The Technical Standards and Commentaries for Port and Harbor Facilities in Japan

The Overseas Coastal Area Development Institute of Japan 3-2-4 Kasumigaseki, Chiyoda-ku, Tokyo, 100-0013, Japan Copyri...

Author: The Overseas Coastal Area Development Institute of Japan - OCDI

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The Overseas Coastal Area Development Institute of Japan 3-2-4 Kasumigaseki, Chiyoda-ku, Tokyo, 100-0013, Japan

Copyright © 2002 by The Overseas Coastal Area Development Institute of Japan Printed by Daikousha Printing Co., Ltd. All rights reserved. No part of this publication may be reproduced, stored in a retrieval systems, transmitted in any form or by any means, electric, mechanical, photocopying, recording or otherwise, without the prior written permission of the publisher. Original Japanese language edition published by the Japan Ports and Harbours Association. Printed in Japan

PREFACE

Preface This book is a translation of the major portion of the Technical Standards and Commentaries of Port and Harbour Facilities in Japan (1999 edition) published by the Japan Port and Harbour Association, stipulated by the Ordinance of the Minister of Transport, which was issued in April 1999. The translation covers about two thirds of the Japanese edition. Japanese islands have a long extension of coastline, measuring about 34,000 km, for the total land area of some 380,000 square kilometers. Throughout her history, Japan has depended on the ports and harbors on daily living and prosperity of people there. Japan did not develop extensive inland canal systems as found in the European Continent because of its mountainous geography, but rather produced many harbors and havens along its coastline in the past. Today, the number of officially designated commercial ports and harbors amounts to about 1,100 and the number of fishing ports exceeds 3,000. After 220 years of isolation from the world civilization from the 17th to 19th centuries, Japan began to modernize its society and civilization rapidly after the Meiji revolution in 1868. Modern technology of port and harbor engineering has been introduced by distinguished engineers from abroad and learned by many ambitious and capable young engineers in Japan. Ports of Yokohama, Kobe, and others began to accommodate large ocean-going vessels in the late 19th century as the Japanese economy had shown a rapid growth. Japanese engineers had drafted an engineering manual on design and construction of port and harbor facilities as early as in 1943. The manual was revised in 1959 with inclusion of new technology such as those of coastal engineering and geotechnical engineering, which were developed during the Second World War or just before it. The Japanese economy that was utterly destroyed by the war had begun to rebuild itself rapidly after the 1950s. There were so many demands for the expansion of port and harbor facilities throughout Japan. Engineers were urged to design and construct facilities after facilities. Japan has built the breakwaters and the quays with the rate of about 20,000 meters each per year throughout the 1960s, 1970s, and 1980s. Such a feat of port development was made possible with provision of sound engineering manuals. The Ministry of Land, Infrastructure and Transport (formerly the Ministry of Transport up to January 2001) which was responsible for port development and operation, revised the basic law on ports and harbors in 1974 so as to take responsibility for provision of technical standards for design, construction, and maintenance of port and harbor facilities. The first official technical standards and commentaries for port and harbor facilities were issued in 1979, and published by the Japan Port and Harbour Association for general use. The technical standards were prepared by a technical committee composed of government engineers within the former Ministry of Transport, including members of the Port and Harbour Research Institute and several District Port Construction Bureaus that were responsible for design and construction in the field. Its English version was published by the Overseas Coastal Area Development Institute in 1980, but it introduced only the skeleton of the Japanese version without giving the details. The Technical Standards and Commentaries for Port and Harbor Facilities in Japan have been revised in 1988 and 1999, each time incorporating new technological developments. The present English translation endeavors to introduce the newest edition of 1999 to the port and harbor engineers overseas. It is a direct translation of essential parts of Japanese edition. Many phrases and expressions reflect the customary, regulatory writings in Japanese, which are often awkward in English. Some sentences after translation may not be fluent enough and give troubles for decipher. The editors in charge of translation request the readers for patience and generosity in their efforts for understanding Japanese technology in port and harbor engineering. With the globalization in every aspect of human activities, indigenous practices and customs are forced to comply with the world standards. Technology by definition is supposed to be universal. Nevertheless, each country has developed its own specialty to suit its local conditions. The overseas readers may find some of Japanese technical standards strange and difficult for adoption for their usage. Such conflicts in technology are the starting points for mutual understanding and further developments in the future. The editors wish wholeheartedly this English version of Japanese technical standards be welcomed by the overseas colleagues and serve for the advancement of port and harbor technology in the world. January 2002 Y. Goda, T. Tabata and S. Yamamoto Editors for translation version -i-

TECHNICAL STANDARDS AND COMMENTARIES FOR PORT AND HARBOUR FACILITIES IN JAPAN

-ii-

CONTENTS

CONTENTS Preface Part I General Chapter 1 General Rules .................................................................................................................................................1 1.1 1.2 1.3

Scope of Application .............................................................................................................................1 Definitions ...............................................................................................................................................2 Usage of SI Units ...................................................................................................................................2

Chapter 2 Datum Level for Construction Work .........................................................................................................4 Chapter 3 Maintenance ....................................................................................................................................................5

Part II Design Conditions Chapter 1 General .............................................................................................................................................................7 Chapter 2 Vessels ..............................................................................................................................................................9 2.1 2.2

Dimensions of Target Vessel ...............................................................................................................9 External Forces Generated by Vessels ...........................................................................................16 2.2.1 2.2.2

2.2.3

2.2.4

General .....................................................................................................................................16 Berthing.....................................................................................................................................16 [1] Berthing Energy..................................................................................................................16 [2] Berthing Velocity ................................................................................................................17 [3] Eccentricity Factor..............................................................................................................20 [4] Virtual Mass Factor ............................................................................................................21 Moored Vessels .......................................................................................................................22 [1] Motions of Moored Vessel..................................................................................................22 [2] Waves Acting on Vessel.....................................................................................................22 [3] Wind Load Acting on Vessel ..............................................................................................23 [4] Current Forces Acting on Vessel........................................................................................24 [5] Load-Deflection Characteristics of Mooring System ..........................................................25 Tractive Force Acting on Mooring Post and Bollard..................................................................25

Chapter 3 Wind and Wind Pressure ..........................................................................................................................28 3.1 3.2 3.3

General..................................................................................................................................................28 Wind .......................................................................................................................................................29 Wind Pressure......................................................................................................................................30

Chapter 4 Waves ..............................................................................................................................................................32 4.1

General..................................................................................................................................................32 4.1.1 4.1.2 4.1.3

4.2 4.3

4.4 4.5

Procedure for Determining the Waves Used in Design.............................................................32 Waves to Be Used in Design ....................................................................................................32 Properties of Waves..................................................................................................................33 [1] Fundamental Properties of Waves .....................................................................................33 [2] Statistical Properties of Waves...........................................................................................37 [3] Wave Spectrum..................................................................................................................38 Method of Determining Wave Conditions to Be Used in Design .................................................40 4.2.1 Principles for Determining the Deepwater Waves Used in Design ...........................................40 4.2.2 Procedure for Obtaining the Parameters of Design Waves ......................................................41 Wave Hindcasting................................................................................................................................42 4.3.1 General .....................................................................................................................................42 4.3.2 Wave Hindcasting in Generating Area ......................................................................................42 4.3.3 Swell Hindcasting......................................................................................................................46 Statistical Processing of Wave Observation and Hindcasted Data .............................................47 Transformations of Waves .................................................................................................................49 4.5.1 General .....................................................................................................................................49 4.5.2 Wave Refraction........................................................................................................................49 4.5.3 Wave Diffraction........................................................................................................................52 [1] Diffraction ...........................................................................................................................52 [2] Combination of Diffraction and Refraction..........................................................................69 4.5.4 Wave Reflection ........................................................................................................................70 -iii-


4.6

4.7 4.8 4.9 4.10

[1] General .............................................................................................................................. 70 [2] Reflection Coefficient ......................................................................................................... 71 [3] Transformation of Waves at Concave Corners, near the Heads of Breakwaters, and around Detached Breakwaters ................................................................................... 72 4.5.5 Wave Shoaling.......................................................................................................................... 74 4.5.6 Wave Breaking ......................................................................................................................... 75 Wave Runup, Overtopping, and Transmission............................................................................... 80 4.6.1 Wave Runup ............................................................................................................................. 80 4.6.2 Wave Overtopping .................................................................................................................... 84 4.6.3 Wave Transmission .................................................................................................................. 90 Wave Setup and Surf Beat ................................................................................................................ 91 4.7.1 Wave Setup .............................................................................................................................. 91 4.7.2 Surf Beat................................................................................................................................... 92 Long-Period Waves and Seiche ....................................................................................................... 93 Waves inside Harbors ........................................................................................................................ 94 4.9.1 Calmness and Disturbances..................................................................................................... 94 4.9.2 Evaluation of Harbor Calmness ................................................................................................ 94 Ship Waves .......................................................................................................................................... 94

Chapter 5 Wave Force ................................................................................................................................................. 100 5.1 5.2

General ............................................................................................................................................... 100 Wave Force Acting on Upright Wall ............................................................................................... 100 5.2.1 5.2.2

5.3 5.4 5.5

General Considerations .......................................................................................................... 100 Wave Forces of Standing and Breaking Waves ..................................................................... 101 [1] Wave Force under Wave Crest........................................................................................ 101 [2] Wave Force under Wave Trough..................................................................................... 105 5.2.3 Impulsive Pressure Due to Breaking Waves .......................................................................... 106 5.2.4 Wave Force on Upright Wall Covered with Wave-Dissipating Concrete Blocks..................... 109 5.2.5 Effect of Alignment of Breakwater on Wave Force ................................................................. 110 5.2.6 Effect of Abrupt Change in Water Depth on Wave Force ....................................................... 110 5.2.7 Wave Force on Upright Wall near Shoreline or on Shore........................................................111 [1] Wave Force at the Seaward Side of Shoreline .................................................................111 [2] Wave Force at the Landward Side of Shoreline ...............................................................111 5.2.8 Wave Force on Upright Wave-Absorbing Caisson ..................................................................111 Mass of Armor Stones and Concrete Blocks ................................................................................ 112 5.3.1 Armor Units on Slope.............................................................................................................. 112 5.3.2 Armor Units on Foundation Mound of Composite Breakwater ............................................... 117 Wave Forces Acting on Cylindrical Members and Large Isolated Structures ......................... 119 5.4.1 Wave Force on Cylindrical Members...................................................................................... 119 5.4.2 Wave Force on Large Isolated Structure ................................................................................ 121 Wave Force Acting on Structure Located near the Still Water Level........................................ 122 5.5.1 Uplift Acting on Horizontal Plate near the Still Water Level .................................................... 122

Chapter 6 Tides and Abnormal Water Levels....................................................................................................... 127 6.1 6.2 6.3 6.4 6.5 6.6

Design Water Level........................................................................................................................... 127 Astronomical Tide ............................................................................................................................. 128 Storm Surge ....................................................................................................................................... 128 Tsunami .............................................................................................................................................. 130 Seiche ................................................................................................................................................. 133 Groundwater Level and Permeation .............................................................................................. 135

Chapter 7 Currents and Current Force ................................................................................................................... 138 7.1 7.2 7.3

General ............................................................................................................................................... 138 Current Forces Acting on Submerged Members and Structures .............................................. 138 Mass of Armor Stones and Concrete Blocks against Currents ................................................. 140

Chapter 8 External Forces Acting on Floating Body and Its Motions ........................................................... 142 8.1 8.2 8.3

General ............................................................................................................................................... 142 External Forces Acting on Floating Body ...................................................................................... 143 Motions of Floating Body and Mooring Force ............................................................................... 145

Chapter 9 Estuarine Hydraulics ................................................................................................................................ 148 9.1

General ............................................................................................................................................... 148

Chapter 10 Littoral Drift .................................................................................................................................................. 154 10.1 General ............................................................................................................................................... 154 10.2 Scouring around Structures ............................................................................................................. 161 10.3 Prediction of Beach Deformation .................................................................................................... 163 -iv-

CONTENTS

Chapter 11 Subsoil ...........................................................................................................................................................167 11.1 Method of Determining Geotechnical Conditions .........................................................................167 11.1.1 11.1.2 11.1.3

11.2

11.3

11.4 11.5 11.6

Principles.................................................................................................................................167 Selection of Soil Investigation Methods ..................................................................................168 Standard Penetration Test ......................................................................................................168 Physical Properties of Soils .............................................................................................................168 11.2.1 Unit Weight of Soil...................................................................................................................168 11.2.2 Classification of Soils ..............................................................................................................169 11.2.3 Coefficient of Permeability of Soil ...........................................................................................169 Mechanical Properties of Soils ........................................................................................................170 11.3.1 Elastic Constants ....................................................................................................................170 11.3.2 Consolidation Properties .........................................................................................................170 11.3.3 Shear Properties .....................................................................................................................173 Angle of Internal Friction by N-value ..............................................................................................175 Application of Soundings Other Than SPT....................................................................................176 Dynamic Properties of Soils .............................................................................................................178 11.6.1 Dynamic Modulus of Deformation ...........................................................................................178 11.6.2 Dynamic Strength Properties ..................................................................................................180

Chapter 12 Earthquakes and Seismic Force ...........................................................................................................182 12.1 12.2 12.3 12.4 12.5 12.6

General................................................................................................................................................182 Earthquake Resistance of Port and Harbor Facilities in Design ................................................182 Seismic Coefficient Method .............................................................................................................184 Design Seismic Coefficient ..............................................................................................................184 Seismic Response Analysis .............................................................................................................190 Seismic Deformation Method ..........................................................................................................192

Chapter 13 Liquefaction .................................................................................................................................................195 13.1 General................................................................................................................................................195 13.2 Prediction of Liquefaction .................................................................................................................195 13.3 Countermeasures against Liquefaction .........................................................................................199

Chapter 14 Earth Pressure and Water Pressure ...................................................................................................200 14.1 Earth Pressure ...................................................................................................................................200 14.2 Earth Pressure under Ordinary Conditions ...................................................................................200 14.2.1 14.2.2

Earth Pressure of Sandy Soil under Ordinary Conditions .......................................................200 Earth Pressure of Cohesive Soil under Ordinary Conditions ..................................................201 14.3 Earth Pressure during Earthquake .................................................................................................202 14.3.1 Earth Pressure of Sandy Soil during Earthquake....................................................................202 14.3.2 Earth Pressure of Cohesive Soil during Earthquake...............................................................204 14.3.3 Apparent Seismic Coefficient ..................................................................................................204 14.4 Water Pressure ..................................................................................................................................205 14.4.1 Residual Water Pressure ........................................................................................................205 14.4.2 Dynamic Water Pressure during Earthquake..........................................................................205

Chapter 15 Loads .............................................................................................................................................................207 15.1 General................................................................................................................................................207 15.2 Deadweight and Surcharge .............................................................................................................207 15.3 Static Load ..........................................................................................................................................207 15.3.1 15.3.2 15.3.3 15.3.4

Static Load under Ordinary Conditions ...................................................................................207 Static Load during Earthquake................................................................................................208 Unevenly Distributed Load ......................................................................................................208 Snow Load ..............................................................................................................................208 15.4 Live Load ............................................................................................................................................209 15.4.1 Train Load ...............................................................................................................................209 15.4.2 Vehicle Load ...........................................................................................................................209 15.4.3 Cargo Handling Equipment Load ............................................................................................209 15.4.4 Sidewalk Live Load .................................................................................................................209

Chapter 16 Coefficient of Friction ................................................................................................................................210 16.1 General................................................................................................................................................210

Part III Materials Chapter 1 General ......................................................................................................................................................... 211 1.1

Selection of Materials........................................................................................................................ 211 -v-


1.2

Safety of Structural Elements .......................................................................................................... 211

Chapter 2 Steel ............................................................................................................................................................... 212 2.1 2.2 2.3

Materials ............................................................................................................................................. 212 Steel Meterial Constants Used in Design Calculation ................................................................. 212 Allowable Stresses ............................................................................................................................ 212 2.3.1 2.3.2 2.3.3 2.3.4 2.3.5 2.3.6 2.3.7

2.4

General ................................................................................................................................... 212 Structural Steel ....................................................................................................................... 212 Steel Piles and Steel Pipe Sheet Piles ................................................................................... 213 Steel Sheet Piles .................................................................................................................... 214 Cast Steel and Forged Steel................................................................................................... 214 Allowable Stresses for Steel at Welded Zones and Spliced Sections .................................... 214 Increase of Allowable Stresses............................................................................................... 215 Corrosion Control .............................................................................................................................. 216 2.4.1 General ................................................................................................................................... 216 2.4.2 Corrosion Rates of Steel Materials ......................................................................................... 216 2.4.3 Corrosion Control Methods..................................................................................................... 217 2.4.4 Cathodic Protection Method ................................................................................................... 217 [1] Range of Application........................................................................................................ 217 [2] Protective Potential .......................................................................................................... 218 [3] Protective Current Density ............................................................................................... 219 2.4.5 Coating Method ...................................................................................................................... 220 [1] Extent of Application ........................................................................................................ 220 [2] Applicable Methods.......................................................................................................... 220 [3] Selection of Method ......................................................................................................... 220

Chapter 3 Concrete ....................................................................................................................................................... 221 3.1 3.2 3.3 3.4 3.5 3.6

General ............................................................................................................................................... 221 Basics of Design Based on the Limit State Design Method ....................................................... 221 Design Based on Allowable Stress Method .................................................................................. 223 Concrete Materials ............................................................................................................................ 224 Concrete Quality and Performance ................................................................................................ 225 Underwater Concrete ....................................................................................................................... 227

Chapter 4 Bituminous Materials ................................................................................................................................ 228 4.1 4.2

General ............................................................................................................................................... 228 Asphalt Mat ........................................................................................................................................ 228 4.2.1 4.2.2 4.2.3

4.3 4.4

General ................................................................................................................................... 228 Materials ................................................................................................................................. 228 Mix Proportioning.................................................................................................................... 229 Paving Materials ................................................................................................................................ 229 Sand Mastic Asphalt ......................................................................................................................... 229 4.4.1 General ................................................................................................................................... 229 4.4.2 Materials ................................................................................................................................. 230 4.4.3 Mix Proportioning.................................................................................................................... 230

Chapter 5 Stone ............................................................................................................................................................. 231 5.1 5.2 5.3 5.4

General ............................................................................................................................................... 231 Rubble for Foundation ...................................................................................................................... 231 Backfilling Materials .......................................................................................................................... 231 Base Course Materials of Pavement ............................................................................................. 232

Chapter 6 Timber ........................................................................................................................................................... 233 6.1

Quality of Timber ............................................................................................................................... 233 6.1.1 6.1.2

6.2 6.3 6.4 6.5

Structural Timber .................................................................................................................... 233 Timber Piles............................................................................................................................ 233 Allowable Stresses of Timber .......................................................................................................... 233 6.2.1 General ................................................................................................................................... 233 6.2.2 Allowable Stresses of Structural Timber ................................................................................. 233 Quality of Glued Laminated Timber ............................................................................................... 233 6.3.1 Allowable Stress for Glued Laminated Timber ....................................................................... 233 Joining of Timber ............................................................................................................................... 233 Maintenance of Timber..................................................................................................................... 233

Chapter 7 Other Materials ........................................................................................................................................... 234 7.1 7.2 7.3

Metals Other Than Steel .................................................................................................................. 234 Plastics and Rubbers ........................................................................................................................ 234 Coating Materials .............................................................................................................................. 236 -vi-

CONTENTS

7.4

Grouting Materials .............................................................................................................................237 7.4.1 7.4.2

General ...................................................................................................................................237 Properties of Grouting Materials .............................................................................................237

Chapter 8 Recyclable Resources .............................................................................................................................238 8.1 8.2 8.3 8.4

General................................................................................................................................................238 Slag ......................................................................................................................................................238 Coal Ash..............................................................................................................................................239 Crashed Concrete .............................................................................................................................240

Part IV Precast Concrete Units Chapter 1 Caissons .......................................................................................................................................................241 1.1 1.2 1.3 1.4

General................................................................................................................................................241 Determination of Dimensions ..........................................................................................................242 Floating Stability ................................................................................................................................242 Design External Forces ....................................................................................................................243 1.4.1 1.4.2 1.4.3 1.4.4 1.4.5

1.5

1.6

Combination of Loads and Load Factors ................................................................................243 External Forces during Fabrication .........................................................................................249 External Forces during Launching and Floating......................................................................249 External Forces during Installation..........................................................................................250 External Forces after Construction..........................................................................................250 [1] Outer Walls.......................................................................................................................250 [2] Bottom Slab......................................................................................................................251 [3] Partition Walls and Others................................................................................................253 Design of Members ...........................................................................................................................254 1.5.1 Outer Wall ...............................................................................................................................254 1.5.2 Partition Wall ...........................................................................................................................254 1.5.3 Bottom Slab.............................................................................................................................254 1.5.4 Others .....................................................................................................................................255 Design of Hooks for Suspension by Crane ...................................................................................255

Chapter 2 L-Shaped Blocks ........................................................................................................................................256 2.1 2.2 2.3

General................................................................................................................................................256 Determination of Dimensions ..........................................................................................................256 Loads Acting on Members ...............................................................................................................257 2.3.1 2.3.2 2.3.3

2.4

2.5

General ...................................................................................................................................257 Earth Pressure ........................................................................................................................258 Converted Loads for Design Calculation.................................................................................258 Design of Members ...........................................................................................................................259 2.4.1 Front Wall................................................................................................................................259 2.4.2 Footing ....................................................................................................................................259 2.4.3 Bottom Slab.............................................................................................................................259 2.4.4 Buttress ...................................................................................................................................260 Design of Hooks for Suspension by Crane ...................................................................................260

Chapter 3 Cellular Blocks ............................................................................................................................................261 3.1 3.2

General................................................................................................................................................261 Determination of Dimensions ..........................................................................................................261 3.2.1 3.2.2

3.3

3.4

Shape of Cellular Blocks .........................................................................................................261 Determination of Dimensions ..................................................................................................261 Loads Acting on Cellular Blocks......................................................................................................262 3.3.1 General ...................................................................................................................................262 3.3.2 Earth Pressure of Filling and Residual Water Pressure..........................................................262 3.3.3 Converted Loads for Design Calculation.................................................................................264 Design of Members ...........................................................................................................................264 3.4.1 Rectangular Cellular Blocks ....................................................................................................264 3.4.2 Other Types of Cellular Blocks................................................................................................265

Chapter 4 Upright Wave-Absorbing Caissons ......................................................................................................267 4.1 4.2 4.3

General................................................................................................................................................267 External Forces Acting on Members ..............................................................................................267 Design of Members ...........................................................................................................................269

Chapter 5 Hybrid Caissons .........................................................................................................................................270 5.1 5.2

General................................................................................................................................................270 Determination of Dimensions ..........................................................................................................270 -vii-


5.3 5.4

Design External Forces .................................................................................................................... 271 Design of Members ........................................................................................................................... 271 5.4.1 5.4.2 5.4.3 5.4.4 5.4.5 5.4.6

5.5

Section Force.......................................................................................................................... 271 Design of Composite Slabs .................................................................................................... 271 Design of SRC Members ........................................................................................................ 271 Design of Partitions................................................................................................................. 271 Design of Corners and Joints ................................................................................................. 271 Safety against Fatigue Failure ................................................................................................ 272 Corrosion Control .............................................................................................................................. 272

Part V Foundations Chapter 1 General ......................................................................................................................................................... 273 Chapter 2 Bearing Capacity of Shallow Foundations ........................................................................................ 274 2.1 2.2 2.3 2.4 2.5

General ............................................................................................................................................... 274 Bearing Capacity of Foundation on Sandy Ground ..................................................................... 274 Bearing Capacity of Foundation on Clayey Ground .................................................................... 275 Bearing Capacity of Multilayered Ground ..................................................................................... 276 Bearing Capacity for Eccentric and Inclined Loads ..................................................................... 277

Chapter 3 Bearing Capacity of Deep Foundations ............................................................................................. 280 3.1 3.2 3.3

General ............................................................................................................................................... 280 Vertical Bearing Capacity................................................................................................................. 280 Lateral Bearing Capacity .................................................................................................................. 281

Chapter 4 Bearing Capacity of Pile Foundations ................................................................................................ 284 4.1

Allowable Axial Bearing Capacity of Piles ..................................................................................... 284 4.1.1 4.1.2 4.1.3 4.1.4 4.1.5 4.1.6 4.1.7 4.1.8 4.1.9 4.1.10 4.1.11

4.2

4.3

4.4

4.5

General ................................................................................................................................... 284 Standard Allowable Axial Bearing Capacity............................................................................ 284 Ultimate Axial Bearing Capacity of Single Piles...................................................................... 285 Estimation of Ultimate Axial Bearing Capacity by Loading Tests ........................................... 285 Estimation of Ultimate Axial Bearing Capacity by Static Bearing Capacity Formulas ............ 286 Examination of Compressive Stress of Pile Materials ............................................................ 288 Decrease of Bearing Capacity Due to Joints .......................................................................... 288 Decrease of Bearing Capacity Due to Slenderness Ratio ...................................................... 288 Bearing Capacity of Pile Group .............................................................................................. 288 Examination of Negative Skin Friction .................................................................................... 290 Examination of Settlement of Piles ......................................................................................... 291 Allowable Pulling Resistance of Piles ............................................................................................ 291 4.2.1 General ................................................................................................................................... 291 4.2.2 Standard Allowable Pulling Resistance .................................................................................. 292 4.2.3 Maximum Pulling Resistance of Single Pile............................................................................ 292 4.2.4 Examination of Tensile Stress of Pile Materials...................................................................... 293 4.2.5 Matters to Be Considered for Obtaining Allowable Pulling Resistance of Piles...................... 293 Allowable Lateral Bearing Capacity of Piles ................................................................................. 293 4.3.1 General ................................................................................................................................... 293 4.3.2 Estimation of Allowable Lateral Bearing Capacity of Piles ..................................................... 295 4.3.3 Estimation of Pile Behavior Using Loading Tests ................................................................... 295 4.3.4 Estimation of Pile Behavior Using Analytical Methods ........................................................... 295 4.3.5 Consideration of Pile Group Action......................................................................................... 301 4.3.6 Lateral Bearing Capacity of Coupled Piles ............................................................................. 301 Pile Design in General ...................................................................................................................... 304 4.4.1 Load Sharing .......................................................................................................................... 304 4.4.2 Load Distribution..................................................................................................................... 305 4.4.3 Distance between Centers of Piles......................................................................................... 305 4.4.4 Allowable Stresses for Pile Materials...................................................................................... 305 Detailed Design ................................................................................................................................. 306 4.5.1 Examination of Loads during Construction ............................................................................. 306 4.5.2 Design of Joints between Piles and Structure ........................................................................ 307 4.5.3 Joints of Piles.......................................................................................................................... 308 4.5.4 Change of Plate Thickness or Materials of Steel Pipe Piles................................................... 308 4.5.5 Other Points for Caution in Design ......................................................................................... 308

Chapter 5 Settlement of Foundations ..................................................................................................................... 310 5.1 5.2

Stress in Soil Mass ........................................................................................................................... 310 Immediate Settlement....................................................................................................................... 310 -viii-

CONTENTS

5.3 5.4 5.5

Consolidation Settlement .................................................................................................................310 Lateral Displacement ........................................................................................................................312 Differential Settlements ....................................................................................................................312

Chapter 6 Stability of Slopes ......................................................................................................................................314 6.1 6.2

General................................................................................................................................................314 Stability Analysis ................................................................................................................................315 6.2.1 6.2.2

Stability Analysis Using Circular Slip Surface Method ............................................................315 Stability Analysis Assuming Slip Surfaces Other Than Circular Arc Slip Surface...................316

Chapter 7 Soil Improvement Methods .....................................................................................................................318 7.1 7.2 7.3

General................................................................................................................................................318 Replacement Method ........................................................................................................................318 Vertical Drain Method .......................................................................................................................318 7.3.1 7.3.2

Principle of Design ..................................................................................................................318 Determination of Height and Width of Fill................................................................................319 [1] Height and Width of Fill Required for Soil Improvement ..................................................319 [2] Height and Width of Fill Required for Stability of Fill Embankment ..................................319 7.3.3 Design of Drain Piles...............................................................................................................319 [1] Drain Piles and Sand Mat.................................................................................................319 [2] Interval of Drain Piles .......................................................................................................320 7.4 Deep Mixing Method .........................................................................................................................322 7.4.1 Principle of Design ..................................................................................................................322 [1] Scope of Application.........................................................................................................322 [2] Basic Concept ..................................................................................................................323 7.4.2 Assumptions for Dimensions of Stabilized Body.....................................................................323 [1] Mixture Design of Stabilized Soil......................................................................................323 [2] Allowable Stress of Stabilized Body .................................................................................324 7.4.3 Calculation of External Forces ................................................................................................325 7.5 Lightweight Treated Soil Method ....................................................................................................326 7.5.1 Outline of Lightweight Treated Soil Method ............................................................................326 7.5.2 Basic Design Concept.............................................................................................................326 7.5.3 Mixture Design of Treated Soil................................................................................................327 7.5.4 Examination of Area to Be Treated .........................................................................................328 7.5.5 Workability Verification Tests ..................................................................................................328 7.6 Replacement Method with Granulated Blast Furnace Slag ........................................................328 7.6.1 Principle of Design ..................................................................................................................328 7.6.2 Physical Properties of Granulated Blast Furnace Slag ...........................................................328 7.7 Premixing Method..............................................................................................................................329 7.7.1 Principle of Design ..................................................................................................................329 [1] Scope of Application.........................................................................................................329 [2] Consideration for Design..................................................................................................329 7.7.2 Preliminary Survey ..................................................................................................................329 7.7.3 Determination of Strength of Treated Soil...............................................................................330 7.7.4 Mixture Design of Treated Soil................................................................................................330 7.7.5 Examination of Area of Improvement......................................................................................331 7.8 Active Earth Pressure of Solidified Geotechnical Materials........................................................333 7.8.1 Scope of Application ...............................................................................................................333 7.8.2 Active Earth Pressure .............................................................................................................333 [1] Outline ..............................................................................................................................333 [2] Strength Parameters ........................................................................................................334 [3] Calculation of Active Earth Pressure................................................................................334 [4] Case of Limited Area of Subsoil Improvement .................................................................335 7.9 Sand Compaction Pile Method (for Sandy Subsoil) .....................................................................336 7.9.1 Principle of Design ..................................................................................................................336 7.9.2 Sand Volume to Be Supplied ..................................................................................................336 7.9.3 Design Based on Trial Execution ............................................................................................338 7.10 Sand Compaction Pile Method (for Cohesive Subsoil) ...............................................................339 7.10.1 Principle of Design ..................................................................................................................339 [1] Scope of Application.........................................................................................................339 [2] Basic Concept ..................................................................................................................339 7.10.2 Strength and Permeability of Sand Piles.................................................................................339 7.10.3 Shear Strength of Improved Subsoil .......................................................................................339 7.10.4 Stability Analysis .....................................................................................................................340 7.10.5 Examining Consolidation.........................................................................................................341

-ix-


Part VI Navigation Channels and Basins Chapter 1 General ......................................................................................................................................................... 345 Chapter 2 Navigation Channels ................................................................................................................................ 346 2.1 2.2 2.3 2.4 2.5 2.6

General ............................................................................................................................................... 346 Alignment of Navigation Channel .................................................................................................. 346 Width of Navigation Channel ........................................................................................................... 347 Depth of Navigation Channel .......................................................................................................... 348 Length of Navigation Channel at Harbor Entrance ...................................................................... 348 Calmness of Navigation Channel ................................................................................................... 348

Chapter 3 Navigation Channels outside Breakwaters ....................................................................................... 350 3.1 3.2 3.3

General ............................................................................................................................................... 350 Width of Navigation Channel ........................................................................................................... 350 Depth of Navigation Channel .......................................................................................................... 350

Chapter 4 Basins............................................................................................................................................................ 351 4.1 4.2

General ............................................................................................................................................... 351 Location and Area of Basin ............................................................................................................. 351 4.2.1 4.2.2 4.2.3

4.3 4.4 4.5

Location .................................................................................................................................. 351 Area of Basin Used for Anchorage or Mooring ....................................................................... 351 Area of Basin Used for Ship Maneuvering.............................................................................. 352 [1] Turning Basin................................................................................................................... 352 [2] Mooring / Unmooring Basin ............................................................................................. 353 Depth of Basin ................................................................................................................................... 353 Calmness of Basin ............................................................................................................................ 353 Timber Sorting Pond ......................................................................................................................... 354

Chapter 5 Small Craft Basins ..................................................................................................................................... 355 Chapter 6 Maintenance of Navigation Channels and Basins .......................................................................... 355 6.1

General ............................................................................................................................................... 355

Part VII Protective Facilities for Harbors Chapter 1 General ......................................................................................................................................................... 357 1.1 1.2

General Consideration ..................................................................................................................... 357 Maintenance....................................................................................................................................... 357

Chapter 2 Breakwaters ................................................................................................................................................ 358 2.1 2.2 2.3 2.4 2.5

General ............................................................................................................................................... 358 Layout of Breakwaters ...................................................................................................................... 358 Design Conditions of Breakwaters ................................................................................................. 359 Selection of Structural Types .......................................................................................................... 359 Determination of Cross Section ...................................................................................................... 362 2.5.1 2.5.2 2.5.3 2.5.4

2.6

2.7

2.8

Upright Breakwater ................................................................................................................. 362 Composite Breakwater ........................................................................................................... 363 Sloping Breakwater................................................................................................................. 363 Caisson Type Breakwater Covered with Wave-Dissipating Concrete Blocks ........................ 364 External Forces for Stability Calculation ........................................................................................ 364 2.6.1 General ................................................................................................................................... 364 2.6.2 Wave Forces........................................................................................................................... 365 2.6.3 Hydrostatic Pressure .............................................................................................................. 365 2.6.4 Buoyancy ................................................................................................................................ 365 2.6.5 Deadweight............................................................................................................................. 365 2.6.6 Stability during Earthuakes ..................................................................................................... 365 Stability Calculation........................................................................................................................... 365 2.7.1 Stability Calculation of Upright Section................................................................................... 365 2.7.2 Stability Calculation of Sloping Section .................................................................................. 369 2.7.3 Stability Calculation of Whole Section .................................................................................... 369 2.7.4 Stability Calculation for Head and Corner of Breakwater ....................................................... 369 Details of Structures ......................................................................................................................... 370 2.8.1 Upright Breakwater ................................................................................................................. 370 2.8.2 Composite Breakwater ........................................................................................................... 371 2.8.3 Sloping Breakwater................................................................................................................. 372 -x-

CONTENTS

2.8.4

Caisson Type Breakwater Covered with Wave-Dissipating Concrete Blocks.........................372

2.9 Detailed Design of Upright Section .................................................................................................372 2.10 Breakwaters for Timber-Handling Facilities ..................................................................................372 2.10.1 2.10.2

Breakwaters for Timber Storage Ponds and Timber Sorting Ponds .......................................372 Fences to Prevent Timber Drifting ..........................................................................................373 2.11 Storm Surge Protection Breakwater ...............................................................................................373 2.12 Tsunami Protection Breakwater ......................................................................................................373

Chapter 3 Other Types of Breakwaters ..................................................................................................................376 3.1 3.2

Selection of Structural Type .............................................................................................................376 Gravity Type Special Breakwaters..................................................................................................377 3.2.1 3.2.2

3.3

General ...................................................................................................................................377 Upright Wave-Absorbing Block Breakwater ............................................................................378 [1] General.............................................................................................................................378 [2] Crest Elevation .................................................................................................................378 [3] Wave Force ......................................................................................................................379 3.2.3 Wave-Absorbing Caisson Breakwater ....................................................................................379 [1] General.............................................................................................................................379 [2] Determination of Target Waves to Be Absorbed..............................................................380 [3] Determination of Dimensions for Wave-Absorbing Section .............................................380 [4] Wave Force for Examination of Structural Stability ..........................................................380 [5] Wave Force for Design of Structural Members ................................................................380 3.2.4 Sloping-Top Caisson Breakwater............................................................................................380 [1] General.............................................................................................................................380 [2] Wave Force ......................................................................................................................381 Non-Gravity Type Breakwaters .......................................................................................................382 3.3.1 Curtain Wall Breakwater .........................................................................................................382 [1] General.............................................................................................................................382 [2] Wave Force ......................................................................................................................384 [3] Design of Piles .................................................................................................................384 3.3.2 Floating Breakwater ................................................................................................................384 [1] General.............................................................................................................................384 [2] Selection of Design Conditions ........................................................................................385 [3] Design of Mooring System ...............................................................................................385 [4] Design of Floating Body Structure....................................................................................386

Chapter 4 Locks..............................................................................................................................................................388 4.1 4.2 4.3

Selection of Location .........................................................................................................................388 Size and Layout of Lock ...................................................................................................................388 Selection of Structural Type .............................................................................................................389 4.3.1 4.3.2

4.4 4.5 4.6

Gate ........................................................................................................................................389 Lock Chamber.........................................................................................................................389 External Forces and Loads Acting on Lock...................................................................................389 Pumping and Drainage System ......................................................................................................389 Auxiliary Facilities ..............................................................................................................................389

Chapter 5 Facilities to Prevent Shoaling and Siltation .......................................................................................390 5.1 5.2

General................................................................................................................................................390 Jetty .....................................................................................................................................................390 5.2.1 5.2.2

5.3 5.4

5.5 5.6

Layout of Jetty.........................................................................................................................390 Details of Jetty.........................................................................................................................391 Group of Groins .................................................................................................................................392 Training Jetties ...................................................................................................................................392 5.4.1 Layout of Training Jetties ........................................................................................................392 5.4.2 Water Depth at Tip of Training Jetty .......................................................................................393 5.4.3 Structure of Training Jetty .......................................................................................................393 Facilities to Trap Littoral Transport and Sediment Flowing out of Rivers .................................393 Countermeasures against Wind-Blown Sand ...............................................................................394 5.6.1 General ...................................................................................................................................394 5.6.2 Selection of Countermeasures................................................................................................394

Chapter 6 Revetments ..................................................................................................................................................396 6.1 6.2 6.3 6.4 6.5

Principle of Design ............................................................................................................................396 Design Conditions .............................................................................................................................396 Structural Stability..............................................................................................................................398 Determination of Cross Section ......................................................................................................398 Details..................................................................................................................................................398 -xi-


Part VIII Mooring Facilities Chapter 1 General ......................................................................................................................................................... 401 1.1 1.2

General Consideration ..................................................................................................................... 401 Maintenance of Mooring Facilities .................................................................................................. 401

Chapter 2 Dimensions of Mooring Facilities.......................................................................................................... 402 2.1 2.2 2.3 2.4 2.5 2.6

Length and Water Depth of Berths ................................................................................................. 402 Crown Heights of Mooring Facilities............................................................................................... 405 Ship Clearance for Mooring Facilities ............................................................................................ 405 Design Water Depth ......................................................................................................................... 405 Protection against Scouring............................................................................................................. 406 Ancillary Facilities .............................................................................................................................. 406

Chapter 3 Structural Types of Mooring Facilities ................................................................................................ 407 Chapter 4 Gravity Type Quaywalls .......................................................................................................................... 408 4.1 4.2 4.3

Principle of Design ............................................................................................................................ 408 External Forces and Loads Acting on Walls ................................................................................. 408 Stability Calculations......................................................................................................................... 410 4.3.1 4.3.2 4.3.3 4.3.4 4.3.5

4.4 4.5 4.6

Items to Be Considered in Stability Calculations .................................................................... 410 Examination against Sliding of Wall........................................................................................ 410 Examination Concerning Bearing Capacity of Foundation ..................................................... 411 Examination Concerning Overturning of Wall......................................................................... 411 Examination on Soft Foundation............................................................................................. 411 Stability Calculations of Cellular Concrete Blocks ....................................................................... 412 Effects of Backfill ............................................................................................................................... 413 Detailed Design ................................................................................................................................. 414

Chapter 5 Sheet Pile Quaywalls ............................................................................................................................... 415 5.1 5.2

General ............................................................................................................................................... 415 External Forces Acting on Sheet Pile Wall ................................................................................... 415

5.3

Design of Sheet Pile Wall ................................................................................................................ 417

5.2.1

External Forces to Be Considered.......................................................................................... 415

5.3.1 5.3.2 5.3.3 5.3.4 5.3.5

5.4 5.5 5.6 5.7

5.8

5.9

Setting Level of Tie Rod ......................................................................................................... 417 Embedded Length of Sheet Piles ........................................................................................... 417 Bending Moment of Sheet Piles and Reaction at Tie Rod Setting Point ................................ 418 Cross Section of Sheet Piles .................................................................................................. 419 Consideration of the Effect of Section Rigidity of Sheet Piles ................................................ 419 Design of Tie Rods ........................................................................................................................... 424 5.4.1 Tension of Tie Rod ................................................................................................................. 424 5.4.2 Cross Section of Tie Rod........................................................................................................ 424 Design of Wale .................................................................................................................................. 425 Examination for Circular Slip ........................................................................................................... 425 Design of Anchorage Work .............................................................................................................. 426 5.7.1 Selection of Structural Type of Anchorage Work.................................................................... 426 5.7.2 Location of Anchorage Work .................................................................................................. 426 5.7.3 Design of Anchorage Work..................................................................................................... 427 Detailed Design ................................................................................................................................. 428 5.8.1 Coping .................................................................................................................................... 428 5.8.2 Fitting of Tie Rods and Wale to Sheet Piles ........................................................................... 429 5.8.3 Tie Rod ................................................................................................................................... 429 5.8.4 Fitting of Tie Rods to Anchorage Work................................................................................... 429 Special Notes for Design of Sheet Pile Wall on Soft Ground..................................................... 429

Chapter 6 Sheet Pile Quaywalls with Relieving Platform ................................................................................. 431 6.1 6.2 6.3 6.4 6.5

Scope of Application ......................................................................................................................... 431 Principles of Design .......................................................................................................................... 431 Determination of Height and Width of Relieving Platform .......................................................... 431 Earth Pressure and Residual Water Pressure Acting on Sheet Piles ...................................... 432 Design of Sheet Pile Wall ................................................................................................................ 432 6.5.1 6.5.2

6.6

Embedded Length of Sheet Piles ........................................................................................... 432 Cross Section of Sheet Piles .................................................................................................. 433 Design of Relieving Platform and Relieving Platform Piles ........................................................ 433 6.6.1 External Forces Acting on Relieving Platform ........................................................................ 433 6.6.2 Design of Relieving Platform .................................................................................................. 433 6.6.3 Design of Piles ........................................................................................................................ 434 -xii-

CONTENTS

6.7 6.8

Examination of Stability as Gravity Type Wall ..............................................................................434 Examination of Stability against Circular Slip................................................................................435

Chapter 7 Steel Sheet Pile Cellular-Bulkhead Quaywalls ................................................................................436 7.1 7.2 7.3

Principle of Design ............................................................................................................................436 External Forces Acting on Steel Sheet Pile Cellular-Bulkhead Quaywall ................................437 Examination of Wall Width against Shear Deformation ..............................................................438 7.3.1 7.3.2 7.3.3 7.3.4

General ...................................................................................................................................438 Equivalent Width of Wall .........................................................................................................439 Calculation of Deformation Moment........................................................................................439 Calculation of Resisting Moment.............................................................................................440 7.4 Examination of Stability of Wall Body as a Whole........................................................................443 7.4.1 General ...................................................................................................................................443 7.4.2 Modulus of Subgrade Reaction...............................................................................................443 7.4.3 Calculation of Subgrade Reaction and Wall Displacement.....................................................443 7.5 Examination of Bearing Capacity of the Ground ..........................................................................448 7.6 Examination against Sliding of Wall ...............................................................................................448 7.7 Examination of Displacement of Wall Top .....................................................................................448 7.8 Examination of Stability against Circular Slip................................................................................449 7.9 Layout of Cells and Arcs ..................................................................................................................449 7.10 Calculation of Hoop Tension............................................................................................................449 7.11 Design of T-Shaped Sheet Pile .......................................................................................................450 7.11.1 General ...................................................................................................................................450 7.11.2 Structure of T-Shaped Sheet Pile ...........................................................................................450 7.12 Detailed Design..................................................................................................................................451 7.12.1 Design of Pile to Support Coping ............................................................................................451 7.12.2 Design of Coping.....................................................................................................................451

Chapter 8 Steel Plate Cellular-Bulkhead Quaywalls ..........................................................................................452 8.1 8.2

Scope of Application .........................................................................................................................452 Placement-Type Steel Plate Cellular-Bulkhead Quaywalls ........................................................452 8.2.1 8.2.2 8.2.3 8.2.4 8.2.5 8.2.6 8.2.7 8.2.8 8.2.9

8.3

Principle of Design ..................................................................................................................452 External Forces Acting on Steel Plate Cellular-Bulkhead .......................................................453 Examination of Wall Width against Shear Deformation ..........................................................453 Examination of Stability of Wall Body as a Whole...................................................................454 Examination of Bearing Capacity of the Ground .....................................................................455 Examination of Stability against Circular Slip..........................................................................455 Determination of Thickness of Steel Plate of Cell Shell ..........................................................455 Layout of Cells and Arcs .........................................................................................................456 Detailed Design.......................................................................................................................456 Embedded-Type Steel Plate Cellular-Bulkhead Quaywalls........................................................456 8.3.1 Principle of Design ..................................................................................................................456 8.3.2 External Forces Acting on Embedded-Type Steel Plate Celluler-Bulkhead............................457 8.3.3 Examination of Wall Width against Shear Deformation ..........................................................457 8.3.4 Examination of Stability of Wall Body as a Whole...................................................................458 8.3.5 Examination of Bearing Capacity of the Ground .....................................................................458 8.3.6 Examination against Sliding of Wall ........................................................................................458 8.3.7 Examination of Displacement of Wall Top ..............................................................................458 8.3.8 Examination of Stability against Circular Slip..........................................................................458 8.3.9 Layout of Cells and Arcs .........................................................................................................458 8.3.10 Determination of Plate Thickness of Cell Shell and Arc Section.............................................458 8.3.11 Joints and Stiffeners................................................................................................................459 8.3.12 Detailed Design.......................................................................................................................459

Chapter 9 Open-Type Wharves on Vertical Piles ................................................................................................460 9.1 9.2

Principle of Design ............................................................................................................................460 Layout and Dimensions ....................................................................................................................462 9.2.1 9.2.2 9.2.3

9.3 9.4 9.5

Size of Deck Block and Layout of Piles...................................................................................462 Dimensions of Superstructure.................................................................................................462 Arrangement of Fenders and Bollards ....................................................................................463 External Forces Acting on Open-Type Wharf ...............................................................................463 9.3.1 Design External Forces...........................................................................................................463 9.3.2 Calculation of Fender Reaction Force.....................................................................................464 Assumptions Concerning Sea Bottom Ground .............................................................................464 9.4.1 Determination of Slope Inclination ..........................................................................................464 9.4.2 Virtual Ground Surface............................................................................................................465 Design of Piles ...................................................................................................................................465 -xiii-


9.5.1 9.5.2 9.5.3 9.5.4 9.5.5 9.5.6 9.5.7 9.5.8 9.5.9

9.6

9.7 9.8 9.9

General ................................................................................................................................... 465 Coefficient of Horizontal Subgrade Reaction.......................................................................... 465 Virtual Fixed Point................................................................................................................... 466 Member Forces Acting on Individual Piles.............................................................................. 466 Cross-Sectional Stresses of Piles........................................................................................... 468 Examination of Embedded Length for Bearing Capacity ........................................................ 468 Examination of Embedded Length for Lateral Resistance...................................................... 468 Examination of Pile Joints....................................................................................................... 468 Change of Plate Thickness or Material of Steel Pipe Pile ...................................................... 468 Examination of Earthquake-Resistant Performance ................................................................... 469 9.6.1 Assumption of Cross Section for Earthquake-Resistant Performance Examination .............. 470 9.6.2 Examination Method of Earthquake-Resistant Performance.................................................. 470 9.6.3 Determination of Seismic Motion for Examination of Earthquake-Resistant Performance..... 471 9.6.4 Examination of Load Carrying Capacity Using Simplified Method.......................................... 473 9.6.5 Examination of Load Carrying Capacity Using Elasto-Plastic Analysis .................................. 475 Design of Earth-Retaining Section ................................................................................................. 477 Examination of Stability against Circular Slip ............................................................................... 477 Detailed Design ................................................................................................................................. 478 9.9.1 Load Combinations for Superstructure Design....................................................................... 478 9.9.2 Calculation of Reinforcing Bar Arrangement of Superstructure .............................................. 478 9.9.3 Design of Pile Head ................................................................................................................ 478

Chapter 10 Open-Type Wharves on Coupled Raking Piles ............................................................................... 480 10.1 Principle of Design ............................................................................................................................ 480 10.2 Layout and Dimensions .................................................................................................................... 481 10.2.1 10.2.2 10.2.3

Size of Deck Block and Layout of Piles .................................................................................. 481 Dimensions of Supersutructure .............................................................................................. 481 Arrangement of Fenders and Bollards.................................................................................... 481 10.3 External Forces Acting on Open-Type Wharf on Coupled Raking Piles .................................. 481 10.3.1 Design External Forces .......................................................................................................... 481 10.3.2 Calculation of Fender Reaction Force .................................................................................... 481 10.4 Assumptions Concerning Sea Bottom Ground............................................................................. 481 10.4.1 Determination of Slope Inclination .......................................................................................... 481 10.4.2 Virtual Ground Surface ........................................................................................................... 481 10.5 Determination of Forces Acting on Piles and Cross Sections of Piles ..................................... 481 10.5.1 Horizontal Force Transmitted to Heads of Coupled Raking Piles........................................... 481 10.5.2 Vertical Load Transmitted to Heads of Coupled Raking Piles ................................................ 483 10.5.3 Pushing-In and Pulling-Out Forces of Coupled Raking Piles ................................................. 483 10.5.4 Cross-Sectional Stresses of Piles........................................................................................... 483 10.6 Examination of Strength of Wharf in the Direction of Its Face Line .......................................... 484 10.7 Embedded Length of Raking Pile ................................................................................................... 484 10.8 Design of Earth-Retaining Section ................................................................................................. 484 10.9 Examination of Stability against Circular Slip ............................................................................... 484 10.10 Detailed Design ................................................................................................................................. 484

Chapter 11 Detached Pier ............................................................................................................................................. 485 11.1 Scope of Application ......................................................................................................................... 485 11.2 Principle of Design ............................................................................................................................ 485 11.3 Design of Detached Pier .................................................................................................................. 485 11.3.1 11.3.2 11.3.3 11.3.4

11.4 11.5

Layout and Dimensions .......................................................................................................... 485 External Forces and Loads..................................................................................................... 485 Design of Piers ....................................................................................................................... 486 Design of Girder...................................................................................................................... 486 Ancillary Equipment .......................................................................................................................... 486 Detailed Design ................................................................................................................................. 486 11.5.1 Superstructure ........................................................................................................................ 486 11.5.2 Gangways .............................................................................................................................. 486

Chapter 12 Floating Piers .............................................................................................................................................. 487 12.1 Scope of Application ......................................................................................................................... 487 12.2 Principle of Design ............................................................................................................................ 488 12.3 Design of Pontoon............................................................................................................................. 488 12.3.1 12.3.2 12.3.3 12.3.4

12.4

Dimensions of Pontoon........................................................................................................... 488 External Forces and Loads Acting on Pontoon ...................................................................... 488 Stability of Pontoon................................................................................................................. 488 Design of Individual Parts of Pontoon..................................................................................... 489 Design of Mooring System............................................................................................................... 490 -xiv-

CONTENTS

12.4.1 12.4.2

12.4.3

12.5 Design 12.5.1 12.5.2 12.5.3

Mooring Method ......................................................................................................................490 Design of Mooring Chain.........................................................................................................490 [1] Design External Forces ....................................................................................................490 [2] Setting of Chain................................................................................................................490 [3] Diameter of Chain ............................................................................................................490 Design of Mooring Anchor.......................................................................................................492 [1] Design External Forces ....................................................................................................492 [2] Design of Mooring Anchor................................................................................................492 of Access Bridge and Gangway ........................................................................................492 Dimensions and Inclination .....................................................................................................492 Design of Access Bridge and Gangway..................................................................................493 Adjusting Tower ......................................................................................................................493

Chapter 13 Dolphins ........................................................................................................................................................494 13.1 13.2 13.3 13.4 13.5 13.6

Principle of Design ............................................................................................................................494 Layout ..................................................................................................................................................494 External Forces Acting on Dolphins ...............................................................................................495 Pile Type Dolphins ............................................................................................................................495 Steel Cellular-Bulkhead Type Dolphins .........................................................................................495 Caisson Type Dolphins .....................................................................................................................496

Chapter 14 Slipways and Shallow Draft Quays ......................................................................................................497 14.1 Slipways ..............................................................................................................................................497 14.1.1 14.1.2 14.1.3

Principle of Design ..................................................................................................................497 Location of Slipway .................................................................................................................497 Dimensions of Individual Parts................................................................................................497 [1] Elevations of Individual Parts ...........................................................................................497 [2] Slipway Length and Background Space...........................................................................498 [3] Water Depth .....................................................................................................................498 [4] Gradient of Slipway ..........................................................................................................498 [5] Basin Area........................................................................................................................498 14.1.4 Front Wall and Pavement........................................................................................................499 [1] Front Wall .........................................................................................................................499 [2] Pavement .........................................................................................................................499 14.2 Shallow Draft Quay ...........................................................................................................................499

Chapter 15 Air-Cushion Vehicle Landing Facilities ...............................................................................................500 15.1 15.2 15.3 15.4

Principle of Design ............................................................................................................................500 Location ...............................................................................................................................................501 Air-Cushion Vehicle Landing Facilities ...........................................................................................501 Dimensions of Individual Parts ........................................................................................................501

Chapter 16 Mooring Buoys and Mooring Posts ......................................................................................................502 16.1 Mooring Buoys ...................................................................................................................................502 16.1.1 16.1.2 16.1.3

Principle of Design ..................................................................................................................502 Tractive Force Acting on Mooring Buoy ..................................................................................503 Design of Individual Parts of Mooring Buoy ............................................................................504 [1] Mooring Anchor ................................................................................................................504 [2] Sinker and Sinker Chain...................................................................................................504 [3] Ground Chain ...................................................................................................................505 [4] Main Chain .......................................................................................................................506 [5] Floating Body ...................................................................................................................507 16.2 Mooring Posts ....................................................................................................................................507

Chapter 17 Other Types of Mooring Facilities.........................................................................................................508 17.1 Quaywall of Wave-Absorbing Type ................................................................................................508 17.1.1 17.1.2

Principle of Design ..................................................................................................................508 Determination of Structural Form ............................................................................................508 17.2 Cantilever Sheet Pile Quaywall .......................................................................................................509 17.2.1 Principle of Design ..................................................................................................................509 17.2.2 External Forces Acting on Sheet Pile Wall..............................................................................510 17.2.3 Determination of Cross Section of Sheet Piles ....................................................................... 511 17.2.4 Determination of Embedded Length of Sheet Piles ................................................................ 511 17.2.5 Examination of Displacement of Sheet Pile Crown................................................................. 511 17.2.6 External Forces during Construction.......................................................................................512 17.2.7 Detailed Design.......................................................................................................................512 17.3 Sheet Pile Quaywall with Batter Anchor Piles ..............................................................................512 17.3.1 Principle of Design ..................................................................................................................512 -xv-


17.3.2 17.3.3 17.3.4 17.3.5 17.3.6

17.4

17.5

External Forces Acting on Sheet Pile Wall with Batter Anchor Piles ...................................... 513 Calculation of Horizontal and Vertical Forces Acting on Connecting Point ............................ 513 Determination of Cross Sections of Sheet Pile and Batter Anchor Pile.................................. 513 Determination of Embedded Lengths of Sheet Pile and Batter Anchor Pile........................... 513 Detailed Design ...................................................................................................................... 513 Sheet Pile Quaywall with Batter Piles in Front ............................................................................. 514 17.4.1 Principle of Design.................................................................................................................. 514 17.4.2 Layout and Dimensions .......................................................................................................... 515 17.4.3 Design of Sheet Pile Wall ....................................................................................................... 515 17.4.4 Design of Open-Type Superstructure ..................................................................................... 515 17.4.5 Embedded Length .................................................................................................................. 516 17.4.6 Detailed Design ...................................................................................................................... 516 Double Sheet Pile Quaywall ............................................................................................................ 516 17.5.1 Principle of Design.................................................................................................................. 516 17.5.2 External Forces Acting on Double Sheet Pile Quaywall ......................................................... 517 17.5.3 Design of Double Sheet Pile Quaywall ................................................................................... 517

Chapter 18 Transitional Parts of Quaywalls ............................................................................................................ 519 18.1 18.2 18.3 18.4

Principle of Design ............................................................................................................................ 519 Transitional Part Where Frontal Water Depth Varies .................................................................. 519 Transitional Part Where Quaywalls of Different Type Are Connected ..................................... 519 Outward Projecting Corner .............................................................................................................. 519

Chapter 19 Ancillary Facilities ...................................................................................................................................... 520 19.1 General ............................................................................................................................................... 520 19.2 Mooring Equipment ........................................................................................................................... 520 19.3 Mooring Posts, Bollards, and Mooring Rings ............................................................................... 520 19.3.1 19.3.2 19.3.3 19.3.4

General ................................................................................................................................... 520 Arrangement of Mooring Posts, Bollards and Mooring Rings................................................. 521 Tractive Force of Vessel ......................................................................................................... 521 Structure ................................................................................................................................. 522 19.4 Fender System .................................................................................................................................. 522 19.4.1 General ................................................................................................................................... 522 19.4.2 Arrangement of Fenders......................................................................................................... 523 19.4.3 Berthing Energy of Vessel ...................................................................................................... 523 19.4.4 Selection of Fender................................................................................................................. 523 19.5 Safety Facilities ................................................................................................................................. 525 19.5.1 General ................................................................................................................................... 525 19.5.2 Skirt Guard.............................................................................................................................. 525 19.5.3 Fence and Rope ..................................................................................................................... 525 19.5.4 Signs or Notices...................................................................................................................... 525 19.5.5 Curbing ................................................................................................................................... 525 19.5.6 Fire Fighting Equipment and Alarm Systems ......................................................................... 525 19.6 Service Facilities ............................................................................................................................... 525 19.6.1 General ................................................................................................................................... 525 19.6.2 Lighting Facilities .................................................................................................................... 525 19.6.3 Facilities for Passenger Embarkation and Disembarkation .................................................... 525 19.6.4 Vehicle Ramp ......................................................................................................................... 526 19.6.5 Water Supply Facilities ........................................................................................................... 526 19.6.6 Drainage Facilities .................................................................................................................. 526 19.6.7 Fueling and Electric Power Supply Facilities .......................................................................... 526 19.6.8 Signs or Notices...................................................................................................................... 527 19.7 Stairways and Ladders ..................................................................................................................... 527 19.8 Lifesaving Facilities ........................................................................................................................... 527 19.9 Curbing ............................................................................................................................................... 527 19.10 Vehicle Ramp..................................................................................................................................... 527 19.11 Signs, Notices and Protective Fences ........................................................................................... 527 19.11.1 General ................................................................................................................................... 527 19.11.2 Provision of Signs ................................................................................................................... 527 19.11.3 Types and Location of Signs .................................................................................................. 528 19.11.4 Position of Sign....................................................................................................................... 528 19.11.5 Structure of Sign ..................................................................................................................... 529 19.11.6 Materials ................................................................................................................................. 530 19.11.7 Maintenance and Management .............................................................................................. 530 19.11.8 Protective Fences ................................................................................................................... 530 19.11.9 Barricades............................................................................................................................... 531 -xvi-

CONTENTS

19.12 Lighting Facilities ...............................................................................................................................531 19.12.1 General ...................................................................................................................................531 19.12.2 Standard Intensity of Illumination ............................................................................................531 [1] Definition ..........................................................................................................................531 [2] Standard Intensity of Illumination for Outdoor Lighting ....................................................531 [3] Standard Intensity of Illumination for Indoor Lighting .......................................................532 19.12.3 Selection of Light Source ........................................................................................................532 19.12.4 Selection of Lighting Equipment..............................................................................................534 [1] Outdoor Lighting...............................................................................................................534 [2] Indoor Lighting..................................................................................................................534 19.12.5 Design of Lighting ...................................................................................................................535 19.12.6 Maintenance and Management...............................................................................................537 [1] Inspections .......................................................................................................................537 [2] Cleaning and Repair.........................................................................................................538

Chapter 20 Aprons ...........................................................................................................................................................540 20.1 Principle of Design ............................................................................................................................540 20.2 Type of Apron .....................................................................................................................................540 20.2.1 20.2.2 20.2.3

20.3 20.4 20.5

20.6

20.7

Width ......................................................................................................................................540 Gradient ..................................................................................................................................540 Type of Pavement ...................................................................................................................540 Countermeasures against Settlement of Apron............................................................................540 Load Conditions .................................................................................................................................541 Design of Concrete Pavement ........................................................................................................541 20.5.1 Design Conditions ...................................................................................................................541 20.5.2 Composition of Pavement .......................................................................................................542 20.5.3 Joints.......................................................................................................................................545 20.5.4 Tie-Bar and Slip-Bar................................................................................................................547 20.5.5 End Protection.........................................................................................................................547 Design of Asphalt Pavement ...........................................................................................................547 20.6.1 Design Conditions ...................................................................................................................547 20.6.2 Composition of Pavement .......................................................................................................548 20.6.3 End Protection.........................................................................................................................551 Design of Concrete Block Pavement..............................................................................................551 20.7.1 Design Conditions ...................................................................................................................551 20.7.2 Composition of Pavement .......................................................................................................552 20.7.3 Joints.......................................................................................................................................553

Chapter 21 Foundation for Cargo Handling Equipment .......................................................................................554 21.1 Principle of Design ............................................................................................................................554 21.2 External Forces Acting on Foundation ...........................................................................................554 21.3 Design of Foundation with Piles ......................................................................................................555 21.3.1 21.3.2

Concrete Beam .......................................................................................................................555 Bearing Capacity of Piles ........................................................................................................555 21.4 Design of Foundation without Piles ................................................................................................556 21.4.1 Examination of Effects on Wharf.............................................................................................556 21.4.2 Concrete Beam .......................................................................................................................556

Part IX Other Port Facilities Chapter 1 Port Traffic Facilities .................................................................................................................................559 1.1

General................................................................................................................................................559 1.1.1 1.1.2

1.2

1.3

Scope of Application ...............................................................................................................559 Operation and Maintenance of Facilities for Land Traffic........................................................559 Roads ..................................................................................................................................................559 1.2.1 General ...................................................................................................................................559 1.2.2 Design Vehicles ......................................................................................................................559 1.2.3 Roadways and Lanes..............................................................................................................559 1.2.4 Clearance Limit .......................................................................................................................560 1.2.5 Widening of Roads at Bends...................................................................................................561 1.2.6 Longitudinal Slope...................................................................................................................561 1.2.7 Level Crossings.......................................................................................................................562 1.2.8 Pavement ................................................................................................................................562 1.2.9 Signs .......................................................................................................................................563 Car Parks ............................................................................................................................................564 1.3.1 General ...................................................................................................................................564 -xvii-


1.3.2

1.4 1.5 1.6

Size and Location ................................................................................................................... 564

Railways ............................................................................................................................................. 567 Heliports .............................................................................................................................................. 567 Tunnels ............................................................................................................................................... 567 1.6.1 1.6.2 1.6.3 1.6.4 1.6.5 1.6.6 1.6.7 1.6.8 1.6.9 1.6.10

1.7

General ................................................................................................................................... 567 Principle of Planning and Design............................................................................................ 567 Depth of Immersion ................................................................................................................ 568 Structure and Length of Immersed Tunnel Elements ............................................................. 568 Ventilation Towers .................................................................................................................. 568 Access Roads ......................................................................................................................... 569 Calculation of Stability of Immersed Tunnel Section .............................................................. 569 Design of Immersed Tunnel Elements.................................................................................... 569 Joints ...................................................................................................................................... 570 Control and Operation Facilities ............................................................................................. 570 Bridges ................................................................................................................................................ 570 1.7.1 General ................................................................................................................................... 570 1.7.2 Design Requirements ............................................................................................................. 570 1.7.3 Structural Durability ................................................................................................................ 571 1.7.4 Fender System ....................................................................................................................... 571

Chapter 2 Cargo Sorting Facilities ........................................................................................................................... 573 2.1 2.2 2.3 2.4

General ............................................................................................................................................... 573 Cargo Sorting Areas ......................................................................................................................... 573 Quay Sheds ....................................................................................................................................... 573 Cargo Handling Equipment ............................................................................................................. 573 2.4.1 2.4.2 2.4.3

2.5 2.6 2.7

General ................................................................................................................................... 573 Oil Handling Equipment .......................................................................................................... 574 Operation and Maintenance of Cargo Handling Equipment ................................................... 574 Timber Sorting Areas ........................................................................................................................ 574 Sorting Facilities for Marine Products ............................................................................................ 575 Sorting Facilities for Hazardous Cargo .......................................................................................... 575

Chapter 3 Storage Facilities ....................................................................................................................................... 576 3.1 3.2 3.3

General ............................................................................................................................................... 576 Yards for Dangerous Cargo and Oil Storage Facilities ............................................................... 576 Other Storage Facilities .................................................................................................................... 576

Chapter 4 Facilities for Ship Services ..................................................................................................................... 577 4.1 4.2

General ............................................................................................................................................... 577 Water Supply Facilities ..................................................................................................................... 577

Chapter 5 Facilities for Passenger ........................................................................................................................... 578 5.1

Facilities for Passenger Boarding ................................................................................................... 578 5.1.1 5.1.2 5.1.3 5.1.4

5.2

General ................................................................................................................................... 578 Structural Types...................................................................................................................... 578 Design of Facilities for Passenger Boarding........................................................................... 578 Ancillary Facilities ................................................................................................................... 578 Passenger Building ........................................................................................................................... 579 5.2.1 General ................................................................................................................................... 579 5.2.2 Design of Passenger Buildings............................................................................................... 579 5.2.3 Ancillary Facilities ................................................................................................................... 579

Part X Special Purpose Wharves Chapter 1 Container Terminals ................................................................................................................................. 581 1.1 1.2

Principle of Design ............................................................................................................................ 581 Design of Mooring Facilities ............................................................................................................ 582 1.2.1 1.2.2 1.2.3

1.3

Length and Water Depth of Berths ......................................................................................... 582 Mooring Equipment................................................................................................................. 582 Fender System ....................................................................................................................... 583 Design of Land Facilities .................................................................................................................. 583 1.3.1 Apron ...................................................................................................................................... 583 1.3.2 Container Cranes.................................................................................................................... 583 1.3.3 Container Yard........................................................................................................................ 583 1.3.4 Container Freight Station........................................................................................................ 583 1.3.5 Maintenance Shop.................................................................................................................. 583 -xviii-

CONTENTS

1.3.6 1.3.7 1.3.8

Administration Building............................................................................................................583 Gates.......................................................................................................................................583 Ancillary Facilities....................................................................................................................583

Chapter 2 Ferry Terminals ..........................................................................................................................................584 2.1 2.2

Principle of Design ............................................................................................................................584 Design of Mooring Facilities .............................................................................................................585 2.2.1 2.2.2 2.2.3 2.2.4

2.3

2.4 2.5

Length and Water Depth of Berths..........................................................................................585 Mooring Equipment .................................................................................................................585 Fender System........................................................................................................................586 Protection Works against Scouring .........................................................................................586 Design of Vehicle Ramp ...................................................................................................................586 2.3.1 Width, Length, Gradient, and Radius of Curvature .................................................................586 2.3.2 Ancillary Facilities and Signs...................................................................................................586 2.3.3 Design of Moving Parts ...........................................................................................................586 Facilities for Passenger Boarding ...................................................................................................586 2.4.1 Width, Length, Gradient, and Ancillary Facilities.....................................................................587 2.4.2 Design of Moving Parts ...........................................................................................................587 Design of Other Facilities .................................................................................................................587 2.5.1 Roads......................................................................................................................................587 2.5.2 Passageways ..........................................................................................................................587 2.5.3 Car Parks ................................................................................................................................587 2.5.4 Passenger Terminals ..............................................................................................................588 2.5.5 Safety Equipment....................................................................................................................588

Part XI Marinas Chapter 1 Introduction ..................................................................................................................................................589 Chapter 2 Main Dimensions of Target Boats ........................................................................................................590 Chapter 3 Navigation Channels and Basins..........................................................................................................591 3.1 3.2 3.3

General................................................................................................................................................591 Navigation Channels .........................................................................................................................591 Mooring Basins ..................................................................................................................................591

Chapter 4 Protective Facilities ...................................................................................................................................592 Chapter 5 Mooring Facilities .......................................................................................................................................593 5.1 5.2 5.3

General................................................................................................................................................593 Design Conditions for Mooring Facilities .......................................................................................593 Floating Piers .....................................................................................................................................595 5.3.1 5.3.2 5.3.3 5.3.4 5.3.5 5.3.6

5.4 5.5

General ...................................................................................................................................595 Structure..................................................................................................................................595 Examination of Safety .............................................................................................................595 Structural Design.....................................................................................................................596 Mooring Method ......................................................................................................................596 Access Bridges .......................................................................................................................596 Ancillary Facilities ..............................................................................................................................597 Lifting / Lowering Frame Facilities ..................................................................................................597

Chapter 6 Facilities for Ship Services......................................................................................................................598 6.1 6.2

General................................................................................................................................................598 Land Storage Facilities .....................................................................................................................598

Chapter 7 Land Traffic Facilities................................................................................................................................599

INDEX

-xix-


-xx-

Part I General

PART I GENERAL

Part I General Chapter 1 General Rules 1.1 Scope of Application The Ministerial Ordinance stipulating the Technical Standards for Port and Harbour Facilities (Ministry of Transport Ordinance No. 30, 1974; hereafter referred to simply as the Ministerial Ordinance) and the Notification stipulating the Details of Technical Standards for Port and Harbour Facilities (Ministry of Transport Notification No. 181, 1999; hereafter referred to simply as the Notification), both of which have been issued in line with Article 56-2 of the “Port and Harbour Law”, shall be applied to the construction, improvement, and maintenance of port and harbor facilities. [Commentary] (1) The Ministerial Ordinance and the Notification (hereafter collectively referred to as the Technical Standards) apply not to the port and harbor facilities stipulated in Article 2 of the “Port and Harbour Law”, but rather to the port and harbor facilities stipulated in Article 19 of the Port and Harbour Law Enforcement Order. Accordingly the Technical Standards also apply to facilities like navigation channels, basins, protective facilities and mooring facilities of the marinas and privately owned ports, which are found in outside of the legally designated port areas. (2) Since the Technical Standards covers a wide rage of facilities, there will be cases where the items shown in the Technical Standards may be inadequate for dealing with planning, designing, constructing, maintaining or repairing of a particular individual structure of a port or harbor. There is also possibility that new items may be added in the future in line with technical developments or innovations. With regard to matters for which there are no stipulations in the Technical Standards, appropriate methods other than those mentioned in the Technical Standards may be adopted, after confirming the safety of a structure in consideration using appropriate methods such as model tests or trustworthy numerical calculations (following the main items of the Technical Standards). (3) Figure C- 1.1.1 shows the statutory structure of the Technical Standards. Port and Harbour Law [Article 56-2] (technical standards for port and harbour facilities)

Port and Harbour Law Enforcement Order [Article 19] (stipulation of facilities covered)

Port and Harbour Law Enforcement Regulations [Article 28] (stipulation of facilities excluded from coverage)

Port and Harbour Law Enforcement Regulations

Port and Harbour Law Enforcement Order

The Technical Standards The Ministerial Ordinance The Notification Fig. C- 1.1.1 Statutory Structure of the Technical Standards for Port and Harbour Facilities

(4) This document is intended to help individuals concerned with correct interpretation of the Technical Standards and to facilitate right application of the Ministerial Ordinance and the Notification. This document is made up of the main items, along with reference sections marked Commentary and Technical Notes, which supplement the main items. The texts in large letters are the main items that describe the parts of the Notification and the basic items that must be obeyed, regarding the items related to the Notification. The sections marked Commentary mainly give the background to and the basis for the Notification, etc. The sections marked Technical Notes provide investigation methods and/or standards that will be of reference value, when executing actual design works, specific examples of structures, and other related materials. (5) Design methods can be broadly classified into the methods that use the safety factors and the methods that use the indices based on probability theory, according to the way of judging the safety of structures. A safety factor is not an index that represents the degree of safety quantitatively. Rather, it is determined through experience to compensate for the uncertainty in a variety of factors. In this document, the safety factors indicate values that are considered by experience to be sufficiently safe under standard conditions. Depending on the conditions, it may be acceptable to lower the values of safety factors, but when doing so it is necessary to make a decision using prudent judgement based on sound reasoning. In the case that the probability distributions of loads and structure strengths can be adequately approximated, it is possible to use a reliability design method. Unlike the more traditional design methods in which safety factors are used, a reliability design method makes it possible to gain a quantitative understanding of the -1-


likelihood of the failure of structure in question and then to keep the likelihood below a certain allowable value. With a reliability design method, design is carried out by using the partial safety factors and reliability indices. Formally speaking, the limit state design method can be classified as one form of reliability design method.

1.2 Definitions The terms used in the Notification are based on the terminology used in the Ministerial Ordinance; in addition, the meanings of the following terms as stipulated in the law or notification are cited. (1) Dangerous articles: This term refers to those that are designated in the Notification stipulating the “Types of Hazardous Goods” for the “Port Regulation Law Enforcement Regulations” (Ministry of Transport Notification No. 547, 1979). (2) Datum level for construction work: This is the standard water level used when constructing, improving or maintaining port and harbor facilities, and is equal to the chart datum level (specifically the chart datum for which the height is determined based on the provisions of Article 9 (8) of the “Law for Hydrographic Activities” (Law No. 102, 1950)). However, in the case of port and harbor facilities in lakes and rivers for which there is little tidal influence, in order to ensure the safe use of the port or harbor in question, the datum level for construction work shall be determined while considering the conditions of extremely low water level that may occur during a drought season. [Commentary] In addition to the terms defined above, the meanings of the following terms are listed below. (1) Super-large vessel: A cargo ship with a deadweight tonnage of 100,000 t or more, except in the case of LPG carriers and LNG carriers, in which case the gross tonnage is 25,000 t or more. (2) Passenger ship: A vessel with a capacity of 13 or more passengers. (3) Pleasure boat: A yacht, motorboat or other vessel used for sport or recreation.

1.3 Usage of SI Units [Commentary] In line with the provisions in the “Measurement Law” (Law No. 51, May 20, 1992), with the aim of executing a smooth switchover to SI units, the Ministry of Agriculture, Forestry and Fisheries, the Ministry of Transport and the Ministry of Construction have concluded to use the International System of Units in their public work projects starting from April 1999.

-2-

PART I GENERAL

Table C- 1.3.1 Conversion Factors from Conventional Units to SI Units Number

Quantity

Non-SI units

SI units

Conversion factor

1

Length

µ

m

1µ ＝ 1µm

2

Mass

kgf•s2/m

kg

1kgf•s2/m ＝ 9.80665kg

3

Acceleration

Gal

m/s2

1Gal ＝ 0.01m/s2

kgf

N

1kgf ＝ 9.80665N

dyn

N

1dyn ＝ 10µN

kgf•m

N•m

1kgf•m ＝ 9.80665N•m

Pa

1kgf/cm2 ＝ 9.80665 × 104Pa ＝ 9.80665 × 10-2MPa 1kgf/cm2 ＝ 9.80665 × 10-2N/mm2

4 5 6

7

Force Moment of a force

Pressure

8

9

10 11

kgf/cm2

N/mm2

mHg

Stress

Work (energy)

kgf/cm2

Pa

1mHg ＝ 133.322kPa

Pa

1kgf/cm2 ＝ 9.80665 × 104Pa ＝ 9.80665 × 10-2MPa 1kgf/cm2 ＝ 9.80665 × 10-2N/mm2

N/mm2

kgf•m

J

1kfg•m ＝ 9.80665J

erg

J

1erg ＝ 100nJ

12

Power

PS HP

W

1PS ＝ 735.499W 1HP ＝ 746.101W

13

Quantity of heat

cal

J W•s

1cal ＝ 4.18605J 1cal ＝ 4.18605W•s

14

Thermal conductivity

cal/(h•m•ºC)

W/(m•ºC)

1cal/(h•m•ºC) ＝ 0.001163W/(m•ºC)

15

Heat conduction coefficient

cal/(h•m2•ºC)

W/(m2•ºC)

1cal/(h•m2•ºC) ＝ 0.001163W/(m2•ºC)

16

Specific heat capacity

cal/(kg•ºC)

J/(kg•ºC)

1cal/(kg•ºC) ＝ 4.18605J/(kg•ºC)

17

Sound pressure level

－

dB

1phon ＝ 1dB

-3-


Chapter 2 Datum Level for Construction Work [Commentary]

The datum level for port and harbor construction work is the standard water level that shall form the basis for the planning, design, and construction of facilities. The chart datum level shall be used as the datum level for construction work. [Technical Notes] (1) Chart Datum Level The chart datum level is set as the level below the mean sea level by the amount equal to or approximately equivalent to the sum of the amplitueds of the four major tidal constituents (M2, S2, K1, and O1 tides), which are obtained from the harmonic analysis of tidal observation data. Here M2 is the principal lunar semi-diurnal tide, S2 is the principal solar semi-diurnal tide, K1 is the luni-solar diurnal tide, and O1 is the principal lunar diurnal tide. Note that the heights of rocks or land marks shown on the nautical charts are the elevation above the mean sea level, which is the long-term average of the hourly sea surface height at the place in question. (In the case that the observation period is short, however, corrections for seasonal fluctuations should be made when determining the mean sea level.) The difference in height between the chart datum level and the mean sea level is referred to as Z0. (2) International Marine Chart Datum The International Hydrographic Organization (IHO) has decided to adopt the Lowest Astronomical Tide (LAT) as the international marine chart datum, and issued a recommendation to this effect to the Hydrographic Departments in various countries throughout the world in June 1997. The LAT is defined as the lowest sea level that is assumed to occur under the combination of average weather conditions and generally conceivable astronomical conditions. In actual practice, tide levels for at least 19 years are calculated using harmonic constants obtained from at least one year’s worth of observations, and then the lowest water level is taken as the LAT. However, in the case of Japan, the chart datum level is obtained using the old method described in (1) above (approximate lowest water level). There will be no switchover to the LAT in the near future in Japan, but it is planned to meet the IHO recommendation by stating the difference between the LAT and the chart datum level in tide tables published by the Hydrographic Department of Maritime Safety Agency, Ministry of Land, Infrastructure, and Transport, Japan.

-4-

PART I GENERAL

Chapter 3 Maintenance In order to maintain the functions of port and harbor facilities at a satisfactory service level and to prevent deterioration in the safety of such facilities, comprehensive maintenance including inspections, evaluations, repairs, etc. shall be carried out, in line with the specific characteristics of the port or harbor in question. [Commentary] (1) “Maintenance” refers to a system consisting of a series of linked activities involving the efficient detection of changes in the state of serviceability of the facilities and the execution of effective measures such as rational evaluation, repair, and reinforcement. (2) Port and harbor facilities must generally remain in service for long periods of time, during which the functions demanded of the facilities must be maintained. It is thus essential not only to give due consideration when initially designing the structures in question, but also to carry out proper maintenance after the facilities have been put into service. (3) A whole variety of data concerning maintenance (specifically, inspections, checks, evaluations, repair, reinforcement work, etc.) must be recorded and stored in a standard format. Maintenance data kept in good systematic order is the basic information necessary for carrying out appropriate evaluation of the level of soundness of the facilities in question, and executing their maintenance and repairs. At the same time the maintenace data is useful when taking measures against the deterioration of the facilities as a whole and when investigating the possibility in the life cycle cost reduction of the facilities. (4) When designing a structure, it is necessary to give due consideration to the system of future maintenance and to select the types of structures and the materials used so that future maintenance will be easily executed, while reflecting this aspect in the detailed design.• [Technical Notes] (1) The concepts of the terms relating to maintenance are as follows: Inspection / checking: • • • •Activities to investigate the state of a structure, the situation regarding damage and the remaining level of function, along with related administrative work: mainly composed of periodic and special inspections

Maintenance

Evaluation: • • • • • • • • • • • • • • • Evaluation of the level of soundness based on the results of inspection / checking, and judgement of the necessity or otherwise of repairs etc. Maintenance: • • • • • • • • • • • • • Activities carried out with the aim of holding back the physical deterioration of a structure and keeping its function within acceptable levels. Repair / reinforcement: • • Activities in which a structure that has deteriorated physically and/ or functionally is partially reconstructed in order to restore the required function and/or structure.

(2) With regard to the procedure for maintenance, it is a good idea to draw up a maintenance plan for each structure while considering factors like the structural form, the tendency to deteriorate and the degree of importance, and then to implement maintenance work based on this plan. (3) For basic and common matters concerning maintenance, refer to the “Manual for Maintenance and Repair of Port and Harbor Structures”.

-5-


-6-

Part II Design Conditions

PART II DESIGN CONDITIONS

Part II Design Conditions Chapter 1 General In designing port and harbor facilities, the design conditions shall be chosen from the items listed below by taking into consideration the natural, service and construction conditions, the characteristics of materials, the environmental impacts, and the social requirements for the facilities. (1) Ship dimensions (2) External forces produced by ships (3) Winds and wind pressure (4) Waves and wave force (5) Tide and extraordinary sea levels (6) Currents and current force (7) External forces acting on floating structures and their motions (8) Estuarine hydraulics and littoral drift (9) Subsoil (10) Earthquakes and seismic force (11) Liquefaction (12) Earth pressure and water pressure (13) Deadweight and surcharge (14) Coefficient of friction (15) Other necessary design conditions [Commentary] The design conditions should be determined carefully, because they exercise great influence upon the safety, functions, and construction cost of the facilities. The design conditions listed above are just those that have a large influence on port and harbor facilities. They are generally determined according to the results of surveys and tests. Thus, the design conditions should be precisely determined upon full understanding of the methods and results of such investigations and tests. In the case of temporary structures, the design conditions may be determined while considering also the length of service life. [Technical Notes] (1) In designing port and harbor facilities, the following matters should be taken into consideration. (a) Functions of the facilities Since facilities often have multiple functions, care should be exercised so that all functions of the facilities will be exploited fully. (b) Importance of the facilities The degree of importance of the facilities should be considered in order to design the facilities by taking appropriate account of safety and broad economic implications. The design criteria influenced by importance of facilities are those of environmental conditions, design seismic coefficient, lifetime, loads, safety factor, etc. In determining the degree of importance of the facilities, the following criteria should be taken into consideration. • Influence upon human lives and property if the facilities are damaged. • Impact on society and its economy if the facilities are damaged. • Influence upon other facilities if the facilities are damaged. • Replaceability of the facilities. (c) Lifetime The length of lifetime should be taken into account in determining the structure and materials of the facilities and also in determining the necessity for and extent of the improvement of the existing facilities. Lifetime of the facilities should be determined by examinig the following: • Operational function of the facilities The number of years until the facilities can no longer be usable due to the occurrence of problems in terms of the function of the facilities, for example the water depth of a mooring basin becoming insufficient owing to the increase in vessel size. • Economic viewpoint of the facilities The number of years until the facilities become economically uncompetitive with other newer facilities (unless some kind of improvements are carried out). -7-


• Social function of the facilities The number of years until the functions of the facilities that constituted the original purpose become unnecessary or until different functions are called for the facilities due to new port planning etc. • Physical property of the facilities The number of years until it is no longer possible to maintain the strength of materials composing the structures at the specified level due to processes such as corrosion or weathering of these materials. (d) Encounter probability The encounter probability is intimately linked with the lifetime length. The encounter probability E1 is obtained using equation (1.1.1) 1) E1 = 1 ( 1 1 ¤ T 1 ) where L1： lifetime length T 1： return period

L1

(1.1.1)

(e) Environmental conditions Not only the wave, seismic, topographical and soil conditions, which have direct influences on the design of the facilities, but also the water quality, bottom material, animal and plant life, atmospheric conditions and rising sea level due to global warming should be taken into consideration. (f) Materials It is necessary to consider the physical external forces, deterioration, lifetime, structural type, construction works, cost, and influence on the environment and landscape when selecting the materials. It is most important to ensure the reguired quality. In recent years, in addition to more traditional materials, new materials such as stainless steels, titanium and new rubbers, and recycled materials such as slag, coal ash and dredged sediment have begun to be used. (g) Construction method In order to carry out design rationally, it is necessary to give sufficient consideration to the construction method. (h) Work accuracy It is necessary to carry out design considering the accuracy of construction works that can be maintained in actual projects. (i) Construction period In the case that the construction period is stipulated, it is necessary to give consideration both to the design and the construction method, in order that it will be possible to complete construction work within the stipulated period. The construction period is generally determined by things like the availability of the materials, the construction equipment, the degree of difficulty of construction, the opening date and the natural conditions. (j) Construction costs etc. Construction costs consist of the initial investment costs and maintenance costs. All of these costs must be considered in design and construction. When doing this, it is necessary to consider the early use of the facilities and to secure an early return on investment. There is also a design approach that the facilities are put into service step by step as the construction progresses, while ensuring the safety of service / construction. Note also that the initial investment costs mentioned here include compensation duties. When carrying out design etc., due consideration must be given to things like the structural type and the construction method, since the construction costs will depend on these things. [Reference] 1) Borgman, L. E.: “Risk criteria”, Proc. ASCE, Vol. 89, No. WW3, 1963, pp.1-35.

-8-


Chapter 2 Vessels 2.1 Dimensions of Target Vessel (Notification Article 21) The principal dimensions of the target vessel shall be set using the following method: (1) In the case that the target vessel can be identified, use the principal dimensions of that vessel. (2) In the case that the target vessel cannot be identified, use appropriate principal dimensions determined by statistical methods. [Technical Notes] (1) Article 1, Clause 2 of the Ministerial Ordinance stipulates that the “target vessel” is the vessel that has the largest gross tonnage out of those that are expected to use the port or harbor facilities in question. Accordingly, in the case that the target vessel can be identified, the principal dimensions of this vessel should be used. (2) In the case that the target vessel cannot be identified in advance, such as in the case of port and harbor facilities for public use, the principal dimensions of the target vessel may be determined by referring to Table T- 2.1.1. In this table, the tonnages (usually either gross or deadweight tonnage) are used as representative indicators. (3) Table T- 2.1.1 lists the “principal dimensions of vessels for the case that the target vessel cannot be identified” by tonnage level. These values have been obtained through methods such as statistical analysis 1),2), and they mainly represent the 75% cover ratio values for each tonnage of vessels. Accordingly, for any given tonnage, there will be some vessels that have principal dimensions that exceed the values in the table. There will also be vessels that have a tonnage greater than that of the target vessel listed in the table, but still have principal dimensions smaller than those of the target vessel. (4) Table T- 2.1.1 has been obtained using the data from “Lloyd’s Maritime Information June ’95” and “Nihon Senpaku Meisaisho” (“Detailed List of Japanese Vessels”; 1995 edition). The definitions of principal dimensions in the table are shown in Fig. T- 2.1.1. (5) Since the principal dimensions of long distance ferries that sail over 300km tend to have different characteristics from those of short-to-medium distance ferries, the principal dimensions are listed separately for “long distance ferries” and “short-to-medium distance ferries.” (6) Since the principal dimensions of Japanese passenger ships tend to have different characteristics from those of foreign passenger ships, the principal dimensions are listed separately for “Japanese passenger ships” and “foreign passenger ships”. (7) The mast height varies considerably even for vessels of the same type with the same tonnage, and so when designing facilities like bridges that pass over navigation routes, it is necessary to carry out a survey on the mast heights of the target vessels. (8) In the case that the target vessel is known to be a small cargo ship but it is not possible to identify precisely the demensions of the ship in advance, the principal dimensions of “small cargo ships” can be obtained by referring to Table T- 2.1.2. The values in Table T- 2.1.2 have been obtained using the same kind of procedure as those in Table T- 2.1.1, but in the case of such small vessels there are large variations in the principal dimensions and so particular care should be exercised when using Table T- 2.1.2. (9) Tonnage The definitions of the various types of tonnage are as follows: (a) Gross tonnage The measurement tonnage of sealed compartments of a vessel, as stipulated in the “Law Concerning the Measurement of the Tonnage of Ships”. The “gross tonnage” is used as an indicator that represents the size of a vessel in Japan’s maritime systems. Note however that there is also the “international gross tonnage”, which, in line with the provisions in treaties etc., is also used as an indicator that represents the size of a vessel, but mainly for vessels that make international sailings. The values of the “gross tonnage” and the “international gross tonnage” can differ from one another; the relationship between the two is stipulated in Article 35 of the “Enforcement Regulations for the Law Concerning the Measurement of the Tonnage of Ships” (Ministerial Ordinance No. 47, 1981). (b) Deadweight tonnage The maximum weight, expressed in tons, of cargo that can be loaded onto a vessel. (c) Displacement tonnage The amount of water, expressed in tons, displaced by a vessel when it is floating at rest. (10) For the sake of consistency, equation (2.1.1) shows the relationship between the deadweight tonnage (DWT) and the gross tonnage (GT) for the types of vessels that use the deadweight tonnage as the representative indicator 1). For each type of vessels, the equation may be applied if the tonnage is within the range shown in Table T- 2.1.1. -9-


GT = 0.541DWT GT = 0.880DWT GT = 0.553DWT GT = 0.808DWT

64748

Cargo ships: Container ships: Oil tankers: Roll-on/roll-off vessels:

(2.1.1)

where GT ： gross tonnage DWT ： deadweight tonnage (11) Tables T-2.1.3 to T-2.1.6 list the frequency distribution of the principal dimensions of general cargo ships, bulk cargo carriers, container ships, and oil tankers, which were analyzed by the Systems Laboratory of Port and Harbour Research Institute (PHRI) using the data from “Lloyd’s Maritime Informations Services (June ’98)”.

Length overall Load water line

Fore perpendicular

Full load draft

Moulded breadth

Moulded depth

After perpendicular

Length between perpendiculars

Fig. T- 2.1.1 Definitions of Principal Dimensions of Vessel Table T- 2.1.1 Principal Dimensions of Vessels for the Case That the Target Vessel Cannot Be Identified 1. Cargo ships Deadweight tonnage (DWT) 1,000 ton 2,000 3,000 5,000 10,000 12,000 18,000 30,000 40,000 55,000 70,000 90,000 100,000 150,000

Length overall (L)

Molded breadth (B) 10.9 m 13.1 14.6 16.8 19.9 21.0 23.6 27.5 29.9 32.3 32.3 38.1 39.3 44.3

67 m 83 94 109 137 144 161 185 200 218 233 249 256 286

Full load draft (d) 3.9 m 4.9 5.6 6.5 8.2 8.6 9.6 11.0 11.8 12.9 13.7 14.7 15.1 16.9

2. Container ships Deadweight tonnage (DWT) 30,000 ton 40,000 50,000 60,000

Length overall (L)

Molded breadth (B)

218 m 244 266 286

30.2 m 32.3 32.3 36.5

-10-

Full load draft (d) 11.1 m 12.2 13.0 13.8


3. Ferries 3-A Short-to-medium distance ferries (sailing distance less than 300km) Gross tonnage (GT) 400 ton 700 1,000 2,500 5,000 10,000

Length overall (L)

Molded breadth (B) 11.8 m 13.5 14.7 18.3 21.6 23.0

50 m 63 72 104 136 148

Full load draft (d) 3.0 m 3.4 3.7 4.6 5.3 5.7

3-B Long distance ferries (sailing distance 300km or more) Gross tonnage (GT) 6,000 ton 10,000 13,000 16,000 20,000 23,000

Length overall (L)


142 m 167 185 192 192 200


4. Roll-on/roll-off vessels Deadweight tonnage (DWT) 400 ton 1,500 2,500 4,000 6,000 10,000

Length overall (L)


75 m 97 115 134 154 182


5. Passenger ships 5-A Japanese passenger ships Gross tonnage (GT) 2,000 ton 4,000 7,000 10,000 20,000 30,000

Length overall (L)


83 m 107 130 147 188 217


5-B Foreign passenger ships Gross tonnage (GT) 20,000 ton 30,000 50,000 70,000

Length overall (L)

Molded breadth (B) 25.7 m 28.4 32.3 35.2

180 m 207 248 278

Full load draft (d) 8.0 m 8.0 8.0 8.0

6. Pure car carriers Gross tonnage (GT) 500 ton 1,500 3,000 5,000 12,000 18,000 25,000

Length overall (L)

Molded breadth (B) 11.8 m 15.7 18.8 21.5 27.0 30.0 32.3

70 m 94 114 130 165 184 200

-11-

Full load draft (d) 3.8 m 5.0 5.8 6.6 8.0 8.8 9.5


7. Oil tankers Deadweight tonnage (DWT) 1,000 ton 2,000 3,000 5,000 10,000 15,000 20,000 30,000 50,000 70,000 90,000

Length overall (L)

Molded breadth (B)

Full load draft (d) 4.0 m 4.9 5.5 6.4 7.9 8.9 9.6 10.9 12.6 13.9 15.0

10.2 m 12.6 14.3 16.8 20.8 23.6 25.8 29.2 32.3 38.0 41.1

61 m 76 87 102 127 144 158 180 211 235 254

Table T- 2.1.2 Principal Dimensions of Small Cargo Ships Deadweight tonnage (DWT) 500 ton 700

Length overall (L)

Molded breadth (B)

51 m 57

Full load draft (d)

9.0 m 9.5

Table T-2.1.3 Frequency Distributions of Principal Dimensions of General Cargo Ships (a) DWT - Length overall L

(b) DWT - Breadth

B

(c) DWT - Full load draft d

-12-

3.3 m 3.4


Table T-2.1.4 Frequency Distributions of Principal Dimensions of Bulk Cargo Carriers (a) DWT - Length overall L

(b) DWT - Breadth

B


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Table T-2.1.5 Frequency Distributions of Principal Dimensions of Container Ships (a) DWT - Length overall L

(b) DWT - Breadth

B


unknown unknown

(d) DWT - TEU

-14-


Table T-2.1.6 Frequency Distributions of Principal Dimensions of Oil Tankers (a) DWT - Length overall L

(b) DWT - Breadth

B


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2.2 External Forces Generated by Vessels 2.2.1 General The external forces acting on the mooring facilities when a vessel is berthing or moored shall be determined using an appropriate method, considering the dimensions of the target vessel, the berthing method and the berthing velocity, the structure of the mooring facilities, the mooring method and the properties of the mooring system, along with the influence of things like the winds, waves and tidal currents. [Commentary] (1) The following loads acting on mooring facilities should be considered when a vessel is berthing or moored: a) Loads caused by berthing of a vessel b) Loads caused by motions of a moored vessel When designing mooring facilities, the berthing force must be considered first. Then the impact forces and tractive forces on the mooring facilities due to the motions of the moored vessel, which are caused by the wave force, wind force and current force, should be considered. In particular, for the cases of the mooring facilities in the ports and harbors that face out onto the open sea with long-period waves expected to come in, of those installed in the open sea or harbor entrances such as offshore terminals, and of those in the harbors where vessels seek refuge during storms, the influence of the wave force acting on a vessel is large and so due consideration must be given to the wave force. (2) As a general rule, the berthing forces acting on the mooring facilities should be calculated based on the berthing energy of the vessel and using the load-deflection characteristics of the fenders. (3) As a general rule, the tractive forces and impact forces generated by the motions of a moored vessel should be obtained by carrying out a numerical simulation of vessel motions taking into account the wave force acting on the vessel, the wind force, the current force, and the load-deflection characteristics of the mooring system.

2.2.2 Berthing [1] Berthing Energy (Notification Article 22, Clause 1) It shall be standard to calculate the external force generated by berthing of a vessel with the following equation: MsV2 (2.2.1) E f = æ -------------ö C e C m C s C c è 2 ø In this equation, E f , M s , V, C e , C m , C s , and C c represent the following: E f： berthing energy of vessel (kJ = kN•m) M s： mass of vessel (t) V： berthing velocity of vessel (m/s) C e： eccentricity factor C m： virtual mass factor C s： softness factor (standard value is 1.0) C c： berth configuration factor (standard value is 1.0) [Commentary] In addition to the kinetic energy method mentioned above, there are also other methods of estimating the berthing energy of a vessel: for example, statistical methods, methods using hydraulic model experiments, and methods using fluid dynamics models 3). However, with these alternative methods, the data necessary for design are insufficient and the values of the various constants used in the calculations may not be sufficiently well known. Thus, the kinetic energy method is generally used. [Technical Notes] (1) If it is assumed that a berthing vessel moves only in the abeam direction, then the kinetic energy E s is equal to ( M s V 2) ¤ 2 . However, when a vessel is berthing at a dolphin, a quaywall, or a berthing beam equipped with fenders, the energy absorbed by the fenders (i.e., the berthing energy E f of the vessel) will become E s ´ f considering the various influencing factors, where f = C e ´ C m ´ C s ´ C c . (2) The vessel mass M s is taken to be the displacement tonnage (DT) of the target vessel. In the case that the target vessel cannot be identified, equation (2.2.2) 1) may be used to give the relationship between the deadweight tonnage (DWT) or the gross tonnage (GT) and the displacement tonnage (DT). -16-


log (DT) ＝ 0.550 ＋ 0.899 log (DWT) log (DT) ＝ 0.511 ＋ 0.913 log (DWT) log (DT) ＝ 0.365 ＋ 0.953 log (DWT) log (DT) ＝ 1.388 ＋ 0.683 log (GT) log (DT) ＝ 0.506 ＋ 0.904 log (GT) log (DT) ＝ 0.657 ＋ 0.909 log (DWT) log (DT) ＝ 0.026 ＋ 0.981 log (GT) log (DT) ＝ 0.341 ＋ 0.891 log (GT) log (DT) ＝ 1.915 ＋ 0.588 log (GT) log (DT) ＝ 0.332 ＋ 0.956 log (DWT)

64444744448

Cargo ships (less than 10,000DWT)： Cargo ships (10,000DWT or more)： Container ships： Ferries (long distance)： Ferries (short-to-medium distance)： Roll-on/roll-off vessels： Passenger ships (Japanese)： Passenger ships (foreign)： Car carriers： Oil tankers：

(2.2.2)

where DT： displacement tonnage (amount of water, in tons, displaced by the vessel when fully loaded) GT： gross tonnage DWT： deadweight tonnage (3) The softness factor C s represents the ratio of the remaining amount of the berthing energy after energy absorption due to deformation of the shell plating of the vessel to the initial berthing energy. It is generally assumed that no energy is absorbed in this way and so the value of C s is often given as 1.0. (4) When a vessel berths, the mass of water between the vessel and the mooring facilities resists to move out and acts just as if a cushion is placed in this space. The energy that must be absorbed by the fenders is thus reduced. This effect is considered when determining the berth configuration factor C c . It is thought that the effect depends on things like the berthing angle, the shape of the vessel’s shell plating, the under-keel clearance, and the berthing velocity, but little research has been carried out to determine it.

[2] Berthing Velocity The berthing velocity of a vessel shall be determined based on the measurement in situ or past data of similar measurements, considering the type of the target vessel, the extent to which the vessel is loaded, the position and structure of the mooring facilities, weather and oceanographic conditions, and the availability or absence of tugboats and their sizes. [Technical Notes] (1) Observing the way in which large cargo ships and large oil tankers make berthing, one notices that such vessels come to a temporary standstill, lined up parallel to the quaywall at a certain distance away from it. They are then gently pushed by several tugboats until they come into contact with the quay. When there is a strong wind blowing toward the quay, such vessels may berth while actually being pulled outwards by the tugboats. When such a berthing method is adopted, it is common to set the berthing velocity to 10 ~ 15 cm/s based on past design examples. (2) Special vessels such as ferries, roll-on/roll-off vessels, and small cargo ships berth under their own power without assistance of tugboats. If there is a ramp at the bow or stern of such a vessel, the vessel may line up perpendicular to the quay. In these cases, a berthing method different from that for larger vessels described in (1) may be used. It is thus necessary to determine berthing velocities carefully based on actualy measured values, paying attention to the type of berthing method employed by the target vessel. (3) Figure T- 2.2.1 shows the relationship between the vessel handling conditions and berthing velocity by vessel size 4); it has been prepared based on the data collected through experience. This figure shows that the larger the vessel, the lower the berthing velocity becomes; moreover, the berthing velocity must be set high if the mooring facilities is not sheltered by breakwaters etc. (4) According to the results of surveys on berthing velocity 5),6), the berthing velocity is usually less than 10 cm/s for general cargo ships, but there are a few cases where it is over 10 cm/s (see Fig. T- 2.2.2). The berthing velocity only occasionally exceeds 10 cm/s for large oil tankers that use offshore terminals (see Fig. T- 2.2.3). Even for ferries which berth under their own power, the majority berth at the velocity of less than 10 cm/s. Nevertheless, there are a few cases in which the berthing velocity is over 15 cm/s and so due care must be taken when designing ferry quays (see Fig. T- 2.2.4). It was also clear from the above-mentioned survey results that the degree to which a vessel is loaded up has a considerable influence on the berthing velocity. In other words, if a vessel is fully loaded, meaning that the under-keel clearance is small, then the berthing velocity tends to be lower, whereas if it is lightly loaded, meaning that the under-keel clearance is large, then the berthing velocity tends to be higher.

-17-


Difficulty of handling vessel / mooring facilities being shelterd or not

Difficult

exposed Good berthing

exposed Easy berthing

exposed Difficult berthing

sheltered Good berthing

sheltered

Berthing velocity (cm/s) Fig. T- 2.2.1 Relationship between Vessel Handling Conditions and Berthing Velocity by Vessel Size 4)

Open type quay

Berthing velocity (cm/s)

Wall type quay (sheet pile, gravity types)

Displacement tonnage

DT (tons)


Fig. T- 2.2.2 Berthing Velocity and Displacement Tonnage for General Cargo Ships 5)


DT (10,000 tons)

Fig. T- 2.2.3 Berthing Velocity and Displacement Tonnage for Large Oil Tankers 6)

Stern berthing


Bow berthing


DT (tons)

Fig. T- 2.2.4 Berthing Velocity and Displacement Tonnage for Longitudinal Berthing of Ferries 5)

-18-


According to the survey by Moriya et al., the average berthing velocities for cargo ships, container ships, and pure car carriers are as listed in Table T- 2.2.1. The relationship between the deadweight tonnage and berthing velocity is shown in Fig. T- 2.2.5. This survey also shows that the larger the vessel, the lower the berthing velocity tends to be. The highest berthing velocities observed were about 15 cm/s for vessels under 10,000 DWT and about 10 cm/s for vessels of 10,000 DWT or over. Table T- 2.2.1 Deadweight Tonnage and Average Berthing Velocity Deadweight tonnage (DWT)

Berthing velocity (cm/s) Cargo ships

Container ships

Pure car carriers

All vessels

1,000 class 5,000 class 10,000 class 15,000 class 30,000 class 50,000 class

8.1 6.7 5.0 4.5 3.9 3.5

7.8 7.2 4.9 4.1 3.4

4.6 4.7 4.4 -

8.1 7.2 5.3 4.6 4.1 3.4

All vessels

5.2

5.0

4.6

5.0 N=738 Poisson distribution m = 3 Poisson distribution m = 4 Weibull distribution Normal distribution

N

V (cm/s)

Cargo ships Container ships Pure car carriers

V (cm/s)

Dead weight tonnage (DWT)

Fig. T- 2.2.5 Relationship between Deadweight Tonnage and Berthing Velocity

Fig. T- 2.2.6 Frequency Distribution of Berthing Velocity 10)

(5) Figure T- 2.2.6 shows a berthing velocity frequency distribution obtained from actual measurement records at offshore terminals used by large oil tankers of around 200,000 DWT. It can be seen that the highest measured berthing velocity was 13 cm/s. If the data are assumed to follow a Weibull distribution, then the probability of the berthing velocity below the value 13 cm/s would be 99.6%. The mean µ is 4.41 cm/s and the standard deviation s is 2.08 cm/s. Application of the Weibull distribution yields the probability density function f ( V ) as expressed in equation (2.2.3): V f ( V ) = ------- exp ( V 1.25 ) 0.8 where V： berthing velocity (cm/s)

(2.2.3)

From this equation, the probability of the berthing velocity exceeding 14.5 cm/s becomes 1/1000. The offshore terminals where the berthing velocity measurements were taken had a design berthing velocity of either 15 cm/s or 20 cm/s 7). (6) Small vessels such as small cargo ships and fishing boats come to berths by controlling their positions under their own power without assistance of tugboats. Consequently, the berthing velocity is generally higher than that for larger vessels, and in some cases it can even exceed 30 cm/s. For small vessels in particular, it is necessary to carefully determine the berthing velocity based on actually measured values etc. (7) In cases where cautious berthing methods such as those described in (1) are not used, or in the case of berthing of small or medium-sized vessels under influence of currents, it is necessary to determine the berthing velocity based on actual measurement data etc., considering the ship drift velocity by currents. (8) When designing mooring facilities that may be used by fishing boats, it is recommended to carry out design works based on the design standards for fishing port facilities and actual states of usage. -19-


[3] Eccentricity Factor (Notification Article 22, Clause 2) The eccentricity factor shall be calculated by the following: 1 C e = -------------------2l 1 + æ --ö è rø

(2.2.4)

where l and r represent the following: l： distance from the point where the vessel touches the mooring facilities to the center of gravity of the vessel as measured along the face line of the mooring facilities (m) r： radius of gyration around the vertical axis passing through the center of gravity of the vessel (m) [Technical Notes] (1) When a vessel is in the middle of berthing operation, it is not aligned perfectly along the face line of the berth. This means that after it comes into contact with the mooring facilities (fenders), it starts yawing and rolling. This results in some of the vessel’s kinetic energy being used up. The amount of energy used up by rolling is small compared with that by yawing and can be ignored. Equation (2.2.4) thus only considers the amount of energy used up by yawing. (2) The radious of gyration r relative to Lpp is a function of the block coefficient C b of the vessel and can be obtained from Fig. T- 2.2.7 8). Alternatively, one may use the linear approximation shown in equation (2.2.5) . (2.2.5) r = ( 0.19C b + 0.11 )L pp where r： radius of gyration; this is related to the moment of inertia I z around the vertical axis of the vessel by the relationship Iz = M s r 2 L pp： length between perpendiculars (m) C b： block coefficient; C b = Ñ /( L pp Bd) ( Ñ : Volume of water displaced by the vessel (m3), B: moulded breadth (m), d: draft (m)) (3) As sketched in Fig. T- 2.2.8, when a vessel comes into contact with the fenders F1 and F2 with the point of the vessel closest to the quaywall being the point P, the distance l from the point of contact to the center of gravity of the vessel as measured parallel to the mooring facilities is given by equation (2.2.6) or (2.2.7); l is taken to be L 1 when k ＜ 0.5 and L 2 when k ＞ 0.5. When k = 0.5, l is taken as whichever of L 1 or L 2 that gives the higher value of C e in equation (2.2.4).

F1

keLpp cos θ

F2

B

αLpp

Lpp

Length between perpendiculars (Lpp)

Radius of gyration in the longitudinal direction (r)

A

eLpp cos θ A

P

B

G

θ Q Block coefficient Cb

Fig. T- 2.2.7 Relationship between the Radius of Gyration around the Vertical Axis and the Block Coefficient (Myers, 1969) 7)

Fig. T- 2.2.8 Vessel Berthing

L 2 = 0.5 a + e ( 1 k ) L pp cos q

(2.2.6)

L 1 = ( 0.5a ek )L pp cos q

(2.2.7)

-20-


where L 1： distance from the point of contact to the center of gravity of the vessel as measured parallel to the mooring facilities when the vessel makes contact with fender F1 L 2： distance from the point of contact to the center of gravity of the vessel as measured parallel to the mooring facilities when the vessel makes contact with fender F2 q： berthing angle (the value of q is set as a design condition; it is usually set somewhere in the range 0 ~ 10º) e： ratio of the distance between the fenders, as measured in the longitudinal direction of the vessel, to the length between perpendiculars a： ratio of the length of the parallel side of the vessel at the height of the point of contact with the fender to the length between perpendiculars; this varies according to factors like the type of vessel, and the block coefficient etc., but is generally in the range 1/3 ~ 1/2. k： parameter that represents the relative location of the point where the vessel comes closest to the mooring facilities between the fenders F1 and F2 ; k varies between 0 and 1, but it is generally taken at k = 0.5.

[4] Virtual Mass Factor (Notification Article 22, Clause 3)

Ñ C b = --------------L pp Bd

64748

It shall be standard to calculate the virtual mass factor using the following equations: d p C m = 1 + --------- ´ --2C b B

(2.2.8)

where Cb,Ñ, Lpp, B, and d represent the following: C b： block coefficient Ñ： volume of water displaced by the vessel (m3) L pp： length between perpendiculars (m) B： moulded breadth (m) d： full load draft (m) [Technical Notes] (1) When a vessel berths, the vessel (which has mass M s ) and the water mass surrounding the vessel (which has mass M w ) both decelerate. Accordingly, the inertial force corresponding to the water mass is added to that of the vessel itself. The virtual coefficient is thus defined as in equation (2.2.9). Ms + M w C m = --------------------Ms where C m： virtual mass factor M s： mass of vessel (t) M w： mass of the water surrounding the vessel (added mass) (t)

(2.2.9)

Ueda 8) proposed equation (2.2.8) based on the results of model experiments and field observations. The second term in equation (2.2.8) corresponds to M w ¤ M s in equation (2.2.9). (2) As a general rule, the actual values of the target vessel are used for the length between perpendiculars ( L pp ), the moulded breadth (B), and the full load draft (d). But when one of the standard ship sizes is used, one may use the principal dimensions given in 2.1 Dimensions of the Target Vessel. Regression equations have been proposed for the relationships between the deadweight tonnage, the moulded breadth and the full load draft 1). It is also possible to use equations (2.2.10), which give the relationship between the deadweight tonnage (DWT) or the gross tonnage (GT) and the length between perpendiculars for different types of vessel 1).

-21-

64444744448

Cargo ships (less than 10,000 DWT)： log (Lpp) ＝ 0.867 ＋ 0.310 log (DWT) Cargo ships (10,000 DWT or more)： log (Lpp) ＝ 0.964 ＋ 0.285 log (DWT) Container ships： log (Lpp) ＝ 0.516 ＋ 0.401 log (DWT) Ferries (long distance, 13,000 GT or less)： log (Lpp) ＝ log (94.6 ＋ 0.00596GT) Ferries (short-to-medium distance, 6,000 t or less)： log (Lpp) ＝ 0.613 ＋ 0.401 log (GT) Roll-on/roll-off vessels： log (Lpp) ＝ 0.840 ＋ 0.349 log (DWT) Passenger ships (Japanese)： log (Lpp) ＝ 0.679 ＋ 0.359 log (GT) Passenger ships (foreign)： log (Lpp) ＝ 0.787 ＋ 0.330 log (GT) Car carriers： log (Lpp) ＝ 1.046 ＋ 0.280 log (GT) Oil tankers： log (Lpp) ＝ 0.793 ＋ 0.322 log (DWT)

(2.2.10)


(3) The volume of water displaced by the vessel Ñ is determined by dividing the displacement tonnage DT by the density of seawater (1.03 t/m3)

2.2.3 Moored Vessels [1] Motions of Moored Vessel (Notification Article 23) As a general rule, the external forces generated by the motions of a moored vessel shall be calculated by carrying out a numerical simulation of vessel motions, with the wave force acting on the vessel, the wind force, the current force due to water currents, etc. being set appropriately. [Commentary] (1) Vessels moored at mooring facilities situated in the open sea or near to harbor entrances, or at mooring facilities inside harbors for which long-period waves are expected to come in, along with vessels moored during stormy weather, are liable to be moved under the influence of loads due to waves, winds, currents, etc. In some cases, the kinetic energy due to such motions can exceed the berthing energy. In such cases, it is thus advisable to give full consideration to the tractive forces and impact forces caused by the motions of vessels when designing bollards and fenders 10). (2) As a general rule, the external forces generated by the motions of a vessel should be obtained by carrying out a numerical simulation of vessel motions, based on the factors such as the wave force acting on the vessel, the wind force, the current force due to currents, and the load-deflection characteristics of the mooring system. [Technical Notes] (1) As a general rule, the motions of a moored vessel should be analyzed through numerical simulation, with consideration given to the random variations of the loads and the nonlinearity of the load-deflection characteristics of the mooring system. However, when such a numerical simulation of vessel motions is not possible, or when the vessel is moored at a system that is considered to be more-or-less symmetrical, one may obtain the displacement of and loads on the mooring system either by using frequency response analysis for regular waves or by referring to results of an motion analysis on a floating body moored at a system that has load-deflection characteristics of bilinear nature 11). (2) The total wave force acting on the hull of a vessel is analyzed by dividing it into the wave exciting force due to incident waves and the radiation force that is generated as the vessel moves. The wave exciting force due to incident waves is the wave force calculated for the case that motions of the vessel are restrained. The radiation force is the wave force exerted on the hull when the vessel undergoes a motion of unit amplitude for each mode of motions. The radiation force can be expressed as the summation of a term that is proportional to the acceleration of the vessel and a term that is proportional to the velocity. Specifically, the former can be expressed as an added mass divided by acceleration, while the latter can be expressed as a damping coefficient divided by velocity 12). In addition, a nonlinear fluid dynamic force that is proportional to the square of the wave height acts on the vessel (see 8.2 External Forces Acting on Floating Body). (3) For vessels that have a block coefficient of 0.7 ~ 0.8 such as large oil tankers, the ship hull can be represented with an elliptical cylinder, allowing an approximate evaluation of the wave force 13). (4) For box-shaped vessels such as working craft, the wave force can be obtained by taking the vessel to be either a floating body with a rectangular cross section or a floating body of a rectangular prism.

[2] Waves Acting on Vessel The wave force acting on a moored vessel shall be calculated using an appropriate method, considering the type of vessel and the wave parameters. [Commentary] The wave force acting on a moored vessel is calculated using an appropriate method, for example the strip method, the source distribution technique, the boundary element method, or the finite element method; the most common method used for vessels is the strip method. [Technical Notes] (1) Wave Force by the Strip Method 11), 12) (a) Wave force of regular waves acting on the hull The wave force acting on the hull is taken to be the summation of the Froude-Kriloff force and the force by the waves that are reflected by the hull (diffraction force).

-22-


(b) Froude-Kriloff force The Froude-Kriloff force is the force derived by integrating the pressure of progressive waves around the circumference of the hull. In the case of a moored vessel in front of a quaywall, it is taken to be the summation of the force of the incident waves and the force of the reflected waves from the quaywall. (c) Diffraction force The diffraction force acting on a vessel is the force that is generated by the change in the pressure field when incident waves are scattered by the vessel’s hull. As an estimate, this change in the pressure field can be replaced by the radiation force (the wave making resistance when the vessel moves at a certain velocity through a calm fluid) for the case that the hull is moved relative to fluid. It is assumed that the velocity of the vessel in this case is equal to the velocity of the cross section of the hull relative to the water particles in the incident waves. This velocity is referred to as the “equivalent relative velocity”. (d) Total force acting on the hull as a whole The total wave force acting on the hull as a whole can be obtained by integrating the Froude-Kriloff force and the diffraction force acting on a cross section of the hull in the longitudinal direction from x = L pp ¤ 2 to x = L pp ¤ 2 . (2) Waves Forces by Diffraction Theory 13) In the case that the vessel in question is very fat (i.e., it has a block coefficient C b of 0.7 ~ 0.8), there are no reflecting structures such as quaywalls behind the vessel, and the motions of the vessel are considered to be very small, the vessel may be represented with an elliptical cylinder and the wave force may be calculated using an equation based on a diffraction theory 13).

[3] Wind Load Acting on Vessel The wind load acting on a moored vessel shall be determined using an appropriate calculation formula. [Commentary] It is desirable to determine the wind load acting on a moored vessel while considering the temporal fluctuation of the wind velocity and the characteristics of the drag coefficients, which depend on the cross-sectional form of the vessel. [Technical Notes] (1) The wind load acting on a vessel should be determined from equations (2.2.11) ~ (2.2.13), using the drag coefficients C X and C Y in the X and Y directions and the pressure moment coefficient C M about the midship. 1 R X = --- r a U 2 A T C X 2 1 R Y = --- r a U 2 A L C Y 2 1 R M = --- r a U 2 A L L pp C M 2 where C X： drag coefficient in the X direction (from the front of the vessel) C Y： drag coefficient in the Y direction (from the side of the vessel) C M： pressure moment coefficient about the midship R X： X component of the wind force (kN) R Y： Y component of the wind force (kN) R M： moment of the wind load about the midship (kN•m) r a： density of air; r a = 1.23 ´ 10 3 (t/m3) U： wind velocity (m/s) A T： front projected area above the water surface (m2) A L： side projected area above the water surface (m2) L pp： length between perpendiculars (m)

(2.2.11) (2.2.12) (2.2.13)

(2) It is desirable to determine the wind force coefficients C X , C Y , and C M through wind tunnel tests or water tank tests on a target vessel. Since such experiments require time and cost, it is acceptable to use the calculation equations for wind force coefficients 14),15) that are based on wind tunnel tests or water tank tests that have been carried out in the past. (3) The maximum wind velocity (10-minute average wind velocity) should be used as the wind velocity U. (4) For the front projected area above the water surface and the side projected area above the water surface, it is desirable to use the values for the target vessel. For standard vessel sizes, one may refer to regression equations 1). (5) Since the wind velocity varies both in time and in space, the wind velocity should be treated as fluctuating in the -23-

TECHNICAL STANDARDS AND COMMENTARIES FOR PORT AND HARBOUR FACILITIES IN JAPAN analysis of the motions of a moored vessel. Davenport 16) and Hino have proposed the frequency spectra for the time fluctuations of the wind velocity. The frequency spectra proposed by Davenport and Hino are given by equations (2.2.14) and (2.2.15), respectively.

64748

X2 2 f S u ( f ) = 4K r U10 --------------------------(1 + X2 )4 ¤ 3 X = 1200f / U 10

5 ¤ 6

æ U 10 aö z 2m a 1 b = 1.169 ´ 10 3 ç -------------÷ æ ------ö è K r ø è 10ø

64748

2

K r U10 ì f 2ü S u ( f ) = 2.856 --------------- í 1 + æ ---ö ý è bø b î þ

(2.2.14)

(2.2.15)

where S u ( f )： frequency spectrum of wind velocity (m2•s) U 10： average wind velocity at the standard height 10 m (m/s) K r： friction coefficient for the surface defined with the wind velocity at the standard height; over the ocean, it is considered that K r = 0.003 is appropriate. a： exponent when the vertical profile of the wind velocity is expressed by a power law [ U µ ( z ¤ 10 ) a ] z： height above the surface of the ground or ocean (m) m： correction factor relating to the stability of the atmosphere; m is taken to be 2 in the case of a storm.

[4] Current Forces Acting on Vessel The flow pressure force due to tidal currents acting on a vessel shall be determined using an appropriate calculation formula. [Technical Notes] (1) Current Pressure Force Due to Currents Coming onto the Bow of Vessel The current pressure force on the vessel due to currents coming onto the bow of a vessel may be calculated using equation (2.2.16). (2.2.16) R f = 0.0014SV 2 where R f： current pressure force (kN) S： wetted surface area (m2) V： flow velocity (m/s) (2) Current Pressure Force Due to Currents Coming onto the Side of Vessel The current pressure force due to a current coming onto the side of a vessel may be calculated using equation (2.2.17). (2.2.17) R = 0.5r 0 CV 2 B where R： current pressure force (kN) r 0： density of seawater (t/m3) (standard value: r 0 = 1.03 t/m3) C： current pressure coefficient V： flow velocity (m/s) B： side projected area of hull below the waterline (m2) (3) The current pressure force due to tidal currents can in principle be divided into frictional resistance and pressure resistance. It is thought that the resistance to currents coming onto the bow of a vessel is predominantly frictional resistance, whereas the resistance to currents coming onto the side of a vessel is predominantly pressure resistance. However, in practice it is difficult to rigorously separate the two resistances and investigate them individually. Equation (2.2.16) is a simplification of the following Froude equation with r w = 1.03, t = 15ºC and l = 0.14: Rf where R f： rw： g： t： S：

= rw gl { 1 + 0.0043 ( 15 t ) }SV 1.825

(2.2.18)

current pressure force (N) specific gravity of seawater (standard value: rw = 1.03) gravitational acceleration (m/s2) temperature (ºC) wetted surface area (m2) -24-


V： flow velocity (m/s) l： coefficient (l = 0.14741 for a 30m-long vessel and l = 0.13783 for a 250m-long vessel) (4) The current pressure coefficient C in equation (2.2.17) varies according to the relative current direction q; the values obtained from Fig. T- 2.2.9 may be used for reference purposes.

Current pressure coefficient

C

(5) Regarding the wetted surface area S and the side projected area below the waterline B, one may use values obtained from a regression equations 3) that have been derived by statistical analysis.

Water depth draft

d

h = 1.1

1.5

7.0

Relative current direction

q(

)

Fig. T- 2.2.9 Current Pressure Coefficient C

[5] Load-Deflection Characteristics of Mooring System When performing a motion analysis of a moored vessel, the load-deflection characteristics of the mooring system (mooring ropes, fenders, etc.) shall be modeled appropriately. [Technical Notes] The load-deflection characteristics of a mooring system (mooring ropes, fenders, etc.) is generally nonlinear. Moreover, with regard to the load-deflection characteristics of a fender, they may show hysteresis, and so it is desirable to model these characteristics appropriately before carrying out the motion analysis of a moored vessel.

2.2.4 Tractive Force Acting on Mooring Post and Bollard (Notification Article 79) (1) It shall be standard to take the values listed in Table 2.2.1 as the tractive forces of vessels acting on mooring posts and bollards. (2) In the case of a mooring post, it shall be standard to assume that the tractive force stipulated in (1) acts horizontally and a tractive force equal to one half of this acts upwards simultaneously. (3) In the case of a bollard, it shall be standard to assume that the tractive force stipulated in (1) acts in all directions. Table 2.2.1 Tractive Forces of Vessels (Notification Article 79, Appended Table 12) Gross tonnage (GT) of vessel (tons)

Tractive force acting on a mooring post (kN)

Tractive force acting on a bollard (kN)

200 ＜ GT ≦ 500

150

150

500 ＜ GT ≦ 1,000

250

250

1,000 ＜ GT ≦ 2,000

350

250

2,000 ＜ GT ≦ 3,000

350

350

3,000 ＜ GT ≦ 5,000

500

350

5,000 ＜ GT ≦ 10,000

700

500

10,000 ＜ GT ≦ 20,000

1,000

700

20,000 ＜ GT ≦ 50,000

1,500

1,000

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Gross tonnage (GT) of vessel (tons) 50,000 ＜ GT ≦ 100,000

Tractive force acting on a mooring post (kN)

Tractive force acting on a bollard (kN)

2,000

1,000

[Commentary] (1) “Mooring posts” are installed away from the waterline, either on or near to the mooring facilities, close to the both ends of a berth so that they may be used for mooring a vessel in a storm. “Bollards”, on the other hand, are installed close to the waterline of the mooring facilities so that they may be used for mooring, berthing, or unberthing a vessel in normal conditions. (2) Regarding the layout and names of mooring ropes to moor a vessel, see Part Ⅷ , 2.1 Length and Water Depth of Berths. (3) Regarding the layout and structure of mooring posts and bollards, see Part Ⅷ , 19.3 Mooring Posts, Bollards, and Mooring Rings. [Technical Notes] (1) It is desirable to calculate the tractive force acting on a mooring post and a bollard based on the breaking strength of the mooring ropes possessed by a vessel arriving at the berth, the meteorological and oceanographic conditions at the place where the mooring facilities are installed, and the dimensions of vessels, and if necessary also considering the force due to a berthing vessel, the wind pressure on a moored vessel, and the force due to motions of a vessel 9), 11). Alternatively, it is also possible to determine the tractive force acting on a mooring post and a bollard in accordance with (2) ~ (6) below. (2) In the case that the gross tonnage of a vessel exceeds 5,000 tons and there is no risk of more than one mooring rope being attached to a bollard that is used for spring lines at the middle of mooring facilities for which the vessel’s berth is fixed, the tractive force acting on a bollard may be taken as one half of the value listed in Table 2.2.1. (3) The tractive force due to a vessel whose gross tonnage is no more than 200 tons or greater than 100,000 tons (i.e., a vessel that is not covered in Table 2.2.1) should be calculated by considering the meteorological and oceanographic conditions, the structure of the mooring facilities, past measurement data on tractive force, etc. The tractive force on mooring facilities at which vessels are moored even in rough weather or mooring facilities that are installed in waters with severe meteorological / oceanographic conditions should also be calculated by considering these conditions. (4) The tractive force acting on a mooring post has been determined based on the wind pressure acting on a vessel in such a way that a lightly loaded vessel should be able to moor safely even when the wind velocity is 25 ~ 30 m/s, with the assumption that the mooring posts are installed at the place away from the wharf waterline by the amount of vessel’s width and that the breast lines are pulled in a direction 45º to the vessel’s longitudinal axis 17),18). The tractive force so obtained corresponds to the breaking strength of one to two mooring ropes, where the breaking strength of a mooring rope is evaluated according to the “Steel Ship Regulations” by the Nippon Kaiji Kyokai. For a small vessel of gross tonnage up to 1,000 tons, the mooring posts can withstand the tractive force under the wind velocity of up to 35 m/s. The tractive force acting on a bollard has been determined based on the wind pressure acting on a vessel in such a way that even a lightly loaded vessel should be able to moor using only bollards in a wind of velocity up to 15 m/s, with the assumption that the ropes at the bow and stern are pulled in a direction at least 25º to the vessel’s axis. The tractive force so obtained corresponds to the breaking strength of one mooring rope for a vessel of gross tonnage up to 5,000 tons and two mooring ropes for a vessel of gross tonnage over 5,000 tons, where the breaking strength of a mooring rope is evaluated according to the “Steel Ship Regulations” by the Nippon Kaiji Kyokai. The tractive force for a bollard that is used for spring lines and is installed at the middle of a berth, for which the vessel’s berthing position is fixed, corresponds to the breaking strength of one mooring rope, where the breaking strength of a mooring rope is evaluated according to the “Steel Ship Regulations” by the Nippon Kaiji Kyokai. Note however that, although there are stipulations concerning synthetic fiber ropes in the “Steel Ship Regulations” by the Nippon Kaiji Kyokai with regard to nylon ropes and type B vinylon ropes (both of which are types of synthetic fiber rope), the required safety factor has been set large owing to the factors such that there is little data on the past usage of such ropes and their abrasion resistance is low, and so both the required rope diameter and the breaking strength are large. Accordingly, in the case of berths for which the mooring vessels use only nylon ropes or type B vinylon ropes, it is not possible to apply the stipulations in (2) above. In the above-mentioned tractive force calculations, in addition to the wind pressure, it has been assumed that there are tidal currents of 2 kt in the longitudinal direction and 0.6 kt in the transverse direction. (5) When determining the tractive force from a small vessel of gross tonnage no more than 200 tons, it is desirable to consider the type of vessel, the berthing situation, the structure of the mooring facilities, etc. During actual

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design of mooring posts and bollards for vessels of gross tonnage no more than 200 tons, it is standard to take the tractive force acting on a mooring posts to be 150 kN and the tractive force acting on a bollard to be 50 kN. (6) When calculating the tractive force in the case of vessels such as ferries, container ships, or passenger ships, caution should be exercised in using Table 2.2.1, because the wind pressure-receiving areas of such vessels are large. [References] 1) Yasuhiro AKAKURA, Hironao TAKAHASHI, Takashi NAKAMOTO: “Statistical analysis of ship dimensions for the size of design ship”, Tech. Note of PHRI, No. 910, 1998 (in Japanese). 2) Yasuhiro AKAKURA and Hironao TAKAHASHI: “Ship dimensions of design ship under given confidence limits”, Technical Note of P.H.R.I., September 1998. 3) PIANC: “Report of the International Commission for Improving the Design of Fender Systems”, Supplement to Bulletine No. 45, 1984. 4) Baker, A. L. L.: “The impact of ships when berthing”, Proc. Int’l Navig. Congr. (PIANC), Rome, Sect II, Quest. 2, 1953, pp. 111-142. 5) Masahito MIZOGUCHI, Tanekiyo NAKAYAMA: “Studies on the berthing velocity, energy of the ships”, Tech. Note of PHRI, No. 170, 1973 (in Japanese). 6) Hirokane OTANI, Shigeru UEDA, Tatsuru ICHIKAWA, Kensei SUGIHARA: “A study on the berthing impact of the big tanker”, Tech. Note of PHRI, No. 176, 1974 (in Japanese). 7) Shigeru UEDA: “Study on berthing impact force of very large crude oil carriers”, Rept. of PHRI, Vol. 20, No. 2, 1981, pp. 169-209 (in Japanese). 8) Myers, J.: “Handbook of Ocean and Underwater Engineering”, McGraw-Hill, New York, 1969. 9) Shigeru UEDA, Eijiro OOI: “On the design of fending systems for mooring facilities in a port”, Tech. Note of PHRI, No. 596, 1987 (in Japanese). 10) Shigeru UEDA, Satoru SHIRAISHI: “On the design of fenders based on the ship oscillations moored to quaywalls”, Tech. Note of PHRI, No. 729, 1992 (in Japanese). 11) Shigeru UEDA: “Analytical method of motions moored to quaywalls and the applications”, Tech. Note of PHRI, No. 504, 1984 (in Japanese). 12) Shigeru UEDA, Satoru SHIRAISI: “Method and its evaluation for computation of moored ship’s motions”, Rept. of PHRI, Vol. 22, No. 4, 1983 pp. 181-218 (in Japanese). 13) Yoshimi GODA, Tomotsuka TAKAYAMA, Tadashi SASADA: “Theoretical and experimental investigation of wave forces on a fixed vessel approximated with an elliptic cylinder”, Rept of PHRI, Vol. 12, No. 4, 1994, pp. 23-74 (in Japanese). 14) R. M. Isherwood: “Wind resistance of merchant ships”, Bulliten of the Royal Inst. Naval Architects, 1972, pp. 327-338. 15) Shigeru UEDA, Satoru SHIRAISHI, Kouhei ASANO, Hiroyuki OSHIMA: “Proposal of equation of wind force coefficient and evaluation of the effect to motions of moored ships”, Tech. Note of PHRI, No. 760, 1993 (in Japanese). 16) Davenport, A. G.: “Gust loading factors”, Proc. of ASCE, ST3, 1967, pp. 11-34. 17) Hirofumi INAGAKI, Koichi YAMAGUCHI, Takeo KATAYAMA: “Standardization of mooring posts and bollards for wharf”, Tech. Note of PHRI, No. 102, 1970 (in Japanese). 18) Iaso FUKUDA, Tadahiko YAGYU: “Tractive force on mooring posts and bollards”, Tech. Note of PHRI, No. 427, 1982 (in Japanese).

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Chapter 3 Wind and Wind Pressure 3.1 General When designing port and harbor facilities, meteorological factors such as winds, air pressure, fog, rainfall, snow depth, and air temperature should be considered. [Commentary] The effects that meteorological factors exert on the design of port and harbor facilities are as follows: (1) Air pressure and its distribution are the factors that govern the generations of winds and storm surge. (2) Wind is a factor that governs the generations of waves and storm surge, it exerts external forces on port and harbor facilities and moored vessels in the form of wind pressure, and it can disrupt port and harbor works such as cargo handling. (3) Rainfall is a factor that determines the required capacity of drainage facilities in ports and harbors, and rain can also disrupt port and harbor works such as cargo handling. (4) Fog is a factor that is an impediment to the navigation of vessels when they are entering or leaving a harbor, and also decreases the productivity of port and harbor facilities. (5) In some cases, snow load is considered as a static load acting on port and harbor facilities. (6) Air temperature affects the stress distribution within structures of port and harbor facilities and may lead to the occurrence of thermal stress in these structures. [Technical Notes] (1) In calculations concerning the generation of storm surge or waves due to a typhoon, it is common to assume that the air pressure distribution follows either Fujita’s equation (3.1.1) or Myers’ equation (3.1.2); the constants in the chosen equation are determined based on actual air pressure measurements in the region of typhoons. Dp p = p ¥ -------------------------------- (Fujita’ formula) 1 + (r ¤ r0 )2

(3.1.1)

r0 (3.1.2) p = p c + Dp exp æ ----ö (Myers’ formula) è rø where p： air pressure at a distance r from the center of typhoon (hPa) r： distance from the center of typhoon (km) p c： air pressure at the center of typhoon (hPa) r 0： estimated distance from the center of typhoon to the point where the wind velocity is maximum (km) Dp： air pressure drop at the center of typhoon (hPa); Dp = p ¥ p c p ¥： air pressure at r = ¥ (hPa); p ¥ = p c + Dp The size of a typhoon varies with time, and so r 0 and Dp must be determined as the functions of time. (2) With regard to wind, see 3.2 Wind. (3) Rain is generally divided into the rain of thunderstorms that have heavy rainfall in a short period of time and the rain that continues for a prolonged period of time (rain by a typhoon is a representative example of the latter). When designing drainage facilities, it is necessary to determine the intensity of rainfall both for the case where the amount of runoff increases very rapidly and for the case where the runoff continues for a prolonged period. In the case of sewage planning whereby the intensity of rainfall during a thunderstorm is a problem, Sherman’s formula or Talbot’s formula is used. a (Sherman’s formula) R = ---nt a (Talbot’s formula) R = ----------t+b where R： intensity of rainfall (mm/h) t： duration of rainfall (min) a, b, n: constants

(3.1.3) (3.1.4)

(4) With regard to snow load acting upon port and harbor facilities, see 15.3.4 Snow Load.

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3.2 Wind (Notification Article 3, Clause 1) It shall be standard to set the wind characteristics for wave estimations and the wind characteristics as the cause of an external force on port and harbor facilities as stipulated in the following: (1) When calculating the wind velocity and wind direction used in estimations of waves and storm surges, either the actual wind measurements or the calculated values for gradient winds are to be used, with all necessary corrections having been made for the heights of measurements, etc. (2) The velocity of the wind acting on port and harbor facilities shall be set based on statistical data for an appropriate period in line with the characteristics of the facilities and structures. [Technical Notes] (1) Gradient Winds (a) The velocity of the gradient wind can be expressed as a function of pressure gradient, radius of curvature of isobars, latitude, and air density as in equation (3.2.1). æ ¶ p ¤ ¶r ö -÷ Vg = rw sin f ç 1 + 1 + ---------------------------r a rw 2 sin 2 fø è

(3.2.1)

where Vg ： velocity of gradient wind (cm/s); in the case of an anticyclone, equation (3.2.1) gives a negative value and so the absolute value should be taken. ¶p -----： pressure gradient (taken to be positive for a cyclone, negative for an anticyclone) (g/cm2/s2) ¶r r： radius of curvature of isobars (cm) w： angular velocity of Earth's rotation ( s 1 ); w = 7.29 ´ 10 5 ¤ s f： latitude (º) ra： density of air (g/cm3) Before performing the calculation, measurement units should first be converted into the CGS units listed above. Note that 1º of latitude corresponds to a distance of approximately 1.11 × 10 7 cm, and an air pressure of 1.0 hPa is 10 3 g/cm/s2. (b) A gradient wind for which the isobars are straight lines (i.e., their radius of curvature in equation (3.2.1) is infinite) is called the geostrophic wind. In this case, the wind velocity is V = ( ¶ p ¤ ¶r ) ¤ ( 2r a rw sin f ) . (2) The actual sea surface wind velocity is generally lower than the value obtained from the gradient wind equation. Moreover, although the direction of a gradient wind is parallel to the isobars in theory, the sea surface wind blows at a certain angle a to the isobars as sketched in Fig. T- 3.2.2. In the northern hemisphere, the wind around a cyclone blows in a counterclockwise direction and inwards, whereas the wind around an anticyclone blows in a clockwise direction and outwards. It is known that the relationship between the velocity of gradient winds and that of the actual sea surface wind varies with the latitude. The relationship under the average conditions is summarized in Table T- 3.2.1. However, this is no more than a guideline; when estimating sea surface winds, it is necessary to make appropriate corrections by comparing estimations with actual measurements taken along the coast and values that have been reported by vessels out at sea (the latter are written on weather charts). Table T- 3.2.1 Relationship between Sea Surface Wind Speed and Gradient Wind Speed

Low

High

Latitude

10º

20º

30º

40º

50º

Angle a

24º

20º

18º

17º

15º

Velocity ratio V s ¤ V g

0.51

0.60

0.64

0.67

0.70

Fig. T- 3.2.2 Wind Direction for a Cyclone (Low) and an Anticyclone (High)

(3) When selecting the design wind velocity for the wind that acts directly on port and harbor facilities and moored vessels, one should estimate the extreme distribution of the wind velocity based on actual measurement data taken over a long period (at least 30 years as a general rule) and then use the wind velocity corresponding to the required return period. It is standard to take the wind parameters to be the direction and velocity, with the wind direction being represented using the sixteen-points bearing system and the wind velocity by the mean wind velocity over 10 minutes. -29-


In the Meteorological Agency’s Technical Observation Notes No. 34, the expected wind velocities with the return periods of 5, 10, 20, 50, 100 and 200 years for 141 government meteorological offices have been estimated from the ten-minute mean wind velocity data of about 35 years, under the assumption that wind velocity follows a double exponential distribution. For locations with topographical conditions different from that of the nearest among the above-mentioned meteorological offices, one should conduct observations for at least one year and then conduct a comparative investigation on topographical effects in order to make it possible to use the aforementioned estimation results. (4) Regarding the wind velocity used in estimating storm surges and waves, it is standard to use the value at a height of 10 m above sea level. The wind velocities obtained at government meteorological offices are the values for a height of approximately 10 m above the ground level. Accordingly, when attempting to use such observed values to estimate sea surface winds, in the case that the elevations of the structural members are considerably different from 10 m, it is necessary to correct the wind velocity with respect to the height. The vertical profile of the wind velocity is generally represented with a power law, and so in current design calculations for all kinds of structures, a power law is simply used: i.e., h n U h = U 0 æè -----öø h0

(3.2.2)

where U h： wind velocity at height h (m/s) U 0： wind velocity at height h 0 (m/s) The value of the exponent varies with the situation with regard to the roughness near to the surface of the ground and the stability of the atmosphere. In structural calculations on land, a value of n = 1/10 ~ 1/4 is used, and it is common to use a value of n ≧ 1/7 out to sea. Statistical data on wind velocity usually consider the ten-minute mean wind velocity. However, for some structures the mean wind velocity over a shorter time period or the maximum instantaneous wind velocity may be used, in which case it is necessary to gain an understanding of the relationship between the mean wind velocity over a certain time period and the maximum wind velocity, and also the characteristics of the gust factor.

3.3 Wind Pressure (Notification Article 3, Clause 2) The wind pressure shall be set appropriately, giving due consideration to the situation with regard to the structural types of the facilities and their locations. [Technical Notes] (1) When calculating the wind pressure acting on a moored vessel, one should refer to 2.2.3 [3] Wind Load Acting on a Vessel. (2) In the case that there are no statutory stipulations concerning the wind pressure acting on a structure, the wind pressure may be calculated using equation (3.3.1). p = cq where p： wind pressure (N/m2) q： velocity pressure (N/m2) c： wind pressure coefficient

(3.3.1)

Equation (3.3.1) expresses the wind pressure, i.e., the force due to the wind per unit area subjected to the wind force. The total force due to the wind acting on a member or structure is thus the wind pressure as given by equation (3.3.1) multiplied by the area of that member or structure affected by the wind in a plane perpendicular to the direction in which the wind acts. The velocity pressure q is defined as in equation (3.3.2). 1 (3.3.2) q = --- r a U 2 2 where q： velocity pressure (N/m2) r a： density of air (kg/m3) r a = 1.23 kg/m3 U： design wind velocity (m/s) The design wind velocity should be taken at 1.2 to 1.5 times the standard wind velocity (ten-minute mean wind velocity at a height of 10 m). This is because the maximum instantaneous wind velocity is about 1.2 to 1.5 times the ten-minute mean wind velocity. The wind pressure coefficient varies depending on the conditions such as the shape of the member or structure, the wind direction, and the Reynolds number. With the exception of cases where it is determined by means of the wind tunnel experiments, it may be set by referring to the Article 87 of the “Enforcement Order -30-


of the Building Standard Law” (Government Ordinance No. 338, 1950) or the “Crane Structure Standards” (Ministry of Labor Notification). With regard to wind direction, it is generally required to consider the wind direction that is most unfavorable to the structure, with the exception of cases where it has been verified that there exists an overwhelmingly prevailing direction of winds.

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Chapter 4 Waves 4.1 General 4.1.1 Procedure for Determining the Waves Used in Design (Notification Article 4, Clause 1) The waves used in the investigation of the stability of protective harbor facilities and other port and harbor facilities, as well as the examination of the degree of calmness of navigation channels and basins shall be set using wave data obtained from either actual wave measurements or wave hindcasting. Wave characteristics shall be obtained by carrying out necessary statistical processing and by analyzing wave transformations owing to sea bottom topography and others. It shall be standard to carry out the wave hindcasting using a method that is based on an appropriate equation for representing the relationship between the wind velocity and the wave spectrum or the significant wave parameters. [Commentary] The size and structural form of facilities are determined by the factors such as the height and period of the waves that act on them. The setting of the wave conditions to be used in design should thus be carried out carefully. The setting of wave conditions should be carried out separately for “ordinary waves” (i.e., waves that occur in ordinary conditions: these are required when estimating the harbor calmness or the net working rate of cargo handling) and “storm waves” (i.e., waves that occur in storm conditions: these are required when estimating the wave force acting on structures). The waves that are obtained by statistically processing data based on either actual measurements or hindcasting are generally deepwater waves that are unaffected by the sea bottom topography. Deepwater waves propagate towards the coast, and once the waves reach to the water depth about one half the wavelength, they start to experience the effects of topography and transform with the result of wave height change. “Wave transformation” includes refraction, diffraction, reflection, shoaling, and breaking. In order to determine the wave conditions at the place where wave data is needed (for instance the place where a structure of interest is located), it is necessary to give appropriate consideration to such wave transformations by means of numerical calculations or model experiments. In the above-mentioned procedure for setting the wave conditions to be used in design, it is necessary to give sufficient consideration to the irregularity of the waves and to treat the waves as being of random nature as much as possible.

Wave data 1) Actual measurement data 2) Hindcasting values Statistical analysis 1) Ordinary waves

2) Storm waves

Wave occurrence rate of deepwater waves

Design deepwater waves

Wave transformation

Wave transformation

Wave occurrence rate at the place of interest

Parameters of design waves 1) Significant wave 2) Highest wave

1) Harbor calmness 2) Net working rate, number of working days 3) Transport energy of incoming waves 4) Others

1) Wave force acting on structures 2) Amount of waves overtopping at seawall and revetments 3) Others

Fig. T- 4.1.1 Procedure for Setting the Waves to Be Used in Design

[Technical Notes] A sample procedure for setting the wave conditions to be used in design is shown in Fig. T- 4.1.1.

4.1.2 Waves to Be Used in Design Significant waves, highest waves, deepwater waves, equivalent deepwater waves and others shall be used in the design of port and harbor facilities. [Commentary] The waves used in the design of structures are generally “significant waves”. The significant wave is a hypothetical wave that is a statistical indicator of an irregular wave group. Significant waves have the dimensions that are approximately equal to the values from visual wave observations, and so they are used in wave hindcasting. It is also known that the period of a significant wave is approximately equal to the period at the peak of the wave spectrum. Because of such advantages, significant waves have been commonly used to represent wave groups. Nevertheless, depending on the purpose, it may be necessary to convert significant waves into other waves such as highest waves and highest one-tenth waves. -32-


[Technical Notes] (1) Definitions of Wave Parameters (a) Significant wave (significant wave height H1/3 and significant wave period T1/3) The waves in a wave group are rearranged in the order of their heights and the highest one-third are selected; the significant wave is then the hypothetical wave whose height and period are the mean height and period of the selected waves. (b) Highest wave (highest wave height Hmax and highest wave period Tmax) The highest wave in a wave group. (c) Highest one-tenth wave (H1/10, T1/10) The wave whose height and period are equal to the mean height and period of the highest one-tenth of the waves in a wave group. (d) Mean wave (mean wave height H , mean period T ) The wave whose height and period are equal to the mean height and period of all of the waves in a wave group. (e) Deepwater waves (deepwater wave height H0 and deepwater wave period T0) The waves at a place where the water depth is at least one half of the wavelength; the wave parameters are expressed with those of the significant wave at this place. (f) Equivalent deepwater wave height (H0¢) A hypothetical wave height that has been corrected for the effects of planar topographic changes such as refraction and diffraction; it is expressed with the significant wave height. (2) Maximum Wave The largest significant wave within a series of significant wave data that was observed during a certain period (for example, one day, one month, or one year) is called the “maximum wave”. In order to clearly specify the length of the observation period, it is advisable to refer to the maximum wave such as the “maximum significant wave over a period of one day (or one month, one year, etc.)”. Moreover, when one wishes to clearly state that one is referring to the significant wave for the largest wave that occurred during a stormy weather, the term “peak wave” is used (see 4.4 Statistical Processing of Wave Observation and Hindcasted Data). The “maximum wave height” is the maximum value of the significant wave height during a certain period; this is different from the definition of the “highest wave height”. (3) Significance of Equivalent Deepwater Waves The wave height at a certain place in the field is determined as the result of transformations by shoaling and breaking, which depend on the water depth at that place, and those by diffraction and refraction, which depend on the planar topographical conditions at that place. However, in hydraulic model experiments on the transformation or overtopping of waves in a two-dimensional channel or in two-dimensional analysis by wave transformation theory, planar topographical changes are not taken into consideration. When applying the results of a two-dimensional model experiment or a theoretical calculation to the field, it is thus necessary to incorporate in advance the special conditions of the place in question, namely the effects of planar topographical changes (specifically the effects of diffraction and refraction), into the deepwater waves for the place in question, thus adjusting the deepwater waves into a form whereby they correspond to the deepwater incident wave height used for the experiment or theoretical calculation. The deepwater wave height obtained by correcting the effects of diffraction and refraction with their coefficients is called the “equivalent deepwater wave height”. The equivalent deepwater wave height at the place for which design is being carried out is given as follows: H0 ¢ = Kd Kr H0

(4.1.1)

where Kr： refraction coefficient for the place in question (see 4.5.2 Wave Refraction) Kd： diffraction coefficient for the place in question (see 4.5.3 Wave Diffraction)

4.1.3 Properties of Waves [1] Fundamental Properties of Waves Fundamental properties of waves such as the wavelength and velocity may be estimated by means of the small amplitude wave theory. However, the height of breaking waves and the runup height shall be estimated while considering the finite amplitude effects.

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[Technical Notes] (1) Small Amplitude Wave Theory The fundamental properties of waves are expressed as the functions of the wave height, period, and water depth. Various characteristics of shallow water waves as obtained as a first approximation by small amplitude wave theory are listed below. Note that, with regard to the coordinates, the positive x direction is taken in the direction of wave travel and the positive z direction vertically upwards with z = 0 corresponding to the still water level. The water depth h is assumed to be constant and wave characteristics are assumed to be uniform in the transverse direction (y direction). (a) Surface elevation (displacement from still water level) (m) 2p 2p H h ( x ,t ) = ---- sin æ ------x ------tö èL 2 T ø where h： H： L： T：

(4.1.2)

surface elevation (m) wave height (m) wavelength (m) period (s)

(b) Wavelength (m) gT 2 2ph L = --------- tanh ---------2p L

(4.1.3)

where h： water depth (m) g： gravitational acceleration (m/s2) (c) Wave velocity (m/s) gT 2ph C = ------ tanh ---------- = 2p L

gL ------ tanh 2ph ---------2p L

(4.1.4)

(d) Water particle velocity (m/s)

644474448

(z + h) cosh 2p ----------------------2p 2p L pH u = ------- ----------------------------------- sin æ ------x ------tö èL T ø 2ph T sinh ---------L 2p ( z + h ) cosh ----------------------2p 2p L pH w = ------- ----------------------------------- cos æ ------x ------tö è T ø 2ph L T sinh ---------L where u： component of water particle velocity in the x direction (m/s) w： component of water particle velocity in the z direction (m/s)

(4.1.5)

(e) Water particle acceleration (m/s)

644474448

2p ( z + h ) cos h ----------------------2p 2p L 2p 2 H du ----------------------------------- cos æ ------x ------tö ------ = ------------2 è 2ph L T ø dt T sinh ---------L 2p ( z + h ) cos h ----------------------2p 2p L 2p 2 H dw ----------------------------------- sin æ ------x ------tö ------- = ------------2 è T ø L 2ph dt T sinh ---------L where du ------： component of water particle acceleration in the x direction (m/s2) dt dw -------： component of water particle acceleration in the z direction (m/s2) dt

-34-

(4.1.6)

PART II DESIGN CONDITIONS (f) Pressure in water when wave acts (N/m2) 2p ( z + h ) cosh ----------------------2p 2p 1 L p = --- r 0 gH ----------------------------------- sin æ ------x ------tö r 0 gz èL T ø 2ph 2 cosh ---------L

(4.1.7)

where r0： density of water (1.01~1.05 × 103 kg/m3 for seawater) (g) Mean energy of wave per unit area of water surface (J) 1 E = E k + E p = --- r 0 gH 2 8

(4.1.8)

where Ek and Ep are the kinetic and potential energy densities respectively, with Ek = Ep. (h) Mean rate of energy transported in the direction of wave travel per unit time per unit width of wave (N • m/m/s) W = CG E = nCE CG = nC

(4.1.9) (4.1.10)

where CG： group velocity of waves (m/s) 4ph ö æ ---------- ÷ 1ç L n = --- ç 1 + ---------------------÷ 2ç 4ph sinh ----------÷ è L ø

(4.1.11)

(2) Characteristics of Deepwater Waves and Wavelength (a) Deepwater waves Waves in water with the depth greater than one-half the wavelength (h/L > 1/2) are called the deepwater waves. Various characteristics of deepwater waves may be obtained from the equations of small amplitude wave theory by letting h/L ® ∞ . The wavelength L0, wave velocity C0, and group velocity CG for deepwater waves thus become as below. Note that the units of period T are seconds (s). L0 ＝ 1.56T 2(m), C0 ＝ 1.56T (m/s) CG＝ 0.78T (m/s) ＝ 1.52T (kt) ＝ 2.81T (km/h)

(4.1.12)

As expressed in equation (4.1.12), the wavelength, wave velocity, and group velocity for deepwater waves depend only on the period and are independent of the water depth. (b) Wavelength of long waves Waves for which the wavelength is extremely long compared with the water depth (h/L < 1/25) are called the long waves. Various characteristics of long waves may be obtained from the equations of small amplitude wave theory by taking h/L to be extremely small. The wavelength, wave velocity, and group velocity for long waves thus become as follows: L = T gh (m) C = CG =

(4.1.13)

gh (m/s)

(3) Consideration of Finite Amplitude Effects The equations shown in (1) are not always accurate for general shallow water waves having a large height, and so it is sometimes necessary to use equations for finite amplitude waves. When carrying out calculations using finite amplitude wave equations, one should refer to “Handbook of Hydraulic Formulas” published by the Japan Society of Civil Engineers. The amount of the errors in calculations that arise from the use of the small amplitude wave theory varies according to the wave steepness H/L and the ratio of water depth to wavelength h ¤ L . Nevertheless, the error in wave parameters is usually no more than 20 ~ 30% with the exception of the horizontal water particle velocity u. One of the finite amplitude effects of waves appears on the crest elevation hc relative to the wave height; the ratio increases as the wave height increases. The definition of the crest elevation hc is shown at the top of Fig. T4.1.2. This figure was drawn up based on wave profile records from the field. It shows the ratio of the highest crest elevation obtained from each observation record to the highest wave height Hmax in that record as the function of relative wave height H1/3/h. -35-


(4) Types of Finite Amplitude Wave Theory The finite amplitude wave theories include the Stokes wave theory, cnoidal wave theory, and others. In the former, the wave steepness is assumed to be relatively low, and the wave profile is represented as a series of (ηc)max Hmax trigonometric functions. A number of researchers have proposed several approximate series solutions. In this theory, however, convergence of the series becomes extremely poor as the water depth to wavelength ratio decreases. This means that the theory cannot be applied if the water depth to wavelength ratio is too small. On the Standard other hand, the cnoidel wave theory is obtained by a perdeviation Number of turbation expansion method with the water depth to data points wavelength ratio assumed to be extremely small, meanMean ing that it is valid when the water depth to wavelength ratio is small. Errors become large, however, when the water depth to wavelength ratio increases. In addition to these two theories, there are also the hyperbolic wave theory, in which a cnoidal wave is approximated as an H1/3 / h expansion of hyperbolic functions, and the solitary wave Fig. T- 4.1.2 Relationship between Maximum theory, which is the asymptotic case of the cnoidal wave Crest Elevation (hc)max/Hmax and theory when the wavelength approaches to infinity. With Relative Wave Height H1/3/h the exception of solitary wave theory, the equations in all of these finite amplitude wave theories are complicated, meaning that calculations are not easy. In particular, with the cnoidal wave theory, the equations contain elliptic integrals, making them very inconvenient to handle. If Dean’s stream function method 1), 2) is adopted, then the wave profile and water particle velocity can be obtained with good accuracy right up to the point where the wave breaks. (5) Application of Finite Amplitude Wave Theories to Structural Designs Nonlinear theories, which include finite amplitude wave theories, are applied to a wide variety of coastal engineering fields. However, there are still a large number of unknowns, and so, in the case of design at present, they are only applied to a limited number of fields such as those discussed below. (a) Maximum horizontal water particle velocity umax at each elevation below the wave crest This information is vital in the estimation of the wave force on a vertical structural member. The equations from the Stokes wave theory are used when the water depth to wavelength ratio is large, and the equations from solitary wave theory are used when the water depth to wavelength ratio is small. An approximate calculation may be carried out using the following empirical equation 3): H 1 ¤ 2 z + h 3 i cos h [ ( 2p ( z + h ) ) ¤ L ] pH u max ( z ) = ------- 1 + a æ ----ö æ -----------ö -----------------------------------------------------è hø è h ø T sinh [ ( 2ph ) ¤ L ]

(4.1.14)

where the coefficient a is given as listed in Table T- 4.1.2. Table T- 4.1.2 Coefficient a for Calculation of Maximum Horizontal Water Particle Velocity h/L

a

h/L

a

0.03 0.05 0.07 0.10 0.14

1.50 1.50 1.43 1.25 0.97

0.2 0.3 0.5 0.7

0.68 0.49 0.25 0.27

(b) Wave shoaling Wave shoaling, which occurs as the water depth decreases, may be calculated using a long wave theory that includes nonlinear terms. Alternatively, the cnoidal wave theory or hyperbolic wave theory may be applied to this phenomenon (see 4.5.5 Wave Shoaling). (c) Rise and drop of the mean water level The mean water level gradually drops as waves approach the breaking point and then rises within the breaker zone toward the shoreline, as can be calculated from the theory of nonlinear interference between waves and currents. This mean water level change is taken into account for the calculation of the wave height change due to random wave breaking (see 4.5.6 Wave Breaking).

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(d) Air gap of offshore structures When determining the amount of air gap of offshore structures above the still water level, it is advisable to consider the relative increase in the wave crest elevation due to the finite amplitude effect such as exhibited in Fig. T-4.1.12.

[2] Statistical Properties of Waves In the design of port and harbor facilities, it shall be standard to consider the statistical properties of the waves with regard to wave heights and periods and to use the Rayleigh distribution for the wave heights of an irregular deepwater wave group. [Commentary] The assumption behind the theory of Rayleigh distribution is a precondition that the wave energy is concentrated in an extremely narrow band around a certain frequency. Problems thus remain with regard to its applicability to ocean waves for which the frequency band is broad. Nevertheless, it has been pointed out that, so long as the waves are defined by the zero-upcrossing method, the Rayleigh distribution can be applied to ocean waves as an acceptable approximation. [Technical Notes] (1) Expression of Rayleigh Distribution The Rayleigh distribution is given by the following equation: ì p H 2ü pH p ( H ¤ H ) = --- ----- exp í --- æ -----ö ý è ø 2H î 4 H þ

(4.1.15)

where p(H/H)： probability density function of wave heights H ： mean wave height (m) According to the Rayleigh distribution, the highest one-tenth wave height H1/10, the significant wave height H 1 ¤ 3 , and the mean wave height H are related to one another by the following equations:

678

H 1 ¤ 10 = 1.27H 1 ¤ 3 H 1 ¤ 3 = 1.60H

(4.1.16)

On average, these relationships agree well with the results of wave observations in situ. The highest wave height Hmax is difficult to determine precisely as will be discussed in (2) below, but in general it may be fixed as in the following relationship: H max = ( 1.6 ～ 2.0 )H 1 ¤ 3

(4.1.17)

The periods are related as follows: T max ≒ T 1 ¤ 3 = ( 1.1 ～ 1.3 )T

(4.1.18)

It should be noted however that as waves approach the coast, waves with the heights greater than the breaking limit begin to break and that their heights are reduced. Thus it is not possible to use the Rayleigh distribution for the wave heights within the breaker zone. (2) Occurrence Probability of the Highest Wave Height The highest wave height Hmax is a statistical quantity that cannot be determined precisely; it is only possible to give its occurrence probability. If the wave height is assumed to follow a Rayleigh distribution, then the expected value Hmax of Hmax , when a large number of samples each composed of N waves are ensembled, is given as follows: 0.5772 (4.1.20) H max = 0.706 æ l n N + ----------------ö H 1 ¤ 3 è 2 l n Nø It should be noted, however, that when Hmax is obtained for each of a large number of samples each containing N waves, there will be a considerable number of cases in which Hmax exceeds Hmax. Thus a simple use of Hmax as the design wave might place structures on a risky side. One can thus envisage the method in which a wave height (Hmax)m with m = 0.05 or 0.1 is used, where (Hmax)m is set such that the probability of the value of Hmax exceeding (Hmax)m is m (i.e., the significance level is m). The value of (Hmax)m for a given significance level m is given by the following equation: N ( H max ) m = 0.706H 1 ¤ 3 l n æ ----------------------------------ö è l n [ 1 ¤ ( 1 m ) ]ø

(4.1.21) -37-


Table T- 4.1.4 lists the values obtained from this equation. Because Hmax is not a definite value but rather a probabilistic variable, the value of Hmax / H1/3 varies greatly with N and m. However, considering the facts that the wave height only approximately follows a Rayleigh distribution and that the wave pressure formula has been derived while containing a certain scatter of experimental data, it is appropriate to use Hmax = (1.6 ~ 2.0) H1/3 by neglecting the very small or large values in the table. Table T- 4.1.4 Relationship between Highest Wave Height Hmax and Significant Wave Height H1/3 Number of waves N

50% significance level (Hmax) 0.5

Mode (Hmax) mode 1.40H1/3 1.52H1/3 1.63H1/3 1.76H1/3 1.86H1/3 1.95H1/3 2.05H1/3 2.12H1/3

50 100 200 500 1,000 2,000 5,000 10,000

Mean (Hmax)

1.46H1/3 1.58H1/3 1.68H1/3 1.81H1/3 1.91H1/3 2.00H1/3 2.10H1/3 2.19H1/3

1.50H1/3 1.61H1/3 1.72H1/3 1.84H1/3 1.94H1/3 2.02H1/3 2.12H1/3 2.19H1/3

10% significance level (Hmax) 0.1 1.76H1/3 1.85H1/3 1.94H1/3 2.06H1/3 2.14H1/3 2.22H1/3 2.31H1/3 2.39H1/3

5% significance level (Hmax) 0.05 1.86H1/3 1.95H1/3 2.03H1/3 2.14H1/3 2.22H1/3 2.30H1/3 3.39H1/3 2.47H1/3

[3] Wave Spectrum In the design of port and harbor facilities, due consideration shall be given to the functional form of the wave spectrum and an appropriate expression shall be used. [Technical Notes] (1) General Form of Wave Spectrum The general form of the wave spectrum is usually represented as in the following equation: S ( f, q ) = S ( f )G ( f, q )

(4.1.22)

where f： frequency q： azimuth from the principal direction of the wave S(f,q)： directional spectrum In the above, S(f) is a function that represents the distribution of the wave energy with respect to frequency; it is called the “frequency spectrum”. G(f,q) is a function that represents the distribution of the wave energy with respect to direction; it is called the “directional spreading function”. The functions expressed in the following equations may be used for S(f) and G(f,q). The frequency spectrum of equation (4.1.23) is called the Bretschneider-Mitsuyasu spectrum, while equation (4.1.24) is called the Mitsuyasu type spreading function. 2

4

4 5

S ( f ) = 0.257H 1 ¤ 3 T1 ¤ 3 f exp [ 1.03 ( T1 ¤ 3 f ) ] q G ( f, q ) = G 0 cos 2S --2 where G0 is a constant of proportionality that satisfies the following normalization condition: i

qmax

òq

G ( f, q ) dq = 1

(4.1.23) (4.1.24)

(4.1.25)

min

f S = S max æ -----ö è f mø

5

:

f ≦ fm

64748

where qmax and qmin are respectively the maximum and minimum angles of deviation from the principal direction. The term S in equation (4.1.24) is a parameter that represents the degree of directional spreading of wave energy. It is given by the following formulas: f 2.5 : f > fm S = S max æ -----ö è f mø (4.1.26)

where fm is the frequency at which the spectrum peak appears. It may be represented in terms of the significant wave period T1/3 as in the following equation: f m = 1 ¤ ( 1.05T 1 ¤ 3 )

(4.1.27)

If the units of H1/3 and T1/3 are meters and seconds respectively, then the units of S(f,q) are m2•s. -38-


S max

(2) Value of Directional Spreading Parameter It is standard to take a value of 10 for the maximum value Smax of the directional spreading parameter in the case of wind waves in deep water. In the case of swell considering the process of wave decay and others, it is appropriate to take a value of 20 or more. Figure T- 4.1.4 shows a graph of approximately estimated values of Smax against wave steepness. Judging by the value of wave steepness, it can be seen that Smax< 20 for wind waves. This graph may be used in order to set an approximate value for Smax. Goda and Suzuki 4) have proposed using as the standard values Smax = 10 for wind waves, Smax = 25 for swell during initial decay, and Smax = 75 for swell that has a long decay distance.

(αp)0

h/L0

Fig. T- 4.1.5 Graph Showing the Change in Smax Due to Refraction Fig. T- 4.1.4 Graph Showing Estimated Values of Smax against Wave Steepness

(3) Change in Smax Due to Refraction The form of the directional spreading function changes as waves undergo the refraction process. When a diffraction calculation on irregular waves is carried out using waves that have been refracted, it is thus very important to consider such changes in the directional spreading function. Figure T- 4.1.5 shows the values of Smax after waves have been refracted at a coastline with straight and parallel depth contour lines. In the figure, (ap)0 is the incident angle of the principal wave direction at the deepwater boundary, i.e., the angle between the principal wave direction and the line normal to the depth contours. (4) Improved Model for Frequency Spectrum If waves are generated in a laboratory flume using the Bretschneider-Mitsuyasu spectrum expressed by equation (4.1.23), the significant wave period of the generated waves often deviates from the target significant wave period. The reason for such a deviation is that the original equation (4.1.23) is given in terms of the peak frequency fm, but this is replaced with the significant wave period T1/3 by using equation (4.1.27). Goda 54) has thus proposed the following standard spectral form for which the significant wave period of the generated waves does not deviate from the target significant wave period. 2

4 5

4

S ( f ) = 0.205H 1 ¤ 3 T1 ¤ 3 f exp [ 0.75 (T1 ¤ 3 f ) ]

(4.1.28)

The peak frequency for equation (4.1.28) is about 8% lower than that for equation (4.1.23), the spectral density at the peak is about 18% higher, and overall the spectrum is shifted towards the low frequency side. At the very least, it is advisable to use the spectral form expressed by equation (4.1.28) for the target spectrum in hydraulic model experiments. (5) Relationship between Wave Spectrum and Typical Values of Wave Characteristics (a) Wave spectrum and typical value of wave height If the probability density function for the occurrence of a wave height H is assumed to follow the Rayleigh distribution, then the relationship between the mean wave height H and the zeroth moment of the wave -39-


spectrum m0 is given by equation (4.1.30), where the n-th moment of the wave spectrum is defined as in equation (4.1.29). mn = H =

¥

ò0 f n S ( f ) df

(4.1.29) (4.1.30)

2pm 0 ≒ 2.5 m 0

Using the relationship H1/3 = 1.60 H , one arrives at the following relationship between the significant wave height and the spectrum. (4.1.31)

H 1 ¤ 3 ≒ 4.0 m 0

According to the results of actual observations, it is often the case that the best relationship is H 1 ¤ 3 = 3.8 m 0 . In the case of wave data from the shallow waters where the wave height is large, the waves are highly nonlinear and so the relationship H 1 ¤ 3 = 4.0 m 0 is satisfied. In either case, there is a very strong correlation between H 1 ¤ 3 and m0. It is thus acceptable to use equation (4.1.31) and calculate the significant wave height from the spectrum. (b) Wave spectrum and typical value of period When waves are defined using the zero-upcrossing method, the mean period Tz is given by the following equation according to Rice’s theory. Tz =

m0 ¤ m2

N = 171 data

Mean ; 0.832 Standard deviation ; 0.072

(4.1.32)

Calculating the mean period using the BretschneiderMitsuyasu type spectrum gives the following relationship: T z = 0.74T 1 ¤ 3

(4.1.33)

Figure T- 4.1.6 shows a comparison between the mean periods T obtained from actually observed wave profiles and the mean periods Tz estimated from spectrum calculations. The values of Tz / T are distributed in the range 0.6 ~ 1.0, with the mean being 0.83. In other words, the mean values obtained from wave profiles tend to be about 20% greater than those calculated from the moments of spectra. The deviation from Rice’s theory is thought to have been caused by the presence of second order nonliner components in the high frequency range of wave spectra.

Fig. T- 4.1.6 Frequency Distribution of the Ratio of Mean Period Tz by Spectral Calculation to Actually Measured Mean Period T

(6) Spectrum for Long-Period Waves The above explanation concerns the spectra for wind waves and swell components that have a relatively short period. For long-period wave components that have a period of tens of seconds or more, see 4.8 Long-period Waves and Seiche.

4.2 Method of Determining Wave Conditions to Be Used in Design 4.2.1 Principles for Determining the Deepwater Waves Used in Design (Notification Article 4, Clause 2) The duration of statistical wave data used in setting the deepwater wave conditions for investigating the stability of the structures of port and harbor facilities etc. shall be determined appropriately, in due consideration to the functions of the port and harbor facilities and the characteristics of the structures. [Commentary] (1) As for actual measurement data, a relatively long period of measurements (10 years or more) is desirable. However, when there is a lack of such actual measurement data, hindcasted values that have been obtained using at least about 30 years’ worth of meteorological data should be used, with these being corrected by means of the available data of actual wave measurement. (2) When hindcasted values obtained from meteorological data are corrected using actual measurement data, it is necessary that the measurement data should cover the period of 3 years at the minimum and contain a considerable number of cases of large storms. However, if waves were recorded during an extraordinary weather that only occurs once every a few tens of years and the values for these waves exceed all the hindcasted values, the observed values may be used to obtain the design deepwater waves. -40-


(3) If there is absolutely no actual measurement data at the site of interest, or if the only measurement data available is for extremely limited conditions, measurement data for a neighboring place with similar natural conditions may be used. In this case, NOWPHAS (Nationwide Ocean Wave Information Network for Ports and Harbors) data may be used. (4) If it is known that an extraordinary storm event occurred in the area before the period for which wave hindcasting using meteorological data is carried out (for example, in a previous decade), the record of such an event should be taken into consideration. (5) When hindcasted values for a hypothetical typhoon are used, it is advisable to sufficiently investigate the magnitudes of past typhoons and the courses that they followed, and to even include an investigation on the occurrence probability of such a typhoon. (6) When estimating deepwater waves using actual measurement data, it is neccessary to take into account the fact that the measured wave height has been affected by refraction and shoaling. Thus the wave height of the deepwater waves should be corrected by dividing the measured height by the refraction coefficient and the shoaling coefficient. In this case, it is also necessary to consider changes in the wave direction. (7) If the significant wave height obtained from actual measurement data is more than one half of the water depth at the measurement location, it is considered that this wave record has been affected by wave breaking. With such wave data, the parameters of the deepwater waves should be estimated by means of wave hindcasting. Note however that, with regard to the hindcasted deepwater waves, significant waves for the measurement location should be estimated as described in 4.5 Transformations of Waves, and a comparison with the actual measurement data should be carried out. (8) It is advisable to determine the deepwater waves that will be used in design with consideration of the encounter probability based on the return period and the lifetime of the structure in question. However, the way in which the encounter probability is interpreted will depend on the functions, importance and return on investment of the structure, and other factors, and so it is not possible to determine it for the general case. It must therefore be determined independently for each individual case by the judgement of the engineer in charge. Here, the “encounter probability” means the probability that waves with a height larger than the return wave height for a given return period occurs at least once during the lifetime of the structure in question. (9) When determing the deepwater waves that will be used in design, it is necessary to examine the external forces on and past damage of existing structures adjacent to the structure under design. (10) It is standard to set deepwater wave parameters separately for each direction of the sixteen-point bearings, although the directions for which the wave height is small and their effects on the structure are readily judged as negligible may be excluded. The wave direction hereby refers to the direction of the irregular wave component that has the highest energy density, in other words, the principal direction. Since the wave force acting on the structure in question will not change greatly when the wave direction changes by only a few degrees, it is acceptable in design to represent the wave direction using the sixteen-point bearing system.

4.2.2 Procedure for Obtaining the Parameters of Design Waves First, deepwater waves shall be determined by following 4.2.1 Principles for Determining the Deepwater Waves Used in Design. Then, transformations due to refraction, diffraction, shoaling, and breaking shall be evaluated. Finally, the waves that have the most unfavorable effects on the structure in question or facilities in the hinterland shall be used as the design waves. [Technical Notes] The parameters of the design waves are determined according to the following procedures: (1) The effects of wave transformation such as refraction, diffraction, shoaling, and breaking are applied to the deepwater waves determined by following 4.2.1 Principles for Determining the Deepwater Waves Used in Design, in order to determine the parameters of the design waves at the design location. (2) If the location in question is subject to special conditions (for example, disturbances from externally reflected waves or an increase in wave height due to concave corners), these should also be taken into account. (3) The wave force and other wave actions on the structure in question such as overtopping are determined for the waves obtained above. (4) According to the various conditions related to wave actions, there can be cases where the wave force becomes largest when the water level is low, and so investigations should be carried out for all conceivable water levels. (5) The above calculations are carried out for each possible direction in which the deepwater waves may come in. The deepwater waves for which the wave action is largest or for which the effects on the structure in question or facilities in the hinterland are most unfavorable are chosen as the design waves.

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4.3 Wave Hindcasting 4.3.1 General Wave hindcasting shall be carried out by using an appropriate hindcasting method. [Commentary] (1) Wave hindcasting should be made in the following two steps: (a) Setting of the wind field (b) Calculation of wave development and attenuation. (2) The field where waves are generated and developed is called the fetch (or wind field), and it is characterized by four parameters: wind velocity, wind direction, fetch length, and wind duration. Where the wind field is set, the wave development and attenuation should be calculated by using the most appropriate hindcasting method for the wind field conditions. [Technical Notes] The wind field is to be set according to the following procedures: (a) (b) (c) (d) (e)

Collection of surface weather charts and meteorological data. Determination of the duration of hindcasting for each case. Calculation of gradient winds from the surface weather charts. Estimation of the sea surface winds by empirical formulas and data of measurement. Preparation of the wind field chart.

4.3.2 Wave Hindcasting in Generating Area For the hindcasting of waves in the generating area, the spectrum methods and the significant wave methods are recommended as standard methods. [Commentary] The reliability of the results of the wave hindcast should be examined through the comparison with the wave measurement data. [Technical Notes] (1) Spectral Methods (a) General Spectral methods can be classified into the spectral component methods that have been developed by assuming that the components of the spectrum for each frequency and direction develop independently until some equilibrium state is reached 6),7), and the parameteric methods that are based on the idea that the development and decay of a wave spectrum can be described by a certain small number of parameters 8),9),10). With the former, the development of waves is described in terms of the influx of energy from the wind into the component waves that make up the spectrum and the weak nonlinear interaction between component waves. With the latter, development of waves are treated as the overall result of strong nonlinear effects and a kind of similarity mechanism is assumed with introduction of a few parameters. Calculations are carried out by formulating and solving the equations that govern the development and transformation processes of waves using the parameters. The accuracy of wave hindcasting by spectral methods has not been sufficiently investigated yet. However, since the accuracy of wave hindcasting depends greatly on the accuracy of estimating ocean winds, at present it is reasonable to believe that the accuracy of spectral methods is comparable to that of significant wave methods. Nevertheless, it should be noted that even for the same wave hindcasting model, results can vary by 10 ~ 20% due to differences in the matters like the calculation mesh, the boundary conditions or empirical constants. Accordingly, it is necessary to investigate the validity and accuracy of hindcasted results by comparing them with observation values (examples of such comparisons are given in references 6)~11)). In particular, an equilibrium spectral form is assigned as the limit of wave development in the current spectral methods. It is thought that the accuracy of the supposed equilibrium spectrum itself affects the results greatly, and so it is a good idea to investigate the accuracy with regard to the functional forms of frequency spectrum or the directional spectrum. This is because the significant wave height is proportional to the square root of the integral of the directional spectrum, meaning that the calculation is such that the significant wave height does not change very much even if the spectral form itself changes somewhat, and so it is considered that the most rigorous way of carrying out evaluation is to examine the spectral form. The spectral methods have the following advantages over the significant wave methods.

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① The effects of the variations of wind speed and direction on wave development are physically well

described. ② Appropriate estimation results on wave heights and periods are obtained even when the wind field moves

together with wave propagation. ③ Wind waves and swell mixed sea conditions can be reproduced in one calculation.

Accordingly, if the results of hindcasting using a significant wave method seem dubious, it is a good idea to make hindcasting again using a spectral method. Incidentally, spectral methods have been researched and developed while primarily focusing on deepwater waves. There are only a few studies concerning shallow water waves, namely Collins 12), Cavaleri 13), Golding 14) and Yamaguchi et al. (b) Details 6),7) Wave forecasting methods by mean of wave spectrum have been developed by many researchers since the 1960s. Those developed by Japanese reserchers include Inoue’s model 6), Isozaki and Uji’s MRI model 7), and Yamaguchi and Tsuchiya’s model. The basis of these models is the following energy balance equation: ¶ ----E ( f, q, t, x ) = C G ( f ) ÑE ( f, q, t, x ) + a ( f, U ) + b ( f, U )E ( F, q, t, x ) + F 3 + F 4 + F 5 (4.3.1) ¶t where E ( F, q, t, x )： energy density of a two-dimensional wave spectrum a ( f, U )： linear amplifying factor in Phillips’ resonance theory 15) b ( f, U )： exponential amplifying factor in Miles’ theory 16) F 3： energy dissipated due to wave breaking F 4： energy loss due to internal friction during wave propagations etc. F 5： energy exchange due to the nonlinear interaction between component waves f, q： component wave frequency and angle t： time x： position vector C G ( f )： group velocity vector U： wind velocity Ñ： differential operator

(2) Significant Wave Methods (a) S-M-B method ① General 19),20)

The S-M-B method is used when the wind field is stationary. The height and period of deepwater significant waves are estimated from the wind velocity and wind duration in the fetch and the fetch length using Fig. T- 4.3.1. Of the wave height obtained from the wind velocity and that from the wind duration, the lower one is adopted as the hindcasted value; likewise for the period. Figure T- 4.3.1 has been drawn based on the relationships by equations (4.3.2), (4.3.3) and (4.3.4), which were rewritten by Wilson 21) in 1965. gH 1 ¤ 3 1 -------------- = 0.30 1 ---------------------------------------------------22 1 ¤ 2ü U ì æ gFö í 1 + 0.004 è -----2-ø ý U î þ

(4.3.2)

gT 1 ¤ 3 1 -------------- = 1.37 1 ---------------------------------------------------52pU 1¤3 ì æ gFö ü í 1 + 0.008 è -----2-ø ý U î þ t =

F

dF

ò 0 C------Gi

=

F

(4.3.3)

dF

-----------------------ò 0 gT 1 ¤ 3 ¤ 4p

(4.3.4)

i

where H 1 ¤ 3： significant wave height (m) T 1 ¤ 3： significant wave period (s) U： wind velocity at 10 m above sea surface (m/s). F： fetch length (m) g： acceleration of gravity (m/s2) (= 9.81 m/s2) t： minimum duration (hr) -43-

Wind Speed


Fetch ( H 1 3 T 1 3 ) 2 = const. t (h) H 1 3 (m) T 1 3 (s) Fig. T - 4.3.1 Wave Forecasting Diagram by the S-M-B Method ② Handling of the effective fetch length

When the fetch width is small relative to the fetch length (for example, in a long bay), the fetch length is determined by the distance to the opposite shore. If the distance to the opposite shore varies greatly when the direction is changed only slightly, it is advisable to use the effective fetch length defined by in equation (4.3.5) 22) when hindcasting is made. 2

SF i cos q i F eff = -----------------------S cos q i

(4.3.5)

where F eff： effective fetch length (km) F i： distance to opposite shore in the i-th direction (km) q i： angle between the direction of Fi and the predominant wind direction (º) (b) Wilson’s method 21), 23) Wilson’s method is an extension of the S-M-B method. It includes improvements that it can be applied even to a moving fetch, for example in the case of a typhoon. Using the H1/3-t-F-T1/3 graph shown in Fig. T- 4.3.2, the propagation of waves is traced in the F-T plane, while the development of the significant wave height and period are traced in the H1/3-t plane and T1/3-t plane, respectively. This figure has been obtained by calculation based on equations (4.3.2), (4.3.3) and (4.3.4). (c) Hindcasting for shallow water waves Methods that consider the influence of the water depth on wave development (i.e., the energy loss due to friction with the sea bottom) include the Sakamoto-Ijima method. It is known from experience that the significant wave period and the significant wave height satisfy the following relationship. (Note however that this applies only for wind waves within the fetch area.) T 1 ¤ 3 = 3.86 H 1 ¤ 3

(4.3.6)

where H 1 ¤ 3： significant wave height (m) T 1 ¤ 3： significant wave period (s)

-44-


Fig. T- 4.3.2 H1/3-t-F-T1/3 Graph (from Wilson's equations (1965))

In the Sakamoto-Ijima method, the ideas in Wilson’s method for deep water waves have been incorporated into the case for shallow water waves, resulting in an H1/3-t-F-CG graph such as shown in Fig. T- 4.3.3. With use of such a graph it possible to carry out the hindcasting of shallow water waves in a variable fetch.

(A) Note: The numbers on the graph show wind velocity (m/s), with water depth (m) in brackets

Fig. T- 4.3.3 H1/3-F-CG Graph for Shallow Water Waves (Sakamoto-Ijima Method) -45-


4.3.3 Swell Hindcasting When swell hindcasting is necessary, it is standard to use the Bretschneider method. [Commentary] Swell hindcasting methods include the Bretschneider method 24), the P-N-J method 5), and spectral methods. With the Bretschneider method, the wave height and period of swell are hindcasted from the parameters of the significant wave. With the P-N-J method, the swell parameters are obtained by estimating the effects of the velocity dispersion and directional spreading of spectral components. With spectral methods as mentioned above, numerical calculations are used; generally, no distinction is made between waves and swell in the generating area, with calculations for the component waves at all of the different frequencies being carried out simultaneously, and the results being the significant wave parameters for the combination of wind waves and swell. If a significant wave method is used in the hindcasting of waves in the generating area, it is necessary to hindcast swell, in which case it is standard to use the Bretschneider method, which is relatively simple and easy to use. Note however that the amount of reliable observation data that has been obtained for swell is insufficient, and so the hindcasting accuracy is lower than that for waves in the generating area. Accordingly, it is necessary to treat swell hindcast values as representing no more than approximate values, and it is advisable to use them only after carrying out a comparative investigation with actual measurement data. [Technical Notes] In the Bretschneider method, swell hindcasting is carried out by using Fig. T- 4.3.4.

Fig. T- 4.3.4 Swell Hindcasting Diagram

The term Fmin in the diagram is the minimum fetch length, D is the decay distance of the swell, HF and TF are the height and period of the significant wave at the end of the fetch respectively, and HD and TD are the height and period of the significant wave at the swell hindcasting point respectively. If the significant wave height and period are determined by the wind velocity and the fetch length in the S-M-B method, the minimum fetch length Fmin is equal to the actual fetch length. If the wave development is governed by the wind duration, then Fmin is the fetch length corresponding to that wind duration and wind velocity. The time t required for waves to propagate over the decay distance D is calculated from the following equation: 4pD D t = ----------- = ----------gT D C GD

(4.3.7)

where C GD：group velocity corresponding to T D (m/s) -46-


4.4 Statistical Processing of Wave Observation and Hindcasted Data (1) Wave characteristics shall be expressed as joint distributions of wave height and period by wave direction using the monthly, seasonal, and annual wave data. (2) Storm wave data shall be sorted by the peaks-over-threshold method so as to yield the data set of extreme wave heights for extreme statistical analysis, and the extreme wave heights shall be expressed in terms of the return perid. [Commentary] (1) The wave distribution characteristics for ordinary conditions are expressed separately for each wave direction as a joint distribution of wave height and period. Observation data are often available for the wave height and the period, and so it is standard to use such data. If observation data are not available, then hindcast data is used. Since waves in ordinary conditions are often affected by the local wind, it is necessary to gain a sufficient understanding of the local wind characteristics. There is generally not much observation data available for the wave direction, and so it is standard to use hindcasting. It is necessary to give sufficient consideration to the effects of swell. (2) It is standard to represent the height of waves used in the design of protective facilities as the “return wave height” for the return period of the “peak waves” using data over a long time period (at least 30 years as a general rule). Since there are only a few places at which observation data extending over such a prolonged time duration are available, generally hindcast data must be used. (3) The peak waves, basic data for estimating the return wave height, are the wave (generally the significant wave) at the time for which the wave height becomes largest during the process of wave development and decay under a certain meteorological condition. It is thought that sampled peak waves are mutually independent in statistical sense. When estimating the return wave height, it is possible to use the time series of data for which the peak waves exceed a certain threshold value during the period in question. Alternatively, it is possible to obtain the maximum value of the “peak waves” for each year, and then use the data as the annual maximum wave. In either case, the theoretical distribution function of the return wave height is not known, and so one should try to fit several distribution functions such as the those of the Gumbel distribution and the Weibull distribution, find the functional form that best fits the data, and then extrapolate it in order to estimate the return wave heights for a number of different return periods (say 50 years, 100 years, etc.). The accuracy of the resulting estimated values depends largely on the reliability of the data used rather than on the statistical processing method. When drawing up the data set of peak waves using wave hindcasting, it is thus necessary to take due care in appropriately selecting the hindcasting method and to closely inspect the hindcasted results. (4) With regard to the wave period corresponding to the return wave height, the relationship between the wave height and the wave period is plotted for the data of peak waves (which have been used in estimating the return wave height), and then the wave period is determined appropriately based on the correlation between the two. [Technical Notes] (1) Estimation of Return Wave Height During statistical processing, the wave heights are rearranged in the descending order, and the probability of each value of wave height not being exceeded is calculated. If there are N data and the m-th largest wave height is denoted with xm,N, then the probability P that the wave height does not exceed xm,N is calculated using the following equation: ma P [ H ≦ x m, N ] = 1 -------------N+b

(4.4.1)

Table T- 4.4.1 Parameters Used in Calculating the Probability not Exceeding a Certain Wave Height Distribution function Gumbel distribution Weibull distribution “ “ “ “ “ “

(k = 0.75) (k = 0.85) (k = 1.0) (k = 1.1) (k = 1.25) (k = 1.5) (k = 2.0)

a

b

0.44 0.54 0.51 0.48 0.46 0.44 0.42 0.39

0.12 0.64 0.59 0.53 0.50 0.47 0.42 0.37

The values used for a and b in this equation depend on the distribution function. Specifically, values such as those in Table T- 4.4.1 are used. The values used for the Gumbel distribution were determined by Gringorten 25) in such a way as to minimize the effects of statistical scatter in the data. The values used for the Weibull distribution were determined by Petruaskas and Aagaard 26) using the same principle. It is commented that the Thomas plot often used in hydrology corresponds to the case a = 0, b = 1, and the Hazen plot corresponds to the case a = 0.5, b = 0. The distribution functions used in hydrology include the Gumbel distribution (double exponential distribution), the logarithmic extreme value distribution, and the normal distribution (in the last case, the data must first be transformed appropriately). Since the data on peak wave heights have not been accumulated over a prolonged period of time, it is not well known which distribution function is most suitable. -47-


Following Petruaskas and Aagaard, we thus introduce the method whereby one tries fitting eight distribution functions, namely the Gumbel distribution function (equation (4.4.2)) and the Weibull distribution function (equation (4.4.3)) with k = 0.75, 0.85, 1.0, 1.1, 1.25, 1.5 and 2.0; the distribution function that best fits the data on any particular data set is then selected as the extreme distribution for that data set. ì xB ü P [ H ≦ x ] = exp exp í æ ------------ö ý è A ø î þ

(Gumbel distribution)

(4.4.2)

ì x B kü (Weibull distribution) (4.4.3) P [ H ≦x ] = 1 exp í æ ------------ö ý è A ø î þ In order to fit the data to the distribution function, the “non-exceedance probability” (probability not exceeding a certain wave height) P is transformed into the variable r v ( = ( x B ) ¤ A ) using equation (4.4.4) or (4.4.5). rv = ln { ln P [ H ≦ x ] } rv = [ ln { 1 P [ H ≦ x ] } ]

1¤k

(Gumbel distribution)

(4.4.4)

(Weibull distribution)

(4.4.5)

If the data fit equation (4.4.2) or (4.4.3) perfectly, then there will be a linear relationship between x and r v . Accordingly, the data are assumed to follow the linear relationship shown in equation (4.4.6). The parameters A and B are determined using the method of least squares, thus giving an equation for estimating the return wave height. x = A^ r v + B^

(4.4.6)

where A^ and B^ are the estimated values of the parameters A and B in equation (4.4.2) or (4.4.3), respectively. The return period Rp of the wave height H is related to the non-exceedance probability P (H ≦ x) as in the following: 1 Rp = K ---- ------------------------------N 1 P ( H≦ £ x) ·

or

(4.4.7)

K P ( H ≦ x ) = 1 ---------NRp where K： number of years during the period for which analysis was carried out N： number of data of peak waves (2) Candidate Distribution Functions and Rejection Eriteria Goda has proposed the following method 51) ~ 53), which is a revised version of the method introduced above. (a) Addition of the Fisher-Tippett type II distribution to the candidate distributions The Fisher-Tippett type II distribution is given by the following equation. k

P [ H ≦x ] = exp [ { 1 + ( x B ) ¤ ( kA ) } ]

(4.4.8)

The following nine functions are employed as the candidate functions to be tried for fitting: the Gumbel distribution function (equation (4.4.2)), the Weibull distribution function (equation (4.4.3)) with k = 0.75, 1.0, 1.4 and 2.0 (four preset values), and the Fisher-Tippett type II distribution function with k = 2.5, 3.33, 5.0 and 10.0 (four preset values). In place of the values listed in Table T- 4.4.1, the following equations are used for a and b in equation (4.4.1): For the Gumbel distribution, a ＝ 0.44, b ＝ 0.12

(4.4.9)

For the Weibull distribution, a = 0.20 + 0.27 k

(4.4.10)

b = 0.20 + 0.23 k For the Fisher-Tippett type II distribution, a = 0.44 + 0.52 ¤ k b = 0.12 0.11 ¤ k

(4.4.11)

-48-


(b) Selection of the best function through introduction of rejection criteria Inappropriate functions are rejected by means of two sets of criterion. The first is the REC criterion. For the residual of the correlation coefficient for each distribution function, the 95% non-exceedance probability level is determined in advance. If the residual of the correlation coefficient exceeds this threshold value for a distribution function when the extreme value data is fitted to that distribution function, the function in question is rejected as being inappropriate. The second is the DOL criterion. The maximum value in the data is made dimensionless using the mean and standard deviation for the whole data. If this value is below the 5% or above the 95% level of the cumulative distriburion of dimensionless deviation of the distribution function being fitted, that function is rejected as being inappropriate. Next, the best function is selected not simply according to the value of the correlation coefficient, but rather according to the MIR criterion, This criterion takes into account the fact that the mean of the residual of the correlation coefficient relative to 1.0 will vary according to the distribution function. The function for which the ratio of the residual of the correlation coefficient of the sample to the mean residual for the fitted distribution is lowest is judged to be the best fitting distribution function.

4.5 Transformations of Waves 4.5.1 General (Notification Article 4, Clause 3) As a general rule, the waves to be considered to exert actions on port and harbor facilities shall be the waves that are most unfavorable in terms of the structure stability or the usage of the port and harbor facilities. In this consideration, appropriate attention shall be given to wave transformations during the propagation of waves from deepwater toward the shore, which include refraction, diffraction, shoaling, breaking, and others. 4.5.2 Wave Refraction The phenomenon of wave refraction occurs in intermediate depth to shallow waters. This is due to the change in local wave velocity caused by the change in water depth. The changes in wave height and wave direction due to refraction shall be considered. [Technical Notes] (1) Refraction Calculations for Regular Waves (a) Refraction phenomenon and refraction coefficient (see Fig. T- 4.5.1) If waves are obliquely incident on a straight boundary where the water depth changes from h1 to h2, waves are refracted at the boundary due to the change in wave velocity caused by the change in water depth. Suppose that the distance between wave rays changes from b1 to b2 as a result. If the change in the wave ray width is not so large, it can be assumed that no wave energy flux cuts across the wave ray and flows out. If other sources of energy loss such as the friction along the sea bottom are ignored, then the continuity in the flux of energy transport results in the change of the wave height H1 at water depth h1 to the wave height H2 at water depth h2 as given by the following equation: H2 ------ = H1

C G1 b 1 ---------- ----C G2 b 2

(4.5.1)

where CG1 , CG2： group velocities at water depths h1 and h2, respectively (m/s) b1 , b2： distances between wave rays at water depths h1 and h2, respectively (m) In the equation,

b 1 ¤ b 2 represents the change in wave height

due to refraction, while C G1 ¤ C G2 represents the change in wave height due to the change in water depth. Using the shoaling coefficient (see 4.5.5 Wave Shoaling),

C G1 ¤ C G2 can be

represented as C G1 ¤ C G2 = K s2 ¤ K s1 , where Ks1 and Ks2 are the shoaling coefficients at water depths h1 and h2, respectively. Suppose that the wave ray width, which is b0 for deepwater waves, changes to b due to the refraction phenomenon. The ratio of the wave height after the change to the original wave height in this case is called the “refraction coefficient”. The refraction coefficient Kr is given by the following equation: Kr =

b0 ¤ b

(4.5.2) -49-

water depth h1 water depth h2

Fig. T- 4.5.1 Schematic Diagram of Wave Refraction


(b) Refraction calculation methods Refraction calculation methods for regular waves include the wave ray methods in which calculations using a computer are made possible, and the numerical wave propagation analysis methods 27) in which surface wave equations are solved by computers using finite difference schemes. An appropriate calculation method is chosen in accordance with the situation. Note however that for a coastline for which the depth contours are straight and parallel to one another, the change in the wave direction and the refraction coefficient can be calculated using the following equations: 2ph sin a = sin a 0 tanh ---------L

(4.5.3)

cos a 0 --------------cos a

(4.5.4)

Kr =

Here, L, a and a0 denote the wavelength at water depth h, the angle of incidence of the wave at water depth h, and the angle of incidence of the wave in deep water, respectively. Figures T- 4.5.2 and T- 4.5.3 show the refraction coefficient and the wave direction, as calculated using equations (4.5.4) and (4.5.3), respectively.

Fig. T- 4.5.2 Refraction Coefficient of Regular Waves at Coast with Straight, Parallel Depth Contours

Fig. T- 4.5.3 Graph Showing the Change in the Wave Direction of Regular Waves at Coast with Straight, Parallel Depth Contours

(2) Range of Application of Refraction Calculations Using Regular Waves Based on the principles behind calculations for regular waves, such calculations are applicable for waves for which there is little directional spreading and the frequency band is narrow; for example, swell-type waves and tsunamis. For waves like wind waves for which there is much directional spreading and the frequency band is broad, it is necessary to carry out refraction calculations for irregular waves. Nevertheless, comparing the graphs showing changes in the refraction coefficient and wave direction for regular waves and irregular waves at a coast -50-


with straight, parallel depth contours, it can be seen that there is only a little difference between regular waves and irregular waves in this case. This means that when the topography of a coastline is monotonous to the extent that the depth contours are considered to be straight and parallel to the shoreline, the difference between the results of refraction calculations for regular waves and irregular waves is usually only slight, and so the results of refraction calculations using regular waves can be used as a good approximation. (3) Refraction Calculations for Irregular Waves (a) Calculation methods Refraction calculation methods for irregular waves include the following: ① the component wave method, whereby the directional wave spectrum is divided into an appropriate number of component waves, a refraction calculation is performed for each component wave, and then the refraction coefficient for the irregular wave is evaluated by making a weighted average of the component wave energies; ② methods in which the wave energy balance equation 28) or the mild-slope wave equation is solved directly using a computer with finite difference schemes. As with the component wave method, the energy balance equation is derived by assuming that wave energy does not cut across wave rays and flow out. This means that the technique is basically the same in both cases. However, with the energy balance equation method, refraction within a small but finite region is calculated, meaning that the refraction coefficient does not become infinite even at a point in which two regular wave rays may converge. On the other hand, the mild-slope wave equation method is a strictly analytical method, but it is difficult to apply it to a large region. When determining the refraction coefficient for irregular waves, it is acceptable to use the component wave method, which involves the linear superposition of refraction coefficients for regular waves and is thus simple and convenient. However, when intersections of wave rays occur during a refraction calculation for a component wave, the energy balance equation method may be used for practical purposes with the exception of the case that the degree of intersection is large. (b) Effects of diffraction When deepwater waves have been diffracted by an island or a headland, the wave spectrum becomes generally different from a standard form that has been assumed initially. Thus it is necessary to use the spectral form after diffraction when performing the refraction calculation. (c) Diagrams of the refraction coefficient and angle for irregular waves at a coast with straight, parallel depth contours Figures T- 4.5.4 and T- 4.5.5 show the refraction coefficient Kr and the principal wave direction ap, respectively, for irregular waves at a coast with straight, parallel depth contours, with the principal direction of deepwater waves (ap)0 as the parameter. The direction (ap)0 is expressed as the angle between the wave direction and the line normal to the boundary of deepwater. Smax is the maximum value of the parameter that expresses the degree of directional spreading of wave energy (see 4.1.3 [3] Wave Spectrum).

Fig. T- 4.5.4 Refraction Coefficient of Irregular Waves at Coast with Straight, Parallel Depth Contours -51-


Fig. T- 4.5.5 Change Due to Refraction in the Principal Direction ap of Irregular Waves at Coast with Straight, Parallel Depth Contours

(4) At places where the water depth is no more than about one half of the deepwater wave height, waves exhibit the characteristics of flow rather than those of undulations. This means that refraction calculations for wave directions and refraction coefficients can only be applied to the water where the depth is at least one half of the deepwater wave height.

4.5.3 Wave Diffraction [1] Diffraction The wave height in regions in which waves are anticipated to be greatly affected by the phenomenon of diffraction caused by obstacles such as breakwaters or islands shall be calculated using an appropriate method. [Commentary] Diffraction is a phenomenon whereby waves wheel into a region that is screened by something like a breakwater. It is the most important phenomenon when determining the wave height in a harbor. The irregularity of waves should be considered in a diffraction calculation. For a harbor within which the water depth is assumed uniform, the diffraction diagrams for irregular waves with regard to a semi-infinite breakwater or a straight breakwater that has just one opening have been prepared. The ratio of the wave height after diffraction to the incident wave height is called the diffraction coefficient Kd. In other words, the diffraction coefficient Kd is given by the following equation: Kd = Hd ¤ Hi

(4.5.10)

where Hi： incident wave height outside harbor Hd： height of wave in harbor after diffraction Diffraction diagrams and diffraction calculation methods assume that the water depth within the harbor is uniform. If there are large variations in water depth within the harbor, the errors will become large, in which case it is advisable to investigate the wave height in the harbor by means of either hydraulic scale model tests or else numerical calculation methods that also take refraction into account. [Technical Notes] (1) Diffraction Diagrams for Irregular Waves Figures T- 4.5.6 (a) ~ (c) show the diffraction diagrams by a semi-infinite breakwater for irregular waves with the directional spreading parameter Smax = 10, 25, and 75. Figures T- 4.5.6 (a) ~ (l) show the diffraction diagrams through an opening of B/L = 1, 2, 4, and 8 for irregular waves with Smax = 10, 25, and 75.

-52-


---- Period ratio

─ Diffraction coefficient

Wave direction

Wave direction Fig. T - 4.5.6(a) Diffraction Diagram by Semi-infinite Breakwaters (θ= 90º) for Smax = 10

-53-


---- Period ratio


Wave direction

Wave direction Fig. T - 4.5.6(b) Diffraction Diagram by Semi-infinite Breakwaters (θ= 90º) for Smax = 25

-54-


---- Period ratio


Wave direction

Wave direction Fig. T - 4.5.6(c) Diffraction Diagram by Semi-infinite Breakwaters (θ= 90º) for Smax = 75

-55-


Period ratio

Diffraction coefficient

Wave direction

Period ratio


Wave direction Fig. T - 4.5.7(a) Diffraction Diagram by Breakwaters with an Opening (B/L= 1.0) for Smax = 10

-56-


Period ratio


Wave direction

Period ratio


Wave direction Fig. T - 4.5.7(b) Diffraction Diagram by Breakwaters with an Opening (B/L= 1.0) for Smax = 25

-57-


Period ratio


Wave direction

Period ratio


Wave direction Fig. T - 4.5.7(c) Diffraction Diagram by Breakwaters with an Opening (B/L= 1.0) for Smax = 75

-58-


Period ratio


Wave direction

Period ratio


Wave direction Fig. T - 4.5.7(d) Diffraction Diagram by Breakwaters with an Opening (B/L= 2.0) for Smax = 10

-59-


Period ratio


Wave direction

Period ratio


Wave direction Fig. T - 4.5.7(e) Diffraction Diagram by Breakwaters with an Opening (B/L= 2.0) for Smax = 25

-60-


Period ratio


Wave direction

Period ratio


Wave direction Fig. T - 4.5.7(f) Diffraction Diagram by Breakwaters with an Opening (B/L= 2.0) for Smax = 75

-61-


Period ratio


Wave direction

Period ratio


Wave direction Fig. T - 4.5.7(g) Diffraction Diagram by Breakwaters with an Opening (B/L= 4.0) for Smax = 10

-62-


Period ratio


Wave direction

Period ratio


Wave direction Fig. T - 4.5.7(h) Diffraction Diagram by Breakwaters with an Opening (B/L= 4.0) for Smax = 25

-63-


Period ratio


Wave direction

Period ratio


Wave direction Fig. T - 4.5.7(i) Diffraction Diagram by Breakwaters with an Opening (B/L= 4.0) for Smax = 75

-64-


Period ratio


Wave direction

Period ratio


Wave direction Fig. T - 4.5.7(j) Diffraction Diagram by Breakwaters with an Opening (B/L= 8.0) for Smax = 10

-65-


Period ratio


Wave direction

eriod ratio


Wave direction Fig. T - 4.5.7(k) Diffraction Diagram by Breakwaters with an Opening (B/L= 8.0) for Smax = 25

-66-


Period ratio


Wave direction

Period ratio


Wave direction Fig. T - 4.5.7(ll) Diffraction Diagram by Breakwaters with an Opening (B/L= 8.0) for Smax = 75

-67-


(2) Treatment of Obliquely Incident Waves When waves are obliquely incident to a breakwater that contains an opening, it is advisable to obtain the diffraction diagram by means of a numerical calculation. When this is not possible, or when the diffraction diagram is only required as a rough guideline, the following approximate method may be used instead. (a) Determining the axis of the diffracted wave When waves are obliquely incident to a breakwater that contains an opening, the direction q ¢ of the axis of the diffracted waves (see Fig. T- 4.5.8) varies slightly from the direction of incidence q. Tables T- 4.5.1 (a) ~ (c) list the direction of the axis of the diffracted waves as a function of the aperture ratio B/L and the direction of incidence. These tables are used to obtain the direction q ¢ of the axis of the diffracted waves, and then the virtual aperture ratio B¢/L corresponding to q ¢ is obtained from the following equation: B¢ ¤ L = ( B ¤ L ) sin q¢

(4.5.11) Table T- 4.5.1 Angle of Axis of Diffracted Waveθ¢

(a) Smax = 10 B/L 1.0 2.0 4.0

Angle between breakwater and incident wave direction q 15º

30º

45º

60º

53º (38º) 46º (31º) 41º (26º)

58º (28º) 53º (23º) 49º (19º)

65º (20º) 62º (17º) 60º (15º)

71º (11º) 70º (10º) 70º (10º)

(a) Smax = 25 B/L 1.0 2.0 4.0

Angle between breakwater and incident wave direction q 15º 49º (34º) 41º (26º) 36º (21º)

30º 52º (22º) 47º (17º) 42º (12º)

45º

60º

61º (16º) 57º (12º) 54º (9º)

70º (10º) 67º (7º) 65º (5º)

(a) Smax =75 B/L 1.0 2.0 4.0

Angle between breakwater and incident wave direction q 15º

30º

45º

60º

41º (26º) 36º (21º) 30º (15º)

45º (15º) 41º (11º) 36º (6º)

55º (10º) 52º (7º) 49º (4º)

66º (6º) 64º (4º) 62º (2º)

Note: Angle in the parentheses is the angle of deflection relative to the angle of incidence

Principal direction of diffracted wave

Principal direction of incident wave

Fig. T- 4.5.8 Virtual Aperture B¢ and Angle of Axis of Diffracted Wave θ¢

(b) Fitting of a diffraction diagram Out of the diffraction diagrams of normal incidence in Figs. T-4.5.7 (a) ~ (l), the diffraction diagram that has an aperture ratio nearly equal to the virtual aperture ratio is selected. This diffraction diagram is next rotated until the direction of incidence matches the direction of the axis of the diffracted waves as determined from Table T- 4.5.1. The diffraction diagram is then copied and taken to be the diffraction diagram for obliquely -68-


incident waves. The errors in this approximate method are largest around the opening in the breakwater; in terms of the diffraction coefficient, the maximum error may amount to around 0.1 in the absolute value. (3) Method for Determining Diffraction Coefficient in a Harbor The diffraction coefficient within a complex shape of harbor is generally estimated by numerical computation with a computer. Diffraction calculation methods include Takayama’s method, which involves linear super position of analytical solutions for detached breakwaters, and calculation methods that use the Green functions. (4) Directional Spreading Method When the length of an island or the width of the entrance of a bay is at least ten times the wavelength of the incident waves, there will not be a large difference between the wave height estimate by the direct diffraction calculation and the estimate using the amount of directional wave energy that arrives directly at the point of interest behind the island or in the bay; the latter is called the directional spreading method. However, if the point of interest is just behind an island or headland, the effects of diffracted waves will be large, and so the directional spreading method cannot be applied. (5) Studies Using Hydraulic Model Experiments Thanks to improvements in multidirectional random wave generating devices, it is easy to reproduce waves that have directional spreading in the laboratory nowadays, meaning that diffraction experiments can be carried out relatively easily. When carrying out a model experiment, an opening in the harbor model is set up within the effective wave making zone, and the wave height is simultaneously measured at a number of points within the harbor. The diffraction coefficient is obtained by dividing the significant wave height in the harbor by the significant wave height at the harbor entrance averaged over at least two observation points.

[2] Combination of Diffraction and Refraction When carrying out diffraction calculations for waves in waters where the water depth varies greatly, wave refraction must also be considered. [Commentary] (1) When the water depth within a harbor is made more-or-less uniform by say dredging (this is often the case with large harbors), the refraction of waves after diffraction can be ignored. In order to determine the wave height in the harbor in this case, it is acceptable to first carry out a calculation considering only refraction and breaking from the deepwater wave hindcasting point to the harbor entrance. Next, a diffraction calculation for the area within the harbor is carried out, taking the incident wave height to be equal to the calculated wave height at the harbor entrance. In this case, the wave height at a point of interest within the harbor is expressed using the following equation: H = Kd Kr Ks H0 where Kd： Kr： Ks： H0：

(4.5.12)

diffraction coefficient at the point of interest within a harbor refraction coefficient at the harbor entrance shoaling coefficient at the harbor entrance (see 4.5.5 Wave Shoaling) deepwater wave height

The energy balance equation method or the improved energy balance equation method in which a term representing dissipation due to wave breaking is added is appropriate as the refraction calculation method for the open sea. Takayama’s harbor calmness calculation method, whereby diffraction solutions for detached breakwaters are superimposed in order to obtain the change in the wave height of irregular waves within the harbor due to diffraction and reflection, can be used for the diffraction calculation for the area within the harbor, provided there are no complex topographic variations within the harbor. (2) When there are large variations in water depth even at places screened by a breakwater (this is often the case with relatively small harbors and coastal areas), it is necessary to simultaneously consider both diffraction and refraction within the harbor. If ignoring wave reflection and just investigating the approximate change in wave height, it is possible to carry out refraction and diffraction calculations separately, and then estimate the change in wave height by multiplying together the refraction and diffraction coefficients obtained. Calculation methods that allow simultaneous consideration of refraction and diffraction of irregular waves include a method that uses time-dependent mild-slope irregular wave equations, a method in which the Boussinesq equation is solved using the finite difference method 29), and the multicomponent coupling method of Nadaoka et al. There are also literatures in which other calculation methods are explained.

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4.5.4 Wave Reflection [1] General In the design of port and harbor facilities, investigations shall be carried out onto the effects of reflected waves from neighboring structures on the facilities in question and also the effects of wave reflection from the facilities in question on neighboring areas. [Commentary] It is necessary to take note of the fact that waves reflected from port and harbor facilities can exercise a large influence on the navigation of vessels and cargo handling. For example, waves reflected from vertical breakwaters can cause disturbances in navigation channels, and multiple-reflected waves from quaywalls can cause agitations within harbors. [Technical Notes] (1) Composition of Reflected Waves and Incident Waves The wave height Hs when incident waves and waves reflected from a number of reflective boundaries coexist (a train of incident waves and those of reflected waves from reflective boundaries are termed the “wave groups”) can be calculated using the following equation: Hs =

2

2

2

H 1 + H2 + ¼ + Hn

(4.5.13)

where Hs: significant wave height when all of the wave groups are taken together H1, H2, ¼ Hn: significant wave heights of wave groups Note however that, if the wave action varies with the wave direction, the differences in the wave directions of various wave groups must be considered. The calculated wave height is valid for places that are at least about 0.7 wavelengths away from a reflecting boundary. Regarding the diffraction and/or refraction of waves for which wave direction is an important factor, the significant wave height is determined separately for each wave group by carrying out whatever calculation is necessary for that wave group, when the wave directions of various wave groups differ. Then the composite wave height is calculated by putting these significant wave heights into equation (4.5.13). An acceptable alternative is to determine the spectrum for each wave group, add these spectra together in order to calculate the spectral form when the wave groups coexist, and then perform direct diffraction and/or refraction calculations using this spectrum. (2) Composition of Periods The significant wave height to be used in calculating the wave force when two wave groups of different periods are superimposed may be determined by the energy composition method (i.e., equation (4.5.13)). The significant wave period T1/3 may be determined using the following equation 30): 2

2

( H 1 ¤ 3 ) I + ( H 1 ¤ 3 )II T 1 ¤ 3 = k -----------------------------------------------------------------------------------------2 2 2 2 ( H 1 ¤ 3 ) I ¤ ( T 1 ¤ 3 ) I + ( H 1 ¤ 3 )II ¤ ( T 1 ¤ 3 )II

(4.5.14)

where k = 1.0 + a ( R H ¤ m ) 0.121Aln ( RH ¤ m)

(4.5.15)

a = 0.08 ( ln R T ) 2 0.15 ln R T

(4.5.16)

ì 0.632 + 0.144 ln RT m = í î 0.6 ì 13.97 + 4.33 ln RT A = í î 10.0

: : : :

0.1 ≦ RT ＜ 0.8 0.8 ≦ RT ＜ 1 0.1 ≦ RT ＜ 0.4 0.4 ≦ RT ＜ 1

(4.5.17) (4.5.18)

R H = ( H 1 ¤ 3 ) I ¤ ( H 1 ¤ 3 ) II

(4.5.19)

R T = ( T 1 ¤ 3 ) I ¤ ( T 1 ¤ 3 ) II

(4.5.20)

(H1/3)I, (H1/3)II : significant wave heights of wave groups I and II before superimposition, respectively (m) (T1/3)I, (T1/3)II : significant wave periods of wave groups I and II before superimposition, respectively (s) Note that, in the above equations, I is assigned to the wave group with the shorter period and Ⅱto that with the longer period.

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(3) Methods for Calculating the Effects of Reflected Waves Calculation methods for investigating the extent of the effects of waves reflected from a structure include the poligonal island reflection method and a simple method by means of diffraction diagrams. (a) Poligonal island reflection method In this calculation method, the theoretical solution that shows the wave transformation around a single convex corner is separated into three terms, representing the incident, the reflected and the scattered waves, respectively. The term for the scattered waves is progressively expanded to obtain an approximate equation, so that the method can be applied to the case where there are a number of convex corners. When there are a number of convex corners, it is assumed as a precondition that the lengths of the sides between convex corners are at least five times the wavelength of the incident waves, so that the convex corners do not interfere with each other. It is necessary to take heed of the fact that errors may become large if the sides are shorter than this. Since another assumption is made such that the water depth is uniform, the refraction of reflected waves cannot be calculated. In general, it is sufficient for practical purposes if the lengths of the sides between convex corners are at least about three times the wavelength of the incident waves. This calculation method can also be applied to the reflection of irregular waves by means of superposing component waves. Although the wave diffraction problems can also be analyzed with this calculation method, there will be large errors if it is applied to the diffraction of waves by thin structures such as breakwaters. (b) Simple method by means of diffraction diagrams Explanation is made for the example shown in Fig. T- 4.5.9. The wave height at a point A on the front face of an upright detached breakwater is estimated when waves are incident on the detached breakwater at an angle a. Instead of the detached breakwater, it is supposed that there are two semi-infinite virtual breakwaters with an opening, such as shown with dashed lines in Fig. T- 4.5.9. Next, one considers the situation whereby waves are incident on the virtual opening from both the wave direction of the incident waves and the direction symmetrical to this with respect to the detached breakwater (i.e., the direction shown by the dashed arrow in Fig. T4.5.9), and draws the diffraction diagram for the Fig. T- 4.5.9 Sketch Showing the Effect opening (dashed lines in Fig. T- 4.5.9). The range of of Reflected Waves influence of the reflected waves is represented by means of the diffraction diagram for the virtual breakwaters with the opening. Accordingly, supposing that the diffraction coefficient at point A is read off as being 0.68, then the wave height ratio with respect to the incident waves at point A is obtained by combining this value of 0.68 with a value of 1.0 representing the incident waves; since it is the energies that are added, the wave height ratio becomes 1 + 0.68 2 = 1.21 . It should be noted, however, that this value of 1.21 represents the mean value of the wave height ratio around the point A. It is not advisable to use this method for points within 0.7 wavelengths of the detached breakwater, because the errors due to a phase coupling effect will be large. For the case of wave reflection by a semi-infinite breakwater, the virtual breakwater also becomes a semiinfinite breakwater in the opposite direction, and so the diffraction diagram for a semi-infinite breakwater is used. When the reflection coefficient of the front face of the breakwater is less than 1.0 due to wave-absorbing work for example, the diffraction coefficient should be multiplied by the reflection coefficient before being used. For example, if the reflection coefficient of the detached breakwater is 0.4 in the previous example, the wave height ratio at the point A becomes 1 + ( 0.4 ´ 0.68 ) 2 = 1.04 .

[2] Reflection Coefficient Reflection coefficients shall be determined appropriately based on the results of field observations, hydraulic model experiments, and past data. [Technical Notes] (1) Approximate Values for Reflection Coefficient It is desirable to evaluate the value of reflection coefficient by means of field observations. However, when it is difficult to carry out observation or when the structure in question has not yet been constructed, it is standard to estimate reflection coefficient by referring to the results of hydraulic model experiments. In this case, it is desirable to use irregular waves as the test waves. The method by Goda et al. 31) may be used for the analysis of irregular wave test data.

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The following is a list of approximate values for the reflection coefficients of several types of structures. Upright wall:

0.7 ~ 1.0 (0.7 is for the case of a low crown with much overtopping) Submerged upright breakwater: 0.5 ~ 0.7 Rubble mound: 0.3 ~ 0.6 Precast wave-dissipating concrete blocks: 0.3 ~ 0.5 Upright wave-absorbing structure: 0.3 ~ 0.6 Natural beach: 0.05 ~ 0.2 With the exception of the upright wall, the lower limits in the above ranges of reflection coefficient correspond to the case of steep waves and the upper limits to waves with low steepness. It should be noted, however, that with the upright wave-absorbing structure, the reflection coefficient varies with the wavelength, and the shape and dimensions of the structure.

[3] Transformation of Waves at Concave Corners, near the Heads of Breakwaters, and around Detached Breakwaters Around the concave corners of structures, near the heads of breakwaters, and around detached breakwaters, the wave height becomes larger than the normal value of standing waves owing to the effects of diffraction and reflection. This increase in wave height shall be investigated thoroughly. Moreover, the irregularity of waves shall be considered in the analysis. [Technical Notes] (1) Influence of Wave Irregularity When the wave height distribution near a concave corner or the head of a breakwater is calculated for regular waves, a distributional form with large undulations is obtained. However, when wave irregularity is incorporated into the calculation, the undulated form of the distribution becomes smoothed out, excluding the region within one wavelength of a concave corner, and the peak value of the wave height becomes smaller. Calculation using regular waves thus overestimates the increase in the wave height around concave corners and the heads of breakwaters. (2) Graphs for Calculating Wave Height Distribution around a Concave Corner Wave height distributions for irregular waves near a concave corner are shown in Fig. T- 4.5.10. This figure exhibits the form of the distribution of the maximum value of the wave height, as obtained from numerical calculations for each principal wave direction. It has been assumed that waves are completely reflected by the breakwater. In the diagram, Kd is the ratio of the wave height at the front face of the main breakwater to the wave height of the incident waves. The irregular waves used in the calculation has a spectral form with Smax = 75, which implies a narrow directional spreading. The long dash-dot line in each graph shows the distribution of the maximum value of the wave height at each point as obtained using an approximate calculation. The length l1 is that of the main breakwater, l2 is that of the wing breakwater, and b is the angle between the main breakwater and the wing breakwater. This figure may be used to calculate the wave height distribution near a concave corner. When it is not easy to use the calculation program, the approximate calculation method may be used. (3) Wave-Height-Reducing Effects of Wave-Absorbing Work When a wave-absorbing work is installed in order to suppress the increase in wave height around a concave corner and if the wave-absorbing work is such that the reflection coefficient of the breakwater becomes no more than 0.4, it is quite acceptable to ignore the increase in wave height due to the presence of concave corner. However, this is only the case when the wave-absorbing work extends along the whole of the breakwater. If the breakwater is long, one cannot expect the wave-absorbing work to be very effective unless it is installed along the entire length of the breakwater, because the effect of waves reflected from the wing breakwater extend even to places at a considerable distance away from the concave corner. The same can be said for the influence of the main breakwater on the wing breakwater.

-72-


Computer method Approximate solution method

Computer method Approximate solution method Fig. T- 4.5.10 Distribution of the Maximum Value of the Wave Height around Concave Corner 32) -73-

(4) Increase in Wave Height at the Head of a Breakwater Near the head of a semi-infinite breakwater or those of breakwaters at a harbor entrance (specifically within a distance of one wavelength from the head), waves diffracted by breakwaters exercise an effect of local wave height increase over the normal standing wave heights. Because the wave height distribution has an undulating form even at the back face of a breakwater, it is necessary to take heed of the fact that the difference in water level between the inside and the outside of the breakwater gives rise to a large wave force. Figure T- 4.5.11 shows an example of the results of a calculation of the wave force ratio (i.e., the ratio of the wave force to that of a standing wave) near the head of a breakwater.

Wave force ratio


Irregular waves Regular waves

0

Fig. T- 4.5.11 Wave Force Distribution along a Semi-Infinite Breakwater 33)

Wave force ratio

(5) Increase in Wave Height around Detached Breakwater Along a detached breakwater, waves with the height greater than that of normal standing waves are produced, and the wave height distribution takes an undulating form even at the back face of the breakwater. This is due to the effect of wave diffraction at the two ends of the breakwater 34). The wave force also becomes large due to the difference between the water levels in the offshore and onshore sides of the breakwater. In particular, it is necessary to take heed of the fact that, with a detached breakwater, the place where the maximum wave force is generated can shift greatly with the wave direction and the ratio of the breakwater length to the wavelength. Figure T- 4.5.12 shows an example of the results of a calculation of the wave force distribution along a detached breakwater for unidirectional irregular waves. In this calculation, the wave direction for which the largest wave force occurrs is a = 30º (i.e., not when the waves are normally incident to the breakwater, but rather when obliquely incident with a relatively shallow angle).

60°

75°

90°

45° α=30°

α x

x

(m)

Fig. T- 4.5.12 Wave Force Distribution along a Detached Breakwater

4.5.5 Wave Shoaling When waves propagate into shallow waters, shoaling shall be considered in addition to refraction and diffraction. It shall be standard to consider the nonlinearity of waves when calculating the shoaling coefficient. [Commentary] Shoaling is one of the important factors that lead to changing of the wave height in coastal waters. It exemplifies the fact that the wave height in shallow waters is also governed by the water depth and the wave period. Figure T- 4.5.13 has been drawn up based on Shuto’s nonlinear long wave theory. It includes the linearized solution by the small amplitude wave theory and enables the estimation of the shoaling coefficient from deep to shallow waters. In the diagram, Ks is the shoaling coefficient, H0¢ is the equivalent deepwater wave height, H is the wave height at water depth h, and L0 is the wavelength in deepwater.

-74-


0

2% dec ay line

0 0 0

0

0

0

Fig. T- 4.5.13 Graph for Evaluation of Shoaling Coefficient

4.5.6 Wave Breaking At places where the water depth is no more than about three times the equivalent deepwater wave height, changing of the wave height due to wave breaking shall be considered. It shall be standard to consider the irregularity of waves when calculating the change in the wave height due to wave breaking. [Commentary] After the height of waves has increased owing to shoaling, waves break at a certain water depth and the wave height decreases rapidly. This phenomenon is called the wave breaking. It is an important factor to be considered when determining the wave conditions exercising on maritime structures. For regular waves, the place at which waves break is always the same: this is referred to as the “wave breaking point”. For irregular waves, the location of wave breaking depends on the height and period of individual waves, and wave breaking thus occurs over a certain distance: this area is referred to as the “breaker zone”. [Technical Notes] (1) Change in Wave Height Due to Wave Breaking The change in wave height due to wave breaking may be determined using Figs. T- 4.5.14 (a) ~ (e) or Figs. T4.5.15 (a) ~ (e). These figures show the change in wave height for irregular waves as calculated by Goda 35), 44) using a theoretical model of wave breaking. For the region to the right of the dash-dot line on each graph, the change in wave height is calculated using the shoaling coefficient (see 4.5.5 Wave Shoaling). For the region to the left of this dash-dot line, the change in wave height due to wave breaking dominates, and so the wave height must be determined using this graph. As for the bottom slope, it is appropriate to use the mean bottom slope over the region where the water depth to equivalent deepwater wave height ratio h/H0¢ is in the range of 1.5 to 2.5. (2) Scope of Application of Graphs of Wave Height Change At places where the water depth is no more than about one half of the equivalent deepwater wave height, a major portion of wave energy is converted to the energy of oscillating flows rather than to that of water level undulation. Therefore, when calculating the wave force acting on a structure in a very shallow water, it is desirable to use the wave height at the place where the water depth is one half of the equivalent deepwater wave height, if the facilities in question are highly important.

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Bottom slope

Bottom slope 0

0

0

deca y lin

e

0

0

0

deca y lin

e

2%

0

2%

0

00

0

h / H '

Fig. T- 4.5.14 (a) Diagram of Significant Wave Height in the Breaker Zone for Bottom Slope of 1/10

Fig. T- 4.5.14 (b) Diagram of Significant Wave Height in the Breaker Zone for Bottom Slope of 1/20

Bottom slope

Bottom slope

0

0 0

0

ine

0

dec

2%

ay l

dec

ine

ay l

0

2%

0

0

0

0

Fig. T- 4.5.14 (c) Diagram of Significant Wave Height in the Breaker Zone for Bottom Slope of 1/30

Fig. T- 4.5.14 (d) Diagram of Significant Wave Height in the Breaker Zone for Bottom Slope of 1/50

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Bottom slope

Bottom slope

0

0

2% d e

cay l ine

0

0 0

0

2

ine

ay l

ec %d

0

0

0

Fig. T- 4.5.14 (e) Diagram of Significant Wave Height in the Breaker Zone for Bottom Slope of 1/100

Fig. T- 4.5.15 (a) Diagram of Highest Wave Height in the Breaker Zone for Bottom Slope of 1/10

Bottom slope

Bottom slope 0

0

0

0

2% d

2% d

0

ecay

ecay

line

line

0

0

0

Fig. T- 4.5.15 (b) Diagram of Highes Wave Height in the Breaker Zone for Bottom Slope of 1/20

Fig. T- 4.5.15 (c) Diagram of Highest Wave Height in the Breaker Zone for Bottom Slope of 1/30

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Bottom slope

0

Bottom slope

0

0

0

2%

deca

2% d e

0

y lin

cay l

e

ine

0

0

0

Fig. T- 4.5.15 (d) Diagram of Highest Wave Height in the Breaker Zone for Bottom Slope of 1/50

Fig. T- 4.5.15 (e) Diagram of Highest Wave Height in the Breaker Zone for Bottom Slope of 1/100

(3) Approximate Calculation Formulas for Breaking Wave Height Calculation of wave height changes based on a theoretical model for wave breaking generally requires use of a computer. However, considering the variability of the phenomenon and the overall accuracy, it is acceptable to calculate wave height changes using the following simple formula 35), 44):

｛

H1 ¤ 3 =

Ks H0 ¢

: h ¤ L0 ≧ ³ 0.2

min { ( b0 H 0 ¢ + b 1 h ), b max H 0 ¢, K s H 0 ¢ } : h ¤ L 0 < 0.2

(4.5.21)

where

64748

b 0 = 0.028 ( H 0 ¢ ¤ L 0 ) 0.38 exp [ 20 ( tan q ) 1.5 ] b 1 = 0.52 exp [ 4.2 tan q ] b max = max { 0.92, 0.32 ( H 0 ¢ ¤ L 0

(4.5.22)

) 0.29 exp [ 2.4 tan q ] }

The shoaling coefficient Ks is determined using Fig. T- 4.5.13, the operators min{ } and max{ } take the minimum and maximum value of the mulitiple quantities within the braces, respectively, and tanq is the bottom slope. Similarly, an approximate calculation formula for the highest wave height Hmax is given as follows:

｛

H max =

1.8K s H 0 ¢ *

: h ¤ L0≧ ³ 0.2 *

*

min { ( b 0 H 0 ¢ + b 1 h ), b max H 0 ¢, 1.8K s H 0 ¢ } : h ¤ L 0 < 0.2

(4.5.23)

where

64748

b*0 = 0.052 ( H 0 ¢ ¤ L 0 ) 0.38 exp [ 20 ( tan q ) 1.5 ] b*1 = 0.63 exp [ 3.8 tan q ] b*

max

(4.5.24)

= max { 1.65, 0.53 ( H 0 ¢ ¤ L 0 ) 0.29 exp [ 2.4 tan q ] }

(4) Graph for Calculating Breaking Wave Height 35) If the maximum value (H1/3)peak of the significant wave height in the breaker zone is taken as representative of the breaking wave height, then the breaker index curve becomes as shown in Fig. T- 4.5.16. If the water depth (h1/3)peak at which the significant wave height is a maximum is taken as representative of the breaker depth, then the graph for calculating the breaker depth becomes as shown in Fig. T- 4.5.17. -78-


Note: (

H

1/3) peak is the water depth at

which

in the breaker zone

o sl

1/3

m

H

1/3 is a maximum

to

of

1/3) peak is the maximum value

H

in the breaker zone

ot

H

B

Note: (

pe

0

o

B o

tt m o

sl p

0

e 0

0

Fig. T- 4.5.16 Diagram of Maximum Value of the Significant 0

0

Fig. T- 4.5.17 Diagram of Water Depth at which the Maximum Value of the Significant Wave Height Occurs

(5) Breaking Wave Height Criterion for Regular Waves Figure T- 4.5.18 shows the breaking wave height criterion for regular waves. This figure can be used to calculate the breaking wave height criterion in hydraulic model experiments using regular waves. The curves in the graph can be approximated with the following equation: Hb ì ü ph ------ = 0.17 1 exp í 1.5 ------ ( 1 + 15 tan 4 ¤ 3 q ) ý L L0 0 î þ

(4.5.25)

where tanq denotes the bottom slope. Figure T- 4.5.18 shows the limiting wave height at the point of first wave breaking. At places where the water is shallow, the water depth increases owing to the wave setup caused by wave breaking. When estimating the limiting wave height in the breaker zone, it is thus necessary to consider this increase in water level. Bottom

slope 0

0

Fig. T- 4.5.18 Breaking Wave Height Criterion for Regular Waves

(6) Change in Wave Height at Reef Coasts At reef coasts where shallow water and a flat sea bottom continue over a prolonged distance, the change in wave height cannot be calculated directly using Figs. T- 4.5.14 and T- 4.5.15. Instead, the following empirical equation may be used 36): Hx h + h¥ ì x ü -------- = B exp í A æ --------ö ý + a ---------------è H 0 ¢ø H0 ¢ H0 ¢ î þ

(4.5.26)

-79-


where H0¢： equivalent deepwater wave height Hx： significant wave height at a distance x from the tip of the reef h： water depth over the reef h ¥： increase in the mean water level at a place sufficiently distant from the tip of the reef The coefficients A and a are 0.05 and 0.33, respectively, according to the results of hydraulic model experiments. However, it is advisable to use the following values that have been obtained from the data of field observations. H0 ¢ ü A = 0.089 ---------------- + 0.015 ï h + h¥ ï ï (4.5.27) ì ý ï 0.20 ( 4 m > H 0 ¢ ³ 2 m ) ï a = í ï ï ï 0.33 ( H 0 ¢ ³ 4 m ) î þ The coefficient B corresponds to the bottom slope at the front of the reef. Using Fig. T- 4.5.14, it is obtained from the significant wave height Hx = 0 at water depth h as follows. Hx = 0 h + h¥ B = -------------- a ---------------H0 ¢ H0 ¢

(4.5.28)

The term (h+ h ¥ )/H0¢ is given by h + h¥ ---------------- = H0 ¢

3 C 0 ¤ æ 1 + --- ba 2ö è ø 8

(4.5.29)

where b = 0.56. From the continuity of the mean water level at the tip of the reef (x = 0), C0 is given by hx = 0 + h 2 3 Hx = 0 2 C 0 = æ -----------------------ö + --- b æ --------------ö è H0¢ ø 8 è H0 ¢ ø

(4.5.30)

The term h x = 0 represents the rise in the mean water level at water depth h, which is controlled by the bottom slope in front of the reef and wave steepness (see 4.7.1 Wave Setup). The calculation method in the above has been derived under the assumption that the water depth h over the reef is small and waves break over the reef. It is thus not possible to apply the method when the water is deep and wave breaking does not occur. Considering the breaking wave height criterion of a solitary wave, the highest wave height Hmax, x at the distance x from the tip of the reef may be obtained as follows. H max, x = min { 0.78 ( h + h x ), 1.8H x }

(4.5.31)

where min{a, b} is the smaller value of a or b, and h x is the rise in the mean water level at the distance x and is given by the following equation: hx + h --------------- = H0¢

3 Hx 2 C 0 --- b æ --------ö 8 è H 0 ¢ø

(4.5.32)

4.6 Wave Runup, Overtopping, and Transmission 4.6.1 Wave Runup Wave runup shall be calculated appropriately by taking into account the configuration and location of the seawall and the sea bottom topgraphy. [Commentary] The phenomenon of wave runup is dependent upon a whole variety of factors, such as the wave characteristics, the configuration and location of the seawall, and the sea bottom topography; thus the runup height varies in a complex way. There are calculation diagrams and equations based on the results of past researches that may be used, although they are applicable only under certain limited conditions. When the seawall and sea bottom are complex in form, it is advisable to determine wave runup heights by carrying out hydraulic model experiments. When designing seawalls of gently sloping type and the like, it is advisable to set the crown elevation of the seawall to be higher than the runup height for regular waves. However, for irregular waves, depending on the wave height, overflow can occur, and so -80-


ultimately the crown elevation and the form of the seawall are determined so as to make the quantity of overtopping (see 4.6.2 Wave Overtopping) no more than a certain permissible value. [Technical Notes] The following is the description of methods for calculating the runup height over smooth impermeable slopes: (1) Simple Cross Section “A simple cross section” refers to the case in which a seawall (including an upright wall) having a front slope of an uniform gradient a is located at a certain place (of water depth h) on the sea bottom with an almost uniform gradient q. (a) Region of standing waves Takada has proposed the following equation for determining the runup height when the water depth h at the foot of the levee is in the range where standing waves exist (i.e., deeper than the depth at the breaker line). He dealt with two cases separately; i.e., the case where wave breaking does not occur on the front slope and the case where such wave breaking does occur. Firstly, according to Miche’s equation, the minimum angle of inclination of the slope ac for which wave breaking does not occur is found as that satisfying the following condition: 2a c sin 2 a c H0 ¢ --------- ---------------- = -------p p L0

(4.6.1)

Accordingly, when the angle of inclination of the slope is greater than ac, wave breaking does not occur over the slope, in which case the runup height is given by the following equation: hs p R -------- = æè ------- + ------ 1 öø K s : a > a c 2a H0 ¢ H1

(4.6.2)

where H0¢ is the equivalent deepwater wave height, Ks is the shoaling coefficient, H1 is the wave height at the water depth at the foot of the slope, hs is the crest elevation, and R is the runup height. Takada used the following equation for hs /H1,which assumes that there is good agreement between the value from Miche’s standing wave theory and experimental data. H1 1 3 h s ¤ H 1 = 1 + p ------ coth kh × æ 1 + --------------------- -----------------------ö è L 4 sinh 2 kh 4 cosh 2 khø

(4.6.3)

When the angle of inclination of the slope is smaller than ac, wave breaking does occur on the front slope. In this case, it is assumed that the runup height is proportional to tan2/3a, leading to the following equation: cot a c 2 ¤ 3 hs ì p ü : R ¤ H 0 ¢ = í --------- + æ ------ 1ö ý Ks æ --------------ö è cot a ø è H1 ø î 2a c þ

a < ac

(4.6.4)

When the water depth is such that standing waves exist, the runup height can be calculated as above. The maximum runup height occurs when a = ac, with the runup height decreasing both when the slope is more steeply inclined than this and when it is more gently inclined. (b) Region where the water is shallower than the breaker depth Takada has given the runup height for regions where the water is sufficiently shallow for wave breaking to occur as follows: h R ¤ H 0 ¢ = ( R max ¤ H 0 ¢ R 0 ¤ H 0 ¢ ) ----- + R 0 ¤ H 0 ¢ hR

(4.6.5)

where R0 is the runup height on the levee body at the shoreline (h = 0). Based on the experimental results of Toyoshima et al., R0/H0¢ is given as follows: ì 0.18 ( H 0 ¢ ¤ L 0 ) 1 ¤ 2 : Bottom slope 1/10 ï R ¤ H 0 ¢ = í 0.075 ( H 0 ¢ ¤ L 0 ) 1 ¤ 2 : Bottom slope 1/20 ï 0.046 ( H ¢ ¤ L ) 1 ¤ 2 : Bottom slope 1/30 0 0 î

(4.6.6)

The term hR in equation (4.6.5) is the water depth at the foot of the levee for which the runup height becomes largest. It is estimated using Fig. T- 4.6.1, which shows the runup height for a vertical wall. The term LR in the figure is the wavelength at water depth hR, while Rmax is the maximum runup height for the region where the water depth is such that standing waves exist (i.e., the runup height when h = hR).

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(2) Complex Cross Section A “complex cross section” refers to the case where the sea bottom topography and the configuration and location of the seawall (on the whole) are as shown in Fig. T- 4.6.2.

0 0

(a) When the cross section can be considered to be complex, the runup height of the seawall is obtained as follows (refer to Fig. T- 4.6.2) 37). ① The wave breaking point B is determined from the deepwater wave characteristics. ② Next, the runup height R is assumed and the point A is set at the maximum runup point. Then, the points A and B are joined by a straight line, Shoaling coefficient and the gradient of this line yields the virtual gradient cota. ③ The runup height for this virtual gradient is calculated using Fig. T- 4.6.3, and the calculated Fig. T- 4.6.1 Graph for Estimating hR for a Vertical Wall height is compared with the initially assumed runup height. If the two do not agree, then a new runup height is assumed, and the estimation are repeated (i.e., the new runup height is used to give a new virtual gradient and so on). This iterative process is repeated until convergence is achieved. ④ The value so obtained is taken to be the runup height for the complex cross section in question. 0

0

0

(b) When the results obtained from this method are compared with actual experimental results for a complex cross section, it is generally found that there is good agreement between the two, with the error usually being no more than 10%. However, if the bottom slope is too gentle, the agreement between the two becomes poor, and so this method should only be used when the bottom slope is steeper than 1/30. (c) Figure T- 4.6.4 shows experimental results obtained for a bottom slope of 1/70. This figure provides a useful reference when estimating the runup height for a complex cross section with a gentle bottom slope.

Maximum runup point Wave breaking point

Actual cross section Virtual gradient

Fig. T- 4.6.2 Complex Cross Section and Virtual Gradient

Fig. T- 4.6.3 Runup Height on a Slope -82-


Fig. T- 4.6.4 Runup Height on a Seawall Located Closer to the Land than the Wave Breaking Point

(3) Oblique Wave Incidence Figure T- 4.6.5 shows the relationship between the incident angle coefficient Kb and the angle b. Here, b is the angle between the wave crest line of the incident waves and the centerline of the seawall, and the incident angle coefficient Kb is the ratio of the runup height for angle b to the runup height when the waves are normally incident (i.e., when b = 0). This figure can be used to estimate the effect of wave incident angle on the runup height. (4) Effects of Wave-Absorbing Work The runup height can be significantly reduced when the front face of a seawall is completely covered with wavedissipating concrete blocks. Figure T- 4.6.6 shows an example. However, the effect of the concrete blocks varies greatly according to the way in which they are laid, and so in general it is advisable to determine the runup height by means of hydraulic model experiments. (5) Estimation Errors It is important to note that the curves for determining the runup height have been obtained by averaging experimental data that show a large scatter. It should also be noted that actual wave runup will frequently exceeds the design crown height because of wave irregularity when the crown height of a seawall is designed against the significant waves, even if the scatter of the experimental data is not considered; in fact, in extreme cases as many as about a half of the waves may exceed this height. Accordingly, the crown height of a seawall should not be decided based purely on the runup height of regular waves; rather, it is necessary to give consideration to the quantity of overtopping (see 4.6.2 Wave Overtopping).

Smooth surface

Holland

0

0

Surface covered with wave-dissipating concrete blocks

0

(Former) Russia

Fig. T- 4.6.5 Relationship between Wave Incident Angle and Runup Height (Full Lines: Experimental Values by Public Works Research Institute, Ministry of Construction)

0

0

Fig. T- 4.6.6 Reduction in Runup Height Due to Wave-Absorbing Work

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4.6.2 Wave Overtopping For structures for which the quantity of overtopping is an important design factor, the overtopping quantity shall be calculated by carrying out hydraulic model experiments or by using data from hydraulic model experiments carried out in the past. In this case, wave irregularity shall be considered. [Commentary] The “quantity of overtopping” is the total volume of overtopped water. The “rate of overtopping”, on the other hand, is the average volume of water overtopping in a unit time; it is obtained by dividing the quantity of overtopping by the time duration of measurement. The quantity of overtopping and the rate of overtopping are generally expressed per unit width. If the quantity of overtopping is large, then not only there will be damage to the seawall body itself, but also damage by flooding to the roads, houses and/or port and harbor facilities behind the levee or seawall, despite that the levee or seawall is intended to protect them. There is further a risk to users of water frontage amenity facilities that they may be drowned or injured. During design, it is necessary to make the quantity of overtopping no more than a certain permissible value that has been determined in line with the characteristics of structures and the situation with regard to their usage. Furthermore, when estimating the quantity of overtopping by means of experiments, it is desirable to consider changes in tidal water level, i.e., to carry out experiments for different water levels. [Technical Notes] (1) Diagrams for Calculating the Rate of Overtopping 38) For an upright or wave-absorbing seawall that has a simple form (i.e., that does not have anything like a toe protection mound or a crown parapet), the rate of overtopping may be estimated using Figs. T- 4.6.7 ~ 4.6.10. These graphs have been drawn up based on experiments employing irregular waves. From the results of a comparison between the experiments and field observations, it is thought that the accuracy of the curves giving the rate of overtopping is within the range listed in Table T- 4.6.1. The rate of overtopping for the waveabsorbing seawall has been obtained under the conditition that the lower armor layer at the crown consists of 2 rows of wave-dissipating concrete blocks. Table T- 4.6.1 Estimated Range for the Actual Rate of Overtopping Relative to the Estimated Value

q ¤ ( 2 g ( H 0' ) 3 ) 10-2 10-3 10-4 10-5

Upright seawall 0.7 ~ 1.5 times 0.4 ~ 2 0.2 ~ 3 0.1 ~ 5

Wave-absorbing seawall 0.5 ~ 2 times 0.2 ~ 3 0.1 ~ 5 0.05 ~ 10

Note that when obtaining rough estimates for the rate of overtopping for irregular waves using Figs. T- 4.6.7 ~ 4.6.10, the following should be considered: (a) If the actual values of the bottom slope and the deepwater wave steepness do not match any of the values on the graphs, the graph for which the values most closely match should be used, or alternatively interpolation should be carried out. (b) The wave-dissipating concrete blocks in the figures are made up of two layers of tetrapods. If a different kind of wave-dissipating concrete block is used, or if the same kind of wave-dissipating concrete block is used but there are differences in the crown width, in the way in which the tetrapods are laid, or in the form of the toe, then there is a risk that the actual rate of overtopping may considerably differ from the value obtained by the figures. (c) If the number of rows of concrete blocks at the crown is increased, the quantity of overtopping tends to decrease 39). (d) When there are difficulties in applying the graphs for estimating the rate of overtopping, the approximate equation of Takayama et al. 40) may be used.

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Fig. T- 4.6.7 Graphs for Estimating the Rate of Overtopping for an Upright Seawall (Bottom Slope 1/30)

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Fig. T- 4.6.8 Graphs for Estimating the Rate of Overtopping for an Upright Seawall (Bottom Slope 1/10)

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0

0

0

0

0

0

0

Co

nc

ret

eb

loc

k

0

nc

0

0

0

ret

eb

loc

k

0

Co

0

0

0

0

0

nc

ret

eb

0

0

loc

0

0

Co

k

0

0 0

0

Fig. T- 4.6.9 Graphs for Estimating the Rate of Overtopping for a Wave-Absorbing Seawall (Bottom Slope 1/30)

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0

0

0

0

0

0

Co

nc

ret

eb

loc

0

0

k

0

0

0

0

0

0

0

Co

nc

ret

eb

loc

k

0

0

nc

ret

0

eb

0

loc

k

0

Co

0

0

0

Fig. T- 4.6.10 Graphs for Estimating the Rate of Overtopping for a Wave-Absorbing Seawall (Bottom Slope 1/10)

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(2) Permissible Rate of Overtopping The permissible rate of overtopping depends on factors such as the structural type of the seawall, the situation with regard to land usage behind the seawall, and the capacity of drainage facilities; it must be set appropriately in line with the individual situation. Although it is thus impossible to give one standard value for the permissible rate of overtopping, Goda 41) nevertheless gave the values for the damage limit rate of overtopping as listed in Table T- 4.6.2 based on past cases of disasters. Furthermore, Nagai et al. have considered the degree of importance of the facilities behind the seawall and have come up with the values for the permissible rate of overtopping as listed in Table T- 4.6.3, using the results of experiments with regular waves. Table T- 4.6.2 Damage Limit Rate of Overtopping Type

Rate of overtopping (m3/m•s)

Covering

Revetment

Apron paved Apron unpaved

0.2 0.05

Levee

Concrete on front slope, crown, and back slope Concrete on front slope and crown, but no concrete on back slope Concrete on front slope only

0.05 0.02 0.005 or less

Table T- 4.6.3 Permissible Rate of Overtopping (m3/m•s) as a Function of the Degree of Importance of the Hinterland Areas where there is a high concentration of houses, public facilities etc. behind the seawall, and so it is anticipated that flooding due to overtopping or spray would cause particularly serious damage

About 0.01

Other important areas

About 0.02

Other areas

0.02 ~ 0.06

(3) Equivalent Crown Height Coefficient The equivalent crown height coefficient can be used as a guideline when setting the quantity of overtopping for a seawall upon which wave-dissipating concrete blocks are laid or for a seawall of wave-absorbing type with vertical slits. The equivalent crown height coefficient is the ratio of the height of the seawall in question to the height of an imaginary upright seawall that results in the same quantity of overtopping, where the conditions in terms of waves and the sea bottom topography are taken to be the same for the both cases. If the equivalent crown height coefficient is less than 1.0, this means that the crown of the seawall under study can be lowered below that of an upright seawall and still give the same quantity of overtopping; in other words, the seawall under study has a form that is effective in reducing the quantity of overtopping. Below are the reference values for the equivalent crown height coefficient b for typical types of seawall. Wave-absorbing seawall with concrete block mound 40): b = 0.9 ~ 0.7 Vertical-slit type seawall 40) Parapet retreating type Stepped

: b = 0.6

seawall 39)

: b = 1.0 ~ 0.5

seawall 39)

When the waves are obliquely incident 42)

: b = 1.7 ~ 1.0 ì 1 sin 2 q : q ≦ £ 30° ü :b = í ý î 1 sin 2 i 30° : q > 30° þ (q is the angle of incidence of the waves; it is 0º when the waves are normally incident on the seawall)

(4) Effect of Winds on the Quantity of Overtopping In general, winds have a relatively large effect on the quantity of overtopping when it is small, although there is a lot of variation. However, the relative effect of winds decreases as the quantity of overtopping increases. Figure T- 4.6.11 shows the results of an investigation on the wind effect on the quantity of overtopping based on field observations. The abscissa shows the spatial gradient of the quantity of overtopping, while the ordinate shows the quantity of overtopping per unit area. As can be seen from the figure, when the quantity of overtopping is small, the larger the wind velocity, the smaller the spatial gradient of the quantity of overtopping becomes. When the quantity of overtopping is large, the spatial gradient of the quantity of overtopping increases. This shows that, when the quantity of overtopping is small, the distance over which a mass of water splash strongly affected by the wind velocity, with a larger distance at a higher wind velocity; however, when the quantity of overtopping is large, the difference in the splash distance becomes small. (5) Overtopping of Multidirectional Random Waves In waters where the multidirectionality of waves is well clarified, the rate of overtopping may be corrected in line with Smax as in reference 42).

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W in

d

ve

lo

cit

y

Quantity of overtopping per unit area q


Gradient

Fig. T- 4.6.11 Wind Effect on Spatial Gradient of Wave Overtopping Quantity

4.6.3 Wave Transmission It shall be standard to calculate the height of waves transmitted behind a breakwater by overtopping and/or permeation through the core or the fundation mound of breakwater by referring to either the results of hydraulic model experiments or the past data. [Commentary] It is necessary to appropriately estimate the transmitted wave height after waves have overtopped and/or passed through a breakwater, because the transmitted waves affect the wave height distribution behind the breakwater. Transmitted waves include waves that have overtopped and/or overflowed, as well as waves that have permeated through a rubble mound breakwater or a foundation mound of composite breakwater. Recently, several breakwaters have been built with caissons (which are originally not permeable) having through-holes in order to enhance the exchange of the seawater within a harbor. In this case, it is necessary to examine the value of wave transmission coefficient, because the coefficient serves as an indicator of the efficiency of the exchange of seawater. [Technical Notes] (1) Transmission Coefficient for a Composite Breakwater Figure T- 4.6.12 may be used to calculate the height of waves that are transmitted into a harbor when they overtop a composite breakwater or permeate through a foundation mound. Even when the waves are irregular, the transmission coefficient agrees pretty well with that shown in Fig. T- 4.6.12. It has also been shown that Fig. T- 4.6.12 is valid not only for the significant wave height, but also for the highest one-tenth wave height and the mean wave height.

Fig. T- 4.6.12 Graph for Calculating the Wave Height Transmission Coefficient -90-


(2) Period of Transmitted Waves for a Composite Breakwater The period of the transmitted waves drops to about 50 ~ 80% of the corresponding incident wave period (this is true both for the significant wave period and the mean period). (3) Experimental Data on Other Types of Breakwater For composite breakwaters covered with wave-dissipating concrete blocks, rubble mound breakwaters armored with wave-dissipating concrete blocks, and other such breakwaters, experiments on the transmitted wave height have been carried out by the Civil Engineering Research Institute of Hokkaido Development Bureau. (4) Transmission Coefficient for Structures Other than Composite Breakwaters (a) For a porous, water-permeable structure such as a rubble mound breakwater or a wave-dissipating concrete block type breakwater, Kondo’s theoretical analysis may be referred to. The following empirical equation may be used to obtain the transmission coefficient of a typical structure. Stone breakwater: K T = 1 ¤ ( 1 + k t H ¤ L )

(4.6.7)

where kt=1.26 (B/d)0.67, B is the crown width of the structure, d is the depth from the water surface to the ground surface of the structure, H is the height of incident waves, and L is the wavelength of transmitted waves. (b) For a curtain wall breakwater, the empirial solutions of Morihira et al. 43) may be used (see Part VII, 3.3.1 Curtain Wall Breakwater). (c) For the transmission coefficient of an upright breakwater of permeable type that has slits in both the front and rear walls, the experimental results are available. (d) Types of breakwater aiming to promote the exchange of seawater include multiple-wing type permeable breakwaters, multiple-vertical cylinder breakwaters, horizontal-plate type permeable breakwaters, and pipe type breakwaters. The transmission coefficients of these types of breakwater have been obtained by hydraulic model tests. (5) Transmission Coefficient for a Submerged Breakwater A submerged breakwater is usually made by piling up natural stones or crushed rock to form a mound, and then covering the surface with concrete blocks to protect underlayers. For a submerged breakwater of crushed rock, a graph showing the relationship between the crown height of the breakwater and the transmission coefficient is available.

4.7 Wave Setup and Surf Beat 4.7.1 Wave Setup When designing structures that will be placed within the breaker zone, it is desirable to consider the phenomenon of wave setup as necessary, which occurs in the breaker zone owing to wave breaking as they approach the coast. [Technical Notes] (1) Diagrams for Estimating the Amount of Wave Setup The changes in the mean water level by random wave breaking on the bottom slopes of 1/100 and 1/10 as calculated by Goda 35), 44) are shown in Figs. T- 4.7.1 and T- 4.7.2. The smaller the wave steepness (H0¢/L0, where H0¢ is the equivalent deepwater wave height and L0 is the wavelength in deepwater ), the larger the rise of mean water level becomes. Moreover, the steeper the bottom slope, the larger the rise of mean water level. Figure T- 4.7.3 shows the rise of mean water level at the shoreline. The effects of wave steepness and bottom slope on the rise of mean water level are clearly shown. When H0¢/L0 is in the range 0.01 ~ 0.05, with the exception of very steep bottom slope, the rise of mean water level near the shoreline is of the order (0.1 ~ 0.15)H0¢. (2) Consideration of the Rise in Mean Water Level in Design A rise of mean water level causes the wave breaking point to shift shoreward and the breaking wave height to increase. The rise of in mean water level is thus important for the accurate calculation of the design wave height in shallow waters.

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0

0 0

Change in mean water level

Change in mean water level η /H 0 0′

η/H ′ 0 0

0

h /H0

′

h /H0 0

Fig. T- 4.7.1 Change in Mean Water Level (Bottom Slope 1/10)

′

Fig. T- 4.7.2 Change in Mean Water Level (Bottom Slope 1/100)

0

Risc in mean water level

0

z rms --------------------(h ) rms 0

Near shoreline Offshore

Oarai Niigata Miyazaki

Wave steepness

0

0

Fig. T- 4.7.3 Rise in Mean Water Level at the Shoreline

1¤2 H ¢ 0 h --------- æ 1 + ---------ö ø L è ¢ H 0 0 0

0

0

Fig. T- 4.7.4 Ratio of Surf Beat Amplitude to Deepwater Wave Amplitude

4.7.2 Surf Beat In shallow waters, surf beat with the period of one to several minutes shall be investigated as necessary. [Technical Notes] Goda’s Formula for Estimating Surf Beat Amplitude Based on the results of field observations of surf beat, Goda has proposed the following relationship 35), 44): 0.04 ( h rms ) 0 0.01H 0 ¢ z rms = ------------------------------------- = ------------------------------------H0¢ H0 ¢ h h -------- æ 1 + --------ö -------- æ 1 + --------ö L0 è L0 è H 0 ¢ø H 0 ¢ø

(4.7.1)

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where zrms is the root mean square amplitude of the surf beat wave profile, (hrms)0 is the root mean square amplitude of the deepwater wave profile, H0¢ is the equivalent deepwater wave height, L0 is the wavelength in deepwater, and h is the water depth. This equation shows that the amplitude of the surf beat is proportional to the deepwater wave height, that it falls as the water depth increases, and that it increases as the deepwater wave steepness H0¢/L0 decreases. Figure T- 4.7.5 shows a comparison between the estimation by equation (4.7.1) and actual observation values.

4.8 Long-Period Waves and Seiche With regard to long-period waves and seiche in harbors, field observations shall be carried out as far as possible, and appropriate measures to control them shall be taken based on the results of these observations. [Commentary] Water level fluctuations with the period between one and several minutes sometimes appear at observation points in harbors and off the shore. Such fluctuations are called long-period waves. If the period of such long-period waves is close to the natural period of oscillation of the vibration system made up of a vessel and its mooring ropes, the phenomenon of resonance can give rise to a large surge motion even if the wave height is small, resulting in large effects on the cargo handling efficiency of the port. If it is clear from observations that long-period waves of significant wave height 10 ~ 15 cm or more frequently arise in a harbor, it is advisable to investigate either hard or soft countermeasures. When conspicuous water level fluctuations with the period several minutes or longer occur at an observation point in a harbor, it can be assumed that the phenomenon of “seiche” is taking place. This phenomenon occurs when small disturbances in water level generated by changes in air pressure out at sea are amplified by the resonant oscillations of the harbor or bay. If the amplitude of such seiche becomes significantly large, inundation at the head of the bay or reverse outflow from municipal drainage channels may occur. Also high current velocities may occur locally in a harbor, resulting in breaking of the mooring ropes of small vessels. When drawing up a harbor plan, it is thus desirable to give consideration to making the shape of the harbor to minimize the seiche motion as much as possible. [Technical Notes] (1) Threshold Height of Long-period Waves for Cargo Handling Works It is necessary to give due consideration to the fact that long-period waves in front of a quaywall can induce ship surging with the amplitude of several meters through resonance. The threshold height of long-period waves for smooth cargo handling works depends on the factors such as the period of the longperiod waves, the dimensions of the vessel in question, the mooring situation, and the loading conditions. Nevertheless, according to field observations carried out in places like Tomakomai Bay 46), it corresponds to a significant wave height of about 10 ~ 15 cm. (2) Calculating the Propagation of Long-period Waves It is desirable to calculate the propagation of longperiod waves into a harbor by setting up incident wave boundary out at sea and then using either the Boussinesq equation 29) or a calculation method that uses long linear wave equations 47).

Observed spectrum Approximate form of standard spectrum

Fig. T- 4.8.1 Comparison between Standard Spectrum

(3) Standard Spectrum for Long-period Waves with Long-period Components and When there has been insufficient field observation Observed Spectrum data of long-period waves out at sea and the longperiod wave conditions that determine the design external forces have not been established, the standard spectrum shown in reference 48) or its approximate expression may be used. Figure T- 4.8.1 shows a comparison between an observed spectrum and an approximate form of the standard spectrum. The term al in the figure is a parameter that represents the energy level of the long-period waves. (4) Calculation Method for Seiche See 6.5 Seiche for a calculation method for seiche.

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4.9 Waves inside Harbors 4.9.1 Calmness and Disturbances When evaluating the harbor calmness, the factors that give rise to disturbances in the harbor shall be set appropriately. [Commentary] The problem of harbor calmness is extremely complex. It involves not only physical factors such as waves, winds, vessel motions, and the wind- and wave-resistance of working machinery, but also the factors requiring human judgement: the latter include the easiness of vessels entering and leaving a harbor, vessel refuge during stormy weather, and the threshold conditions of works at sea. The harbor calmness is further related with the economic factors, such as the efficiency of cargo handling works, the operating rate of vessels, and the cost of constructing the various facilities required to improve the harbor calmness. The factors that lead to wave disturbances in harbors, which form the basis of the criteria for determining the harbor calmness, include the following: (a) (b) (c) (d) (e)

Waves penetrating through the harbor entrance Transmission of waves into the harbor over breakwaters Reflected waves Long-period waves Seiche

In large harbors, wind waves generated within the harbor may require attention, and the ship-generated waves by larger vessels may cause troubles for small vessels.

4.9.2 Evaluation of Harbor Calmness The harbor calmness shall be evaluated by considering individual wave components estimated separately for respective factors that cause disturbances in the harbor. [Technical Notes] The following method may be used for evaluating the harbor calmness: (1) To estimate the waves in the harbor, first establish the joint distribution of the height and direction of deepwater waves. (2) Next, calculate the wave transformations by refraction and breaking that takes place between the deepwater wave observation and/or hindcasting point and the harbor entrance, using say the energy balance equation method, and thus obtain the wave conditions at the harbor entrance. (3) Obtain the wave height in the harbor, focusing mainly on diffraction and reflection. If necessary, carry out an investigation on wave transmission at this time. (4) The wave height in the harbor can be estimated by taking the squares of each of the diffracted wave height, the reflected wave height and the transmitted wave height, adding the results, and then taking the square root. For harbors where the effects of transmitted waves are relatively slight, the wave period in the harbor may be taken to be the same as the period of the diffracted waves. Note that the wave height in the harbor should be investigated for each wave direction for various classes of wave heights with the occurrence probability outside the harbor. (5) It is standard to express the occurrence rate of waves in a harbor as the percentage of the waves exceeding 0.5 m or 1.0 m in height or in terms of the number of days. However, depending on the usage purpose, it is also acceptable to take into consideration the exceedance probability for other wave height classes. The harbor calmness is obtained by subtracting from 100% the occurrence probability (in percentage) that the wave height in the harbor exceeds the threshold level for cargo handling works at the berth in question. It is not possible to determine a value for the threshold wave height for cargo handling works that is valid universally; rather, it depends on the purpose for which the wharf facilities are used, the dimensions of vessels, and the period and direction of waves, etc. Nevertheless, a value of 0.5 to 1.0 m (significant wave height) may be used as a reference value. However, it should be noted that the critical wave height for cargo handling is lower for waves of long-periodicity such as swell 49), and so care is required when evaluating the net working rate for ports and harbors that face out onto the open sea.

4.10 Ship Waves In canals and navigation channels, it is desirable to examine the influence of waves generated by moving vessels.

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[Technical Notes] (1) Pattern of Ship Waves as Viewed from Top If ship waves are viewed from top, it appears as shown in Fig. T- 4.10.1. Specifically, it is composed of two groups of waves. One group of waves spread out in a shape like 八 (the Chinese character for 8) from a point slightly ahead of the bow of the vessel. The other group of waves are behind the vessel and are such that the wave crest is perpendicular to the vessel’s sailing line. The former waves are termed the “divergent waves”, while the latter are termed the “transverse waves”. The divergent waves form concave curves; the closer to the sailing line, the smaller the gap between waves. On the other hand, the transverse waves are approximately arcshaped, with the gap between waves being constant (i.e., independent of the distance from the sailing line). In deep water, the area over which the ship waves extend is limited within the area bounded by the two cusplines with the angles ± 19º28' from the sailing line and starting from the origin (i.e., the point from which the cusp lines diverge) lying somewhat in front of the bow of the vessel. The divergent waves cross the transverse waves just inside the cusplines; this is where the wave height is largest. The wave steepness is smaller for the transverse waves than for the divergent waves, implying that the transverse waves often cannot be discerned from an aerial photograph.

Vessel's sailing line

Fig. T- 4.10.1 Plan View of Ship Waves

(2) Wavelength and Period of Ship Waves The wavelength and period of ship waves differ for the divergent waves and the transverse waves, with the latter having both a longer wavelength and a longer period. Amongst the divergent waves, the wavelength and period are both longest for the first wave and then become progressively shorter. (a) The wavelength of the transverse waves can be obtained by the numerical solution of the following equation, which is derived from the condition that the celerity of the transverse waves must be the same as the velocity at which the vessel is sailing forward. gL t 2ph ------- tanh --------- = V 2 : (provided V < g h ) 2p Lt

(4.10.1)

where Lt： wavelength of transverse waves (m) h： water depth (m) V： sailing speed of vessel (m/s) Note however that when the water is sufficiently deep, the wavelength of the transverse waves is given by the following equation: 2p L 0 = ------V 2 = 0.169V k2 (4.10.2) g where L0： wavelength of transverse waves at places where the water is sufficiently deep (m) Vk： sailing speed of vessel (kt); Vk = 1.946V (m/s) (b) The period of the transverse waves is equal to the period of progresseive waves with the wavelength Lt (or L0) in water of depth h. It is given by equation (4.10.3) or (4.10.4). Tt =

2p 2ph 2ph ------L t coth æ ----------ö = T 0 coth æ ----------ö è Lt ø è Lt ø g

(4.10.3)

2p T 0 = ------V = 0.330V k g

(4.10.4) -95-


where Tt： period of transverse waves in water of depth h (s) T0： period of transverse waves at places where the water is sufficiently deep (s) (c) The wavelength and period of the divergent waves are given by equations (4.10.5) and (4.10.6), which are derived from the condition that the component of the vessel’s speed in the direction of travel of the divergent waves must be equal to the velocity of the divergent waves. L d = L t cos 2 q

(4.10.5)

T d = T t cos q

(4.10.6)

where Ld： wavelength of divergent waves as measured in the direction of travel (m) Td： period of divergent waves (s) q： angle between the direction of travel of the divergent waves and the sailing line (º)

0

Ratio of the period of the divergent waves to that of the transverse waves Td / T0

Angle between the direction of travel of the divergent waves and the sailing line q

According to Kelvin’s theory of wave generation at places where the water is sufficiently deep, the angle of travel q of the divergent waves can be obtained as shown in Fig. T- 4.10.2, as a function of the position of the place under study relative to the vessel. Note however that for actual vessels the minimum value of q is generally about 40º, and q is usually about 50º ~ 55º for the point on a particular divergent wave at which the wave height is a maximum. Note also that, as shown in the illustration in the figure, the angle q directs the location of the source point Q from where the divergent wave has been generated; a is the angle between the cusp line and the sailing line.

0

Relative position of observation point

x / s

Fig. T- 4.10.2 Wave Direction and Period of

at Places Where the Water is Sufficiently Deep

(3) Shoaling Effect on Ship Waves As common with water waves in general, ship waves are affected by the water depth, and their properties vary when the water depth decreases relative to the wavelength of ship waves. This shoaling effect on ship waves may be ignored if the following condition is satisfied: V ≦ 0.7 gh

(4.10.7)

The critical water depth above which ship waves may be regarded as deepwater waves is calculated by equation (4.10.7), as listed in Table T-4.10.1. As can be seen from this table, the waves generated by vessels in normal conditions can generally be regarded as deepwater waves. Situations in which they must be regarded as shallow water waves include the following cases: a high-speed ferry travels through relatively shallow waters, a motorboat travels through shallow waters, and ship waves propagate into shallow waters. Note that ship waves in shallow water have a longer wavelength and period than those generated by the vessel sailing in deep water at the same speed.

-96-


Table T-4.10.1 Conditions under Which Ship Waves Can Be Regarded as Deepwater Waves Speed of vessel Vk (kt) Water depth h (m) ≧ Period of transverse waves T0 (s)

5.0 1.4 1.7

7.5 3.1 2.5

10.0 5.5 3.3

12.5 8.6 4.1

15.0 12.4 5.0

17.5 16.9 5.8

20.0 22.0 6.6

25.0 34.4 8.3

30.0 49.6 9.9

(4) Height of Ship Waves The Ship Wave Research Committee of the Japan Association for Preventing Maritime Accidents has proposed the following equation for giving a rough estimate of the height of ship waves: Ls 1 ¤ 3 E HPW -----------------------H 0 = æ ---------ö è 100ø 1620L s V K

(4.10.8)

where H0： characteristic wave height of ship waves (m), or the maximum wave height observed at a distance of 100 m from the sailing line when a vessel is sailing at its full-load cruising speed Ls： length of the vessel (m) V K： full-load cruising speed (kt) EHPW： wave-making power (W) The wave-making power EHPW is calculated as follows. E HPW = E HP E HPF

(4.10.9)

E HP = 0.6S HPm

(4.10.10)

1 E HPF = --- rSV 03 C F 2

(4.10.11)

S ≒ 2.5 ÑL s

(4.10.12)

V0 Ls C F = 0.075 ¤ æ log ---------- 2ö è ø n where SHPm： r0： V0： CF： n： Ñ：

2

(4.10.13)

continuous maximum shaft power (W) density of seawater (kg/m3); r0 = 1030 (kg/m3) full-load cruising speed (m/s); V0 = 0.514VK frictional resistance coefficient coefficient of kinematic viscosity of water (m2/s); n ≒ 1.2 × 10-6 (m2/s) full-load displacement of vessel (m3)

Equation (4.10.8) has been obtained by assuming that the energy consumed through wave making resistance is equal to the propagation energy of ship waves, while the values of the coefficients have been determined as averages from the data from ship towing tank tests. The characteristic wave height H0 varies from vessel to vessel, although for medium-sized and large vessels it is about 1.0 ~ 2.0 m. Tugboats sailing at full speed produce relatively large waves. It is considered that the wave height decays as s-1/3, where s is the distance of the observation point from the sailing line. It is also considered that the wave height is proportional to the cube of the cruising speed of the vessel. Accordingly: 100 H max = H 0 æ ---------ö è s ø

1¤3

V 3 æ ------k ö è V Kø

(4.10.14)

where Hmax：maximum height of ship waves at any chosen observation point (m) s：distance from the observation point to the sailing line (m) Vk：actual cruising speed of the vessel (kt) Equation (4.10.14) cannot be applied if s is too small; specifically, the approximate minimum value of s for which equation (4.10.14) can be applied is either the vessel length Ls or 100 m, whichever is the smaller. The upper limit of the height of ship waves occurs when the breaking criterion is satisfied; this criterion is expresed as the steepness Hmax/Lt of the highest divergent wave being equal to 0.14. If the angle between the wave direction and the sailing line is assumed to be q = 50º at the point on a divergent wave where the wave height becomes largest, the upper limit of the wave height at any given point is given by equation (4.10.15). This also assumes, however, that the conditions for deepwater waves are satisfied. H limit = 0.010V k2

(4.10.15)

where -97-


Hlimit：upper limit of the height of ship waves as determined by the wave breaking conditions (m) (5) Propagation of Ship Waves (a) Among two groups of ship waves, the transverse waves propagate in the direction of vessel’s sailing line, and continue to propagate even if the vessel changes course or stops. In this case, the waves have a typical nature of regular waves (with the period being given by equation (4.10.3), and they propagate at the group velocity, undergoing transformation such as refraction and others. Takeuchi and Nanasawa gave an example of such transformations. Note however that as the waves propagate, the length of wave crest spreads out (the wave crest gets longer), and even when the water is of uniform depth, the wave height decays in a manner inversely proportional to the square root of the distance traveled. (b) The direction of propagation of a divergent wave varies from point to point on the wave crest. According to Kelvin’s theory of wave generation, the angle between the direction of propagation and the sailing line is q = 35.3º at the outer edge of a divergent wave. As one moves inwards along the wave crest, the value of q approaches 90º. The first (c) arriving at a any particular point has the angle q = 35.3º, while q getting gradually larger for subsequent waves. This spatial change in the direction of propagation of the divergent waves can be estimated using Fig. T- 4.10.2. (d) The propagation velocity of a divergent wave at any point on the wave crest is the group velocity corresponding to the period Td at that point (see equation (4.10.6)). In the illustration in Fig. T- 4.10.2, the time needed for a component wave to propagate at the group velocity from the point Q at wave source to the point P is equal to the time taken for the vessel to travel at the speed V from the point Q to the point O. Since each wave profile propagates at the wave velocity (phase velocity), the waves appear to pass beyond the cuspline and vanish one after the other at the outer edge of the divergent waves. (6) Generation of Solitary Waves. When a vessel sails through shallow waters, solitary waves are generated in front of the vessel if the cruising speed Vk (m/s) approaches gh . Around the mouths of rivers, there is a possibility of small vessels being affected by such solitary waves generated by other large vessels 50). [References] 1) Dean, G. R.: “Stream function wave theory and application”, Handbook of Coastal and Ocean Engineering, Volume 1, Gulf Pub., 1991, pp. 63-94. 2) Dean G. R. and R. A. Dalrymple: “Water Wave Mechanics for Engineers and Scientists”, World Scientific, 1991, pp. 305-309 3) Goda, Y.: “Wave forces on a vertical circular cylinder: Experiments and proposed method of wave force computation”, Rept. of PHRI, No. 8, 1964, 74 p. 4) Yoshimi GODA, Yasumasa SUZUKI: “Computation of refraction and diffraction of sea waves with Mitsuyasu’s directional spectrum”, Tech. Note of PHRI, No. 230, 1975 (in Japanese). 5) Pierson, W. J. Jr., G. Neumann and R. W. James: “Practical methods for observing and forecasting ocean waves by means of wave spectra and statistics”, U. S. Navy Hydrographic Office, Pub. No. 603, 1955. 6) Inoue, T.: “On the growth of the spectra of a wind gererated sea according to a modified Miles-Phillips mechanism and its application to wave forecasting”, Geophysical Science Lab., TR-67-5, New York Univ., 1967, pp. 1-74. 7) Isozaki, I. and T. Uji: “Numerical prediction of ocean wind waves”, Papers of Meteorology and Geophysics, Vol. 24 No. 2, 1973, pp. 207-231. 8) Joseph, P. S., S. Kawai and Y. Toba: “Ocean wave prediction by a hybrid model combination of single-parameterized wind waves with spectrally treated swells”, Sci. Rept. Tohoku Univ., Ser. 5, (Tohoku Geophys. Jour.), Vol. 28, No. 1, 1981. 9) Uji, T.: “A coupled discrete wave model MRI-II”, Jour. Oceanogr, Society of Japan, Vol. 40, 1985, pp. 303-313. 10) Günther, H. et al: “A hybrid parametrical wave prediction model”, Jour. Geophys. Res., Vol. 84, 1979, pp. 5727-5738. 11) Takeshi SOEJIMA, Tomoharu TAKAHASHI: “A comparison on wave hidcasting methods”, Tech. Note of PHRI, No. 210, 1975, pp. 1-24 (in Japanese). 12) Collins, J. I.: “Prediction of shallow water wares”, Jour. Geopyhs. Res., Vol. 77, No. 15, 1972, pp. 2693-2702. 13) Cavaleri, L. and P. H. Rizzoli: “Wind wave prediction in shallow water: theory and applications”, Jour. Geophys. Res., Vol. 86 No. C11, 1981, pp. 10961-10973. 14) Golding. B: “A wave perdiction system for realtime sea state forecasting”, Quat. Jour. Royal Meleorol. Soc., Vol. 109, 1983, pp. 393-416. 15) Phillips, O. M.: “On the generation of waves by turbulent wind”, J. F. M., Vol. 2, 1957, pp. 417-445. 16) Miles, J. W.: “On the generation of surface waves by shear flows”, J. F. M., Vol. 6, 1959, pp. 568-582. 17) Hasselmann, K.: “Weak-interaction of ocean waves”, Basic Developments in Fluid Dynamics, Vol. 2, Academic Press Inc., New York., 1968. 18) Hasselmann, S and K. Hasselamann: “Computations and parameterizations of the nonlinear energy transfer in a gravity wave spectrum, Part I: A new method for efficient computations of the exact nonlinear transfer integral”, J. Phys. Oceanogr., Vol. 15, 1985, pp. 1369-1377. 19) Sverdrup, H. U. and W. H. Munk: “Wind Sea and Swell, Theory of Relations for Forecasting”, U. S. Hydrographic Office, Pub. No. 601, 1947. 20) Bretschneider, C. L.: “The generation and decay of waves in deep water”, Trans. A. G. U., Vol. 37, No. 3, 1952. 21) Wilson, B. W.: “Numerical prediction of ocean waves in the North Atlantic for December 1959”, Deut. Hydro. Zeit, Jahrg. 18, Ht. 3. 1965. -98-


22) Saville. T.: “The effect of fetch width on wave generation”, Tech. Memo., B. E. B., No. 70. 23) Wilson, B. W.: “Graphical approach to the forecasting of waves in moving fetches”, Tech, Memo., B. E. B., No. 73, 1955. 24) Bretschneider. C. L.: “Decay of ocean waves: Fundamentals of ocean engineering - Part 8b”, Ocean Industry, 1968, pp. 4550. 25) Gringorten, I. I.: “A plotting rule for extreme probability paper”, J. Geophysical Res., Vol. 68 No. 3, 1963, pp. 813-814. 26) Petruaskas, C. and P. M. Aagaard: “Extrapolation of historical storm data for estimating design wave heights”, Preprints 2nd OTC, No. 1190, 1970, pp. I-409-428. 27) Yoshiyuki ITO, Katsutoshi TANIMOTO, Shoichi YAMAMOTO: “Wave height distribution in the region of ray crossing application of the numerical analysis method of wave propagation -”, Rept of PHRI, Vol. 11, No. 3, 1972, pp. 87-110 (in Japanese). 28) Tomotsuka TAKAYAMA, Naota IKEDA, Tetsuya HIRAISHI: “Practical computation method of directional random wave transformation”, Rept. of PHRI, Vol. 30, No. 1, 1991, pp. 21-67 (in Japanese). 29) Tetsuya HIRAISHI, Isao UEHARA, Yasumasa SUZUKI: “Applicability of wave transformation model in boussinesq equation”, Tech. Note of PHRI, No. 814, 1995, 22 p. (in Japanese). 30) Katsutoshi TANIMOTO, Katsutoshi KIMURA, Antonio Paulo dos Santos Pinto: “Random wave forces and design wave periods of composite breakwaters under the action of double peaked spectral waves”, Rept. of PHRI, Vol. 25, No. 2, 1986, pp. 3-25 (in Japanese). 31) Yoshimi GODA, Yasumasa SUZUKI, Yasuharu KISHIRA, Osamu KIKUCHI: “Estimation of incident and reflected waves in random wave experimen”, Tech. Note of PHRI, No. 248, 1976, 24 p. (in Japanese). 32) Koji KOBUNE, Mutsuo OSATO: “A study of wave height distribution along a breakwater with a corner”, Rept. of PHRI, Vol. 15, No. 2, 1976 (in Japanese). 33) Yoshiyuki ITO, Katsutoshi TANIMOTO: “Meandering damages of composite type breakwaters”, Tech. Note of PHRI, No. 112, 1971 (in Japanese). 34) Yoshimi GODA, Tomotsuka YOSHIMURA: “Wave force computation for structures of large diameter, isolated in the offshore”, Rept. of PHRI, Vol. 10, No. 4, 1971 (in Japanese). 35) Yoshimi GODA: “Deformation of irregular waves due to depth-controlled wave breaking” Rept. of PHRI, Vol. 14, No. 3, 1975 (in Japanese). 36) Tomotsuka TAKAYAMA, Yutaka KAMIYAMA, Osamu KIKUCHI: “Wave transformation on a reef ”, Tech. Note of PHRI, No. 278, 1977, 32 p. (in Japanese). 37) Saville, T. Jr.: “Wave run-up on composite slopes”, Proc. 6th Conf. on Coastal Eng., 1958, pp. 691-699. 38) Yoshimi GODA, Yasuharu KISHIRA, Yutaka KAMIYAMA: “Laboratory investigation on the overtopping rate of seawalls by irregular waves”, Rept. of PHRI, Vol. 14, No. 4, 1975, pp. 3-44 (in Japanese). 39) Yoshimi GODA, Yasuharu KISHIRA: “Experiments on irregular wave overtopping characteristics of low crest types”, Tech. Note of PHRI, No. 242, 1976, 28 p. (in Japanese). 40) Tomotsuka TAKAYAMA, Toshihiko NAGAI, Kazuhiko NISHIDA: “Decrease of wave overtopping amount due to seawalls of low crest types”, Rept. of PHRI, Vol. 21, No. 2, 1982, pp. 151-205 (in Japanese). 41) Yoshimi GODA: “Estimation of the rate of irregular wave overtopping of seawalls”, Rept. of PHRI, Vol. 9, No. 4, 1970, pp. 3-41 (in Japanese). 42) Tetsuya HIRAISHI, Norio MOCHIZUKI, Kazuo SATO, Haruhiro MARUYAMA, Tsuyoshi KANAZAWA, Tatsuya MASUMOTO: “Effect of wave directionality on overtopping at seawall”, Rept. of PHRI, Vol. 35, No. 1, 1996, pp. 39-64 (in Japanese). 43) Michio MORIHIRA, Shusaku KAKIZAKI, Yoshimi GODA: “Experimental investigation of a curtain-wall breakwater”, Rept. of PHRI, Vol. 3, No. 1, 1964, pp. 1-27 (in Japanese). 44) Yoshimi Goda: “Irregular wave deformation in the surf zone”, Constal Engineering in Japan, JSCE, Vol. 18, 1975, pp. 13-26. 45) Kazumasa KATOH, Satoshi NAKAMURA, Naota IKEDA: “Estimation of infragravity waves in consideration of wave groups - An examination on basis of field observation at HORF -”, Rep. of PHRI, Vol. 30, No. 1, 1991, pp. 137-163 (in Japanese). 46) Tetsuya HIRAISHI, Atsuhiro TADOKORO, Hideyoshi FUJISAKU: “Characteristics of long period waves observed in port”, Rept. of PHRI Vol. 35, No. 3, 1996, pp. 3-36 (in Japanese). 47) Tomotsuka TAKAYAMA, Tetsuya HIRAISHI: “Amplification mechanism of harbor oscillation derived from field observation and numerical simulation”, Tech. Note of PHRI, No. 636, 1988, 70 p. (in Japanese). 48) Tetsuya HIRAISHI, Tokuhiro TADOKORO, Shigenori TAMAKI, Junzo HASEGAWA: “Standard frequency spectrum of long-period waves for design of port and harbor facilities,” porc. 44th Japanese Coastal Eng. Corof., 1997, pp. 246-250 (in Japanese). 49) Shigeru UEDA, Satoru SHIRAISHI, Hiroyuki OSHIMA, Kohei ASANO: “Allowable wave height and wharf operation efficiency based on the oscillations of ships moored to quay walls”, Tech. Note of PHRI, No. 779, 1994, 44 p. (in Japanese). 50) Ertekin, R. C., W. C. Webster and J. V. Wehausen: “Ship generated solitions”, Proc. 15th Symp. Nav. Hydrodyn., 1985, pp. 347-364. 51) Yoshimi GODA: “On the methodology of selecting design wave height”, Proc. 21st Int. Conf. Coastal Eng., ASCE, 1988, pp. 899-913. 52) Yoshimi GODA and Koji KOBUNE: “Distribution function fitting to storm waves”, Proc. 22nd Int. Conf. Coastal Eng., ASCE, 1990, pp. 18-31. 53) Yoshimi GODA: “Random Waves and Design of Maritime Structures (2nd Edition)”, World Scientific, Singapore, 2000, Chapt. 11 (Statistical Analysis of Extreme Waves). 54) Yoshimi GODA: “Statistical variability of sea state parameters as a function of wave spectrum,” Coastal Engineering in Japan, JSCE, Vol. 31, No. 1, 1988, pp. 39-52.

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Chapter 5 Wave Force 5.1 General (Notification Article 5, Clause 1) The wave force acting on a structure shall be determined using appropriate hydraulic model experiments or design methods described in 5.2 Wave Force Acting on Upright Wall, with the design waves determined by the procedures described in Chapter 4 Waves. [Commentary] (1) Structure Type and Wave Forces Wave forces can be generally classified by the type of structure as follows: (a) (b) (c) (d)

Wave force acting on a wall-type structure Wave force acting on armor stones or concrete blocks Wave force acting on submerged members Wave force acting on structures near the water surface

The wave force is different for each type of structure. It is thus necessary to use an appropriate calculation method in accordance with the structural type. For some types of structures with a few experiences of construction, their wave forces have not been sufficiently elucidated, and therefore it is desirable to carry out studies including hydraulic model experiments for such structures. (2) Wave Irregularity and Wave Force Sea waves are irregular with the wave height and period varying from wave to wave. Depending on the water depth and the topography of the sea bottom, there may appear waves that have not broken, waves that are just breaking, or waves that have already broken. When calculating the wave force, it is important to include the waves that cause the severest effect on the structure. It is necessary to give sufficient consideration to wave irregularity and to the characteristics of the wave force being produced in accordance with the type of structure. In general, it may be assumed that the larger the wave height, the greater the wave force becomes. It is thus acceptable to focus on the wave force of the highest wave among a train of irregular waves attacking the structure. However, with regard to the stabilities of floating structures and cylindrical structures with small rigidity, and those of concrete blocks or armor stones on the slope, it is desirable to consider the effect of the successive action of the irregular waves. (3) Calculation of Wave Force Using Hydraulic Model Experiments When studying wave force using hydraulic model experiments, it is necessary to give sufficient consideration to the failure process of the structure and to use an appropriate measurement method. It is also necessary to give sufficient consideration to the irregularity of waves. In particular, when carrying out experiments using regular waves, an investigation against the highest wave should be included.

5.2 Wave Force Acting on Upright Wall 5.2.1 General Considerations The wave force acting on an upright wall varies with the wave conditions, as well as the tidal level, the water depth, the sea bottom topography, the cross-sectional form of the structure, and the configuration of the alignment of the structure. The wave force acting on an upright wall shall thus be calculated appropriately while considering these items. An upright wall on a steep seabed or a high mound is often subjected to a strong impulsive wave breaking force, so that sufficient attention shall be paid to the conditions under which such a force is generated when calculating the wave force. [Commentary] (1) Parameters Affecting Wave Force on an Upright Wall 1) The major parameters that affect the wave force acting on an upright wall are the wave period, the wave height, the wave direction, the water level, the water depth, the bottom slope, the water depth on and the berm width of the mound, the crown height of upright wall, and the water depth at the base of upright wall. In addition, it is also necessary to consider the effect of the wall alignment. The wave force on an upright wall with a concaved alignment may be larger than that on an upright, straight wall of infinite length. Furthermore, if the front face of upright wall is covered with a mound of wave-dissipating concrete blocks, the characteristics of these blocks and the crown height and width of the mound will affect the wave force. (2) Types of Wave Force The wave force acting on an upright wall can be classified according to the type of waves as a standing wave force, a breaking wave force, or a wave force due to a broken wave. It is considered that the wave force changes -100-


its type continuously with the variation in the offshore wave height. A standing wave force is produced by waves whose height is small compared with the water depth, and the change in the wave pressure over time is gradual. As the wave height increases, the wave force also increases. In general, the largest wave force is generated by the waves breaking just a little off the upright wall. Accordingly, with the exception of very shallow water conditions, the force exerted by waves breaking just in front of an upright wall is larger than the wave force by higher waves that have already broken well. It is necessary to note that when breaking waves act on an upright wall on a steep seabed, or on an upright wall set on a high mound (even if built on a gentle seabed), a very strong impulsive breaking wave force may be generated.

5.2.2 Wave Forces of Standing and Breaking Waves [1] Wave Force under Wave Crest (Notification Article 5, Clause 1, Number 1) (1) Wave Pressure on the Front Face of an Upright Wall Assuming a linear distribution of wave pressure with a maximum value p1 at the still water level, 0 at the height h* above the still water level, and p2 at the sea bottom, the wave pressure from the bottom to the crown of the upright wall shall be calculated by the following equations: h* = 0.75 ( 1 + cos b )l 1 H D

(5.2.1)

p 1 = 0.5 ( 1 + cos b ) ( a 1 l 1 + a 2 l 2 cos 2 b )r 0 g H D

(5.2.2)

p1 p 2 = ---------------------------------cosh ( 2 p h ¤ L ) p3 = a3 p1 where h*： p1： p2： p3： r0： g： b： l1, l2：

h： L： HD：

(5.2.3) (5.2.4)

height above still water level at which intensity of wave pressure is 0 (m) intensity of wave pressure at still water level (kN/m2) intensity of wave pressure at sea bottom (kN/m2) intensity of wave pressure at toe of upright wall (kN/m2) density of water (t/m3) gravitational acceleration (m/s2) angle between the line normal to the upright wall and the direction of wave approach. The angle shall be reduced by 15º, but the resultant angle shall be no less than 0º. This correction provides a safety margin against uncertainty in the wave direction. wave pressure modification factors (1.0 is the standard value) water depth in front of upright wall (m) wavelength at water depth h used in calculation as specified in the item (3) below (m) wave height used in calculation as specified in the item (3) below (m)

1 ì 4 ph ¤ L ü a 1 = 0.6 + --- í --------------------------------- ý 2 î sinh ( 4 p h ¤ L ) þ

2

(5.2.5)

ì h b d H D 2 2d ü x 2 = min í æ --------------ö æ -------ö , ------- ý è ø è d ø HD î 3h b þ

(5.2.6)

ü h¢ ì 1 (5.2.7) a 3 = 1 ---- í 1 ---------------------------------- ý hî cosh ( 2 p h ¤ L ) þ where hb： water depth at an offshore distance of 5 times the significant wave height from the upright wall (m) d： water depth at the crest of either the foot protection works or the mound armoring units of whichever is higher (m) h¢： water depth at toe of upright wall (m) min {a,b}： smaller value of a or b (2) Uplift beneath Upright Wall The uplift acting on the bottom of an upright wall is described by a triangular distribution, with the pressure intensity at the front toe pu given by the following equation and 0 at the rear toe. p u = 0.5 ( 1 + cos b )a 1 a 3 l 3 r0 g H D

(5.2.8) -101-


where pu： uplift pressure acting at front toe of upright wall (kN/m2) l3： uplift pressure modification factor (1.0 is the standard value) (3) Wave Height and Wavelength Used in the Wave Pressure Calculation The wave height HD and the wavelength L are the height and wavelength of the highest wave. The wavelength of the highest wave is that corresponding to the significant wave period, while the height of the highest wave is as follows: (a) When the upright wall is located off the breaking zone:

678

HD = Hmax (5.2.9) Hmax = 1.8H1/3 where Hmax： highest wave height of incident waves at the water depth at the upright wall (m) H1/3： significant wave height of incident waves at the water depth at the upright wall (m) (b) When the upright wall is located within the breaking zone: HD is the maximum wave height considering the breaking of irregular waves (m) [Commentary] It is standard to calculate the maximum horizontal wave force acting on an upright wall and the simultaneous uplift pressure using the extended Goda equation. The extended Goda pressure formula is that proposed by Goda and modified to include the effects of wave direction and others. Its single-equation formula enables to calculate the wave force from the standing to breaking wave conditions without making any abrupt transition. However, where the upright wall is located on a steep seabed, or built on a high mound, and is subjected to a strong impulsive wave pressure due to breaking waves, the formula may underestimate the wave force. It should therefore be carefully applied with consideration of the possibility of occurrence of impulsive wave pressure due to breaking waves (see 5.2.3 Impulsive Pressure Due to Breaking Waves). The wave pressure given by the Goda formula takes the hydrostatic pressure at the still water condition as the reference value. Any hydrostatic pressure difference between the offshore and onshore sides of the wall, if presents, should be considered separately. Further, the equation is designed to examine the stability of the whole body of vertical wall. When breaking wave actions exist, the equation does not necessarily express the local maximum wave pressure at the respective positions; thus such should be considered during examination of the stress of structural members. [Technical Notes] (1) Wave Pressure on the Front Face According to the Extended Goda Formula Figure T- 5.2.1 illustrates the distribution of wave pressure acting on an upright section of a breakwater. The correction to the incident wave angle b is exemplified in Fig. T- 5.2.2. p

1

η* hc

d h'

Buoyancy

h

p

u

p

p

2

3

Fig. T- 5.2.1 Wave Pressure Distribution Used in Design Calculation

-102-


15

°

90 °

β

Pr inc

ipa

ld

ire c

tio n

of

wa

ve

Normal Line to the alignment

Fig. T- 5.2.2 Way of Obtaining the Incident Wave Angle b

(2) Highest Wave In breakwater designs in general, it is necessary to evaluate the largest wave force that can be given by the Goda formula using the highest wave. The appearance of the highest wave in an irregular wave group is probabilistic, and so it is not possible to determine the highest wave explicitly. Nevertheless, after examination of the results of applying the current method to breakwaters in the field, it has been made standard to use 1.8 times the significant wave height as the height of the highest wave when the upright wall is located off the breaking zone. It has also been made standard to use the wavelength corresponding to the significant wave period as the wavelength of the highest wave. In order to determine whether or not the highest wave is subject to wave breaking, the graphs for determining the highest wave height (Fig. T- 4.5.15 (a)~(e) in 4.5.6 Wave Breaking) should be used by referring to the location of the peak wave height in the zone in the onshore side of the 2% decay line. It is acceptable to consider that the highest wave is not subject to wave breaking when the water is deeper than that at the peak height, but that it is subject to wave breaking when the water is shallower than this. If the highest wave height is to be obtained using the approximate equation (4.5.23) in 4.5.6 Wave Breaking, hb should be substituted as h in the first term in the braces { } on the right-hand side of the equation. If using a value other than 1.8 as the coefficient on the right-hand side of equation (5.2.9), it is necessary to conduct sufficient investigations into the occurrence of the highest wave and then choose an appropriate value (see 4.1.3 [2] Statistical Properties of Waves). (3) Correction Factors l1, l2, l3 Equations (5.2.1) ~ (5.2.8) are the extended version of the Goda formula. It contains three correction factors so that it can be applied to walls of different shapes and conditions. For an upright wall, the correction factors are of course 1.0. The wave pressure acting on other types of wall such as a caisson covered with a mound of wavedissipating concrete blocks or a perforated-wall caisson may be expressed using the extended Goda formula with appropriate correction factors (see 5.2.4 Wave Force on Upright Wall Covered with Wave-Dissipating Concrete Blocks). (4) Application of Other Theoretical and Calculation Equations When the ratio of the wave height to the water depth is small and a standing wave force is obviously exerted on a upright wall, a high-accuracy, standing wave theory may be applied. In this case, however, it is necessary to give sufficient consideration to the irregularity of waves in the field, and to evaluate the force due to the highest wave. Moreover, when the applicability can be verified based on past results for existing breakwaters, the Sainflou formula 3) and the Hiroi formula may also be used for a design wave force calculation. (5) Features and Application Limits of the Goda Formula The first feature of the Goda formula is that the wave force from standing waves to breaking waves can be evaluated continuously, including the effect of period. The parameter a1 given by equation (5.2.5) expresses the effect of the period (strictly speaking h/L); it takes the limiting values of 1.1 for shallow water waves and 0.6 for deepwater waves. The effect of period also appear when evaluating the maximum wave height to be used in the calculation; for a constant deepwater wave height, the longer the period, the larger the maximum wave height in the surf zone. Since the Goda formula incorporates the effect of period on the wave pressure as well as on the maximum wave height, it is necessary to take sufficient care when determining the period in the design conditions. Another feature of the Goda formula is that the change in the wave force with the mound height and the bottom slope is considered by means of the parameter a2. As can be seen from equation (5.2.6), as the mound height is gradually increased from zero (i.e., d = h), a2 gradually increases from zero to its maximum value. After reaching its maximum value, a2 then decreases until it reaches zero again when d = 0. The maximum value of a2 is 1.1; combining this with the maximum value of a1 of 1.1, the intensity of the wave pressure p1 at the still water level is given by 2.2r0gHD. With regard to the effect of the bottom slope, hb within the equation for a2 is taken as the water depth at the distance of 5 times the design significant wave height from the upright wall. Because of this artifice, a steep bottom slope results in the same effect as having a high mound. The effect of the bottom slope also appear when evaluating the maximum wave height to be used in the calculation. In the wave breaking zone, the steeper the bottom slope, the larger the wave height, because the wave height used in the calculation is the maximum wave -103-


height at the a distance 5H1/3 offshore from the upright wall. The bottom slope thus has a strong influence on the wave force, and so care must be taken when setting the bottom slope in the design conditions. As explained above, the Goda formula considers the effects of the mound height and the bottom slope on the wave pressure. Nevertheless, for an upright wall on a high mound or a steep sea bed, a large impulsive breaking wave force may act, and under such conditions the Goda formula may underestimate the wave force. When applying the Goda formula, it is thus necessary to pay attention to the risk of an impulsive breaking wave force arising. In particular, with a high mound, it is necessary to consider not only a2 in equation (5.2.6) but also the impulsive breaking wave force coefficient aI by Takahashi et al. (see 5.2.3 Impulsive Pressure Due to Breaking Waves), and to use aI in place of a2 when aI is the larger of the two. One more problem with the Goda formula concerns its applicability to extremely shallow waters, for example near to the shoreline. The Goda formula cannot be applied accurately for broken waves. It is difficult, however, to clearly define where the limit of applicability lies. For cases such as the wave force acting on an upright wall near the shoreline, it is advisable to use other calculation equations together with the Goda formula. (see 5.2.7 Wave Force on Upright Wall near Shoreline or on Shore). (6) Modification of the Original Goda Formula for Wave Direction Although results from a number of experiments on the effect of wave direction on the wave force are available, there are still many points that are unclear. Traditionally, for standing waves, no correction has been made for wave direction to the wave force. The effects of wave direction have been considered only for breaking waves, by multiplying the wave force by cos2b, where b is the angle between the direction of wave approach and the line normal to upright wall alignment. However, this has resulted in the irrational situation whereby the breaking wave force is assumed to decrease as the wave angle b increases, reaching zero at the limiting value b = 90º, and yet standing waves are assumed to maintain as the perfect standing wave condition. One explanation is such that because actual breakwaters are finite in extension, when the incident angle is large (i.e., oblique wave incidence), it takes a considerably large distance from the tip of breakwater until the wave height becomes two times the incident height. As b approaches to the limiting value of b = 90º, the distance to the place where the wave height becomes two fold tends to go to infinity. In other words, in this case, it is appropriate to consider that the wave pressure of progressive waves acts on the upright wall. Considering these points and application to breakwaters in the field, it has been decided to correct equation (5.2.2) for wave direction by multiplying a2 (which represents mound effects) with cos2b, and then multiplying the whole term by 0.5(1+cosb). (7) Wave Force and Significant Wave Period for Waves Composed of Two Groups of Different Periods Examples of two wave groups of different periods being superimposed are such a case that waves enter a bay from the outer sea and another group of waves are generated within the bay by local winds. Another case is the superposition of diffracted waves coming from the entrance of a harbor and waves transmitted by means of overtopping. In such cases, the spectrum is bimodal (i.e., having two peaks), and there are actual cases of such observations in the field. Tanimoto et al. 4) carried out experiments on the wave force acting on the upright section of a composite breakwater by using waves with a bimodal spectrum, and verified that the Goda formula can be applied even in such a case. They also proposed a method for calculating the significant wave period to be used in the wave force calculation (see 4.5.4 Wave Reflection). If each frequency spectrum of the two wave groups before superimposition can be considered to be a Bretschneider-Mitsuyasu type, the significant wave period after superimposition may be obtained using the method by Tanimoto et al. Then this significant wave period may be used in wave force calculation. (8) Wave Force for Low Crested Upright Wall According to results of model experiments, the stability of upright wall tends to increase as the crown height is reduced. Nakata and Terauchi have proposed a method for calculating the wave force for a breakwater with a low crown height. In the method, the horizontal wave pressure and the uplift pressure from the Goda formula are multiplied by a modification factor lh, thus reducing the wave force. (9) Wave Force for High Crested Upright Wall When the crown of the upright wall is considerably higher than that for a normal breakwater, there will be no overtopping, meaning that the wave force may be larger than that given by the Goda formula. Mizuno and Sugimoto carried out experiments into the wave force acting on a breakwater with a high crown. (10) Wave Force on Inclined Walls When the wall is slightly inclined, the horizontal wave force is more-or-less the same as that for a perfectly upright wall. However, it is necessary to consider the vertical component of the wave force acting on the inclined surface, along with the reduction in uplift pressure and others. Tanimoto and Kimura 5) have carried out experiments on the wave force for trapezoidal caisson walls, and have proposed a method for calculating the wave force. For a caisson in which the upper part of the upright section is inclined (sloping-top caisson), the horizontal wave force is reduced not only for the sloping part but also for the vertical part. It is also necessary to consider the vertical component of the wave force for the sloping part for stability analysis of breakwaters. Morihira et al. were the first to propose a method for calculating the wave force in such a case. Hosoyamada et al. have come up with a method that is based on the approach by Morihira et al., but the method by Hosoyamada is more general and can be applied for a wider variety of sloping-top caissons (see Part VII, 3.2.4 Sloping-Top Caisson Breakwater). -104-


(11) Uplift on a Caisson with a Footing When a caisson has a footing, a wave force acts downwards on the upper surface of the footing on the seaside, and an uplift pressure of p¢u acts at the front toe, while the uplift pressure at the rear toe is zero. Nevertheless, in general the resultant force is not significantly different to that without the footing. It is thus acceptable to ignore the footing, and to assume that the uplift pressure has a triangular distribution as shown in Fig. T- 5.2.3, with the uplift pressure pu at the front toe being given by equation (5.2.8), and the uplift pressure at the rear toe being zero. If the footing is extremely long, however, it is necessary to calculate the uplift pressure appropriately, considering the change in the uplift pressure p¢u at the front toe of the footing.

Fig. T- 5.2.3 Uplift Pressure When There Is a Footing

(12) Wide Mound Berm in Front of the Upright Wall The wave force acting on the upright wall of a composite breakwater varies not only with the mound height but also with the berm width and the front slope of mound (see 5.2.3 Impulsive Pressure Due to Breaking Waves). As explained, of these three factors, the Goda formula incorporates only the effect of the mound height. Consequently, if the width and/or slope of the mound are considerably different from normal, it is advisable to carry out studies using hydraulic model experiments. Note however that if the berm is sufficiently wide, it may be considered as a part of the topography of the sea bottom. Even with the standard formula, if the width is more than one half of the wavelength, it is thus standard to use the water depth on the mound for evaluation of both the wave height and the wavelength to be used in the wave force calculation. (13) Wave Force Acting on an Upright Wall Comprised of a Row of Vertical Cylinders Nagai et al. and Hayashi et al. have carried out studies on the wave force acting on an upright wall comprised of a row of cylinders (a pile breakwater). Through their researches, it has been verified that the wave force is not greatly different from that acting on an upright wall with a flat face. It is thus acceptable to treat an upright wall comprised of a row of cylinders as having a flat face and calculate the wave force using the Goda formula.

[2] Wave Force under Wave Trough (Notification Article 5, Clause 1, Number 2) The negative wave force at the time of wave trough acting at a wall shall be calculated using either appropriate hydraulic model experiments or an appropriate calculation formula. [Commentary] When the trough of a wave is acting at a wall, a negative wave force acts corresponding to the trough depth of the water surface from the still water level. A “negative wave force” is the force directed seaward. It is necessary to note that the negative wave force may be comparable in magnitude to a positive wave force when the water is deep and the wavelength is short. [Technical Notes] (1) Negative Wave Pressure Distribution The negative wave pressure acting on an upright wall at the wave trough can be approximately estimated as shown in Fig. T- 5.2.4. Specifically, it can be assumed that a wave pressure acts toward the sea, with the magnitude of this wave pressure being zero at the still water level and having a constant value of pn from a depth 0.5HD below the still water level right down to the toe of the wall. Here, pn is given as follows: p n = 0.5r 0 gH D where pn： intensity of wave pressure in constant region (kN/m2) r0： density of seawater (usually 1.03 t/m3) g： gravitational acceleration (9.81 m/s2) HD： wave height used in design calculation (m)

(5.2.10) Seaward

0.5

Shoreward

HD

In addition, the negative uplift pressure acting on the bottom of the upright wall can be assumed to act as shown in Fig. T- 5.2.4. Specifically, it can be assumed that an uplift pressure acts downwards with its intensity being pn (as given by equation (5.2.10)) T- 5.2.4 Negative Wave Pressure Distribution at the front toe, zero at the rear toe, and having a triangular distribution in-between. Incidentally, it is necessary to use the highest wave height as the wave height HD used in the design calculation. -105-


(2) Negative Wave Force by Finite Amplitude Wave Theory Goda and Kakizaki 6) have carried out a wave force calculation based on the fourth order approximate solutions of a finite amplitude standing wave theory, and presented calculation diagrams for negative wave pressure. It has been verified that their calculation results agree well with experimental results. When the water is deep and standing waves are clearly formed, it is acceptable to use the results of this finite amplitude standing wave theory of higher order approximation. It should be noted that, for a deepwater breakwater, the negative wave force at the wave trough may become larger than the positive wave force at the wave crest, and that the upright wall may slide toward offshore.

5.2.3 Impulsive Pressure Due to Breaking Waves (1) When it is apprehended that an impulsive pressure due to breaking waves may be generated, a study including hydraulic model experiments shall be carried out as a general rule. (2) It is desirable to avoid the adoption of cross-sectional forms and structure type that may induce the generation of large impulsive pressure due to breaking waves. If a large impulsive pressure due to breaking waves cannot be avoided, it is desirable to redesign the structure such that the wave force will be reduced, for example by providing appropriate wave-absorbing works. [Commentary] An impulsive pressure is generated when the wave front of a breaking wave strikes a wall surface. It has been shown from model experiments that under certain conditions the maximum wave pressure may rise as much as several tens of times the hydrostatic pressure corresponding to the wave height (1.0r0 gHD). However, such a wave pressure acts only locally and for a very short time, and even slight changes in conditions lead to marked reduction in the wave pressure. Because of the impulsive nature of the wave force, the effects on stability and the stress in structural elements vary according to the dynamic properties of the structure. Accordingly, when there is a risk of a large impulsive pressure due to breaking waves being generated, it is necessary to take appropriate countermeasures by understanding the conditions of the impulsive pressure generation and the wave force characteristics by means of hydraulic model experiments. [Technical Notes] (1) Conditions of Impulsive Pressure Due to Breaking Waves A whole variety of factors contribute to generation of an impulsive pressure due to breaking waves, and so it is difficult to describe the conditions in general. Nevertheless, based on the results of a variety of experiments, it can be said that an impulsive pressure is liable to occur in the following cases when the wave angle b is less than 20º. (a) Steep bottom slope When the three conditions (the bottom slope is steeper than about 1/30; there are waves that break slightly off the upright wall; and their equivalent deepwater wave steepness is less than 0.03) are satisfied simultaneously, then an impulsive pressure is liable to be generated. (b) High mound Even if the bottom slope is gentle, the shape of the rubble mound may cause an impulsive pressure to be generated. In this case, in addition to the wave conditions, the crown height, the berm width and the slope gradient of the mound all play a part, and so it is hard to determine the conditions under which such an impulsive pressure will be generated. In general, an impulsive pressure will be generated when the mound is relatively high, the berm width is suitably wide or the slope gradient is gentle, and breaking waves form a vertical wall of water at the slope or at the top of the mound 7). When the seabed slope is gentler than about 1 ¤ 50 and the ratio of the depth of water above the top of the mound (including any armor work) to the water depth above the seabed is greater than 0.6, it may be assumed that a large impulsive pressure will not be generated. (2) Countermeasures If a large impulsive pressure due to breaking waves acts on an upright wall, the wave force can be greatly reduced by sufficiently covering the front face with a mound of wave-dissipating concrete blocks. In particular, with a high mound, a sufficient covering with wave-dissipating concrete blocks can stop the occurrence of the impulsive pressure itself. In some cases the action of an impulsive pressure can also be avoided by using special caissons such as perforated-wall caissons or sloping-top caissons 7). The wave direction also has a large effect on the occurrence of an impulsive pressure, and therefore, one possible countermeasure is to ensure that the wave direction is not perpendicular to the breakwater alignment. (3) Investigating Wave Force Using Model Experiments When investigating the wave force using hydraulic model experiments for the case that an impulsive pressure due to breaking wave acts, it is necessary to give consideration to the response characteristics of the structure to

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the impact force. It is better to study the stability of the upright wall as a whole by sliding tests, and to study the strength of structural elements such as parapets by stress and strain measurements. (4) Impulsive Pressure Due to Breaking Waves Acting on an Upright Wall on a Steep Seabed. (a) Water depth that produce a maximum wave pressure and the mean intensity of wave pressure Mitsuyasu 8), Hom-ma et al., Morihira et al. 9), Goda and Haranaka 10), Horikawa and Noguchi, and Fujisaki and Sasada have all carried out studies on the impulsive pressure due to breaking waves acting on an upright wall on a steeply sloping sea bottom. In particular, Mitsuyasu carried out a wide range of experiments using regular waves whereby he studied the breaking wave force acting on an upright wall on uniform slopes of gradient 1/50, 1/25, and 1/15 for a variety of water depths. He investigated the change in the total wave force with the water depth at the location of the upright wall, and obtained an equation for calculating the water depth hM at the upright wall for which the impulsive wave force is largest. When the Mitsuyasu equation is rewritten in terms of the deepwater wavelength, it becomes as follows: hM ------ = H0 where CM = H0： L0： tanq：

H 0 1 / 4 C M æ ------ö è L0 ø

(5.2.11)

0.59 3.2 tan q deepwater wave height (m) deepwater wavelength (m) gradient of uniform slope

(5.2.12)

Hom-ma, Horikawa and Hase have proposed a slightly different value for CM based on the results of experiments with a gradient of 1/15 and other data. In any case, the impulsive wave pressure is largest when the structure is located slightly shoreward of the wave breaking point for progressive waves. Figure T- 5.2.5 shows the total wave force when the impulsive wave force is largest for a number of slope gradients, as based on the results of Mitsuyasu’s experiments. In this figure, the mean intensity of the wave pressure p has been obtained and then divided by r0gHD to make it dimensionless; it has then been plotted against the deepwater wave steepness. It is possible to gain an understanding of the overall trend from this figure. Specifically, it can be seen that the smaller the wave steepness, the larger the impulsive pressure is generated. Also, as the slope gradient becomes smaller, the intensity of the maximum impulsive pressure decreases. (b) Conditions for generation of impulsive breaking wave pressure The conditions for the occurrence of an impulsive pressure on a steep seabed, as described in (1) (a), have been set by primarily employing Fig. T- 5.2.5 as a gross guideline. For irregular waves in the sea, the wave steepness can be evaluated as the ratio of the equivalent deepwater wave height corresponding to the highest wave height Hmax to the deepwater wavelength corresponding to the significant wave period: the wave height Hmax is to be evaluated at the distance 5H1/3 from the upright wall. One may refer to Fig. T- 5.2.5 in order to obtain an approximate estimate of Fig. T- 5.2.5 Mean Intensity of Wave Pressure the mean intensity of the wave pressure for for the Severest Wave Breaking this equivalent deepwater wave steepness. (Upright Wall on a Steep Slope) In this case, Hb should be taken to be the aforementioned Hmax. One can also envisage an installation of a breakwater at a place where the risk of impulsive pressure generation is not large for the design waves. However, when placing an upright wall closer to the shore where waves already broken act upon, it becomes important to carry out investigations for waves with a height lesser than that of the design waves. (c) Impulsive wave force acting on an upright wall on a horizontal floor adjoining a steep slope Takahashi et al. 11) have carried out studies on the impulsive wave pressure acting on an upright wall on a horizontal floor that is joined to a steep slope. They employed a horizontal berm connected to a slope of gradient 1/10 or 3/100 in a wave channel, and then measured the wave pressure that acts on an upright wall at a variety of positions with regular waves. They have proposed an equation (valid for certain wave conditions) for calculating the upright wall position at which the wave force is largest and the maximum wave force in that condition.

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(5) Impulsive Wave Pressure Acting on a Composite Breakwater (a) Effect of the mound shape (impulsive breaking wave pressure coefficient) Takahashi et al. have proposed, based on the results of sliding experiments 7), the impulsive breaking wave pressure coefficient aI. This is a coefficient that represents the extent of the impulsive pressure due to breaking waves when the mound is high. It is expressed as the function of the ratio of the wave height to the depth of water above the mound in front of the caisson H/d, the ratio of the depth of water above the mound to the original water depth at the upright wall d/h, and the ratio of the berm width of the mound to the wavelength at this place BM/L. Note that the wave height H is the design wave height (highest wave height). The impulsive breaking wave pressure coefficient aI is expressed as the product of aI0 and aI1 as in the following equations: a I = a I0 a I1

(5.2.13)

£2 ìH ¤ d : H ¤ d ≦ a I0 = í : H¤d>2 î2

(5.2.14)

Figure T- 5.2.6 shows the distribution of aI1. It attains the maximum value of 1 when d/h is 0.4 and BM /L is 0.12. The impulsive breaking wave pressure coefficient aI takes values between 0 and 2; the larger the value of aI, the larger the impulsive breaking wave force is. When calculating the wave force using Goda’s formula, one should use aI in place of a2 (equation (5.2.6)) if aI is larger than a2. Note that equation (5.2.13) for aI has been derived for the case of H/h being equal to 0.60 or greater, based on the results of sliding experiments. This coefficient aI may be used when examining the sliding of an upright wall against the waves of relatively large height.

1.0

α

I1

0.1

0.8 0.2 0.4 0.6 0.8 0.9

0.6

hd h 0.4

0.0

(b) Effect of the crown height of the upright wall The higher the crown height, the greater the risk of an impulsive breaking wave force being generated. This is because the steep front of a breaking wave often takes a nearly vertical cliff of water above the still water level, and if there is an upright wall at this place, the impact of the wave front results in the generation of an impulsive force. For example, Mizuno et al. have pointed out the tendency that, when the crown is high, an impulsive breaking wave force is generated even when the mound is relatively low.

α

0.2

I

=

α

α α

I0

I0

=

BH I1

H:H d d H 2 : d

d

< 2 = > 2

0 0

0.1

0.2

0.3

0.4

BM L Fig. T- 5.2.6 Impulsive Breaking Wave Pressure Coefficient aI1

(c) Effect of the wave direction According to the results of the sliding experiments of Tanimoto et al. 7), even if conditions are such that a large impulsive pressure is generated when the wave angle b is 0º, there is a rapid drop in the magnitude of the wave force as b increases to 30º or 45º. By considering the fluctuation in the wave direction, it is reasonable to assume that the condition for the generation of an impulsive wave force is that b is less than 20º. (d) Dynamic response of the upright section to an impulsive force and the sliding of upright section When an impulsive pressure due to breaking waves acts on an upright section, the instantaneous local pressure can rise up to several tens of times the hydrostatic pressure corresponding to the wave height, although the duration time of the impulsive pressure is very short. The impulsive peak pressures fluctuate significantly, but the fluctuations in the impulse are not large. It is necessary to evaluate the contribution of the impulsive breaking wave force to sliding in terms of the dynamic response, considering deformation of the mound and the subsoil. Goda 12) as well as Takahashi and Shimosako, have carried out calculations of the shear force at the bottom of an upright section using dynamic models. Judging by the results of these calculations and the results of various sliding experiments, it would seem reasonable to take the mean intensity of the wave pressure equivalent to the sliding shear force to be (2.5 ~ 3.0) r0gH. The impulsive breaking wave pressure coefficient aI has been introduced, based on the results of sliding experiments with consideration of such dynamic response effects.

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5.2.4 Wave Force on Upright Wall Covered with Wave-Dissipating Concrete Blocks The wave force acting on an upright wall covered with a mound of wave-dissipating concrete blocks shall be evaluated based on hydraulic model experiments or an appropriate calculation equation, considering the crown height and width of the wave-absorbing work as well as the characteristics of the wave-dissipating concrete blocks. [Commentary] If the front face of an upright wall is covered with a mound of wave-dissipating precast concrete blocks, the features of wave force acting on the wall are changed. The extent of this change depends on the characteristics of incident waves, along with the crown height and width of the wave-absorbing work, the type of wave-dissipating concrete blocks used, and the composition of the wave-absorbing work. In general, when nonbreaking waves act on an upright wall, the change in wave force upon the upright wall covered with wave-dissipating blocks is not large. However, when a large impulsive breaking wave force acts, the wave force can be reduced significantly by covering the upright wall with a mound of wave-dissipating blocks. Nevertheless, such a reduction in the wave force is only achieved when the wave-absorbing work has a sufficient width and crown height; in particular, it should be noted that if the top of the wave-absorbing work is below the design water level, the wave-absorbing work often invites an increase in the wave force. [Technical Notes] (1) Wave Force Calculation Formula for Upright Wall Sufficiently Covered with Wave-Dissipating Concrete Blocks The wave force acting on an upright wall covered with a mound of wave-dissipating concrete blocks varies depending on the composition of the wave-absorbing work, and therefore it should be evaluated using the results of model experiments corresponding to the design conditions. However, if the crown elevation of the waveabsorbing work is as high as the top of the upright wall and the wave-dissipating concrete blocks are sufficiently stable against the wave actions, the wave force acting on the upright wall may be calculated using the extended Goda formula. In this method with the standard formula given in 5.2.2 Wave Forces of Standing and Breaking Waves, the values of h*, p1, and pu given by equations (5.2.1), (5.2.2), and (5.2.8) are used respectively, but it is necessary to assign appropriate values to the wave pressure modification factors l1, l2, and l3 in accordance with the design conditions. (2) Modification Factors for the Extended Goda Formula The method using the extended Goda formula can be applied by assigning appropriate values to the modification factors l1, l2, and l3. Studies have been carried out by Tanimoto et al. 13), Takahashi et al. 14), Sekino and Kakuno, and Tanaka and Abe amongst others and have revealed the following: (a) Wave-dissipating concrete blocks cause a considerable reduction in the breaking wave pressure, and so it is generally acceptable to set the breaking wave pressure modification factor l2 to zero. (b) The larger the wave height, the smaller the modification factor l1 for standing wave type pressure and the modification factor l3 for uplift pressure become. (c) The larger the ratio of the block mound width to the wavelength, the smaller the modification factors l1 and l3 become. (d) If even a small portion of the upper part of the upright section is left uncovered, there is a risk of the wave force here becoming an impulsive breaking force.

678

Based on such experimental results, Takahashi et al. 14) have summarized that in general, when the upright wall is sufficiently covered with wave-dissipating concrete blocks, the wave pressure reduction factor l2 may be taken to be zero, while the values of l1 and l3 depend primarily on the wave height H (the highest wave height). They have thus proposed the following equations:

l1 =

1.0 1.2 - (2/3)(H/h) 0.8

: H/h ≦ 0.3 : 0.3 ＜ H/h ≦ 0.6 : H/h ＞ 0.6

l3 = l1 l2 = 0

(5.2.15) (5.2.16) (5.2.17)

In the breaker zone, where breakwaters covered with wave-dissipating concrete blocks are generally used, the above equations give l1 = l3 = 0.8.

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5.2.5 Effect of Alignment of Breakwater on Wave Force In the case when the distribution of wave heights along the face line of a breakwater is not uniform, the wave force shall be calculated by giving appropriate consideration to this aspect of wave height distribution. [Commentary] When the extension of breakwater is not infinitely long, the distribution of the wave height along the face line of breakwater becomes non-uniform due to the effects of wave reflection and diffraction. Ito and Tanimoto 16) have pointed out that most damaged breakwaters having been struck by storm waves equivalent to design waves show a pattern of meandering distribution of sliding distance (they have termed this “meandering damage”), and that one of the causes of this type of damage is the differences in the local wave forces induced by the non-uniform wave height distribution. The variation of wave heights along the breakwater is particularly prominent when the breakwater alignment contains a corner that is concave with respect to the direction of wave incidence (see 4.5.4 [3] Transformation of Waves at Concave Corners, near the Heads of Breakwaters, and around Detached Breakwaters). Variations in wave heights along the breakwater alignment may also occur near the head of the breakwater. In particular, for a detached breakwater that extends over a short length only, diffracted waves from the two ends may cause large variations in wave heights 17). [Technical Notes] Wave force calculation methods that consider the effects of the shape of the breakwater alignment have not reached to the level of reasonable reliability yet. It is thus desirable to carry out an investigation using hydraulic model experiments. Nevertheless, there is a good correlation between the increase in the wave height owing to the shape of the breakwater alignment and the increase in the wave force. It is thus acceptable to increase the wave height for the design calculations in accordance with the extent of the effect of the shape of the breakwater alignment as in equation (5.2.18), and then calculate the wave force based on the standard calculation formula. HD¢ = min {KcHD, KcbHb}

(5.2.18)

where HD¢： wave height to be used in the wave force calculation in consideration of the effect of the shape of breakwater alignment (m) Kc： coefficient for the increase in wave height due to the effect of the shape of breakwater alignment; Kc ≧ 1.0 Kcb ≒ 1.4 Kcb： limit value of the height increase coefficient for breaking limit waves; HD： wave height used in the wave force calculation when the effects of the shape of breakwater alignment are not considered (m) Hb： breaking wave height at the offshore location with the distance of 5 times the significant height of progressive waves from the upright wall (m) The height increase coefficient Kc in equation (5.2.18) is generally expressed as in equation (5.2.19). It should be appropriately determined based on the distribution of the standing wave height (see 4.5.4 [3] Transformation of Waves at Concave Corners, near the Heads of Breakwaters, and around Detached Breakwaters) along the face line of breakwater as determined under the condition that the waves do not break. Kc = HS / {HI (1 + KR)}

(5.2.19)

where HS： standing wave height along the front wall of breakwater (m) HI： incident wave height (m) KR： reflection coefficient for the breakwater in question If the waves are treated as being of regular trains, then the coefficient for wave height increase varies considerably along the breakwater. Moreover, the height increase coefficient is very sensitive to the period of the incident waves and the direction of incidence. It is thus reasonable to consider the irregularity of the period and the direction of incident wave. It should be noted that the value of Kc obtained in this way varies along the breakwater and that there may be regions where Kc < 1.0. However, the wave height to be used in design must not be less than the original incident wave height. The limit value Kcb of the height increase coefficient for breaking waves has not been clarified in details. Nevertheless, it may be considered to be about 1.4 based on experimental results up to the present time.

5.2.6 Effect of Abrupt Change in Water Depth on Wave Force For an upright wall located in a place where the water depth changes abruptly owing to the presence of reefs and others, it is desirable to calculate the wave force acting on the upright wall based on hydraulic model experiments, by taking the rapid transformation of waves into consideration. -110-


[Technical Notes] Ito et al. 18) have carried out experiments on the wave force acting on an upright wall located on or behind a reef where the water depth is more-or-less uniform, with the offshore slope of the shoal having a gradient of about 1/10.

5.2.7 Wave Force on Upright Wall near Shoreline or on Shore [1] Wave Force at the Seaward Side of Shoreline It is desirable to calculate the wave force acting on an upright wall in shallow water near the shoreline based on hydraulic model experiments, considering the effects of water level changes due to surf beat etc. and the complex processes of random wave breaking. [Technical Notes] A number of different wave force formulas have been proposed for upright walls near the shoreline and on shore. It is necessary to carry out an appropriate wave force calculation in line with the design conditions. Very roughly speaking, the standard formula in 5.2.2 Wave Forces of Standing and Breaking Waves are applicable in the regions where the seabed slope is gentle and the water is relatively deep. The formula of Tominaga and Kutsumi is applicable in the regions near the shoreline. The formula of Hom-ma, Horikawa and Hase is applicable in the regions where the seabed slope is steep and the water is of intermediate depth. When applying the standard wave pressure formula to the places where the water depth is less than one half the equivalent deepwater wave height, it may be appropriate to use the values for the wavelength and wave height at the water depth equal to one half the equivalent deepwater wave height in the calculation.

[2] Wave Force at the Landward Side of Shoreline It is desirable to calculate the wave force on an upright wall situated on the landward side of the shoreline based on hydraulic model experiments, considering increases in the water level due to surf beat and wave setup as well as wave runup. [Technical Notes] For an upright wall situated on the landward side of the shoreline, the formulas by the US Army Coastal Engineering Research Center (CERC) 19) are available. Moreover, one may refer to the research that has been carried out by Tominaga and Kutsumi on the wave force acting on an upright wall situated on the landward side of the shoreline.

5.2.8 Wave Force on Upright Wave-Absorbing Caisson The wave force acting on an upright wave-absorbing caisson shall be calculated based on hydraulic model experiments or appropriate calculation formulas, considering changes in the wave force due to the structure of the wave-absorbing compartment. [Commentary] The wave force acting on an upright wave-absorbing caisson (perforated-wall caisson etc.) varies in a complex way. Specifically, it varies with the wave characteristics, the water level, the water depth, the topography of sea bottom and the shape of the mound as with the case of a normal upright wall, but it also varies with the structure of the waveabsorbing compartment. It is thus difficult to designate a general calculation method that can be used in all cases. Consequently, if the calculation method that is sufficiently reliable for the structure in question has not been proposed, it is necessary to carry out studies using hydraulic model experiments matched to the individual conditions. It is necessary to sufficiently investigate not only the wave force to be used in the stability investigation but also the wave force acting on structural members. Moreover, it should be noted that the wave force varies significantly according to whether or not the top of wave chamber is covered with a ceiling slab. [Technical Notes] (1) Wave Force without a Ceiling Slab in the Wave Chamber The wave force acting on an upright wave-absorbing caisson varies depending on the structural conditions of the wave-absorbing compartment, and so it is not possible to calculate this wave force for all general cases involved. Nevertheless, for the normal case where there is no ceiling slab in the wave chamber, one can use the extended Goda formula to calculate the wave force, provided necessary modifications are made. Takahashi et al. 20) have carried out experiments on a vertical-slit wall caisson, and have presented a method for calculating the wave pressure acting on the slit and rear walls for four representative phases, whereby the wave pressure given by the extended Goda formula is multiplied by a modification factor l for the vertical-slit wall caisson; they give specific values for the modification factor for the slit and rear walls for each phase. This method can be used to

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give not only the wave force that is severest in terms of the sliding or overturning of the caisson, but also the wave force that is severest in terms of the design of the elements for each wall. (2) Wave Force with a Ceiling Slab in the Wave Chamber When the top of the wave chamber is closed off with provision of a ceiling slab, an impulsive pressure is generated at the instant when the air layer in the upper part of the wave chamber is trapped in by the rise of water surface. It is thus necessary to give consideration to this impulsive pressure in particular with regard to the wave pressure used in design of structural elements. This impulsive pressure can be reduced by providing suitable air holes. However, it should be noted that if these air holes are too large, the rising water surface will directly hit the ceiling slab without air cushion, meaning that the wave force may actually increase 22), 23).

5.3 Mass of Armor Stones and Concrete Blocks 5.3.1 Armor Units on Slope (Notification Article 48, Clause 5) It shall be standard to calculate the mass of rubble stones or concrete blocks necessary to cover the front slope of a sloping structure that is subject to wave forces, by means of appropriate hydraulic model experiments or the following equation:

rr H 3 M = --------------------------3 NS ( S r 1 ) 3 where M： minimum mass of rubble stones or concrete blocks (t) rr： density of rubble stones or concrete blocks (t/m3) H： wave height used in the stability calculation (m) NS： stability number Sr： specific gravity of rubble stones or concrete blocks relative to sea water

(5.3.1)

[Commentary] The armor layer for the slope of a rubble mound breakwater protects the rubble stones in the inside, and so it is necessary to ensure that an armor unit has a mass sufficient to be stable against wave actions so that it does not scatter itself. The mass required to produce such stability can be calculated using a suitable calculation formula. For example, for the armor units on the slope of a rubble mound breakwater, the required mass was calculated in the past by Hudson’s formula with an appropriate coefficient (KD value), but recently it has become common to use Hudson’s formula with a stability number. The latter is more general in that it can also be applied to other cases, such as the armor units on the mound of a composite breakwater. [Technical Notes] (1) Hudson’s Formula The required mass of armor units on a slope can be expressed using the Hudson formula with a stability number (this is also referred to as the generalized Hudson formula) 24) (see equation (5.3.1)). (2) Stability Number and Nominal Diameter The stability number directly corresponds to the necessary size (nominal diameter) of the armor stones or concrete blocks for a given wave height. In other words, by introducing the nominal diameter Dn = (M/rr)1/3 and the term D = Sr - 1 and substituting them into equation (5.3.1), the following relatively simple equation is obtained: H/(DDn) = NS (5.3.2) It can be seen that the nominal diameter is proportional to the wave height with the constant of proportionality being 1/DNS. (3) Design Wave Height The Hudson formula was proposed based on the results of experiments that used regular waves. When applying it to the action of actual waves (which are irregular), there is thus a problem of which defcinition of wave heights should be used. However, with structures that are made of rubble stones or concrete blocks, there is a tendency for damage to occur not when one single wave having the maximum height H among an irregular wave train attacks the armor layer, but rather for damage to progress gradually under the continuous action of waves of various heights. Considering this fact and past experiences, it has been decided to make it standard to use the significant wave height of incident waves at the place where the slope is located as the wave height H in equation (5.3.1), because the significant wave height is representative of the overall scale of an irregular wave train. Consequently, it is also standard to use the significant wave height when using the generalized Hudson formula. Note however that for places where the water depth is less than one half the equivalent deepwater wave height, the significant wave height at the water depth equal to one half the equivalent deepwater wave height should be used. -112-


(4) Parameters Affecting the Stability Number As shown in equation (5.3.1), the required mass of armor stones or concrete blocks varies with the wave height and the density of the armor units, and also the stability number NS. The NS value is a coefficient that represents the effects of the characteristics of structure, those of armor units, wave characteristics and other factors on the stability. The main factors that influence the NS value are as follows. (a) Characteristics of the structure ① Type of structure (rubble mound breakwater, breakwater covered with wave-dissipating concrete blocks,

composite breakwater, etc.) ② Gradient of the armored slope ③ Position of armor units (breakwater head, breakwater trunk, position relative to still water level, front face

and top of slope, back face, berm, etc.) ④ Crown height and width, and shape of superstructure ⑤ Inner layer (its coefficient of permeability, thickness, and degree of surface roughness)

(b) Characteristics of the armor units ① Shape of armor units (shape of armor stones or concrete blocks; for armor stones, their diameter

distribution) ② Placement of armor units (number of layers, regular laying or random placement, etc.) ③ Strength of armor material

(c) Wave characteristics ① Number of waves acting on armor layers ② Wave steepness ③ Form of sea bottom (bottom slope, existence of reef, etc.) ④ Ratio of wave height to water depth (as indices of nonbreaking or breaking wave condition, breaker type,

etc.) ⑤ Wave direction, wave spectrum, wave grouping characteristics

(d) Extent of damage (damage rate, damage level, relative damage) Consequently, the NS value used in design must be determined appropriately based on hydraulic model experiments in line with the respective design conditions. By comparing the results of regular wave experiments with those of irregular wave experiments, it was found that the ratio of the height of regular waves to the significant height of irregular waves that gave the same damage ratio (within the error of 10%) varied in the range of 1.0 to 2.0 (depending on the conditions). In other words, there was a tendency for the irregular wave action to be more destructive than the action of regular waves. It is thus better to employ irregular waves in experiments. (5) Stability number NS and KD value In 1959, Hudson published the so-called Hudson formula 24), replacing the previous Iribarren-Hudson formula. Hudson developed equation (5.3.1) by himself using K D cot a instead of NS, i.e. N S3 = K D cot a

(5.3.3)

where a： angle of the slope from the horizontal line (º) KD： constant determined primarily by the shape of the armor units and the damage ratio The Hudson formula was based on the results of a wide range of model experiments and has proved itself well in usage in the prototype design. In the past, this formula (i.e., the one using the KD value) has thus been used in the calculation of the required mass of armor units on a slope. However, the generalized Hudson formula that uses the stability number (equation (5.3.1)) has been used for quite a while for calculating the required mass of armor units on the mound of a composite breakwater (to be discussed later), and is also used for the armor units of other structures such as submerged breakwaters. It is thus now more commonly used than the old formula with the KD value, and so the generalized Hudson formula with the stability number can be considered as being the standard equation for calculating the required mass of armor units on a slope. The stability number NS can be derived from the KD value and the angle a of the slope from the horizontal line by using equation (5.3.3.) There is no problem with this process if the KD value is an established one and the slope angle is within a range of normal design. However, most of the KD values obtained up to the present time have not sufficiently incorporated various factors like the characteristics of the structure and the waves. Thus, this method of determining the stability number NS from the KD value cannot be guaranteed to yield economical design always. In order to calculate more reasonable values for the required mass, it is thus desirable to use the results of experiments matched to the conditions in question, or else to use calculation formulas (calculation diagrams) that include the various relevant factors as described below.

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(6) Van der Meer’s Formula for Armor Stones In 1987, van der Meer carried out systematic experiments concerning the armor stones on the slope of a rubble mound breakwater with a high crown. He proposed the following calculation formula for the stability number, which considers not only the slope gradient, but also the wave steepness, the number of waves, and the damage level 25). Note however that the following formula has been slightly altered in comparison with van der Meer’s original one in order to make calculations easier. For example, the wave height H2% for which the probability of exceedance is 2% has been replaced by H1/20. NS = max {Nspl, Nssr} Nspl = 6.2CHP 0.18 (S 0.2 / N 0.1) Ir- 0.5 Nssr = CHP-0.13 (S 0.2 / N 0.1) (cota)0.5IrP where Nspl： Nssr： Ir： Som： L0： T1/3： CH： H1/3： H1/20： a： Dn50： M50： P： S： A： N：

(5.3.4) (5.3.5) (5.3.6)

stability number for plunging breakers stability number for surging breaker Iribarren number (tan a/Som0.5) (also called the surf similarity parameter) wave steepness (H1/3/L0) deepwater wavelength (L0 = gT1/32/2p, g = 9.81m/s2) significant wave period modification factor due to wave breaking [=1.4 / (H1/20/H1/3) ] (=1.0 in the region where wave breaking does not occur) significant wave height highest one-twentieth wave height (see Fig. T- 5.3.1) angle of slope from the horizontal line (º) nominal diameter of armor stone (=(M50/rr)1/3) 50% value of the mass distribution curve of an armor stone (required mass of an armor stone) permeability coefficient of the inner layer (see Fig. T- 5.3.2) deformation level (S = A / Dn502) (see Table T- 5.3.1) erosion area of cross section (see Fig. T- 5.3.3) number of waves (in storm duration)

The wave height H1/20 in Fig. T- 5.3.1 is for a point at a distance 5H1/3 from the breakwater, and H0’ is the equivalent deepwater wave height. The deformation level S is an index that represents the amount of deformation of the armor stones, and it is a kind of damage ratio. It is defined as the result of the area A eroded by waves (see Fig. T- 5.3.3) being divided by the square of the nominal diameter Dn50 of the armor stones. As shown in Table T- 5.3.1, three stages are defined with regard to the deformation level of the armor stones: initial damage, intermediate damage, and failure. With the standard design, it is common to use the deformation level for initial damage for N = 1000 waves. However, with design where a certain amount of deformation is permitted, usage of the value for intermediate damage can also be envisaged. Table T- 5.3.1 Deformation Level S for Each Failure Stage for a Two-layered Armor Slope

Initial damage

Intermediate damage

1 : 1.5 1:2 1:3 1:4 1:6

2 2 2 3 3

3～5 4～6 6～9 8 ～12 8 ～12

Failure 8 8 12 17 17

(7) Stability Number for Armor Concrete Units of Rubble Mound Breakwater Van der Meer has carried out model experiments on several kinds of precast concrete blocks, and proposed the formulas for calculating the stability number NS 26). In addition, other people are also proceeding with research into establishing calculation formulas for precast concrete blocks. For example, Burcharth and Liu 27) have proposed a calculation formula. However, it should be noted that these are based on the results of experiments for a rubble mound breakwater with a high crown. (8) Stability Number for Concrete Units of the Wave-Dissipating Block Mound in Front of Upright Walls (horizontallycomposite breakwater) The wave-dissipating concrete block mound of a horizontally-composite breakwater may have various crosssectional forms. In particular, when all the front face of an upright wall is covered by wave-dissipating concrete blocks, the stability is higher than for the normal case of armor concrete units covering a rubble mound breakwater because the permeability is high. In Japan, much research has been carried out on the stability of breakwaters covered with wave-dissipating concrete blocks. For example, Tanimoto et al. 28), Kajima et al., and Hanzawa et al. have carried out systematic research on the stability of wave-absorbing concrete blocks. In addition, Takahashi et al. 29) have proposed the following equation for wave-dissipating concrete blocks that are randomly placed in the mound covering the whole of upright wall. -114-


Sea Bottom slope

1/100

Sea Bottom slope

1/50

Sea Bottom slope

1/30

H0′/L0 H0′ : Equivalent deepwater wave height

h/H0¢ Fig. T- 5.3.1 Ratio of H1/20 to H1/3 (H1/20 Values are at a Distance 5H1/3 from the Breakwater)

able erme

er or lay

layer

Imp

Arm

Filter layer

yer or la Arm ter layer Fil

Area of eroded part r r laye Armo Core

Fig. T- 5.3.3 Erosion Area A

r

r laye

Armo

No filter, no core

Nominal diameter of armor stones Nominal diameter of filter material Nominal diameter of core material

Fig. T- 5.3.2 Permeability Coefficient P

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N S = C H { a ( N 0 ¤ N 0.5 ) 0.2 + b }

(5.3.7)

where N0： relative damage (a kind of damage ratio that represents the extent of damage: it is defined as the number of concrete blocks that have moved within a width Dn in the direction of the breakwater alignment, where Dn is the nominal diameter of the concrete blocks: Dn = (M/rr)1/3, where M is the mass of a concrete block) CH： modification factor due to wave breaking; CH = 1.4 / (H1/20 / H1/3) (in the region where wave breaking does not occur, H1/20 / H1/3 = 1.4, and so CH = 1.0) a, b： coefficients that depend on the shape of the concrete blocks and the slope angle (for concrete blocks with the KD value of 8.3, a = 2.32 and b = 1.33, if cot a = 4/3; a = 2.32 and b = 1.42, if cot a = 1.5) Takahashi et al. 29) have further presented a method for calculating the cumulative relative damage (the expected relative damage) over the lifetime of a breakwater. In the future, reliability design methods that consider the expected relative damage will become important in the advanced design methodology. In the region where wave breaking does not occur, if the number of waves is 1000 and the relative damage N0 is 0.3, the design mass as calculated using the method of Takahashi et al. is more-or-less the same as that calculated using the KD value in the past. The value of N0 = 0.3 corresponds to the conventionally-used damage ratio of 1%. (9) Breakwater Head Waves attack the head of a breakwater from a whole angle of directions, and there is a greater risk of the armor units on the top of the slope falling not so much forward but rather toward the rear side. Stones or concrete blocks to be used at the head of a breakwater must thus have a mass greater than the value given by equation (5.3.1). Hudson suggested to raise the mass by 10% in the case of stones and 30% in the case of concrete blocks. However, it is thought to be insufficient. It would be desirable to use the mass at least 1.5 times the value given by equation (5.3.1) for both stones and concrete blocks. (10) Submerged Armor Units Since the action of waves on a rubble mound breakwater is weaker midwater than around the waterline, stones or concrete blocks of reduced mass may be used at depths more than 1.5H1/3 below the still water level. (11) Effect of Wave Direction The extent to which the incident wave angle affects the stability of the armor stones has not been investigated sufficiently. Nevertheless, according to the results of experiments carried out by van de Kreeke 30) in which the wave angle was changed between 0º (i.e., direction of incidence is perpendicular to the breakwater alignment), 30º, 45º, 60º and 90º, the damage ratio for a wave direction of 45º or smaller is more-or-less the same as that when the wave direction is 0º; when the wave direction is more than 60º, the damage ratio drops. Based on these results, it is considered that when the wave angle b (see Fig. T- 5.2.2 in 5.2.2 [1] Wave Force under Wave Crest) is 45º or less, the minimum mass should not be corrected for wave direction. Moreover, Christensen et al. 31) have shown that the stability increases when the directional spreading of random waves is large. (12) Integrity of Concrete Blocks With a precast concrete block, it is necessary not only to ensure that the block has a mass sufficient to be stable against the design waves, but also to confirm that the block itself has sufficient structural strength. (13) Armor Units in Reef Area In general, a reef rises up at a steep slope from the relatively deep sea, and forms a relatively flat and shallow sea bottom. Consequently, when a large wave arrives at such a reef, it breaks around the tip of the reef, and then the regenerated waves propagate over the reef in the form of surge. The characteristics of waves over a reef are strongly dependent on not only the incident wave conditions but also the water depth over the reef and the distance from the tip of the reef. The stability of wave-dissipating concrete blocks situated on a reef also varies greatly for the same reasons, making the situation more complicated than that in general cases. The stability of wave-dissipating concrete blocks situated on a reef must thus be investigated based either on model experiments matching the conditions in question or on field experiences for sites having similar conditions. (14) Armor Units of Low Crest Breakwater For a rubble mound breakwater with a low crown, it is necessary to note that the concrete blocks around its crown (in particular on the shoreward side) are easily damaged. For example, for detached breakwater composed of wave-dissipating concrete blocks, unlike a caisson breakwater covered with wave-dissipating concrete blocks, there is no supporting wall at the back and the crown is not high. This means that the concrete blocks near the crown (in particular at the rear) are easily damaged, and indeed such cases of block damage have been reported. (15) Effect of Steep Slope Bed When the bottom slope is steep and waves break in plunging form, a large wave force may act on concrete blocks, subject to their shapes. It is thus necessary to carry out appropriate investigations while considering the possibility of large wave force (see Takeda et al.).

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(16) High-density Blocks The minimum mass of blocks that are made of high-density aggregate may also be determined using the Hudson formula with the stability number (equation (5.3.1)). As shown by the equation, high-density blocks have a high stability, so a stable armor layer can be made using relatively small blocks of high density. (17) Effect of Placement The stability of wave-dissipating concrete blocks also varies with the method of placement (random placement or regular placement etc.). According to the results of experiments carried out to compare the random placement over the whole cross section and that of regular two-layer placement upon a stone core, the stability of the regular placement with well-interlocking was markedly improved for most of the cases tested. Moreover, the stability is also affected by the crown height and width of the mound of wave-dissipating concrete blocks. According to the results of a number of experiments, for example, there is a tendency of greater stability when the crown is high and wide. (18) Standard Method of Hydraulic Model Experiments The stability of concrete blocks is influenced by a very large number of factors, and so it has still not been sufficiently elucidated. This means that it is necessary to carry out studies using model experiments for the design of prototype breakwaters, and it is needed to progressively accumulate the results of such experiments. The following points should be noted when carrying out studies using model experiments. (a) It is standard to carry out experiments using irregular waves. (b) For each particular set of conditions, the experiment should be repeated at least three times (i.e., with three different wave trains). However, when experiments are carried out by systematically varying the mass and other factors and a large amount of data can be acquired, one run for each test condition will suffice. (c) It is standard to study the action of 1000 waves in total of three runs for each wave height level. Even for the systematic experiments, it is desirable to apply more than 500 waves or so. (d) For the description of the extent of damage, in addition to the damage ratio which has been commonly used in the past, the deformation level or the degree of damage may also be used. The deformation level is suitable when it is difficult to count the number of armor stones or concrete blocks that have moved, while the degree of damage is suitable when one wishes to represent the damage to wave-dissipating concrete blocks. The damage ratio is the ratio of the number of damaged armor units in an inspection area to the total number of armor units in the same inspection area. The inspection area is taken from the elevation of wave runup to the depth of 1.5H below the still water level or to the bottom elevation of the armor layer (take a shallower depth), where the wave height H is inversely derived from the Hudson formula with the mass of armor units as the input. However, for the deformation level and the degree of damage, there is no need to define the inspection area. For evaluating the damage ratio, an armor unit is judged to be damaged if it has moved over a distance of more than about 1/2 to 1.0 times its height.

5.3.2 Armor Units on Foundation Mound of Composite Breakwater (Notification Article 48, Clause 5) It shall be standard to calculate the mass of armor stones or concrete blocks for the foundation mound of a composite breakwater, by means of appropriate hydraulic model experiments or the following equation:

rr H3 M = --------------------------3 NS ( S r 1 ) 3 where M： rr： H： NS ： Sr：

(5.3.1)

minimum mass of rubble stones or concrete blocks (t) density of rubble stones or concrete blocks (t/m3) wave height used in the stability calculation (m) stability number specific gravity of rubble stones or concrete blocks relative to sea water

[Commentary] The mass required for an armor unit covering the foundation mound of a composite breakwater varies according to the wave characteristics, the water depth, the shape of the mound (thickness, berm width, slope angle, etc.), and the type of armor unit, its placement method, and its position (breakwater head, breakwater trunk, etc.). In particular, the effects of the wave characteristics and the mound shape are more pronounced than those in the case of the armor units covering the surface of sloped breakwater in 5.3.1 Armor Units on Slope. It is thus necessary to appropriately determine the mass, considering the results of past studies, research, and actual experience in the field, and carrying out model experiments if necessary. Moreover, it is necessary to take sufficient heed of the effects of wave irregularity.

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Note however that the stability of the armor units covering the foundation mound of a composite breakwater is not necessarily determined purely by their sizes. Depending on the structure and the layout of armor units, it may be possible to achieve stability even when the armor units are relatively small. [Technical Notes] (1) Generalized Hudson’s Formula for Calculating the Required Mass Similarly with the stable mass of armor units on a slope, the required mass of armor units covering the foundation mound of a composite breakwater can be calculated using the generalized Hudson formula (the Hudson formula with the stability number), i.e., equation (5.3.1). Ever since Brebner and Donnelly 32) used it as the basic equation for calculating the required mass of the armor stones of the rubble mound for an upright wall, the generalized Hudson formula has been used widely, and in Japan it is also known as the Brebner-Donnelly formula. Because it has a certain degree of validity even from a theoretical standpoint, the generalized Hudson formula may also be used as the basic formula for calculating the minimum mass of armor units for the breakwater mound 33). Note however that the stability number NS varies not only with the water depth, the wave characteristics, the shape of the mound, and the characteristics of the armor units, but also with their position of the placement (breakwater trunk, breakwater head, etc.). It is thus necessary to assign the stability number NS appropriately through model experiments corresponding to the design conditions. Moreover, the wave height used in the design calculation is generally the significant wave height, and the waves used in the model experiments should be irregular. (2) Stability Number for Armor Stones The stability number NS can be determined using the method of Inagaki and Katayama 34), which is based upon the work of Brebner and Donnelly and past experience of damage. However, the following formulas by Tanimoto et al. 33) are based on the flow velocity near the mound and allow the incorporation of a variety of conditions, and they have been extended by Takahashi, Kimura, and Tanimoto 35) to include the effects of wave direction. The extended Tanimoto formulas have thus been made the standard formulas. (a) Extended Tanimoto formulas ì 1 k h¢ ( 1 k ) 2 h¢ ü - ------------ + 1.8 exp 1.5 ------------------ ------------ ý : B M ¤ L¢ < 0.25 N S = max í 1.8, 1.3 ----------1 / 3 H k k 1 / 3 H1 ¤ 3 þ 1¤3 î

(5.3.8)

k = k 1 ( k 2 )B

(5.3.9)

4ph¢ ¤ L¢ k 1 = -----------------------------------sinh ( 4ph¢ ¤ L¢ ) ( k 2 )B = max { a s sin 2 b cos 2 ( 2p l cos b ¤ L¢ ), cos 2 b sin 2 ( 2p l cos b ¤ L¢ ) } s

a

(5.3.10) (5.3.11)

where h¢： water depth on top of rubble mound foundation (excluding the armor layer) (m) (see Fig. T- 5.3.4) l： in the case of normal wave incidence, the berm width BM (m) In the case of oblique wave incidence, either BM or BM¢, whichever gives the larger value of (k2)B (see Fig. T- 5.3.4) L： wavelength corresponding to the design significant wave period at the water depth h¢ (m) as： correction factor for when the armor layer is horizontal (= 0.45) b： incident wave angle (see Fig. T- 5.3.5) H1/3： design significant wave height (m) The validity of the above formulas have been verified for the breakwater trunk for oblique wave incidence with an angle of incidence of up to 60º. Seaward

Shoreward BM

d h'

BM '

Foot protection blocks

Upright section

h

C

Foot protection blocks

h

Armor material Rubble mound

Armor material

Fig. T- 5.3.4 Standard Cross Section of a Composite Breakwater and Notations

(b) Stability Number When a Certain Amount of Damage is Permitted Sudo et al. have carried out stability experiments for the special case such that the mound is low and no wave breaking occurs. They investigated the relationship between the number of waves N and the damage ratio, and proposed the following equation that gives the stability number NS* for any given number of waves N and any given damage ratio DN (%). -118-


N S * = N S [ D N ¤ exp { 0.3 ( 1 500 ¤ N ) } ] 0.25 (5.3.12) where NS is the stability number given by the Tanimoto formula when N = 500 and the damage ratio is 1%. In design it is necessary to take N = 1000 considering the progress of damage, while the damage ratio 3% to 5% can be allowed for a 2-layer armoring. If N = 1000 and DN = 5%, then NS* = 1.44NS. This means that the required mass decreases to about 1/3 of that required for N = 500 and DN = 1%. (3) Stability Number for Concrete Units The stability number NS for concrete blocks varies according to the shape of the block and the method of placement. It is thus desirable to evaluate the stability number by means of hydraulic model experiments. When carrying out such experiments, it is best to employ irregular waves.

nk tru ate r Br eak w

(4) Conditions for Applying the Stability Number for Armor Stones on Foundation Mound When the water above the armor units covering a mound is shallow, wave breaking often causes the armor stones to become unstable. It is thus appropriate to use the stability number for armor stones on a mound only when h¢/H1/3 ≧ 1: when h¢/H1/3 10 4 , it is acceptable to take z ≒ 20.

Ud R e = ------n Fig. T- 6.6.3 Relationship between Resistance Coefficient and Reynolds Number

[References] 1) IPCC: “Climate Change 1995”, IPCC Second Assessment Report, The Science of Climate Change, 1995, 572p. 2) Toshihiko NAGAI, Kazuteru SUGAHARA, Hiroshi WATANABE, Koji KAWAGUCHI: “Long team observation of the mean tide level and lond waves at the Kurihama-Bay”, Rept of PHRI, Vol. 35, No. 4, 1996. (in Japanese). 3) Katsutoshi TANIMOTO, Tomotsuka TAKAYAMA, Kazuo MURAKAMI, Shigeru MURATA, Hiroiti TSURUYA, Shigeo TAKAHASHI, Masayuki MORIKAWA, Yasutoshi YOSHIMOTO, Susumu NAKANO, Tetsuya HIRAISHI: “Field and laboratory investigations of the tsunami caused by 1983 Nihonkai Chubu Earthquake”, Tech. Note of PHRI, No. 470, 1983, 299p. (in Japanese). 4) Chiaki GOTO, Kazuo SATO: “Development of tsunami numerical simulation system for Sanriku Coast in Japan”, Rept of PHRI, Vol. 32, No. 2, 1995. (in Japanese). 5) Tomotsuka TAKAYAMA, Tetsuya HIRAISHI: “Hydraulic model tests on tsunamis at Suzaki Port”, Tech. Note of PHRI, No. 549, 1986, 131p. (in Japanese). 6) Tomotsuka TAKAYAMA, Norihiro NAGAI, Tetsuya HIRAISHI: “The numerical calculation of tsunami in Tokyo Bay”, Tech. Note of PHRI, No. 454, 1986, 131p.(in Japanese). 7) Toshihoko NAGAI, Noriaki HASHIMOTO, Tetsuya HIRAISHI, Katsuyoshi SHIMIZU: “Characteristics of the HokkaidoEast-off-Earthquake Tsunami”, Tech. Note of PHRI, No. 802, 1995, 97p. (in Japanese). 8) Koji KOBUNE, Toshihiko NAGAI, Noriaki HASHIMOTO, Tetsuya HIRAISHI, Katsuyoshi SHIMIZU “Characteristics of the Irianjaya Earthquake Tsunami in 1996”, Tech. Note of PHRI, No. 842, 1996, 96p. (in Japanese).

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PART II DESIGN CONDITIONS 9) Tomotsuka TAKAYAMA, Tetsuya HIRAISHI: “Amplification mechanism of harbor oscillation derived from field observation and numerical simulation”, Tech. Note of PHRI, No. 636, 1988, 70p. (in Japanese). 10) Honda, K., T. Terada, and D. Ishitani: “Secondary undulation of oceanic tides”, Philosophical Magazine, Vol.15,1908, pp.88126. 11) Ippen, A.T. and Y. Goda: “Wave-induced oscillations in harbors: the solution for a rectangular harbor connected to the open sea,” M.I.T. Hydrodynamics Lab. Report No.59, 1963, 90p. 12) Todd, D. K.: “Groundwater Hydrology”, John Wiley & Sons, Inc., 1963. 13) Yoshihiro SHOJI, Masaharu KUMEDA, Yukiharu TOMITA: “Experiments on seepage through interlocking joints of sheet pile”, Rept of PHRI, Vol. 21, No. 4, 1982, pp. 41-82 (in Japanese).

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Chapter 7 Currents and Current Force 7.1 General (1) The current parameters that shall be used in the design of port and harbor facilities are the velocity and the direction. The severest conditions shall be set, based on either the field measurements at the installation location of the facilities in question or the numerical estimation. (2) For the current force, consideration shall be given to the drag and lift, depending on the type of the facilities in question and the structural form. [Commentary] For structures that are located in a place where there is strong currents such as a tidal currents or river flow, it is necessary to carry out investigations on the forces produced by the currents with the largest velocity from the most unfavorable direction. Depending on the type of structures or members, it may also be necessary to consider the vertical distribution of the current velocity. When waves coexist with currents, it is necessary to use the current velocity and direction in the state of coexistence. Types of currents in the sea area include ocean currents, tidal currents, and wind drift currents, which are described in the [Technical Notes] below, along with density currents caused by the density differences due to salinity or water temperature. In addition, in the coastal area, there are longshore currents and rip currents caused by waves. [Technical Notes] (1) Ocean Currents Ocean currents are the phenomenon involving the circulation of seawater around the ocean as a whole. They are the result of a combination of the following currents: a) density currents that are based on local differences in the density of seawater, b) wind-driven drift currents that are caused by the wind, and c) gradient currents that accompany spatial inequalities in the atmospheric pressure, along with d) compensation currents (upwelling currents and or sinking currents) that supplement the aforementioned currents. Ocean currents maintain the almost steady direction and strength over prolonged periods of time. (2) Tidal Currents (a) The nature and strength of tidal currents vary with the geographical conditions of the sea area in question and the celestial movements. In order to analyze the harmonic components of tidal currents, it is necessary to carry out continuous observation for at least 25 hours or advisably for full 15 days. In particular, if the topography of a place is going to be changed considerably, for example when carrying out large-scale land reclamation in shallow coastal waters, it is desirable to examine the resultant changes in tidal currents at the planning stage. (b) The tidal currents are the flow of seawater in the horizontal direction that accompanies a tidal variation of sea level. This variation consists of the tidal components (diurnal tide, semi-diurnal tide, etc.) of the water level and is thus periodic. (3) Wind-Driven Currents When a wind blows over the sea surface, the friction on the boundary between the air and the sea surface produces a shear stress that causes to induce a flow on the sea surface. As this flow develops, the turbulent eddy viscosity of the seawater causes the lower layers to start to be pulled along by the upper layers. If the wind velocity and direction remain constant for a prolonged period of time, a steady state of currents is eventually reached. Such the currents are referred to as the wind-driven currents. (4) Nearshore Currents In the surf zone, there exist special currents called the nearshore currents induced by waves. Because the nearshore currents are induced within the surf zone, they transport suspended sediments and cause topographical change of beaches. Consequently, an understanding of the pattern of nearshore currents leads to a deeper perception of topographical change.

7.2 Current Forces Acting on Submerged Members and Structures (Notification Article 7) It shall be standard to calculate the drag and lift forces caused by currents acting on a member or a structure that is submerged or near the water surface using the following equations: (1) Drag Force 1 F D = --- C D r 0 AU 2 2

(7.2.1)

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where F D： CD： r0： A： U：

drag force acting on the object in the direction of the current (kN) drag coefficient density of water (t/m3) projected area of the object in the direction of the current (m2) flow velocity (m/s)

(2) Lift Force 1 F L = --- C L r 0 A L U 2 2 where FL： lift force acting on the object in the direction perpendicular to the current (kN) CL： lift coefficient AL： projected area of the object in the direction perpendicular to the current (m2)

(7.2.2)

[Commentary] The fluid force due to the currents acting on members of a pile-supported structure such as a pier, a pipeline, or the armor units of a mound is proportional to the square of the flow velocity. It may be divided into the drag force acting in the direction of the current and the lift force acting in the direction perpendicular to this. Note also that a thin, tubelike object in the water may be subject to vibrations excited by current-induced vortices. [Technical Notes] (1) Drag Coefficient The drag to a submerged object due to currents is expressed as the sum of the surface resistance due to friction and the form drag due to pressure difference around the object. The drag coefficient varies according to the shape of the object, the roughness, the direction of the current, and the Reynolds number, and thus the value appropriate to the conditions in question must be used. When the Reynolds number is greater than about 103, the values listed in Table T- 7.2.1 may be used as standard values for the drag coefficient. Note that for a circular cylinder or sphere with a smooth surface, there is a phenomenon whereby the value of the drag coefficient drops suddenly when the Reynolds number is around 105. However, for a circular cylinder with a rough surface, this drop in drag coefficient is not particularly large, and the drag coefficient settles down to a constant value that depends on the relative roughness. For the values of the drag coefficient when a prism or L-shaped member is oriented diagonally relative to the current, search for references. The data for the cube have been obtained from wave force experiments carried out by Hamada, Mitsuyasu and Hase. Table T- 7.2.1 Drag Coefficients

Shape Circular cylinder (rough surface)

Projected area

Drag coefficient

D

1.0

D

B

2.0

B

πD2 4

1.2

ab

a /b = 1 a /b = 2 a /b = 4 a/b= 10 a/b= 18 a /b =

πD2 4

0.5

0.2

D2

1.3

1.6

D

Rectangular prism

B

Circular disc

D

Flat plate

b

a

Sphere

D

Cube

D

D

D

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1.12 1.15 1.19 1.29 1.40 2.01


(2) Lift Coefficient As with the drag coefficient, the lift coefficient varies with the shape of the object, the direction of the current, and the Reynolds number. However, the lift coefficient is not well understood (see 5.4.1 Wave Force on Submerged Members). (3) Current Force Acting on Submerged Breakwater As for the force acting on the coping of the submerged section at the opening of tsunami protection breakwater, Iwasaki et al. have measured the pressure on the coping due to the currents. They obtained the values of 0.94 for the drag coefficient and 0.48 for the lift force coefficient. Tanimoto et al. have carried out similar measurements, obtaining the values 1.0 to 1.5 for the drag coefficient and 0.5 to 0.8 for the lift coefficient. They have also pointed out, however, that when the flow velocity in the breakwater opening is large, the presence of the water surface gradient causes the coefficient values to increase.

7.3 Mass of Armor Stones and Concrete Blocks against Currents (Notification Article 48, Clause 6) It shall be standard to calculate the required mass for the armor units (rubble etc.) on a rubble mound to be stable against currents by means of either appropriate hydraulic model experiments or else the following equation: prr U 6 M = ------------------------------------------------------------------------------(7.3.1) ( 48 )g 3 y 6 ( S r 1 ) 3 ( cos q sin q ) 3 where M： rr： U： g： y： Sr： q：

minimum mass of armor stones and blocks (t) density of armor stones and blocks (t/m3) current velocity above armor stones and blocks (m/s) gravitational acceleration (= 9.81 m/s2) Isbash’s constant (1.20 for embedded stones; 0.86 for exposed stones) specific gravity of armor stones and blocks relative to water slope angle in the axial direction of the channel bed (º)

[Technical Notes] (1) Isbash’s Equation With regard to the mass of rubble stone that is stable against currents, the US Army Coastal Engineering Research Center (CERC) has presented equation (7.3.1) for the mass that a rubble stone must have in order to prevent scouring by tidal currents 8). (2) Isbash’s Constant Equation (7.3.1) has been derived by considering the balance between the drag caused by a flow acting on a spherical object on a sloped surface and the frictional resistance of the object. The coefficient y is termed Isbash’s constant. It would appear that the values of 1.20 and 0.86 for embedded stones and exposed stones, respectively, were determined by Isbash, but the details were not documented. Since equation (7.3.1) has been obtained by considering the balance of forces for steady flow, for places where it is anticipated that strong vortices will be generated, it is necessary to use rubble stones of larger mass. (3) Armor Units for the Mound at the Opening of Tsunami Protection Breakwaters Iwasaki et al. have carried out two-dimensional steady flow experiments in which they used precast concrete blocks as the armor for the mound in the opening of breakwaters designed to protect harbors and coastal area against tsunamis. They obtained a value of 1.08 for Isbash’s constant in equation (7.3.1). Tanimoto et al. have carried out three-dimensional experiments on the opening of a tsunami breakwater. They clarified the structure of the three-dimensional flow near the opening, and revealed the relationship between the damage ratio and Isbash's constant when stones or precast concrete blocks were used as the covering material. [References] 1) Kazuo MURAKAMI, Masayuki MORIKAWA, Tatsuya SAKAGUCHI: “Wind effect and water discharge effect on constant flow - discussion using observation data at off-Sennan (1978-1981) -”, Rept of PHRI, Vol. 21, No. 4, 1982, pp. 3-39 (in Japanese). 2) Masch, F. D.: “Mixing and dispersion of wastes by wind and wave action”, ‘Advances in Water Pollution Research,’ Proc. Int. Conf., Vol. 3, 1962, pp. 145-168. 3) Longuet-Higgins, M.S. and R.W. Stewart: “Radiation stress and mass transport in gravity waves, with application to ‘surf beat’”, J. Fluid Mech., Vol. 13, 1962, pp. 481-504. 4) Bowen, A. J., D. L. Inman, and V. P. Simons: “Wave ‘set-down’ and ‘set-up’”, J. Geophs. Res. Vol. 73, 1968, pp. 2569-2577. 5) Kazumasa KATOH, Shin-ichi YANAGISHIMA, Tomoyoshi ISOGAMI, Hiroyuki MURAKAMI: “Wave set-up near the shoreline - field observation at HORF -”, Rept of PHRI, Vol. 28, No. 1, 1989, pp. 3-41 (in Japanese).

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6) Yoshimi GODA: “Deformation of irregular waves due to depth-controlled wave breaking”, Rept of PHRI, Vol. 14, No. 3, 1975, pp. 59-106 (in Japanese), also “Irregular wave deformation in the surf zone”, Coastal Engineering in Japan, JSCE, Vol.18, 1975, pp.13-26. 7) Katsutoshi TANIMOTO, Katsutoshi KIMURA, Keiji MIYAZAKI: “Study on stability of submerged disk at the opening section of tsunami protection breakwaters”, Rept of PHRI, Vol. 27, No. 4, 1988, pp. 93-121 (in Japanese). 8) Coastal Engineering Research Center: “Shore Protection Manual”, Vol. II, U.S. Army Corps of Engineering, 1977

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Chapter 8 External Forces Acting on Floating Body and Its Motions 8.1 General The motions of a floating body produced by external forces such as those due to winds, currents and waves, along with the mooring force, shall be given due consideration in design of the floating body and related facilities. [Commentary] (1) Floating Body In general, a floating body refers to a structure that is buoyant in water and its motions within a certain range is permitted during use. When designing a floating body, it is necessary to investigate both its functions that are going to be demanded and its safety. In general, the design conditions for the investigation of its function differ from those for the examination of its safety. (2) Mooring Equipment Mooring equipment comes in a whole variety of types and is generally composed of a combination of mooring lines, mooring anchors, sinkers, intermediate weights, intermediate buoys, mooring rods, connection joints, and fenders. The mooring equipment has a large influence on the motions of a floating body, and so it is important to design this equipment safely and appropriately. [Technical Notes] (1) Classification of Floating Bodies The floating bodies used as port and harbor facilities can be divided into floating terminals, offshore petroleum stockpiling bases, floating breakwaters, mooring buoys, and floating bridges. Moreover, researches for development of extra large floating structures (mega-float) are being carried out. (2) Classification of Mooring Methods and Characteristic Features of Each Method Floating bodies can also be classified by the type of mooring methods. As described below, mooring methods include catenary mooring (slack mooring), taut mooring, and dolphin mooring. (a) Catenary mooring (Fig. T- 8.1.1(a)) This is the most common mooring method. With this method, the chains or whatever used in the mooring are given sufficient lengths to make them slack. This means that the force restraining the motions of the floating body is small, but nevertheless the mooring system fulfills the function of keeping the floating body in moreor-less the same position. There are various types of catenary mooring, depending on factors like the material of the mooring lines, the number of mooring lines, and the presence or absence of intermediate buoys and sinkers. (b) Taut mooring (Fig. T- 8.1.1(b)) This is a mooring method that reduces the motions of the floating body greatly; a tension leg platform (TLP) is an example. With this method, the mooring lines are given a large initial tension so that they do not become slack even when the floating body moves. The advantages of this mooring method are that the floating body does not move much, and only a small area is needed for installing the mooring lines. However, it is necessary to take note of the fact that because a large tensile force is generated in the mooring lines, the design of the lines becomes the critical factor on the safety of the floating body. (c) Dolphin mooring (Fig. T- 8.1.1(c)) With this method, mooring is maintained using either a pile-type dolphin or a gravity-type dolphin. In general, this method is suitable for restraining the motions of a floating body in the horizontal direction, but a large mooring force acts on the dolphin. This method has been used for mooring floating units of offshore petroleum stockpiling bases. (d) Mooring method using a universal joint (Fig. T- 8.1.1(d)) The mooring system shown in the figure is an example of a mooring method that can be used to moor a large offshore floating body. Examples of mooring systems that use a universal joint on the sea bottom include a SALM (Single Anchor Leg Mooring) type mooring buoy and a MAFCO (MAritime Facility of Cylindrical cOnstruction) tower.

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Fender

Chain

Dolphin

Mooring anchor

(a) Catenary mooring

(c) Dolphin mooring

Damper

Universal joint

(b) Taut mooring

(d) Mooring by universal joint

Fig. T- 8.1.1 Examples of Mooring Methods for Floating Body

8.2 External Forces Acting on Floating Body (Notification Article 26, Clause 1) When a port or harbor facility is made of a floating structure, it shall be standard to take the following forces in design calculation: wind drag force, current drag force, wave-exciting force, wave-drift force, wave-making resistance, restoring force, and mooring force. These forces shall be calculated by means of an appropriate analytical method or hydraulic model experiments, in accordance with the mooring method for the floating body and the size of facility. [Technical Notes] (1) Wind Drag Force With a structure for which a part of the floating body is above the sea surface, winds exert a force on the structure. This force is called the wind drag force (or wind pressure), and is composed of a pressure drag and a friction drag. If the floating body is relatively small in size, the pressure drag is dominant. The pressure drag is proportional to the square of the wind velocity and is expressed as in the following equation: 1 (8.2.1) F w = --- r a C DW A W UW2 2 where Fw： wind drag force (N) ra： density of air (1.23 kg/m3) AW： projected area of the part of the floating body above the sea surface as viewed from the direction in which the wind is blowing (m2) UW： wind velocity (m/s) CDW： wind drag coefficient The wind drag coefficient is a proportionality constant and is also known as the wind pressure coefficient. It may be determined by means of wind tunnel experiments or the like. However, it is also acceptable to use a value that has been obtained in the past experiments for a structure with a shape similar to the structure under current study. Values such as those listed in Table T- 8.2.1 have been proposed as the wind drag coefficients of objects in the uniform flow. As can be seen from this table, the wind drag coefficient varies with the shape of the floating body, but it is also affected by the wind direction and the Reynolds number. Note that it is considered that the wind pressure acts in the direction of the wind flow, with the point of application being the centroid of the projection of the part of the floating body that is above the water surface. However, it is necessary to take heed of the fact that this may not necessarily be the case if the floating body is large. Moreover, the velocity of the actual wind is not uniform in the vertical direction, and so the value of the wind velocity UW used in the wind pressure calculation is set as that at the elevation of 10 m above the sea surface.

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Table T- 8.2.1 Wind Pressure Coefficient

Square cross-section

2.0 1.6

1 2

Rectangular cross-section 2 (ratio of side lengths = 1:2)

2.3 1.5

1

(when one face is in contact with the ground) Circular cross-section

(smooth surface)

1.2 1.2

(2) Current Drag Force When there is currents such as tidal currents, these currents will exert a force on the submerged part of the floating body. This force is referred to as the flow pressure or the current drag force. Like the wind drag force, it is proportional to the square of the flow velocity. Note however that since the velocity of the current is generally small, the current drag force is actually expressed as being proportional to the square of the velocity of the current relative to the velocity of motion of the floating body as in the following equation: 1 (8.2.2) F C = --- r 0 C DC A C | U C U | ( U C U ) 2 where FC： current drag force (N) r0： density of fluid (for seawater, 1030 kg/m3) AC： projected area of the submerged part of the floating body as viewed from the direction of the currents (m2) UC： velocity of the currents (m/s) U： velocity of motion of the floating body (m/s) CDC： drag coefficient with respect to the currents The drag coefficient CDC is a function of the Reynolds number. When the Reynolds number is large, however, the values for steady flow in Table T- 7.2.1 in 7.2 Current Forces Acting on Submerged Members and Structures may be used. The drag coefficient for the currents varies with the shape of the floating body and the direction of the currents. As with the wind pressure, the direction of the force exerted by the currents and the direction of the currents itself are not necessarily the same. In general, the deeper the draft of the floating body relative to the water depth, the larger the drag coefficient for the currents becomes. This is referred to as the shoaling effect, and the drag coefficient increases because the smaller the gap between the sea bottom and the base of the floating body, the harder it is for water to flow through this gap. (3) Wave-Exciting Force The wave-exciting force is the force exerted by incident waves on the floating body when the floating body is considered to be fixed in the water. It is composed of a linear force that is proportional to the amplitude of the incident waves and a nonlinear force that is proportional to the square of the amplitude of the incident waves. The linear force is the force that the floating body receives from the incident waves as reaction when the floating body deforms the incident waves. The velocity potential for the deformed wave motion is obtained using wave diffraction theory. The nonlinear force, on the other hand, is composed of a force that accompanies the finite amplitude nature of waves and a force that is proportional to the square of the flow velocity. The former force due to finite amplitude effect can be analyzed theoretically, but in practice it is often ignored. The latter force that is proportional to the square of the flow velocity becomes large, in particular when the diameter of the floating body is small relative to the wavelength; it is necessary to determine this force experimentally. (4) Wave-Drift Force When waves act on a floating body, the center of the floating body’s motion gradually shifts in the direction of wave propagation. The force that causes this shift is called the wave-drift force. If it is assumed that the floating body is two-dimensional and the wave energy is not dissipated, then the wave-drift force is given by the following equations: 1 (8.2.3) F d = --- r 0 gHi2R 8

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ì 4ph ¤ L ü R = K R2 í 1 + -------------------------------- ý ( 4ph ¤ L ) þ sinh î where r0： density of seawater (kg/m3) Fd： wave drift force per unit width (N/m) Hi： incident wave height (m) KR： reflection coefficient R： drift force coefficient

(8.2.4)

If the dimensions of the floating body are extremely small relative to the wavelength, the wave drift force may be ignored as being much smaller than the wave-exciting force. However, as the floating body becomes larger, the wave drift force becomes dominant. When irregular waves act on a floating body moored at a system having only a small restraining force, such as a single point mooring buoy designed for use of supertankers, the wavedrift force becomes a dominant factor as it may give rise to slow drift motions. (5) Wave-Making Resistance When a floating body moves in still water, the floating body exerts a force on the surrounding water, and the floating body receives a corresponding reaction force from the water; this reaction force is called the wavemaking resistance. This force may be determined by forcing the floating body to move through the still water and measuring the force acting on the floating body. In general, however, an analytical method is used whereby each mode of the floating body motions is assumed to be realized separately, and the velocity potential, which represents the motion of the fluid around the floating body, is obtained. Only the forces that are proportional to the motion of the floating body may be determined analytically; the nonlinear forces that are proportional to the square of the motion cannot be determined analytically. Out of the linear forces (i.e., that proportional to the motion of the floating body), the term that is proportional to the acceleration of the floating body is called the added mass term, while the term that is proportional to the velocity is called the wave damping term. (6) Restoring Force The static restoring force is the force that makes a floating body to return to its original position when the floating body moves in still water. It is generated by buoyancy and gravity, when the floating body heaves, rolls or pitches. This force is generally treated as being proportional to the amplitude of the motion of the floating body, although this proportionality is lost if the amplitude becomes too large. (7) Mooring Force The mooring force (restraining force) is the force that is generated in order to restrain the motion of the floating body. The magintude of this force depends greatly on the displacement-restoration characteristics of the mooring system. (8) Solution Method for Wave-Exciting Force and Wave-Making Resistance Using Velocity Potential The method adopted for calculating the wave-exciting force and the wave-making resistance involves deriviation of the velocity potential, which represents the motion of the fluid, and then calculating the waveexciting force and the wave-making resistance from the potential. The analytical method with the velocity potential is the same for both the wave-exciting force and the wave-making resistance, the only difference being the boundary conditions. The velocity potential may be obtained using any of a number of methods, such as a region segmentation method, an integral equation method, a strip method, or a finite element method. (9) Wave Force Acting on Fixed Floating Body with Rectangular Cross Section When a floating body is fixed in position, the velocity potential that satisfies the boundary conditions at the sea bottom and around the floating body can yield the wave force. The wave force acting on a floating body with a long rectangular cross section such as a floating breakwater can be determined using the approximation theory of Ito and Chiba 2). (10) Materials for Mooring For the materials used in mooring and their characteristic features, search for appropriate references. (11) Forces Acting on an Extra Large Floating Structure For an extra large floating structure (mega-float), the external forces described in (1) ~ (10) above are different from those for a smaller floating body, because of its large size and elastic response characteristics of the floating body structure. It is thus necessary to carry out sufficient investigations on the motions and elastic response characteristics of the floaty body structure.

8.3 Motions of Floating Body and Mooring Force (Notification Article 26, Clause 2) The motions of a floating body and the mooring force shall be calculated by means of an appropriate analytical method or hydraulic model experiments, in accordance with the shape of the floating body and the characteristics of the external forces and the mooring system.

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[Commentary] The motions of a floating body can be determined by Heaving solving the dynamic equilibrium equation, with the Yawing Rolling external forces taken to be the forces due to winds and Surging waves, the restoring force of the floating body itself, and the reaction forces of the mooring lines and fenders. If the floating body is assumed to be a rigid body, then its motions are comprised of the six Pitching components shown in Fig. T-8.3.1, namely surging, swaying, heaving, pitching, rolling and yawing. Out of Swaying these, the modes that represent motions within the horizontal plane, namely surging, swaying and yawing, may show long-period oscillations with the period of a Fig. T- 8.3.1 Components of Vessel’s Motion few minutes or more. Such long-period oscillations have a large influence on the occupancy area of a vessel at a mooring buoy and the design of the mooring system. One may thus give separate consideration to the long-period oscillations, taking only the wave-drift force and the long-period oscillation components of the winds and waves as the external forces when doing analysis. If the floating body is very long, elastic deformation may accompany the motions of floaty body and this should be investigated as necessary. [Technical Notes] (1) Methods of Solving the Equations of Motions (a) Steady state solution method for nonlinear equations of motion The equations of motions for a floating body are nonlinear, meaning that it is not easy to obtain solutions. Nevertheless, if it is assumed that the motion amplitudes are small and the equations of motion are linearized by using linear approximations for the nonlinear terms, the solutions can be obtained relatively easily. For example, for a three-dimensional floating body, one ends up with a system of six simultaneous linear equations involving the amplitudes and phases of the six modes of motions. Note that if the floating body is assumed to be a rigid body and its motions are linear, then the motions are proportional to the external forces. In particular, if there are no currents or wind, then the motions are proportional to the wave height. (b) Numerical simulation of nonlinear motions The wind drag force and the current drag force are in general nonlinear, and moreover the restraining forces of mooring equipment are also often nonlinear. In this case, an effective solution method is to use a numerical simulation whereby the equations of motion are progressively solved for a series of time steps. Such numerical simulation is commonplace nowadays. First, the time series data (which will be used as the external forces) are obtained for the wave-exciting force and the flow velocity due to the waves from the input of incident wave spectrum, as well as the fluctuating wind speed from the wind spectrum. The external forces obtained from these time series data are then put into the equations of motions for the floating body, and the time series data for the motions of the floating body and the mooring force are calculated. Numerical simulations are used for analyzing the motions of all kinds of floating bodies. For example, Ueda and Shiraishi 3) have carried out numerical simulations on the motions of a moored vessel, and Suzuki and Moroishi 4) have analyzed the swinging motion of a vessel moored at a buoy. Note that the following is usually assumed as preconditions in a numerical simulation: ① the fluid is an ideal fluid; ② the amplitudes of motions of the floating body are small; ③ the incident waves are linear and their superposition is allowed. If these assumptions cannot be held, it is necessary to carry out hydraulic model experiments. (2) Hydraulic Model Experiments Hydraulic model experiments provide a powerful technique for determining the motions of a floating body and the mooring force. Up to the present time, hydraulic model experiments have been carried out for all kinds of floating body. For examples, see references 5) and 6). (3) Law of Similarity for Mooring Systems The characteristics of the motions of a floating body vary greatly with the mooring method. When carrying out hydraulic model experiments on a floating body, it is thus particularly important to give appropriate consideration to the laws of similarity for the displacement and reaction force characteristics of the mooring equipment. For example, with a mooring rope, if the material used in the hydraulic model experiments is kept the same as that used in the field and the size is simply scaled down while maintaining the same shape, then the law of similarity will not hold; rather it is necessary to scale down the elastic modulus of the material used in the models relative to that used in the prototype. In practice, however, it will probably be unable to find such a material, in which case various other contrivances must be used.

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[References] 1) Tomotsuka TAKAYAMA, Tetsuya HIRAISHI, Masami FURUKAWA, Kunihisa SAO, Shin-ichiro TACHINO: “Feild observation of motions of a SALM buoy and tensions of mooring hawsers”, Tech. Note of PHRI, No. 542, 1985, 38 p. (in Japanese). 2) Yoshiyuki ITO, Shigeru CHIBA: “An approximate theory of floating breakwaters”, Rept of PHRI, Vol. 11, No. 2, 1972, pp. 15-28 (in Japanese). 3) Shigeru UEDA, Satoru SHIRAISHI: “Method and its evaluation for computation of moored ship’s motions”, Rept of PHRI, Vol. 22, No. 4, 1983, pp. 181-218 (in Japanese). 4) Yasumasa SUZUKI, Kazuyuki MOROISHI: “On the motions of ships moored to single-point mooring systems”, Rept of PHRI, Vol. 21, No. 2, 1982, pp. 107-150 (in Japanese). 5) Yasumasa SUZUKI: “Study on the design of single point buoy mooring”, Tech. Note of PHRI, No. 829, 1996, 48 p. (in Japanese). 6) Sigeru UEDA: “Analytical method of ship motions moored to quay walls and the aplications”, Tech. Note of PHRI, No. 504, 1984, 372 p. (in Japanese).

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Chapter 9 Estuarine Hydraulics 9.1 General (Notification Article 8) In planning and designing port and harbor facilities in an estuary where a river flows into the sea, estuarine hydraulic phenomena such as the flow situation at the times of high water and low water in the river, the bedload, tidal changes and density currents, and the coexistence of waves and river flow shall be estimated appropriately. [Commentary] In addition to the effect of outflow of fresh water during floods and droughts and the sediment transport from rivers, estuaries are also affected by the tide level changes, waves, tidal currents, longshore currents, and littoral drift. As a result, several hydraulic phenomena occur such as the periodic changes in water level and current speed, the formation of density currents, and the settling and deposition of sediment. These phenomena have a large influence on the flow regime in the estuary and the transport of sediment and others. It is thus necessary to give consideration to both conditions of rivers and the sea when handling estuarine hydraulics. [Technical Notes] (1) Tides in River The surface water level in a river channel can be calculated using either equation (9.1.1) or equation (9.1.2). (a) When tides are negligible (see Fig. T- 9.1.1) 2 2 aQ æ 1 1 ö Q æ 1 1 ö -------------Dh = h 1 h 2 = z 2 z 1 Dx -------- -------- - ç -------+ -------2 ÷ 2gs çè A 2 A 2 ÷ø 2 è K2 K2 ø 2 1 1

(9.1.1)

(b) When tides are considered

64748

2

2

1 ¶Q 2QB ¶H Q B æ ¶Hö Q ¶B ¶H Q Q ------ ------- ----------2- ------- ---------i + ------- --------3- ( H z ) ------ + ------- + ----------=0 3 è 2 ø ¶t ¶x ¶x gA ¶t ¶x (9.1.2) gA gA gA K ¶A ¶Q ------ + ------- = 0 ¶t ¶x where Dh： difference in water depth between two cross sections (m) h1： water depth at cross section 1 (m) h2： water depth at cross section 2 (m) z1： height of river bed above an arbitrary datum level at cross section 1 (m) z2： height of river bed above the arbitrary datum level at cross section 2 (m) z： height of river bed above the arbitrary datum level (m) a： velocity coefficient a ≒ 1.0 Q： flow rate (m3/s) A： cross-sectional area (m2) K： flow carrying capacity of cross section (m3/s), K 2 = A 2 R 1 / 3 / n 2 R： hydraulic radius (m) n： Manning’s roughness coefficient (s/ m1/3) Dx： distance between two cross sections (m) t： time (s) B： river width (m) H： water level from an arbitrary datum level (m), H = h + z Cross Section 2 Cross Section 1 i： channel bottom slope g： gravitational acceleration (g = 9.81 Fig. T- 9.1.1 Diagram Showing Water Level Curves m/s2) Equation (9.1.1) is a modified form of the basic equation for non-uniform flow in a channel of arbitrary cross section. Consequently, it cannot be applied to an estuary where there are strong tidal effects and a reverse, upstream flow occurs during a flood tide. However, it can be applied to an estuary where the tidal range is small (less than 20 cm) and the tidal compartmut is not long (say up to about 3 to 4 km upstream). Even so, it should only be used for the order estimate of hydraulic quantities during planning, because the calculation is only an approximation while ignoring tides. -148-


Equations (9.1.2) represents the equations of motions and continuity having been modified from the basic equations for unsteady flow in a river, where the flow rate and water level are the variables. In order to estimate the surface water level and flow rate due to the tidal action and propagation of tsunami into an estuary, simultaneous solutions can be obtained by equations (9.1.2) with appropriate boundary conditions. However, for a channel with a variable cross section, it is not so easy to solve equations (9.1.2) numerically. (2) Waves Entering an Estuary Upon entering a river mouth, waves are deformed by the currents. In addition to refraction due to the water depth, refraction due to the difference in the directions between waves and currents causes the attenuation of wave height. When the direction of waves is exactly opposite to that of river flow, however, wave height may increase through energy exchange through the river flow’s stopping action or radiation stress. When waves with an increased height run up the river channel, the wave height gradually decreases due to the effects of internal and external frictions, and turbulence of currents. These opposing effects are related to the properties of river flow and waves, and the mechanism of wave height change is very complex.

2

C 2 / C 1 = [1 - ( u /C1) sin a 1 ]

1

H2 / H 1 =

L1

C1

Wave direction cre

Zone 1: Still water

st

u=0

α1

α2

C2 L2

t res

L 2 / L 1 = [1 - ( u /C1) sin a 1 ]

6447448

sina 2 = sin a 1 /[1 - ( u /C1) sin a 1 ]2

Wa ve

c ve Wa

(a) Deformation of waves by currents (deepwater waves) As shown in Fig. T- 9.1.2, when waves propagates at an angle a across the straight boundary of discontinuity between the zone I where the water is still and the zone II where the water is flowing with a uniform velocity, refraction occurs at the boundary, changing the wave celerity and wavelength. If waves can be regarded as deepwater waves (i.e., the water is sufficiently deep relative to the wavelength in both the zones I and II), the wave celerity equation C = g L 2 leads to equations (9.1.3) and (9.1.4).

Zone 2: Flowing water

u

Uniform flow

Fig. T- 9.1.2 Refraction of Waves Due to River Flow

(9.1.3)

sin 2a 2 / sin 2a 1

(9.1.4)

The deformation of deepwater waves propagating on exactly opposite currents is given by equation (9.1.5).

L 2 / L 1 = ( 1 + m )2/4 H 2 / H 1 = 1/ 1 + 4u / C 1 m =

1 + 4u / C 1

644474448

C 2 / C 1 = ( 1 + m ) /2

(9.1.5)

where a： angle between the boundary line and the wave crest (º) u： uniform flow velocity in zone II (m/s) (positive when the flow is following the direction of propagation of the waves, negative when it is against) L： wavelength (m) C： wave celerity (m/s) H： wave height (m) Note that the subscript 1 denotes zone I (still water), while the subscript 2 denotes zone II (flowing water). Equations (9.1.3) was proposed by Johnson 1), while equation (9.1.4) was presented by Longuet-Higgins and Stewart 2). Equations (9.1.5) is a relationship that was obtained by Yu 3). According to equation (9.1.5), the wave height should increase in the exactly opposite current, and waves breaking theoretically occur when u = -C1/4. However, according to Yu’s experiments, wave partially breaks around u = -C1/7, and the wave height decreases. Incidentally, it should be noted that equation (9.1.5) cannot be applied to waves after breaking. (b) Deformation of waves by currents (finite water depth) Near a river mouth, where the water depth is relatively shallow compared with the wavelength of the incoming waves, the deformation of waves depends on the properties of both waves and river flow, along with the nonlinear interaction between them. It is thus not easy to estimate the wave height. Arthur 4) has carried out calculations whereby he specified the sea bottom bathymetry and the flow velocity distribution. He assumed the linear long waves, where the phase velocity of waves relative to the river flow is given as g h and it is not affected by the river flow. However, the velocity of waves is generally affected by currents and is different from the case of no river flow.

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For wave deformation near a river mouth, Iwagaki et al. have proposed a method for calculating the wave refraction in a current field on uneaven bottom. However, since the phase velocity and the group velocity of waves relative to the currents cannot be given in advance, any quantitative discussion was not made. Sakai et al. 5) have proposed a numerical calculation method for obtaining the directional spectrum of irregular waves near a coast where the water depth changes and currents are present. They show several cases of calculations. For example, there is a tendency that the change in the principal wave directions are affected mainly by the water depth and that wave components with frequency higher than the peak freakency are affected by currents. However, there is much room for further study on the way in which the wave breaking conditions are given. With regard to the nature of waves immediately after they have entered an estuary and come up against the river flow, Hamada has determined the changes that steady shallow water waves undergo while running up an estuary for both the cases that the vertical velocity distributions are uniform and parabolic. According to Hamada’s calculations, when h = 15 cm, T = 1.2 s and u = 20 cm/s, for both the distributions the wave height increases by about 5% in comparison with the case of no river flow. However, the rates of the wave height increase and the wavelength decrease with the river flow were slightly larger for the parabolic velocity distribution than for the uniform distribution. (3) Siltation and Channel Maintenance (a) Siltation When constructing a harbor, it is often necessary to carry out dredging, i.e., to excavate the sea bottom to deepen it in order to create navigation channels, mooring basins, and small craft basins. Even with an existing harbor, if the coming vessels are going to increase in size, it is necessary to carry out dredging in order to increase the water depth of the navigation channels and mooring basins. The sediment on the sea bottom is usually subject to external forces such as currents and waves. This means that even after construction of a harbor has been completed, it is necessary to continue dredging in order to maintain the functionality of the harbor because siltation occurs. The sediment in the estuarine part of a bay is often composed of fine particles such as clay and silt (hereafter referred to as mud). The phenomenon whereby such fine sediment is picked up, transported, and accumulates at the sea bottom is referred to as “siltation”. In Japan, Kumamoto Port and Miike Port in the Ariake Sea, and several harbors in the Suo Nada Sea area are faced with the problem of siltation. There are also many harbors in Europe, Southeast Asia, China, and South America that have similar problems. The siltation phenomenon can be divided into three stages: picking-up and transport of bottom mud by currents; mutual interference between waves and the bottom mud layer; and settling, accumulation and consolidation. In estuaries, both waves and currents exist simultaneously, and flocculation is promoted in the zone where saltwater and fresh water mix. The major difference between siltation and littoral drift (mainly sand) comes from different grain sizes. The mud that leads to siltation has a tendency to flocculate due to the mixing of the river water and the seawater in an estuary. Flocculation causes large changes in the settling characteristics of fine sediment. Fine particles of mud that have settled onto the sea bottom experience a process of dewatering and form the bottom sediment, and then its strength gradually increases over a long period of time through consolidation. Consequently, the resistence characteristic of mud against erosion by external forces such as waves and currents vary, strictly speaking, depending on the characteristic features of the mud (duration over time after settling (the level of consolidation), the texture, water content, organic matter content, etc.). This is the major difference between siltation and littoral drift, whereby the sand is generally treated as individual grain. Near to the sea bottom, the density of the mud generally varies with depth. In harbors that suffer from heavy siltation, much efforts are being made for the measurement, maintenance, and control of water depth, including that of navigation channels. For places where siltation is particularly pronounced, the water depth of navigation channels is quite changeable, and so it is necessary to monitor the bottom level constantly. If the required water depth is not sufficient for the safe navigation of vessels, bottom sediment need to be removed immediately. In most of the harbors in the world that suffer from heavy siltation, the sea bottom in approach channels is covered with fluid mud with a density of 1.05~1.3 g/cm3. In such a case, it is important to define the water depth which ensure safe navigation, because this definition directly affects the timing and quantity of dredging. Bathymetry measurement using a sounding lead or echo-sounders has been carried out for a long time for the purpose of managing and maintaining navigation channels. In the echo-sounding, a fluid mud layer can be detected by using different frequencies. The two frequencies commonly used are say 210 kHz (sound waves of this frequency are reflected from the surface of the fluid mud) and 33 kHz (sound waves of this frequency passes through the fluid mud but is reflected from sand or higher-density mud). In some of the large European ports such as Rotterdam, Zeebrugge, Dunkirk, Bordeaux and Nantes, as well as in estuaries in Brazil, Venezuela and Indonesia, it is said that the difference between the surface detected by 210 kHz sound waves and that detected by 33 kHz sound waves can be as much as a few meters 6). Note, however, that it is not really sufficient to fix the water depth for which navigation is possible simply by using such equipment. In Europe, where many navigation channels have the problem of heavy siltation (in particular in the Rotterdam Europort area), the safe nautical depth is specified as being the depth at which the density of bottom material is no more than 1.2 g/cm3 7). -150-


In addition to this density requirement, the following two criteria must be satisfied. ① Even if the draft of a vessel approaches the nautical depth, there should be no damage to the hull. ② The viscous drag induced by the underside of a moving vessel (i.e., the rheology characteristic) and the

internal waves generated at the mud/water boundary do not cause any change in the water depth. The criteria mentioned above can be considered that the water depth has been defined from a physical standpoint. Although it is ultimately desirable to carry out direct evaluation by means of a viscosity meter, it can be considered that at the present technical stage, the water depth has been specfied using the density value that is most reliable in terms of measurement technology. New measuring equipment for the sediment density in navigation channels using the g rays has been developed in Europe 6). In Japan, Ishizuka and Nemoto 8) have developed a density measuring device that uses the g rays. (b) Formation of fluid mud The fluid mud layer is often found in estuaries or on the continental shelf close to the coast. The layer contains a very high concentration of mud in fluid condition and easy to move. The mud concentration in the fluid mud layer is of the order of 10,000~300,000 mg/l 9). In fact, Krone1 9) defines a fluid mud layer to have a mud concentration of at least 10,000 mg/l. Kirby and Parker1 10) have obtained the vertical distribution of the density within a fluid mud layer using a density measuring device that makes use of the scattering properties of the g rays. By comparing with the results of echo sounding measurements, they have concluded that the density of fluid mud lies in the range 1.05~1.3 g/cm3. (c) Effect of submerged dykes In Kumamoto Port, which is currently being constructed on very gentle mud flats of the Ariake Sea, it is expected that navigation channels and mooring basins will be subject to siltation. Large-scale field observations are thus being carried out to investigate siltation process in this area, and proposals of countermeasure are being investigated. Before construction was commenced, i.e., when there was nothing at all in this sea area, three test trenches for siltation experiments were constructed. One (Trench No. 1) was located where the water depth is 4m, and two (Trenches No. 2 and No. 3; separated from each other by 100m) were located where the water depth is 2m. The three trenches were all of the same size, but Trench No. 3 was made different from the other two in that it was surrounded by 1 m-high submerged dykes. The locations of the trenches are shown in Fig. T- 9.1.3.

Fig. T- 9.1.3 Locations of Trenches Used in Siltation Experiments

Figure T- 9.1.4 shows the time series of the amount of siltation in each trench, as measured at the center of each of the three trenches. During two large storms in 1987, Trenches No.1 and No.2 silted up rapidly, with over 60 cm of siltation occurring in just one day. However, in Trench No. 3, which was surrounded by submerged dykes, hardly any siltation occurred at all; there was also no trace of any significant mud accumulation along the outsides of the submerged dykes.

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-100

0

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Area 6

92.9

93.3

93.10 94.2

Fig. T- 9.1.5 Time History of Siltation in Various Areas of Kumamoto Port

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3/1 cm 150

Trench No.1 (DL-4m) Trench No.2 (DL-2m) Trench No.3 (DL-2m) 100

Fig. T- 9.1.4 Time History of Siltation in the Central Parts of Three Test Trenches

N

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0

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0

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,

,

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Units : cm


Figure T- 9.1.5 shows the change over time in siltation for a number of areas of Kumamoto Port after it was partially opened; some of the areas are in a navigation channel, while others are in a mooring basin. Submerged dykes of 1 m to 1.5 m in height were installed on the both sides of the navigation channel. For the part of navigation channel (Areas 6 to 10), hardly any silt accumulation was observed from October 1993 onwards, thus showing the effect of the submerged dykes in effectively preventing siltation. For the mooring basin part (Areas 1’~5), on the other hand, there was progressive sedimentation. It is thought that this was caused by fluid mud flowing in from the lower section of the pier. The places from where the fluid mud flowed in are currently being plugged up, and continuous observation is being carried out. (d) Numerical simulation In order to forecast siltation, it is necessary to predict external forces such as currents and waves, and develop the computational models that incorporate the above-mentioned flocculation, settling, and erosion mechanisms, with the fluid mud model that describes the formation of fluid mud and its flow mechanism. Tsuruya 12) has applied a multi-layered level siltation forecasting model to Kumamoto Port, and then Tsuruya et al. have extended it into a new model that includes fluid mud layer mechanism. They have shown that this model accurately represents the siltation of Kumamoto Port during a typhoon. [References] 1) J. W. Johnson: “The refraction of surface waves by current”, Trans. A. G. V., Vol. 28, No. 6, 1947. 2) Longuet-Higgins, M. S., and R. W. Stewart: “The change in amplitude of short gravity waves on steady non-uniform currents”, J. Fluid Mech., Vol. 10, 1952, pp. 529-549. 3) Yu, Yi-Yuan: “Breaking of waves by an opposing current”, Trans. A. G. U., Vol. 33, No. 1, 1950. 4) Arthur, R. S.: “Refraction of shallow water wave - The combined effect of current and underwater topography -”, Trans. A. G. U., Vol. 31, 1950, pp. 549-552. 5) Sakai, T., F. Hirosue, and Y. Inagaki: “Wave directional spectra change due to underwater topography and current”, Proc. Fifth International Offshore Mechanics and Arctic Engineerring (OMAE) Symposium, Vol. 1, 1986, pp. 59-65. 6) De Vlieger, H. and J. De Cloedt: “Navitracker: a giant step forward in tactics and economics of maintenance dredging”, Terra et Aqua 35, December 1987, pp. 2-18. 7) PIANC: “Navigation in muddy areas”, Excerpt from Bulletin No. 43(1982Å - 1983), Report of Working Group 3-a, 1983, PIANC, Brussels. 8) Koji ISHIZUKA, Takashi NEMOTO: “Development of mud layer density meter using radio isotope”, Rept of PHRI, Vol. 30, No. 4, 1991, pp. 85-109 (in Japanese). 9) Van Leussen, W. and J. Dronkers: “Physical process in estuaries: An introduction”, in J. Dronkers and W. van Leussen (ed), “Physical Processes in Estuaries”, Springer-Verlag, 1988, 560 p. 10) Krone, R. B.: “Flume studies of the transport of sediment in estuarial shoaling processes”, University of California, Hydraulic Engneering Lab. and Sanitary Research Lab., Barkeley, 1962, 110 p. 11) Kirby, R. and W. R. Parker: “Seabed density measurements related to echo sounder records”, The Dock and Harbour Authority, Vol. LIV, No. 641, 1974, pp. 423-424. 12) Tsuruya, H.: “Cohesive sediment transport model and its application to approach channels and anchorages in estuarine ports”, Seminar Text on Maritime Infrastructure Development, Langkawi, Malaysia, November 1995, pp. 39-53.

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Chapter 10 Littoral Drift 10.1 General (Notification Article 9) When port and harbor facilities are to be affected by the littoral drift phenomenon, characteristic values of littoral drift shall be established appropriately for sediment grain size, threshold depth of sediment movement, longshore sediment transport rate, and predominant direction of longshore transport. [Commentary] (1) Littoral drift refers to either the phenomenon that the sediment composing a sea coast or lakeshore is moved by the action of some force such as waves or currents, or material that is moved by the above process. (2) Although the movement of sand by winds and the sand that is thus moved are referred to as the wind-blown sand, in the broad definition the littoral drift is also considered to include wind-blown sand at beaches. (3) Sediment that forms a beach is supplied from nearby rivers, coastal cliffs, and the adjacent coastline. The sediment is exposed to the action of waves or currents during the supply process or after it has accreted on the beach. This is why the sediment shows characteristics that reflect the characteristics of external forces such as waves and currents. This is referred to as the sediment sorting action by external forces. (4) As a natural beach is repeatedly subjected to process of erosion when storm waves attack and that of accretion during periods when waves are moderate, it achieves a relatively balanced topography over a long period. This balance may be lost by a reduction in the supply of sand owing to river improvements, by changes in sand supply conditions following construction of coastal structures, and by changes in external forces such as waves and currents. Then beach deformation will occur as the beach moves toward new equilibrium conditions. When building structures such as breakwaters, groins, detached breakwaters, and training jetties, careful attention should be paid to the changes that the construction works will bring about in the balance of the beach. Topographical changes that might be induced by a construction project should be sufficiently investigated in advance. In addition, careful attention should be paid to the deformation conditions of the beach both during construction and following completion of any structure, and appropriate coastal protection countermeasures should be taken any time when there are concerns about the possibility for disaster triggered by coastal erosion. (5) When waves approach a coast from offshore, the movement of water particles near the sea floor does not have the force to move the sediment in places where the water depth is sufficiently deep. At a certain water depth, however, the sediment begins to move. The water depth at this boundary where sediment begins to move is called the threshold depth of sediment movement. Sato 1) studied the movement of sediment by placing radioactive glass sand on the sea floor and investigating the distribution of their movement. From this study, he defined two conditions that are called the surface layer movement and the total movement, respectively. He applied the former term to the situation in which the sand in the surface layer on the sea floor is moved collectively in the direction of wave movement. The latter term he applied to the situation that the sand shows striking movement with a distinctly visible change in water depth (see [Technical Notes] (4)). (6) Longshore sediment transport rate refers to the rate of littoral drift in the direction parallel to the coast that is caused by waves obliquely incident to a coast. (7) Longshore sediment moves in either the right or left direction along a coast, corresponding to the direction of incoming waves. The direction with the larger volume of movement during a year is called the predominant direction. (8) The littoral drift in the direction parallel to the coastline (shoreline) is designated as the longshore sediment transport. It contributes to the advance and retreat of a shoreline. Because the process is normally nonreversible, it is often accompanied by beach erosion or accretion over a long period of time. [Technical Notes] (1) Coast Topography (a) Terminology for various sections of a beach profile Various sections of a sandy beach are defined with the terminology shown in Fig. T- 10.1.1. The “offshore” is the area on the ocean side where waves do not break normally, and in many cases the bottom slope is comparatively gentle. The “inshore” (nearshore) refers to the area between the offshore and the low tide shoreline, where waves break and longshore bars or steps are formed. The “foreshore” is the zone from the low tide shoreline to the location where waves will reach normally, and the “backshore” is the zone from the landward boundary of foreshore to the coastline, where waves will reach during stormy weather with the rise of water level.

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Shore Backshore Foreshore

Beach scarp

Inshore

Low tide shoreline

Bluffs

High tide shoreline

Coastline

Coast

Offshore

Mean high water Mean Low water

Fig. T- 10.1.1 Terminology of Beach Profile

(b) Step type beach (ordinary beach) and bar type beach (storm beach) When a model beach is constructed with natural sand in a wave channel and exposed to waves over a long period of time, the beach profile will approach an equilibrium condition corresponding to the waves acting upon it. This equilibrium condition of beach is broadly classified into two types as shown in Fig. T- 10.1.2 (a) and (b), called a step type beach and a bar type beach, respectively. Bar type beach

Step type beach

(Storm beach)

(Normal beach)

Forebeach

Forebeach Step

Trough

(a) Step type beach

bar

(b) Bar type beach

Fig. T- 10.1.2 Step Type Beach and Bar Type Beach

(2) Characteristics of Sediment The grain size characteristics of sediment are normally expressed with the indices given below. ① Median diameter (d50): diameter corresponding to a cumulative percentage p = 50% on the grain size

distribution curve

100

å dDp

② Average grain diameter (dm):

p=0

d m = ---------------------

(10.1.1)

100

å Dp

p=0

③ Sorting coefficient

(S0):

S 0 = d 75 ¤ d 25

④ Skewness parameter

(Sk):

S k = d 75 ´ d 25 ¤ ( d 50 )

where p： Dp： d25： d75：

(10.1.2) 2

(10.1.3)

cumulative percentage (%) increase in the cumulative percentage grain size corresponding to 25% of the cumulative percentage grain size corresponding to 75% of the cumulative percentage

(3) Form of Littoral Drift Movement Littoral drift is classified into two categories of bedload and suspended sediment, according to the modes of sediment movement. The bedload includes sheet flow. (a) Bedload: littoral drift that moves by tumbling, sliding or bouncing along the surface of the sea floor through the direct action of waves and currents. (b) Suspended sediment: littoral drift that is suspended in seawater by turbulence of breakers and others and transported by currents. -155-


(c) Sheet flow: littoral drift that moves as a layer of high density flow near the bed surface. Shallow water zones can be classified into three regions as shown in Fig. T- 10.1.3, depending upon the physical properties of waves that provide the external forces for the littoral drift phenomenon. The dominant mode of the littoral drift movement in each region is as follows. ① Offshore zone

In order for sand to be moved by the action of fluid motion (oscillatory movement), the current velocity of the fluid must exceed a certain value. This condition is generally called the threshold of movement. For littoral drift the threshold of movement is defined with the water depth (threshold depth of sediment movement). When the water depth is shallower than the threshold depth of sediment movement, regular, small undulating topographic contours that are called the sand ripples will form on the sea floor surface. When sand ripples form, vortices are generated by the fluid motion in the vicinity of the sand ripples and movement of suspended sediment trapped in the vortices occur. As the water depth becomes shallower, sand ripples are extinguished, and a sheet flow condition occur in which sediment moves in stratified layers extending several layers below the sea bed surface. ② Surf zone

Inside the surf zone, high-density suspension of sediment is formed by the severe agitation and action of large-scale vortices that are generated by breakers. The volume of sand that moves near the bed surface in a bedload state also increases. For convenience the sand movement inside the surf zone is divided into a component that moves parallel to the shoreline (coastline) (called the longshore sediment transport) and a component that is perpendicular to the shoreline (called the cross-shore sediment transport). While the time frame for beach deformation caused by longshore sediment transport is long, the time frame for cross-shore sediment transport is relatively short (from a few days to about one week), like that for periods of passing storms. ③ Swash zone

Bed surface current speed amplitude

Wave height H

The sand movement in a swash zone differs for the times of wave runup and downflow. During the time of wave runup sand is put in suspension by the agitation at the front of a wave and transported by running-up water, whereas during the downflow sand is carried in a bedload mode.

Threshold depth of sediment movement

Suspension above sand ripples

Bedload movement

Offshore zone

Sheet flow

Suspension from breakers

Breaker zone

Movement in the wave runup zone Agitation from breakers Suspension above sand ripples Undertow Nearshore currents

Swash zone

Fig. T- 10.1.3 Changes in Sediment Movement Modes in Cross-Shore Direction

(4) Physical Meaning of and Estimation Formulas for the Threshold Depth of Sediment Movement To determine the extension of breakwater (water depth at the head) and the required threshold depth of sediment movement when seeking the offshore boundary of beach deformation, conducted a number of field surveys -156-


using radioactive glass sand as a tracer. Based upon their observed results they defined the littoral drift movement conditions as follows. (a) Surface layer movement: As shown in Fig. T- 10.1.4 (a), the elongation of the isocount lines that show the distribution of radioactive glass sand after waves acted upon it on the sea floor demonstrates that all of the sand has moved in the direction of the waves. But the location of the highest count remained at the input point of glass sand, indicating no movement. This corresponds to a situation in which the surface layer sand is moved collectively by traction, parallel to the wave direction. (b) Total movement: As shown in Fig. T- 10.1.4 (b), this refers to a situation in which both the isocount lines and the portion of the highest count move in the wave direction. This corresponds to a situation of distinct sand movement with the result of apparent change in water depth. The threshold depth of total sediment movement is often calculated when the threshold depth of sediment movement is investigated for engineering purposes.

Waves

Waves Input point

Input point

Highest count

Highest count Isocount line

Isocount line

(a) Surface layer movement

(b) Total movement

Fig. T- 10.1.4 Spread of Radioactive Glass Sand in Surface Layer Movement and Total Movement

Based upon their field data, 2) proposed two equations for estimating the threshold depth of surface layer sediment movement and that of total sediment movement. (c) Threshold depth of surface layer sediment movement 2ph i H 0 H0 d 1¤3 sinh ----------- × ----------- = 1.35 æ -----ö è ø L H L L0 0

(10.1.4)

(d) Threshold depth of total sediment movement H0 -----L0 where L0： H0： L： H： d： hi：

2ph i H 0 d 1¤3 = 2.40 æ -----ö sinh ----------- × -----è L 0ø L H

(10.1.5)

deepwater wavelength (m) equivalent deepwater wave height (m) wavelength at water depth hi (m) wave height at water depth hi (m) sediment grain size (average grain size or median diameter) (m) threshold depth of sediment movement (m)

Repeated calculations are required to estimate the threshold water depths using equations (10.1.4) and (10.1.5). Calculation diagrams like those in Fig. T- 10.1.5 (a) and (b) have been prepared so that the depths can be easily estimated. By specifing d/L0 and H0/L0, it is possible to determine hi/L0. (5) Longshore Sediment Transport The predominant direction of longshore sediment transport is determined using the following information: (a) (b) (c) (d)

Topography of the natural coast and that around coastal structures (see Fig. T- 10.1.6) Alongshore distribution of the sediment characteristics (median diameter, mineral composition, etc.) Direction of movement of fluorescent sand tracers Direction of incident wave energy flux

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Fig. T- 10.1.5 (a) Calculation Diagram for Threshold Depth of Surface Layer Sediment Movement

Fig. T- 10.1.5 (b) Calculation Diagram for Threshold Depth of Total Sediment Movement

Predominant direction

Predominant direction Jetties Predominant direction

River current Predominant direction

Harbor interior Eroding coast protected Littoral drift accretion by a coastal revetment Beach is wide Beach is narrow Predominant direction Detached breakwaters Beach is narrow

Predominant direction Sand spit

Beach is wide

Fig. T- 10.1.6 Typical Coastal Topography Showing the Predominant Direction of Littoral Drift

To estimate the longshore sediment transport rate, the following various data must be prepared and sufficiently investigated: (a) Continuous observation data of the change in sediment volume in the area around a coastal structure (b) Data on the alongshore component of wave energy flex -158-


(c) (d) (e) (f)

Data concerning the littoral drift rate at the surrounding coast Data on past dredging quantities Continuous observation data on deposition volume at the location of experimental dredging Data on the volume of movement of fluorescent sand tracers placed within the surf zone

Various equations can be used to estimate an approximate value of longshore sediment transport rate 3), 4). Longshore sediment estimation equations are normally given in the expression shown in equation (10.1.6), with the coefficient a for various equations being as given in Table T- 10.1.1. 2

Ex = where Qx ： Ex： Kr： nA： w 0： r 0： g： HA ： LA： T： a b：

2 SK r

æ n A w 0 H A L Aö -÷ sin a b × cos a b ç ---------------------------8T è ø

64748

Q x = aE x

(10.1.6)

longshore sediment transport rate (m3/s) alongshore component of wave energy flux (kN･m/m/s) refraction coefficient between the wave observation point and the breaking point ratio of group velocity to wave celerity at the wave observation point = r0g seawater density (t/m3) (1030 kg/m3) gravitational acceleration (m/s2) (9.81 m/s2) wave height at the wave observation point (m) wavelength at the wave observation point (m) wave period (s) angle of wave incidence at the breaking point (º) Table T- 10.1.1 Coefficient a for Longshore Sediment Transport Rate Equation Savage 3)

Sato and Tanaka 1)

U.S. Army Corps of Engineers 4)

0.022

0.03

0.04

(6) Littoral Drift Phenomenon in the Surf Zone Inside the surf zone, large quantities of sand are put in motion by the increase of the water particle velocity near the bottom, by turbulence caused by wave breaking, and by the existence of nearshore currents. The sediment movement when suspended sediment is predominant can be examined by dividing the movement into two types. (a) Sediment suspension process caused by systematic vortices formed by wave breaking. (b) Settling process during which sediment is buffeted by random external forces following breakup of organized vortices into small eddies.

Density

Current velocity in the direction of the shore

Figure T- 10.1.7 gives the results of continuous observations of suspended sediment concentration and horizontal current speed that were carried out by Katoh et al. 5) inside a breaker zone in the field. The white arrows in the figure point out the waves that broke on the offshore side of the observation point and the black arrows point out the waves that passed the observation point and broke on the coast side. It is clear that the suspended sediment concentration increased rapidly when waves broke on the offshore side. This result indicates that sediment suspension is related to the organized vortices (particularly obliquely descending vortices) that occur after waves break.

Time

Fig. T- 10.1.7 Example of Field Observation of Suspended Sediment Density 10)

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(7) Topographical (Shoreline) Deformation in the Swash Zone Horikawa et al. investigated the criteria for shoreline advance and retreat occurring as a result of sand movement in the swash zone based on laboratory experiments, and proposed equation (10.1.7) which is also applicable for the field condition. H0 0.27 æ d ö 0.67 ----(10.1.7) ------ = C s ( tan b ) è L 0ø L0 where H 0： L0： tanb： d： Cs：

deepwater wave height (m) deepwater wavelength (m) average bottom slope from the shoreline to a water depth of 20 m sediment grain size (m) coefficient

Based on equation (10.1.7), a shoreline will retreat when Cs ≧ 18 (see Fig. T- 10.1.8).

Retreat Index based upon experimental results or

Retreat Advance

Advance

Fig. T- 10.1.8 Advance and Retreat of Shorelines in Field

Distance in offshore

HoRF 3/12/86 9/11/86 D.L.+1.4m Offshore direction

Calculated value Measured values

Shore direction

Month seaward Fig. T- 10.1.9 Comparison of Calculation and Actual Measurements of Shoreline Location

Katoh et al. revised equation (10.1.7) using deepwater wave energy flux and presented a model to calculate the change of daily shoreline locations. Figure T- 10.1.9 is a comparison of the calculated and actually measured results of shoreline location. (8) Relationship between Foreshore Topographical Changes and Groundwater Level The topographical changes that accompany the change in the foreshore tide level can be explained as follows by using Fig. T- 10.1.10 6). When the tide level changes, the beach groundwater level also changes as a response. But because of the delay in response time, the groundwater level differs from the foreshore tide level during the time of flood tides and that of ebb tides. During the flood tide the groundwater level is low, and it is easy for the seawater running up on the beach to permeate underground. Thus the sediment carried by the seawater when it runs up on the beach will accrete there. On the other hand, during the time of ebb tide the groundwater level is high and it is difficult for seawater to run up on the beach and to permeate underground. At certain conditions, the groundwater may flow out of the beach surface during the ebb tide. As shown in Fig. T- 10.1.10 (b), the -160-


result is that the sediment that accreted during the time of flood tide will be eroded, and return to its original location. When waves run up to a high level on a beach during storm periods, a high groundwater level condition continues throughout the stormy weather period because the run-up seawater permeates into the beach, and the condition becomes as shown in Fig. T- 10.1.10 (b). Occurrence of rapid foreshore erosion during such the condition has been confirmed by the field data. Some shore protection methods make use of this relationship between the foreshore groundwater level and sand movement; i.e., lowering the groundwater level by forced means or gravity and thus halting erosion. In the gravity methods, a highly water-permeable layer is installed in the foreshore sand to cause the groundwater flow down offshore and to lower the groundwater level. With this method it is possible to preserve beach conditions close to those of a natural beach because no structures are visible above the beach floor.

Accretion

Sand movement towards the coast

Permeation Groundwater surface

Erosion

(a) Flood Tide

Groundwater surface Sand movement towards the offshore Water outflow Accretion (b) Ebb Tide Fig. T- 10.1.10 Relationship between Foreshore Topographical Changes and Groundwater Level

10.2 Scouring around Structures Scouring shall be taken into consideration when there are concerns that scouring around structures such as breakwaters, groins, and training jetties may affect the safety and integrity of structures. [Commentary] Wave characteristics that act on natural beaches can be considered as nearly fixed over a long period of time. Topographies that form in response to these characteristics are nearly stable as well. Scouring will occur when structures are erected and the equilibrium between external forces and topography will be disturbed locally or over a broad area. Because the mechanism and amount of scouring will also change when the conditions of wave action on a beach vary with construction of structures, these must be considered carefully when choosing countermeasures against scouring. [Technical Notes]

K

(1) Beach Scouring in Front of Coastal Revetment It is well known that beach scouring in front of coastal revetment has a close relationship with wave reflection. For example, Fig. T- 10.2.1 has been proposed for determining scouring or accretion by means of the reflection coefficient K and the parameter (H0 / L0)(l /d50) sina, which is defined with the wave steepness H0 / L0, median sediment diameter d50, slope gradient of coastal revetment a (for a vertical breakwater, a = 90º), and the distance l from the wave runup point on an equilibrium profile to the location of the coastal revetment. The diagram indicates that all other conditions being equal, it is advantageous to make the front surface of revetment inclined against beach scouring in front of revetment.

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Scouring Accretion

H0 L0 d50

sin α

Fig. T- 10.2.1 Threshold Conditions between Scouring and Accretion in Front of Coastal Revetment


(2) Local Scouring around Breakwaters (a) Scouring in the surf zone ① Local scouring at the breakwater head

West breakwater Inner breakwater East breakwater North breakwater Inner breakwater South breakwater

Legend

West breakwater Port of Kanazawa Inner breakwater New south breakwater Port ofAkita ofAkita Port Breakwater Port of Mikuni Sankoku

Scouring depth

Scouring depth

Water depth

Kashima Niigata New Port Port of Kagoshima

Figure T- 10.2.2 shows the local scouring conditions around a breakwater head, as analyzed by Tanaka. The maximum scouring depth is found to be nearly equal to the maximum significant wave height (H1/3)max during the period up to 15 days prior to the time of scouring measurements. In addition, Fig. T10.2.3 shows the relationship between the water depth around a breakwater head and the scouring depth. The scouring depth becomes largest when the water depth at breakwater head is about 3 m to 5 m (breaker zone).

Relationship to (H1/3)max during the prior15 days

Fig. T- 10.2.2 Relationship between Scouring Depth at Breakwater Head and Maximum Significant Wave Height during the Prior 15 Days

Water depth around the breakwater head Fig. T- 10.2.3 Relationship between Scouring Depth around Breakwater Head and Water Depth

② Scouring along the face line of breakwaters

Scouring depth (m)

Figure T- 10.2.4 shows the relationship between the scouring depth along the face line of a breakwater and the water depth, which is based on the field data of a large harbor. The closed cirdes in the figure show the scouring depth in the section of the south breakwater which is oriented obliquely to the coastline. The alignment of south breakwater changed from perpendiaular to oblique to the coastline at the water depth of about 7 m. We can see that the scouring depth becomes largest at this inflection point and then gradually decreases as one moves offshore. The scouring depth along the sections of breakwaters perpendicular to the coastline are shown with open circles. It reaches a maximum value at the point of water depth about 2 m, and decreases as the water depth becomes shallower or deeper than that point. The location where the scouring depth is largest corresponds to the location where a longshore bar exist.

North breakwater and south breakwater vertical sections South breakwater slope section

Water depth (m)

Fig. T- 10.2.4 Relationship between Scouring Depth and Water Depth

(b) Scouring in standing wave regime Scouring depth in front of a vertical wall tends to decrease as the initial water depth in front of the wall increases and the wave condition is shifted into the standing wave regime. In case of composite type breakwaters, where the toe of rubble mound is somewhat away from the wave reflection surface of the upright section, scouring at the toe of rubble mound by standing waves sometimes become a problem. Irie et al. 7) carried out experiments concerning this type of scouring and highlighted the following issues: -162-


horizonal velocity of water particles at the bottom by incident waves Ub to the settling velocity of sedimentω. When Ub /ω > 10, sediment will move from the location of the node of standing waves to the location of the antinode, with scouring occurring at the node and accretion taking place at the antinode (L-type scouring). When Ub / ω < 10, the opposite phenomenon (N-type scouring) will occur (refer to Fig. T- 10.2.6). The L-type scouring refers to the phenomenon where accretion occurs at the antinode of standing waves and scouring occurs at the node, whereas the N-type scouring refers to the opposite phenomenon where scouring occurs at the antinode and accretion occurs at the node.

② There are many instances in the field where Ub /ω > 10,

Breakwater

① The basic parameter is Ub /ω, the ratio of the maximum

Node

Antinode N-type

Rubble mound

L-type

Fig. T- 10.2.6 Sketch of Scouring by Standing Waves 15)

and generally scouring at the node of standing waves is predominant. Normally, because a toe of rubble mound is located at the distance of about 1/4 wavelength or so from the upright section, scouring and subsidence of the rubble mound of breakwater will occur at its toe as the sediment there move toward the location of antinode at one half wavelength from the upright section.

10.3 Prediction of Beach Deformation All the related factors shall be thoroughly investigated when predicting beach deformation, with consideration given to factors such as the predicted results by an appropriate forecasting method and the data of past beach deformation at the site in question. [Commentary] Various methods exist as procedures for predicting beach deformation, including empirical prediction techniques, estimation based on hydraulic model experiments (especially with movable bed model experiments), and numerical simulations. Because beach deformation is strongly governed by the characteristics of the region in question, however, it is inappropriate to rely on any single method. Efforts should be made to predict beach deformation by combining two or more procedures and by investigating the local data and information as comprehensively as possible. [Technical Notes] (1) Empirical Prediction Techniques The empirical method is a procedure that, on the basis of collection and analysis of past examples of beach deformation, the layout and structural characteristics of structures to be built are compared with past examples of similar nature. Based upon the similarities, the potential for beach deformation to be caused by construction of structures is judged. Tanaka has carried out research on modeling of the complicated topographical changes that occur after construction of structures. He classified characteristics of typical topographical changes in numerous examples of beach deformation. As a result of this research, it is possible to understand the topographical changes in the vicinity of Japan’s ports and harbors in several representative patterns (Fig. T10.3.1). Exceptions to these patterns are relatively rare. By judging which pattern in Fig. T- 10.3.1 is applicable to the coast under investigation, a qualitative prediction of beach deformation becomes possible. (2) Hydraulic Model Experiment (particularly, Movable Bed Model Experiments) The capability of predicting beach deformation based on hydraulic model experiments, particularly movable bed model experiments, is limited because the problem of similarity remains unresolved. But the advantage of model experiments is such that specific topographical changes can be reproduced in a laboratory basin and the phenomenon to be forecasted can be understood visually. Because of the similarity problems being unresolved, experiments are carried out with partially distorting model scales, and by focusing attention on the reproducibility of the area of most concern, based on the comparison of several formulas on beach deformation simililarity, and a topographic model that is judged most reliable is introduced in a laboratory basin. Before predicting future beach deformation, it will be necessary to verify the model for reproducibility of the topographical changes that occurred in the past in the study area and to confirm the model’s kinematic similarity. The degree of kinematic similarity will be judged by the accuracy of reproduction. The reproductive accuracy of the experiment, therefore, cannot exceed the accuracy of the data collected on beach deformations in the past.

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Convex coast

Concave coast

Effect of coastline shape

Long jetty

Short jetty

Updrift accretion and downdrift erosion by obstructing littoral transport Effect of jetty length

a

Without lee breakwater

Formation of one-side salient

Change in salient by breakwater extension

Completion stage

Tombolo formation and associated shoreline changes Formation of a tombolo Inithial stage

Accretion in calm zone

Prevention of local erosion and accretion caused by Mach-stem waves Breakwaters extending obliquely from the shoreline

Changes in beach and water depth inside a harbor

Change in a shoreline accompanying change in the direction and shape of harbor entrance

With lee breakwater

Fig. T- 10.3.1 Classification of Patterns of Topographical Changes after Construction of Structures

Formation of a sand bar at the head of a main breakwater and harbor entrance

Accretion on both sides of training jetty of the center of a long concave coast

Accretion towards the training jetty of a river month at the edge of a beach

At the side surface of a straight section

At the front of an oblique section

At a narrow mouth

At a bend

Local scouring in the vicinity of a breakwater At the breakwater head

Accretion between a cape and a breakwater built on the downdrift side of the cape

Long, oblique breakwater



(3) Prediction Based on Numerical Simulations At the present time, numerical simulations are divided into two models: those that predict changes in the shoreline location, and those that predict three-dimensional changes in water depth; i.e., beach topographical changes. The former model is called the one-line theory, and its overview is given below. Beach sediment is transported by waves and currents both in the offshore and onshore directions and in the alongshore direction. Because littoral drift is caused mainly from the direct action of waves, littoral drift during storm periods will be predominantly towards offshore, and the coast will be eroded with a retrent of shoreline. When the sea becomes calm, however, the sediment will be carried towards the shore and the shoreline will advance. Along with these movements the beach profile will also change. This topographical change in the shoreline location and beach profiles caused by the onshore-offshore transport is normally a seasonal one, however. When looked at on the average profile over a long period of time, the changes caused by onshoreoffshore transport can mostly be ignored when compared with those caused by longshore transport. Thus, when focusing on beach erosion or accretion over a period of several years, one can assume there is no change on the shape of beach profile and that beach erosion and accretion will correspond to the retreat and advance of the shoreline. A prediction of changes in the shoreline location can then be based on the deposititon and removal of sediment volume primarily from longshore transport. Figure T- 10.3.2 sketches the calculation principles of a shoreline change prediction model. As shown in the figure, the coastline should be split along the longshore direction of the shoreline into sections having the width Dy, and the inflow and outflow of sediment volume between those widths are considered. That is, when the ¶Q inflow of sediment volume QDt and the outflow of sediment volume æ Q + -------Dyö Dt during time period Dt are è ¶y ø compared, accretion will occur if the former is larger and erosion will take place if the latter is larger. By introducing the assumption that the beach profile remains unchanged over time and any imbalance in the sediment inflow and outflow simply shifts the beach profile offshore or onshore, it is possible to express the advance and retreat of the shoreline as the result of the imbalance. When this is expressed in the continuity of sediment flux, the result is equation (10.3.1). ¶x s 1 ¶Q ------- + ------ × æ ------- qö = 0 ø ¶t D s è ¶y

Offsh

Offshore direction coordinate

ore d

irecti

on co

ordin

ate

Longshore coordinate

gsh

Lon

e

inat

ord

co ore

(10.3.1)

Model view

Plan view

Fig. T- 10.3.2 Relationship between Volume of Shoreline Change and Sand Movement

where xs ： t： y： Ds ： Q： q：

shoreline location (m) time (s) coordinate in longshore direction (m) width of the littoral drift movement zone (m) littoral sediment transport rate (m3/s) cross-shore inflow (q > 0) or outflow (q < 0) of the littoral sediment transport rate across the onshoreoffshore boundary per unit width in the longshore direction (m3/m/s)

To find the longshore sediment transport rate Q, use should be made of the littoral sediment transport rate equation related to the component of the incident wave energy flux at the breaker point in the longshore current direction. Using this equation, the longshore sediment transport rate can be determined using just the wave height and wave direction at the breaker point as the input. Normally when beach deformation is predicted, however, the beach will have some structures that produce an area sheltered from incoming waves. Because of this sheltered area, the wave height varies alongshore and longshore currents are induced. An equation based on Ozasa and Brampton 8) that incorporates this kind of effect is frequently used. 2

HB Cg B ¶H B 2K 2 Q = -------------------------- × æ K 1 sin 2 q B ---------- × cos q B × ----------ö 16s ( 1 l ) è ¶y ø tan b

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(10.3.2)


where HB ： CgB： q B： tanb： s： r s： r 0： l： K1, K2：

breaker height (m) group velocity at the breaker point (m/s) angle formed by the wave front when waves break and the shoreline (º) equilibrium beach slope s = ( rs r0 ) ¤ r0 density of sediment (g/cm3) density of seawater (g/cm3) void ratio of sediment coefficients

The width of sediment movement zone Ds is the distance perpendicular to the shoreline from the wave runup point on the beach to the offshore boundary where longshore transport activity becomes insignificant. The distance Ds is determined basically by investigating the volume of beach profile area change from the bathymetric data of the coast in question. When the available data are inadequate, an energy-averaged representative wave is estimated and its dimensions are substituted into the equations for the runup height and the threshold depth of sediment movement as a method to conveniently find the distance Ds. Because equation (10.3.2) cannot be solved analytically except in extremely simple cases, a computer is required to perform the numerical computation. In the numerical computation Q must be evaluated at each measuring line. For this purpose the breaker height, angle of incidence to the shoreline, and breaker depth at each measuring line must be calculated using a separate wave deformation calculation. Many kinds of models (profile change, shoreline change, three-dimensional change) have been proposed in the past for prediction of various topographical changes. Several representative models have been tested for mutual comparison. The results of comparative tests indicate that when applying the various models to a field site, they must be used properly to fit the mechanism of beach deformation at the site chosen and the scale of time and area. [References] 1) Shoji SATO: “A study of littoral drift related to harbor construction”, Tech. Note of PHRI, No. 5, 1962, 156p. (in Japanese). 2) Sato, S. and N. Tanaka: “Field investigation on sand drift at Port Kashima facing the Pacific Ocean”, Proc. 10th Int. Conf. Coastal Engrg., ASCE, 1966. 3) Savage, R. P.: “Laboratory determination of littoral-transport rate”, Proc. of ASCE., Vol. 88, No. WW2, 1962. 4) U. S. Army Corps of Engineers, Coastal Engineering Research Center: “Shore Protection Manual”, Vol. 1, 1977. 5) Katoh, K., N. Tanaka and I. Irie: “Field observation on suspended-load in the surf zone”, Proc. 19th Int. Conf. Coastal Eng., 1984, pp. 1846-1862. 6) Komar, P. D.: “Beach Process and Sedimentation”, Prentice-Hall, Inc., 1976. 7) Isao IRIE, Kazuo NADAOKA, Takamichi KONDO, Kenji TERASAKI: “Two dimensional seabed scour in front of breakwaters by standing waves - a study from the standpoint of bedload movement -”, Rept. PHRI, Vol. 23, No. 1, 1984, pp. 3-52 (in Japanese). 8) Hiroaki OZASA, Alan H. BRAMPTOM: “Models for predicting the shoreline evolution of beaches backed by seawalls”, Rept. PHRI, Vol. 18, No. 4, 1979, pp. 77-103 (in Japanese).

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Chapter 11 Subsoil 11.1 Method of Determining Geotechnical Conditions 11.1.1 Principles Geotechincal parameters shall be determined appropriately through site investigation and soil tests. [Commentary] The geotechnical conditions for construction include the depth of the bearing layer, thickness of soft layers, density condition, shear properties, deformation characteristics, consolidation characteristics, permeability, groundwater level (residual water level), etc. Geotechical properties of subsoil drastically change as a result of the passage of consolidation time or the change in load pressure. When using old information obtained at the reference site, it is important to confirm that the ground conditions have not changed due to the changes in load pressure or degree of consolidation. [Technical Notes] The location, interval, and depth of soil investigation should be determined by taking into consideration the size of the structure, the stress distribution in subsoil due to the weight of the structure, and the uniformity of the subsoil conditions. It is difficult to determine the number of investigation points or their depths a priori, because they depend on construction cost or importance of the structure. Subsoil uniformity is the most important factor for determining the number of investigation points. Existing data such as the results of past investigations, land topography, and acoustic survey are very useful to verify subsoil uniformity. Mechanically determining the intervals of investigation points without consideration of local conditions should be avoided. Guidelines for the intervals of boring or sounding investigation points are given in Table T- 11.1.1 for reference. The investigation depth should be sufficiently deep to confirm presence of the strata that has sufficient bearing capacity. In the case of a structure of a relatively small size or a structure other than pile foundation, the investigation is to be completed upon confirmation of several meters of the layer with N≧30, and with N≧50 in the case of a large structure where the pile foundation is expected, where N refers to the number of blows in the standard penetration test. Table T- 11.1.1 Guidelines for Intervals of Boring and Sounding Points (a) When stratification is relatively uniform both horizontally and vertically (Units: m) Along the face line of structure

Perpendicular to the face line of structure

Intervals

Preliminery investigation

Boring

Sounding

Largescale area

300 ～ 500

100～　300

Smallscale area

50 ～ 100

20 ～ 50

50 ～ 100

20 ～ 50

Detailed investigation

Maximum distance from the face line of structure

Intervals Boring

Sounding

50

25

20 ～ 30

10 ～ 15

Boring

Sounding

50 ～ 100

(b) When stratification is complicated (Units: m) Along the face line of structure

Perpendicular to the face line of structure

Intervals

Preliminery investigation Detailed investigation

Maximum distance from the face line of structure

Intervals

Boring

Sounding

Boring

Sounding

～ 50

15 ～ 20

20 ～ 30

10 ～ 15

10 ～ 30

5 ～ 10

10 ～ 20

5 ～ 10

Boring

Sounding

50 ～ 100

Note: Some soundings require boring holes, while some do not. The values shown for soundings in the table are those not requiring a bore hole. For soundings requiring a bore hole, values in the columns of “boring” are to be used. -167-


11.1.2 Selection of Soil Investigation Methods Soil investigation methods shall be so determined that soil parameters necessary for design and construction plan of the structures can be secured, while considering the type, size, and importance of the structures as well as the soil properties in the neighboring sites. [Commentary] Soil investigation methods should be selected most appropriately for their purpose, considering the area and cost of investigations, and the importance of the structure. [Technical Notes] The soil investigation methods are classified by the investigation purpose and the soil parameters in question as listed in Table T- 11.1.2. Table T- 11.1.2 Soil Investigation Methods and Soil Parameters. Purpose

Investigation method

Soil parameters

Boring

Depth of bearing layer

Sounding

Thickness of soft layer

Geophysical exploration

Stratification

Bearing capacity

Undisturbed sampling

Unconfined compressive strength

qu

Slope stability

Sounding

Shear strength

Field test

Angle of shear resistance

tf f

Relative density

Dr

Verification of conditions of stratification

Earth pressure Consolidation characteristics

Undisturbed sampling

Coefficient of consolidation Coefficient of volume compressibility

cv mv

Permeability

Undisturbed sampling Field test

Coefficient of permeability

k

Compaction characteristics

Disturbed sampling is allowed Field test

Maximum dry density Optimum moisture content CBR

Classification

Undisturbed sampling (Disturbed sampling is allowed except for g t )

Unit weight Moisture content Soil particle density Gradation Consistency

g dmax wopt

gt

w

rs

wL, wP

11.1.3 Standard Penetration Test (Notification Article 10, Clause 5) The N-values of the “standard penetration test” for soils shall be determined following the JIS “Standard Penetration Test Method for Soils”. [Commentary] The N-value determined by the “standard penetration test” is extensively used in Japan. The N-value is measured according to the JIS A 1219 “Standard Penetration Test Method for Soils”. The standard penetration test can be applied to various soils except those such as bedrock, cobblestones and coarse gravel. The test is less precise, however, for a layer of soft clay or for a layer containing gravel with a grain size of 10 mm or greater. Since there are suitable investigation methods for alluvial clay, such as laboratory tests with undisturbed sample or vane shear test in situ, evaluation of the soil parameters of alluvial clay by means of the N-value should be avoided.

11.2 Physical Properties of Soils 11.2.1 Unit Weight of Soil The unit weight of soil shall be determined using undisturbed soil samples or shall be measured directly in situ.

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11.2.2 Classification of Soils (Notification Article 10, Clause 2) Soil classification shall be made by gradation for coarse soils and by consistency for fine soils. [Commentary] Mechanical properties of soil such as strength or deformability have a close relationship with the soil gradation for coarse soils, and with the consistency for fine soils. [Technical Notes] Engineering Classification Method for Subsoil Materials (Japanese Unified Soil Classification System) The classifying method of soil and rock, and their nomenclature should be in accordance with the engineering classification method for subsoil material prescribed by the JGS 0051-2000 “Japanese Unified Soil Classification System” of the Japanese Geotechnical Society. The grain size classifications and their names are shown in Fig. T11.2.1 1). The coarse-grained soil refers to soil composed mainly of coarse fraction with a grain size ranging from 75 mm to 75 mm. Soil consisting of components with a grain size less than 75 mm is called the fine-grained soil. 5 mm Clay

75mm Silt

Fine materials

250mm Fine sand

425mm

850mm

Medium sand Sand

2mm

Coarse sand

4.75mm

Fine gravel

19mm

Medium gravel Gravel

75mm

Coarse gravel

Coarse materials

300mm (grain size)

Cobblestone (cobble)

Large rocks (boulders) Rock Rock materials

Fig. T- 11.2.1 Grain Size Classification and Nomenclature 1)

The uniformity coefficient is an index showing the grain size characteristics of sandy soil and is defined by equation (11.2.1). (11.2.1) U c = D 60 ¤ D 10 where Uc： uniformity coefficient D60： grain size correrponding to 60 perent passing of mass in grain size distribution curve (mm) D10： grain size corresponding to 10 perent passing of mass in grain size distribution curve (mm) A large uniformity coefficient means that the grain size is broadly distributed, and such a soil is labeled as “well graded”. In contrast, a small value of Uc means that the grain size distribution is narrow or the grain size is uniform. Such a soil is labeled as “pooly graded”. In the “Japanese unified soil classification system”, coarse soil where fine contents are less than 5% of the total mass is further divided into “broadly-distributed soil” and “uniformed soil”. Broadly-distributed soil: Uc ≧　10 Uniformed soil: Uc < 10

11.2.3 Coefficient of Permeability of Soil When the seepage flow in a completely saturated ground is a steady laminar flow, the coefficient of permeability shall be estimated by using Darcy’s law. [Commentary] The coefficient of permeability k is calculated by equation (11.2.2). q k = --------i×A where k： coefficient of permeability (cm/s) q： volume of water flow in soil in unit time (cm3/s) i： hydraulic gradient, i = h ¤ L h： head loss (cm) L： length of the seepage path (cm) A： cross-sectional area (cm2)

(11.2.2)

The measurement of k can be carried out for sampled soil by a laboratory permeability test, or a in-situ permeability test. [Technical Notes] An approximate value for the coefficient of permeability can be obtained as follows. Hazen showed that the effective grain size D10 and the permeability of sand k are related, and gave equation (11.2.3) to calculate k of relatively -169-


uniform sand with the uniformity coefficient of Uc less than 5, and the effective grain size D10 from 0.1 mm to 0.3 mm. (11.2.3) k = CD 10 2 where k： coefficient of permeability (cm/s) C： constant (C = 100 (1/cm/s)) D10： grain size corresponding to 10 percentage passing of mass in grain size distribution curve (called as the effective grain size) (cm) Terzaghi has pointed out that equation (11.2.3) can also be applied to cohesive soils by using C ≒ 2. The approximate values of the coefficient of permeability are listed in Table T- 11.2.4 2). Table T- 11.2.4 Approximate Values of Coefficient of Permeability 2) Soil Coefficient of permeability

Sand

Silt

Clay

10-2 cm/s

10-5 cm/s

10-7 cm/s

11.3 Mechanical Properties of Soils 11.3.1 Elastic Constants When analyzing a ground as an elastic body, the elastic constants shall be determined with due consideration for the nonlinearity of stress-strain relation of soils. [Commentary] When analyzing a ground as an elastic body, the deformation modulus and Poisson’s ratio are normally used as the elastic constants. Because of the strong nonlinearity of stress-strain relation of soil, the elastic constants in design must be determined by considering the strain level of the ground to be analyzed. [Technical Notes] (1) Strain Dependency of Deformation Modulus The stress-strain relation of soil usually shows a strong nonlinearity. When the strain level is within a range of 10-5 or less, the deformation modulus is largest and nearly constant. This maximum value Emax is called the dynamic elasticity modulus, because it has been measured by the dynamic testing methods such as the elastic wave exploration. As the strain level increases, the elasticity modulus decreases. The secant modulus E50, determined from a conventional unconfined compression test or a triaxial compression test, is the deformation modulus when the strain is of the order of 10-3. When conducting an elastic analysis of ground, it is necessary to determine the elastic constant by considering the strain level of the ground. (2) Relationship between Undrained Shear Strength and Deformation Modulus For cohesive soils, the approximate values for the initial tangent modulus Ei, which can be regarded as the deformation modulus corresponding to a strain level of 0.2 ~ 0.5%, and the secant modulus E50 can be determined by using equation (11.3.1) and equation (11.3.2) 3). Ei E50 where Ei ： E50： cu ：

= 210cu = 180cu

(11.3.1) (11.3.2)

initial tangent modulus (kN/m2) secant modulus (kN/m2) undrained shear strength (kN/m2)

The equation (11.3.1) is applicable only for highly structured marine clay with high plasticity. (3) Poisson’s Ratio For determining Poisson’s ratio of soil, there is no established method currently, although a number of methods have been proposed. Practically, v = 1/2 is used for undrained conditions of saturated soil, and v = 1/3 ~ 1/2 is used for many other situations.

11.3.2 Consolidation Properties (Notification Article 10, Clause 3) The coefficient of volume compressibility and other parameters that are used for the prediction of consolidation settlement of soft ground based on the theory of consolidation shall be obtained by laboratory consolidation tests of undisturbed samples according to the JIS A1217-193 “Test Method for One-Dimensional Consolidation Properties of Soils”. -170-


[Commentary] The standard laboratory consolidation test is prescribed by the JIS A 1217-1993 “Test Method for One-Dimensional Consolidation Properties of Soils. [Technical Notes] The consolidation settlement consists of the primary consolidation and the secondary (delayed) consolidation. The former is the settlement that accompanies the dissipation of excess pore water pressure generated due to the increase of overburden pressure. The latter is the settlement that continues after the dissipation of the excess pore water pressure. When port and harbor structures are constructed on normally-consolidated marine deposits, the most of the settlement is due to the primary consolidation and the effect of secondary consolidation is negligible. In some cases, however, the prediction of secondary settlement is necessary. The primary consolidation settlement is determined by the calculation of final settlement and the settlement rate. (1) Calculation of Final Settlement by Consolidation By plotting the consolidation pressure p and the void ratio e after the consolidation is completed on semi-logarithmic coordinates, a socalled “e-log p curve” is obtained as shown in Fig. T- 11.3.1. The relationship between the void ratio e for the segment abc in Fig. T- 11.3.1 and the pressure p is expressed by equation (11.3.3). p2 (11.3.3) e 2 = e 1 C c log 10 ----p1 where Cc is the degree of inclination of the segment abc and is called the compression index. Using equation (11.3.3), the decrease in void ratio Δe, when consolidation pressure increases from p1 to p2, is given by equation (11.3.4). p2 (11.3.4) Fig. T- 11.3.1 e-log p Relationship of Soil De = e 1 e 2 = C c log 10 ----p1 When the consolidation pressure is applied to soft ground, the final consolidation settlement can be calculated using three methods: e-log p curve method, m v (coefficient of volume compressibility) method, and Cc method. In the e-log p curve method, Δe is determined from equation (11.3.4) and the settlement S is calculated by the following equation: De (11.3.5) S = h -------------1 + e0 In the m v method, S is calculated by the following equation: S = mv・Dp ・ h (11.3.6) where S： final consolidation settement (m) mv： coefficient of volume compressibility when the consolidation pressure is (p0+1/2Δp) (m2/kN) e0： void ratio of soil in situ p0： overburden pressure in situ (kN/m2) Δp： increase in consolidation pressure (kN/m2) h： thickness of layer (m) When the soil is normally consolidated, the value of m v decreases with the increase of consolidation pressure, showing an almost straight line when plotted on logarithmic coordinates. The m v method has been commonly used in practice, because the determination of m v is easy when the subsoil is normally-consolidated clay. However, in the case of Pleistocene clay layer, the determination of m v value should be carried out carefully, because the subsoil is usually at the boundary between the over-consolidated stage and the normally consolidated stage. In the Cc method, S is calculated by the following equation; Cc p 0 + Dp S = h -------------- log10 -----------------p0 1 + e0 where Cc： compression index

(11.3.7)

(2) Rate of Settlement The consolidation is the time-dependent settlement phenomenon. The rate of consolidation for an entire clay layer is represented with the parameter U for the average degree of consolidation. The relationship between U and the nondimensional time factor Tv is obtained by the theory of consolidation. The relationship between the nondimensional time factor Tv and the actual time t is shown by the following equation:

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c v ·• t T v = ----------H* 2 where Tv： time factor cv： coefficient of consolidation (cm2/d) t ： time after the consolidation starts (d) H*： maximum drainage distance (cm)

(11.3.8)

Degree of consolidation

Average degree of consolidation U (%)

Fig. T- 11.3.2 Consolidation Isochrones

Time factor Tv

Fig. T- 11.3.3 Theoretical Relationship between Average Degree of Consolidation and Time Factor

When the permeable layer exist at both sides of the clay layer, the maximum drainage distance H* is the same as H (half of the thickness of the layer). However, when the permeable layer only exists in one side, H* is equal to 2H. The degree of consolidation at each depth is shown by the consolidation isochrones in Fig. T- 11.3.2. Furthermore, Fig. T- 11.3.3 shows the theoretical relationship between the average degree of consolidation and the time factor. (3) Secondary Consolidation The progress of consolidation with the lapse of time is exemplified in Fig. T-11.3.4. In the following cases, secondary consolidation must be taken into consideration at design stage. (a) The ground settlement may give an serious influence on structures in the long term after the primary consolidation is completed. (b) The consolidation pressure exceeds the consolidation yield stress of the soil layer but not by a large margin, (e.g., in the case of the settlement of Pleistocene clay in deep layers), and the contribution of secondary consolidation is not negligible. -172-

Average degree of consolidation U (%)


Primary consolidation Theoretical curve Actual curve Secondary consolidation

t 100

Fig. T- 11.3.4 Primary Consolidation and Secondary Consolidation

(4) Consolidation Settlement in Very Soft Clay When the landfill is carried out with dredging or disposed sludge, it is necessary to predict the consolidation settlement of extremely soft deposits. Mikasa’s consolidation theory can be applied to this problem, because the theory takes into account the effect of deadweight of clay layer and the changes in layer thickness that are ignored in the conventional theory of consolidation 4). To use Mikasa’s consolidation theory, the time-settlement relationship must be calculated with the finite difference method. To obtain the consolidation parameters for extremely soft deposits correctly, the constant strain rate consolidation test can be utilized 5).

11.3.3 Shear Properties (Notification Article 10, Clause 4) To determine the shear strength parameters of soil, soil shall be classified into sandy soil and cohesive soil. The shear strength of sandy soil shall be determined under drained conditions, while the shear strength for cohesive soil shall be determined under undrained conditions. [Commentary] In general, the coefficient of permeability of sandy soil is 103 ~ 105 times that of cohesive soil. For sandy soil layer, the excess water in pores is considered to be completely drained during construction. For cohesive soil layer, on the other hand, almost no drainage is expected during construction because the soil permeability is low. Thus in many cases for sandy soil layer the shear strength is evaluated using the frictional angle in drained condition fd and the cohesion in drained condition cd. Because the value of cd is usually very small, practically cd is ignored and only fd is used as the strength parameter. In the case of saturated cohesive soil layer, the shear strength of the layer undergoes almost no change between before and after construction, as the drainage cannot take place during construction. The undrained shear strength before construction is therefore used as the strength parameter. For intermediate soil that has the permeability somewhere between those of sandy soil and cohesive soil, the soil should be viewed as sandy soil or cohesive soil based on the coefficient of permeability and construction conditions. The shear strength is then determined using an appropriate testing procedure. [Technical Notes] (1) Shear Strength Shear strength of soil is greatly dependent upon whether a volume change of soil (in the case of saturated soil, suction or drainage of pore water) takes place during the shear or not. The drainage condition is classified into the following three categories and the different strength parameters are used for each case: ① Unconsolidated, undrained condition (UU condition) ② Consolidated, undrained condition (CU condition) ③ Consolidated, drained condition (CD condition) The shear strength used for ground design should be that for the most dangerous drainage conditions expected under the given load. The drainage condition and shear strength are then as in the following: (a) When loading takes place rapidly on the cohesive soil ground: Because consolidation progresses and shear strength increases with the passage of time, the most dangerous time is immediately after the loading when almost no drainage has occurred (this is called the short-term stability problem). The shear strength tf at this time is the shear strength cu that is determined from unconsolidated undrained (UU) tests using the sample before loading. The parameter cu is also called the apparent cohesion, and the analysis using ?u is also called the “f = 0 method”. Constructions of seawalls or breakwaters (without excavation), landfill, and embankments on soft cohesive soil ground are in this category.

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(b) When ground permeability is large or when drainage from consolidated layer is almost completed during construction period because the loading is carried out very slowly: Because drainage from the layer occurs simultaneously with loading and an increase in strength of the layer is expected along with the loading, design of structures should be carried out using cd and fd determined under consolidated, drained (CD) conditions. Constructions of seawalls or breakwaters (when there is no excavation), landfill and embankments on sandy soil belong to this category. (c) When ground permeability is poor and the load is removed to decrease the stress s normal to the shear plane: In this case, the most dangerous situation is after a long time is elapsed, when the soil absorbs water, expands, and loses its shear strength (this is called the long-term stability problem). Among the three categories the consolidated, undrained shear strength becomes lowest after water absorption and soil expansion when the overconsolidation ratio is small (s is a little less than pc). In this situation, therefore, the cu value should be used with consideration of soil expansion. Earth retaining and excavation in clayey ground or removal of preloading on cohesive soil ground belongs to this category. On the other hand, in the case of heavily overconsolidated ground where s is very small compared to pc, the parameters cd and fd are used for design because the shear strength under consolidated, drained condition is the smallest. In almost all cases for normal construction conditions of port and harbor structures, the undrained strength in UU conditions of (a) is used in design for cohesive soils and the strength parameter in the CD conditions of (b) is used for sandy soils. The following equations show the strength calculation methods respectively: For cohesive soil (the sand content is less than 50%) t = cu where t： shear strength (kN/m2) cu： undrained shear strength (kN/m2) For sandy soil (the sand content is higher than 80%) t = ( s u ) tan f d where t： shear strength (kN/m2) s： normal stress to shear plane (kN/m2) u： steady water pressure at the site (kN/m2) fd： angle of shear resistance for drained conditions (º)

(11.3.9)

(11.3.10)

Furthermore, because soil with a sand fraction ranging from 50% ~ 80% displays intermediate characteristics between sandy soil and cohesive soil, it is called the intermediate soil. The evaluation of shear strength of intermediate soil is difficult compared with that of sandy soil or cohesive soil. Hence, the shear strength for such soil should be evaluated carefully by referring to the most recent research results. (2) Shear Strength of Sand Because sandy soil has high permeability and is regarded in completely drained condition, the shear strength of sand is represented by equation (11.3.10). The angle of shear resistance fd for drained conditions can be determined using a triaxial CD (consolidated, drained) test. Because the value of fd becomes large when sand’s void ratio becomes small and its density becomes high, the void ratio e0 in situ must be accurately determined. Therefore, it is best to take and test an undisturbed sample. Although the fd values of sand with the same density will vary a little with the shear conditions, the value of fd determined by a triaxial CD test, which is conducted with the confining pressure corresponding to design conditions with undisturbed sample, can be used as the design parameter for stability analysis. However, in the case of bearing cabacity problem for foundation, which is much influenced by progressive failure, the bearing capacity is overestimated in some cases if the value of fd determined by a triaxial CD test is directly used as the design parameter. Compared with the case of cohesive soil, sampling of undisturbed sand samples is technically difficult and also very expensive. This is the reason that the shear strength for sandy soil is frequently determined from the Nvalue of “Standard Penetration Test” rather than from a laboratory soil test. For the equation to determine fd from N-values, 11.4 Angle of Internal Friction by N-value is referred to. (3) Shear Strength of Cohesive Soil Soil in which the clay and silt fraction by percentage is greater than 50% is regarded as cohesive soil. There are several methods, as presented below, to determine the undrained shear strength cu of cohesive soil. An appropriate method should be chosen in consideration of such factors as past experiences, subsoil characteristics, the importance of the structures, etc. (a) qu method This method uses the average value of unconfined compressive strength qu determined from undisturbed samples. The undrained shear strength cu used for design is given by the following equation: (11.3.11) cu = qu ¤ 2 -174-

PART II DESIGN CONDITIONS In this equation, q u is the average value of unconfined compressive strength (kN/m2). (b) Methods combining unconfined compressive strength and strength from triaxial compression tests One problem with the qu method is that the test’s reliability depends on the skill of the technician, because the test is subject to the influence of disturbance during sampling. To resolve this problem, a combination method can be used to determine the strength by comparing the qu of undisturbed samples with the strength from a triaxial CU test and evaluating the quality of the sample. This method is used for natural soil ground and cannot be applied to unconsolidated, reclaimed ground. For more details see the references 6) and 7). (c) Method using strength from a box shear test This method uses the strength tDS (units in kN/m2) determined by a box shear test, after undisturbed sample is consolidated one-dimensionally under in-situ effective overburden pressure. The box shear test is conducted according to the JGS T 560-1997 of the Japanese Geotechnical Society 8). The undrained shear strength cu used for design is given by the following equation: (11.3.12) c u = 0.85t DS In this equation, 0.85 is a correction factor related to shear rate effect. (d) Method determing undrained shear strength from an in-situ vane shear test Vane shear test is described in 11.5 Application of Soundings Other Than SPT. The average value of the obtained shear strength cu (v) can be used in design as the undrained shear strength cu 9). An in-situ vane shear test can be carried out rather easily with mobility at a field site. The test can determine the shear strength for very soft clay for which an unconfined compression test cannot be performed due to the difficuly in making a specimen freestanding. It can thus be applied, for example, to site management where soil is being improved using vertical drains.

c

(4) Increase in Cohesive Soil Strength due to Consolidation The undrained strength of cohesive soil will increase with the progress of consolidation. For soil improvement methods such as the vertical drain method or sand compaction pile method, the rate of strength increase cu/p by consolidation is an important parameter because the strength is increased by the drainage of pore water by consolidation. The larger the value of cu /p, the larger the increase rate of the strength and the more effective soil improvement are expected. From the past experiences in the field and research results for marine clay in Japan, the value of cu /p lies in a range shown by the following equation, regardless of plasticity (refer to Fig. T- 11.3.5) 9). c u ¤ p = 0.28 ~ 0.30

:

triaxial compression

:

direct shear

:

triaxial extension

Plasticity index Ip

(11.3.13) Fig. T- 11.3.5 Relationship between Plasticity Index and cu/p

(5) Strength of Intermediate Soil Soil with a sand content in the range of 50% ~ 80% is intermediate soil between sandy soil and cohesive soil 10). For this type of soil, the coefficient of permeability and design conditions are taken into consideration to determine whether the soil is sandy soil or cohesive soil. Then the shear strength is determined accordingly. For intermediate soil with a large sand fraction or with coral gravel, the coefficient of permeability determined from a standard consolidation test generally gives an underestimated value, because of the limitations of test conditions. It is desirable not only to improve the test procedure, but also to conduct an in-situ permeability test or an electrical cone test to determine the coefficient of permeability 11). When the coefficient of permeability determined by this kind of procedures is greater than 10-4 cm/s, the ground is regarded permeable. Hence, the value of fd determined from an electrical cone penetration resistance or a triaxial CD test can be used as design parameters, regarding cd = 0. According to the experience on investigating the properties of intermediate soils in Japan, the value of fd is greater than 30º in many cases 12) ~ 14). When the coefficient of permeability is less than 10-4 cm/s, the intermediate soil should be regarded as a cohesive soil. Because the influence of stress release during sampling in intermediate soil is much greater than that in cohesive soil, the shear strength determined by qu method is underestimated. A correction method is used for the strength of such intermadiate soil with a large sand fraction by means of clay fraction and plasticity index 15). However, it is desirable that the combined method with unconfined compression test and triaxial compression test or the box shear test be used as the method for evaluating the strength of intermediate soil 16).

11.4 Angle of Internal Friction by N-value (Notification Article 10, Clause 6) The angle of internal friction for sandy soils shall be calculated using the following equation from a standard penetration test value. -175-


100N f = 25 + 3.2 ---------------------70 + p n o ¢

(11.4.1)

where f： angle of internal friction of sand (º) N： standard penetration test value pvo¢： effective overburden pressure at the time the standard penetration test value is obtained (kN/m2) [Commentary] Relationships between the N-value and many soil parameters have been established by the data at various sites. When using these relationships, however, it is necessary to consider the background of their derivation and the ground conditions of the data and to confirm the range of their applicability. In the “Technical Standards for Port and Harbour Facilities in Japan” (1989), the value of f was determined directly from the N-values without considering the effective overburden pressure pvo¢, as can be seen in Danham’s equation. Because the relative density Dr varies with pvo¢ as seen in Fig. C- 11.4.1, pvo¢ must be taken into consideration to determine f from an N-value.

Terzaghi

Gibbs, Holtz (dry sand, wet sand)

N-Value

Yanase (wet fine sand)

Yanase (saturated fine sand)

Overburden pressure

Relative density Dr Fig. C- 11.4.1 Influence of Effective Overburden Pressure and Relative Density on N-Values (Meyerhof) 17)

11.5 Application of Soundings Other Than SPT (Notification Article 10, Clause 7) When conducting soundings other than the “Standard Penetration Test”, a method shall be appropriately selected considering the subsoil properties, the required soil constants, and the level of precision. [Technical Notes] (1) Types of Soundings As listed in Table T- 11.5.1, there are many different types of soundings 18). When preparing a ground investigation plan, a combination of the laboratory tests and the soundings should be appropriatelly studied considering the characteristics of the subsoil of ground, the required soil parameters, and the level of precision required for design or construction. (2) Vane Shear Test A vane shear test is a test in which a cross-shaped vane is pushed into the ground and the undrained shear strength of the subsoil is determined from the vane’s torque for rotation. This test is suitable to determine the strength of subsoil in case the specimens are incapable of freestanding, such as soft cohesive soil. (3) Electrical Static Cone Penetration Test The notable characteristic of this test is its function to measure soil parameters almost continuously in the vertical direction, and thus the subsoil conditions can be estimated more precisely. This is the method of ground investigation most applicable for the case when sandy layers and cohesive soil layers are intricately mixed. -176-


Equation (11.5.1) is used to determine the undrained shear strength cu of clay from a cone penetration test. (11.5.1) c u = ( q t s v0 ) ¤ N kt where cu： undrained shear strength (kN/m2) qt： cone penetration resistance (kN/m2) s v0： overburden pressure in terms of total stress (kN/m2) Nkt： cone parameter From investigations conducted on marine clay in Japan, the value of Nkt extends over a range from 8 to 15, as shown in Fig. T- 11.5.2 19). It is well understood from past testings that the value of Nkt varies depending upon many factors. When determining cu from a cone test, therefore, a laboratory test and an electrical static cone penetration test for at least one location should be conducted to determine the value of Nkt. Table T- 11.5.1 Characteristics and Applicable Subsoils for Sounding Methods

Method

Continuity

Measured parameters

Estimated values from measured parameters

Applicable subsoils

Continuous

Settlement by each load (Wsw) Number of onehalf revolution per 1 m penetration (Nsw)

Converted Nvalue or unconfined compressive strength qu value

All subsoils except 15 m level cobblestone and gravel

Continuous

Penetration resistance

Unconfined compressive Cohesive soils strength, and or humic soils cohesion of clay

5 m level

Double tube electrical static Continuous cone penetration test

Point resistance qc, Pore water pressure u

Shear strength, soil classfication, and consolidation properties

Depends upon the capacity of Data reliability the penetration is high apparatus or fixing apparatus

In situ vane shear test

Discontinuous

Maximum resisting moment for rotation

Undrained Soft cohesive shear strength soils of cohesive soil

Discontinuous

Pressure, Hole wall displacement, Creep

Deformation modulus, initial pressure, yield stress, and undrained shear strength

All subsoils and bedrocks where the borehole Basically no wall surface is limits smooth and freestanding

Mechanical meaning of the estimated value is very clear

Discontinuous N-value Standard Minimum (number of Penetration Test interval is 50 cm hammering)

Sand density, strength, friction angle, shear modulus, bearing capacity, clay cohesion, unconfined compressive strength

All subsoils except cobblestone or boulder stone

Basically no limits

Widely employed and can be used for almost all subsurface investigations

Simple dynamic cone penetration Continuous test

Nd = (1 ~ 2) N (regarded as equivalent to the N-value)

Same as above

15 m level (lod friction becomes larger as depth increases)

Operation is simple compared to the Standard Penetration Test

Designation

Weight penetration test (Swedish sounding)

Portable cone penetration test

Static

Borehole horizontal loading test

Dynamic

Nd (number of hammering)

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Cohesive soils or sandy soils

Applicable depth (m)

15 m level

Characteristics

Operation is simple compared to the Standard Penetration Test

Simple test, very quick

Specifically for soft cohesive soil; direct measurement of cu


Kurihama

Izumo

Ogi-shima

Hachirogata

Kuwana

Shinonome

Nkt

Tamano

Ip

Fig. T- 11.5.2 Cone Parameters of Marine Clay in Japan (Undrained shear strength was determined from unconfined compression tests) 19)

11.6 Dynamic Properties of Soils 11.6.1 Dynamic Modulus of Deformation (Notification Article 10, Clause 8) For seismic response analysis, an appropriate dynamic modulus of deformation of soils shall be determined to prescribe the relationship between the shear stress and shear strain of soil. [Commentary] Seismic design methods can be broadly classified into the static design methods and the dynamic design methods. One example of static design methods is the seismic coefficient method. In this method the seismic force is assumed to act on the ground or structures in the form of a static inertia force, and stability is investigated from the equilibrium of forces. In dynamic design methods, on the other hand, dynamic magnification factors or amplification values of acceleration, speed, and deformation of subsoils or structures against basement rock are calculated to examine the stability of ground or structures. As for the seismic response analysis method, both the time domain analysis and the frequency domain analysis are used. For either method, the relationship between the shear stress and shear strain of the soils is required. Normally the relationship between the shear stress and shear strain in ground subjected to dynamic loading is described by a skeleton curve and a hysteresis curve, as shown in Fig. T- 11.6.1. A skeleton curve will display remarkable nonlinearity as the shear strain amplitude becomes larger. Since the dynamic modulus of deformation prescribes this relationship between the shear stress and shear strain, it must be appropriately determined when conducting a seismic response analysis. [Technical Notes] (1) Relationship between Dynamic Shear Stress and Shear Strain of Soil. There are many models to apply the shear stress and shear strain curves of soil into analysis, such as the equivalent linear model, the bilinear model, the hyperbolic model (Hardin-Dornevich model), and the RambergOsgood model 20). Of these models the equivalent linear calculation method is frequently used for seismic response analyses from the standpoint of calculation time and solution stability. The applicable range of the equivalent linear model, however, is up to a strain level of 10-3. When the strain level exceeds 10-3, the results must be carefully examined. (2) Modeling of Deformation Properties in the Equivalent Linear Model To estimate the behavior of ground during an earthquake, the nonlinearity of the relationship between the dynamic stress and strain of soil for a wide range of the shear strain amplitude must be appropriately assessed and modeled. The relationship of the dynamic stress and strain of soil is expressed with two parameters: the shear modulus and the damping constant in the equivalent linner model. The shear modulus G and the damping constant h are defined with the shear strain amplitude by equation (11.6.1) and equation (11.6.2) as shown in Fig. T- 11.6.2. -178-

G is the shear modulus

Skeleton curve

h is the damping constant Shear stress τ

Shear stren τ


Shear strain γ Shear strain γ

Hysteresis curve

Fig. T- 11.6.2 Shear Modulus and Damping Constant

GG /

/

GG

Fig. T- 11.6.1 Stress Strain Curve

Shear strain amplitude

γ

Fig. T- 11.6.3 Shear Modulus, Damping Constant and Shear Strain Amplitude

t G = g

(11.6.1)

DW h = ---------------2p × W where G： shear modulus (kN/m2) t： shear stress amplitude (kN/m2) g： shear strain amplitude h： damping constant W： strain energy (kN/m2) DW： damping energy (kN/m2)

(11.6.2)

Since the values of shear modulus G and damping constant h vary nonlinearly depending on the value of g, a G/G0 ~ g curve and a h ~ g curve are normally drawn as shown in Fig T- 11.6.3, where G0 is the shear modulus at g ≒ 10-6. (3) Measurement of the Shear Modulus and the Damping Constant The shear modulus and the damping constant must be determined by laboratory tests such as the resonance test or cyclic triaxial test, or by the in-situ tests using elastic waves such as the PS logging method or the cross hole method. The laboratory tests can be used to measure the shear modulus and damping constant for a wide range of shear strain amplitudes from 10-6 to 10-2 although undisturbed samples from the field must be obtained. The tests can also be used to evaluate the change in the modulus of dynamic deformation due to construction of structures. With the cyclic triaxial test, the shear modulus is determined from equation (11.6.3) with Poisson’s ratio v. sa G = ------------------------(11.6.3) 2e a ( 1 + v ) where s a： axial stress amplitude (kN/m2) e a： axial strain amplitude For v, the value of 0.33 is normally used for a drainage condition and 0.45 is used for an undrained condition. The damping constant is calculated from equation (11.6.2) with W and DW obained from the stress - strain curve such as shown in Fig. T- 11.6.2. -179-

TECHNICAL STANDARDS AND COMMENTARIES FOR PORT AND HARBOUR FACILITIES IN JAPAN In-situ tests are limited to measurements of the shear modulus that only corresponds to 10-6 level of shear strain amplitude. Such tests have not been put to practical application to measure the shear modulus and damping constant for the large shear strain amplitude. But the tests do possess the advantage of being able to measure the values in situ directly. They are also used to calibrate the shear modulus taken from laboratory tests. The elastic constant of subsoil is obtained by equations (11.6.4) to (11.6.6) from the data of elastic wave velocity measurements by a seismic exploration using bore holes. gt 2 2 (11.6.4) G 0 = rV s = --- V s g

E 0 = 2 ( 1 + v )G 0

(11.6.5)

V 2 æ -----p-ö 2 è Vsø v = --------------------------------ì Vp 2 ü 2 í æ ------ö 1 ý è Vsø î þ where V p： longitudinal wave velocity (m/s) V s： transverse wave velocity (m/s) G0： shear modulus (kN/m2) E0： Young’s modulus (kN/m2) v： Poisson’s ratio r： density (t/m3) g t： wet unit weight (kN/m3) g： gravitational acceleration (m/s2)

(11.6.6)

11.6.2 Dynamic Strength Properties Soil strength against dynamic external forces shall be determined through laboratory tests. When doing so, the properties of the external forces and the subsoil conditions shall be appropriately set in. [Commentary] The typical dynamic external forces encountered in ports and harbors are seismic force and wave force. Seismic forces are characterized by a short period and few cyclic repetitions, while wave forces are characterized by a long period and many cyclic repetitions. Under present circumstance these dynamic external forces are normally converted into static loads like in the seismic coefficient method. There are situations, however, in which it is necessary to handle them as dynamic loads like in liquefaction analysis or in strength decrease analysis of cohesive soil beneath structures exposed to waves. In such cases the dynamic strength of soils are normally obtained by cyclic triaxial tests. When conducting cyclic triaxial tests, the cyclic undrained triaxial test method explained in the “Soil Testing Methods and Commentary” of the Japanese Geotechnical Society should be used 21). When applying the results of cyclic triaxial tests to the liquefaction analysis of ground during an earthquake, refer to the “Prediction and Determination of Liquefaction” in the “Handbook on Liquefaction Remediation of Reclaimed Land” (Revised Edition) 22). [References] 1) Japanese Geotechnical Society: “Revised Standard of Japanese Geotechnical Society and Commentary, Engineering Classification Method for Subsoil Material (Japanese Unified Soil Classification System)”, Nov., 1996 (in Japanese). 2) K. Terzaghi, and P. B. Peck: “Soil Mechanics in Engineering Practice”, John Wiley and Sons Inc., New York, 1948, p. 44. 3) Akio NAKASE, Masaki KOBAYASHI, Akio KANECHIKA: “Undrained shear strength and deformation modulus of clays”, Rept. of PHRI, Vol. 11, No. 2, 1972, pp. 243-259 (in Japanese). 4) Masato MIKASA: “Consolidation of Soft Clay”, Kajima Publisher, 1996 (in Japanese). 5) Yasufumi UMEHARA: “Study on the consolidation characteristics of soils and consolidation test methods”, Tech. Note of PHRI, No. 469, 1983 (in Japanese). 6) Takashi TSUCHIDA, Jun-ichi MIZUKAMI, Ken OIKAWA, Yoshio MORI: “New method for determining undrained strength of clayey ground by means of unconfined compression test and triaxial test”, Rept. of PHRI, Vol. 23, No. 3, 1989, pp. 81-145 (in Japanese). 7) Takashi TSUCHIDA: “Study on determination of underained strength of clayey ground by mean of triaxial tests”, Tech. Note of PHRI, No. 688, 1991 (in Japanese). 8) Japanese Geotechnical Society: “Method for Consolidated Undrained Box Shear Test of Soil” in “Newly Established Standard of Japanese Geotechnical Society and Commentary Ⅳ (1997 version)”, pp.15-58 (in Japanese). 9) Hiroyuki TANAKA, Masanori TANAKA: “Detemination of undrained shear strength of clayey ground measured by vane shear tests”, Rept. of PHRI, Vol. 33, No. 4, 1994, pp. 1-17 (in Japanese). 10) Susumu KURATA and Toshio FUJISHITA: Studies on the engineering properties of sand-clay mixture, Rept. of Transportation Technical Research Institute, Vol. 11, No. 9, 1961, 36 p. (in Japanese).

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11) Hiroyuki TANAKA, Motoo SAKAKIBARA, Kenji GOTO, Kouji SUZUKI, Takeshi FUKAZAWA: “Properties of Japanese normally consolidated marine clays obtained from static piezocone penetration test”, Rept. of PHRI, Vol. 31, No. 4, 1992, pp. 61-92 (in Japanese). 12) Masaki KOBAYASHI, Takashi TSUCHIDA and Takeshi KAMEI: “Intermediate soil-sand or clay?-”, Geotech. Note 2, Japanese Geotechnical Society, 1992 (in Japanese). 13) Takashi TSUCHIDA, Masaki KOBAYASHI, Shusuke IFUKU and Isao FUKUDA: “Engineering properties of coral soils in Japanese South Western Islands”, Int. Conf. on Calcarious Soils, 1988. 14) Takashi TSUCHIDA: “Consolidation, compression and permeability characteristics of intermediate soil and mixture”, Soils and Foundations, Vol. 41, No. 7, 1993 (in Japanese). 15) Akio NAKASE, Masaru KATSUNO, Masaki KOBAYASHI: “Unconfined compression strength of soils of intermediate grading between sand and clay”, Rept. of PHRI, Vol. 11, No. 4, 1972 (in Japanese). 16) Hiroyuki TANAKA, Masanori TANAKA and Takashi TSUCHIDA: “Strength characteristics of naturally deposited intermediate soil”, Jour. JSCE, No. 589/ Ⅲ-42, 1998 (in Japanese). 17) G. G. Meyerhof: “Discussion on soil properties and their measurement”, Discussion2, Proc. of the 4th International Conference on Soil Mechanics and Foundation Engineering, Vol. 3, 1957, p. 110. 18) Japanese Geotechnical Society: “Method of Ground Investigation”, 1995, 191 p. (in Japanese). 19) Hiroyuki TANAKA, Masanori TANAKA: “A site investigation method using cone penetration and dilatometer tests”, Tech. Note of PHRI, No. 837, 1996 (in Japanese). 20) Kenji ISHIHARA: “Fundamentels of Geodynamics”, Kajima Publishers, pp. 7-34, 1976 (in Japanese). 21) Japanese Geotechnical Society: “Method of Soil Testing and Commentary”, 1990, pp. 421-450. (in Japanese). 22) Coastal Development Institute of Technology: “Handbook on Liquefaction Remediation of Reclaimed Land (Revised Edition)”, 1997, pp. 114-136. (in Japanese).

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Chapter 12 Earthquakes and Seismic Force 12.1 General In the design of port and harbor facilities, the effect of earthquakes shall be carefully examined so that the facilities retain appropriate seismic resistance. [Commentary] (1) The following matters should be taken into consideration when investigating the seismic resistance of structures: (a) (b) (c) (d)

Level of seismicity in the region, earthquakes to be examined in the analysis, earthquake ground motion, etc. Ground conditions at the construction site Importance of the facilities (based on a comprehensive evaluation of social and economic importance) Earthquake resistance of the facilities

(2) Investigation items related to analysis of seismic resistance are as follows: (a) (b) (c) (d) (e)

Stability of the entire structure Stability against subsoil sliding Influence of liquefaction on the subsoil stability and structures Stress within structural elements From the standpoint of serviceability, the relative displacement of the various parts of a structure or that between adjacent structures or subsoil

(3) When the Port of Kobe was hit by the “Hyogoken-Nanbu Earthquake”, most of the quaywalls were suffered identical damage because almost all of the quaywalls were built in an identical structural type. If structural types were different, their seismic response characteristics would have differed, and the extent of damage to structures would have been diverse. This experience during the “Hyogoken-Nanbu Earthquake” strongly suggests that structured types should be diversified to ensure different seismic response characteristics in the facilities when selecting the structural types of port and harbor facilities. [Technical Notes] The seismic resistance of port and harbor facility must be examined with consideration of the dynamic characteristics of the structures using the methods shown in 12.3 Seismic Coefficient Method, 12.5 Seismic Response Analysis, 12.6 Seismic Deformation Method, or combinations of the methods. For structures such as gravity type quaywalls that are comparatively rigid and their amplitudes of vibration is small compared with the ground motion during an earthquake, the seismic resistance should be examined using the seismic coefficient from 12.4 Design Seismic Coefficient and the method shown in 12.3 Seismic Coefficient Method. For structures having a natural period close to the predominant period of seismic motion and a small damping characteristics, or for structures having a comparatively long natural period, the seismic resistance should be examined using the modified seismic coefficient method shown in 12.3 Seismic Coefficient Method in consideration of the dynamic response characteristics of structures. The seismic design method for open-type wharves on with vertical piles using the modified seismic coefficient method is described in Part Ⅷ　, 9.6 Examination of Earthquake-Resistance Performance. For structures whose stability is subject to the deformation of the surrounding earth, such as immersed tunnels, pipelines and other pipe-type structures that are buried in the earth, the seismic resistance should be examined in accordance with 12.6 Seismic Deformation Method. For very important structures or structures for which past examples of damage are rare, the seismic resistance using seismic response analysis after appropriately modeling related conditions such as the structure, ground or earthquakes should be examined, in addition to an investigation based on the seismic coefficient method, modified seismic coefficient method, or seismic deformation method.

12.2 Earthquake Resistance of Port and Harbor Facilities in Design (Notification Article 13 and Article16) (1) Earthquake Resistance of Port and Harbor Facilities (a) Port and harbor facilities shall be capable of retaining their required structural stability without losing their function when subjected to the “Level 1” earthquake motion (earthquake motion with a high probability of occurrence during the lifetime of facilities). (b) High seismic resistant structures (particularly important facilities whose seismic resistance is to be reinforced) shall be the structures that will sustain only slight damage during the “Level 2” earthquake motion (earthquake motion that has a vety low probability of occurrence during the lifetime of facilities, but which is very large when it occurs) and whose functions can be quickly -182-


restored after a Level 2 earthquake and are able to retain their expected function througout the rest of its lifetime. (2) Examination of Earthquake Resistance of High Seismic Resistant Structure (a) In any investigation of the structural stability of high seismic resistant structures, the earthquake resistance shall be examined to ensure required earthquake resistance against seismic load, based upon the provisions in 12.4 Design Seismic Coefficient. The earthquake motion used for such examination shall be the Level 2 earthquake motion defined in (1)(b) above. (b) For examination of the earthquake resistance, the subsoils and the structures shall be appropriately modeled corresponding to the earthquake motion indicated in (2)(a) above and the resistance should be examined using an appropriate method considering the structural characteristics. (c) The earthquake motion used for examination of earthquake resistance shall be determined based on a dynamic analysis of the ground. [Commentary] For the seismic design of port and harbor facilities, earthquake motion with a 75-year return period should be used as the “Level 1” earthquake motion. Earthquake motion from an inter-plate earthquake or a plate earthquake near the coast should be used as “Level 2” earthquake motion, of which the return period will be several hundred years or more. “High seismic resistant structures” refers to facilities such as high seismic resistant quaywalls whose seismic resistance has been reinforced to ensure the shipment of emergency supplies immediately after an earthquake in order to maintain social and economic activities. It also refers to high seismic resistant revetments designed for the area designated for refuge of people and rescue operation after an earthquake. “Without losing their function” means that the facility preserves its initial structural stability. “To retain their expected function” means that even if damage occurs to the facility it is minor and the facility’s function can quickly be restored after the earthquake. These specifications are summarized in Table C- 12.2.1. Table C- 12.2.1 Earthquake Motion and Earthquake Resistance of Port and Harbor Facilities To Be Considered for Design Ground motion level

Ground motion considered for seismic design

Applicable facilities

Earthquake resistance

Level 1

Ground motion with a 75-year return period

All facilities (except facilities that are regulated according to other standards)

Do not lose their function

Level 2

Ground motion from intra-plate earthquake or ground motion from inter-plate earthquake. The return period is anticipated to be several hundred years or more

High seismic resistant structures (high seismic resistant quaywalls, high seismic resistant revetments for refuge and rescure centers, etc.). In addition, port and harbor facilities such as bridges and immersed tunnels that must take the Level 2 earthquake motion into consideration in design.

Retain their expected function

[Technical Notes] (1) The capability of high seismic resistant structures to retain their expected function against the Level 2 earthquake motion is assured by examining structural deformation and other changes through an appropriate seismic response analysis. This is because an analysis of stability of structures or subsoil using just the seismic coefficient method is not sufficient in many cases when the earthquake motion such as the Level 2 earthquake exercises immense forces. (2) The judgment whether high seismic resistant structures can retain their expected function is made by a comprehensive consideration of structural stability, function and the ease or difficulty of the quick restoration. In order to make this judgment it is necessary to determine the allowable amount of deformation of a quaywall or the like, but its precise determination is difficult at the current state of knowledge. The deformation values listed in Table T- 12.2.1 and Table T- 12.2.2 are used as reference to estimate deformation of quaywalls which allows the temporary use immediately after an earthquake. Because the values in the tables do not take into consideration the structural stability and function of cranes built on the quays, however, another study must be conducted for such cases. Among the gravity type quaywalls that were damaged by the Hyogoken-Nanbu Earthquake, some quaywalls that sustained a deformation ratio (swelling of the quaywall divided by the height) of 10% to 20% were quickly restored and put back into service 1).

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Table T- 12.2.1 Quaywall Deformation Rough Standards from the Viewpoint of Temporaly Service Suffered deformation: maximum swelling or maximum apron settlement Structural type Quaywall depth Normal use Restricted use

Gravity type quaywall -7.5 m or greater

Less than -7.5 m

0 ～ 30 cm 30 ～ 100 cm

0 ～ 20 cm 20 ～ 50 cm

Sheet pile type quaywall -7.5 m or greater 0 ～ 30 cm 30 ～ 50 cm

Less than -7.5 m 0 ～ 20 cm 20 ～ 30 cm

Table T- 12.2.2 Quaywall Deformation Rough Standards from the Viewpoint of Functionality Structure body itself

Settlement of the entire apron: Inclination: Deviation of the swelling:

20 ~ 30 cm 3o ~ 5o 20 ~ 30 cm

Apron

Settlement; differences in height of apron surface: difference in height between apron and yard: Inclination; normal slope 3% ~ 5%, reverse 0%

3 ~ 10 cm 30 ~ 70 cm

12.3 Seismic Coefficient Method (Notification Article 14) (1) It shall be standard to use the seismic coefficient method for determining the seismic load for structures having a comparatively short natural period and large damping factor. In this case, the seismic load shall be determined using the seismic coefficient prescribed in 12.4 Design Seismic Coefficient, and either one calculated by (a) or (b) below that is most damaging to port and harbor facilities shall be applied to the centers of gravity of facilities. (a) Seismic load = seismic coefficient × deadweight (b) Seismic load = seismic coefficient × (deadweight + surcharge) (2) For structures such as immersed tunnels for which the seismic coefficient method is not applicable, the seismic load shall be determined appropriately considering characteristics of the respective structures. [Commentary] The influence of the vertical component of earthquake motion on structural stability is complex, intertwining with both the structural type and the horizontal component. Thus it is a more rigorous examination procedure to use a vertical seismic coefficient to introduce seismic load in a vertical direction. For practical seismic design, however, the vertical seismic coefficient is not taken into account in order to avoid the computational complexity, because it has been shown from the results of observations that, except for the region near seismic center, the vertical component is not so large compared with the horizontal component. Based upon past studies and experience, moreover, the horizontal seismic coefficient in 12.4 Design Seismic Coefficient is considered to be determined taking into account the effect of vertical earthquake motion. It is therefore sufficient to use the horizontal seismic coefficient in order to dertermine seismic load in design of ordinary port and harbor structures 2), 3).

12.4 Design Seismic Coefficient (Notification Article 15 and Article 16) (1) The seismic coefficient shall be the horizontal seismic coefficient determined by the equation below. The coefficient value shall be expressed in two-digit figures by rounding up if the third digit is five or larger or dropping the third digit when it is less than five. Seismic coefficient = regional seismic coefficient × subsoil condition factor × importance factor In this case, the regional seismic coefficient shall be the values listed in Table 12.4.1 corresponding to the region where the port and harbor facilities are located. The factor for subsoil condition shall be the values listed in Table 12.4.2 corresponding to the type of subsoil given in Table 12.4.3, and the importance factor shall be the values listed in Table 12.4.4 corresponding to the characteristics of structures.

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Table 12.4.1 Regional Seismic Coefficient (Notification Article 15, Appendix Table 1) Regional seismic coefficient

Regional classification

Region A

Counties of Nemuro, Kushiro, Tokachi and Hidaka in Hokkaido Prefecture; Saitama Prefecture, Chiba Prefecture; Tokyo, except for Hachijo-jima and Ogasawara Islands; Kanagawa Prefecture, Yamanashi Prefecture, Shizuoka Prefecture, Aichi Prefecture, Gifu Prefecture, Fukui Prefecture, Shiga Prefecture, Mie Prefecture, Nara Prefecture, Wakayama Prefecture, Osaka Prefecture, Hyogo Prefecture

0.15

Region B

Pacific Ocean coast south of Shiriya-zaki of Aomori Prefecture; Iwate Prefecture, Miyagi Prefecture, Fukushima Prefecture, Ibaraki Prefecture, Tochigi Prefecture, Gunma Prefecture, Nagano Prefecture, Kyoto Prefecture, Kochi Prefecture, Tokushima Prefecture

0.13

Region C

Counties of Iburi, Oshima and Hiyama in Hokkaido Prefecture; Aomori Prefecture except for the Pacific Ocean coast south of Shiriya-zaki; Akita Prefecture, Yamagata Prefecture, Niigata Prefecture, Toyama Prefecture, Ishikawa Prefecture, Tottori Prefecture, Hiroshima Prefecture, Ehime Prefecture, Oita Prefecture, Miyazaki Prefecture, Kumamoto Prefecture; Amami Islands in Kagoshima Prefecture

0.12

Region D

Counties of Abashiri, Shiribeshi, Ishikari, Sorachi, Rumoi and Kamikawa in Hokkaido Prefecture; Okayama Prefecture, Shimane Prefecture, Kagawa Prefecture, Nagasaki Prefecture except for Goto Islands, Iki-no-shima and Tsushima; Saga Prefecture, Kagoshima Prefecture except for Amami Islands, Okinawa Prefecture except for Daito Islands

0.11

Region E

Counties of Soya in Hokkaido Prefecture; Hachijo-jima and Ogasawara Islands in Tokyo; Yamaguchi Prefecture, Fukuoka Prefecture; Goto Retto, Iki-no-shima and Tsushima in Nagasaki Prefecture; Daito Islands in Okinawa Prefecture

0.08

Table 12.4.2 Subsoil Condition Factor (Notification Article 15, Appendix Table 2) Type of subsoil Factor for subsoil conditions

Class Ⅰ

Class Ⅱ

Class Ⅲ

0.8

1.0

1.2

Table 12.4.3 Classification by Type of Subsoil (Notification Article 15 Appendix Table 3) Type of subsoil

Gravel stratum

Ordinary sandy soil and cohesive soil

Soft subsoil

5 meters or less

Class Ⅰ

Class Ⅰ

Class Ⅱ

More than 5 meters and less than 25 meters

Class Ⅰ

Class Ⅱ

Class Ⅲ

25 meters or greater

Class Ⅱ

Class Ⅲ

Class Ⅲ

Thickness of quaternary strata

Notes: In these tables, “soft subsoil” refers to sandy soil subsoil with the N-value of SPT test being less than 4 or cohesive soil with the unconfined compressive strength being less than 20 kN/m2. When the subsoil is composed of two or more soil layers, the type of soil should be set by the layer with predominant thickness. If the subsoil is composed of two or more layer with the almost equal thickness, the type of soil should be set by the layer having the largest value of the subsoil condition factor among all the layers.

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Table 12.4.4 Importance Factor (Notification Article 15, Appendix Table 4) Type of structure

Structure characteristics

Importance factor

Special class

Among structures in the category of Class A, those having a remarkable risk specified in the item 1, a grave consequence in the item 2, or a serious threat in the item 4, or those having a particular importance in the item 3

1.5

Class A

1 Having high level of risk to cause large loss of human lives and property if the structure is damaged by an earthquake 2 Having serious economic and social consequences if the structure is damaged by an earthquake 3 Having an important role in restoration works after earthquake disaster 4 Having threat of huge loss of human lives and property, if the structure is damaged by an earthquake (ie., structures handling toxic or hazardous substances) 5 Having difficulty in restoration if the structure is damaged by an earthquake

1.2

Class B

Structures other than those classified as Special, Class A or Class C

1.0

Class C

Slight economic and social consequences if the structure is damaged by an earthquake, or easy restoration among structures other than those classified as Special or Class A

0.8

(2) When the examination using the vertical seismic coefficient is required in the seismic coefficient method, the vertical seismic coefficient shall be determined appropriately by considering the structural characteristics, subsoil properties, etc. (3) The design seismic coefficient to high seismic resistant structures in the seismic coefficient method shall be determined after comprehensive judgment of the result calculated by the equation in (1) and the horizontal seismic coefficient calculated by the equations listed below. In the former calculation, the importance factor shall be set to 1.5. For the latter calculation, the peak ground acceleration shall be estimated by the dynamic analysis against the Level 2 ground motion of the potential earthquake stipulated in the “Regional Disaster Prevention Plan” specified by the Article 2, Item 10 of the “Fundamental Law for Countermeasures against Natural Disaster” (Law No.223, 1961). kh = a /g (b) When a is larger than 200 Gal 1 k h = --- ( a ¤ g ) 1 / 3 3

644474448

(a) When a is 200 Gal or less

(12.4.1)

where kh： horizontal seismic coefficient a： peak ground acceleration at the surface (Gal) g： gravitational acceleration (Gal) [Commentary] (1) General Cases of Port and Harbor Facilities (a) Factors to determine the design seismic coefficients of structures include the regional seismicity, the subsoil conditions, the dynamic properties of the structures, and the importance of the structures. However, the dynamic response of the structure needs not be taken into consideration in design, because the majority of port and harbors structures have comparatively short natural periods as well as large damping factors. In general, the dynamic response of structures is ignored and seismic design is conducted based on the procedure given in 12.3 Seismic Coefficient Method, using the design seismic coefficient stipulated above. (b) In cases where it is possible to accurately determine the seismic coefficient by investigating the factors such as the regional seismicity, ground motion characteristics, and the amplification of ground motion due to subsoil response, the value may be used as a design seismic coefficient. For example, when a dynamic response -186-


analysis of the ground is conducted with an input earthquake ground motion based on earthquake data or strong-motion records at the construction site, or when conducting a seismic response analysis of structures in order to consider their dynamic response to earthquake ground motion, it is possible to determine the design seismic coefficient based on the results of those analyses. (c) The importance factor of structures is not a value that could be applied uniformly depending upon the use, type or size of the facilities, but must be determined by giving due consideration to the social and economic characteristics of the facilities along with consideration of the following items: ① Earthquake resistance required for the respective function of facilites ② Magnitude of damage when the facility is struck by an earthquake, difficulty in restoration of the damaged facility, and residual structural strength of the facility after damage. ③ Cargo handling capacity of the facility after damage. (d) When calculating the seismic load in 12.3 Seismic Coefficient Method, the design seismic coefficient must be multiplied by the net deadweight without deducting buoyancy. In the calculation of earth pressure, however, the apparent seismic coefficient should be used, because the submerged unit weight that takes buoyancy into account is used in this case. (2) High Seismic Resistant Structures (a) The target earthquake for the seismic design of high seismic resistant structures should be the potential earthquake preseribed in the “Regional Disester Prevention Plan”, and the input earthquake ground motion at the construction site should be determined based on this potential earthquake. (b) The peak ground acceleration to calculate the design seismic coefficient of high seismic resistant structures can be determined by the multiple reflection model given in 12.5 Seismic Response Analysis [Technical Notes] (2). (c) Noda et al.4) have given a detailed explanation of equation (12.4.1) for calculating the design seismic coefficient of high seismic resistant structures from earthquake motion. (d) Based upon the experience with the serious damage suffered by the Port of Kobe during the Hyogoken-Nanbu Earthquake, the design seismic coefficient should be at least 0.25 when high seismic resistant structures be constructed in locations near an active fault plane (when the site is supposed to be in an epicentral area). (e) The seismic response analysis as well as the seismic coefficient method should be used as the design methods for the Level 2 earthquake motion, based on a comprehensive evaluation of structural type, ground motion and subsoil properties. In those cases, the facilities must secure a certain level of safety factor that is required to ensure the soundness of the facility against the Level 1 ground motion. [Technical Notes] (1) “Level 1” Earthquake Motion for All Port and Harbor Facilities (a) The regional seismic coefficients listed in Table 12.4.1 were set forth using the distribution of expected peak acceleration 8) corresponding to a 75-year return period for coastal regions. The expression “75-year return period” is based on the theory of probability. It does not mean that, should an earthquake motion equivalent to the 75-year return period have occurred, another earthquake motion of the same magnitude will not occur during the following 75 years. For example, the probability that a structure with a 50-year lifetime will encounter an earthquake motion with a 75-year return period or greater during the lifetime is quite high, with a value near 0.5. In the case of a facility whose lifetime can be set shorter than 50 years, the return period of the design earthquake becomes shorter than 75 years if the encounter probability is made equally to around 0.5. Information regarding the relationship between the lifetime and the encounter probability is given in Chapter 1 General. (b) The expected peak acceleration of bedrock with a 75-year return period is listed in Table T- 12.4.1 by regional classification. The regional seismic coefficients in Table 12.4.1 are stipulated from the expected peak acceleration using the average relationship in Fig. T- 12.4.1. “Bedrock” as used here means soil type of Class Ⅰ . (2) “Level 2” Earthquake Motion for High Seismic Resistant Facilities (a) When a potential earthquake has not been stipulated in “Regional Disaster Prevention Plan”, or when it is appropriate to determine specially an earthquake to be used in the design, the earthquake that will create the most hazardous ground motion at the construction site among the potential earthquakes should be selected as the target earthquake. The earthquakes to be considered are such as the largest past earthquake, the earthquake by an “Active Fault with High Probability Level 1” or the earthquake by an “Active Fault under Special Attention”. When researching such earthquakes the “Japan’s Active Fault Distribution Maps and Materials (New Edition)”, “Complete Listing of Damaging Earthquakes in Japan (New Edition)” and the “Japan Earthquake Fault Parameter Handbook” are available as references. The magnitude of an earthquake on an active fault can be estimated from the following equation: -187-


log 10 L = 0.6M 2.9 where L： length of the earthquake fault (km) M： magnitude of earthquake

(12.4.2)

Table T- 12.4.1 Regional Seismic Coefficient and Peak Bedrock Acceleration with 75-Year Return Period by Regional Classification Regional classification

Regional seismic coefficient

Peak bedrock acceleration with 75-year return period (Gal)

A

0.15

350

B

0.13

250

C

0.12

200

D

0.11

150

E

0.08

100

1 3

( Asmac g )

Seismic coefficient

kh

kh = 31

( Asmac g )

Kh =

Upper and lower limits of the seismic coefficient estimated for each quaywall Upper and lower limits of the seismic coefficient estimated for each port (Lines connecting the upper and lower limits determine the range of the seismic coefficient)

SMAC peak acceleration (Gal)

M = 6.0 M = 6.5 M = 7.0 M = 7.5 M = 8.0

SMAC peak acceleration

ASMAC (Gal)

Fig. T- 12.4.1 Relationship between Seismic Coefficient and Peak Ground Acceleration

O data of M 6.0

8.0

Fault distance X (km) Fig. T- 12.4.2 Relationship between SMAC Peak Accelerations and Fault Distance by Magnitude )

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(b) The bedrock acceleration to be used for the analysis of ground seismic response to determine the design seismic coefficient of high seismic resistant facilities is calculated using the following equation: 0.53M

(12.4.3) log 10 A SMAC = 0.53M log 10 ( X + 0.0062 ´ 10 ) 0.00169X + 0.524 where ASMAC： peak bedrock acceleration measured by a SMAC-type strong motion seismograph (Gal) M： magnitude of earthquake X： fault distance (km) The term “fault distance” refers to the minimum distance from the fault plane to the site of interest. The results of calculation by equation (12.4.3) are shown in Fig. T- 12.4.2. When the fault plane is not known, the minimum distance from the surface fault should be used as the fault distance. The term “bedrock” means rock mass, sandy soil layer with the N-value of 50 or greater, clay layer with the unconfined compressive strength qu of 650 kN/m2 or greater, or subsoil with a shear wave velocity of 300m/s or higher. (c) Potential earthquakes are divided into two classes called intra-plate earthquakes and inter-plate earthquakes, depending upon the mechanism of their occurrence. When an intra-plate earthquake is assumed, the Port Island bedrock incident waves computed from the records at the Port Island of Kobe Port during the Hyogoken-Nanbu Earthquake should be used as the input earthquake waves for the ground seismic response analysis. When assuming an inter-plate earthquake, waves such as the Hachinohe bedrock incident waves (S252NS Base) computed from records at the Port of Hachinohe during the 1968 Tokachi-Oki Earthquake or the Ofunato bedrock incident waves (S-1210 E41S) at the Port of Ofunato during the 1978 Miyagi-Ken-Oki Earthquake should be used for the input earthquake waves. When a high seismic resistant facility is to be constructed within a hypocenter area and the earthquake is supposed to be large one occurring in that area, however, the Port Island bedrock incident waves (PI-79NS Base) should be used as the input ground motion, even if the one being considered is an inter-plate earthquake. Figure T- 12.4.3 shows these three wave profiles. The Fourier spectra of these wawes are shown in Fig. T- 12.4.4. (d) The judgment whether the construction site is near the earthquake fault or not should be made according to Fig. T- 12.4.5. If the construction site is in the zone A on the chart, then the site should be considered as being near to the fault plane (located within the hypocenter area).

Acceleration

Hachinohe bedrock incident waves

Time

Acceleration

Ofunato bedrock incident waves

Time

Acceleration

Port Island bedrock incident waves

Time

Fig. T- 12.4.3 Incident Waves for Bedrock

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Fourier spectrum (Gal s)

(a) S-252 NS Base

Distance to fault plane x (km)

Parzen window Band = 0.8 Hz

Frequency (Hz)


(b) S-1210 E 41 S

Magnitude M


Fig. T- 12.4.5 Classification of Earthquakes by Active Faults Note: Sites falling in zone A are regarded as being within hypocenter areas

Frequency (Hz)


(c) Pl-79 NS Base


Frequency (Hz)

Fig. T- 12.4.4 Fourier Spectra of Incident Waves for Bedrock

12.5 Seismic Response Analysis The seismic resistance of port and harbor facilities that are particularly important or that are of structural types having few past precedents shall be examined based on a seismic response analysis in addition to the methods in 12.3 Seismic Coefficient Method or 12.6 Seismic Deformation Method. [Commentary] (1) General Many types of structures that are different from or are much larger than those of structures built in the past have been designed and built as port and harbor facilities in recent years. There are also many cases that structures must be built in locations where ground conditions are extremely poor. Moreover, it is now necessary to study structural deformation to examine the function of high seismic resistant facilities even when subjected to the ground motion near the fault plane, such as the Hyogoken-Nanbu Earthquake that heavily damaged the Port of Kobe. When the type of structure was rare in the past or when the importance for the target structure is extremely high, seismic response analysis should be carried out to examine the deformation of structure by the Level 2 earthquake motion, in which the behavior of structure during an earthquake can be fully studied. (2) Seismic Response Analysis To conduct a seismic response analysis, first a suitable analytical method is selected and the structure is modeled according the analytical procedure, specifying the materials constants for the model. Then the input earthquake ground motion (peak amplitude, wave profiles) is selected and computation of the earthquake response is carried out. The validity of the calculated results must be evaluated based upon an adequate examination of the applicable range and limiting conditions of the analytical procedure, the details of the modeling, and reliability of the material constants.

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[Technical Notes] (1) Input Earthquake Ground Motion (a) The ground motion used for seismic design should basically be determined from the results of past observations or from computations of ground seismic response. The records of strong earthquake motions of the port and harbor regions in Japan are continuously collected by a nationwide observation network. The important records have been published annually since 1963 5). (b) The magnitude of the earthquake, the fault mechanism, the distance from the fault plane, and the seismic response characteristics of ground should be considered in order to select the input ground motion. (c) When the ground conditions of the design site differ from those at which the strong-motion records to be used for the seismic response analysis were obtained 9), the observed records of the surface ground motion are first converted to the bedrock incident wave profiles. Then the bedrock incident waves are inputted to the bedrock of the design site and the surface ground motion can be calculated. The ground seismic response analysis method based on the multiple reflection theory can be used to calculate the surface ground motion from the incident waves of the bedrock and vise versa. In many cases, however, the multiple reflection theory is based on the equivalent-linear method, and the application of this method is limited to a strain level of 1% or less. When used for seismic response analysis for strong ground motion such as the Level 2 earthquake motion, therefore, careful attention should be paid to the limitations of this method. (d) The items indicated in (b) above must be taken into consideration when specifying the peak amplitude of design ground motion. Generally the peak amplitude of ground motion is a function of magnitude and distance, and the following equations have been proposed: log 10 A COR = 0.55M log 10 ( X + 0.0050 ´ 10

0.55M

log 10 A SMAC = 0.53M log 10 ( X + 0.0062 ´ 10 log 10 V = 0.48M log 10 ( X + 0.014 ´ 10

0.43M

log 10 D = 0.62M log 10 ( X + 0.018 ´ 10 where ACOR： ASMAC： V： D： M： X：

) 0.00122X + 0.502

0.53M

) 0.00169X + 0.524

) 0.00060X 0.324

0.43M

) 0.00067X 1.886

(12.5.1) (12.5.2) (12.5.3) (12.5.4)

corrected peak bedrock acceleration (Gal) peak bedrock acceleration for SMSC-type strong motion seismograph (Gal) peak bedrock velocity (cm/s) peak bedrock displacement (cm) magnitude of earthquake distance from fault plane (km) Table T- 12.5.1 Numerical Methods of Seismic Response Analyses

Analysis method (for saturated ground)

Effective stress analysis (solid and liquid phases), Total stress analysis (solid phase)

Dimensions

One-dimensional, Two-dimensional, Three-dimensional

Models for analysis

Multiple reflection model, Spring-mass model, Finite element model

Materials characteristics

Linear, Equivalent-linear, Nonlinear

Domains for computation

Time-domain analysis, Frequency-domain analysis

(2) Seismic Response Analysis Methods Seismic response analysis methods are broadly classified into two types: procedures based on numerical calculations using computers, and vibration experiments using apparatus such as a shaking table. (a) Seismic response analysis based on numerical calculations Seismic response analyses based on numerical calculations can be classified as listed in Table T- 12.5.1. Those methods are briefly explained below. ① Effective stress analysis and total stress analysis Seismic response analyses can be divided into the methods based on the effective stress and those based on the total stress. When the ground liquefies, the excess pore water pressure is induced and the effective stress declines. As a result, the restoring force or damping property of soils will change and the response properties of ground will also change. With the effective stress analysis method, these situations can be analyzed and the excess pore water pressure that occurs in the ground can be determined directly from calculations. On the other hand, the total stress analysis is unable to calculate the excess pore water pressure from the calculation process. So the change in seismic response due to the change in the effective stress cannot be taken into consideration. For situations where the excessive pore water pressure above a certain -191-


level will occur (where the excessive pore water pressure ratio is generally 0.5 or greater), therefore, it is much likely that the calculation results based on the total stress method will differ greatly from the actual seismic response. Nevertheless, the total stress analysis is simple and is often used for design purposes. It is well understood that in general the response values calculated by the effective stress analysis (shear stress or acceleration) are smaller than those by total stress analysis. Therefore, it is considered that total stress analysis will give safer results in design. ② Models for analysis (i) Multiple reflection model This calculation model regards the ground as a stack of horizontal soil layers. When shear waves propagate perpendicularly from the bedrock, it is repeatedly transmitted and reflected at the boundaries between soil layers. With this method the relationship between the soil stress and strain is treated as being linear. Multiple reflection models using equivalent linearization that makes it possible to deal with quasi-nonlinearity have been widely used in recent years. The computer program “SHAKE” 13) based on this model is often used. (ii) Spring-mass model In this model the ground is replaced with a combination of multiple masses, springs, and dampers. With this method the calculation procedure is comparatively simple and it is possible to introduce a nonlinear relationship between displacement and restoring force. (iii) Finite element models These models are not restricted to just ground motion analysis but are used widely in many fields. The ground is divided into a number of finite elements. The finite element models are characterized by their capability to analyze and display with ease the two-dimensional changes of layer thickness and ground conditions. Examples of practical application programs include “FLUSH” 14), “BEAD” 15), and “FLIP” 16), 17). Based on the experience of analyses of damaged quaywalls at the Port of Kobe during the Hyogoken-Nanbu Earthquake, “FLIP” is being used for verification of the deformation level of high seismic resistant facilities and other purposes. In addition, “FLAC” is used as a finite difference analysis program using the explicit solution method. (iv) Continuum model This model regards a structure (or ground) as a cantilever beam possessing uniform characteristics or properties that vary in a uniform manner. The shear beam model is normally selected for modeling. Constants required for calculations are the dimensions of structure, the unit weight of each part, the shear modulus and its depth-wise rate of variation, and the damping constant. (b) Vibration experiments using a shaking table This method takes the dynamic similitude into consideration and applies vibrations to a model of structure. It is an effective means for understanding the overall behavior of a structure and the ground. In this regard, a fairly high level of experimental technique is required in making a model that adequately reproduces the dynamic characteristics of the prototype. The methods by vibration experiments with apparatus such as a shaking table include the following methods. ① Model vibration experiments using a shaking table For these experiments, a model is prepared that will satisfy the similitudes of the geometrics and dynamic properties of the target structure and the ground. The assumed earthquake motion is then applied to the model with a shaking table. ② Model vibration experiments using a centrifuge For these experiments, a centrifuge is used to reproduce stress conditions in the model that are identical to actual stress. The assumed earthquake motion is then applied with a vibration test device loaded on the centrifuge to satisfy the similitude.

12.6 Seismic Deformation Method The displacements and stresses of tubular structures buried in the ground during an earthquake shall be investigated using the seismic deformation method, because the displacements of these types of structures are controlled by the deformation of the surrounding earth. [Commentary] For an investigation of the seismic resistance of tubular structures buried in the ground such as immersed tunnels or oil pipelines that are very long compared to their cross-sectional area, it is very important to investigate the relative displacement of subsoil around the structure controlled by the ground deformation during an earthquake. The relative displacement that will occur in the subsoil around tubular structures will vary depending on the nature of earthquake as well as other factors such as the ground conditions. -192-


[Technical Notes] The “Immersed Tunnel Technology Manual” is refered to design immersed tunnels. [References] 1) Hironao TAKAHASHI, Takashi NAKAMOTO, Hisanori YOSHIMURA: “Analysis of maritime transportation in KOBE Port after the 1995 HYOGOKEN-NANBU Earthquake”, Tech. Note of PHRI, No. 861, 1997 (in Japanese). 2) Tatsuo UWABE, Setsuo NODA, Eiichi KURATA: “Characteristics of vertical components of strong-motion accelerograms and effects of vertical ground motion on stability of gravity type quaywalls”, Rept. of PHRI, Vol. 15, No. 2, 1976 (in Japanese). 3) Atsushi NOZU, Tatsuo UWABE, Yukihiro SATO, Takumi SHINOZAWA: “Relation between seismic coefficient and peak ground acceleration estimated from attenuation relations”, Tech. Note of PHRI, No. 893, 1997 (in Japanese). 4) Setuo NODA, Tatsuo UWABE, Tadaki CHIBA: “Relation between seismic coefficient and ground acceleration for gravity quaywall”, Rept. of PHRI, Vol. 14, No. 4, 1975 (in Japanese). 5) Masafumi MIYATA, Yukihiro SATO, Koji ICHII, Toshikazu MORITA, Susumu IAI: “Annual report on strong-motion earthquake records in Japanese ports (1994)”, Tech. Note of PHRI, No. 840, 1996 (in Japanese). 6) Hajime TSUCHIDA, Eiichi KURATA, Tokuzo ISHIZAKA, Satoshi HAYASHI: “Average response spectra for various site conditions”, Rept. of PHRI, Vol. 12, No. 4, 1973 (in Japanese). 7) Shoichi KITAJIMA, Tatsuo UWABE: “Analysis on seismic damage in anchored sheet-piling bulkheads”, Rept of PHRI, Vol. 18, No. 1, 1979 (in Japanese). 8) Sosuke KITAZAWA, Tatsuo UWABE, Norihiro HIGAKI: “Expected values of maximum base rock accelerations along coasts of Japan”, Tech. Note of PHRI, No. 486, 1984 (in Japanese). 9) Eiichi KURATA, Tokuzo ISHIZAKA, Hajime TSUCHIDA: “Site characteristics of strong-motion earthquake stations in ports and harbours in Japan (Part Ⅲ )”, Tech. Note of PHRI, No. 156, 1973 (in Japanese). 10) Hajime TSUCHIDA, Susumu IAI, Eiichi KURATA: “Analysis of earthquake ground motions observed with two dimensional seismometer array (First Report) - North Tokyu Bay Earthquake of June 4, 1977 -”, Rept. of PHRI, Vol. 22, No. 2, 1983 (in Japanese). 11) Susumu IAI, Eiichi KURATA, Hajime TSUCHIDA: “Digitization and corrections of strong-motion accelerograms”, Tech. Note of PHRI, No. 286, 1978 (in Japanese). 12) Tatsuo UWABE: “Base rock motion around the pacific coasts in Tohoku district - Design earthquake ground motion of Kamaishi breakwater”, Tech. Note of PHRI, No. 412, 1982 (in Japanese). 13) Schnabel, P. B., Lysmer, J., and Seed, H. B.: “SHAKE - A computer program for earthquake response analysis of horizontally layered site”, Report No. EERC72-12, College of Engineering, University of California, Berkeley, 1972. 14) Lysmer, J., Udaka, T, Tsai, C. F., and Seed, H. B.: “FLUSH - A computer program of approximate 3-D analysis of soil structure interaction problems”, Report No. EERC 75-30, University of California at Berkeley, 1975. 15) Tatsuo UWABE, Setsuo NODA, Tadaki CHIBA, Norihiro HAGAKI: “Coupled hydrodynamic response characteristics and water pressures of large composite breakwaters”, Rept of PHRI, Vol. 20, No. 4, 1981 (in Japanese). 16) Susumu IAI, Yasuo MATSUNAGA, Tomohiro KAMEOKA: “Parameter identification for a cyclic mobility model”, Rept of PHRI, Vol. 29, No. 4, 1990 (in Japanese). 17) Susumu IAI, Yasuo MATSUNAGA, Tomohiro KAMEOKA: “Strain space plasticity model for cyclic mobility”, Rept of PHRI, Vol. 29, No. 4, 1990 (in Japanese). 18) Takamasa INATOMI, Ikuki YOKOHAMA, Motoki KANAZAWA: “One-dimensional nonlinear dynamic ground response analyses”, Tech. Note of PHRI, No. 561, 1986 (in Japanese). 19) ISHIHARA, K., YOSHIDA, N. and TSUJINO, S.: “Modelling of stress - strain relations of soils in cyclic loading”, Proc. of 5th Conf. on Numerical Methods in Geomechanics, Nagoya, Vol. 1, 1985, pp. 373-380. 20) Finn, W. D. L., Martin, G. R., and Lee, M. K. W.: “Comparison of dynamic analyses for saturated sands”, Proc. ASCE, Earthquake Engineering and Soil Dynamics, Vol. 1, 1978, pp. 472-491. 21) HOUSNER, G. W.: “Behavior of structures during an earthquake”, Proc. of ASCE, Vol. 85, No. EM4, 1959. 22) Susumu IAI, Tomohiro KAMEOKA: “Analysis of deformation in sheet pile quaywall due to liquefaction”, Rept of PHRI, Vol. 30, No. 2, 1991 (in Japanese). 23) Susumu IAI, Koji ICHII, Toshikazu MORITA: “ Effective stress analysis on a caisson type quaywall - mechanism of damage to port facilities during 1995 Hyogoken-Nanbu Earthquake (Part Ⅶ)”, Tech. Note of PHRI, No. 813, pp. 253-280 (in Japanese) 24) Noda, S., Tsuchid, H. and Kurata, E.: “Dynamic tests of soil embankments”, 5th WCEE, 1973. 25) Osamu KIYOMIYA, Shigeo NAKAYAMA, Hajime TSUCHIDA: “Observations of dynamic response of Kinuura submerged tunnel during an earthquakes and dynamic response analysis”, Tech. Note of PHRI, No. 221, 1975 (in Japanese). 26) Takamasa INATOMI, Satoshi HAYASHI, Ikuhiko YAMASHITA: “Vibration characteristics of the open type steel piled wharf with container crane”, Rept. of PHRI, Vol. 12, No. 2, 1973 (in Japanese). 27) Susumi IAI, Hajime TSUCHIDA: “Earthquake response analysis of floating type structures”, Tech. Note of PHRI, No. 337, 1980 (in Japanese). 28) Tatsuo UWABE, Hajime TSUCHIDA, Eiichi KURATA: “Coupled hydrodynamic response analysis based on strong motion earthquake records of fill type breakwater in deep sea”, Rept. of PHRI, Vol. 22, No. 2, 1983 (in Japanese). 29) Motoki KAZAMA, Takamasa INATOMI, Toshihiro IMAMURA: “Observation and analysis of seismic response grid type improved ground by deep mixing method”, Rept of PHRI, Vol. 22, No. 4, 1983 (in Japanese). 30) Hiromasa FUKUUCHI, Nobuo MIYAJIMA, Ikuo YAMASHITA: “Studies on the vibration characteristics of fill-type embankments”, Rept of PHRI, Vol. 5, No. 3, 1966 (in Japanese). 31) Hideo ARAI, Yasuhumi UMEHARA: “Vibration of dry sand layers”, Rept. of PHRI, Vol. 6, No. 5, 1969 (in Japanese). 32) Hideo ARAI, Yasuhumi UMEHARA: “Vibration of saturated sand layers”, Rept. of PHRI, Vol. 8, No. 1, 1969 (in Japanese).

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33) Hideo ARAI, Toshiyuki YOKOI: “On the earthquake resistance of anchored sheet-pile walls (1st Report) -　Model vibration tests of anchored sheet-pile walls in dry sand”, Rept of PHRI, Vol. 4, No. 9, 1965 (in Japanese) 34) Tatsuo UWABE, Sosuke KITAZAWA, Norihiro HIGAKI: “Shaking table tests and circular arc analysis for large models of embankment on saturated sand layers”, Rept of PHRI, Vol. 24, No. 2, 1985 (in Japanese) 35) Takuji NAKANO, Kenji MORI: “Earthquake-resistant calculation and dynamic model test on trench type tunnel”, Tech. Note of PHRI, No. 172, 1973 (in Japanese) 36) Takamasa INATOMI, Motoki KAZAMA, Toshihiro IMAMURA: “An experimental study on the earthquake resistance of wall type improved ground by deep mixing method”, Rept. of PHRI, Vol. 22, No. 3, 1983 (in Japanese). 37) Setsuo NODA, Sosuke KITAZAWA, Takeshi IIDA, Nobuo MORI, Hiroshi TABUCHI: “An experimental study on the earthquake of steel plate cellular-bulkheads with embedment”, Rept of PHRI, Vol. 21, No. 3, 1982 (in Japanese). 38) Takashi SUGANO, Masaaki MITOH, Ken OIKAWA: “Mechanism of damage to port facilities during 1995 Hyogoken-Nanbu Earthquake (Part Ⅵ ) - experimental study on the behavior of caisson type quay wall during an earthquake using underwater shaking table”, Tech. Note of PHRI, No. 813, 1995 (in Japanese). 39) Osamu KIYOMIYA, Hideo NISHIZAWA, Hiroshi YOKOTA: “Field observation and response analysis at Kawasaki Port submerged tunnel”, Rept of PHRI, Vol. 22, No. 3, 1983 (in Japanese). 40) Setsuo NODA, Tatsuo UWABE: “Microtremor measurement on sea banks”, Tech. Note of PHRI, No. 205, 1975 (in Japanese). 41) Takamasa INATOMI, Motoki KAZAMA, Toshihiro IMAMURA: “An experimental study on the earthquake resistance of wall type improved ground by deep mixing method”, Rept. of PHRI, Vol. 22, No. 3, 1983 (in Japanese). 42) Newmark, N. M.: “Effects of earthquakes on dams and embankments”, Geotechnique, Vol. 15, No. 2, 1965. 43) Hajime TSUCHIDA, Ken-ichiro MINAMI, Osamu KIYOMIYA, Eiichi KURATA, Hideo NISHIZAWA: “Stress of buried pipe during an earthquake based on two dimensional seismometer array observation”, Rept. of PHRI, Vol. 20, No. 4, 1981 (in Japanese). 44) Tatsuo UWABE, Norihiro HIGAKI: “An experimental study on sliding block in water during an earthquake”, Rept. of PHRI, Vol. 23, No. 3, 1984 (in Japanese).

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Chapter 13 Liquefaction 13.1 General Saturated loose sandy subsoils tend to liquefy during an earthquake, causing damage to structures. When designing structures, the effects of liquefaction shall be taken into consideration to the extent as necessary. [Technical Notes] The “Handbook on Liquefaction Remediation of Reclaimed Land” (Revised Edition) can be referred to examine liquefaction of the subsoils.

13.2 Prediction of Liquefaction (Notification Article 17) In principle, predictions of subsoil liquefaction shall be carried out according to the following: (1) When the subsoil is composed of soils such as saturated loose sandy soil, a prediction and judgment of liquefaction occurrence of the subsoil shall be made. (2) Predictions and judgment of liquefaction occurrence of subsoil shall be carried out by choosing an appropriate method that uses the gradation and Standard Penetration Test results or the cyclic triaxial test results. [Commentary] (1) Types of Liquefaction Prediction and Judgment There are two types of the methods for prediction of liquefaction occurrence. One method is based on the gradation and N-values, and another uses the results of a cyclic triaxial test. The former method based on the gradation and N-values is the simplest method and can be applied in general. The latter method using cyclic triaxial test results is a detailed procedure and is used in the situations where prediction according to the method based on the gradation and N-values is difficult. (2) Depth of Investigation The subsoils subject to liquefaction prediction are those down to a depth of 20 m below the ground surface (or below the seabed for structures in the sea). However, when it is predicted that structures will suffer serious damage if liquefaction occur in a layer at a depth of more than 20 m, or when the soil layer continues below the depth of 20m, liquefaction predictions should include these layers as well. [Technical Notes] (1) Prediction of Liquefaction Using Gradation and N-values 1) (a) Judgment based on gradation The subsoils should be classified according to gradation, by referring to Fig. T- 13.2.1, which is divided into two sub-figures according to the value of the uniformity coefficient. The threshold value of the uniformity coefficient (Uc= D60 /D10) is 3.5, where Uc is the uniformity coefficient, and D60 and D10 denote the grain sizes corresponding to 60% and 10% passing, respectively. Soil is judged not to liquefy when the grain size distribution curve is not included in the range “possibility of liquefaction”. When the grain size distribution curve spans the “possibility of liquefaction” range, a suitable approach is required to examine the possibility of liquefaction. For the soil with grain size distribution of large fine portion a cyclic triaxial test should be carried out. For that of large gravel portion, the soil is determined not to liquefy when the coefficient of permeability is 3 cm/s or greater. When there are subsoils with poor permeability such as clay or silt on top of the target subsoil in this case, however, it should be treated as soil that falls within the range of “possibility of liquefaction”.

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For soil with large uniformity coefficient U

Percentage passing by mass

Very large possibility of liquefaction Possibility of liquefaction

Clay

Silt

Grain size Sand

Gravel

Fig. T- 13.2.1(a) Range of Possible Liquefaction (Uc ≧ 3.5)

For soil with small uniformity ecefficient U

Percentage passing by mass

Very large possibility of liquefaction Possibility of liquefaction

Clay

Silt

Grain size Sand

Gravel

Fig. T- 13.2.1(b) Range of Possible Liquefaction (Uc < 3.5)

(b) Prediction of liquefaction using equivalent N-values and equivalent acceleration For the subsoil with a gradation that falls within the range “possibility of liquefaction” shown in Fig. T13.2.1, further investigations should be carried by the descriptions below. ① Equivalent N-value The equivalent N-value should be calculated from equation (13.2.1). N 0.019 (s v ¢ 65 ) ( N ) 65 = -----------------------------------------------------0.0041 (s v ¢ 65 ) + 1.0

(13.2.1)

where (N)65： equivalent N-value N： N-value of the subsoil s v ¢： effective overburden pressure of the subsoil (kN/m2) (the effective overburden pressure used here should be calculated with respect to the ground elevation at the time of the Standard Penetration Test.) The equivalent N-value refers to the N-value corrected for the effective overburden pressure of 65 kN/m2. This conversion reflects the practice that liquefaction prediction was previously made on the basis of the N-value of a soil layer near a groundwater surface (see the “Technical Standards and Commentaries for Port and Harbour Facilities in Japan” (1979)). In cases where equation (13.2.3) in (c) below is used, however, the equivalent N-value should be the same as the N-value of the layer without correction. ② Equivalent acceleration The equivalent acceleration should be calculated using equation (13.2.2). The maximum shear stress determined from the seismic response analysis is used to determine the equivalent acceleration for each soil layer. The input motion to the bedrock in the seismic response analysis is determined by consulting with the reference 5). t max (13.2.2) a eq = 0.7 ´ ---------- g sv ¢ -196-


where a eq： equivalent acceleration (Gal) t max： maximum shear stress (kN/m2) s v ¢： effective overburden pressure (kN/m2) (the effective overburden pressure should be determined based on the ground elevation during an earthquake.) g： gravitational acceleration (980 Gal)

Equivalent

N-value

③ Predictions using the equivalent N-value and equivalent acceleration The soil layer should be classified according to the ranges labeled I ~ IV in Fig. T- 13.2.2, using the equivalent N-value and the equivalent acceleration of the soil layer. The meaning of the ranges I ~ IV is explained in Table T-13.2.1 to appear later.

Equivalent acceleration (Gal)

Fig. T- 13.2.2 Classification of Soil Layer with Equivalent N-Value and Equivalent Acceleration

(c) Correction of N-values and predictions when the fraction of fines content is relatively large. When the fines content (grain size is 75 mm or less) is 5% or greater, the equivalent N-value should be corrected before applying Fig. T- 13.2.2. Corrections of the equivalent N-value are divided into the following three cases.

N

Compensation factor for equivalent -value

cN

① Case 1: when the plasticity index is less than 10 or cannot be determined, or when the fines content is less than 15% The equivalent N-value (after correction) should be set as (N)65/cN. The compensation factor cN is given in Fig. T- 13.2.3. The equivalent N-value (after correction) and the equivalent acceleration are used to determine the range in Fig. T- 13.2.2.

Fines content

FC (%)

Fig. T- 13.2.3 Compensation Factor of Equivalent N-Value Corresponding to Fine Contents

② Case 2: when the plasticity index is greater than 10 but less than 20, and the fines content is 15% or higher The equivalent N-value (after correction) should be set as both (N)65/0.5 and N + DN, and the range should be determined according to the following situations, where the value for DN is given by the following equation: (13.2.3) DN = 8 + 0.4 ´ ( I p 10 ) i ) When N + DN falls within the range I, use range I. ii ) When N + DN falls within the range II, use range II. iii) When N + DN falls within the range III or IV and (N)65/0.5 is within range I, II or III, use range III. -197-


iv) When N + DN falls within range III or IV and (N)65/0.5 is within range IV, use range IV. Here, the range III is used for the case iii) even when the equivalent N-value (after correction) with (N)65/0.5 is in the range I or II, because the results from the fines content correction are too conservative. The reason that the range IV is not used for the case iii) even when range IV is given by a correction N + DN is that the reliability of the plasticity index in the equation is low when the value is 10 ~ 20. Therefore, judging the subsoil as the range IV “possibility of liquefaction is very low” is considered as risky. ③ Case 3: when the plasticity index is 20 or greater, and the fines content is 15% or higher The equivalent N-value (after correction) should be set as N + DN. The range should be determined according to the equivalent N-value (after correction) and the equivalent acceleration. Figure T-13.2.4 illustrates the ranges of applications of the cases ① to ③ .

No corrections

Correction with

cN in Fig. T- 13.2.3 N in Equation (13.2.3)

Plasticity index

Ip

Correction with ∆

Correction with

cN in Fig. T- 13.2.3 and ∆N

by equation (13.2.3)

Fine contents

Fc (%)

Fig. T- 13.2.4 N-Value Correction Methods by Fine Contents and Plasticity Index

(d) Liquefaction predictions Since liquefaction predictions must also consider the factors other than physical phenomena such as what degree of safety should be maintained in the structures, it is not possible to unconditionally establish any criteron for judgments regarding various prediction results. The rule of judgment of liquefaction occurrence for the results of prediction that is considered as standard are listed in Table T- 13.2.1. In this table the term “prediction of liquefaction” refers to the high or low possibility of liquefaction as a physical phenomenon. In contrast, the term “judgment of liquefaction” refers to the consideration of the high or low possibility of liquefaction and judgment of whether or not the ground will liquefy. Table T- 13.2.1 Predictions and Judgments of Liquefaction for Soil Layer According to Ranges Ⅰ to Ⅳ Range shown in Fig. T- 13.2.3

Prediction of liquefaction

Ⅰ

Possibility of liquefaction occurrence is very high

Liquefaction will occur

Ⅱ

Possibility of liquefaction occurrence is high

Either to judge that liquefaction will occur or to conduct further evaluation based on cyclic triaxial tests.

Ⅲ

Possibility of liquefaction is low

Either to judge that liquefaction will not occur or to conduct further evaluation based on cyclic triaxial tests. For a very important structure, either to judge that liquefaction will occur or to conduct further evaluation based upon cyclic triaxial tests.

Ⅳ

Possibility of liquefaction is very low

Liquefaction will not occur

Judgment of liquefaction

(2) Prediction and Judgment Based on the Results of Cyclic Triaxial Tests In some cases it may be difficult to predict or judge the possibility of subsoil liquefaction from the results of gradation and N-values. In such cases a prediction and judgment for subsoil liquefaction should be made with the results of a seismic response analysis and cyclic triaxial tests conducted on undisturbed samples. -198-


(3) Judgmant of Overall Liquefaction In the judgment of overall subsoil liquefaction for a site consisting of soil layers, the decision should be made based on a judgement for each layer of subsoil.

13.3 Countermeasures against Liquefaction When implementing liquefaction countermeasures, appropriate countermeasure works shall be selected to maintain the functions of structures after an earthquake. [Commentary] (1) When designing liquefaction countermeasure works, appropriate decisions regarding the following items must be made: (a) Construction methods of countermeasure works (b) Dimensions of countermeasure works (area and depth) (c) Specific design of the countermeasure works (2) Items (a) to (c) below are considered to be liquefaction countermeasure works. (a) To prevent the occurrence of excessive pore water pressure ① Compaction ② Solidification ③ Replacement (replacement with easily compacted sand) (b) To dissipate excessive pore water pressure ① Drain placement ② Replacement (replacement with coarse sand, gravel, etc.) (c) Combinations of (a) and (b) ① Simple combinations of measures in (a) and (b) ② Combinations of measures in (a) and (b) after giving consideration to their relationship with surrounding structures (3) The dimensions of soil improvements to be used as liquefaction countermeasures should be determined to the extent to maintain the functions of structures. (4) When using compaction as a liquefaction countermeasure, the subsoil should be compacted so that the N-value after compaction reaches a value at which liquefaction will be judged not to occur (see 13.2 Prediction of Liquefaction). When the subsoil that will liquefy is located adjacent to the improved subsoil, an area under the influence of the liquefaction should be reserved in the improved subsoil area. The equivalent N-value for the reserved area should be 16 or greater. (5) Liquefaction countermeasure works should be carried out by taking due consideration to the design targets of subsoil improvement, and the influence on existing structures and on the surrounding area. [Reference] 1) Hiroyuki YAMAZAKI, Kouki ZEN, Fumikatsu KOIKE: “Study of the liquefaction prediction based on the grain distribution and the SPT N-value”, Tech. Note of PHRI, No. 914, 1998 (in Japanese).

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Chapter 14 Earth Pressure and Water Pressure 14.1 Earth Pressure The earth pressure acting on the structure shall be calculated in correspondence to various conditions such as sandy soil and cohesive soil properties, active and passive states of earth pressure due to the displacement mode of structure, the situation in ordinary condition or earthquake condition. [Commentary] The earth pressure is the force that is exerted by the soil mass and acts on a structure, for instance, quaywall. The magnitude of the earth pressure is known to vary considerably with the displacement of structure and with the nature of the soil such as gradation, void ratio, water content, and stress path. The earth pressure discussed in this chapter is the pressure acting on a structure by ordinary soil. The earth pressure caused by liquefied soil, improved soil, or reinforced soil is not discussed in this chapter. The actual phenomenon of the earth pressure during an earthquake is caused by dynamic interaction between backfill soil, structure and water, which is different from the earth pressure discussed in this chapter. However, many analyses of past damage due to earth pressures during earthquakes have enabled to formulate the practical calculation method of earth pressure during an earthquake for designs. The hydrostatic pressure and dynamic water pressure acting on a structure should be evaluated separately.

14.2 Earth Pressure under Ordinary Conditions 14.2.1 Earth Pressure of Sandy Soil under Ordinary Conditions (Notification Article 11, Clause 1, Number 1) The earth pressure of sandy soil acting on the backface wall of structure and the angle of sliding surface shall be calculated by the following equations: (1) Active Earth Pressure and the Angle of Failure Surface.

p ai = K ai éêSg i hi + w cosy ùú cosy

(14.2.1)

y - b)û

ë

cos(

y + d ) sin(f + d ) cos(y - b ) sin(f - b ) cos(

z - b ) = - tan(f + d + y - b ) + sec(f + d + y - b )

cot(

i

i

i

i

(14.2.2)

i

where

fi - y )

2

cos (

K ai = cos

2

é y cos(d + y ) ê1 + êë

fi + d ) sin(fi - b ) ù ú cos(d + y ) cos(y - b ) ú û

2

sin(

(2) Passive Earth Pressure and the Angle of Failure Surface.

p pi = Kpi éêSg i hi + w cosy ùú cosy

(14.2.3)

y - b)û

ë

cos(

z - b ) = tan(f - d - y + b ) + sec(f - d - y + b )

cot(

i

i

i

y + d ) sin(f - d ) cos(y - b ) sin(f + b ) cos(

i

(14.2.4)

i

where 2

fi + y )

cos (

K pi = cos

2

é y cos(d + y ) ê1 êë

fi - d ) sin(fii + b ) ù ú cos(d + y ) cos(y - b ) ú û

2

sin(

with pai , ppi： active and passive earth pressure, respectively, acting on the bottom level of the i-th soil layer (kN/m2) fi： angle of internal friction of the i-th soil layer (°) gi： unit weight of the i-th soil layer (kN/m3) hi： thickness of the i-th soil layer (m) Kai , Kpi： coefficients of active and passive earth pressures, respectively, in the i-th soil layer y： angle of batter of backface wall from vertical line (°) b： angle of backfill ground surface from horizontal line (°) -200-


d： angle of friction between backfilling material and backface wall (°) zi： angle of failure surface of the i-th soil layer (°)

w： uniformly distributed surcharge (kN/m2)

w

z1

b

z2

y y

y+d d zi

Fig. 14.2.1 Schematic Diagram of Earth Pressure Acting on Retaining Wall

[Commentary] (1) Angle of Internal Friction of Soil The angle of internal friction of backfill soil normally has a value of 30°. In case of especially good backfilling material, it can be set as large as 40°. (2) Angle of Friction between Backfilling Material and Backface Wall The angle of friction between backfilling material and backface wall normally has a value of ±15° ~ 20°. It may be estimated as one-half of the angle of internal friction of backfilling material. (3) Unit Weight of Soil. The unit weight of soil normally has a value of 18 kN/m3 as unsaturated soil such as a soil above the residual water level, and 10 kN/m3 as saturated soil below it.

14.2.2 Earth Pressure of Cohesive Soil under Ordinary Conditions (Notification Article 11, Clause 1, Number 2) The earth pressure of cohesive soil acting on the backface wall of structure and the angle of failure surface shall be calculated by following equations: (1) Active Earth Pressure pa = Sg i hi + w -2c (2) Passive Earth Pressure p p = Sg i hi + w + 2c where pa： active earth pressure acting on the bottom level of the i-th soil layer (kN/m2) pp： passive earth pressure acts on the bottom level of the i-th soil layer (kN/m2) gi： unit weight of the i-th soil layer (kN/m3) hi： thickness of the i-th soil layer (m) w： uniformly distributed surcharge (kN/m2) c： cohesion of soil in the i-th layer (kN/m2)

(14.2.5) (14.2.6)

[Commentary] (1) Active earth pressure is calculated using equation (14.2.5). If a negative earth pressure is obtained by calculation, then the pressure should be set as zero. (2) Cohesion of soil should be determined using an appropriate method (refer to 11.3.3 Shear Properties). (3) In case of cohesive soil, the cohesion between backfill and backface wall should be ignored. (4) The unit weight of cohesive soil should be estimated by soil test. The wet unit weight gt should be used for soils above the residual water level, and the submerged unit weight g ¢ be used for soils below the residual water level. -201-


14.3 Earth Pressure during Earthquake 14.3.1 Earth Pressure of Sandy Soil during Earthquake (Notification Article 18, Clause 1, Number 1) The earth pressure of sandy soil acting on a backface wall of structure during an earthquake and the angle of failure surface shall be calculated by following equations: (1) Active Earth Pressure and the Angle of Failure Surface.

pai = K ai éêSgi hi + ë

y ù ú cosy cos(y - b ) û w cos

(14.3.1)

z - b ) = - tan(f + d + y - b ) + sec(f + d + y - b )

cot(

i

i

i

y + d + q ) sin(f + d ) cos(y - b ) sin(f - b - q ) cos(

i

(14.3.2)

i

where

fi - y - q ) é sin(fi + d ) sin(fi - b - q ) ù y cos(d + y + q ) ê1 + ú cos(d + y + q ) cos(y - b ) ú êë û 2

cos (

K ai = cos

q cos

2

2

(2) Passive Earth Pressure and the Angle of Failure Surface.

ppi = K pi éêSg i hi + ë

y ù ú cosy cos(y - b ) û w cos

(14.3.3)

z - b ) = tan(f - d - y + b ) + sec(f - d - y + b )

cot(

i

i

i

y + d - q ) sin(f - d ) y - b ) sin(f + b - q )

cos(

i

cos(

(14.3.4)

i

where

fi + y - q ) é sin(fi - d ) sin(fi + b - q ) ù y cos(d + y - q ) ê1 ú cos(d + y - q ) cos(y - b ) ú êë û 2

cos (

K pi = cos

q cos

2

2

where q： composite seismic angle (°) given by the following equations: q = tan-1k (above water level) q = tan-1k¢ (below water level) k： seismic coefficient k¢ ： apparent seismic coefficient

The notations other than q, k, and k¢ are the same as those defined in 14.2.1 Earth Pressure of Sandy Soil under Ordinary Conditions. The apparent seismic coefficient k¢ is described in 14.3.3 Apparent Seismic Coefficient. [Commentary] (1) Earth pressure during an earthquake is based on the theories proposed by Mononobe (1917) and Okabe (1924). (2) Angle of friction between backfilling material and backface wall normally has a value of ± 15° ~ 20°. It may be estimated as one-half of the angle of internal friction of backfilling material. (3) The composite seismic angle k is used for soils above the residual water level, and k¢ is used below it. (4) The coefficient of earth pressure and angle of failure surface as illustrated can be obtained from the diagrams in Fig. C- 14.3.1. (5) The earth pressure theory assumes that the soil and the pore water behave as contained in one unit. Thus the equations mentioned above cannot be applied to liquefied soil. It is necessary for liquefied soil to treat carefully with dynamic effective stress analysis or model tests.

-202-


70 °

δ = 15°

1.0

δ = 15°

60 °

0.8

50 °

φ=

φ= 40 °

° 25

φ=

° 30

φ=

° 35

4 φ=

=

30 °

0°

K acos δ

0.4

=

ζa

30 °

φ

φ=

φ

0.6

40 °

35 °

25 °

20 ° 30° φ= 35° φ=

5°

φ=2

0.2

10 °

K asin δ

0° φ=4

0

0 0

0.1

0.2

0.3

0.4

0.5

0

0.1

0.2

0.3

k

10.0

0.4

0.5

k

δ = –15°

30 °

δ = –15° 25 °

8.0

K p cos δ

φ=

40°

φ = 25 ° φ = 30 ° φ = 35°

20 °

6.0

φ=

ζp

35°

φ = 40°

15 °

φ=3 0°

4.0

φ=2 K p sin δ

10 °

5°

φ = 40° φ = 35° φ = 30°

2.0

5°

φ = 25° 0

0 0

0.1

0.2

0.3

0.4

0.5

0

0.1

0.2

0.3

k

0.4

0.5

k

60 °

δ = 15° 50 °

φ= 25° φ=3 0° φ = 35 °

3.0

δ = 15°

40 °

K p cos δ

φ = 40°

2.0

φ = 40°

30 °

φ = 35°

ζp

φ = 30°

20 °

φ = 25°

10 °

1.0 K p sin δ

φ = 40° φ = 35°

0

φ = 30°

φ = 25° –10 °

0 0

0.1

0.2

0.3

0.4

0.5

0

0.1

0.2

0.3 k

k

Fig. C- 14.3.1 Coefficient of Earth Pressure and Failure Angle -203-

0.4

0.5


14.3.2 Earth Pressure of Cohesive Soil during Earthquake (Notification Article 18, Clause 1, Number 2) The earth pressure of cohesive soil acting on a backface wall of structure during an earthquake shall be calculated by the following: (1) Active Earth Pressure Active earth pressure shall be calculated using an appropriate earth pressure equation which takes the seismic coefficient into account so that the structural stability will be secured during an earthquake. (2) Passive Earth Pressure Passive earth pressure shall be calculated using an appropriate earth pressure equation so that the structural stability will be secured during an earthquake. [Commentary] (1) The active earth pressure acting on a backface wall of structure during an earthquake and the angle of sliding surface should be calculated by following equations: pa

=

(Sgi

hi

+ w)

sin (z a

+q )

cos q sin z a

-

c

(14.3.5)

cos z a sin z a

æ Sg h + 2w ö z a = tan -1 1 - ç i i ÷ tan q 2c è ø

where p a： gi： hi： z a： w： c： q： k： k¢：

(14.3.6)

active earth pressure, acting on the bottom level of the i-th soil layer (kN/m2) unit weight of the i-th soil layer (kN/m3) thickness of the i-th soil layer (m) angle of failure surface of the i-th soil layer (°) uniformly distributed surcharge (kN/m2) cohesion of the soil (kN/m2) 1 1 composite seismic angle (°) ( q = tan k or q = tan k¢ ) seismic coefficient apparent seismic coefficient

(2) There are many unknown items concerning the method for determining the passive earth pressure of cohesive soil during an earthquake. From the practical point of view, the passive earth pressure in ordinary condition discribed in 14.2.2 Earth Pressure of Cohesive Soil under Ordinary Conditions can be used as passive earth pressure during an earthquake. (3) The apparent seismic coefficient should be used to calculate the earth pressure of cohesive soil down to the sea bottom during an earthquake. The apparent seismic coefficient may be set as zero when calculating the earth pressure at the depth of 10 m from the sea bottom or deeper. The earth pressure between these two depths is determined assuming that the earth pressure is linearly distributed between them. If the earth pressure at the depth of 10 m below the sea bottom becomes less than the earth pressure at the sea bottom, the latter should be applied down to the depth of 10 m.

14.3.3 Apparent Seismic Coefficient (Notification Article 19) It shall be standard to calculate the earth pressure acting below the residual water level during an earthquake according to the procedures given in 14.3.1 Earth Pressure of Sandy Soil during Earthquake and 14.3.2 Earth Pressure of Cohesive Soil during Earthquake, by using the apparent seismic coefficient that is determined by equation (14.3.7). k

where k¢： gt： g： w： hi： hj： h： k：

¢

=

2 ( Sg t hi 2 [Sg t hi

+ Sg

hj

+w) +g h

+ S (g - 10 ) h j +w ] + ( g - 10 ) h

k

apparent seismic coefficient unit weight of soil layer above the residual water level (kN/m3) unit weight (in the air) of saturated soil layer below the residual water level (kN/m3) uniform external load at the ground surface (kN/m2) thickness of the i-th soil layer above the residual water level (m) thickness of the j-th soil layer below the residual water level (m) thickness of soil layer to calculate earth pressure below the residual watere level (m) seismic coefficient -204-

(14.3.7)


[Commentary] (1) In case of stability analysis of quaywall with use of equation (14.3.7), the dynamic water pressure during an earthquake should be applied to the wall in the seaward direction. (2) The concept of the apparent seismic coefficient k¢ is expressed by the following equation: g t ´ k = ( g 10 ) ´ k¢ (3) A product of unit weight of a soil layer (in the air) and seismic coefficient becomes equal to the product of submerged unit weight of a soil layer and the apparent coefficient for the soil below the water level.

First stratum

First i stratum Residual water level R.W.L. First stratum Second stratum

First j stratum

Stratum for which earth pressure will be calculated

14.4 Water Pressure 14.4.1 Residual Water Pressure (Notification Article 12)

Fig. C- 14.3.2 Cross Section of Soil Layers and Symbols.

The residual water pressure caused by the time delay of water level changes between the sea level and the residual water level shall be calculated using the following equation: (14.4.1)

(2) When y is equal to or greater than hw pw = gw hw

(14.4.2)

where p w： gw： y： h w：

residual water pressure (kN/m2) unit weight of water (kN/m3) depth from the residual water level (m) water level difference (m)

Residual Water Press

(1) When y is less than hw pw = g w y

Residual Water Level

Fig. C-14.4.1 Schematic Diagram of the Residual Water Pressure

[Commentary] For practical design, the residual water pressure calculated by equation (14.4.1) or (14.4.2) should be used as a water pressure acting on the backface wall (see Fig. C-14.4.1). The residual water level is affected by various conditions such as permeability of backfill soil, and tidal range. Normally the height hw will be 1/3 ~ 2/3 of the tidal range.

14.4.2 Dynamic Water Pressure during Earthquake (Notification Article 20) (1) The dynamic water pressure during an earthquake shall be calculated using the following equation:

pdw = ± 7 kg w Hy where p dw： k： gw： H： y：

(14.4.3)

8

dynamic water pressure (kN/m2) seismic coefficient unit weight of water (kN/m3) height of structure below the still water level (m) depth from the still water level (m)

(2) The resultant force of dynamic water pressure and its acting depth shall be calculated by the following equation:

pdw = ± 7 kg w H 2 æç = ò H 7 kg w Hydy ö÷ 0 12

è

8

ø

hdw = 3 H 5 -205-

(14.4.4)


where p dw： resultant force of dynamic water pressure (kN/m) hdw： depth of the acting point of the resultant force from the still water level (m) [Commentary] Equation (14.4.3) has been derived for the dynamic water pressures induced by stationary oscillations of water 1). In case of a structure that have the free water surface in both sides such as a breakwater, the magnitude of dynamic water pressure should be two times that calculated above. [Reference] 1) H. M. Westergaard: “Water pressures on dams during earthquakes”, Transactions of ASCE, No. 1835, 1933, pp. 418-472

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Chapter 15 Loads 15.1 General When designing port and harbor facilities, loads shall be taken into consideration as necessary. [Technical Notes] For port and harbor facilities design, loads are divided into deadweight and surcharge. (1) Deadweight: the weight of the structure itself (2) Surcharge: the weight loaded on top of the structure, and it is divided into static load and live load. (a) Static load The load such as bulk cargo loaded onto aprons, cargo in transit sheds and warehouses, etc. are included in static load. In regions with heavy snowfall, the snow load also is a kind of static load. (b) Live load The following must be considered as live load, when designing port and harbor facilities. ① Train load ② Vehicle load ③ Cargo handling equipment load ④ Sidewalk live load

15.2 Deadweight and Surcharge (Notification Article 24) (1) The unit weights of materials in design of port and harbor facilities shall be the values given in Table 15.2.1, except for those cases where another unit weight can be specified by means of advance surveys and others. (2) The surcharge acting on port and harbor facilities shall be appropriately determined with due consideration for the service conditions of the port and harbor facilities such as the type and volume of cargo handled and the handling conditions. The types of surcharge to be taken into consideration include static load, snow load, train load, vehicle load, cargo handling equipment load, sidewalk live load, and any other loads that will have an effect on the design of the port and harbor facilities. Table 15.2.1 Unit Weights of Materials (Notification Article 24, Clause 5) Unit weight (kN/m3)

Material Steel and casting steel

77.0

Casting iron

71.0

Aluminum

27.5

Reinforced concrete

24.0

Plain concrete

22.6

Timber

7.8

Asphalt concrete

22.6

Stone (granite)

26.0

Stone (sandstone)

25.0

Sand, gravel, and rubble (dry)

16.0

Sand, gravel, and rubble (wet)

18.0

Sand, gravel, and rubble (saturated)

20.0

15.3 Static Load 15.3.1 Static Load under Ordinary Conditions Static load under ordinary conditions shall be determined based on adequate consideration of the factors such as the type, shape and volume of cargo handled, the handling method, and the loading period.

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[Technical Notes] (1) The static load of apron is 10 ~ 30 kN/m2 in many cases. For aprons that handle cargo of large weight such as steel materials, the value of the static load should be determined after investigating the cargo handing condition. (2) The unit weights for bulk cargo have been obtained based on surveys of past actual conditions, which are listed in Table T- 15.3.1. Table T- 15.3.1 Unit Weights of Bulk Cargo (Units: kN/m3) Commodity

Unit weight

Coke

4.9

Coal (bulk)

8.8 ～ 9.8

Coal (dust)

9.8 ～ 11.0

Iron ore

20.0 ～ 29.0

Cement

15.0

Sand, gravel and rubble

19.0

15.3.2 Static Load during Earthquake Static load during an earthquake shall be determined by considering the condition of the static load at the time when the earthquake considered in the design should occur. [Technical Notes] The conditions of static load during an earthquake should be examined separately for the respective types of facilities such as transit sheds, warehouses, open storage yards and aprons. Static load to be used in design during an earthquake should be determined according to their conditions of usage. On the other hand, for facilities such as aprons used as cargo handling facilities where cargo is only placed temporarily, the static load will vary tremendously whether cargo loading operations are underway or not. For this reason in design of quaywall the static load on an apron during an earthquake is normally assumed to be 1/2 of that under service conditions.

15.3.3 Unevenly Distributed Load When calculating the stability of a structure as a whole, the unevenly distributed load may be converted to an uniform load in an area of an apron, transit shed or warehouse. However, where a large concentrated load is likely to act on the structure, this concentrated load shall be taken into consideration without converting to an uniform load. 15.3.4 Snow Load For regions where snowfall is large, the snow load should be determined according to the regional conditions. [Technical Notes] (1) For quaywalls where snow removal operations will be carried out, it is often sufficient to determine the snow load with the accumulated weight of snow over one night. In this case the snow load should be determined by taking into consideration the trend in snowfall, snow quality and snow removal operation, based on thorough consideration of past snowfall records. (2) In most cases the snow load is set as 1 kN/m2. This is equivalent to, for example, approximately 70 ~ 100 cm thickness of dry, new powder snow. (3) The relationship between normal snow conditions and snow unit weight, described in the “Railway Structure Design Standards and Commentary”, is shown in Table T- 15.3.2. Table T- 15.3.2 Normal Snow Conditions and Unit Weight of Snow (Units: kN/m3) Normal snow conditions

Unit weight

Dry powder snow Dry powder snow accumulated by winds Snow with medium water content Snow with high water content

1.2 1.7 4.4 8.3

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15.4 Live Load 15.4.1 Train Load Train load shall be applied in such a way to induce the maximum effect on the structures or their members, by taking into consideration the net car weight, loaded weight, and axle arrangement of wheels. Train load shall be applied in principle as the full set of multiple loads in succession without dividing it into two or more separate sets. 15.4.2 Vehicle Load Vehicle load shall be determined according to the “Highway Bridge Specification and Commentary”. [Technical Notes] The regulations concerning the dimensions and maximum gross mass of containers used for marine transport have been set by the “International Organization for Standardization (ISO)” as listed in Table T- 15.4.1.

Designation

Table T- 15.4.1 Standard Demension of Containers Length (L) mm

Tolerance ft mm

Width (W) in

Tolerance in

mm

Tolerance mm

ft

Tolerance in

1 AAA 1 AA 12,192 1A

0 -10

0 -3/8

40

2,438

0 -5

Maximum gross mass

Height (H)

8

0 -3/16

Tolerance mm

ft

in

Tolerance in

2,896

0 -5

9

6²

0 -3/16

2,591

0 -5

8

6²

0 -3/16

2,438

0 -5

8

mm

0 -3/16

1 AX

0 V

(1.3.1)

where V： displacement (m3) I： moment of inertia of the cross-sectional area at the still water level about the long axis (m4) C： center of buoyancy G： center of gravity M： metacenter For safety the distance GM should be 5% of the draft or greater. (2) Equation (1.3.1) applies when the cross section of caisson is symmetrical and no significant tilt is expected.

Fig. T- 1.3.1 Stability of Caisson

(3) Equation (1.3.2) or (1.3.3) applies when a counterballast is used. (a) When using water as the counterballast:

I -----¢(I ¢ S i ) C ¢ G ¢ > 0 V

(1.3.2)

(b) When using sand, stone, or fresh concrete as the counterballast:

I¢ -----¢ C ¢ G ¢> 0 V

(1.3.3)

where i： moment of inertia of the water surfaces of inside chambers about the centerline parallel to the rotating axis of the caisson (m4) V ¢, I ¢, C¢, G¢： displacement, moment of inertia, center of buoyarcy, and center of gravity, of caisson with a counterballast respectively

1.4 Design External Forces 1.4.1 Combination of Loads and Load Factors The combination of loads and the load factors shall be appropriately considered under the following conditions, respectively: (1) Breakwater ① under no waves ② under waves ③ during execution (2) Quaywall ① under ordinary conditions ② during an earthquake ③ during execution [Technical Notes] (1) Table T- 1.4.1 lists the ultimate limit state safety factors and the serviceability limit state influence coefficients on crack width (see Part Ⅲ, 3.2 Basics of Design Based on the Limit State Design Method) to be multiplied to the load characteristics for various combinations of loads to be considered in design. Footings may be treated in the same way as bottom slabs.

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Table T- 1.4.1 Loads and Load Factors (a) Breakwaters Conditions

Deadweight

Hydrostatic pressure

0.9 (1.0)

1.1 (1.0)

Internal earth pressure

Bottom slab reaction

Internal water pressure

Uplift pressure

Variable bottom slab reaction

Variable internal water pressure

Wave force

Hydrostatic head between chambers

1.1 (1.0)

Remarks

Bottom slab

Under no waves

1.1 (1.0)

1.1 [0.9] (1.0)

1.1 [0.9] (1.0)

1.1 (1.0)

outer wall

1.1 [0.9] (1.0)

1.3 [0.7] (1.0)

1.2 [0.8] (1.0)

Bottom slab

0.9 (1.0)

Under waves

1.3 (1.0) outer wall

1.1 (1.0)

0.9 (0.5)

During execution

1.1 (1.0)

1.2 (1.0)

1.1 (0.5)

Bottom slab (while afloat)

1.1 (0.5)

outer wall (while afloat)

1.1 (0.5)

Partition wall (during installation)

(b) Quaywalls Conditions

Under ordinary conditions

During an earthquake

Deadweight


0.9 (1.1)

1.1 (1.0)


1.1 (1.0)

1.0 (-)


Bottom slab plate reaction under permanent load

Surcharge

1.1 (1.0)

0.8 (0.5)

Bottom slab reaction during an earthquake

Hydrostatic pressure during installation Load during execution

Hydrostatic pressure in still water

Remarks

Bottom slab (bottom slab reaction considering surcharge during no earthquake)

1.1 (1.0)

outer wall

1.0 (-)

1.0 (-)

Bottom slab (bottom slab reaction considering surcharge during an earthquake)

1.0 (-)

0.9 (0.5)

During execution

1.1 (0.5)

1.1 (0.5)

Bottom Slab (while afloat)

1.1 (0.5)

outer wall (while afloat)

Partition wall (during installation)

Note: When considering seismic force, the characteristic loads are calculated according to Part Ⅱ, Chapter 12 Earthquakes and Seismic Force.

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PART IV PRECAST CONCRETE UNITS

(2) The values in the upper rows of the respective boxes in Table T-1.4.1 are the load factors to be used when studying the ultimate limit state. The values in the brackets [ ] are the load factors to be used when a smaller load yields the larger design load of the members. The values in the parentheses ( ) in the lower rows show the influence coefficient on crack width in the serviceability limit state. (3) Loads during execution have a shorter duration of action than other conditions, and only occur during execution. Therefore, the influence coefficient on crack width (kp, kr) in the serviceability limit state can be set at 0.5. (4) The design loads for outer walls of breakwater caissons are shown in Figs. T- 1.4.1 ~ 1.4.3. The load factors and the influence coefficients on crack width are listed in Tables T- 1.4.2 ~ 1.4.4. (a) Front wall (parallel to centerline: seaside) Table T- 1.4.2 Load Factors and Influence Coefficient on the Crack Width for Front Wall (Breakwater) Direction of load Load from outside Load from inside

Design conditions

Ultimate limit state

Serviceability limit state

Under wave crest

1.3H - 0.9D

1.0H - 1.0D

While afloat

1.1Sf

0.5Sf

Under wave trough

1.1D + 1.1S + 1.2DS

1.0D + 1.0S + 1.0DS

Note: 1) The load from outside is chosen as the larger one of the above two load conditions. 2) For the symbols in the table, see Fig .T-1.4.1.

Fig. T- 1.4.1 Design Loads for Front Wall (Breakwater)

(b) Rear wall (parallel to center line: land side) Table T- 1.4.3 Load Factors and Influence Coefficient on Crack Width for Rear Wall (Breakwater) Direction of load

Design conditions



Load from outside

While afloat

1.1Sf

0.5Sf

Load from inside

In the lowest water level under no wave

1.1D ＋ 1.1S

1.0D ＋ 1.0S

Note: For the symbols in the table, see Fig. T-1.4.2.

-245-


Under no waves During execution (while afloat) Hydrostatic pressure

Internal water pressure Internal earth pressure

in the lowest water level under no wave

Fig. T- 1.4.2 Design Loads for Rear Wall (Breakwater)

(c) Sidewall (perpendicular to levee normal) Table T- 1.4.4 Load Factors and Influence Coefficient on Crack Width for Sidewall (Breakwater) Direction of load

Design conditions



Load from outside

While afloat

1.1Sf

0.5Sf

Load from inside

Under wave trough

1.1D ＋ 1.1S ＋ 1.2DS

1.0D ＋ 1.0S ＋ 1.0DS


Under waves (trough) During exception (while afloat) Hydrostatic pressure

Internal water

Internal earth

pressure in the lowest

pressure

water level

Variable internal water pressure caused by the fluctuation of water surface

Fig. T- 1.4.3 Design Loads for Sidewall (Breakwater)

(5) The design loads for the sidewalls of quaywall caissons are shown in Fig. T- 1.4.4. The load factors are listed in Table T- 1.4.5. Table T- 1.4.5 Load Factors and Influence Coefficient on Crack Width for Outer Wall (Quaywall) Direction of load

Design conditions



Load from inside

In the lowest water level

1.1D ＋ 1.1S

1.0D ＋ 1.0S

Load from outside

Wile afloat

1.1Sf

0.5Sf


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(a) Under ordinary conditions (loads from inside)



Fig. T- 1.4.4 (a) Design Loads for Sidewalls (Quaywall)

(b) While afloat (loads from outside)


Draft + 1.0m

Fig. T- 1.4.4 (b) Design Loads for Sidewalls (Quaywall)

(6) The loads for the bottom slabs of breakwater caissons while Table T- 1.4.6 Load Categories under afloat are calculated by multiplying the load characteristic Wave Force (Breakwater) by the load factors and the influence coefficient on the crack Load category Load width given in Table T- 1.4.1. The loads acting on the bottom slabs of breakwater Composite load under no Permanent caissons after construction are shown in Fig. T- 1.4.5. The waves (D0) composite load under no waves (D0) is treated as a Variable bottom slab reaction Variable permanent load. The composite load under waves consists (ΔR), uplift pressure (U) of the composite load under no waves (D0), the variable bottom slab reaction (ΔR ), and the uplift pressure (U) as shown in Fig. T- 1.4.5. The load can be calculated by using the equations listed in Table T- 1.4.7 in accordance with the classification in Table T- 1.4.6. Under no waves

Under waves

Weight of filling Weight of concrete lid Weight of bottom slab

Weight of filling Weight of concrete lid Weight of bottom slab


Hydrostatic pressure Bottom slab reaction under wave force

Bottom slab reaction under no waves Composite load under no wave forces

Variation bottom slab reaction Uplift pressure Fig. T- 1.4.5 Design Load of Bottom Slab (Breakwater) -247-


Table T- 1.4.7 Combination of Load with Load Factors or Influence Coefficient on Crack Width (Breakwater) Limit state

Direction of ΔR and W

Condition

ΔR ↑ Wave crest

ΔR ↓


ΔR ↑ Wave trough

ΔR ↓ Serviceability limit state

All

Load factors and combination of load

W↑

1.1D0 + 1.2ΔR + 1.3U

W↑

1.1D0 + 0.8ΔR + 1.3U

W↓

0.9D0 + 1.2ΔR + 0.7U

W↑

1.1D0 + 1.2ΔR + 0.7U

W↓

0.9D0 + 0.8ΔR + 1.3U

W↑

1.1D0 + 0.8ΔR + 0.7U

W↓

0.9D0 + 1.2ΔR + 1.3U

(*)

(*)

1.0D0 + 1.0ΔR + 1.0U

－

Note: 1) W = D0 + ΔR + U. The summation of the loads D0, ΔR and U should be done by considering the direction of each load. 2) (*) When variable bottom slab reaction (ΔR) acts downwards, the value of 1.2 |ΔR| cannot exceed that of 1.1 | R|. Therefore, if 1.2|ΔR|>1.1| R|, the combination of loads is replaced by the equation 0.9D0+1.1| R|+0.7 (1.3) U.s

(7) The design loads of bottom slabs of quaywall caissons while afloat are calculated by multiplying the load characteristic by the load factors given in Table T- 1.4.1. Loads acting on the bottom slab are shown in Fig. T- 1.4.6. The resultant force that is composed of the weight of filling and concrete lid, hydrostatic pressure and bottom slab reaction is treated as the permanent load. Surcharge is treated as the variable load and the bottom slab reaction in an earthquake is also treated as the variable load. The design loads can be calculated by the equation listed in Table T- 1.4.8. Table T- 1.4.8 Load Combinations (Quaywall) Ultimate limit state

Serviceability limit state (*)

Under ordinary conditions

0.9D + 1.1R + 1.1F + 0.8W

1.0D + 1.0R + 1.0F + 0.5W


0.9D + 1.0F + 1.0R¢ + 1.0W ¢

Study not required

While afloat

0.9Df + 1.1Sf

0.5Df + 0.5Sf

Note: For the symbols in the table, see Fig. T-1.4.6. (While afloat)

(During earthquake)

(Under ordinary conditions)

Weight of bottom slab

Weight of concrete lid Weight of filling Weight of bottom slab



Surcharge during

Surcharge

an earthquake

in normal condition

R



under ordinary

during an earthquake

condition

Fig. T- 1.4.6 Design Loads of Bottom Slabs (Quaywall)

(8) For examination of the bending moment for partition walls, the design load is the hydrostatic head between chambers during installation. The design values are as follows: Ultimate limit state: Serviceability limit state:

1.1S 0.5S

(1.1 for load factor: gf ) (0.5 for influence coefficient on crack width: kp)

The symbol S represents the load characteristic. For examination of dislodging, the design load should be determined in the condition where the dislodging load is largest among the design loads indicated for the bottom slab and sidewall.

-248-


1.4.2 External Forces during Fabrication When a caisson is being fabricated on a dry dock or floating dock, external forces during fabrication may not be considered. However, when the caisson is raised by a jack or is placed on a launch truck to be moved to a slipway or caisson platform, concentrated loads of selfweight shall be applied on the caisson as design loads. [Technical Notes] The whole caisson should be regarded as a simple beam to examine the section force during fabrication.

1.4.3 External Forces during Launching and Floating When launching and floating a caisson fabricated on a dry dock, floating dock, or ordinary slipway, the hydrostatic pressure at the design draft shall be increased by an allowance as a safety factor to determine the external force. If the hydrostatic pressure more than this might act on a caisson temporarily during launching, it shall be considered separately. [Technical Notes] (1) Outer Wall The design water pressure distribution on the outer wall should be considered as follows (see Fig. T- 1.4.7):

Allowance 1.0m

① The draft of caisson is increased by 1.0 m.

③ The water pressure distribution is taken as a triangle shape with the bottom value calculated above and extending to the crest of outer wall. (2) Bottom Slab The external load acting on the bottom slab should be the value obtained by subtracting the deadweight of the bottom slab from the hydrostatic pressure at the bottom slab. (see Fig. T- 1.4.8)

Draft

② The water pressure on bottom is calculated with the draft mentioned above.

w Fig. T- 1.4.7 Water Pressure Acting on Outer Wall

p2 = pw - w = w0H0 - w

(1.4.1)

where p2： characteristic pressure acting on the bottom slab (kN/m2) pw： characteristic hydrostatic pressure acting on the bottom slab with an allowance of about 1.0 m added to the design draft of caisson (kN/m2) w： characteristic deadweight of the bottom slab (including the weight of filling material as counterballast, if any), without subtracting buoyancy (kN/m2) w0： characteristic specific weight of sea water (kN/m3) H0： water depth with an allowance of about 1.0 m added to the design draft of caisson.

Hydrostatic head

Fig. T- 1.4.8 External Force Acting on the Bottom Slab

-249-

Fig. T- 1.4.11 Hydrostatic Head across Chambers


1.4.4 External Forces during Installation [Technical Notes] (1) Hydrostatic pressures acting on the sidewalls and bottom slab are not considered as the external force during installation. (2) The hydrostatic head between chambers should be applied to the partition walls as the external forces. (see Fig. T- 1.4.11).

1.4.5 External Forces after Construction [1] Outer Walls The earth pressure of filling and the internal water pressure shall be considered as the external forces on the outer walls. The wave forces acting on the front wall shall also be considered in the design of the caissons for breakwaters. [Technical Notes] (1) Earth Pressure of Filling (a) The distribution of the composite load often takes an irregular shape. For design purposes, however, the irregular distribution can be modified as an equivalent uniform or triangular distribution. (b) The coefficient of earth pressure is set at be 0.6. However, the earth pressure is disregarded when the filling consists of concrete blocks or fresh concrete. (c) It should be assumed that the earth pressure increases to the depth equal to the width of the chamber, beyond which it remains at a constant value. (see Fig. T- 1.4.12).

Fig. T- 1.4.12 Earth Pressure of Filling

The symbols in Fig. T-1.4.12 are defined as follows: q ： characteristic surcharge on top of filling (kN/m2) g '： characteristic submerged specific weight of filling material (kN/m3); generally g '=10 kN/m3 can be used. K： coefficient of earth pressure of filling; K = 0.6 b： width of chamber (m) ; b = H (d) When a concrete lid or crown is placed so tightly above the caisson, the effect of surcharge on the concrete lid or crown can be neglected. (2) Internal Water Pressure The internal water pressure should be the head difference between the water levels in the caisson and the lowest water level (LWL) outside the caisson. However, in case of wave troughs acting on the front walls parallel to the centerline of breakwater or the sidewalls perpendicular to the centerline, the external water level should be taken at the depth (H1/3)/2 below LWL. (3) Wave Force For the front walls of caissons parallel to the centerline of the breakwater, wave force should be taken into account when wave crests act on the walls. (4) Distribution of Internal Earth and Water Pressures The distributions of internal earth and water pressures for the walls of several caissons are shown in Fig. T1.4.13. A breakwater caisson covered with a mound of wave-dissipating concrete blocks is also subject to impacts of wave-dissipating blocks to the front walls. Depending on the locality, it is also affected by collisions of drifting ice or driftwoods, and by sea water freezing. The effects of these are not known. Therefore, the upper sections of the walls should be reinforced as a precautionary measure.

-250-




Composite load

(a) Breakwaters (front walls)

Internal earth Internal water Composite load pressure pressure

(b) Breakwaters (rear walls parallel to the centerline or sidewalls perpendicular to the centerline)


Internal water Composite load pressure

(c) Quays (outer walls parallel or perpendicular to the centerline)

Wave pressure


Composite load

(d) Wave forces Fig. T- 1.4.13 Internal Earth and Water Pressures Acting on Caisson Walls

[2] Bottom Slab (1) For the bottom slab fixed with the outer walls and partition walls, the bottom reactions, hydrostatic pressure, uplift, weight of filling, weight of concrete lid, weight of bottom slab, and surcharge shall be considered as the external forces. (2) Footings The bottom reaction, weight of footing with buoyancy considered, and surcharge on the footing shall be considered as the external forces acting on the footing. [Technical Notes] (1) Bottom slab (a) The distribution of the composite load often takes an irregular shape. For design purposes, however, the -251-


irregular distribution can be modified as an equivalent uniform or triangular distribution. (b) Bottom reaction The characteristic bottom reaction should be calculated according to equations (1.4.3) and (1.4.4) (see Fig. T1.4.14). The value e is calculated using equation (1.4.5) 1 ①If e ≦ --- b 6 b e = --- x 2

(1.4.3)

Mw M h x = --------------------V

6447448

6e V p 2 = æ 1 ------ö --è bøb

6447448

6e V p 1 = æ 1 + ------ö --è bøb

(1.4.5)

1 ②If e > --- b 6

b b ¢ = 3 æ --- eö è2 ø where p1： p2： V：

644474448

2 V p 1 = --- ----------------3æb ö --- e è2 ø

(1.4.4)

characteristic reaction at the front toe (kN/m2) characteristic reaction at the rear toe (kN/m2) characteristic vertical resultant force per unit length (kN/m) H： characteristic horizontal resultant force per unit length (kN/m) e： eccentricity of resultant force of V and H (m) b： width of bottom (m) 1 b¢： distribution width of bottom reactions in case of e > --- b (m) 6 Mw： characteristic moment for point A by the vertical resultant force (kN•m/m) Mh： characteristic moment for point A by the horizontal resultant force (kN•m/m)

(c) Hydrostatic pressure The hydrostatic pressure acting on the bottom slab should be considered at the design tide level. (d) Uplift pressure When wave forces act on the caisson, the uplift pressure should be considered. For calculating the uplift, Chapter 5 Wave Force should be referred to. (e) Weight of filling The specific weight of the filling material is normally determined by the tests of the material to be used.

If

If

Fig. T- 1.4.14 Bottom Reaction

(f) Weight of concrete lid The weight of the concrete lid should be the dry weight without influence of buoyancy. For design calculations, the characteristic secific weight can be set as 22.6 kN/m3 for plain concrete and 24.0 kN/m3 for reinforced concrete. (g) Weight of bottom slab The weight of the bottom slab should be the dry weight without influence of buoyancy. For design calculations, the characteristic specific weight can be set as 24.0 kN/m3 (h) Surcharge The weight of soil on top of the caisson and superimposed load should be considered for the surcharge acting on the bottom slab. However, when the concrete lid or concrete crown is placed so tightly above the caisson, the effect of the surcharge on top of the concrete lid or concrete crown upon the bottom slab through the filling can be neglected.

-252-


(2) Footing (a) The design loads acting on the footing should be determined with respect to the load distributions as shown in Fig. T- 1.4.15 (b) Bottom reaction The bottom reactions acting on the footing should be the values calculated according to equations (1.4.3) or (1.4.4).

p : Bottom reaction (kN/m2)

(c) Weight of footing The weight of footing is the submerged weight with buoyancy considered. The characteristic value of the specific weight of footing in air can be set as 24.0 kN/m3.

w : Weight of footing (with buoyancy considered) (kN/m3) w1: Surcharge on the footing (kN/m2)

(d) Surcharge The weight of wave-dissipating concrete blocks of breakwaters, the weight of overburden soil and/or surcharge of quaywalls should be considered for the surcharge acting on the footing.

p1: Composite load (kN/m2)

Fig. T- 1.4.15 Design Load of Footing

[3] Partition Walls and Others (1) Partition Walls (a) In the examination of dislodging of an outer wall from the partition walls due to tensile failure, the earth pressure of filling and the internal water pressure acting on the outer wall shall be considered, assuming that they act in the joint between the partition walls and the outer wall. (b) In the examination of dislodging of the bottom slab from the partition walls, the following loads shall be considered: ① ② ③ ④ ⑤ ⑥ ⑦

weight of filling surcharge weight of bottom slab weight of concrete lid bottom slab reaction uplift pressure hydrostatic pressure

(2) When a load is possibly generated due to the unevenness of the ground bearing capacity, it shall be examined. [Technical Notes] (1) The design load to examine dislodging of an outer wall from the partition walls should be determined with respect to the load distribution shown in Fig. T- 1.4.16. (2) The design load to examine dislodging of the bottom slab from the partition walls should be determined with respect to the load distribution shown in Fig. T- 1.4.17 (3) Partition Wall The load sharing divisions are based on those for ordinary slab supported by beams.

Converted design load Composite load

Internal earth pressure + internal water pressure (kN/ m2)

Internal earth pressure + internal water pressure between partition wall and outer wall.

Fig. T- 1.4.16 Design Load to Examine Dislodging of Outer Wall from Partition Walls. -253-


w : Design load acting on bottom slab (kN/m2)

P : Load acting in joint

between partition wall and bottom slab (kN/m)

Pd: Converted design load (kN/m)

Fig. T- 1.4.17 Design Load to Examine Dislodging of Bottom Slab from Partition Walls

(4) Examination of External Load by Unevenness of Ground Bearing Capacity When examining the external load by the unevenness of ground bearing capactiy, the caisson can be assumed as a cantilever beam with the span equivalent to 1/3 of the length or width of the caisson (see Fig. T- 1.4.18).

1.5 Design of Members Fig. T-1.4.18 Examination on the Unevenness of Ground Bearing Capacity

1.5.1 Outer Wall [Technical Notes]

(1) An outer wall should be designed as a slab fixed on three sides and free on one side. (2) The span used in calculations is the central interval. (3) The covering of main reinforcements should not be less than the following values in principle: Outer side: 7 cm Inner side: 5 cm

1.5.2 Partition Wall [Technical Notes] (1) The partition wall should be designed as a slab fixed on three sides and free on one side. (2) After installation, safety in the ultimate limit state should be verified for dislodging failure of outer walls or bottom slab from partition walls. (3) The span used in calculations is the central interval. (4) The covering of reinforcements should not be less than 5 cm.

1.5.3 Bottom Slab [Technical Notes] (1) The bottom slab surrounded by outer walls and partition walls should be designed as a slab fixed on four sides. (2) The covering of main reinforcements should not be less than the following values in principle: Outer side: 7 cm Inner side: 5 cm -254-


(3) Footing should be calculated as a cantilever slab. (4) The span of bottom slab used in calculation should be set on the central interval (see Fig. T- 1.5.2). (5) The cross section used to examine bending of footing should be the front surface of the outer wall (see Section A-A of Fig. T- 1.5.2). The cross section for examining shear of footing should be located at the distance equivalent to 1/2 of the footing height from the front of the caisson. To calculate the height of footing, a part of the haunch with the gradient steeper than 1:3 should be disregarded.

Footing

Portion fixed on four sides

Fig. T- 1.5.2 Span Used to Design Bottom Slab and Footing

1.5.4 Others [Technical Notes] The whole caisson should be examined as a simple beam when lifted with jacks for transportation or when making analysis for differential settlement after installation.

1.6 Design of Hooks for Suspension by Crane The load acting on the hooks shall appropriately be calculated by considering the weight of the caisson and the bonding of the bottom slab to the floor of caisson fabrication yard.

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Chapter 2 L-Shaped Blocks 2.1 General (1) The provisions in this chapter shall be applied to the design of ordinary L-shaped blocks. (2) Design shall follow the limit state design method. [Technical Notes] (1) The design of L-shaped blocks can be made according to the sequence depicted in Fig. T-2.1.1. Assumption of dimensions of L-shaped block members ↓ Determination of design external forces During suspension After construction ↓ Design of members

Design conditions

(Design of bottom slab)

↓ Calculate section force ↓ Examine bending

(Design of front wall)

↓ Calculate section force ↓ Examine bending

(Design of buttress)

ultimate limit state and { [1] [2] serviceability limit state

↓ Calculate section force ↓ Examine bending ↓ Examine dislodging

(Design of footing)

ultimate limit state and { [1] [2] serviceability limit state

ultimate limit state and { [1] [2] serviceability limit state [1] ultimate limit state

↓ Calculate section force ↓ Examine bending ↓ Examine shear

ultimate limit state and { [1] [2] serviceability limit state ultimate limit state and { [1] [2] serviceability limit state

Fig. T- 2.1.1 Sequence of Design of L-Shaped Block

2.2 Determination of Dimensions The dimensions of the members L-shaped block shall be determined by considering the following: (1) Capacity of the facilities for fabricating L-shaped blocks (2) Hoisting capabilities of the crane (3) Water depth in which L-shaped blocks are installed to form a quaywall (4) Tidal range (5) Elevation of coping [Commentary] The crown height of an L-shaped block is determined for ease of execution of the superstructure, considering the water depth and tidal range. [Technical Notes] Figure T- 2.2.1 shows the relationship between the block width and the wall height as well as the L-shaped block height, based on the past examples of construction. The design seismic coefficient in (a) and (b) is taken as 0.05 and 0.10, respectively. As the design seismic coefficient increases, the width increases compared with the height. -256-


h

15

1.54

1

b

0.26

15

14

14

13

13

h

2

1.36

b

0.75

9

h

h

m

h

10

1

h

10

2

0.75

2

Height

8

7

b

9

h

8

h

h

m

b

1.02

2.53

11

11

Height

1

12

12

1.97

1

2

7

6

6

h

5

h1

5

1

h

2

h2

4

4

3

3

b

2

2

1

1

0

h

0 1

2

3

4

5

6

Width

7

8

b

9

10

11

12

13

14

15

b

1

2

3

4

5

6

Width

m

(a) Design seismic coefficient kh = 0.05

7

8

b

9

10

11

12

13

14

15

m

(b) Design seismic coefficient kh = 0.10

Fig. T- 2.2.1 Relationship between Height and Width of L-Shaped Blocks

2.3 Loads Acting on Members 2.3.1 General The following shall be considered as the loads acting on the structural members of L-shaped blocks:. (1) Front Wall (a) Surcharge and earth pressure by backfilling stones or soil (b) Residual water pressure (2) Footing (a) Bottom reaction (b) Deadweight of footing (3) Bottom Slab (a) Surcharge (b) Weight of backfill (c) Deadweight of bottom slab (d) Bottom reaction (4) Buttress (a) Reaction from the front wall (b) Reaction from the bottom slab (c) Earth pressure and residual water pressure acting on the rear of buttress. In addition, the following loads shall be considered during execution. (d) Load during suspension of the L-shaped block (e) Load while setting up the L-shaped block [Technical Notes] (1) Load combinations and load factors are the same as described in 1.4.1 Combination of Loads and Load Factors. (2) Loads acting on the structural members of L-shaped blocks are illustrated in Fig. T- 2.3.1. (3) The calculation methods of earth pressure acting on L-shaped block members are described in PartⅡ Ⅱ, Chapter 14 Earth Pressure and Water Pressure. The angle of friction of the wall should be taken as δ= 15º. (4) The calculation methods of bottom reaction are described in 1.4.5 External Forces after Construction. (5) There are vertical and horizontal methods for placing concrete when manufacturing L-shaped blocks. The latter involves setting-up of the block before the installation work. Therefore, the load at setting up the block examined be examined in the design. -257-


q

K 1q

h3

Dead weight Bottom reaction

(a) Load acting on front wall and buttresses

(b) Load acting on footings

(c) Load acting on bottom slab

Notation q: surcharge (kN/m2) g1: specific weight of soil above residual water level (kN/m3) g2: specific weight of soil below residual water level (kN/m3) gw: specific weight of water (kN/m3) h1: thickness of soil layer above residual water level (m) h2: thickness of soil layer below residual water level (m) h3: tidal range (m) h4: thickness of bottom slab (m) K1: coefficient of earth pressure of soil above residual water level K2: coefficient of earth pressure of soil below residual water level w: deadweight of bottom slab (kN/m2) Fig. T- 2.3.1 Loads Acting on L-Shaped Blocks

2.3.2 Earth Pressure The earth pressure to be used for design of the structural members of L-shaped blocks shall be the earth pressure used for stability calculation of the quaywall. 2.3.3 Converted Loads for Design Calculation The loads acting on an L-shaped block can be converted to multiple loads of uniform distribution for the convenience of calculation. When converting loads, the structural safety of members in design shall be taken into account. [Technical Notes]

Converted load distribution for member design

Converted load distribution for member design

Footing

Bottom slab

Load distribution

Load distribution

Normally, the loads acting on L-shaped blocks are not uniform loads. This is inconvenient when designing the members. Therefore, the loads may be converted into multiple uniform loads. When doing so, care should be taken not to produce weak points in the strength of members. Fig. T- 2.3.2 shows examples of converting loads.

(i)

( ii )

(a) Earth pressure

(b) External loads on footing and bottom slab

Fig. T- 2.3.2 Converting Methods of Load Distribution

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2.4 Design of Members 2.4.1 Front Wall [Technical Notes] (1) Front wall should be designed as a slab supported by buttresses. If one buttress is considered as a support, the front wall should be designed as a cantilever slab supported by a buttress. If two or more buttresses are considered as supports, the front wall should be designed as a continuous slab. (2) The span of a front wall should be measured between the centers of the buttresses. (3) The loads should be assumed to act on the whole span of the front wall. (4) The covering of main reinforcements of a front wall should not be less than the following values in principle: Sea side: Land side:

7 cm 5 cm

(5) The span of the front wall and the loads acting on it can be taken as shown in Fig. T- 2.4.1. (6) Structurally, the front wall is supported by the bottom slab as well as by buttresses. Therefore, the front wall may be regarded as being supported on two or three sides. Generally, the front wall of L-shaped blocks with large heights are not greatly affected by support of the bottom slab, while the arrangement of reinforcing at the bottom slab attachment becomes complicated. Thus, in principle, the front wall can be designed as a cantilever slab or a continuous slab supported by buttresses. When it is advantageous in design to treat the front wall as a slab supported on two or three sides, however, the descriptions in this document are not necessarily applied. p

1

earth pressure + residual water pressure

1

,

2

span of member

Buttress

span of member 2

Buttress

Front wall

p

Buttress

Front wall

earth pressure + residual water pressure

Support

Support

1

(a) For one buttress

(b) For two or more buttresses

Fig. T- 2.4.1 Assumption of Span of Member and Load

2.4.2 Footing [Technical Notes] Front wall

(1) Footing should be designed as a cantilever slab supported by the front wall. (2) The length and load of footing should be as shown in Fig. T- 2.4.2. (3) The covering of reinforcements should not be less than 7 cm in principle.

a Footing

2.4.3 Bottom Slab

a

[Technical Notes]

: length of footing

a (1) The bottom slab should be designed as a slab fixed on one or more butp tresses. If one buttress is considered as a support, the bottom slab should a be designed as a cantilever slab supported by the buttress, and if two or p = (Bottom reaction) - (Weight of footing) more buttresses are considered as supports, it should be designed as a continuous slab. Fig. T- 2.4.2 Length and Load of Footing

-259-


(2) The span of a bottom slab should be measured between the centers of the buttresses. (3) The load should be assumed to act on the whole span of the bottom slab. (4) The covering of main reinforcements of the bottom slab should not be less than the following values in principle: Bottom side: 7 cm Upper side: 5 cm (5) Structurally, the bottom slab may be regarded as supported by the front wall as well as by buttresses. Therefore, the bottom slab may be designed as a slab with supports on two or three sides. For the same reason as stated in 2.4.1 Front Wall [Technical Notes] (6), it should be designed as a cantilever slab or a continuous slab supported by buttresses as standard. However, when advantageous in design to treat the bottom slab as a slab with supports on two or three sides, the descriptions in (1) is not necessarily applied. (6) Of the loads acting on the bottom slab, the bottom reaction acts on the whole span. The loads of backfill and surcharge act only on the clear span of the bottom slab. However, since it is troublesome to consider it precisely in design and it does not greatly affect the design of bottom slab, the loads of backfill and surcharge may be taken as acting on the whole span.

2.4.4 Buttress [Technical Notes] (1) Buttress should be designed against the reaction from the bottom slab and front wall. (2) Buttress should be designed as a T-beam combined with the front wall. (3) Buttress should be designed as a cantilever beam supported at the bottom slab against the reaction from the front wall. (4) Buttress should be designed in the cross sections parallel to the bottom slab. (5) Buttress, front wall, and bottom slab should be tightly connected. The amount of reinforcement for the connection should be calculated independently from that of stirrups against shear stresses. (6) The covering of reinforcements of buttress should be 5 cm or larger. (7) If the front wall and bottom slab are designed as specified in this chapter, the load behind the buttress can be ignored. (8) The element length of buttress should be the total height of block including the bottom slab, as shown in Fig. T2.4.3. The load acting on the buttress includes that of the superstructure. crown Superstructure

Notation p： lh： b： H：

p H

sum of earth pressure and residual water pressure element length of buttress width of block height of block

b

Fig. T- 2.4.3 Length and Load of Buttress

2.5 Design of Hooks for Suspension by Crane Hooks for suspension shall be designed in accordance with 1.6 Design of Hooks for Suspension by Crane.

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Chapter 3 Cellular Blocks 3.1 General (1) The provisions in this chapter shall be applied to the design of ordinary cellular blocks. (2) Design shall follow the limit state design method. [Technical Notes] (1) Cellular blocks using the limit state design method should generally be designed in the sequence depicted in Fig. T-3.1.1. (2) For the design of individual members for various types of cellular blocks, the design methods in Chapter 1 Caissons or Chapter 2 L-Shaped Blocks can be used according to respective types. When using cellular blocks for breakwaters or revetments subject to actions of for wave forces, the fatigue limit state should be examined separately. Determination design external forces During suspension During execution After construction ↓

Design of members

Design conditions ↓ Front wall ↓ Rear wall ↓ Sidewall ↓ Partition wall

[1] ultimate limit state [2] serviceability limit state

↓ Bottom slab ↓ Footing ↓ Hooks for suspension

Fig. T- 3.1.1 Sequence of Design of Cellular Blocks

3.2 Determination of Dimensions 3.2.1 Shape of Cellular Blocks The shape of cellular blocks shall be determined to insure the stability of the entire structure. [Commentary] “Cellular blocks” generally refer to blocks consisting of sidewalls without bottom slabs such. They can be used as a wall by piling up several blocks. Some special types have bottom slabs such as those placed on the lowest level. When designing cellular blocks, an appropriate design method has to be adopted after fully ascertaining the characteristics of the block shape.

3.2.2 Determination of Dimensions The dimensions of cellular blocks shall be determined by considering the following items: (1) Capability of the facility for fabricating cellular blocks (2) Hoisting capability of the crane (3) Water depth in which cellular blocks are piled up to form a wall (4) Tidal range -261-


(5) Elevation of coping (6) Solidification of whole blocks when piled up in stages

3.3 Loads Acting on Cellular Blocks 3.3.1 General Loads acting on cellular blocks are as follows: (1) Earth pressure of filling (2) Residual water pressure (3) Deadweight of blocks (4) Weight of coping (5) Surcharge (6) Bottom reaction (7) Loads during execution (8) Wave force [Technical Notes] (1) The rear wall is subject to backfill earth pressure and residual water pressure. But these are mutually cancelled out by the earth pressure of the inner filling, and therefore such loads need not be examined. (2) The earth pressure of the filling and the residual water pressure acting on cellular blocks are shown in Fig. T3.3.1. When fully backfilled as a wall, the stress of sidewalls by the inner filling is considerably reduced by the active earth pressure and residual water pressure in the backfill. Therefore, sidewalls should be designed for the condition during the execution process when the inner filling alone is executed before start of backfill. Sidewall (front wall)

Earth pressure of filling

Sidewall

Sidewall

+ residual water pressure

Sidewall (rear wall)

Fig. T- 3.3.1 Loads Acting on Cellular Blocks

(3) Wave pressure is only taken into consideration when an impulsive wave pressure acts on the blocks. (4) The loads in Chapter 2 L-Shaped Blocks can be used for the loads during execution. (5) For the combinations of loads and the load factors to be taken into account in design, the descriptions in 1.4.1. Combination of Loads and Load Factors can be referred to.

3.3.2 Earth Pressure of Filling and Residual Water Pressure (1) Front Wall, Rear Wall, and Sidewalls The earth pressure of filling and residual water pressure shall be considered when designing the front wall, rear wall and sidewalls. If concrete is placed so tightly above the cellular block, the effect of surcharge on top of the concrete lid or crown can be neglected. (2) Partition Wall The partition wall shall be designed against dislodging failure of the sidewalls from the partition wall due to the earth pressure of filling and residual water pressure.

-262-


[Technical Notes] (1) Front Wall, Rear Wall, and Sidewalls (a) Earth pressure of filling ① The coefficient of earth pressure should be set as 0.6. However, if the filling is blocks or fresh concrete, the earth pressure of the filling should not be considered ② The earth pressure should be assumed to increase from the crown of block to a height equal to the inner width b1 of the cellular block, and it should be constant beyond it. ③ The earth pressure acting on cellular blocks piled up in stages is calculated in a manner as shown in Fig. T3.3.2. However, when the inner width of the lower cellular block is less than that of the upper block (in case of cellular block partitioned by walls), the earth pressure obtained for the upper block can be extended to the lower block. The symbols in Fig. T- 3.3.2 are as follows: q： characteristic value of surcharge (kN/m2) g1： specific weight of filling material above the residual water level (kN/m3) g2： specific weight of filling material below the residual water level (kN/m3) in general, g1=18 kN/m3 and g2 =10 kN/m3 can be used. K： coefficient of earth pressure of filling; K = 0.6 b1： inner width of block chamber (m) ; b1 = H1 (b) The earth pressure of the filling of cellular block should be in accordance with that of caissons (see 1.4.5 [1] Outer Walls.)

Residual water level

Fig. T- 3.3.2 Calculating Earth Pressure of Filling

(c) Residual water pressure ① For quaywalls Residual water pressure is calculated from the head difference between the residual water level and LWL. ②For breakwaters When used as breakwaters or revetments and the wave trough acts on the front of the block, the increase of the residual water level difference should be examined. Part Ⅱ, 5.2 Wave Force Acting on Upright Wall can be referred to for a calculation of water pressure in this case. (2) Partition Wall The characteristic values of loads against dislodging failure of partition walls and sidewalls should be those of the earth pressure acting on the shaded parts in Fig. T- 3.3.3.

-263-


Sidewall (front wall)

Sidewall Partitition

(rear wall)

Load bearing area





per 1 m of width

per

lm

Fig. T- 3.3.3 Load for Examination of Dislodging Failure of Sidewalls from Partition Wall

3.3.3 Converted Loads for Design Calculation The load acting on a cellular block can be converted to multiple loads of uniform distribution for the convenience of calculation in accordance with 2.3.3 Converted Loads for Design Calculation.

3.4 Design of Members 3.4.1 Rectangular Cellular Blocks The members of rectangular cellular blocks shall be designed as appropriate in view of their structural types. [Technical Notes] (1) Sidewalls (a) The covering of main reinforcements should not be less than the following values in principle: Outer side: 7 cm Inner side: 5 cm (b) The section force generated in a rectangular cellular block is solved by assuming the block as a rigid box frame for each unit height against the equivalent uniform load converted from the actual load distribution. (c) The span used for calculations is measured between the centers of the connected walls. (2) Partition Wall (a) The section forces acting on partition walls are calculated in the same way as that of sidewalls. (b) When any difference of filling height between neighboring chambers may occur during execution, the partition wall should be designed against the earth pressure caused by the difference. (c) The span used for calculations is measured between the centers of the connected walls. (d) The covering of main reinforcements should not be less than 5 cm in principle. (3) Footings (a) Footings may be designed as cantilever slabs supported by the sidewalls. (b) The span of footing is the distance from the front of the sidewall to the tip of the footing. (c) The covering of main reinforcements should not be less than 7 cm in principle.

-264-


3.4.2 Other Types of Cellular Blocks The members of other types of cellular blocks shall be designed as appropriate for their structural shapes. [Commentary] Examples of other types of cellular blocks are shown in Fig. T- 3.4.3. (units: m)

Caisson type with

Rectangular type

I-shaped type

bottom slab

with flange

with bottom slab

I-shaped type

Fig. T- 3.4.3 Examples of Other Types of Cellular Blocks

[Technical Notes] (1) Front Wall (a) The front wall should be designed as a slab supported by the sidewalls. When the front wall protrudes beyond the both sides of the frame, the unbalanced moment at the supports is regarded to be transferred to the sidewalls. (b) The span of the front wall should be measured between the centers of sidewalls connected to it. (c) The load from the back of a front wall should act in the whole span of the front wall (see Fig. T- 3.4.4.).

Front wall

Sidewall

member span load strength

Fig. T- 3.4.4 Load on Cantilever Portion of Front Wall

(d) The covering of main reinforcements should not be less than the following values in principle: Sea side: 7 cm Land side: 5 cm

-265-


(2) Rear Wall (a) Filling is often done first during the execution, and the design conditions of rear walls are the same as for the front wall. Therefore, the rear wall may be designed in the same way as for the front wall. (b) The load to the rear wall should be the earth pressure of filling. The active earth pressure behind the wall is not considered in principle. (c) The covering of main reinforcements should not be less then 5 cm. When using cellular blocks for breakwaters, the wall come into direct contact with seawater. In these cases therefore, the covering of main reinforcements on the outside should preferably be 7cm or larger. (3) Sidewalls (a) Sidewalls are in principle designed against the reactions and transferred moments from the front and rear walls (see Fig. T3.4.5).

Front wall

The symbols used in Fig. T- 3.4.5 are as follows.

Rear wall

Side wall

PF： characteristic value of reaction from the front wall (kN) MF： characteristic value of moment transmitted from the front wall (kN•m) PB： characteristic value of reaction from the rear wall (kN) MB： characteristic value of moment transmitted from the rear wall (kN•m) l： distance between supports (m) (b) The span of the sidewalls is the distance between the centers of the front and rear walls. (c) The covering of main reinforcements should not be less than 5 cm in principle. (d) When the height difference of the filling is expected during execution, the strength of the sidewalls should be examined for the earth pressure due to the height difference.

Fig. T- 3.4.5 Forces and Moments on Sidewall of the Cellular Black

(4) Bottom Slab When a bottom slab is used for a cellular block, it may be designed against the following loads: ① Loads on the bottom such as the weight of filling material and surcharge ② Deadweight of the bottom slab ③ Bottom reaction

-266-


Chapter 4 Upright Wave-Absorbing Caissons 4.1 General The provisions in this chapter shall be applied to the design of upright wave-absorbing caissons used for quaywalls, breakwaters, and revetments. The limit state design method shall be used. [Commentary] Upright wave-absorbing caissons having slits in the front wall and internal water retaining chambers with wave dissipating function are now being used for constructing quays and breakwaters. Various shapes and types of construction are now available for wave-absorbing caissons, but they can be roughly divided into permeable and impermeable types. As to the slit shape, the vertical slit type is the most widely used. In addition, there are the horizontal slit and perforated wall types. The general design method for these upright wave-absorbing caissons has not been fully established. In designing the structural members, the characteristics of various structures should be fully surveyed, hydraulic model experiments suited to the conditions should be performed, and then appropriate design should be carried out. [Technical Notes] (1) Upright wave-absorbing caissons can be designed in the sequence given in 1.1 General [Technical Notes] (1). (2) The names of members of the relatively common vertical slit caisson are shown in Fig. T- 4.1.11.

Lower beam

Ordinary caisson part

Side wall slit column Back wall

Slit column

Upper beam

Ceiling slab Front wall

Upper beam

Slit column Partition slit column Lower beam

Lower wall Bottom slab

Lower wall

Lower wall

Base plate

Base plate

Side wall Partition

Sectios (B - B)

Front view (C - C)

Sidewall Water chamber Partition

Plan view

Ordinary caisson part

Side wall slit column Slit column Slit column Partition slit column

(A - A)

Fig. T- 4.1.1 Names of Vertical Slit Caisson Members

4.2 External Forces Acting on Members The following external forces acting on the members of wave-absorbing caissons shall be taken into consideration for each limit state as necessary. (1) Earth pressure (2) Earth pressure of filling (3) Residual water pressure (4) Wave force (5) Uplift pressure -267-


(6) Collision force from drifting objects (7) Weight of caisson and filling (8) Weight of superstructure (9) Surcharge (10) Reaction from fender (11) Bottom reaction (12) Loads during execution Of these, seismic effects may be ignored when calculating the earth pressure of filling, wave force, uplift pressure, reaction from fender, and external forces during execution. [Commentary] Wave forces acting on the members of slit caissons vary significantly, depending on the structure of the water chamber and whether or not it has a ceiling slab 1). Therefore, as well as referring to past cases of implementation, appropriate hydraulic model experiments are recommended in accordance with the individual conditions prior to design. [Technical Notes] (1) Load combinations and load factors are the same as described in 1.4.1 Combination of Loads and Load Factors. (2) Part Ⅱ , 5.2.8 Wave Force on Upright Wave-Absorbing Caisson should be referred to for external forces acting on structural members. (3) If the top of the water chamber is completely sealed by the ceiling slab, an impulsive pressure may be generated by the compression of the air trapped woth the top at the instant when the front of incoming wave shuts off the slits or perforations. Provision of ventilation holes with a suitable opening ratio in the ceiling slab can reduce impulsive pressure due to air compression. The opening ratio of these holes should be carefully designed. If too great, the wave surface collides directly with the ceiling slab, and this could produce a greater impulsive uplift pressure than that of no ventilation 2), 3).

Symbol

(4) Figure T- 4.2.2 3) shows an example of changes in the uplift pressure intensity pe1 when the opening ratio of ventilation holes ε1 is changed in a model experiment. Normally, the air pressure acting on the ceiling slab can be reduced by 30% to 50% compared with that on an unventilated slab by providing ventilation holes with the opening ratio of around 0.5% to 1.0%.

Ventilation opening ratio Uplift pressure when 1 = 0 Uplift pressure corresponding to

1

1

Fig. T- 4.2.2 Experimental Results on Changes in Uplift Pressure due to Opening Ratio of Ventilation Holes

(5) Table T- 4.2.1 lists the external forces taken into account when designing the members of the water chamber of wave-absorbing caisson.

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Table T- 4.2.1 External Forces for Design of Members of Water Chamber of Wave-absorbing Caisson Member

Member number

Design load

①

Partition slit column

②

Wave pressure (including wave force transmitted from partition wall)

Sidewall slit column

③

Water pressure while afloat (including wave force transmitted from sidewalls) Wave pressure ( ditto )

Upper beam

④

Vertical loads from above and below Water pressure while afloat (reaction transmitted from slit column) Wave pressure (wave force acting on the beam itself and slit column reaction)

Lower beam

⑤

Water pressure while afloat (reaction from slit column and lower wall, load acting on the beam itself) Wave pressure (ditto)

Lower wall

⑥

Water pressure while afloat Wave pressure

Sidewall

⑦

Water pressure while afloat Wave pressure

Partition

⑧

Design wave pressure acts on both sides separately in the direction parallel to centerline Fender reaction

Rear wall

⑨

Wave pressure Earth pressure, residual water pressure

Bottom slab

⑩

Bottom reaction, bottom slab weight, water head difference, and water pressure while afloat in various cases of loads.

Ceiling slab

⑪

Wave pressure (upwards, downwards) Surcharge Deadweight

Front wall

Slit column

Water pressure while afloat Wave pressure (parallel/perpendicular to centerline) Impact force from driftwood and other drifting objects Axial force transmitted from upper beam

Note: Member numbers are those shown in Fig. T- 4.1.1

4.3 Design of Members [Technical Notes] (1) The span used in calculations is the distance between the centers of support members. (2) The covering of main reinforcements should not be less than the following values in principle: Parts washed by seawater: 7 cm Other parts: 5 cm [References] 1) Shigeo TAKAHASHI, Ken-ichirou SHIMOSAKO, Hitoshi SASAKI: “Experimental study on wave forces acting on perforated wall caisson breakwaters”, Rept of PHRI, Vol. 30, No. 4, 1991 (in Japanese). 2) Sigeo TAKAHASHI, Katsutoshi TANIMOTO: “Uplift forces on a ceiling slab of wave dissipating caisson with a permeable front wall - Field data analysis -”, Rept of PHRI, Vol. 23, No. 2, 1984 (in Japanese). 3) Katsutoshi TANIMOTO, Shigeo TAKAHASHI, Tsutomu MURANAGA: “Uplift forces on a ceiling slab of wave dissipating caisson with a permeable front wall - Analytical model for compression of an enclosed air layer -”, Rept of PHRI, Vol.19., No. 1, 1980, pp. 3-31 (in Japanese).

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Chapter 5 Hybrid Caissons 5.1 General The provisions in this chapter shall be applied to the design of a hybrid caisson that is a composite structure of steel plates and concrete. The limit state design method shall be used. [Commentary] In this chapter, caissons with a composite structural type of steel plates and concrete are defined as hybrid caissons. By combining several different materials, composite structures achieve superior structural strength properties that are not possible using a single material alone. In “composite structures”, the member sections consist of a combination of different materials to achieve the functions of the structure. Hybrid caissons, like conventional steel reinforced concrete caissons, are used in breakwaters, quays, and coastal revetments. Figure C- 5.1.1 shows two types of structure members of hybrid caissons commonly used in the port and harbor structures. One is a composite member structure with steel plates arranged on one side only. The other is an SRC structure with H-shaped steel embedded inside it. In this chapter the term “hybrid caisson” is used as general term for caissons using these two structural types. Concrete

Concrete

Shear connectors (example of headed studs)

Steel reinforcement

Steel reinforcements

Steel plate

Steel frame

Fig. C- 5.1.1 Hybrid Structure Members

[Technical Notes] (1) The “Hybrid Caisson Design Manual” 1) and others may be used as reference for the design of hybrid caissons. (2) Figure T- 5.1.2 shows an example of a hybrid caisson structure.

Partition (steel stiffened plate) Steel Steel plate

reinforcement

Studs

Side wall (composite slab)

Footing (SRC slab)

Concrete

Base steel frame

Base plate (SRC slab)

Fig. T- 5.1.2 Example of a Hybrid Caisson Structure

5.2 Determination of Dimensions Dimensions of hybrid caissons shall be determined in accordance with 1.2 Determination of Dimensions. -270-


5.3 Design External Forces Design external forces shall be the same as described in 1.4 Design External Forces.

5.4 Design of Members 5.4.1 Section Force The section force to be examined in the design of members shall be the same as described in 1.5 Design of Members. 5.4.2 Design of Composite Slabs Composite slabs shall be designed in consideration of the following: (1) Bending moment (2) Shearing force (3) Integration of steel and concrete [Technical Notes] (1) Bending Moment For the bending moment, the section stress of composite slabs can be calculated as a double reinforced concrete member by converting the steel plates to equivalent reinforcements. (2) Shearing Force The shearing force of composite slabs can be analyzed in the same manner as that of reinforced concrete slabs. (3) Integration of Steel and Concrete Shear connectors are particularly important structural elements for the integration of materials in a hybrid structure. In composite slabs, headed studs and shape steel are most commonly used as the shear connectors. The required quantity and arrangement of shear connectors should be designed in consideration of preventing the steel plate separating from the concrete (especially when compressive stress is active) and securing the transmission of horizontal shear force occurring on the interface between steel plate and concrete.

5.4.3 Design of SRC Members The steel and reinforced concrete (SRC) members shall be designed against the bending moment and shearing force, by taking full account of the structural characteristics due to differences in the structural type of the steel frame. [Technical Notes] (1) SRC members can normally be classified as follows, depending on the structural type of steel frames: (a) Full-web type (b) Truss web type (2) For the bending moment, the section stress can be calculated as a reinforced concrete member by converting steel frames to equivalent reinforcements. When the fixing of steel frame ends with concrete is insufficient in full-web type, it should be calculated as a composite of the independent steel frame member and the reinforced concrete member. (3) For shearing force, if the web is of truss type, the shear stress can be calculated as a reinforced concrete by converting steel frames to equivalent reinforcements. If it is of full-web type, steel frames themselves can resist against the shearing force, and they can be duly considered in design.

5.4.4 Design of Partitions Partitions of hybrid caissons shall be designed to be sufficiently safe against external forces acting on them and to function as supporting members for sidewalls and base plates. 5.4.5 Design of Corners and Joints Corners and joints shall be designed to smoothly and firmly transmit section forces, and to be easily fabricated and executed.

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[Technical Notes] To secure sufficient strength at corners and joints, it is desirable to firmly connect the steel materials on the tensile side to those of the compressive side. It is also desirable to provide shear reinforced steel materials (haunches) against concrete tensile stress of the inside of joints.

5.4.6 Safety against Fatigue Failure Hybrid caissons shall be sufficiently safe against fatigue failure. [Technical Notes] Hybrid caissons use a large number of welded joints for connecting steel plates, and attaching shear connectors and shear resistance steel. Therefore, where the members are frequently subject to repeated load, the fatigue strength in welded parts should be examined. For coastal revetments and quaywalls, the influence from such repeated load is small. When designing breakwaters, however, the safety of the hybrid caisson against fatigue strength should be examined.

5.5 Corrosion Control Corrosion control in hybrid caissons shall be employed as appropriate in consideration of their structures and the design and execution conditions. [Technical Notes] Steel materials used on the outside of hybrid caissons are generally covered with concrete or asphalt mats. The inside is isolated from the external atmosphere by means of concrete lids. It is also in contact with filling sand in a static state and with residual seawater. Thus, when designing hybrid caissons, direct contact between the steel plates of members and the marine environment is generally avoided. For corrosion control, it is usual to set steel plate on the inside and concrete on the outside so to avoid direct contact of steel plate to fresh seawater. If steel plates are in direct contact with seawater, corrosion control should be applied such as coating methods to splash zone or tidal zone and cathodic protection methods in seawater. [References] 1) Coastal Development Institute of Technology (CDIT): “Design Manual for Hybrid Caissons”, 1999, ISBN-4-900302-44 (in Japanese). 2) Hiroshi YOKOTA: “Study on mechanical properties of steel-concrete composite structures and their applicability to marine structures”, Tech. Note of PHRI, No.750, 1993 (in Japanese).

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Part V Foundations

PART V FOUNDATIONS

Part V Foundations Chapter 1 General (1) In the design of ports and harbor facilities, appropriate types of foundations shall be selected by taking account of the importance of structures and soil conditions of the ground. (2) When the foundation ground consists of soft clayey soil, the stability and the settlement of foundation shall be carefully examined. When the foundation ground consists of loose sandy soil, the effect of liquefaction due to an earthquake shall be also studied. (3) When the bearing capacity of foundation ground is not large enough for supporting the structures, countermeasures such as pile foundation and soil improvement shall be applied as necessary. [Commentary] (1) When structures are constructed on the soft foundation ground, excessive settlement or deformation may arise owing to the lack of the bearing capacity. When the foundation ground consists of loose sandy soil, liquefaction due to an earthquake can make the structure collapse or significantly damage its functions. In such cases, the weight of the structures should be reduced or the foundation ground should be improved. (2) For stability of foundations, see Chapter 2 Bearing Capacity of Shallow Foundations, Chapter 3 Bearing Capacity of Deep Foundations, or Chapter 6 Stability of Slopes. For settlement of foundations, see Chapter 5 Settlement of Foundations. For liquefaction due to an earthquake, see Part II, Chapter 13 Liquefaction. For the design of pile foundations, see Chapter 4 Bearing Capacity of Pile Foundations. (3) For soil improvement methods, see Chapter 7 Soil Improvement Methods

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Chapter 2 Bearing Capacity of Shallow Foundations 2.1 General (Notification Article 41, Clause 1 ) When the embedded length of a foundation is less than the smallest width of the foundation, the foundation shall be analyzed in principle as a shallow foundation. [Commentary] (1) Generally, the bearing capacity of a foundation is the sum of bottom bearing capacity and side resistance. Bottom bearing capacity is evaluated with the intensity of the pressure to make the foundation bottom pushed into the ground and to cause plastic flow of soil. The side resistance of a foundation is the friction or the adhesion resistance between the sides of the foundation and the soil. In comparison with studies on the bottom bearing capacity of foundations, there are not many on the side resistance. If the embedded length of a shallow foundation is less than the smallest width of the foundation, the side resistance is much smaller than the bottom bearing capacity and may be ignored in the design. (2) When an eccentric or inclined load acts on a foundation, see 2.5 Bearing Capacity for Eccentric and Inclined Loads.

2.2 Bearing Capacity of Foundation on Sandy Ground (Notification Article 41, Clause 2 ) The following equation shall be used to calculate the allowable bearing capacity of foundation on sandy ground. In this case, the safety factor shall be given an appropriate value after due consideration of the characteristics of the structure. 1 q a = ----- ( bg 1 BN r + g 2 DN q ) + g 2 D (2.2.1) Fs where qa： Fs： b： g1： B： Nr , Nq： g 2： D：

allowable bearing capacity of foundation considering the buoyancy of underwater part (kN/m2) safety factor for the bearing capacity of sandy ground shape factor of foundation unit weight of soil below the level of foundation bottom (or submerged unit weight if submerged) (kN/m3) smallest width of foundation (m) bearing capacity factors unit weight of soil above the level of foundation bottom (or submerged unit weight if submerged) (kN/m3) embedded length of foundation (m)

[Commentary] When loads on foundations increase, settlement of foundations occurs in proportion to the loads. When the load becomes sufficiently great and reaches a certain value, settlement suddenly increases and shear failure of the ground occurs. The intensity of load required for this shear failure is called the ultimate bearing capacity. The allowable bearing capacity of a foundation is calculated by dividing the ultimate bearing capacity by a safety factor. [Technical Notes] (1) The shape factor β is given in Table T-2.2.1 for several shapes of foundations. The bearing capacity factors Nr and Nq are determined by the internal friction angle fd as shown in Fig. T-2.2.1. Table T- 2.2.1 Shape Factors Shape of foundations

Continuous

Square

Circular

Rectangular

β

0.5

0.4

0.3

0.5 ~ 0.1B/L

B ; short side length, L; long side length

-274-

Bearing capacity factor

PART V FOUNDATIONS

Internal friction angle Fig. T- 2.2.1 Relationship between Bearing Capacity Factors Nr and Nq and Internal Friction Angle φd

(2) Allowable Bearing Capacity on Sandy Ground When equation (2.2.1) is used to calculate the allowable bearing capacity of foundation on sandy ground, the safety factor should not be less than 2.5 as a general rule.

2.3 Bearing Capacity of Foundation on Clayey Ground (Notification Article 41, Clause 3) When the undrained shear strength of clayey ground increases in proportion to the depth of subsoil, it shall be standard to calculate the allowable bearing capacity of a foundation on clayey ground with equation (2.3.1). An appropriate value of safety factor shall be selected in consideration of the characteristics of the structure. B c0 (2.3.1) q a = N c0 æ 1 + n ---ö ----- + g 2 D è Lø F s where qa： Nc0： n： B： L： c0： F s： g 2：

allowable bearing capacity of foundation considering the buoyancy of underwater part (kN/m2) bearing capacity factor for foundation shape factor of foundation smallest width of foundation (m) length of foundation (m) undrained shear strength of cohesive soil at the foundation bottom (kN/m2) safety factor for the bearing capacity of clayey ground unit weight of soil above the level of foundation bottom (submerged unit weight if submerged) (kN/m3) D： embedded length of the foundation (m)

[Commentary] As the undrained shear strength of clayey ground in coastal areas usually increases linearly with depth, the bearing capacity of foundation should be calculated by the equation that takes account of the effect of shear strength increase.

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[Technical Notes] (1) Safety Factor for Bearing Capacity of Clayey Ground Whose Strength Increases with Depth. The safety factor of bearing bearing capacity should not be less than1.5 as a general rule. When a slightest settlement or deformation of ground significantly impairs the functions of superstructures such as bridge cranes, the safety factor of the bearing capacity needs to be 2.5 or greater.

Load intensity

0

0

0

Fig. T- 2.3.1 Bearing Capacity Factor Nco and Shape Factor n of Clayey Ground Having Strength Increase with Depth

(2) Practical Equation for Calculating Allowable Bearing Capacity Equation (2.3.2) can be used to calculate the allowable bearing capacity of clayey ground within a range of kB/c0 ≦ 4 in the case of continuous foundations, on the basis of the data of bearing capacity factors shown in Fig. T2.3.1, where k is the undrained shear strength increment per unit depth. 1 (2.3.2) (kB/c0 ≦ 4) q a = --- ( 1.018kB + 5.14c 0 ) + g 2 D F where the symbols are the same as those in equation (2.3.1).

2.4 Bearing Capacity of Multilayered Ground (Notification Article 41, Clause4) When a foundation ground consists of multiple soil layers, the bearing capacity shall be examined by the circular arc analysis. In this case, an appropriate value of safety factor shall be used for the design in consideration of the characteristics of the ground and structure. [Technical Notes] The bearing capacity of the ground consisting of multiple soil layers is generally calculated by the circular arc analysis, with the condition that the overburden pressure above the level of foundation bottom is taken as surcharge load as shown in Fig. T- 2.4.1. The safety factor for the circular arc passing through the bottom edge of the foundation is calculated with the circular arc analysis of the modified Fellenius method. The safety factor should not be less than1.5 as a general rule. The safety factor should be increased to 2.5 in such cases as crane foundations, where excessive settlement could significantly damage the functions of the superstructures. If the clayey layer thickness H is significantly less than the smallest width of the foundation B (i.e., H < 0.5B), a punching shear failure, in which the clayey layer is squeezed out between the surcharge plane and the bottom of clayey layer, is liable to occur. The bearing capacity against this kind of squeezed-out failure can be calculated by the following equation 1):

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B

Fig. T- 2.4.1 Circular Arc Analysis to Calculate the Bearing Capacity of Multilayered Ground

PART V FOUNDATIONS

cu B- ) ---- +g2D (2.4.1) qa = (4.0 + 0.5 --H F where qa： allowable bearing capacity of foundation considering the buoyancy of the underwater part (kN/m2) B： smallest width of foundation (m) H： thickness of clay layer (m) cu： mean undrained shear strength in layer of thickness H (kN/m2) g 2： unit weight of soil above the level of foundation bottom (or submerged unit weight if submerged) (kN/m3) F： safety factor D： embedded length of foundation (m)

2.5 Bearing Capacity for Eccentric and Inclined Loads (Notification Article 41, Clause 5) The circular arc analysis is used as a standard method when examining the bearing capacity for eccentric and inclined loads acting on a foundation of gravity type structures. The circular arc analysis based on the simplified Bishop method shall be used in this case. The safety factor calculated by the following equation shall have an appropriate value in view of the characteristics of the structure. The strength constants of the ground as well as action formats of external forces and loads shall be determined as appropriate in view of the structural characteristics of facilities. 1 ( cb + W ¢ tan f ) sec a (2.5.1) F s = ------------------------------------------------ S ---------------------------------------------1 + ( tan a tan f ) ¤ F s 1ö æ S W sin a + è ---ø S Ha R where Fs： safety factor against circular failure according to the simplified Bishop method W： total weight of slice element per unit length (kN/m) a： angle at which the bottom of a slice element intersects the horizontal plane (º) R： radius of circular arc slip circle (m) H： horizontal external force acting on soil inside the circular slip surface (kN/m) a： arm length from the center of circular slip surface to the acting position of horizontal external force H (m) c： undrained shear strength in clayey ground, or apparent cohesion under drained condition in sandy ground (kN/m2) b： width of a slice element W¢： effective weight of a slice element per unit length (sum of soil weight and surcharge) (the submerged unit weight when submerged) (kN/m) f： internal friction angle under drained condition for sandy ground (º); the value is 0º for clayey ground [Commentary] Gravity type quaywalls and breakwaters are subject to external forces such as the deadweight, earth pressure, seismic force, and wave force. The combined result of these forces usually yields an eccentric and inclined load. Therefore, for calculating the bearing capacity of foundations, the effects of eccentric and inclined loads should also be taken into consideration. The eccentric and inclined loads mean that the load inclination ratio is equal to or greater than 0.1. Normal gravity type structures are supported with a two-layered system such that a rubble mound layer is set on the foundation ground. Thus the method of calculating the bearing capacity should fully reflect the characteristics of two-layer system. It is known that the circular arc calculation based on the simplified Bishop method can well evaluate the bearing capacity of this type of foundation. This has been proved by a series of researches including laboratory model experiments, in situ prototype experiments, and case studies on existing breakwaters and quaywalls 2). [Technical Notes] (1) Analysis of Bearing Capacity by Circular Arc Calculation Based on the Simplified Bishop Method Analysis through circular arc calculation based on Bishop method is more precise than the analysis based on the modified Fellenius method, except when a vertical load acts on horizontally layered sandy ground. Therefore, this method is applied under the condition that eccentric and inclined loads act. As shown in Fig. T- 2.5.1 (a), the start point of the slip surface is set symmetrical about the acting point of resultant load to one of the foundation edges that is closer to the load acting point. In this case, the vertical load acting on the rubble mound is converted into a uniformly distributed load acting on the width 2b¢ as specified in Fig. T- 2.5.1 (b) and (c). The horizontal force is assumed to act at the bottom of structure. When calculating the bearing capacity during an earthquake, seismic force is assumed not to act on the rubble mound and the ground. -277-


(2) Safety Factor The safety factor is expressed as the ratio of the resisting moment due to shear resistance to the sliding moment due to external forces and soil weight, as in the other cases of circular arc analyses. See [Technical Note] (2) in 6.2.1 Stability Analysis Using Circular Slip Surface Method. Standard values of the safety factor are listed in Table T- 2.5.1 Table T- 2.5.1 Safety Factors for Bearing Capacity for Eccentric and Inclined Loads (Simplified Bishop’s Circular Arc Method) Quaywalls

Breakwaters

Ordinary condition

1.2 or greater

-


1.0 or greater

-

Under storm waves

-

1.0 or greater

When subgrade reaction has a trapezoidal distribution When subgrade reaction has a triangular distribution

Combined force of load

Rubble mound Subsoil

Fig. T- 2.5.1 Analysis of Bearing Capacity for Eccentric and Inclined Loads

(3) Strength Parameters for Mound Materials and Foundation Ground (a) Mound materials Model and field experiments on bearing capacity Experimental values subject to eccentric and inclined loads have verified c 20 that high precision results can be obtained by conducting circular arc analyses based on the simplified Bishop method, applying the strength parameters obtained by triaxial compression tests 2). Large-scale triaxial compression test results of crushed stone have confirmed that the strength parameters of large diameter particles are approximately equal to those d0 obtained from similar gradation materials with the same coefficient of uniformity 3). Therefore, triaxial compression tests using samples with similar gradation should be conducted in order to estimate the strength parameters of rubbles accurately. If strength tests are not conducted, the values of apparant cohesion cd = 20 kN/m2 and internal friction angle fd = 35º are applied as the standard strength parameters for rubbles generally used in port construction

Lateral pressure (kN/m2) works. The above standard values have been obtained Fig. T- 2.5.2 Relationship between fd0 and Lateral on the safe side consideration based on the results of Confining Pressure s3 large-scale triaxial compression tests of crushed stones and the bearing capacity analysis of existing breakwaters and quaywalls. It should be noted that the strength parameter for cohesion cd = 20 kN/m2 is the apparent cohesion, taking account of changes of the internal friction angle fd of crushed stones due to the confining pressure. Figure T- 2.5.2 shows the result of triaxial compression tests on various types of crushed stones and rubbles 2). It shows that as the confining pressure increases, fd0 decreases due to particle crushing. The continuous solid line in the figure represents the value under the assumption that the apparent cohesion is cd = 20 kN/m2 and the internal friction angle is fd = 35º. Here, the dependency of fd0 on the confining pressure d

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PART V FOUNDATIONS

is well described by taking the apparent cohesion into account. These standard values can be applied only to the stone material with an unconfined compressive strength in the parent rock of 30 MN/m2 or more, according to the past studies. If weak stones with a compressive strength of less than 30 MN/m2 are used as a part of the mound, the strength parameters will be around cd = 20 kN/m2 and fd = 30º 4). (b) Foundation ground Foundations subject to eccentric and inclined loads often cause shallow surface slip failure. In these cases, it is important to evaluate the strength near the surface of foundation ground. If the foundation ground is sandy, the friction angle fd is usually estimated from the N-value. The estimation formulas employed up to now have tended to underestimate fd in case of shallow sandy grounds. This is because no correction has been made regarding the effective surcharge pressure in situ. Figure T- 2.5.3 collates the results of triaxial compression tests on undisturbed sand in Japan (dotted points) and presents a comparative study of the formulas proposed in the past. Even the N-values are less than 10, internal friction angles of around 40º have been obtained. In many cases, the bearing capacity for eccentric and inclined loads is problematic for design not under normal conditions but under dynamic external forces such as wave pressure and seismic force. Based on the results of bearing capacity analysis of the structures damaged in the past, the values given below are applied as the standard values of fd in foundation ground. Sandy ground with N-value of less than 10: Sandy ground with N-value of 10 or more:

fd = 40º fd = 45º

If the ground consists of cohesive soil, the strength may be determined by the method indicated in Part Ⅱ, 11.3.3 Shear Properties.

Range according to Meyerhof

Triaxial test results

φ =

20N + 15 according to Osaki

N-value Fig. T- 2.5.3 Relationship between N-value and φd Obtained by Triaxial Tests on Undisturbed Sand Samples

[References] 1) Hakujyu YAMAGUCHI: “Soil Merchanics (New Editon)”, Chapter 9 Bearing Capacity, GIHOUDOU, 1985, pp. 273-274 (in Japanese). 2) Masaki KOBAYASHI, Masaaki TERASHI, Kunio TAKAHASHI, Kenjirou NAKASHIMA, Hiraku ODANI: “A new method for calculating the bearing capacity of rubble mounds”, Rept. of PHRI, Vol. 26, No. 2, 1987 (in Japanese). 3) Yoshihiro SHOJI: “Study on shearing properties of rubble with large scale triaxial compression test”. Rept. of PHRI, Vol. 22, No.4, 1983 (in Japanese). 4) Jun-ichi MIZUKAMI, Masaki KOBAYASHI. “Strength characteristics of rubble by large scale triaxial compressin test”, Tech. Note of PHRI, No. 699, 1991 (in Japanese).

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Chapter 3 Bearing Capacity of Deep Foundations 3.1 General (Notification Article 42) When the embedded length of a foundation is larger than the smallest width of the foundation, the foundation shall be examined in principle as a deep foundation. In such a case, the bearing capacity of the deep foundation shall be calculated by an appropriate method in view of the characteristics of both the ground and the structure.

3.2 Vertical Bearing Capacity The allowable vertical bearing capacity of a deep foundation shall be determined as appropriate in view of the structural type, methods of construction, and ground conditions. [Commentary] (1) Allowable Vertical Bearing Capacity of Deep Foundations Generally, the allowable vertical bearing capacity of a deep foundations is calculated as the sum of the allowable bearing capacity due to the side resistance of foundation and the allowable bearing capacity of foundation bottom, as expressed by equation (3.2.1). However, if the amount of structural deformation is to be examined, the deformation of deep foundation should be estimated by assuming the ground having characteristics of spring. (3.2.1) q a = q a1 + Dq a where qa: allowable bearing capacity of deep foundation (kN/m2) qa1: allowable bearing capacity of foundation bottom (kN/m2) (see 2.2 Bearing Capacity of Foundation on Sandy Ground, 2.3 Bearing Capacity of Foundation on Clayey Ground) Dqa: increment of allowable bearing capacity due to resistance of foundation sides (kN/m2) (2) Side Resistance of Deep Foundations Caution is required concerning evaluation of the side resistance of deep foundations. This is because, depending on structural type and method of construction, the surrounding ground may be disturbed during construction and the adequate amount of bearing capacity due to side resistance may not always be obtained. [Technical Notes] (1) Frictional Resistance at the Sides of Foundations in Sandy Ground Equation (3.2.2) is used to calculate the increment of allowable bearing capacity due to the frictional resistance at foundation sides in sandy ground. 1 B D2 (3.2.2) Dq a = --- æ 1 + ---ö ------K a g 2 m Fè Lø B where F: safety factor (the same value as used for qa1) Ka: coefficient of active earth pressure (d = 0º) (see Part Ⅱ, Chapter 14 Earth Pressure and Water Pressure) g 2 : unit weight of soil above the level of foundation bottom (or submerged unit weight if submerged) (kN/ m3) D: embedded length of foundation (m) m: coefficient of friction between foundation sides and sandy soil, m = tan(2/3)f B: width of foundation (m) L: length of foundation (m) In equation (3.2.2), the increment of allowable bearing capacity Dqa is obtained by dividing the total friction resistance by the bottom area of foundation and the safety factor. The total friction resistance is calculated as the product of the mean side friction strength f multiplying with the embedded length D and the total contact surface area between the sandy soil and foundation sides. Equation (3.2.3) is generally used to calculate the mean side friction strength f corresponding to the embedded length D. 1 1 D (3.2.3) f = ---- ò0 g zK a m dz = --- K a gDm 2 D The friction angle between the foundation sides and sandy soil should not be greater than the internal friction angle of soil f, and it may be taken as (2/3)f for the case between concrete and sandy soil. (2) Cohesive Resistance of Foundation Sides in Clayey Ground Equation (3.2.4) may be used to calculate the increment of allowable bearing capacity due to the cohesive resistance at foundation sides in clayey ground. -280-

PART V FOUNDATIONS

2 B Dc Dq a = --- æ 1 + ---ö ------ c a è F Lø B where c a : mean adhesion (mean value in embedded part below the ground water level) (kN/m2) Dc: embedded length of foundation below the groundwater level (m)

(3.2.4)

In case of deep foundations in clayey ground, the soil above the groundwater level is subject to dry shrinkage in summer. This means that the soil is unsuitable to be regarded as effective contact surface. Therefore, the mean adhesion c a in equation (3.2.4) refers to the mean value in the embedded part below the ground water level. See Table T- 3.2.1 for the practical values of mean adhesion in clayey soil. Table T- 3.2.1 Mean Adhesion Types of ground

(Units: kN/m2)

ca

qu

Soft cohesive soil

25 ~ 50

*

Medium cohesive soil

50 ~ 100

6 ~ 12

Hard cohesive soil

100 ~ 200

12 ~ 25

Very hard cohesive soil

200 ~ 400

25 ~ 30

400 or greater

30 or greater

Compacted cohesive soil

* : With soft cohesive soil, side resistance should not be taken into account. (3) Safety Factor The safety factor to be applied in using equations (3.2.2) and (3.2.4) should be 2.5 or greater for important structures and 1.5 or greater for others.

3.3 Lateral Bearing Capacity The lateral bearing capacity of a deep foundation shall be determined as appropriate in view of ground conditions, structural characteristics, and the method of construction. [Commentary] The lateral bearing capacity of a deep foundation is governed by the horizontal subgrade reaction of the foundation sides and the vertical subgrade reaction at the the bottom of foundation. [Technical Notes] (1) When a resultant force at a bottom of fandation acts inside the core (when the eccentricity of total resultant force acting at the bottom of foundation is within one-sixth of the foundation width from the central axis of the foundation) and the distributions of horizontal and vertical subgrade reaction are assumed as in Fig. T- 3.3.1, the maximum horizontal subgrade reaction p1 and maximum vertical subgrade reaction q1 can be estimated by assigning an adequate value of safety factor against the passive earth pressure and the ultimate bearing capacity in their respective positions. (2) Assumption on the Distribution of Subgrade Reaction The distribution of horizontal subgrade reaction shown in Fig. T- 3.3.1 may be assumed as being a quadratic parabola with 0 at the ground surface. This assumption is equivalent to the relationship between the displacement y and the subgrade reaction p of equation (3.3.1) when the foundation rotates as a rigid body. (3.3.1) p = Kxy where p： subgrade reaction (kN/m2) K： rate of increase in coefficient of horizontal subgrade reaction with depth (kN/m4) x： depth (m) y： horizontal displacement at depth x (m)

N0 P0 M0

h p1

kW

l

W q2

q1 2a

Fig. T- 3.3.1 When Resultant Force Is inside the Core

A linear distribution has been assumed for vertical subgrade reaction. Therefore, when a resultant force acting at the bottom of foundation is inside the core, the distribution becomes trapezoidal as shown in Fig. T- 3.3.1. -281-


(3) When Resultant Force at the Bottom Is inside the Core The condition under which the resultant force at the bottom is inside the core is expressed as in equation (3.3.2). N 0 + w 1 l 3aK¢ ( kw 1 l 2 + 4P 0 l + 6M 0 ) ---------------------- ≧ -------------------------------------------------------------------(3.3.2) A b ( l 3 + 24 a K¢a 3 ) The maximum horizontal subgrade reaction p1 (kN/m2) and the maximum vertical subgrade reaction q1 (kN/m2) in this case are obtained by equations (3.3.3) and (3.3.4), respectively. 3 [ kw 1 l 4 + 3P 0 l 3 + 4M 0 l 2 + 8 a K¢a 3 ( kw 1 l + P 0 ) ] 2 (3.3.3) p 1 = ---------------------------------------------------------------------------------------------------------------------------4bl 3 ( l 3 + 24aK¢a 3 ) ( kw 1 l 2 + 4P 0 l + 6M 0 ) N 0 + w 1 l 3aK¢ ( kw 1 l 2 + 4P 0 l + 6M 0 ) (3.3.4) q 1 = ---------------------- + -------------------------------------------------------------------A b ( l 3 + 24aK¢a 3 ) When examining the lateral bearing capacity of deep foundations, the values of p1 and q1 obtained by equations (3.3.3) and (3.3.4) must satisfy equations (3.3.5) and (3.3.6), respectively. 1 (3.3.5) p 1 ≦ --- pp F q1 ≦ q a (3.3.6) where l： 2b： 2a： A： P 0： M 0： N 0： k： K¢： K 1： K 2：

embedded length (m) maximum width (perpendicular to horizontal force) (m) maximum length (m) bottom area (m2) horizontal force acting on the structure above ground level (kN) moment due to P0 at ground level (kN m) vertical force acting at ground level (kN) horizontal seismic coefficient coefficient ratio; K¢= K2 /K1 rate of increase in the coefficient of vertical subgrade reaction (kN/m4) rate of increase by depth in the confficient of horizontal subgrade reaction (kN/m4 ) (see equation (3.3.1)) w1： deadweight of deep foundation per unit depth (kN/m) a： coefficient determined by bottom shape (a = 1.0 for rectangular shape and a = 0.588 for circular shape) pp： maximum passive earth pressure at the depth h (kN/m2) (see Part Ⅱ, Cahpter 14 Earth Pressure and Water Pressure). h： depth at which horizontal subgrade reaction becomes maximum (m) (see equation (3.3.7)) qa： vertical bearing capacity at bottom level (kN/m2) (see equation (3.2.1)) F： safety factor for horizontal bearing capacity kw 1 l 4 + 3P 0 l 3 + 4M 0 l 2 + 8aK¢a 3 ( kw 1 l + P 0 ) h = -----------------------------------------------------------------------------------------------------------------(3.3.7) 2l ( kw 1 l 2 + 4P 0 l + 6M 0 ) ·

(4) When Resulant Force at the Bottom Is outside the Core When the resultant force acting at the base of foundation is not inside the core, a triangular distribution of vertical subgrade reaction is assumed as shown in Fig. T- 3.3.2 1). It is because the tensile resistance cannot take place between the of foundation bottom and the ground. In this case, the maximum subgrade reaction p1 (kN/m2) in the frontal side is obtained from equation (3.3.8) 3 ( kWl + 4M 0 4N 0 e 4We + 3P 0 l ) 2 (3.3.8) p 1 = --------------------------------------------------------------------------------------------------4bl 2 ( kWl + 6M 0 6N 0 e 6We + 4P 0 l ) The value of p1 calculated by equation (3.3.8) should satisfy equation (3.3.5). In this case, h is obtained by equation (3.3.9). l ( kWl + 4M 0 4N 0 e 4We + 3P 0 l ) (3.3.9) h = ------------------------------------------------------------------------------------------2 ( kWl + 6M 0 6N 0 e 6We + 4P 0 l ) where h： depth at which horizontal subgrade reaction becomes maximum (m) (see Fig. T- 3.3.2) W： deadweight of foundation (kN) e： eccentric distance (m) The distance e is defined as shown in Fig. T- 3.3.2. When the foundation bottom is rectangular with the length of 2a (m) and the width of 2b (m), the value of e is calculated by equation (3.3.10). -282-

Fig. T- 3.3.2 When Resultant Force Is Not inside the Core

PART V FOUNDATIONS

64748

W + N0 (3.3.10) e = a ----------------4bq a In the case of a circular foundation bottom, the calculation may be made by replacing it with a rectangular foundation bottom having length 2a and width 2b defined by equation (3.3.11). p 2a = --- D 3 (3.3.11) 3 2b = --- D 4 where D: diameter of circle (m) In this way, the horizontal bearing capacity can be estimated at a safer side by approximately 10%. However, this substitution should be applied on the basis of the appropriate judgement, by referring to reference 1). (5) Safety Factor When applying the above caluculation methods, the safety factor should generally be 1.5 or greater for important structures and 1.1 for others. [Reference] 1) Kunio TAKAHASHI, Masatoshi SAWAGUCHI: “Experimental study on the lateral resistance of a well” Rept. of PHRI, Vol. 16, No. 4, 1977, pp.3-34 (in Japanese).

-283-


Chapter 4 Bearing Capacity of Pile Foundations 4.1 Allowable Axial Bearing Capacity of Piles 4.1.1 General (Notification Article 43, Clause 1) The allowable axial bearing capacity of piles shall be determined on the basis of the ultimate axial bearing capacity of single piles divided by the safety factor as the standard value. Furthermore it shall be appropriately determined by taking the following factors into consideration to the extent as necessary. The safety factor shall be determined as appropriate in view of the characteristics of the structures and the ground. (1) Allowable compressive stress of pile material (2) Decrease of allowable stress at pile joints (3) Decrease of allowable stress due to slenderness ratio of piles (4) Effect of piles group (5) Negative skin friction of piles (6) Settlement of piles [Commentary] (1) The above provision presents a guideline for judging the axial bearing capacity of piles in pile foundations. First, the ultimate axial bearing capacity of single piles is determined, and then it is divided by the safety factor to obtain the standard allowable axial bearing capacity. Next, the factors (1) to (6) listed above are examined and the standard allowable axial bearing capacity should be lowered if necessary. The final result is the allowable axial bearing capacity of piles to be used in the design of pile foundations. (2) The bearing capacity of piles is greatly affected by the methods and execution of construction. Therefore, before the start of construction execution, test piles should be driven and the design should be confirmed through various studies. The pile length and construction method should be modified to reflect the test results. (3) Piles group means the group of piles in which the bearing capacity and deformation of each pile are reciprocally affected by others.

4.1.2 Standard Allowable Axial Bearing Capacity The standard allowable axial bearing capacity shall be calculated by dividing the ultimate axial bearing capacity of single piles by an appropriate value of safety factor. [Commentary] The safety factor used when calculating the standard allowable axial bearing capacity from the ultimate axial bearing capacity takes account of deviations in ground conditions, pile conditions, loading conditions, etc. The safety factor should have the value that guarantees the safety of piles against ground failure. When calculating the ultimate axial bearing load of single piles by loading tests and static bearing capacity formulas, the standard allowable axial bearing capacity should be the value obtained from the ultimate axial bearing capacity of single piles being divided by the safety factor. [Technical Notes] (1) Table T- 4.1.1 lists the guideline for the minimum values of safety factors. The values are based on the following principles: (a) The minimum value of 2.5 for the safety factor under ordinary conditions means that the safety factor is 2.5 against the phenomena of ultimate ground failure. The safety factor against the ground yield phenomena will be 1.5 to 2.0. (b) The minimum safety factor during an earthquake may be set lower than that under ordinary conditions. This is because the duration of earthquakes is comparatively short and the strength of soil against impact load increases generally. (2) Further examinations would be required if permanent deformation of the ground is expected to remain after an earthquake. Regarding the dynamic properties of soil, many unknown matters remain. Therefore, it may be doubtful if a high strength of soil during an earthquake can be employed for foundation design. For example, sensitive clay may lose its strength due to violent shaking. In the past cases of earthquake damage, it is known that liquefaction had occurred in loose sand strata, and this greatly reduced the bearing capacity of piles. Friction piles are easily affected by such phenomena. Therefore, the safety factor of friction piles during an earthquake should be given a greater value than that of bearing piles. -284-

PART V FOUNDATIONS

(3) If the values of safety factor in Table T- 4.1.1 are applied, ordinary port and harbor structures under ordinary design conditions should be safe. For the structures that are particularly important or are critical in terms of human life, however, larger values of the safety factor may have to be applied. On the other hand, it may be possible to apply the value of safety factor lower than the minimum values on the basis of adequate research and prudent judgement. An example is the case that detailed soil surveys and loading tests are conducted, and the pile behavior estimated on this basis is consistent with detailed observation results on actual structures of the same type in the vicinity. In such cases, a lower value of safety factor is permissible, provided that the judgement by experts in the field of geotechnical engineering supports it. Table T- 4.1.1 Guidelines for Minimum Values of Safety Factor Ordinary condition During an earthquake

2.5

Bearing piles

1.5

Friction piles

2.0

4.1.3 Ultimate Axial Bearing Capacity of Single Piles It is desirable to obtain the ultimate axial bearing capacity of single piles by conducting loading tests in situ. If it is difficult to conduct the loading tests, the ultimate axial bearing capacity of single piles may be estimated by static bearing capacity formulas. [Commentary] Estimating methods of the ultimate axial bearing capacity of single piles are as follows: (1) Estimation by loading tests (2) Estimation by static bearing capacity formulas (3) Estimation by existing data It is desirable to estimate the ultimate axial bearing capacity of single piles by conducting axial loading tests in situ. This is the most rational method to estimate the allowable bearing capacity in the design. Sometimes it may be difficult to conduct loading tests prior to design, due to limit of construction period and/or economical constraint. In such cases, estimation by static bearing capacity formulas is permissible. Even when estimating the ultimate axial bearing capacity by the methods other than loading tests and deciding the allowable bearing capacity on this basis, loading tests should be conducted at an early stage of construction and the appropriateness of the allowable bearing capacity adopted in the design should be confirmed.

4.1.4 Estimation of Ultimate Axial Bearing Capacity by Loading Tests When the ultimate load can be confirmed by the load - settlement curve, the value shall be taken as the ultimate axial bearing capacity. When the ultimate load cannot be confirmed from the load - settlement curve, the yield load shall be confirmed and the ultimate load may be estimated from the yield load. [Commentary]

Load

Pile loading tests are a kind of full scale experiments that can obtain the ultimate bearing capacity directly. However, in loading tests, the loading conditions are different from the actual conditions in respect of the number of piles and the duration of the loading.

(1) Yield Load and Ultimate Load Relationship between the load and the total settlement by a static loading test is indicated schematically in Fig. 4.1.1. The yield load is indicated by P1 at the point A, where the initial gentle inclination becomes clearly steeper and the curve shows the maximum curvature. The ultimate load is indicated by P2 at the point where the load - settlement curve becomes vertical. Piles embedded in the sandlayer do not usually reach such the load. In such cases, the ultimate load may be taken as P3 at the point B. Here, the load settlement curve suddenly turns downward and a small increase of load produces a large amount of settlement.

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Total settlement

[Technical Notes]

Fig. T- 4.1.1 Yield Load and Ultimate Load


(2) Ultimate Load It is not always possible to confirm the ultimate load by loading tests due to the constraint of capacity of test equipment. In such cases the ultimate load is calculated as 1.2 times the yield load obtained from loading tests. This judgement is based on the results of research by Yamakata and Nagai on steel pipe piles and statistical studies by Kitajima et al. 1) When yielding does not occur during loading tests, the ultimate load is taken as 1.2 times the maximum test load. In either case, it is necessary to have a reasonable assurance that the actual ultimate load will be larger than the ultimate load estimated in this way.

4.1.5 Estimation of Ultimate Axial Bearing Capacity by Static Bearing Capacity Formulas When estimating the ultimate axial bearing capacity by static bearing capacity formulas, attention shall be paid to the ground and pile conditions, construction methods, and the application limitation of formulas. [Technical Notes] (1) When using bearing capacity formulas, particular attention should be paid to differences in the construction methods. (a) Pile driven by hammer ① Equation (4.1.1) may be used when estimating the ultimate bearing capacity of the piles driven into sandy ground by hammer. R u = 300NA p + 2NAs where Ru： ultimate bearing capacity of pile (kN) Ap： toe area of pile (m2) As： total circumferential area of pile (m2) N： N-value of the ground around pile toe N： mean N-value for total penetration length of pile

(4.1.1)

The N-value is calculated by equation (4.1.2) N 1 + N2 N = ------------------2 where N1： N-value at the toe of pile N2： mean N-value in the range from the toe of pile to the level 4B above B： diameter or width of pile (m)

(4.1.2)

Caution is required here for estimating the ultimate bearing capacity of piles supported by the ground with an N-value of 50 or more, since the N-values larger than 50 may not be reliable. Furthermore, it remains to be confirmed whether the first term in the right-hand side of equation (4.1.1) can be applied directly to this kind of hard ground. ② Equation (4.1.3) may be used when estimating the ultimate bearing capacity of piles driven into clayey ground by hammer.

R u = 8c p A p + ca A s where cp： ca：

(4.1.3)

cohesion at pile toe (kN/m2) mean adhesion for total embedded length of pile (kN/m2)

678

The adhesion value may be calculated as follows: : c £ 100 kN ¤ m 2 ca = c

(4.1.4) c a = 100 kN ¤ m 2 : c > 100 kN ¤ m 2 Here, due attention should be paid to the ground characteristics and the pile conditions, and the value of obtainable adhesion should be examined. This is because there are theoretical problems2) in calculating pile adhesion from the ground cohesion c or the unconfined compressive strength. (b) The vibratory pile driving method (vibrohammer method) is increasingly being used for driving piles because of the capacity increase of pile-driving machinery in recent years. As the principles of this method differ from those of pile driving by hammer, the bearing capacity should be carefully estimated. When using this method, the ground should be compacted by the method of hammer pile driving instead of vibratory pile driving in the course of final driving, or vertical loading tests should be conducted to confirm the characteristics of bearing capacity of the ground in question.

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PART V FOUNDATIONS

(c) In recent years, the use of pile installation method not involving pile driving by hammer has been increasing in port and harbor construction projects; an example is pile installation by inner excavation. In such cases, the characteristics of the bearing capacity of piles in question should be confirmed by vertical loading tests. (2) Even if there is no shoe on the pile toe, the end bearing area of steel piles can be considered closed, as shown by the shaded areas in Fig. T- 4.1.2. In this case, the outer edge alone of the closed area is taken as the perimeter. This is based on the following principle. Soil enters the interior of steel pipes or the space between the flanges of H-shaped steel during the pile driving, until the internal friction between the soil and the surface of steel pile becomes equal to the end resistance of pile. This balance prevents soil from entering and has the same effect as the case when the open end section is closed. But complete closure cannot be expected in the case of large-diameter piles. In such cases the closed area ratio should be examined.

Fig. T- 4.1.2 End Bearing Area of Steel Piles

(3) Closed Area Ratio The closed area ratio is not determined simply by the diameter or the width of piles. It also depends on the penetration length of piles, properties of the ground, and the method of installation. Past data suggest the closed area ratio of 100% when the diameter of steel pipe piles is less than 60 cm or the short side of H-shaped steel piles is less than 40 cm. For piles with a larger diameter or width, there have been several theoretical calculation methods and laboratory experiment results. The closed area ratio varies greatly depending on the properties of the ground and the method of pile installation. The state of plugging of actual piles depends on the penetration depth and associated ground stress. It is thus difficult to obtain the ratio by theoretical calculation. The Steel Pipe Pile Association collated measured examples of the closed area ratio, as shown in Fig. T- 4.1.3. Here, piles with a diameter of 1117.6 mm or less are shown and the closed area ratio is estimated by assuming that the first term of equation (4.1.1) indicates the end bearing capacity in the case of complete plugging. According to this figure, the closed area ratio ranges from 30% to 140%. This means that the ratio cannot be judged only by the diameter of steel pipes or the embedded length ratio to the bearing stratum, or the soil classification of the bearing stratum. In recent years, steel pipe piles of 1,100 mm diameter or more have been commonly used. However, any data for the closed area ratio of these pipes have not been reported and therefore the closed area ratio for the pipe piles mentioned above are unknown. Since the closed area ratio is affected by the installation and ground conditions in actual sites, it should first be ascertained through loading tests. (4) Bearing Capacity of Soft Rock When piles are supported on soft rock or hard clay, the bearing capacity may be calculated by equation (4.1.1). If unconfined compressive strength qu (kN/m2) has been measured by undisturbed samples, equation (4.1.5) may alternatively be used. (4.1.5) R u = 5q u A p Further, the value of qu should be reduced to 1/2 or 1/3 of the measurement depending on the progress of cracking in natural ground. In any event, however, the value of qu should not exceed 2´10 4 kN/m2.

Ratio of pile end load in loading tests to 300·N·Ap

Outer diameter 700 mm or more Outer diameter 650 mm or less

Ratio of embedded length to bearing stratum to inner diameter of pile

Fig. T- 4.1.3 Plug Effect of Open-ended Pile -287-


4.1.6 Examination of Compressive Stress of Pile Materials (Notification Article 43, Clause 2) The allowable axial bearing capacity of piles shall not exceed the value obtained by multiplying the effective section area of piles by the allowable compressive stress of pile materials, which is mentioned in 4.1.1 General. 4.1.7 Decrease of Bearing Capacity Due to Joints (1) If it is necessary to splice piles, the splicing work shall be executed under appropriate supervision and reliability of joints of spliced pile shall be confirmed by appropriate inspection. (2) If joints are sufficiently reliable, it may not be necessary to decrease the axial allowable bearing capacity due to existence of joints. [Technical Notes] (1) When spliced piles are used, the joints are the weak points of the piles. Therefore, the allowable axial bearing capacity should be decreased in consideration of the effect of the joints against the overall bearing capacity of the pile foundation. (2) On-site circular welding by semi-automatic methods is generally employed for the splicing of steel pipe piles used in the field of port and harbor construction works. When such highly reliable jointing methods are applied under appropriate supervision and the reliability of the joints has been confirmed by inspection, it is not necessary to decrease the allowable axial bearing capacity. (3) Provisions for the joints can be found in 4.5.3 Joints.

4.1.8 Decrease of Bearing Capacity Due to Slenderness Ratio For piles with a very large ratio of length to diameter, the allowable axial bearing capacity of piles shall be decreased in consideration of the accuracy of installation, unless the safety of bearing capacity confirmed by loading tests. [Technical Notes] (1) This provision takes account of the fact that the inclination of piles during installation reduces their bearing capacity. If loading tests are conducted on foundation piles, the ultimate bearing capacity can be determined, including the necessity of decrease of bearing capacity due to installation accuracy. Therefore, in this case the decrease due to the slenderness ratio may not necessarily be taken into account. (2) When decreasing the bearing capacity due to the slenderness of piles, the following values may be used as references: (a) For non-steel piles ì ï ï a = í ï ï î

0

:

l --- 60 : d

l --- £ 60 d l --- > 60 d

(4.1.6)

(b) For steel piles ì l ï 0 : --- £ 120 ï d a = í l l ï ----: - > 120 ï 2d- 60 -d î where a： rate of reduction (%) l： pile length (m) d： pile diameter (m)

(4.1.7)

4.1.9 Bearing Capacity of Pile Group When a group of piles are examined as a pile group, the bearing capacity of pile group may be studied as a single, deep foundation formed with the envelope surface surrounding the outermost piles in the group of piles. -288-

PART V FOUNDATIONS

[Technical Notes] Terzaghi and Peck state that a failure of a pile group foundation does not mean the failure of individual piles but the failure as a single block 3), 4). This is based on the principle that the soil and piles inside the shaded area in Fig. T4.1.4 work as a single unit when the intervals between the piles are small. The ultimate bearing capacity of pile group in this case is expressed by equation (4.1.8). (4.1.8) R g u = q d A g + sUL where Rgu： bearing capacity of pile group as a single block (kN) qd： ultimate bearing capacity at the block bottom (kN/m2) (see 2.2 Bearing Capacity of Foundation on Sandy Ground, 2.3 Bearing Capacity of Foundation on Clayey Ground) Ag： bottom area of pile group block (m2) U： perimeter of pile group block (m2) L： embedded length of piles (m) s： mean shear strength of soil contiguous with piles (kN/m2) The allowable bearing capacity per pile is expressed by equation (4.1.9). ì1 ü1 (4.1.9) R a = í --- ( R g u g 2¢ A g L ) ý --F î þn where Ra： allowable bearing capacity per pile against failure as a block (kN) g2'： mean unit weight of the whole block including piles and soil (kN/m3) (mean unit weight is calculated using submerged unit weight below the groundwater level (taking account of buoyancy) and using the wet unit weight above ground water level.) n： number of piles in pile group F： safety factor (see 4.1.2 Standard Allowable Axial Bearing Capacity) In the case of cohesive soil, equation (4.1.9) is replaced by equation (4.1.10), where c is cohesion and g2' ≒ g2 (g2： mean unit weight of soil above the pile tip level). ü B 1 ì R a = ------- í 5.7cA g æ 1 + 0.3 ------ö + cUL g 2 A gL ý è B 1ø nF î þ where B： short side length of pile group block (m) B1： long side length of pile group block (m)

(4.1.10)

The allowable bearing capacity of each pile when used in a pile group is the smaller of the allowable bearing capacity of single piles or the allowable bearing capacity against block failure given by equation (4.1.9) or (4.1.10). The upper limit of the interval between rows of piles to which the above assumptions apply cannot be uniformly defined. This is because it varies depending on the properties of the ground and the arrangement of piles.

Negative skin friction

Weak layer

Perimeter U

Positive skin friction

Consolidation settlement

L

Bearing stratum

Fig. T- 4.1.4 Pile Group Foundation

(a)

(b)

Fig. C- 4.1.1 Negative Skin Friction -289-


4.1.10 Examination of Negative Skin Friction If bearing piles penetrate through a soil layer that is susceptible to consolidation, negative skin friction shall be taken into account when calculating the allowable axial bearing capacity of piles. [Commentary] When a pile penetrates through a clayey soft layer to reach a bearing stratum, the friction force from the soft layer acts upwards initially and bears a part of the load acting on the pile head. When the clayey soft layer is consolidated, the pile itself is supported by the bearing stratum and hardly settles, the direction of the friction force is reversed, as shown in Fig. C- 4.1.1. The friction force on the pile circumference now ceases to resist the load acting on the pile head, but instead turns into a load downwards and places a large burden on the toe of the pile. This friction force acting downwards on the pile circumference is called the negative skin friction or negative friction. [Technical Notes] (1) Although the actual value of negative skin friction is not well known yet, the maximum value may be obtained from equation (4.1.11). R nf, max where Rnf, max： j： L2： f s：

= jL2 f s

(4.1.11)

maximum negative skin friction (for single pile) (kN) circumference of piles (perimeter of closed area in the case of H-shaped steel piles) (m) length of piles in the consolidating layer (m) mean skin friction intensity in the consolidating layer (kN/m2)

(2) In the above, f s in clayey ground is sometimes taken at qu/2. If a sand layer is located between consolidating layers, or if a sand layer lies on top of consolidating layer, the thickness of the sand layer should be included in L2. The skin friction in the sand layer is sometimes taken into account for f s . The maximum value of negative skin friction in such cases is expressed by equation (4.1.12). qu Lc R nf, max = æ 2N s2 L s2 + -----------ö j è 2 ø where Ls2： thickness of sand layer included in L2 (m) Lc： thickness of clay layer included in L2 (m) L s2 + L c = L 2 N s2： mean N-value of the sand layer of thickness Ls2 q u： mean unconfined compressive strength of clayey layer of thickness Lc (kN/m2)

sUH + A g gL 2 (4.1.13) R nf, max = --------------------------------n where Rnf, max： maximum negative skin friction per pile (for piles group) (kN) U： perimeter of pile group (m) H： depth from the ground level to the bottom of consolidating layer (m) s ： mean shear strength of soil in the range of H (kN/m2) Ag： bottom area of pile group (m2) g： mean unit weight of soil in the range of L2 (kN/m3) n： number of piles in pile group

Consolidated layer

(3) In case of pile group, the negative skin friction may be calculated by assuming the piles group as a single, deep foundation. The negative skin friction per pile is then calculated by dividing it by the number of piles (see Fig. T- 4.1.5).

(4.1.12)

Fig. T- 4.1.5 Skin Friction of Pile Group

Equations (4.1.11) to (4.1.13) give the maximum value for negative skin friction. The actual values will be affected by the amount of consolidation settlement and the speed of consolidation, the creep characteristics of the soft clayey layer, and the deformation characteristics of the bearing stratum. (4) When calculating the allowable axial bearing capacity of piles, there is some uncertainty in evaluating the influence of negative skin friction. In one method, the influence of negative skin friction is examined by checking whether the force transmitted to the tip of the pile exceeds both the yield load value of the ground at the tip of the pile and the yield compressive strength of cross section of the pile. That is, by denoting the allowable axial bearing capacity under normal conditions as Ra , it should satisfy equations (4.1.14) and (4.1.15) as well as guaranteeing the required safety factor for ordinary loads.

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PART V FOUNDATIONS

1 (4.1.14) R a ≦ ------- R p R nf, max 1.2 R a ≦ s f A e R nf, max (4.1.15) where Ra： allowable axial bearing capacity (ordinary) (kN) Rp： end bearing capacity of pile (ultimate value) (kN) Rnf,max： maximum negative skin friction (kN) (the smaller of the values for single pile or pile group) sf： compressive stress of pile at yield point (kN/m2) Ae： effective section area of pile (m2) The value for end bearing capacity Rp may be taken at 300NAp in equation (4.1.1). If the pile has penetrated into the bearing stratum, the skin friction in the bearing stratum may be included in the end bearing capacity (see Fig. T- 4.1.6). (4.1.16) R p = 300NA p + 2N s1 L s1 j where Rp： end bearing capacity of pile (ultimate value) (kN) N： N-value of ground at pile tip Bearing straum Ap： area of pile tip (m2) L s1 = L 1： length of pile inside the bearing straum (sandy ground) (m) N s1： mean N-value for the zone of Ls1 j： circumference of pile (m) Fig. T- 4.1.6 End Bearing Capacity

4.1.11 Examination of Settlement of Piles The allowable axial bearing capacity of pile shall be determined in such a way that an estimated settlement of pile head does not exceed the allowable settlement determined for superstructures.

4.2 Allowable Pulling Resistance of Piles 4.2.1 General (Notification Article 43, Clause 3) The standard allowable pulling resistance of piles shall be calculated by estimating the maximum pulling resistance by an appropriate method and dividing it by the safety factor. The factors listed below shall be taken into consideration to the extent as necessary. In such cases, the safety factor shall take an appropriate value in view of the characteristics of structure and the strength of ground. (1) Allowable tensile stress of pile material (2) Effect of pile joints (3) Action of pile group (4) Upward displacement of piles by pulling

Deadweight

The standard allowable pulling resistance of piles is first obtained by dividing the maximum pulling resistance of single piles by an appropriate value of safety factor. Then, the allowable pulling resistance of piles is evaluated on the basis of it by taking account of the stress of pile materials, the influence of joints, the actions of pile group, and the upward displacement. Here, “maximum pulling resistance” means the maximum value of the pulling load shown in Fig. C- 4.2.1. In a pulling test of piles, the load decreases once the upward displacement has exceeded a certain limit, and finally only the deadweight of the pile remains. Caution should be paid to the difference in magnitude between the maximum load and the ultimate load.

Displacement

[Commentary]

Maximum pulling resistance

Pulling resistance

Fig. C- 4.2.1 Maximum Pulling Resistance

-291-


4.2.2 Standard Allowable Pulling Resistance The standard allowable pulling resistance shall be calculated by dividing the maximum pulling resistance of a single pile by an appropriate value of safety factor. [Technical Notes] (1) Table T- 4.2.1 gives guidelines for the minimum values of safety factor when calculating the standard allowable pulling resistance from the maximum pulling resistance of a single pile. (2) The minimum value of safety factor during an earthquake has been established at 2.5. This has been set lower than the value for ordinary conditions because the duration of seismic load is short. When there is a risk of liquefaction of the sand layer during an earthquake, the allowable pulling resistance shall be determined by taking the liquefaction into consideration. Table T- 4.2.1 Guidelines for Minimum Values of Safety Factor Ordinary condition

3.0


2.5

(3) The deadweight of pile acts always as a pulling resistance together with the weight of soil packed inside it. Therefore, the pile deadweight needs not be divided by the safety factor, and it is reasonable to calculate the standard allowable pulling resistance from the maximum pulling resistance by the equations below. When the deadweight of pile is relatively small, however, this process of adding the pile deadweight is generally be omitted. If the diameter of pile is too large, the soil packed into the pile is not always lifted up together with the pile but may be expected to slip-down from the pile. (a) When the maximum pulling resistance is calculated by pulling tests R ut1 W p R at = W p + ----------------------F

(4.2.1)

(b) When the maximum pulling resistance is calculated by a static bearing capacity equation R ut2 R at = W p + ---------F where Rat： allowable pulling resistance of piles (kN) Wp： deadweight of piles (with buoyancy subtracted) (kN) Rut1： maximum pulling resistance of pile (from pulling tests) (kN) Rut2： maximum pulling resistance of pile (from static bearing capacity formulas) (kN) F： safety factor

(4.2.2)

(4) Lowering of Safety Factor If the ground consists of well compacted high quality sandy soil and the bearing capacity can be estimated with sufficient accuracy on the basis of reliable data such as the results of accurate soil investigations and pulling tests as well as the predictions by reliable analysis methods, the value of safety factor may be lowered.

4.2.3 Maximum Pulling Resistance of Single Pile The maximum pulling resistance of a single pile shall be obtained on the basis of the results of pulling tests. [Commentary] Unlike axial bearing capacity, there are few comparative data for pulling resistance, and indirect estimations may entail some risk. Thus pulling tests should be conducted to calculate the maximum pulling resistance of a single pile. However, in the case of relatively soft cohesive soil, skin friction during driving of a pile is considered to be virtually the same as that during pulling of piles. Therefore, the maximum pulling resistance may be estimated from the results of loading tests (pushing direction) and static bearing capacity equations. [Technical Notes] (1) Estimation of maximum pulling resistance by static bearing capacity formulas may follow the explanation given in 4.1.5. Estimation of Ultimate Axial Bearing Capacity by Static Bearing Capacity Formulas. In this case, the end bearing capacity in the first term of equations (4.1.1) and (4.1.3) should be ignored. Thus, for piles driven by hammer, the following equations may be used.

-292-

PART V FOUNDATIONS

(a) Sandy ground R ut = 2NA s

(4.2.3)

(b) Clayey ground R ut where Rut： N： As： c a：

= ca As

(4.2.4)

maximum pulling resistance of pile (kN) mean N-value for total embedded length of pile total circumference area of pile (m2) mean adhesion for total embedded length of pile (kN/m2) (see 4.1.5 Estimation of Ultimate Axial Bearing Capacity by Static Bearing Capacity Formulas)

(2) Terzaghi’s equation is sometimes used as a static bearing capacity formula to estimate the maximum pulling resistance of piles, as shown in equation (4.2.5). In this case, the values calculated by equations (4.2.3) and (4.2.4) are to be compared with the value that calculated by using Terzaghi’s equation to obtain the most appropriate value. R ut = R f = jLf s

(4.2.5)

S ( c ai + K s q i m )l i f s = ---------------------------------------L where Rut： maximum pulling resistance of pile (kN) Rf： skin friction of pile (kN) j： circumference of pile (m) L： embedded length of pile (m) f s： mean skin friction intensity (kN/m2) cai： adhesion between soil and pile in the i-th layer (kN/m2) Ks： coefficient of horizontal earth pressure acting on pile qi： mean effective overburden pressure in the i-th layer (kN/m2) m： coefficient of friction between pile and soil li： thickness of the i-th layer (m)

(4.2.6)

For ca and m, see 4.1.5 Estimation of Ultimate Axial Bearing Capacity by Static Bearing Capacity Formulas. The value of the coefficient of horizontal earth pressure Ks in the case of pulling is considered to be smaller than the value in the case of bearing. A value between 0.3 and 0.7 (close to the coefficient of earth pressure at rest) is generally used.

4.2.4 Examination of Tensile Stress of Pile Materials (Notification Article 43, Clause 4) The allowable pulling resistance of pile shall not exceed the value calculated by multiplying the allowable tensile stress of pile materials by the effective cross-sectional area of pile. 4.2.5 Matters to Be Considered for Obtaining Allowable Pulling Resistance of Piles When calculating the allowable pulling resistance of piles, the following shall be taken into account: (1) In case of spliced piles, pulling resistance of piles below joints shall be ignored. When high-quality joints are installed in steel piles, however, pulling resistance of lower piles may be taken into account within the allowable tensile strength of the joints, after confirming the reliability of the joint. (2) In case of pile group, pulling resistance shall be calculated as that of a single block surrounded with the envelope surface of outermost piles in the group of piles. (3) When determining the allowable pulling resistance of piles, the amount of allowable upward displacement of pile heads imposed by the superstructures shall be taken into account.

4.3 Allowable Lateral Bearing Capacity of Piles 4.3.1 General The allowable lateral bearing capacity of a single pile shall be determined as appropriate on the basis of the behavior of the pile when it is subject to lateral forces.

-293-


[Commentary] (1) For the axial bearing capacity, the standard value for allowable bearing capacity is calculated by dividing the ultimate bearing capacity by the safety factor. But for the lateral bearing capacity, its allowable value is calculated directly from the behavior of the piles without referring to the ultimate bearing capacity. (2) The allowable lateral bearing capacity of a pile should be determined to satisfy the following two conditions: (a) The bending stress arising in a pile should not exceed the allowable bending stress of pile material. (b) The displacement of pile head (horizontal displacement) should not exceed the allowable displacement imposed by the superstructure. [Technical Notes] (1) Embedded Length of Piles The length of embedded part of pile that yields effective resistance against external forces is called the effective length. Piles are called the long pile when the embedded length is longer than their effective length. Piles are called the short pile when the embedded length is shorter than their effective length. (2) Piles Subject to Lateral Force The resistance when a pile is subject to lateral forces (horizontal or nearly horizontal external forces) is called the lateral resistance. It is categorized into the three basic forms shown in Fig. T- 4.3.1 5). (a) When one vertical pile is subject to horizontal force When a horizontal external force acts on a vertical pile, only the lateral resistance occurs and axial resistance is not involved. This is the simplest form of lateral resistance and called the lateral resistance of pile in a narrow sense. (b) When one batter pile is subject to horizontal force In this case, some part of external force is supported by the axial resistance. The burden ratio between the lateral and axial resistances is almost wholly determined by the inclination angle of the piles. Therefore, the bearing capacity may be divided into the lateral and axial resistances and examined separately. (c) When coupled piles are subject to horizontal force Coupled piles are those in which two or more piles with differing axial directions are combined. The simplest form of coupled piles is shown in Fig. T- 4.3.1 (c). In this case, most of the external force is supported by the axial resistance of coupled batter piles. Therefore, when the free length of the piles is long, the lateral resistance is usually ignored in estimating the bearing capacity. Instead, the axial bearing capacity alone is considered. With coupled piles, it is quite difficult to calculate the pile head displacement. A number of methods have been proposed to date, but none of them is sufficient to apply to the calculation of pile head displacement (see 4.3.6 Lateral Bearing Capacity of Coupled Piles [Technical Notes]). However, since the displacement of coupled piles is far smaller than that of single piles, displacement rarely becomes a problem.

(a) When one vertical pile is subject to horizontal force

(b) When one batter pile is subject to horizontal force

(c) When coupled piles are subject to horizontal force

Fig. T- 4.3.1 Piles Subject to Lateral Force

-294-

PART V FOUNDATIONS

4.3.2 Estimation of Allowable Lateral Bearing Capacity of Piles (Notification Article 43, Clause 5) The allowable lateral bearing capacity of piles shall be determined at an appropriate value on the basis of loading tests or analytical methods, or estimation methods by combining these results. 4.3.3 Estimation of Pile Behavior Using Loading Tests When loading tests are conducted to estimate behavior of a single pile subject to lateral force, all the due considerations shall be given to the differences in the pile and load conditions in between those of actual structures and loading tests. [Technical Notes] When loading tests have been conducted under the conditions same as those in actual structures, the allowable bearing capacity may be obtained from the loading test results by the following method. The load and pile head displacement curve in lateral loading tests generally takes a curved form from the beginning. Therefore, it is difficult to read clearly the yield load or the ultimate load on the load and pile head displacement curve (except for short piles). It is because a pile with long embedded length is only subject to small-scale and progressive ground failure phenomena but not to overall failure. Therefore, the load and pile head displacement curve is not used to obtain the yield load or the ultimate load but to confirm the pile head displacement itself. In other words, once the allowable pile head displacement is determined, the load corresponding to this displacement on the load and pile head displacement curve defines the allowable lateral bearing capacity. Furthermore, the bending stress corresponding to this allowable bearing capacity also needs to be accounted for. That is, when a load equal to the allowable bearing capacity is applied, the maximum bending stress arising in the pile must not exceed the allowable bending stress of the pile material (see 4.4.4 Allowable Stress for Pile Materials). To calculate the allowable lateral bearing capacity of short piles, overturning of piles must be considered, in addition to the pile head displacement and bending stress mentioned already. When the overturning load cannot be ascertained, the maximum test load may be used instead of the overturning load.

4.3.4 Estimation of Pile Behavior Using Analytical Methods When estimating behavior of a single pile subject to lateral force by using analytical methods, it shall be standard to analyze the pile as a beam rested on an elastic foundation. [Commentary] Methods of analytically estimating the behavior of a single pile subject to lateral force as a beam rested on an elastic foundation include the relatively simple Chang’s method as well as the Port & Harbor Research Institute (PHRI) method 6). The latter is preferred as the method of analysis, because it can accurately express the actual behavior of single piles. However, Chang’s method may be used when no significant difference is expected between the two methods. [Technical Notes] (1) Basic Equation for Beam on Elastic Foundation Equation (4.3.1) is the basic equation for analytically estimating behavior of a pile as a beam on an elastic foundation. d4y (4.3.1) EI -------4- = P = pB dx where EI： flexural rigidity of pile (kN·m2) x： depth from ground level (m) y： displacement of pile at depth x (m) P： subgrade reaction per unit length of pile at depth x (kN/m) P p： subgrade reaction per unit area of pile at depth x (kN/m2); p = --B B： pile width (m) Shinohara, Kubo, and Hayashi proposed the PHRI method as an analytical method considering the nonlinear elastic behavior of the ground 6). This method has a significant merit that it can describe the behavior of actual piles more faithfully than other methods. The PHRI method uses equation (4.3.2) to describe the relationship between the subgrade reaction and the pile displacement. (4.3.2) p = kx m y 0.5 where k： constant of lateral resistance of ground (kN/m3.5 or kN/m2.5) m： index 1 or 0 -295-


(2) PHRI Method (a) Characteristics of the PHRI method In the PHRI method, the ground is classified into the S type and the C type. The relationship between the subgrade reaction and the pile displacement for each ground is assumed by equation (4.3.3) and (4.3.4), respectively. S-type ground: (4.3.3) p = k s xy 0.5 C-type ground: p = kc y 0.5 where ks： constant of lateral resistance in S-type ground (kN/m3.5) kc： constant of lateral resistance in C-type ground (kN/m2.5)

(4.3.4)

The identification of S-type or C-type ground and the estimation of ks and kc are based on the results of loading tests and soil investigation. In the PHRI method, the nonlinear relationships between p and y are introduced as given by equations (4.3.3) and (4.3.4) to reflect the actual state of subgrade reaction. Therefore, the solutions under individual conditions would remain unattainable without help of numerical calculation, and the principle of superposition could not be applied. The results of many full-scale tests have confirmed that this method reflects the behavior of piles more accurately than the conventional methods. It is commented here that for piles to behave as long piles, they must be at least as long as 1.5 lm1 (lm1: depth of the first zero point of bending moment in the PHRI method). (b) Constants of lateral resistance of the ground The two ground types in the PHRI method are defined as follows. ① S-type ground (i)The relationship between p-y is expressed as p = ks xy0.5 (see equation (4.3.3)) (ii) The N-value by the standard penetration test increases in proportion to the depth. (iii) Actual examples: sandy ground with uniform density, and normally-consolidated clayey ground. ② C-type ground (i) The p-y relationship is expressed as p = kc y0.5 (see equation (4.3.4)) (ii) The N-value by the standard penetration test is constant regardless of depth. (iii) Actual examples: sandy ground with compacted surface, and heavily-preconsolidated clayey ground. In the S-type ground, the relationship between the increment N of the N-value per meter in depth and the lateral resistance constant ks has been obtained as shown in Fig. T- 4.3.2. Even if the distribution of N-values with depth does not have the value 0 at the ground level, the increment N may be determined from the mean gradient of N-values passing through the zero point at the ground level. In the C-type ground, the relationship between the N-value itself and kc has been obtained as shown in Fig. T- 4.3.37). In this way, the approximate value of ks or kc may be judged from the distribution of N-values with depth. (c) Estimation of lateral resistance constants by loading tests Estimations of the lateral resistance constants by using the N-value can only provide approximate values. Loading tests should be conducted to obtain more accurate values. The constants ks and kc are determined from the ground conditions alone, and are unaffected by other conditions unlike Chang’s Es. Therefore, if ks or kc can be obtained by a loading test, its value can be applied to other conditions as well. (d) Effect of pile width There are two ways of considering the effect of pile width. The first is to regard that the pile width B has no effect on the relationship between the subgrade reaction p per unit area and the displacement y. The second, as proposed by Terzaghi, is to assume that the value of p corresponding to a given y value is inversely proportional to B. Shinohara, Kubo and Sawaguchi conducted model experiments on the relationship between the ks value in sandy ground (PHRI method) and B. The results are shown in Fig. T- 4.3.4. It seems to support a combination of the two theories mentioned above in such a way that the first theory holds true if the pile width B is sufficiently large. On the basis of these results, it was decided not to consider the effect of pile width in the PHRI method. (e) Effect of pile inclination For batter piles, the relationship between the inclination angle of the piles and the ratio of the lateral resistance constant of batter piles to that of vertical piles has been obtained as shown in Fig. T- 4.3.5.

-296-

PART V FOUNDATIONS

1. Alton, Illinois (FEAGIN)

2. Winfield, Montana (GLESER)

3. Port Hueneme (MASON)

4.5. Hakkenbori No.1, No.2

6. Ibaragigawa (GOTO)

7. Osaka National Railways (BEPPU)

8.9. Tobata No.6, No.9

10. Tobata K-I (PHRI)

11. Tobata K-II (PHRI)

12. Tobata L-II (PHRI)

13. Kurihama model experiment

14. Shin-Kasai Bridge (TATEISHI)

15. Yamanoshita (IGUCHI)

N Fig. T- 4.3.2 Relationship between N and ks

1. Tobata K-I (TTRI)

2. Tobata K-III (TTRI)

3. Tobata K-IV (TTRI)

4. Tobata L-II (TTRI)

5. Tobata L-IV (TTRI)

6. Hakkenbori No.1

7. Hakkenbori No.2

8. Osaka National Railways

9. Yahata Seitetsu No.6

10. Yahata Seitetsu No.9

11. Tobata preliminary test-1 (TTRI)-1

12. Tobata preliminary test-2 (TTRI)-2

13. Wagner (Callif.) No.15

14. Wagner (Callif.) No.25

15. Wagner-1 (Alaska)-1

16. Wagner-1 (Alaska)-2

17. Tokyo National Railways b

18. Tokyo National Railways A4

19. Tokyo National Railways B

N-value Fig. T- 4.3.3 Relationship between N-value and kc

-297-

Lateral resistance constant

ks

(kN/m3.5)


1st Series 2nd Series 3rd Series

Legend maximum pile head displace p-y curve bending moment -ment

Pile width (cm)

Fig. T- 4.3.4 Relationship between ks and Pile Width

= k/k 0

k0

Indoor experiments

Field experiments Value of , when θ = 0

k

Fig. T- 4.3.5 Relationship between Pile Inclination Angle and Ratio k

-298-

y

y0

-299-

M0

S

M

lm

M s,max

M s,max

l0

L

K 1, K 2, K 3, K 4

Pile head rigidity factor

angle zero point

Depth of deflection

point

Depth of 1st steady

occurs

Depth at which

of embedded parts

Maximum bending moment

moment

Pile head bending

pile members

Shear strength of

pile members

Bending moment of

θt

Pile head inclination

displacement

Ground level

yt

Pile head displacement

Deflection curve

diagram

Bending moment

diagram

Deflection curve

Situation of pile

symbols

and explanation of

of deflection curve

0

h

(y x)

0

(y x)

(yz)

θt Mt

Mm

–M t

(M x)

(M z)

βh + 1 l 0 = 1 tan – 1 β βh – 1

1 + βh 0 l 0 = 1 tan – 1 β βh 0

12EI β 3 1 + βh K2 = K3 = K1 ⋅ 2β ( 1 + βh ) 3 + 2 4EI β ( 1 + βh ) 3 + 1/2 K4 = ⋅ ( 1 + βh ) ( 1 + βh ) 3 + 2

K1 =

L = 1 tan – 1 ( – βh ) β

l m = 1 tan – 1 1 β βh

Ht 1 + ( βh ) 2 ⋅ exp ( – βl m ) 2β

1 + βh K Ht = 2 Ht K1 2β

M s,max = –

M0 =

H M z = t ( 1 + βh – 2βz ) 2β H M x = t e – β x [ ( 1 – βh ) cos βh – ( 1 + βh ) sin β x ] 2β Sz = – H t S x = – H t e – β x [ cos β x – βh sin β x ]

1 l m = 1 tan – 1 1 + 2βh 0 β

L = 1 tan – 1 [ – ( 1 + 2βh 0 ) ] β

M

∞

0

H t= H 0

yt θ t

(y x)

Mm (M x)

= 0)

K 1 = 4EI β 3 K 2 = K 3 = 2EI β 2 K 4 = 2EI β

L = 3π = 2.3562 β 4β

l 0 = π = 1.5708 2β β

l m = π = 0.7854 β 4β

M s,max = – 0.3224

M0 = 0 Ht β

S x = – H t e – β x ( cos β x – sin β x )

h =0

∞

M0

yt = y0

(M x)

M

Ht βH t = 4EI β 3 Bkh

Ht e – β x ( cos β x + sin β x ) 4EI β 3

H t – βx e ( cos β x – sin β x ) 2β

Ht 2β

L = π = 3.1416 β β

l 0 = 3π = 2.3562 β 4β

l m = π = 1.5708 β 2β

M s,max = – 0.2079

M0 =

Ht β

S x = – H t e – β x cos β x

Mx =

θt = 0

y0 = yt

yt =

yx =

Ht H0

If pile head does not rotate

Embedded underground

Mt

H t – βx e sin β x β

Ht 2EI β 2

Mx = –

θt =

Ht 2βH t = 2EI β 3 Bk h

y0 = yt

θt = 0

( 1 + 2βh 0 ) 2 + 1 ⋅ exp ( – βl m )

(M z)

(M z)

1 + βh Ht 4EI β 3

( 1 + βh ) 2 1 + βh Ht + Mt EI β 2EI β 2

Ht 2β

M0

yt =

y0 =

M s,max = –

yt

Basic system (but

H ( 1 + βh ) 3 + 2 Ht = t K1 12EI β 3

1 + βh 0 Ht 2EI β 3

M0 = – Mt

∞

y0 0

Ht

4

3

: Coefficient of horizontal subgrade reaction (kN/m ) : Height of pile head above ground (m)

Bk h / 4EI (m –1) M h0 = h + t Ht

kh h β

Ht ( 1 + βh ) H t 2 H t 3 yx = e – β x cos β x yz = yt – z + z 6EI 4EI β 2EI β 3 Ht yx = e – β x [ ( 1 + βh ) cos β x + ( 1 – βh ) sin β x ] If t 0, use equations in putting 4EI β 3 h 0 = M t / H t : the same applies below yt =

Sz = – H t S x = – H t e – β x [ cos β x – ( 1 + 2βh 0 ) sin β x ]

0

H t : Lateral force on pile head (kN) M t : External force moment on pile head (kN.m) B : Pile diameter (m) EI : Flexural rigidity (kN.m2)

does not rotate

( 1 + βh ) 3 + 1/2 ( 1 + βh ) 2 Ht + Mt 2EI β 2 3EI β 3

M z = – M t – H tz H M x = – t e – β x [ βh 0 cos β x + ( 1 + βh 0 ) sin β x [ β

θt =

y0 =

yt =

h

If pile head

Protruding above ground

M H y z = y t – θ tz + t z 2 + t z 3 2EI 6EI Ht yx = e – β x [ ( 1 + βh 0 ) cos β x – βh 0 sin β x ] 2EI β 3

∞

Ht y0

yt

d 4 yx + Bk h y x = 0 dx 4

EI

Embedded sections:

Basic formation

x

L

[Symbols]

l

d 4 yz =0 dz 4

L

EI

l

h

lm

h lm

Exposed sections:

l

(M t)

L

(yz)

lm

z

l

Ht

L

Differential equations

lm

Table T- 4.3.1 Calculations for Piles of Semi-Infinite Length if kh Is Constant

PART V FOUNDATIONS


(3) Chang’s Method

64748

(a) Calculation equation Using the elasticity modulus of the ground Es = Bkh, the elasticity equation of piles are expressed as follows. d4 y1 - = 0 Exposed section: (0 ≧ x ≧ -h) EI ---------dx 4 (4.3.5) d4 y2 - + Bk h y 2 = 0 (x ≧ 0) Embedded section: EI ---------dx 4 By calculating these general solutions (with Bkh as a constant) and inputting the boundary conditions, the solution for piles of semi-infinite length can be obtained (see Table T- 4.3.1). According to Yokoyama, piles of finite length can also be calculated in a similar way to piles of infinite length as long as bL ≧ p. If shorter than this, a pile must be treated as having a finite length. Diagrams are available to simplify this process. (b) Estimation of kh in Chang’s method ① Terzaghi’s proposal8) Terzaghi proposed the following values for the coefficient of lateral subgrade reaction in cohesive and sandy soil: (i) In case of clayey soil 0.2 k h = ------- k h1 B Es = where kh ： B： kh1：

(4.3.6)

Bk h = 0.2k h1

(4.3.7)

coefficient of lateral subgrade reaction (kN/m3) pile width (m) value listed in Table T- 4.3.2

(ii) In case of sandy soil x k h = n h --B E s = Bk h = n h x where x： depth (m) B： pile width (m) nh： value listed in Table T- 4.3.3

(4.3.8) (4.3.9)

In sandy soil, Es is a function of depth and thus cannot be introduced directly into Chang’s method. For such cases, Chang states that Es should be taken at one-third of the value at the depth of ly1 (depth of the first zero-displacement point). However, ly1 itself is a function of Es, and repeated calculations have to be made to obtain the value of Es. The book by Yokoyama 9) describes the method of calculation without repetition. Terzaghi assumes that the value of kh is inversely proportional to the pile width B, as shown in equations (4.3.6) and (4.3.8). Other opinions suggest that pile width is irrelevant (see [Technical Notes] (2) (d)). Table T- 4.3.2 Coefficient of Lateral Subgrade Reaction Consistency of clay Unconfined compressive strength qu Range of kh1

(kN/m3)

Proposed value of kh1

Hard

Very hard

Solid

100 ～ 200

200 ～ 400

400 or greater

16,000 ～ 32,000

32,000 ～ 64,000

64,000 or greater

24,000

48,000

96,000

(kN/m2)

(kN/m3)

Table T- 4.3.3 Value of n h Relative density of sand nh for dry or wet sand

Loose

Medium

Dense

(kN/m3)

2,200

6,600

17,600

(kN/m3)

1,300

4,400

10,800

nh for submerged sand

② Yokoyama’s proposal Yokoyama 9) has collated the results of lateral loading tests on steel piles conducted in Japan and he reversely calculated kh by using these results. Figure T- 4.3.6 shows the relationship between these values and the mean N-values at depths down to b -1 from the ground level. In this case, Es = khB is taken as valid for both sandy and clayey soil, while kh itself is assumed not to be affected by B. Although the reversely-calculated values of kh by using the actually observed data decrease in -300-

PART V FOUNDATIONS

proportion to the increase of load, Fig. T- 4.3.6 shows the values of kh corresponding to the load at which the bending stress of steel material reaches 100 ~ 150 MN/m2. This diagram may be used when approximating the value of Es from soil conditions alone without conducting loading tests in situ.

1. Yamaborigawa

2. Tobata

3. Tobata K-I

4. Tobata L-II

5. Tobata K-II

6. Tobata K-III

7. Tobata L-IV

8. Tobata K-IV

9. Shell Ogishima

10. Ibaragigawa

11. Takagawa

12. Tokyo Supply Warehouse

13. Kasai Bridge

14. Aoyama

15. Den-en

N-value Fig. T- 4.3.6 Reversely-Calculated Values of kh from Horizontal Loading Tests on Piles

4.3.5 Consideration of Pile Group Action When piles are used as a pile group, the effect of pile group action on the behavior of individual piles shall be considered. [Technical Notes] When the interval of the driven piles exceeds the value listed in Table T- 4.3.4, the effect of the pile group on the lateral bearing capacity of individual piles may be ignored. In the table, “transverse” means in the direction perpendicular to that of the external force, and “longitudinal” means in the direction of the external force. Table T- 4.3.4 Minimum Pile Interval for Lateral Bearing Capacity of Individual Piles Sandy soil Cohesive soil

transverse

1.5 times the pile diameter

longitudinal


transverse


longitudinal


4.3.6 Lateral Bearing Capacity of Coupled Piles The lateral bearing capacity of a foundation of the structure with coupled piles shall be determined as appropriate in view of structural characteristics of the foundation. [Commentary] (1) Distribution of Horizontal Force in Foundation with a Combination of Vertical and Coupled Piles When a horizontal force acts on a foundation with a combination of vertical and coupled piles, the force borne by vertical piles is far smaller than that borne by coupled piles under the condition of equal horizontal displacement. It may generally be assumed that all of the horizontal force is borne by the coupled piles.

-301-


(2) Lateral Bearing Capacity of Coupled Piles There are two groups of calculation methods for the lateral bearing capacity of coupled piles. The first group only takes account of the resistance from the axial bearing capacity of each pile. The second group takes account of the resistance from the axial bearing capacity of each pile and the lateral bearing capacity of each pile in consideration of the bending resistance of piles. [Technical Notes]

64748

(1) Estimation of Lateral Bearing Capacity of Coupled Piles Considering the Axial Bearing Capacity of Piles Only When the resistance forces are borne by axial bearing capacity only, as shown in Fig. T- 4.3.7, the vertical and horizontal external loads acting on the head of a pair of coupled piles are divided into the axial force of each pile. This means that the force acting in the axial direction of each pile should be less than the allowable axial bearing capacity (or allowable pulling resistance). The axial force is calculated by using equation (4.3.10) or a graphic solution (see Fig. T- 4.3.7). V i sin q 2 + H i cos q 2 P 1 = -------------------------------------------sin ( q 1 + q 2 ) (4.3.10) V i sin q 1 H i cos q 1 P 2 = -------------------------------------------sin ( q 1 + q 2 ) where P1, P2： pushing force acting on each pile (or pulling force when negative) (kN) q1, q2： inclination angle of each pile (º) Vi： vertical external load of coupled piles (kN) Hi： horizontal external load of coupled piles (kN)

(Out-batter pile)

Fig. T- 4.3.7 Axial Forces of Coupled Piles

(In-batter pile)

Fig. T- 4.3.8 Coupled Piles in View of Soil Resistance Due to Deflection Caused by Pile Bending Moment

(2) Estimation of Lateral Bearing Capacity of Coupled Piles Considering Lateral Bearing Capacity of Individual Piles There are various ways of calculating the lateral bearing capacity of the coupled piles, taking account of the lateral bearing capacity of individual piles. For example: ① Setting conditions whereby the displacement of each pile is always the same at the point of connection of coupled piles, on the assumption that the spring characteristics in the axial and lateral directions at the pile head are elastic9). ② Calculating the ultimate resistance of the coupled piles on the assumption that the axial and lateral bearing capacities of piles indicate an elasto-plasticity characteristic10). ③ Calculating the load and displacement at the pile heads, or the settlement and the upward displacement of -302-

PART V FOUNDATIONS piles in the case of ② on the basis of empirical equations11). ④ Using the results of loading tests on single piles12). ⑤ Assuming that the yield state of each pile will occur in turn and the resistance of the yield member will be constant thereafter until the resistance of coupled piles reach the ultimate bearing capacity.

w2 N 2 = ------ { [ m 2 cos q 2 + m 1 cos q 1 cos ( q 1 + q 1 ) + w 1 sin q 1 sin ( q 1 + q 2 ) ]V D [ m 2 sin q 2 m 1 sin q 1 cos ( q 1 + q 2 ) + w 1 cos q 1 sin ( q 1 + q 2 ) ]H } m1 H 1 = ----- { [ w 1 sin q 1 w 2 sin q 2 cos ( q 1 + q 2 ) + m 2 cos q 2 sin ( q 1 + q 2 ) ] V D + [ w 1 cos q 1 + w 2 cos q 2 cos ( q 1 + q 2 ) + m 2 sin q 2 sin ( q 1 + q 2 ) ]H } m2 H 2 = ----- { [ w 2 sin q 2 w 1 sin q 1 cos ( q 1 + q 2 ) + m 1 cos q 1 sin ( q 1 + q 2 ) ]V D +[ w 2 cos q 2 + w 1 cos q 1 cos ( q 1 + q 2 ) + m 1 sin q 1 sin ( q 1 + q 2 ) ]H } D = ( w 1 + w 2 ) ( m 1 + m 2 ) + ( w 1 m 1 ) ( w 2 m 2 ) sin 2 ( q 1 + q 2 )

6444444744444448

The method ① is for calculating the distribution of horizontal force to each pile on the assumption that the axial and lateral resistances of a pile have elastic properties 9). In the coupled piles shown in Figure T- 4.3.8, the settlement of each pile at the pile head is proportional to the axial force acting on that pile and also the lateral displacement is proportional to the lateral force acting on that pile. On this assumption, the axial and lateral forces acting on each pile of the coupled piles can be calculated using equation (4.3.11), derived from the conditions of force equilibrium and compatibility of displacements. w1 N 1 = ------ { [ m 1 cos q 1 + m 2 cos q 2 cos ( q 1 + q 2 ) + w 2 sin q 2 sin ( q 1 + q 2 ) ]V D + [ m 1 sin q 1 m 2 sin q 2 cos ( q 1 + q 2 ) + w 2 cos q 2 sin ( q 1 + q 2 ) ]H }

(4.3.11)

1 + --- [ ( m 1 w 1 ) sin 2q 1 + ( w 2 m 2 ) sin 2q 2 ]H } 2 1 ì1 h 1 ¢ = h 2 ¢ = --- í --- [ ( m 1 w 1 ) sin 2q 1 + ( w 2 m 2 ) sin 2q 2 ]V Dî2 ü + [ m 1 sin 2 q 1 + w 1 cos 2 q 1 + m 2 sin 2 q 2 + w 2 cos 2 q 2 ] ý þ where N1, N2： H1, H2： V： H： q1, q2： w1, w2： m1, m2： d1¢,d2¢： h1¢,h2¢：

644474448

Vertical and horizontal displacements of the pile head are calculated by equation (4.3.12) 1 d 1 ¢ = d 2 ¢ = --- { [ w 1 sin 2 q 1 + m 1 cos 2 q 1 + w 2 sin 2 q 2 + m 2 cos 2 q 2 ] V D

(4.3.12)

axial force acting on each pile (compressive force is indicated by positive value) (kN) lateral force acting on each pile (kN) vertical load per pair of coupled piles (kN) horizontal load per a pair of coupled piles (kN) inclination angle of each pile (º) axial spring constant of each pile head (kN/m) lateral spring constant of each pile head (kN/m) vertical displacement of each pile head (m) horizontal displacement of each pile head (m)

The subscript numbers attached to the symbols, as shown in Fig. T- 4.3.8, are “1” for the pushed pile and “2” for the pulled pile when only a horizontal load acts. The values listed in Table T- 4.3.5 may be used for the spring constants of pile head. The symbols used in Table T- 4.3.5 are defined below. ( 1 + bl ) 3 + 1 ¤ 2 j D( bl ) = --------------------------------------( bl ) 3 ( 1 + bl ) 3 + 2 j D( bl ) = --------------------------------( bl ) 3 b =

4

Es --------4EI -303-


where l： l： E： A： I： Es： B： k h：

embedded length of piles (m) exposed length of piles (m) Young’s modulus of pile material (kN/m2) pile section area (m2) moment of inertia of pile (m4) elastic modulus of subsoil (kN/m2), Es =khB pile width (m) coefficient of lateral subgrade reaction (kN/m3)

The coefficient of lateral subgrade reaction kh may be calculated by multiplying the value of kh obtained in [Technical Notes] (3) (b) in 4.3.4 Estimation of Pile Behavior Using Analytical Methods by the factor obtained from Fig. T- 4.3.5, in accordance with the inclination of piles. Table T- 4.3.5 Spring Constants of Pile Head Bearing piles Axial spring constant of pile head (ω)

Cohesive soil

2AE w = --------------2l + l

Sandy soil

3AE w = ------------------3l + 2l

Without exposed section (λ ＝ 0)

ES m = 2EIb 3 = -----2b

With exposed section (λ ≠ 0)

3EI m = -----------------------3j l D ( bl )

Without exposed section (λ ＝ 0)

ES 3 m = 4EIb = -----b

With exposed section (λ ≠ 0)

12EI m = -----------------------3j l D ( bl )

Friction piles

Pile head hinged Lateral axial spring constant of pile head (μ)

AE w = -----------l+l

Pile head fixed

4.4 Pile Design in General 4.4.1 Load Sharing (1) Vertical loads shall be supported by piles alone. No bearing capacity shall be expected for the ground touching with the bottom of structures supported by piles. (2) Horizontal loads shall be supported by piles alone in principle. However, when enough resistance can be obtained by the earth pressure acting on the embedded section of structure alone, this resistance may be assumed to support the horizontal load. [Technical Notes] (1) Vertical Loads Some gap may appear between the bottom of the structure supported by piles and the ground underneath it as time passes, even if the bottom is constructed by touching the ground. Thus, for the sake of safety the bearing capacity of the ground under a structure should be ignored. (2) Horizontal Loads In principle, horizontal loads should be supported by piles alone. If the resistance due to passive earth pressure against the embedded section of structure can be expected, this resistance may be added. However it is generally difficult to calculate the resistance due to passive earth pressure in this case. There is no simple way of determining whether the passive earth pressure of the ground reaches its ultimate value in response to the pile head displacement corresponding to the allowable lateral bearing capacity of piles. In some cases, if a structure is displaced to the extent of passive earth pressure obtained by using Coulomb’s equation, this displacement of the structure could cause the pile bending failure. These factors must be fully considered when calculating this resistance due to passive earth pressure at the embedded section.

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PART V FOUNDATIONS

4.4.2 Load Distribution Within the same foundation, piles shall be so arranged that vertical and horizontal forces acting on individual piles are as equal as possible. 4.4.3 Distance between Centers of Piles When determining the distance between the centers of piles to be driven, the workability, deformation behavior of surrounding ground, and behavior as a pile group shall be taken into account. 4.4.4 Allowable Stresses for Pile Materials Allowable stresses for pile materials shall be determined as appropriate in view of their characteristics. [Technical Notes] (1) Pile Materials Piles used for the construction of port and harbor facilities are made of timber, concrete, or steel. For timber and steel piles, see the relevant sections of Part III, 6.2 Allowable Stress for Timber and Part III, 2.3 Allowable Stresses, respectively. Allowable stresses for concrete are defined in Part III, 3.3 Design Based on Allowable Stress Method. In the construction of structure supported by concrete piles, these piles entail problems of driving or underground construction. These kinds of conditions are different from the conditions in the construction of other concrete structures. Therefore, the allowable stresses for concrete piles should be lower than those for concrete members used for other structures. (2) Allowable Stresses for Concrete Piles Table T- 4.4.1 lists examples of the allowable stresses for concrete piles. Table T- 4.4.1 Allowable Stress for Concrete Piles Allowable stress

Type of pile

Reinforced concrete piles (RC piles)

Precast concrete piles formed via centrifugal force

Concrete piles

Prestressed high-strength concrete piles (PHC piles)

Type

Amount

Allowable compressive stress used when determining axial bearing capacity

30% of design standard strength, and not exceeding 12 MN/m2

Others

As specified in Part III, 3.3 Design Based on Allowable Stress Method

Allowable compressive stress used when determining axial bearing capacity


Allowable bending compressive stress


Allowable axial tensile stress and allowable bending tensile stress

3.0 MN/m2 (Type A) 5.0 MN/m2 (Types B, C)

Allowable bending compressive stress (including cases with axial force)

25% of design standard strength, and not exceeding 6.0 MN/m2

Others

70% of the allowable stress for concrete specified in Part III, 3.3 Design Based on Allowable Stress Method

Allowable bending compressive stress (including cases with axial force)

20% of design standard strength, and not exceeding 5.0 MN/m2

Others

50% of the allowable stress for concrete defined in Part III, 3.3 Design Based on Allowable Stress Method

Cast-in-place concrete piles with outer casing

Cast-in-place concrete piles with no outer casing

The values given above may be increased up to 1.5 times when considering short-term load or earthquake impact

-305-


4.5 Detailed Design 4.5.1 Examination of Loads during Construction In the design of piles, it is recommended to examine not only the loads acting after completion of construction but also those during transportation, positioning, and driving. [Technical Notes] Pile driving formulas, designed to calculate the static ultimate bearing capacity of piles from dynamic penetration resistance, are difficult to make good use. Although estimations of bearing capacity using pile driving formulas have the advantage of being very simple, the problem lies in their accuracy. In Fig. T- 4.5.1 by Sawaguchi, the ultimate bearing capacity obtained from the pile driving formula for steel piles is compared with the results of loading tests in a form of the ratio in percentage of the former to the latter. The figure reveals major discrepancy and dispersion between the two. In clayey soil, soil is disturbed during pile driving and skin friction temporarily decreases. Therefore, the ultimate bearing capacity cannot be estimated by pile driving formulas. In sandy soil, meanwhile, pile driving formulas are said to be inaccurate for estimating the ultimate bearing capacity of friction piles. The limits of applicability of pile driving formulas are discussed in reference 13). Nevertheless, when driving a large number of piles into almost identical ground, pile driving formulas can be used as a reference for estimating the relative differences in bearing capacity per each driven pile. Thus, the application of these formulas should be restricted to construction management. However, they may also be used as reference to confirm variation in the bearing capacity of each pile or to finish the driving of each pile so that they are all governed by the same condition.

Hiley’s formula

Weisbach’s formula

Janbu’s formula

Danish formula

Smith’s formula

Fig. T- 4.5.1 Distribution of Ratios of the Estimation Values by Pile Driving Formulas to the Results of Loading Tests

Hiley’s equation is the most common pile driving formula, and is expressed by equations (4.5.1) and (4.5.2). W p ( 1 e 2 ) R du C 1 R du C 2 R du C 3 (4.5.1) R du S = ef F F -------------------------- --------------- --------------- --------------W H + Wp 2 2 2 The invidual terms of this equation represent the following: the term in the left-hand side is for “work necessary for penetrating piles”, and the four terms in the right-hand side are for “impact loss”, “loss due to elastic deformation of the pile”, “loss due to elastic deformation of the ground”, and “loss due to cushion”, respectively. ef F WH + e 2 W p (4.5.2) R du = ----------------------------------------- --------------------------C1 + C2 + C3 WH + Wp S + -------------------------------2 where Rdu： ultimate pile-driving resistance; i.e., dynamic ultimate bearing capacity (kN) WH： weight of hammer (ram) (kN) WP： weight of pile (including pile head attachments) (kN) F： impact energy (kJ) -306-

PART V FOUNDATIONS

ef： e： S： C1： C2： C3：

efficiency of hammer, ranging from 0.6 to 1.0, depending on the type of hammer14) rebound coefficient (e = 1 if completely elastic, and e = 0 if completely non-elastic) final penetration length of pile (m) elastic deformation of pile (m) elastic deformation of ground (m) elastic deformation of pile head cushion (m)

Most pile driving formulas are obtained by replacing C1, C2, C3, ef, e, etc. in equation (4.5.2) with appropriate values. Equation (4.5.3) is considered relatively well-suited to steel piles. Assuming the impact between hammer and pile to be elastic, i.e., e = 1, the following is derived: ef F (4.5.3) R du = ----------------------------------------C 1 + C2 + C 3 S + -------------------------------2 The term C1 + C2 + C3 in the above is the sum of elastic deformation of ground, pile, and pile head cushion. Of these, C1 + C2 are equal to the rebound K measured at the pile head in pile driving tests (see Fig. T- 4.5.2). With steel piles, elastic deformation C1 is dominant while C3 is generally smaller. Thus, if C3 is neglected, the following can be assumed: C1 + C2 + C3 ≒ C1 + C2 = K thus, (4.5.4)

678

ef F R du = ------------K S + ---2 where Rdu： dynamic ultimate bearing capacity of pile (kN) ef： efficiency of hammer, set at 0.5 in case of equation (4.5.4) S： penetration length of pile (cm) drop hammers: mean penetration per strike for the final 5 ~ 10 strikes (m) other hammers: mean penetration per strike for the final 10 ~ 20 strikes (m) K： rebound (m) F： impact energy (kJ) Drop hammer: F = WH H Single action steam hammer: Double action steam hammer: F = ( ap + WH )H F = 2WH H Diesel hammer: H： drop height of hammer (m) WH： weight of hammer (kN) a： cross-sectional area of cylinder (m2) p： steam pressure, air pressure (kN/m2) The allowable bearing capacity Rda is obtained by dividing Rdu by a safety factor of 3. Thus, 1 R da = --- R du 3

(4.5.5)

Pile

Pencil

Metal clamp Pencil

Elastic compression of pile and ground (K)

Penetration of pile (S)

Fig. T- 4.5.2 Rebound Measurement

4.5.2 Design of Joints between Piles and Structure Joints between piles and structures shall be designed to ensure safety against the stresses arising in them.

-307-


[Commentary] It is not easy to decide whether joints between pile heads and the structure should be designed as rigid joints or hinge joints, because each joint type has its advantages and disadvantages. In actual designs, judgements should be based on a firm understanding of the characteristics of each, as described below: (1) There is no difference between a rigid joint and a hinge joint as far as the response to vertical loads is concerned. When buckling must be considered, the rigid joint is preferable because the buckling length of rigid joint is shorter. (2) When subject to a horizontal load, the horizontal displacement of the pile head with rigid joints is much smaller than that with hinge joint. (3) When a horizontal load is applied to a pile with rigid joint, a fixed end moment occurs at the pile head. (4) In case of rigid joints, the pile head moment changes owing to the rotation of the structure. (5) In dynamic analysis of joints, analysis of hinge joints is simpler and clearer than that of rigid joints. But in normal construction it is difficult to use hinge structures for all joints between pile heads and the structure.

4.5.3 Joints of Piles (1) Joints of piles shall be sufficiently safe against loads after completion as well as during construction. (2) Joints shall be placed at the position where there is a sufficient margin in cross-sectional strength and relatively free from corrosion. [Commentary] Depending on the position of joints, the forces acting on joints after completion of a structure are sometimes far smaller than the total strength of the piles. However, considerations should be taken to ensure the safety of joints against the pile-driving stress during construction, load increases in future, and unexpected stresses arising within the cross section of joints.

4.5.4 Change of Plate Thickness or Materials of Steel Pipe Piles When changing plate thickness or material of steel pipe piles, all due considerations shall be given to the workability and the distribution of section force of piles. [Commentary] The section force of steel pipe piles varies with depth, generally decreasing as the penetrated depth of piles becomes large. Therefore, plate thickness or material of steel pipe piles is sometimes changed from economical point of view. [Technical Notes] (1) When changing plate thickness or material of steel pipe piles, the position of the change should be at the depth where the section force arising in the piles does not increase. Caution is also required because such a change may not be allowed if a large negative skin friction is active. (2) Jointing piles with different thickness and materials should be done by shop circular welding. The shape of the welded section should comply with JIS A 5525.

4.5.5 Other Points for Caution in Design [Technical Notes] Generally, there is no risk of buckling during pile driving if the impact stress is below the yield stress of steel piles. Kishida and Takano proposed equation (4.5.6) to express the influence of plate thickness on yield stress. s py t (4.5.6) -------- = 0.69 + 2.2 æ - ö è rø sy where spy： sy： t： r：

yield stress of steel piles taking account of plate thickness (kN/m2) yield stress of steel piles against static load (kN/m2) plate thickness of pipe pile (mm) radius of pipe pile (mm)

In either case, steel pipe piles outside the range shown in Fig. T- 4.5.3 should not be used. If there is a risk of buckling, the steel pipe should be reinforced with steel bands or thicker piles should be used.

-308-

PART V FOUNDATIONS

Range of use of steel pipe piles

Cut-away pipes resistance

Electric welded pipes UOE Roller-bending Spiral pipes Electric

welded resistance pipes

Electric resistance welded pipes

t 2r

Fig. T- 4.5.3 Relationship between Buckling Stress (smax) and Ratio of Sheet Thickness to Diameter ( ----- )

[References] 1) Shoichi KITAJIMA, Shusaku KAKIZAKI, Yoshio HANAKI, Toshiko TAHARA: “On the axial bearing capacity of single Piles”, Tech. Note of PHRI, No. 36, 1967 (in Japanese). 2) Kunio TAKAHASHI: “Behavior of single piles in subsiding ground”, Tech. Note of PHRI, No. 533, 1985, 17p. (in Japanese). 3) Terzaghi, K., Peck, R. B., and Mesri, G.: “Soil Mechanics in Engineering Practice (Third Edition)”, John Wiley, 1995, pp. 435-436. 4) Peck, R. B., Hanson, W. E., and Thornburn, T. H.: “Foundation Engineering”, John Wiley, 1953, p. 260. 5) Transportalion Tech. Res. Inst. and Yahata Steel: “Study on Lateral Resistance of H-shaped Piles”, 1963, pp.345-353 (in Japanese). 6) Koichi KUBO: “A new method for the estimation of lateral resistance of piles”, Rept. of PHRI, Vol. 2, No. 3, 1964, pp. 1-37 (in Japanese). 7) Masatoshi SAWAGUCHI: “Soil constants for piles”, Rept. of PHRI, Vol. 7, No. 2, 1968, p. 551 (in Japanese). 8) Terzaghi, K.: “Evaluation of coefficient of subgrade reaction”, Geotechnique, Vol. 5, No. 4, 1955, pp. 316-319. 9) Yukimitu YOKOYAMA: “Calculation Methods and Examples for Pile Structures”, SANKAIDO, 1977, pp. 193-197 (in Japanese). 10) Yoshinori AOKI: “Design of coupled batter piles subject to horizontal forces”, Tsuchi-to-Kiso, Vol. 18, No. 8, 1970, pp. 2732 (in Japanese). 11) Muneaki SEGAWA, Toyohiro UCHIDA, Takeo KATAYAMA: “Design of coupled batter piles (part 2) - two batter piles hinged at their tops -”, Tech. Note of PHRI, No. 110, 1970, pp. 1-14 (in Japanese). 12) Masatoshi SAWAGUCHI: “Experimental investigation on the horizontal resistance of coupled piles”, Rept. of PHRI, Vol. 9, No. 1, 1970, pp. 11-13 (in Japanese). 13) Hoityker T.,: “Design of Pile Foundations”, SYOKOKUSYA, 1978, pp. 37-41. (Translated in Japanese). 14) Chellis, R. D.: “Pile Foundations”, McGraw-Hill, 1961, pp. 29-32.s

-309-


Chapter 5 ySettlement of Foundations 5.1 Stress in Soil Mass The stress induced in a soil mass due to load on a foundation shall be estimated by assuming that the soil mass be an elastic body. However, for uniformly distributed load, the stress in the soil mass may be estimated by simply assuming that the stress disperse linearly with depth. [Commentary] When a structure built on the ground has a sufficient margin of safety factor against shear failure of subsoils, the stress distribution in the soil mass can be rationally approximated by assuming the soil mass to be an elastic body. The elastic solution obtained by Boussinesq is commonly used in calculation of stress distribution in a soil mass. Boussinesq’s solution is based on the case that a vertical concentrated load acts on the surface of an isotropic and homogeneous semi-infinite elastic body. By superposing this solution, it is possible to calculate the stress distribution in the soil mass when a line load or spatially-distributed load acts on the ground surface. In addition to this elastic solution, the method of Koegler that assumes the stress to disperse linearly with depth can be used for estimating the stress in the soil mass when a strip load or a rectangular load acts on the ground.1)

5.2 Immediate Settlement Immediate settlement shall be estimated with the theory of elasticity by using an appropriate value of the elastic modulus of the ground. [Commentary] Immediate settlement, unlike consolidation settlement that will be described later, is caused by shear deformation and occurs at the instance of loading. The immediate settlement of sandy ground can be considered as the total settlement since there is no long-term consolidation settlement. On the other hand, the immediate settlement of clayey ground is caused by the settlement by undrained shear deformation and by the lateral movement by plastic flow. In the case of soft clayey ground, the immediate settlement may be ignored sometimes in design because it is smaller than the consolidation settlement. In order to calculate the immediate settlement, the ground is usually assumed to be an elastic body, and the theory of elasticity with elastisc modulus E and Poisson’s ratio v is used. As the elastic modulus of soil is greatly dependent on the strain level of soil, it is important to use the elastic modulus that corresponds to the actual strain level taking place in the soil mass. For example, the strains induced in the soft ground with a small margin of safety factor against shear failure are around 0.5% to 1.5%, while those in the examination of deformation of the stiff ground under excavation or in the case of deformation of foundation are below 0.1%. The relationship between the strain level and the elastic modulus is described in Part II, 11.3.1 Elastic Constants.

5.3 Consolidation Settlement Settlements of foundations that are caused by consolidation of subsoil shall be examined in accordance with the procedures described in Part II, 11.3.2 Consolidation Properties. Design parameters for the subsoils shall be determined by using an appropriate method based on the results of consolidation tests. [Commentary] Calculations of settlements due to consolidation are based on the results of consolidation tests on undisturbed samples of cohesive soils. The final consolidation settlement, which is the amount of settlement when consolidation caused by a load has finally completed, is determined by the compressibility properties of the soil skeleton, and can be estimated directly from the results of consolidation tests. Time-dependent changes in settlement up to the final consolidation settlement of a foundation are calculated based on the theory of consolidation. [Technical Notes] (1) Calculation of Final Consolidation Settlement of Foundation Final consolidation settlement of foundation is calculated by using the method described in PartⅡ Ⅱ, 11.3.2 Consolidation Properties. (2) Calculation of Time-Settlement Relationship The rate of consolidation settlement is calculated from the relationship between the average degree of consolidation U and the time factor T that is obtained from Terzaghi’s consolidation theory, where the dissipation of excess pore water pressure is expressed as a partial differential equation of thermal conductivity type. The amount of settlement S(t) at a given time t can be calculated from the average degree of consolidation U(t) by the following equation: S(t) = S0 U(t)

(5.3.1) -310-

PART V FOUNDATIONS

The finite element analysis with visco-elasto-plasticity model for cohesive soil can be utilized for accurate analysis of the consolidation settlement that takes account of inhomogeneity on consolidation properties of the subsoil, the effect of self weight of cohesive soil layer, time-related changes in consolidation load, and others. (3) Division of Cohesive Soil Layer Subject to Consolidation When calculating the final consolidation settlement, the cohesive soil layer is usually divided into a number of sub-layers as shown in Fig. T- 5.3.1. This is because the consolidation pressure Ds z and the coefficient of volume compressibility mv vary with depth. With the mv method, the final consolidation settlement of foundation may be calculated using equation (5.3.2). (5.3.2) S 0 = Sm v DhDs z where S0： final consolidation settlement (m) Ds z： increase in consolidation pressure at the center of a sub-layer (kN/m2) mv： coefficient of volume compressibility for the consolidation pressure equal to (s z 0 + Ds z /2), (m2/kN) where s z 0 is the effective overburden pressure at the center of a sub-layer before consolidation Δh： thickness of a sub-layer in the consolidated layer (m)

Fig. T- 5.3.1 Calculation of Consolidation Settlement

Since mv and Ds z generally decrease with depth, the compression in each sub-layer becomes smaller as the depth increases. The thickness of sub-layer Δh is usually set at 3 ~ 5 m. It should be noted that the consolidation settlement of soft cohesive soil will be underestimated when Δh is taken too large, because the value of mv of the surface layer is very large and it governs the total settlement. The increment of consolidation pressure Ds z is calculated at the center of each sub-layer using the vertical stress distribution with depth, which is described in 5.1 Stress in Soil Mass. The term Ds z is the increment in vertical stress due to loading. In the natural ground, it is usually assumed that consolidation due to the existing overburden pressure has completely finished. Although the distribution of subgrade reaction at the bottom of foundation is not same as that of the acting load due to the rigidity of foundation, the rigid foundation settles unifomly and the stress distribution of subsoil at a certain depth becomes irrelevant to the distribution of reaction immediately below the foundation bottom. Therefore, the vertical stress distribution in the subsoil beneath a rigid foundation can be estimated by considering only the load distribution acting on the foundation bottom. (4) Vertical Coefficient of Consolidation cv and Horizontal Coefficient of Consolidation ch When pore water of ground flows vertically during consolidation, the vertical coefficient of consolidation cv is used. But when vertical drains are installed, drained water of ground flows mainly to the horizontal direction and the horizontal coefficient of consolidation ch should be used. The value of ch obtained from experiments on the clay in Japanese port areas ranges from 1.0 to 2.0 times the value of cv2). However, in design ch ≒ cv is acceptable when considering a decrease in ch due to disturbance caused by installation of vertical drains, inhomogeneous consolidation properties in the ground, and others. (5) Coefficient of Consolidation cv of Overconsolidated Clay 3) The coefficient of consolidation of clay in overcosolidated state is generally larger than that in normally consolidated state. When the clay seems to be clearly in overconsolidated state, the value of cv used for design should be the one at the mean consolidation pressure between the existing effective overburden pressure and the final pressure after consolidation. However, rather than simply calculating cv at the mean consolidation pressure, it would be better to determine a weighted mean value of cv considering the settlement based on the e-log p curve obtained by consolidation tests. (6) Rate of Consolidation Settlement in Inhomogeneous Ground When the ground consists of alternate layers with different cv values, the rate of consolidation settlement is analyzed using the equivalent-thickness method 4) or numerical analysis such as the finite difference method 5) or the finite element method 6) , 7). The equivalent-thickness method is used as a simplified method, but it -311-


sometimes yields significant errors. When the ground is inhomogenous to a large extent or when accurate estimation is required, it is recommended to use the finite element method. (7) Settlement Due to Secondary Consolidation The shape of the settlement - time curve in long-term consolidation tests on cohesive soil is consistent with Terzaghi’s consolidation theory up to the degree of consolidation of around 80%. When the consolidation passes this level, the settlement increases linearly with logarithm of time. This is due to the secondary consolidation that arises with the time-dependent properties of soil skeleton under consolidation load, beside the primary consolidation that causes the settlement accompanying dissipation of excess pore water pressure induced in the cohesive soil due to consolidation load. The settlement due to secondary consolidation is particularly significant in peat and other organic soils. In ordinary alluvial clay layers, the consolidation pressure caused by loading is often several times greater than the consolidation yield stress of the subsoil. Under such conditions, the settlement due to secondary consolidation is smaller than that due to the primary consolidation, and is not significant in design. But when the consolidation pressure acting on the ground due to loading does not greatly exceed consolidation yield stress, the settlement due to secondary consolidation tends to continue over a long time, even though the settlement due to primary consolidation may be small. In this case, the secondary consolidation settlement must be fully taken into account in design. The settlement due to secondary consolidation may be generally calculated using the following equation: Ca (5.3.3) S s = -------------- h log 10 ( t ¤ t 0 ) 1 + e0 where Ss： settlement due to secondary consolidation (m) C a： coefficient of secondary compression t ： time (d) t0： start time of secondary consolidation (d) h： clay layer thickness (m) ·

The coefficient of secondary compression C a is obtained from conventional consolidation tests. It can also be estimated from the relationship between C a and the compression index Cc that is generally expressed in the following equation 8): (5.3.4)

C a = (0.03 ～ 0.05) Cc

5.4 Lateral Displacement In quaywalls or revetments constructed on soft clayey ground, countermeasures shall be taken when lateral displacements due to shear deformation of the ground have an adverse effect on structures. [Commentary] (1) In quaywalls or revetments on soft ground, it is sometimes necessary to estimate lateral displacements caused by shear deformation of the ground. The lateral displacements take two forms. The first one is the displacement accompanying the settlement occurring immediately after loading. The second one is the displacement that occurs gradually with time thereafter. If a load is significantly smaller than the ultimate bearing capacity of the ground, the lateral displacement accompanying the immediate settlement can be predicted by assuming the ground to be an elastic body. (2) A common problem with soft ground is the case that the safety factor of the whole ground is set relatively low at around 1.3, and lateral displacements occur as a combination of consolidation and creep deformation by shear stress of the ground. In order to predict whether this kind of lateral displacement will occur or not, a practical method based on past experiences 9) has been proposed. A computer program of finite element method that applies elasto-plasticity or visco-elasto-plasticity models for clayey ground has been used to predict the progresses of settlement and latreral displacement with time in detail. An appropriate method of calculation should be chosen in consideration of the importance of the structure under lateral displacement.

5.5 Differential Settlements When constructing structures on a soft clayey ground, differential settlements of the ground shall be taken into account and appropriate countermeasures shall be taken when differential settlements have an adverse effect on structures. [Technical Notes] A simplified method has been proposed for estimating differential settlements in reclaimed lands in port areas. It classifies the reclaimed lands into the following four types: -312-

PART V FOUNDATIONS

(1) Extremely inhomogeneous ground (2) Inhomogeneous ground (3) Homogeneous ground (4) Extremely homogeneous ground Figure T- 5.5.1 shows the mean differential settlement ratios for each type of reclaimed land. The raios of the mean differential settlements between two arbitrarily chosen points to the mean total settlement for reclaimed land. For example, the mean differential settlement ratio for the distance between two points separated by 50 meters in inhomogeneous ground is 0.12. Thus, when a total settlement of x cm occurs at a certain time from the beginning, the differential settlement occurring on average in the distance of 50 meters is calculated as 0.12x. When applying this method to actual design, the values given in Fig. T- 5.5.1 should be corrected with respect to the time period and the depth of the layer to be consolidated 10).

Mean differential settlement ratio

0.5

Extremely inhomogeneous ground Inhomogeneous ground Homogeneous ground Extremely homogeneous ground

0.4

0.3

0.2

0.1

0 0

20

50

100

Distance between two points (m)

Fig. T- 5.5.1 Relationship between Mean Distance and Mean Differential Settlement Ratio in Reclaimed Lands

[References] 1) Koichi AKAI: “Bearing Capacity and Settlement of Soils”, Sankai-do, 1964 (in Japanese). 2) Masaki KOBAYASHI, Junichi MIZUKAMI, Takashi TSUCHIDA: “Estimation method of horizontal coefficient of consolidation for cohesive soils”, Rept. of PHRI Vol. 29, No. 2, 1990 (in Japanese). 3) Akio NAKASE, Masaki KOBAYASHI, Akio KANECHIKA: “Consolidation parameters of overconsolidated clays”, Rept. of PHRI, Vol. 12, No. 1, 1973, pp. 123-139 (in Japanese). 4) L. A. Palmer and P. P. Brown: “Settlement analysis for areas of continuing subsidence”, Proc. 4th. Int. Conf. S. M. F. E., Vol. 1, 1957, pp. 395-398 5) R. L. Schiffman and R. E. Gibson: “Consolidation of nonhomogeneous clay layers”, Journal of S. M. F. E., ASCE, Vol. 90, No. SM5, 1964, pp. 1-30 6) Masaki KOBAYASHI: “Numerical analysis of one-dimensional consolidation problems”, Rept. of PHRI, Vol. 21, No. 1, 1982 (in Japanese). 7) Masaki KOBAYASHI: “Finite element analysis of the effectiveness of sand drains”, Rept. of PHRI, Vol. 30, No. 2, 1991 (in Japanese). 8) Mesri, G.: “Coefficient of secondary compression”, Proc. ASCE, Vol. 99, No. SM1, 1973, pp. 123-137 9) Yasuo KASUGAI, Ken-ichiro MINAMI, Hiroyuki TANAKA: “The prediction of the lateral flow of port and harbour structures”, Tech. Note of PHRI, No. 726, 1992 (in Japanese). 10) Takashi TSUCHIDA, Kenji ONO: “Evaluation of differential settlements with numerical simulation and its application to airport pavement design”, Rept. of PHRI, Vol. 27, No. 4, 1988 (in Japanese).

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Chapter 6 Stability of Slopes 6.1 General (1) Stability of slopes against slip failure caused by self weight of soil and/or surcharge shall be analyzed as a two-dimensional problem, assuming a circular arc slip surface or a planar slip surface. (2) Slope stability analysis shall be carried out on the condition that the slope becomes least stable. [Commentary] (1) General Slope stability analysis is to calculate a safety factor at a limit equilibrium state when a soil mass of slope becomes unstable due to self weight of soil and/or surcharge. Calculation methods used in slope stability analysis may also be used to study bearing capacity of foundations, because they are intended to examine the stability of soil mass. (2) Shapes of Slip Surface (a) Types of shapes of slip surfaces Theoretically speaking, shapes of slip surfaces in slope stability analysis are combinations of planar, logarithmic spiral, and/or circular arc shapes 1). In practice, however, planar or circular arc slip surfaces are assumed. When there is a particularly weak layer and a slip surface is expected to pass over it, that slip surface or other appropriate slip surfaces may sometimes be assumed. An assumed slip surface in general should be the one along which the slip of the soil mass smoothly takes place. Thus, a slip surface with sharp bends or curves that seems to be kinematically unnatural should not be used. (b) Slope failure of sandy ground Slip failure of dry or saturated sand slope usually takes a mode that the slope inclination decreases by collapsing. Thus, its shape should be considered as a planar plane rather than a circular arc. Even when assuming a circular arc slip surface, its form is close to a planar one. The inclination of sandy slope in equilibrium is termed as the angle of repose, which is equivalent to an internal friction angle that corresponds to the void ratio of the sand in the slope. The angle of repose of unsaturated sand, having apparent cohesion due to surface tension of pore water in the sand, becomes larger than that of dry or saturated sands. (c) Slope failure of clayey ground The actual slip failure surface of clayey ground is close to a circular arc, and a deep slip called the base failure often takes place, while a shallow slip appears near the surface layer in sandy slope. Slope stability analysis is usually dealt as a two-dimensional problem. Although actual slip surfaces in slopes with long extention take the form of three-dimensional curved surfaces, a two dimensional analysis gives a solution on the safer side. When the stability is expected to decrease due to surcharge over a finite extention, however, the lateral resistance of both the ends of a cylindrical failure surface may be taken into account. (3) External Forces in Slope Stability Analysis Important causes of slip failures are self weight of soil, surcharge, water pressure and others. Beside them, cyclic loads such as seismic force, wave force, and others may be included. Resistance against the slip is given by shear resistance of soil and counter weight. Because the shear strength of soil is related with time, the stability problems on soil mass are classified into two cases; loading on the ground in normally consolidated state, and unloading by excavation etc. The former is refered to as a short-term stability problem and the latter a long-term one. In either case the shear strength that is suitably evaluated should be used (see Part II, 11.3.3 Shear Properties). [Technical Notes] A safety factor in slope stability problems is the ratio of the shear strength of soil to the shear stress arising on an assumed slip surface. As the value of safety factor differs from an assumed slip surface to another, the safety factor against slip failure of the slope is the smallest of those obtained by assuming different slip surfaces under the given conditions. As a standard, a safety factor more than 1.30 for ordinary conditions should be maintained in slope stability analyse for design puposes. However, a safety factor between 1.10 and 1.30 may be acceptable under the condition that the design parameters are thought to be highly reliable based on the actual performance in the similar subsoil conditions, and that construction works are carried out with detailed monitoring of displacements and stresses of the ground 2).

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PART V FOUNDATIONS

6.2 Stability Analysis 6.2.1 Stability Analysis Using Circular Slip Surface Method (Notification Article 44) Depending on the ground characteristics, slope stability shall be examined by circular arc analysis using the modified Fellenius method as shown in the following equation, or else by an appropriate method in accordance with 2.5 Bearing Capacity for Eccentric and Inclined Loads. In the analysis as described above, the safety factor shall maintain an appropriate value taking account of the characteristics of the ground and structure. R S ( cl + W ¢ cos a tan f ) F s = -------------------------------------------------------S Wx + S Ha (6.2.1) S ( cb + W ¢ cos 2 a tan f ) sec a = ------------------------------------------------------------------1 S W sin a + ---S Ha R where Fs： safety factor against circular slip failure according to the modified Fellenius method R： radius of slip circle (m) c： undrained shear strength of cohesive soil, or apparent cohesion of sandy soil in drained condition (kN/m2) l： length of base of a segment (m) W'： effective weight of a segment per unit length (the sum of soil self weight and surcharge; for submerged part, use the submerged unit weight) (kN/m) a： angle of segment base to the horizontal (º) f： 0 for cohesive soil, or angle of internal friction of sandy soil in drained condition (º) W： total weight of a segment per unit length (the sum of soil self weight, water weight and surcharge) (kN/m) x： horizontal distance between the center of gravity of a segment and the center of slip circle (m) H： horizontal external force acting on a soil mass inside a slip circle (water pressure, seismic force, wave force and others) (kN/m) a： arm length of horizontal external force H with respect to the center of a slip circle (m) b： width of segment (m) [Commentary] In slope stability analysis, slip failure may be caused by self weight of soil, surcharge, water pressure, wave pressure, seismic force and others. Resistance to slip failure is the shear resistance of soil, counter weight and others. A safety factor against slip failure of slope is defined as the ratio of the shear strength of soil to the shear stress arising on an assumed slip surface. When assuming a circular slip surface, it is calculated as the ratio of resisting moment to overturning moment around the center of the slip circle. In the slice method used for the circular arc analysis, the soil mass inside the slip circle is divided into a number of segments by vertical planes. The shear stress and resistant stress of soil on the base of each segment are then calculated. Summing up the stresses all over the segments, a safety factor is obtained. For the slice method, the statically determinate condition has to be assumed in order to solve for the forces in equilibrium between the segments. Various methods have been proposed depending on the assumption used. Generally the modified Fellenius method and the simplified Bishop method are used in many cases. [Technical Notes] (1) Stability Analysis Method Using Modified Fellenius Method 1), 3) Various calculation methods have been proposed for the slice method. They vary in their assumptions on the forces acting on the vertical planes between the segments. The modified Fellenius method assumes that the direction of resultant force acting on vertical planes between the segments is parallel to the base of the segments. This method is also referred to as the simplified method or Tschebotarioff method. When a circular arc and a segment are as shown in Fig. T6.2.1, a safety factor according to the modified Fellenius method can be obtained using equation (6.2.1). For conducting slope stability analysis, the center of a slip circle is set first of all. Of the slip circles that Fig. T- 6.2.1 Circular Arc Analysis Using Modified take this point as their center, the one that gives the Fellenius Method -315-


minimum safety factor is sought for. This safety factor then becomes the safety factor for that center point. The safety factors for the other centers are then found using the same method. The minimum value obtained via contours for safety factors is defined as the safety factor against slip failure of the slope. (2) Stability Analysis Method Using Bishop Method 4) Bishop proposed an equation for calculating a safety factor taking account of vertical shear force and horizontal force acting on the vertical planes of segments. In actual calculations, a commonly used method assumes vertical shear forces to be in equilibrium. This is termed the simplified Bishop method. The safety factor by the simplified Bishop method Fs is calculated using equation (6.2.2). 1

( cb + W¢ tan f ) × sec a 1 + ( tan a tan f ) ¤ F s

Fs = ------------------------------------------- S ----------------------------------------------------

1 SW sin a + ---SHa R

(6.2.2)

where F s： safety factor against circular slip failure according to the simplified Bishop method R： radius of slip circle (m) W'： effective weight of a segment per unit length (the sum of soil self weight and surcharge; for submerged part, use the submerged unit weight) (kN/m) W： total weight of a segment per unit length (the sum of soil self weight, water weight and surcharge) (kN/m) b： width of segment (m) a： angle of segment base to the horizontal (º) c： undrained shear strength of cohesive soil, or apparent cohesion of sandy soil in drained condition (kN/m2) f： 0 for cohesive soil; angle of shear resistance of sandy soil in drained condition (º) H： horizontal external force acting on a soil mass inside a slip circle (water pressure, seismic force, wave force and others) (kN/m) a： arm length of horizontal external force H with respect to the center of a slip circle (m) (3) Applicability of Stability Analysis Methods 5) The safety factors obtained from the modified Fellenius method and the simplified Bishop method are in agreement for cohesive soil in which f = 0, while they differ when the circular arc passes through sandy soil layers. The modified Fellenius method is commonly used for circular arc analysis both in Japan and abroad. This is because it has been reported that the modified Fellenius method reasonably explains the actual behaviors of slope failure based on the results of stability analysis on case histories of slip failures in port and harbor areas in Japan 3), and also because it gives a safety factor on the safer side in sandy soil layers. However, when the foundation ground consists of only sandy soil layers, or when a circular arc cuts through the ground consisting of a thick sandy layer at the top and a cohesive layer at the bottom, it has been confirmed that the modified Fellenius method underestimates the safety factor. The simplified Bishop method gives solutions of higher accuracy under such conditions, being judged from the viewpoint of the basic principles of stability calculation method. Thus the simplified Bishop method should be used particularly for bearing capacity problems of rubble mounds, in which an eccentric and inclined loads are applied (see 2.5 Bearing Capacity for Eccentric and Inclined Loads). The simplified Bishop method has a shortcoming that it overestimates a safety factor when a surcharge is applied vertically on sandy soil layer that lies almost horizontally. For such a case, a stability calculation method may be carried out whereby the ratio of the vertical to the horizontal forces between the segments is assumed to be 1/3.5 of the angle of segment base to the horizontal. The safety factor in this calculation is evaluated using the following equation: ì ü ( ncb + W¢ × tan f ) sec a 1 (6.2.3) F = ------------------------------------------- S í --------------------------------------------------------------------------------- ý n + [ tan a tan ( ba ) ] × ( tan f ) ¤ F 1 þ SW sin a + ---SHa î R where n = 1 + tan a • tan (ba). The parameter b defines the ratio of the vertical to the horizontal forces acting on the vertical planes of segments, and taken as b = 1/3.5. The other symbols are the same as in equation (6.2.2).

6.2.2 Stability Analysis Assuming Slip Surfaces Other Than Circular Arc Slip Surface Despite the provisions stated in the previous sections, a planar or a compounded slip surface shall be assumed in stability analysis when it is more appropriate to assume a slip surface other than a circular arc slip surfaces according to the ground conditions. [Technical Notes] When assuming a planar slip surface, a safety factor against slip failure of slope is calculated using the following equation: S { cl + ( W¢ cos a H sin a ) tan f } (6.2.4) F = --------------------------------------------------------------------------------sin aSW¢ + cos aSH -316-

PART V FOUNDATIONS

where F： c： f： l： W'：

safety factor against planar slip failure cohesion of soil (kN/m2) internal friction angle of soil (º) length of a segment base(m) effective weight of a segment per unit length (for submerged part, use the submerged unit weight) (kN/m) Fig. T- 6.2.2 Slope Stability Analysis Using a： inclination of segment base (taking positive in the case Planar Slip Surface shown in Fig. T- 6.2.2 ) (º) H： horizontal external force per unit length applied to a slope (water pressure, seismic force, wave pressure and others) (kN/m)

The minimum safety factor against slip failure should be 1.2 under ordinary conditions and 1.0 during an earthquake. [References] 1) R. F. Scott: “Principles of soil mechanics”, Addison Wesley, 1972, p. 431. 2) Takashi TSUCHIDA, Tang YIXIN: “The optimum safety factor for stability analyses of harbor structures by use of the circular arc ship method”, Rept. of PHRI, Vol. 35, No. 1, 1996 (in Japanese). 3) Akio NAKASE: “The f = 0 analysis of stability and unconfined compressive strength”, Soils and Foundations, Vol. 7, No. 2, 1967, pp33-50. 4) A. W. Bishop: “The use of the slip circle in the stability analysis of slopes”, Geotechnique, Vol. 5 No. 1, 1955, pp. 7-17. 5) Kenji NOMURA, Yoshinobu HAYAFUJI, Fumiaki NAGATOMO: “Comparison between Bishop’s method and Tschebotarioff's method in slope stability analysis”, Rept. of PHRI, Vol. 7, No. 4, 1968, pp. 133-175 (in Japanese).

-317-


Chapter 7 Soil Improvement Methods 7.1 General When carrying out soil improvement as a countermeasure against possible failures of soft ground, an appropriate method shall be selected in view of the characteristics of foundation subsoil, type and scale of structure, ease and period of construction, economic factors, influence on the environment, and others.

7.2 Replacement Method Design for replacement method shall be conducted by considering stability against circular slip failure, settlement of subsoil, and workability of replacement. [Commentary] Replacement methods are classified into the removal and replacement method, and the forced displacement method. The latter includes the forced displacements by self weight of embankment, by blasting, and by the sand compaction pile method1). [Technical Notes] This section deals with the removal and replacement method, which is commonly used in marine construction works. (1) Stability Analysis Stability calculation method and safety factors for cilcular arc analysis are described in Chapter 6 Stability of Slopes. To calculate earth pressure on sheet piles or anchorage inside the replaced area, an analysis incorporating a compounded slip surface should be made in addition to conventional earth pressure calculation. When the whole soft layer is replaced but the bearing stratum underlying the replaced area is inclined, the stability against a compounded slip failure including the slip failure along the bearing plane should be examined. (2) Examination of Settlement When cohesive soil is left beneath the replaced sandy soil (in cases of partial replacement or slope excavation), the consolidation settlement may be expected, and thus its influence on a structure should be examined. (3) Replaced Sand and Angle of Internal Friction The angle of internal friction of replaced sand can usually be assumed to be around 30 degrees. However, its value depends on the particle size, size distribution, execution procedure and sequence of replacement, elapsing time, surcharge and others. Attention must be paid to the condition of the replaced sand, because it is sometimes very loose. (4) Examination of Liquefaction Liquefaction should be assessed based on the particle size distribution and the N-values of the replaced sand. When being difficult to evaluate, the liquefaction should be examined by cyclic triaxial test (see Part Ⅱ, Chapter 13 Liquefaction). When liguefaction is the governing facter in design, it should be considered from the stage of selecting the replaced material. If a sufficient strength in terms of the N-value cannot be obtained, the replaced sand should be compacted. (5) The replaced sand should have an appropriate particle size distribution with low content of silt and clay. Usually the fines content is kept to be less than 15%. (6) The N-values of the replaced sand are affected by its particle size and size distribution, execution procedure and sequence of replacement, elapsing time, surcharge and others. According to some case histories, the N-values of the replaced sand were around 10 when instantaneously dumped in large volume from large-capacity hopper barges with bottom doors, around 5 when discharged with grab buckets from sand carriers, and even smaller values when spreaded by pump dredger. Some case histories show that the N-values of the loose replaced sand increased with the application of surcharge and the elapsing time (after discharging the replaced sand, after rubble deposition, after placing caisson and others).

7.3 Vertical Drain Method 7.3.1 Principle of Design Design for vertical drain method shall be conducted by considering the strength increment to be required, the allowable settlement for structure, the area and depth to be improved, and others.

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PART V FOUNDATIONS

7.3.2 Determination of Height and Width of Fill [1] Height and Width of Fill Required for Soil Improvement The height and width of fill shall be determined by considering the strength increment of soil required for stability of sutructures to be constructed as well as the allowable settlement, influence on the surrounding area, and others. [Commentary] It is desirable to set the fill top width larger than the width required for soil improvement (see Fig. C- 7.3.1).

Mean width

fill top width (fill)

[2] Height and Width of Fill Required for Stability of Fill Embankment

improved area

Stability of fill embankment itself shall be confirmed by a circular arc analysis or other appropriate methods for the dimensions of height and width of fill determined.

(permeable layer)

[Technical Notes]

Fig. C- 7.3.1 Fill Width for Vertical Drain Method

The strength increment of soil and the settlement due to fill may be determined using equations (7.3.1) and (7.3.2). Dc = ( Dc ¤ D p ) ( ag ¢h p c )U

(7.3.1)

S = m v ( ag ¢h p c )HU (7.3.2) where h： fill height (m) H： thickness of clay layer (m) mv ： coefficient of volume compressibility (m2/kN) pc ： preconsolidation pressure (kN/m2) S ： settlement (m) U ： degree of consolidation a ： coefficient of stress distribution (ratio of vertical stress distributed inside subsoil to fill pressure) g ' ： effective unit weight of fill material (kN/m3) Δc： increment of undrained shear strength (kN/m2) Δc /Δp： rate of strength increase Since surcharge is usually applied in several stages in the vertical drain method, the degree of consolidation U to be substituted in equations (7.3.1) and (7.3.2) differs at each surcharge stage. However, strength increment may often be calculated by assuming a uniform degree of consolidation of approximately 80%.

7.3.3 Design of Drain Piles In designing drain piles, consolidation process shall be calculated by taking account of drain pile interval, drain pile diameter, and drainage conditions at the top and bottom of clay layer as well as the characteristics of drain materials and sand mat, and thickness of sand mat. [1] Drain Piles and Sand Mat Drain piles and sand mat shall have the required drainage capacity. [Technical Notes] (1) Consolidation Rate and Drain Pile Diameter Consolidation rate is approximately proportional to drain pile diameter and inversely proportional to the square of drain pile interval. Generally, the amount of drain pile materials can be reduced by using sand piles of small diameter with a small interval rather than using those of large diameter with a large interval. However, the small diameter sand pile may suffer from clogging with clayey particles and breaking failure of the pile caused by deformation due to surcharge and consolidation. According to case histories, a diameter of approximately 40 cm is prevaling, while the diameter generally varies between 30 and 50 cm. The fabri-packed drain method, in which sand piles wrapped by geotextile with a diameter of 12 cm are drived in, has often been used for extremely soft subsoil in land construction works. Usually, a set of four sand piles are installed simultaneously -319-


by a small size pile driving machine. In marine works, fabri-packed drains with a diameter of 40 cm or more have been commonly used to improve extremely soft subsoil. (2) Sand Used for Sand Piles Sand used for sand piles should have high permeability as well as suitable particle sizes to prevent clogging with clayey particles. The particle size distribution of sand used in past works are shown in Fig. T- 7.3.1. Sand with slightly higher fines content have also been used in recent years. Fine sand

Coarse sand

Gravel

Passing weight percentage (%)

Silt

Japan New York

Particle size Fig. T- 7.3.1 Examples of Sand Used in Sand Piles

(3) Materials for Plastic-Board Drain In lieu of sand piles, strip type drains with materials such as plastic-board drains are sometimes employed. In the design of the strip type drains, the strip is converted to a sand pile with a diameter that produces the perimeter equal to that of the strip type drain in order to apply the conventional design method. By setting the safety factor at a value higher than that of the conventional method, the design has been conducted by assuming that a strip type drain of 10 cm wide and 5 mm thick is equivalent to a 5 cm-diameter sand pile. When the drainage capacity is low, it should be considered that a possibe delay of consolidation could occur especially at the tip of the vertical drain: i.e, at the bottom of the consolidation layer. (4) Sand Mat Thickness of sand mat is usually set to be approximately 1.0 m to 1.5 m for marine works and 0.5 m to 1.0 m for land works. A thick layer of sand mat may cause troubles in drain pile driving, while a thin layer of sand mat may lose its permeability partially due to infiltration of clayey particles. When the drainage capacity of sand mat is low, a delay in consolidation may occur due to the head loss within drains. In such a case, the delay in consolidation is more noticeable around the center of the area spreaded with the sand mat. Thus, the sand mat materials must have high permeability. For a case that the delay is anticipated because of low permeability of sand mat or large area of soil improvement work, an approximate solution may be employed to evaluate the delay.

[2] Interval of Drain Piles Interval of drain piles shall be so determined that the required degree of consolidation can be obtained during a given construction period. [Technical Notes] (1) General The vertical drain method is applied when the rate of one-dimensional consolidation by the preloading method is too small under the constraint of time limit of the construction period. Figure T- 7.3.2 shows the relationship between the required consolidation time t80 (day), drainage distance H (m), and coefficient of consolidation cv (cm2/min), which is calculated for the condition of 80% consolidation of a clayey layer for the preloading method or the vacuum consolidation method without vertical drains. (2) Determination of Interval of Drain Piles The interval of drain piles should be determined by means of Fig. T- 7.3.3 and equation (7.3.3). If the interval of drain piles is too small, consolidation may be delayed due to the effect of smear (the disturbance of clay layer by the process of drain pile driving) and others 2).

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PART V FOUNDATIONS

10 years

5 years 4 years 3 years

t

2 years 80

(d) 1 year

6 months

Permeable layer

Clay Permeable layer Permeable layer Clay Im-permeable layer

H (m) Fig. T- 7.3.2 Required Days for 80% Consolidation of Clay Layer

D where D: b : n: De : Dw : T h¢ : cvh : t:

= bnD w

(7.3.3)

interval of drain piles (cm) factor (b = 0.886 for square grid pattern, and b = 0.952 for triangular grid pattern) diameter ratio of De/Dw (n is read off from Fig. T- 7.3.3) effective diameter of drain area (cm) diameter of drain pile (cm) parameter similar to time factor (Th¢ = cvh・t/Dw2) horizontal coefficient of consolidation (cm2/min) consolidation time (min)

Note: the time t used in Figs. T- 7.3.3 and 7.3.4 is expressed in the units of days. 1 100

50

nn

Uh =

n = DDwe T'h = Dcvh2w t

5

0.

6 7 0. 0. 0.8 0.9 95 0.

10

5

cvh ; (cm /min) t ; (d) Dw ; (cm) 2

1 0.0001

0.001

0.01

T h'

0.1

1

Fig. T- 7.3.3 Calculation Chart for n

(3) Calculation of the Degree of Consolidation After determining the interval of drain piles, the accurate value of the degree of consolidation Uh is calculated by equations (7.3.4) and (7.3.5) and Fig. T- 7.3.4. -321-


c vh t T h = --------2De

(7.3.4)

De n = ------Dw

(7.3.5)

where Th : cvh : t: De : Dw :

time factor of horizontal consolidation horizontal coefficient of consolidation (cm2/min) elapsed time from start of consolidation (min) effective diameter of drain area (cm) diameter of drain pile (cm)

cvh ; (cm /min) t ; (d) De ; (m) 2

50

2 n= n=

0

n=3

n=2

n = 16 n = 12 n = 10 n= 8 n= 7 n= 6 n= 5 n= 4

Fig. T- 7.3.4 Horizontal Consolidation Calculation Chart

(4) Effective Diameter The effective diameter of drain area De is the diameter of an equivalent circle that has the same area as the soil being drained by a sand pile. The relationship between De and interval of the drain pile D is as follows: De = 1.128D for square grid pattern. De = 1.050D for equilateral triangular grid pattern.

7.4 Deep Mixing Method 7.4.1 Principle of Design [1] Scope of Application (1) The design method described in this section shall be applied to improvement of subsoil beneath gravity type structures such as breakwaters, quaywalls, and revetments. (2) The design method shall be applied to the block type and the wall type improvement works. [Commentary] (1) The deep mixing method dealt with in this section is the one in which the soil in situ is mixed mechanically with cement. (2) Large-scaled soil improvement with the deep mixing method in port and harbor construction works have been mostly applied to the subsoils for caisson type breakwaters, quaywalls, or revetments. There are few cases of application to other types of structures. Thus, the sope of this section is set as specified above. (3) When applying the deep mixing method to port and harbor structures, a rigid underground structure is formed by overlapping stabilized soil columns treated by mixing machines. The pattern of improvement is selected depending on the superstructure or the properties of the subsoil. The block type and the wall type shown in Fig. C- 7.4.1 are typical improvement patterns in port and harbor construction works. In this section, these two patterns are discussed. -322-

PART V FOUNDATIONS

(4) The wall type improvement consists of long and short walls as shown in Fig. C- 7.4.1 (b). The basic concept of the design is that the long walls function to transfer the external forces to the stiff layer, while the short walls function to increase the uniformity of the improved ground. Sea surface

Sea surface

Soft subsoil Soft subsoil Sea surface

Sea surface

Sea bed

Sea bed Soft subsoil

Improved subsoil Improved subsoil

Soft subsoil

Long wall Short wall

(a) block type improvement

(b) wall type improvement

Fig. C- 7.4.1 Typical Improvement Patterns in the Deep Mixing Method

[2] Basic Concept Design for subsoil improvement by the deep mixing method shall be conducted by examining external stability, internal stability, and displacement of the stabilized body. [Commentary] (1) Terminology Definitions of principlal terms are described below: Stabilized soil:

soil treated by the deep mixing method.

Stabilized body:

a sort of underground structure consisting of the stabilized soil (in the wall type improvement, this includes original soil between the long walls).

Improved subsoil system: a system consisting of the improved subsoil and the upper structure between the vertial planes passing through the toe and heel of the improved subsoil. External stability:

examination of stability when the stabilized body and the upper structure as a whole behave as a rigid body.

Internal stability:

examination of internal failure within the stabilized body under the condition that the external safety stability is secured.

Fixed type:

subsoil improvement is carried out for the whole soft layer so that the stabilized body rests on the bearing layer and the external forces are directly transferred there.

Floating type:

subsoil improvement is stopped midway of the soft layer so that the stabilized body does not reach the bearing layer but rests on top of the soft subsoil.

(2) Generally, the stabilized soil by the deep mixing method has extremely high strength and elastic modulus, and extremely small strain at failure compared with those of untreated soil3). Therefore, a block of stabilized soil can be considered as a sort of underground structure. For this reason, the stability analysis shall be conducted with respect to the stability as a whole structure (external stability), the strength of the stabilized body itself (internal stability), and the settlement, horizontal displacement and overturning of the stabilized body as a rigid body. [Technical Notes] In the design of the deep mixing method, the “Technical Manual for the Deep Mixing Method in Marine Construction Works” and the “Technical Manual for the Deep Mixing Method in Land Construction Works” may be referred to.

7.4.2 Assumptions for Dimensions of Stabilized Body [1] Mixture Design of Stabilized Soil The mixing proportions of soil and cement for stabilized soil shall be determined from laboratory mixing tests or in-situ tests. -323-


[2] Allowable Stress of Stabilized Body Allowable stress of the stabilized body shall be appropriately determined for the examination of internal stability. [Technical Notes] (1) The allowable compressive stress s ca is expressed by equation (7.4.1) based on the unconfined compressive strength. 1 s ca = ---abg q uf F where sca : allowable compressive stress (kN/m2) F : safety factor for material a : factor for effective cross-sectional area of stabilized soil column b : reliability factor of overlapping g : correction factor for scattered strength q uf : average unconfined compressive strength of in-situ stabilized soil (kN/m2)

(7.4.1)

(2) The induced stresses within the stabilized body should be lower than the allowable compressive stress sca, the allowable shear stress t a and the allowable tensile stress sta, which are given by the following equations: 1 t a = --- s ca 2

(7.4.2)

s ta = 0.15s ca ≦ 200 kN ¤ m

2

(7.4.3)

(3) In the design, the stabilized body is assumed to be a rigid structure with homogeneous strength. In the field, however, the inhomogeneous body may be formed, depending on the operation of construction machines and the method to overlap the stabilized columns to form the stabilized body. The factor of safety and other factors shown in equation (7.4.1) are the reduction factors to deal with inhomogeneity of the improved subsoil so as to assume the subsoil as homogeneous with respect to strength. (a) Safety factor for material (F) Since the allowable compressive stress sca is based on the unconfined compressive strength, an appropriate value for safety factor should be determined by incorporating influence of creep and cyclic loading, importance of the structure, types of loading, design calculation method and reliability of the materials. In past works, the safety factor was often taken as 3.0 for ordinary conditions and 2.0 for seismic conditions. (b) Factor of effective cross-sectional area of stabilized soil column (a) When stabilized by a deep mixing (DM) machine with several mixing blades, the cross section of the stabilized body consists of a group of circles, as shown in Fig. T- 7.4.1. In the block type and wall type improvements, the stabilized columns are overlapped with each other to form the stabilized body as shown in Fig. T- 7.4.2. In there two cases, the length of connected portion along the overlapping line is smaller than those along other lines. The factor of effective cross-sectional area of stabilized soil column a is so defined to correct for this untreated portion.

Dx

R D

lx

width of

d overlapping

Dy

connecting surface

l L

lx

Fig. T- 7.4.2 Connecting Surfaces

Fig. T- 7.4.1 Specific Effective Width of Deep Mixing Machine

(c) Reliability factor of overlapping (b) An overlapped part is made by connecting a fresh stabilized column with an existing one. Therefore, accuracy of execution and reduction in strength should be taken into account. The reliability coefficient of overlapping is the ratio of the strength of overlapped part to that of the stabilized soil. It depends on time interval of overlapping, mixing capacity of the machine, supplying method of stabilizing agent and others. The factor b may be set approximately between 0.8 and 0.9. (d) Correction factor for scattered strength ( g ) The strength of in-situ stabilized soil shows large scattering in terms of unconfined compressive strength. The correction factor for scattered strength takes this scatter of strength into account. -324-

PART V FOUNDATIONS

(4) Strength of In-Situ and Laboratory Stabilized Soils. The average unconfined compressive strength of in-situ stabilized soil q uf can be related with the average unconfined compressive strength of laboratory stabilized soil q ul using equation (7.4.4) (7.4.4) q uf = l q ul The correction factor l may be taken as l = 1 in many marine works, but it is recommended to determine its value through in-situ tests. (5) The values of the factors used in design should be determined by considering the situation that they are not independent of each other but are closely related. According to the past works, the ratio of the allowable compressive stress in situ to the laboratory strength was often set between 1/6 and 1/10, considering all factors including l .

7.4.3 Calculation of External Forces External forces acting on the stabilized body shall be appropriately determined for external stability, internal stability and displacements of the stabilized body, respectively. [Technical Notes] (1) Figure T- 7.4.3 shows a schematic diagram of external forces acting on the stabilized body 4) in the case of a gravity type quaywall. (2) Since the wall type stabilized body contains untreated soil, the external forces should be determined individually on untreated and stabilized soils depending on the item of examination. (3) For the analysis on external stability, Pa and Pp should be determined as the active and passive earth pressures respectively as described in Part Ⅱ, Chapter 14 Earth Pressure and Water Pressure. For the analysis on internal stability, Pa represents the active earth pressure, while Pp should be appropriately determined within the range of passive earth pressure to earth pressure at rest, considering the external stability. (4) It has been confirmed in the experiments that adhesion acts on the vertical planes of the active and passive sides of the stabilized body. In the case of backfilling, negative skin friction due to consolidation settlement of untreated soil acts downwards on the vertical plane of the active side of the stabilized body. Thus, these adhesions described above may be taken into account for ordinary conditions 5). On the other hand, for seismic conditions, Pav and Ppv may be assumed to act in the favorable directions for analysis of both external and internal stability, because the safe side design consideration has been made such that both seismic mass force of the stabilized body and the seismic earth pressure act simultaneously during an earthquake.

Pav Pa

Pp

' Ppv '

In case of wall type stabilization

Fig. T- 7.4.3 External Forces Acting on Stabilized Body -325-

Pw


The symbols used in Fig. T- 7.4.3 are as follows: Pa： Pav： Pw ： Pp： Ppv： Pdw： R： T： t1, t2： W1： HK1： W2 ： HK2 ： W3 ： HK3 : W4： HK4：

resultant earth pressure per unit length acting on the vertical plane of the active side (kN/m) resultant vertical adhesion per unit length acting on the vertical plane of the active side (kN/m) resultant residual water pressure per unit length (kN/m) resultant earth pressure per unit length acting on the vertical plane of the passive side (kN/m) resultant vertical adhesion per unit length acting on the vertical plane of the passive side (kN/m) resultant dynamic water pressure during an earthquake (kN/m) shear resistance per unit length acting on the bottom of stabilized body (kN/m) reaction force per unit length acting on the bottom of stabilized body (kN/m) reaction pressure per unit length at toe and heel of stabilized body, respectively (kN/m) weight per unit length of upper structure (surcharge, overburden, upper structure, and others) (kN/m) seismic inertia force per unit length acting on structure (kN/m) weight per unit length of rubble stones (kN/m) seismic inertia force per unit length acting on rubble stones (kN/m) weight per unit length of backfill (including surcharge) (kN/m) seismic inertia force per unit length acting on backfill (kN/m) weight per unit length of stabilized body (kN/m) seismic inertia force per unit length acting on stabilized body (kN/m)

The following external forces are considered only in the case of wall type improvement: W5¢： effective weight per unit length of untreated soil in between long walls (kN/m) HK5¢： seismic inertia force per unit length acting on untreated soil in between long walls (kN/m) For evaluation of external forces for the liquefied subsoil during an earthquake, the dynamic water pressure during an earthquake should be considered. The evaluation of dynamic water pressure is described in Part Ⅱ　 , 14.4.2 Dynamic Water Pressure during Earthquake.

7.5 Lightweight Treated Soil Method 7.5.1 Outline of Lightweight Treated Soil Method The provisions in this section shall be applied to the design for the lightweight treated soil method. [Commentary] The lightweight treated soil method is to produce artificial lightweight and stable subsoil by adding lightening materials and hardening agents to slurry-state earth of dredged soil or excavated soil from construction works, and then using the mixture as materials for landfill or backfilling. When using air foam as the lightening material, it is named the foam treated soil, and when using EPS (expanded polistyrol) beads, it is named the beads treated soil. The lightweight treated soil has the following characteristics: (1) The weight is approximately one half of ordinary sand in the air and approximately one fifth in the seawater. This lightness can prevent or reduce ground subsidence due to landfill or backfill. (2) Due to its light weight and high strength, the earth pressure during an earthquake is reduced. This enables to create highly earthquake-resistant structures or reclaimed lands. (3) It uses dredged spoils, which are regularly produced and treated as waste in ports and harbors, or waste soils that are generated by land-based construction works. Thus, employment of the lightweight treated soil can reduce the amount of waste materials to be dealt with at waste disposal sites. [Technical Notes] Refer to the “Technical Manual for the Lightweight Treated Soil Method in Ports, Harbors and Airports” for further details on the design based on this method.

7.5.2 Basic Design Concept Lightweight treated soil, which is a light and stabilized geotechnical material, shall comply with the design methods for earth structures described in Chapter 2 Bearing Capacity of Shallow Foundations to Chapter 6 Stability of Slopes. [Commentary] Apart from mixing tests, the design method for lightweight treated soil is basically the same with that for other earth structures6).

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PART V FOUNDATIONS

[Technical Notes] The properties of lightweight treated soil should be evaluated by means of laboratory tests that take account of the environmental and construction conditions of the site. They may be evaluated as follows: (1) Unit Weight The unit weight g t may be set within a range of 6 to 15 kN/m3 by adjusting the amount of lightening material and water. When used in port and harbor facilities, casting work may be difficult and the risk of flotation may be anticipated in the case of the rise of sea water level if the unit weight is less than that of seawater. Therefore, the following values should ordinarily be used for the design unit weight: Below the water level: g t = 12 kN/m3 In the air: g t = 10 kN/m3 In the mixture design the fact should take into account that the unit weight of lightweight treated soil greatly depends on the environmental conditions particularly on the intensity of hydrostatic pressure7), 8), during and after casting. (2) Shear Strength9) The static strength of lightweight treated soil mainly owes to the solidified strength yielded by cement-based hardeners. The lightweight treated soil is generally used in the strength range of 100 ~ 500 kN/m2 in terms of unconfined compressive strength qu. Because of inclusion of air foam or EPS beads, no increase in strength can be expected with respect to an increase in confining pressure, but the residual strength is approximately 70% of the peak strength. The safety factor F defined as the ratio of laboratory strength to design strength may usually be set between 2 and 3. (3) Consolidation Yield Stress The consolidation yield stress py may be calculated using the following equation: py = 1.4 qu

(7.5.1)

(4) Coefficient of Friction For analysis on external stability, assuming the lightweight treated soil to be a rigid soil mass, the following values may generally be adopted as the coefficient of friction between the lightweight treated soil and sand (or rubble): Lightweight treated soil - sand: Lightweight treated soil - rubble:

μ = 0.55 ~ 0.60 μ = 0.75 ~ 0.80

(5) Deformation Modulus E50 When tests are conducted for the samples manufactured with great cares such as accurate finishing at both ends of them and detailed measurements of small deformations, the secant modulus E50 from the test may be used as an deformation modulus in design. When such tests are not possible, the modulus may be estimated from the unconfined compressive strength qu using the following equation: E50 = (100 ~ 200) qu

(7.5.2)

The deformation modulus shown above corresponds to a strain level of 0.05% ~ 0.2%. (6) Poisson’s Ratio v Poisson’s ratio v of lightweight treated soil varies depending on the stress level and the state before or after the attainment of peak strength. When the surcharge is less than the consolidation yield stress of treated soil, the following mean values may be used: Foamed treated soil: Beads treated soil:

v = 0.10 v = 0.15

(7) Dynamic Properties The shear modulus G, damping factor h, strain dependency of G and h, and Poisson’s ratio v used in dynamic analysis should be obtained from laboratory tests. They may be estimated from a correlation between unconfined compressive strength and the results of ultrasonic propagation test as a simplified method. When calculating earth pressure of lightweight treated soil under ordinary and seismic conditions, the method described in 7.8 Active Earth Pressure of Solidified Geotechnical Material should be applied.

7.5.3 Mixture Design of Treated Soil (1) Mixture design shall be conducted to obtain the strength and the unit weight required in the field. (2) Types of hardening and lightening agents shall be determined after their efficiency has been confirmed in tests.

-327-


7.5.4 Examination of Area to Be Treated The area to be filled with the lightweight treated soil shall be determined as appropriate in view of the type of structure to be built there and the conditions of external forces as well as the stability of the structure and the subsoil as a whole. [Commentary] The extent of filling area with lightweight treated soil usually depends on the intended degree of lightweightness. That is, when the method is applied to control settlement or lateral displacement, it is determined from the allowable conditions for settlement or displacement; to secure stability, it is determined from the condition of slope stability; to reduce earth pressure, it is determined from the required conditions for earth pressure reduction10).

7.5.5 Workability Verification Tests When no experience of construction is available as reference, or when the execution is to be conducted under special conditions, verification tests shall be conducted to confirm workability before starting execution.

7.6 Replacement Method with Granulated Blast Furnace Slag 7.6.1 Principle of Design When using granulated blast furnace slag as backfill for quaywalls or revetments, landfill, or surface covering for soft subsoil, the characteristics of the materials shall be duly taken into account. [Commentary] Granulated blast furnace slag is a granular material having latent hydraulic property that is not seen in natural sand, and it solidifies and stabilizes with time. When comparing its granular state with solidified state, the granular state generally presents less stable design conditions. In some occasion, however, the solidified state may become critical for stability of structure. An adequate examination should be carried out for each condition.

7.6.2 Physical Properties of Granulated Blast Furnace Slag When using granulated blast furnace slag, its physical properties shall be ascertained in advance. [Commentary] Granulated blast furnace slag is in a state like coarse sand when shipped from plants. The important characteristics of physical properties are its small unit weight and dormant hydraulicity. [Technical Notes] (1) Grain Size Distribution The grain size distribution of granulated blast furnace slag is generally within the range listed in Table T- 7.6.1. The grain size of general granulated blast furnace slag is 4.75 mm or less and its fine content is extremely small. The grain size distribution is poorly graded or almost single-grained, mostly in the range of coarse sand, with a uniformity coefficient of 2.5 to 4.2 and a coefficient of curvature of 0.9 to 1.4. Table T- 7.6.1 General Grain Size Distribution of Granulated Blast Furnace Slag Passing weight percentage (%) 4.75 mm

2.00 mm

0.85 mm

0.425 mm

0.250 mm

0.106 mm

0.075 mm

99~100

80~100

30~75

7~30

3~15

1~6

0~4

(2) Unit Weight Granulated blast furnace slag is lighter than natural sand because it has air bubbles inside its particles and has a large void ratio owing to angular shape and single-grained particle size. According to past records, the wet unit weight of granulated blast furnace slag ranges from 8 to 13 kN/m3 and the submerged unit weight is approximately 7 kN/m3. (3) Permeability The coefficient of permeability in a granular state depends on the void ratio. Generally, it ranges from 1 × 10-1 to 1×100cm/s. The coefficient of permeability decreases with solidification down to approximately 1×10-2cm/s 11). (4) Compressibility The time-dependent change of compressibility of granulated blast furnace slag used for backfill, landfill, or surface covering can be ignored. -328-

PART V FOUNDATIONS

(5) Internal Friction Angle and Cohesion In a granular state, cohesion can be ignored. The angle of internal friction in this state is 35 degrees or greater. The shear strength increases by its solidification. (6) Liquefaction during an Earthquake When granulated blast furnace slag is used in backfill, it solidifies in many cases in about 6 months because of its latent hydraulic property. When solidification can be expected, liquefaction can be ignored. However, there is a risk of liquefaction for granulated blast furnace slag that has not yet solidified. Therefore in this case, the possibility of liquefaction should be examined, treating the granulated blast furnace slag as a granular material.

7.7 Premixing Method 7.7.1 Principle of Design [1] Scope of Application The design method described in this section shall be applied to the subsoil treated by the premixing method aimed at reducing earth pressure and preventing liquefaction. [Commentary] (1) The meaning of the main terms used for this method are as follows: Treated soil:

soil improved by stabilizer.

Treated subsoil:

subsoil improved by filling with treated soil.

Area of improvement:

area of treated subsoil by the premixing method.

Stabilizer content:

weight ratio of stabilizer to dry weight of parent material (soil), expressed in percentage.

Reduction of earth pressure: measures designed to reduce earth pressure against wall (active earth pressure) under ordinary and seismic conditions. (2) In the premixing method, stabilizer and antisegregation agent are added into soil, mixed in advance and used as landfill materials. The treated soil thus produced is then placed beneath the water to create a stable subsoil. The subsoil improvement is materialized in principle that cement-based stabilizers add cohesion to the soil used in landfill by means of chemical solidification action between soil and stabilizer. This method can be applied to backfill behind quaywalls and revetments, filling of cellular-bulkhead, replacement after base excavation, refilling, etc. (3) Soils applicable to the treatment are sand and sandy soils, excluding clay. This is because the mechanical properties of the treated soil differ considerably depending on the parent material. (4) Besides reducing earth pressure and preventing liquefaction, this method can also be used to improve the soil strength necessary for construction of facilities on reclaimed lands. In this case, the strength of treated ground should be assessed appropriately. [Technical Notes] Refer to the “Technical Manual for Premixing Method” published by the Coastal Development Institute of Technology for details of other matters concerning the design and construction using the premixing method.

[2] Consideration for Design (1) In designing the subsoil treatment by the premixing method, the required strength of treated soil, and the stabilizer content, and the area to be treated shall be determined as appropriate. (2) When evaluating the earth pressure reduction effect or examining the stability of the subsoil against circular slip failure and others, the treated soil should be regarded as a “c - f material”. (3) The treated subsoil may be thought to slide as a rigid body during an earthquake because the treated subsoil has a rigidity considerably greater than that of the surrounding untreated subsoil. Therefore, when determining the area of improvement, the stability against sliding failure of the subsoil including upper structures (total stability) shall also be examined. 7.7.2 Preliminary Survey The characteristics of soil used in the premixing method shall be evaluated as appropriate by preliminary surveys and tests. -329-


[Commentary] (1) Preliminary surveys and tests should include soil tests on particle density, water content, grain size distribution, maximum and minimum densities of soils to be used for filling, and surveys on records of soil properties and field tests of existing reclaimed ground nearby. (2) The water content and the fines content of soils for filling have influence on manifestation of strength of treated soil after mixing and selection of the method to mix soils with stabilizer. (3) The density of treated soil after landfilling must be adequately estimated in advance, since the density, has a major effect on the test results in laboratory mixing tests.

7.7.3 Determination of Strength of Treated Soil The strength of treated soil shall be determined in such a way to yield the required improvement effects, by taking account of the purpose and conditions of application of this method. [Commentary] (1) For the purpose to reduce the earth pressure, the cohesion c of treated soil should be determined such that the earth pressure is reduced to the required value. (2) For the purpose to prevent liquefaction, the strength of treated soil should be determined such that the treated soil will not liquefy. [Technical Notes] (1) There is a significant relationship between the liquefaction strength and the unconfined compressive strength of treated soils. It is reported that treated soils with the unconfined compressive strength of 100 kN/m2 or more will not liquefy. Therefore, when aiming to prevent liquefaction, the unconfined compressive strength should be set at 100 kN/m2. When the unconfined compressive strength of treated soil is set at less than 100 kN/m2, cyclic triaxial tests should be conducted to confirm that the soil will not liquefy. (2) In determing the cohesion of treated soil, the internal friction angle f of treated soil is first estimated. Then, the cohesion is determined by reverse calculation using an earth pressure calculation formula that takes account of cohesion and internal friction angle with the target reduced earth pressure and the estimated internal friction angle f. (3) According to the results of consolidated and drained triaxial compression tests on treated soil with stabilizer content of less than 10%, the internal friction angle of treated soil is equal to or slightly larger than that of untreated soil. Therefore, the internal friction angle of the treated soil may be assumed to be the same as that of the untreated soil to keep a design on the safer side. (4) The internal friction angle may be obtained from consolidated and drained triaxial compression tests based on the presumed density and effective overburden pressure of the subsoil after landfilling. The internal friction angle f used in design is usually set at the value by 5 to 10 degrees smaller than that obtained from the tests. Without conducting triaxial tests, it may be obtained from the estimated N-value of the subsoil after landfilling (note that the N-value of untreated subsoil should be used).

7.7.4 Mixture Design of Treated Soil Mix proportioning of treated soil shall be determined by conducting appropriate laboratory mixing tests. A reduction of strength shall be taken into account because the in-situ strength may be lower than the strength obtained from laboratory mixing tests. [Commentary] The purpose of laboratory mixing tests is to obtain the relationship between the strength of treated soil and the stabilizer content, and to determine the stabilizer content so as to obtain the required strength of treated soil. The relationship between the strength of treated soil and the stabilizer content is greatly affected by the soil type and the density of soil. Therefore, test conditions of laboratory mixing tests should be as similar to field conditions as possible. [Technical Notes] (1) For the purpose to reduce earth pressure, consolidated and drained triaxial compression tests should be carried out to obtain the relationship among the cohesion c, the internal friction angle f, and the stabilizer content. For the purpose to prevent liquefaction, unconfined compression tests should be conducted to obtain the relationship between the unconfined compressive strength and the stabilizer content. (2) It is important to grasp the difference between in-situ and laboratory strengths when setting the increase factor for mixture design in the field. According to the past experiences, the laboratory strength is larger than the in-330-

PART V FOUNDATIONS

situ strength, and the increase factor of a = 1.1 to 2.2 is used. Here, the increase factor a is defined as the ratio of the laboratory to the field strengths in terms of unconfined compressive strength.

7.7.5 Examination of Area of Improvement The area to be improved by the premixing method shall be determined as appropriate in view of the type of structure to be constructed and the magnitudes of external forces as well as the stability of subsoil and structures as a whole. [Commentary] (1) For the purpose to reduce earth pressure, the area of improvement should be determined in such a way that the earth pressure of treated subsoil acting on a structure should be small enough to guarantee stability of the structure. (2) For the purpose to prevent liquefaction, the area of improvement should be determined in such a way that liquefaction in the adjacent untreated subsoil will not affect the stability of structure. [Technical Notes] (1) Figure T- 7.7.1 shows the external and resisting forces acting on a structure and the treated subsoil when the untreated subsoil behind the treated subsoil is not liquefied, while Fig. T-7.7.2 shows the external forces in the case of liquefaction. (2) For either purpose to reduce earth pressure or to prevent liquefaction, sliding during an earthquake should be examined for the treated subsoil and the structure. Also, stability against circular slip failure should be examined for the subsoil and the structure as a whole. (a) Examination of sliding during an earthquake Sliding during an earthquake is to be examined because of a risk that the treated subsoil may slide as a rigid body. A safety factor against sliding during an earthquake in this case should be equal to or greater than 1.0. The friction coefficient at the bottom of treated subsoil should be 0.6. When the layer beneath treated subsoil is cohesive soil, however, the cohesion of soil in that layer should be used in evaluation of sliding resistance at the bottom of treated subsoil. The resultant earth pressure in equation (7.7.1), which is used to calculate a safety factor for sliding in the case that the untreated subsoil is not liquefied, represents a simple case in which the residual water level is at the ground surface. When the untreated subsoil behind the treated subsoil is liquefied, the fluid pressure from the liquefied subsoil acts upwards on the treated subsoil if the configuration of the treated subsoil is such that as shown in Fig. T- 7.7.1. This reduces the effective weight of the treated subsoil. Thus, the shape of treated subsoil shown in Fig. T- 7.7.1 is disadvantageous to sliding failure compared with that shown in Fig. T- 7.7.2. When the residual water level is below the ground surface and the untreated subsoil is liquefied, the soil layer above the residual water level up to the ground surface should also be treated as liquefied due to propagation of excess pore water pressure from the lower layer. ① When the untreated subsoil is not liquefied By defining the positive directions of external and resisting forces as shown in Fig. T- 7.7.1, the safety factor F s for sliding can be calculated using equation (7.7.1). R1 + R2 + Pw1 (7.7.1) F s = ---------------------------------------------------------------H1 + H2 + Ph + Pw2 + Pw3 where P w 1 = ( 1 ¤ 2 )g w h 1

2

P w 2 = ( 7 ¤ 12 )kg w h 1 P w 3 = ( 1 ¤ 2 )g w h 2

2

2

H 1 = kW 1 H 2 = kW 2 2

P h = ( 1 ¤ 2 )K a g ¢h 2 cos ( d + y ) ¤ cos y P v = P h tan ( d + y ) R1 = f1 W 1 ¢ R2 = f2 ( W 2 ¢ Pv )

(when the layer beneath the treated subsoil is sandy soil)

R 2 = cl bc

(when the layer beneath the treated subsoil is cohesive soil) -331-


g w： g ¢： k： Ka： h1： h2： d： y： f1： f2： c： lbc：

unit weight of seawater (kN/m3) submerged unit weight of untreated soil (kN/m3) seismic coefficient coefficient of active earth pressure of untreated subsoil during an earthquake height of water level from the sea bottom at the front of the structure (m) height of residual water level from the sea bottom (m) (residual water level in Fig. T- 7.2.1 is taken at the subsoil surface for simplicity) angle of wall friction between treated and untreated subsoils (cd) (°) angle of the back of treated subsoil (cd) to the vertical (°) (positive for anticlockwise; y in Fig. T- 7.7.1 is negative) friction coefficient at the bottom of structure friction coefficient at the bottom of treated subsoil (= 0.6) cohesion of the cohesive layer beneath the treated subsoil (kN/m2) bottom length of treated subsoil (bc) (m)

Structure

Treated subsoil

(7.2.1)

Untreated subsoil (not liquefied)

H1 : seismic force acting on structure (abef) (kN/m)

H2 : seismic force acting on treated subsoil (bcde) (kN/m)

Pw1 : static water pressure acting on the front of structure (af) (kN/m)

Pw2 : dynamic water pressure acting on the front of structure (af) (kN/m)

Pw3 : static water pressure acting on the back of treated subsoil (cd) (kN/m)

Ph : horizontal component of resultant force by active earth pressure during an earthquake of the untreated subsoil acting on the back of treated subsoil (cd) (kN/m)

Pv : vertical component of resultant farce by active earth pressure during an earthquake of the untreated subsoil acting on the back of treated subsoil (cd) (kN/m)

W1, W2 : weight of structure (abef) and treated subsoil (bcde) (not taking account of buoyancy due to static water pressure) (kN/m) W1', W2' : effective weight of structure (abef) and treated subsoil (bcde) (taking account of buoyancy) (kN/m)

R1 : friction resistance at the bottom of structure (ab) (kN/m)

R2 : friction resistance at the bottom of treated subsoil (bc) (kN/m)

Fig. T- 7.7.1 External Forces Acting When Untreated Subsoil Is Not Liquefied (Reduction of Earth Pressure) ② When the untreated subsoil is liquefied By defining the positive directions of external and resisting forces as shown in Fig. T- 7.7.2, the safety factor F s for sliding can be calculated using equation (7.7.2). When the untreated subsoil behind the treated subsoil is liquefied, the static and dynamic pressures from the untreated subsoil should be exerted on the back of the treated subsoil as shown in Fig. T- 7.7.2. The static pressure should be calculated by adding the static water pressure to the earth pressure with the earth pressure coefficient of 1.0. The dynamic pressure should be calculated using equations (14.4.1) and (14.4.2) in Part II, 14.4.2 Dynamic Water Pressure during Earthquake. The unit weight of water in equations (14.4.1) and (14.4.2) should be replaced by the saturated unit weight of soil. R1 + R2 + P w1 (7.7.2) F s = -----------------------------------------------H 1 + H2 + Ph + Pw2 where P w 1 = ( 1 ¤ 2 )g w h 1

2

P w 2 = ( 7 ¤ 12 )kg w h 1

2

H 1 = kW 1 H 2 = kW 2 2

2

P h = ( 1 ¤ 2 )gh 2 + ( 7 ¤ 12 )kg h 2 P v = P h tan y R1 = f1 W 1 ¢

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PART V FOUNDATIONS

2

R 2 = f 2 { W 2 ¢ + [ P v ( 1 ¤ 2 )g w h 2 tan y ] } (when the layer beneath the treated subsoil is sandy soil) (when the layer beneath the treated subsoil is cohesive soil) R 2 = cl bc y： angle of the back of treated ground (cd) to the vertical (º) (positive sfor anticlockwise; y in Fig. T- 7.7.2 is positive) The other symbols are the same as ① .

Structure

Treated subsoil

Untreated subsoil (liquefied)

Static pressure Dynamic pressure (earth + water) (earth + water)

H1 : seismic force acting on structure (abef) (kN/m)

H2 : seismic force acting on treated subsoil (bcde) (kN/m)

Pw1 : static water pressure acting on the front of structure (af) (kN/m)

Pw2 : dynamic water pressure acting on the front of structure (af) (kN/m)

Ph : horizontal component of the resultant force by static and dynamic pressure of the liquefied subsoil acting on the back of treated subsoil (cd) (kN/m)

Pv : vertical component of the resultant force by static and dynamic pressure of the liquefied subsoil acting on the back of treated subsoil (cd) (kN/m) W1, W2 : weight of structure (abef) and treated subsoil (bcde) (not taking account of buoyancy) (kN/m) W1', W2' : effective weight of structure (abef) and treated subsoil (bcde) (taking account of buoyancy due to static water pressure) (kN/m)

R1 : friction resistance at the bottom of structure (ab) (kN/m)

R2 : friction resistance at the bottom of treated subsoil (bc) (kN/m)

T- 7.7.2 External Forces Acting When Untreated Subsoil Is Liquefied (Prevention of Liquefaction) Fig. (b) Examination of stability against circular slip failure The stability against circular slip failure should be examined according to 6.2.1 Stability Analysis Using Circular Slip Surface Method.

7.8 Active Earth Pressure of Solidified Geotechnical Materials 7.8.1 Scope of Application [Technical Notes] This section describes design principles for calculation of active earth pressure when using geotechnical materials solidified by stabilizers such as cement as backfill materials. Solidified geotechnical materials considered in this section include those that harden naturally and others that are hardened artificially by adding cement or other stabilizer. Materials developed to date are listed below. The variety of materials is certain to increase in future. (1) Premixed soil (treated soil by premixing method) (2) Lightweight treated soil (3) Cement-mixed soils other than the above two (4) Solidified coal ash (5) Self-hardening coal ash (6) Granulated blast furnace slag used in a solidified state

7.8.2 Active Earth Pressure [1] Outline [Technical Notes] (1) When using solidified geotechnical materials, the material properties and the characteristics of earthquake motion should be appropriately taken account in calculations of active earth pressure on a structure. (2) When calculating active earth pressure during an earthquake, the seismic coefficient method may generally be used. When detailed examination of earth pressure during an earthquake is required however, response analysis and others must be carried out. Methods to calculate earth pressure using the seismic coefficient method considering material properties are described later in this section. -333-


(3) Generally, when solidified geotechnical materials are thought to have sufficiently large cohesion, liquefaction in the treated area needs not be considered. If the unconfined compressive strength qu is more than around 50 ~ 100 kN/m2, the generation of excess pore water pressure inside the treated subsoil during an earthquake may be ignored though it may depend on the magnitude of seismic forces.

[2] Strength Parameters [Technical Notes] The method of determining strength parameters for geotechnical materials depends on the materials used. The cohesion and angle of internal friction are to be evaluated in accordance with the characteristics of the material used. In general, deep mixed soil (treated soil by the deep mixing method), lightweight treated soil, and solidified coal ash are assumed to be cohesive materials. Premixed soil is thought to be a material with both the cohesion and internal friction. Granulated blast furnace slag is usually treated as a granular material with the internal friction, but it may also be treated as a cohesive material if the solidification process is positively introduced when used.

[3] Calculation of Active Earth Pressure [Technical Notes] (1) Generally, the earth pressure may be evaluated based on the provisions in Part II, Chapter 14 Earth Pressure and Water Pressure. The principle for calculation of earth pressure may be regarded as the Mononobe-Okabe principle. In this method, the earth pressure is calculated by an equilibrium of forces in accordance with Coulomb’s concept of earth pressure by assuming that the subsoil fails while forming a wedge. (2) Many factors remain unknown about the earth pressure during an earthquake. This is particularly true of earth pressure during an earthquake in submerged subsoils. Nevertheless, the principle of earth pressure in Part II, Chapter 14 Earth Pressure and Water Pressure has so far been adopted in the design of many structures with satisfactory results. (3) Equation (7.8.1), an expansion of the earth pressure equation in Part II, Chapter 14 Earth Pressure and Water Pressure, can be applied to materials having both the cohesion c and internal friction angle f (see Fig. T- 7.8.1).

2z i = y + f i m i + 90° B i C i + A i B i2 A i2 + C i2 m i = tan 1 ---------------------------------------------------------B i2 A i2 A i = sin ( d + b + q ) B i = sin ( y + f i + d b ) × cos q [ sin ( y f i + q ) × cos ( d + b ) ] 2c i × cos ( y b ) × cos f i × cos ( d + b ) × cos q + -------------------------------------------------------------------------------------------------------( Sg i × h 1 ) × cos ( y b ) ------------------------------------------------------ + w 2 cos y C i = sin ( y + f i + d b ) × sin q + [ sin ( y f i + q ) × sin ( d + b ) ]

where pai： ci ： f i： g i： hi： y： b： d： z i：

2c i × cos ( y b ) × cos f i × sin ( d + b ) × cos q ------------------------------------------------------------------------------------------------------( Sg i × h 1 ) × cos ( y b ) ------------------------------------------------------ + w 2 cos y active earth pressure acting on wall by the i-th layer (kN/m2) cohesion of soil in the i-th layer (kN/m2) angle of internal friction in the i-th layer (º) unit weight of the i-th layer (kN/m3) thickness of the i-th layer (m) angle of wall to the vertical (º) angle of ground surface to the horizontal (º) angle of wall friction (º) angle of failure surface of the i-th layer to the horizontal (º) -334-

64444444444744444444448

sin ( z i f i + q ) × cos ( y z i ) ì ( Sg i × h i ) cos ( y b ) ü p ai = í ------------------------------------------------- + w ý × -----------------------------------------------------------------------------------------------× cos ( y z i + f i + d ) × sin ( z i b ) cos y cos q î þ c i × cos ( y b ) × cos f i --------------------------------------------------------------------------cos ( y z i + f i + d ) sin ( z i b )

(7.8.1)

PART V FOUNDATIONS

w： q： k： k ¢：

surcharge per unit area of ground surface (kN/m2) resultant seismic angle (º); q = tan 1 k or q = tan 1 k ¢ seismic coefficient apparent seismic coefficient

Fig. T- 7.8.1 Earth Pressure

(4) Equation (7.8.1) is an extended form of Okabe’s equation12). This extension lacks such rigorousness that Okabe solved the equilibrium of forces. However, when the soil is exclusively granular material with no cohesion or exclusively cohesive material with no angle of internal friction, it is consistent with the equations in Part II, Chapter 14 Earth Pressure and Water Pressure. (5) The earth pressure and the angle of failure surface should be calculated separately at each soil layer with different soil properties, while the earth pressure distribution and the failure line inside each layer are treated as linear. Actually within a soil layer, the earth pressure and the failure line sometimes become curved when calculated for divided sublayers. This contradicts the original assumption in Okabe’s equation that is based on a linear slip on the premise of Coulomb’s earth pressure. (6) When using the equations above, the existence of cracks sometimes has to be considered in accordance with the characteristics of the geotechnical materials used.

[4] Case of Limited Area of Subsoil Improvement [Technical Notes] (1) When the area treated by solidified geotechnical materials is limited and Mononobe-Okabe’s equation cannot be applied simply, the earth pressure is evaluated by a suitable method that allows the influence of the treated area to be assessed. When the treated area is limited, the earth pressure can be evaluated by the slice method. This method of calculating earth pressure assumes a slip surface behind a structure. The soil mass sandwiched between the slip surface and the wall surface is divided into segments by vetical planes, and the earth pressure is calculated from the equilibrium of forces by self weight, buoyancy, shear force along the slip surface, and seismic force of the respective sliced segments. Although the actual failure surface of the subsoil does not agree with this rationale, this method may be used when there is no other suitable method available. (2) The characteristics of earth pressure calculation using the slice method desribed here are as follows: (a) In the semi-infinite multilayered subsoil, the earth pressure calculated by this method almost agrees with that obtained from Part II, Chapter 14 Earth Pressure and Water Pressure. -335-


(b) In the finite multilayered subsoil, the calculated earth pressure is consistent with Mononobe-Okabe’s rationale on earth pressure. (c) When the subsoil includes granular soil layers, the angle of wall friction is set at 15 degrees. When the subsoil consists of only cohesive soil layers, the angle is 0 degrees. (d) The acting position of resultant earth pressure cannot be obtained. Therefore, the following method is used to obtain the acting position: ① The resultant earth pressures are obtained at two points along the wall, the distance between is small enough. ② The intensity of earth pressure between two points is calculated by dividing the difference of two resultant earth pressures by the distance of two points. ③ The procedures in ① and ② are repeated through the whole height of the wall. Thus, the earth pressure distribution will be obtained. ④ The acting position is calculated using the earth pressure distribution. (e) The failure mode when calculating the distribution of earth pressure in the vertical direction is sometimes conflicting with the failure mode for calculating the resultant earth pressure as a whole. In this case, caution is required for determination of the earth pressure distribution used in design. (f) With the slice method, three modes of failure are examined (see Fig. T- 7.8.2). Mode 1: when a uniform slip surface is formed in the whole backfill (shear resistance mode) Mode 2: when a cracks down to the bottom of the solidified soil layer is developed (crack failure mode) Mode 3: when a slip surface is formed along the edge line of the solidified range (friction resistance mode) Note: Among Mode 1, the case in which the slip surface does not pass the solidified body is categorized as Mode 0.

Mode 2

Mode 3

Mode 1

Mode 0

Fig. T- 7.8.2 Three Failure Modes Considered in the Slice Method

(g) The earth pressure distribution is calculated by assuming that the difference between the resultant earth pressures at adjacent depths is the earth pressure for the corresponding depth.

7.9 Sand Compaction Pile Method (for Sandy Subsoil) 7.9.1 Principle of Design Design of sand compaction pile method to densify sandy soils shall be conducted appropriately after examining the characteristics of subsoil properties and construction methods, as well as by taking account of the past construction records and the results of test execution. [Technical Notes] (1) Purpose of Improvement The purpose of improving loose sandy subsoil can be classified into (a) improving liquefaction strength, (b) reducing settlement, and (c) improving the stability of slopes or bearing capacity. (2) Compaction of Deep Layer Vibration or impact from the ground surface is usually not effective to compact a deep sandy layer. The improvement method generally adopted is to drive sand or gravel piles or to drive vibrating rods into a deep loose sandy layer.

7.9.2 Sand Volume to Be Supplied Design of sand volume to be supplied shall be made appropriately, examining the characteristics of the subsoil, the required relative density, and the N-value. [Technical Notes] (1) Sand volume to be supplied can be calculated by equation (7.9.1). e0 e v = -------------1 + e0 -336-

(7.9.1)

PART V FOUNDATIONS

where e： design void ratio of the improved soil (between sand piles) e0： void ratio of the original soil v： volume of supplied sand required per unit volume of the original soil (m3/m3) (2) The N-value after improvement is greatly affected by the N-value of the original subsoil and the volume of sand supplied as well as the particle size distribution of the original soil and the overburden pressure. In particular, the improvement effect might decrease in the subsoil with a high content of fine particles less than 75μm (Fc). Therefore, the reduction ratio b due to fines content should be incorporated in the design based on Fig. T- 7.9.1. (3) Design for compaction of loose sandly subsoil by the vibro-composer method can be conducted based on a number of accumulated case records on this method. Design for other compaction methods may be conducted based on similar data if available. (4) For improving loose sandy subsoil with a fines content of 20% or less, the sand supply rate Fv (= n) can be calculated based on the past records on the N-value of the original subsoil and the required design N-value. The relationship between the N-value of the original subsoil (N0) and that of the improved ground at the sand pile center (Np) is shown in Fig. T- 7.9.2 with the parameter of the sand supply rate Fv. Similarly, the N-value of the improved subsoil at the intermediate position between the sand piles (N1) is shown in Fig. T- 7.9.3. These diagrams are obtained from the accumulated case histories on land-based construction. Determining the N-value of the original subsoil, N0, from site surveys and the design N-value after improvement, N1, from the conditions on upper structure. STEP 1 Estimating emax and emin from the fines content Fc (%) using the following equations:

e max = 0.02F c + 1.0

e min = 0.008F c + 0.6 STEP 2 Calculating the relative density Dr0 and e0 using the following equation by substituting the N-value of the original subsoil, N0, and the confining pressure, s v (kN/m2).

D r0 = 21 100 × N 0 ¤ ( s v + 70 )

(%)

D r0 e 0 = e max --------- ( e max e min ) 100 STEP 3 Calculating the reduction ratio b for the increment of N-value due to fine content. b = 1.0 0.5 log F c ( for Fc > 1.0 ) STEP 4 Calculating the corrrected N-value, N1', from the calculated N-value, N1, without any fine content, considering the reduction ratio b.

N1 ¢ = N 0 + ( N1 N0 ) ¤ b STEP 5 Calculating e1 for N 1 ¢ by substituting N 1 ¢ into the equation in STEP 2 instead of N 0 . STEP 6

e0 e1

Calculating the density increase ratio as by substituting e0 and e1 into a s = ---------------- . 1 + e0 STEP 7

As

2 As 3x

Determining the interval of sand piles x from a s = ----2- or a s = ------- ----2- in which A s is the sectional area

x

of a sand pile. Fig. T- 7.9.1 Design Procedure for Sand Compaction Pile Method -337-

N-value at pile center after improvement N

p


Legend

N-value of original subsoil N-value at pile center after improvement sand supply rate

N-value of original subsoil N

0

between sand piles after improvement

N-value at intermediate position

Fig. T- 7.9.2 Relationship between N-value of Original Subsoil (N0) and N-value at Pile Center after Improvement (Np) (Fines Content ≦ 20%)

Legend

N-value of original subsoil N-value at the intermediate position

between the sand piles after improvement sand supply rate

N-value of original subsoil N

0

Fig. T- 7.9.3 Relationship between N-value of Original Subsoil (N0) and N-value at Intermediate Position after Improvement (N1) (Fines Content ≦ 20%)

7.9.3 Design Based on Trial Execution When reliable past records are not available or the soil has a high silt content, trial execution shall be conducted before designing sandy soil improvement. Trial execution shall be planned comprehensively by examining the required degree of compaction estimated by means of void ratio, actual performance, and others.

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PART V FOUNDATIONS

7.10 Sand Compaction Pile Method (for Cohesive Subsoil) 7.10.1 Principle of Design [1] Scope of Application The design method described in this section shall be applied to improvement of cohesive subsoils beneath gravity type structures such as breakwaters, quaywalls, and revetments. [2] Basic Concept The improved subsoil by the sand compaction pile method is a composite ground formed by driving sand piles into soft cohesive soil. Therefore, the design shall be conducted appropriately considering the composite nature of the subsoil and the influence of execution on the subsoil. 7.10.2 Strength and Permeability of Sand Piles Sand piles shall have appropriate strength and permeability. [Commentary] Materials for sand pile should have high permeability, low fines content (less than 75mm), well-graded particle size distribution, ease of compaction, and sufficient strength as well as ease of discharge out of casing. When the sand piles with a low replacement area ratio are positively expected to function as drain piles to accelerate consolidation of clayey layer, the permeability of the sand pile material and prevention of clogging are important. The permeability requirement is relatively less important in the case of improvement with a high replacement area ratio, that is close to the replacement by displacement. Therefore, materials for sand pile should be selected considering the replacement area ratio and the purpose of improvement. [Technical Notes] There are no particular specifications on materials to be used for sand piles. Any sand material that can be supplied near the site may be used from economical viewpoint as far as it satisfies the requirements in [Commentary]. Figure T-7.10.1 shows several examples of sands used in the past. Recently, sand with a slightly higher fines content have often been used. Silt

Fine sand

Coarse sand Fine gravel Medium gravel Coarse gravel

Case 1

Passing weight percentage

Case 2 Case 3

Case 4 Case 5

Particle size

Fig. T- 7.10.1 Examples of Particle Size Distribution of Sands Used for Sand Compaction Piles

7.10.3 Shear Strength of Improved Subsoil Shear strength of improved subsoil shall be determined appropriately by considering the stability analysis method and the replacement area ratio. [Technical Notes] (1) Formula for Shear Strength of Improved Subsoil Several formulas have been proposed to calculate the shear strength of improved subsoil that consists of sand piles and soft cohesive soil. Equation (7.10.1) is most commonly used in practice, regardless of the replacement area ratio (see Fig. T- 7.10.2). 2

t = ( 1 a s ) ( c 0 + kz + Ds z m c Dc ¤ DpU ) + ( g s z + m s Ds z )a s tan f s cos q -339-

(7.10.1)


where as： replacement area ratio of sand piles (cross-sectional area of a sand pile divided by the effective cross-sectional area improved by the sand pile) c0： undrained shear strength at z = 0 (kN/m2) c0 + kz： undrained shear strength of clay in natural subsoil (kN/m2) k： rate of increase in strength with depth (kN/m3) n： stress sharing ratio (n = Ds s ¤ Ds c ) U： average degree of consolidation z： vertical coordinate (m) t： average shear strength of the improved subsoil mobilized at a slip surface (kN/m2) m s： stress concentration coefficient in a sand pile ( m s = Ds s ¤ Ds z = n ¤ [ 1 + ( n 1 )a s ] ) m c： stress reduction coefficient in clay part ( m c = Ds c ¤ Ds z = 1 ¤ [ 1 + ( n 1 )a s ] ) g s： unit weight of sand pile (or effective unit weight below the groundwater level) (kN/m3) g c： unit weight of cohesive soil (or effective unit weight below the groundwater level) (kN/m3) f s： internal friction angle of sand in pile (°) q： angle of slip line to the horizontal (°) Ds z： mean increment of vertical stress due to external forces at a slip surface (kN/m2) Ds s： increment of vertical stress due to external forces in a sand pile at a slip surface (kN/m2) Ds c： increment of vertical stress due to external forces in clay part between sand piles at a slip surface (kN/m2) Dc ¤ Dp：ratio of strength increase of cohesive soil in original subsoil. Slip line Clay

Sand pile

Fig. T- 7.10.2 Shear Strength of Composite Subsoil

(2) Design Parameters Design parameters used in equation (7.10.1) varied to a certain degree in the past case histories. Design parameters should be set by examining the strength of original subsoil, safety factor to be used, method of calculation (see 7.10.4 Stability Analysis), and construction speed. The followings are standard values of the stress sharing ratio and internal friction angle, which are obtained from the case studies of previous design and construction records by equation (7.10.1)13). as ≦ 0.4: n = 3, fs = 30º 0.4 ≦ as ≦ 0.7: n = 2, fs = 30º as ≧ 0.7: n = 1, fs = 35º When the value of as is equal to or larger than 0.7, the first term in the right-hand side of equation (7.10.1) is commonly neglected, or the improved area is often assumed as an uniform sandy soil with f = 30º.

7.10.4 Stability Analysis Stability analysis shall be conducted with an appropriate circular slip surface analysis, considering the past records of executions. [Commentary] The modified Fellenius method is often used for circular slip surface analysis to evaluate the stability of improved subsoil by the sand compaction pile method. In this method the subsoil and upper structures are divided into a number of segments and the normal stress on the slip surface is calculated without incorporating the forces acting on the vertical planes of segments. That is, only external forces acting on a segment is assumed to contribute to the normal stress on the slip surface of the segment. Hereinafter in this section, this calculation method is called the “slice method”. In actual subsoils, on the other hand, the stress due to external forces is distributed to a certain extent. To take into account the effects of this stress distribution in the circular slip surface analysis, the vertical stress increment Ds z at an arbitrary point on a slip surface may be obtained by using Boussinesq’s equation, which is then applied for use in the modified Fellenius method. Hereafter, this method is called the “stress distribution method”. Either the slice method or the stress distribution method is used for the stability analysis of subsoil improved by the sand compaction pile method. -340-

PART V FOUNDATIONS

[Technical Notes] In case histories in which equation (7.10.1) was applied, a safety factor of 1.2 to 1.4 is adopted including the case of examination on stability during execution. The calculated value of safety factor is strongly dependent on the shear strength parameters and the equation selected. Therefore, the combination of a calculation equation and design parameters should be appropriately determined by examining previous design data and construction records. The influence of the shear strength formula, design parameters, and calculation methods on the calculated safety factor are described in reference 13), which also includes an evaluation of safety factor based on the reverse analyse of case histories of marine construction.

7.10.5 Examining Consolidation Consolidation of improved subsoil shall be examined appropriately by considering the characteristics of composite subsoil. [Commentary] Consolidation of improved subsoil by the sand compaction pile method is calculated in the similar way to that of sand drain method except for the following two points. In design of sand drain method, the diameter and interval of sand piles (expressed as the replacement area ratio in the case of sand compaction piles) can be determined only by the length of construction period, neglecting the effect of sand piles on stability of subsoil. In design of the sand compaction pile method, it is usual to evaluate consolidation of improved ground after the replacement area ratio is determined by stability analysis, because the replacement area ratio contributes to the stability of the subsoil greatly and the consolidation usually completes in a short period. Another point is that the ground settlement may decrease and the consoliation rate may change compared with that by the sand drain method, owing to stress concentration to the sand piles in the sand compaction pile method. [Technical Notes] (1) Calculating Consolidation The final settlement of improved subsoil Sf and the final settlement of original subsoil S 0f can be calculated by the following equations: S f = e z H = bS 0f

(7.10.2)

S 0f = m v DpH where H : thickness of consolidated layer (m) mv : coefficient of volume compressbility of original subsoil (m2/kN) b : settlement ratio of improved subsoil to original subsoil: b = S f ¤ S 0f e z : vertical strain Δp : average consolidation pressure (kN/m2)

(7.10.3)

The consolidation of improved subsoil is calculated by the following procedures: (a) Calculating the relationship between the degree of consolidation and the elapsed time by the method as in 7.3 Vertical Drain Method for the diameter, interval and arrangement of sand piles obtained from the replacement area ratio determined by stability analysis. (b) Calculating the final settlement of improved subsoil by multiplying the calculated final settlement of original subsoil by the settlement ratio b . (c) Calculating the relationship between time and settlement. The increment of shear strength of clay in between sand piles due to consolidation, Δc, is obtained by the following equation, as shown within the first term in the right-hand side of equation (7.10.1). Dc = Ds z m c ( Dc ¤ Dp )U

(7.10.4)

In the calculation, a correction to the consolidation coefficient cv obtained by consolidation tests as well as appropriate selection of settlement ratio b and stress reduction coefficient mc should be made by considering the past records as described later. (2) Comparison between Calculated Settlement and Field Measurements The final settlement of improved subsoil is calculated by multiplying the predicted final settlement of original subsoil by the settlement ratio b as shown in equation (7.10.2). The settlement ratio b is generally expressed in the form similar to the stress reduction coefficient mc. Figure T- 7.10.3 shows the comparison of the calculated settlement ratio and the field measurement data. Here, the values of b by field measurements in the figure have been obtained as the ratio of the final settlement of improved subsoils, which was estimated by approximating the time-settlement relations with hyperbolic curves, to the calculated value of final settlement of the original subsoil. -341-


The figure also indicates the settlement ratio of b = 1 - as, which is used experientially in the case of high replacement area ratio. It is found in the figure that the reduction of settlement due to improvement is remarkable, and that the effect is influenced by the replacement area ratio. Moreover, although there is a large scatter in the measured values, it is found that the measured settlements are close to those estimated with the stress sharing ratio of approximately n = 4.

marine construction land construction Case 3 Case 1

Settlement ratio

β

Case 2

Case 4

b=

1+ (n

1

- 1) as Case 1,2

Replacement area ratio

as

Fig. T- 7.10.3 Relationship between Settlement Ratio and Replacement Area Ratio

(3) Comparison between Calculated and Measured Consolidation Time The consolidation rate of subsoil improved by the sand compaction pile method tends to be delayed compared to that predicted by Barron’s equation. Figure T- 7.10.4 based on previous construction data shows the delay in consolidation in terms of the coefficient of consolidation as a major parameter. In the figure, cv is the coefficient of consolidation reverse-analyzed from actual measurements for the time - settlement relationship, and cv0 is the coefficient of consolidation obtained from laboratory tests. It can be seen that the time delay in consolidation becomes greater with the increase in the replacement area ratio. Legend land construction

c v /c v

o

marine construction

Replacement area ratio

as

Fig. T- 7.10.4 Delay in Consolidation of Subsoil Improved by Sand Compaction Pile Method

(4) Comparison between Calculated and Actual Strength Increments The increment of strength of clay in between sand piles Dc is estimated by equation (7.10.4). Figure T- 7.10.5 shows the comparison between calculated and actual strength increments, in which m c is analyzed with the measured increments of clay strength between sand piles. The measured value of m c (= Dc a / Dc c ) is obtained from the measured increment of clay strength Dc a in the improved ground by the sand compaction pile method and the calculated increment of clay strength Dc c =[ Ds z ( Dc / Dp ) U] in the original subsoil. It is found that the data of field measurements range around the curve with the stress sharing ratio n of approximately 3 to 4.

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PART V FOUNDATIONS

calculated increment of clay strength measured increment of clay strength

m c ( D c a /D c c )

at sea

on land

Kasai oki

Replacement area ratio as Fig. T- 7.10.5 Strength Increase of Clay between Sand Piles in Improved Subsoil

[References] 1) Japanese Geotechnical Society: “Countermeasures for Soft Ground - Investigation, Design and Execution,” 1988, pp.317321 (in Japanese). 2) Hiroyuki TANAKA, Kazumi OHTA, Takahide MARUYAMA: “Performance of vertical drains for soft and ununiform soils”, Rept. of PHRI, Vol. 30, No. 2, 1991, pp.211-227 (in Japanese). 3) Masaaki TERASHI, Hiroyuki TANAKA, Tsukasa MITSUMOTO, Yuji NIIDOME, Sadakichi HONDA: “Fundamental properties of lime and cement treated soils (2nd Report)”, Rept. of PHRI, Vol. 19, No. 1, 1980, pp. 33-62 (in Japanese). 4) Masaaki TERASHI, Masaki Kitazume, Takeshi NAKAMURA: “External forces acting on a stiff soil mass improved by DMM”, Rept. of PHRI, Vol. 27, No. 2, 1988, pp. 147-184 (in Japanese). 5) Masaki KITAZUME: “Model and analytical studies on stability of improved ground by deep mixing method”, Tech. Note of PHRI, No. 774, 1994, 73p. (in Japanese). 6) Takashi TSUCHIDA, Yuji YOKOYAMA, Jun-ichi MIZUKAMI, Kazuya SHIMIZU, Jutaro KASAI: “Field test of lightweight geomaterials for harbor structures”, Tech. Note of PHRI, No. 833, 1996 (in Japanese). 7) Takashi TSUCHIDA, Jutaro KASAI, Jun-ichi MIZUKAMI, Yuji YOKOYAMA, Kazunori TSUCHIDA: “Effect of curing condition on mechanical properties of lightweight soils”, Tech. Note of PHRI, No. 834, 1996 (in Japanese). 8) Takashi TSUCHIDA, Koji NAGAI, Masayuki YUKAWA, Takao KISHIDA, Minoru YAMAMOTO: “Properties of lightweight soil used for backfill of pier”, Tech. Note of PHRI, No. 835, 1996 (in Japanese). 9) Yi Xin TANG, Takashi TSUCHIDA, Daisuke TAKEUCHI, Masataka KAGAMIDA, Noboru NISHIDA: “Mechanical properties of light-weight tested by triaxial apparatus”, Tech. Note of PHRI, No. 845, 1996 (in Japanese). 10) Takashi TSUCHIDA, Yoshiaki KIKUCHI, Tetsuo FUKUHARA, Takeo WAKO, Kazuhiro YAMAMURA: “Slice method for earth pressure analysis and its application to lightweight fill”, Tech. Note of PHRI, No. 924, 1999 (in Japanese). 11) Yoshiaki KIKUCHI, Kunio TAKAHASHI: “Change of mechanical characteristics of the granulated blast furnace slag according to age”, Tech. Note of PHRI, No. 915, 1998, 26p. (in Japanese). 12) Sabro OKABE: “General theory on earth pressure and seismic stability of retaining wall and dam”, Journal of J.S.C.E. Vol. 10 No. 6, 1924, pp. 1277-1323. 13) Katsumi KANDA, Masaki TERASHI: “Practical equation for the composite ground improved by sand compaction pile method”, Tech. Note of PHRI, No. 669, 1990, pp. 1-52 (in Japanese).

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-344-

Part VI Navigation Channels and Basins

PART VI NAVIGATION CHANNELS AND BASINS

Part VI Navigation Channels and Basins Chapter 1 General In planning and design of navigation channels and basins, careful considerations shall be given to their relationship with mooring facilities, protective facilities, and others. It is also necessary to examine their effects after completion upon nearby waters, adjacent facilities, topography, flow regimes, and other environmental conditions. The expected course of the port’s future development shall also be taken into consideration. In particular, special care shall be given to the planning and design of navigation channels and basins for ships that mainly transport hazardous goods. [Commentary] (1) With regard to harbor facilities such as navigation channels, port entrances, and ship maneuvering basins where ships maneuvering problems may arise, it is necessary to hear opinions of persons concerned including captains, mates, and pilots. (2) In selecting the locations for basins exclusively used to dangerous cargo vessels, the following should be considered: (a) To minimize an encounter with general ships, especially passenger ships. (b) To isolate them from the facilities of which surrounding environment should be preserved, such as housing areas, schools and hospitals. (c) To be capable of easily coping with accidents including hazardous goods spill. (3) From the viewpoint of securing safety and efficiency in navigation and cargo handling, it is desirable to separate the basins for passenger ships, ferries, and fishing boats and small craft basins from those for other types of vessels. (4) In principle, it is desirable to separate timber handling facilities as a specialized terminal from other general facilities.

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Chapter 2 Navigation Channels 2.1 General In planning and design of a navigation channel, considerations shall be given to the safety of navigation, the easiness of ship maneuvering, the topographic, meteorological and marine conditions, and the conformity with related facilities. [Commentary] (1) A navigation channel can be difined as a waterway with sufficient depth and width to allow the smooth passage of vessels. A good navigation channel should satisfy the following requirements: (a) The alignment of the navigation channel is close to a straight line. (b) The width and depth are sufficient in consideration of the effects of the shape of the channel’s bank, the sea bottom topography, and ship-generated waves on the navigation of other vessels. (c) Meteorological and marine conditions including winds and tidal currents are good for safe navigation. (d) A sufficient number of good navigation aids and signaling facilities have been provided. (2) When setting a navigation channel, it is necessary to analyze the tracks of vessels entering and leaving the port using examples of similar existing ports and harbors as references. It is also necessary to hear opinions from the people in the local maritime organizations. Further considerations should be given to the status of provisions of navigation aids and the marine traffic control system within the harbor, the distances from the adjacent basins to the harbor, the navigation channel division methods employed for the harbor (e.g., large and small ships, inbound and outbound traffic), the angle of approach to the harbor, and whether or not tugboats are used. (3) For the area of water that is mainly used for the navigation of vessels, measures should be taken to avoid anchorage or turning of vessels within such waters even when there is no designation for navigation channels.

2.2 Alignment of Navigation Channel (Notification Article 28, Clause 1) When a navigation channel has a bend, the intersection angle of centerlines of channels at the bend shall be made as small as possible. [Commentary] (1) In the determination of the intersection angle at a bend of navigation channel, it is necessary to consider the turning diameters, the sailing speeds, the ratio of vessel’s draft to water depth, the number of navigation aids deployed, etc. (2) When the direction of wind or tidal currents is nearly perpendicular to a navigation channel, ship maneuvering is greatly affected by the wind or tidal currents. Therefore, it is necessary to consider these effects where wind and/ or tidal currents are strong. (3) It is advisable that the intersection angle of channel centerlines at a bend will not exceed approximately 30º. When the angle exceeds 30º, the centerline of the channel at the bend should be made as an arc having the radius of curvature larger than about four times the length overall of the target vessel. The width of the channel at such a bend should be equal to or larger than the required width, except for cases where the target vessel has good turning performance, such as the cases of yachts, motorboats, and other ships used for sports or recreational purposes, or where a sufficient number of signals or other facilities have been provided to Fig. C- 2.2.1 Width Increase with Corner Cut at Bend Section allow safe and smooth ship maneuvering. (4) At a bend of a double-way channel where the intersection angle is 30º or greater and the channel width is equal to L (the length overall of the target vessel), it is necessary to increase the channel width at the bend with a corner cut as sketched in Fig. C- 2.2.1.

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(5) When a bridge is to be built across a navigation channel, it is necessary to secure a sufficient clearance in both the height and width. When determining clearance, the following items should be considered: (a) In the determination of clearance height ① Ship mast height and trim of ship ② Tide level and wave height ③ Psychological effects on ship captains and mates (b) In the determination of clearance width ① Prevailing wind, tidal currents, and tidal current changes by piers (depending on their shapes) ② Speed, maneuvering performance, and stopping performance of ship ③ Psychological effects on ship captains and mates

2.3 Width of Navigation Channel In the determination of the width of a navigation channel, due consideration shall be given to the types and dimensions of target vessels, the traffic volume and length of the channel, and natural conditions including meteorological and marine conditions. In such cases that the use of tugboats is prescribed, a refuge area for vessels is provided, or the length of channel is very short, however, the channel width can be reduced to the extent not to hinder safe navigation of vessels. [Commentary] (1) For ordinary navigation channels, the following values are adopted as the standard width (L indicates the length overall of the target vessel): (a) For a double-way channel, an appropriate width that is 1.0L or larger is adopted, except for: ① cases in which the length of the navigation channel is relatively long: ② cases in which the target vessels frequently pass in both ways through the channel: ③ cases in which the target vessels frequently pass in both ways through the channel and the length of the channel is relatively long:

1.5L 1.5L 2.0L

(b) For a one-way channel, an appropriate width that is 0.5L or larger is adopted. When the width becomes less than 1.0L, it is desirable to introduce sufficient safety measures such as the provision of facilities to assist navigation. (2) For special navigation channels (navigation channels with extremely heavy traffic, navigation channels which are crossed by navigating vessels, navigation channels for ultralarge vessels, navigation channels in which meteorological and/or marine conditions are severe, etc.), the channel width should be determined by adding margins as necessary to the above-mentioned standard width. (3) For navigation channels for fishing boats or ships of less than 500 GT, appropriate channel widths are determined in consideration of the usage conditions. [Technical Notes] Previous studies on navigation channel width are listed in Table T- 2.3.1. The table lists up major studies and analyses on the navigation channel width that have been published in Japan and overseas. As the table only gives a summary, refer to individual books and reports for details. The relationship between the dimensions L and B of cargo vessels according to the data of “Lloyd’s Maritime Information ’95.6” is as follows: Average value (50% value): L = 6.6B (25% value: L = 6.1B; 75% value: L = 7.2B) where L: length overall (m) B: molded breadth (m)

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Table T- 2.3.1 Studies and Analyses on Navigation Channel Width Author/originator

Double-way channel

One-way channel

Literature name

7.2B ～ 8.2B

4.6B ～ 5.1B

General Theory of Ship Maneruvering (in Japanese)

8B ～ 10B

5B ～ 6B

Basics of Ship Maneuvering in Bays and Harbors - Controllability and Ship Maneuvering for Entering Basins - (in Japanese)

7B + 30 (m)

5B

Joint Working Group PIANC and IAPH, in cooperation with IMPA and IALA

4.2B ～ 14.2B

1.9B ～ 7.2B

Approach Channels: A Guide for Design

Gregory P. Tshinker

6.2B ～ 9.0B

3.6B ～ 6.0B

Handbook of Port and Harbor Engineering

Keinosuke Honda

Akira Iwai

United Nations Conference on Trade and Development

Port Development: A Handbook for Planners in Developing Countries

Note 1: B = molded breadth of the target vessel Note 2: Since these are sample values provided for the sake of comparison, the above figures are not specifically given in the books and reports.

2.4 Depth of Navigation Channel The depth of navigation channel shall be determined in such a way that the basin depth as specified in 4.3 Depth of Basin is secured and appropriate allowances are added as necessary in consideration of the type of the bottom material, vessel motions, trim and hull sinking of vessels, errors of charts and surveying data, and the accuracy of dredging. In the case that the target vessels always navigate with shallower drafts than the respective full drafts, such as in the channels for vessel’s approach from / to docks and in those for vessels that always call after finishing unloading at another port, however, the channel depth may be determined otherwise. [Technical Notes] In the determination of keel clearance, books by Iwai, by Honda and others may be used as references. For estimating the approximate amount of hull sinking depending on the speed and size of vessel, the equation suggested by J.P. Hooft may be refered to.

2.5 Length of Navigation Channel at Harbor Entrance (Notification Article 28, Clause 2) The length of a navigation channel at the harbor entrance and the area of basin neighboring the channel shall be determined appropriately in consideration of the stopping distance of the target vessel. [Commentary] Vessels must keep a certain minimum speed while navigating in waters outside of protective facilities such as breakwaters to enter a harbor, in order to avoid adverse effects of winds and tidal currents. Therefore it is desirable to determine the length of the navigation channel from the tip of breakwaters to the wharf and the area of the neighboring basin, in a manner of ensuring the adequate stopping distances of vessels. [Technical Notes] The distance required for deceleration varies according to the size and initial speed of vessel. In the determination of this distance, books by Iwai and by Honda may be used as references. For ultralarge vessels, the report on the result of the study conducted by the Japan Maritime Disasters Prevention Association may be used. With regard to very large vessels, it is desirable to consider the use of tugboats.

2.6 Calmness of Navigation Channel (Notification Article 29, Clause 1) In planning and design of navigation channels, appropriate calmness of navigation channels shall be achieved in consideration of the safe navigation of vessels, presence of pilots on board, and use of tugboats.

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[Commentary] (1) The effect of waves on small vessels while navigating is large when there is a beam sea or following sea condition. The effect of waves on large vessels while navigating is large when there is a following sea condition. Navigation channels should be designed in such a way that the period of waves does not correspond with the vessel’s natural period of rolling and the vessel is not subjected to waves with the wavelength that is the same as or close to the length overall of the vessel. In addition, if a large vessel is subjected to following waves while navigating in an area close to the harbor entrance at a low speed, the speed of the vessel relative to the wave speed may become small and this may hinder the steering of the vessel and generate large yawing. Therefore it is desirable to avoid the course of a navigation channel that would compel vessels to navigate around the harbor entrance in the following wave condition with an angle of 45º or smaller from the astern direction. (2) When considering measures and methods to secure the desired degree of calmness for a navigation channel, it is necessary to consider the effects of waves running along and/or reflected from breakwaters and quaywalls in addition to the effect of incident waves. (3) For navigation channels in waters where wave conditions are severe, it is necessary to achieve the degree of calmness that matches the controllability of vessels, in consideration of the entry speeds and stopping distance of vessels. (4) When there is no alternative other than to provide a navigation channel in a water that would allow only a small keel clearance, it is necessary to raise the degree of calmness so that the hull sinking due to the motions of vessels can be minimized. [References] 1) 2) 3) 4)

United Nations Conference on Trade and Development: “Port Development”. The Joint Working Group PIANC and IAPH: “Approach Channels: A Guide for Design”, 1997. Gregory Tsinker: “Handbook of Port and Harbor Engineering”, 1996. Hooft, J. P: “The behavior of a ship in head waves at restricted water depths”, Int. Ship Building Progress, Vol.21, No. 244.

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Chapter 3 Navigation Channels outside Breakwaters 3.1 General In planning and design of a navigation channel outside breakwaters, the provisions given in Chapter 2 Navigation Channels shall be followed, and the safe vessel navigation at the normal speed shall be taken into consideration. [Commentary] The term “navigation channel” used in this chapter means a waterway outside breakwaters through which vessels navigate at normal speeds.

3.2 Width of Navigation Channel A navigation channel outside breakwaters shall have a width equal to or greater than the width specified in 2.3 Width of Navigation Channel. In the determination of the width of a navigation channel, considerations shall be given to the natural conditions including waves, winds, tidal currents and topography, and the sailing speed of vessels. [Commentary] For navigation channels in which two vessels may navigate side by side or a vessel may pass by another, it is necessary to consider the interaction between two vessels, the vessel’s yawing, and the psychological effects of the channel width on the vessel’s captain and mates. [Technical Notes] (1) When two vessels navigate alongside or a vessel pass by another in close proximity to each other, the symmetric flow patterns around both vessels are lost and this generates the suction and repulsive forces between them. The effect of these forces on vessel maneuvering is called the “interaction between two ships”. (2) In the determination of the interaction between two vessels, books by Iwai and by Honda may be used as references.

3.3 Depth of Navigation Channel For navigation channels outside breakwaters, the depth specified in 2.4 Depth of Navigation Channel shall be secured. As navigation channels outside breakwaters are generally characterized by severe marine conditions and high vessel speeds, careful consideration shall be given to sufficient depth allowances for the vessel motions, hull sinking and trim of vessel, type of bottom sediment, errors of charts and surveying data, and accuracy of dredging. [Commentary] (1) The motions of vessel hull due to waves are governed by the length overall, width and speed of the vessel, and the wavelength and wave height. For the keel clearance to be able to accommodate these vessel motions, it is generally considered sufficient to allow about two thirds of the wave height for small and medium-sized ships and about one half of the wave height for large-sized vessels. In the determination of keel clearance, it is desirable to use the data of existing navigation channels with similar navigation conditions. (2) In shallow water areas and navigation channels with a small cross section, the water around the hull of a sailing vessel is accelerated and the pressure is reduced. As a result, the water level around the vessel hull goes down, thereby lowering the vessel hull itself and in many cases changing vessel attitude into “trim by bow”. In this case, the turning performance of the vessel is reduced, but the course keeping performance is improved. (3) In the determination of the depth of a navigation channel in which sand waves often appear or a navigation channel whose bottom material is muddy, it is necessary to consider the influence of these factors.

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Chapter 4 Basins 4.1 General In planning and design of basins, considerations shall be given to the safety in anchorage, the easiness of ship maneuvering, the cargo handling efficiency, the meteorological and marine conditions, the effects of reflected waves and ship-generated waves on vessels in the harbor, and the conformity with related facilities.

4.2 Location and Area of Basin 4.2.1 Location The location of basin shall be determined appropriately in consideration of the layout of breakwaters, wharves and navigation channels, and the calmness requirement. 4.2.2 Area of Basin Used for Anchorage or Mooring (1) In the determination of the area of a basin used for anchorage or buoy mooring, due considerations shall be given to the purpose of the use, anchorage method, sea bottom material, wind speed, and water depth. (2) For buoy mooring, considerations shall be given to the type of the use, and to the horizontal movement of the buoy when the tidal range is large. [Commentary] (1) Swinging mooring (Fig. C- 4.2.1(a)) and mooring with two anchors (Fig. C- 4.2.1(b)) are the two most frequently used anchorage methods. (2) The required length of anchor chain varies depending on the type of vessel, anchorage method, and meteorological and marine conditions. Therefore, it is necessary to determine the chain length in such a way that the holding powers of the mooring anchor and the chain lying on the sea bottom can resist the forces acting on the vessel under expected conditions. In general, the stability of the mooring system increases as the length of the anchor chain becomes longer. (3) The area of anchorage area is defined as a circle having a radius equivalent to the sum of the vessel’s length and the horizontal distance between the bow and the center of rotation. (4) When the dimensions required to calculate the length of the anchor chain are unknown, Table C- 4.2.1 may normally be used as a reference to determine the area of anchorage area. (5) Figure C- 4.2.1 (c) shows a vessel moored in a single-buoy mooring, and Figure C- 4.2.1 (d) shows a vessel moored in a double-buoy mooring with the buoys located in bow and stern of the vessel. In this double-buoy mooring, it is necessary to locate the buoys in such a way that the line connecting the two buoys become parallel with the directions of tidal currents and winds. In the determination of the area of these types of buoy mooring areas, Table C- 4.2.2 may be used as a reference. (6) The width of a basin to be established between two piers in cases where many finger piers run in parallel may be determined using the following rule as reference: (a) When the number of berths on one side of the pier is 3 or less: 1.0L (b) When the number of berths on one side of the pier is 4 or more: 1.5L (L: length overall of target vessel) In cases where the basin is to be used by bunker boats or barges, or where the area around the end of the pier is to be used as a basin for small crafts, it is desirable to add margins to the width in consideration of the respective usage pattern. Table C- 4.2.1 Anchorage Area Purpose of the use of the basin

Anchorage method Swinging mooring

Offshore waiting or cargo handling Mooring with two anchors Note:

Sea bottom material or wind speed Good anchoring

L + 6D

Poor anchoring

L + 6D + 30 m

Good anchoring

L + 4.5D

Poor anchoring

L + 4.5D + 25 m

L; length overall of target vessel (m), D; water depth (m) -351-

Radius


Table C- 4.2.2 Area of Basin Used for Buoy Mooring Anchorage method

Area

Single-buoy mooring

Circle having a radius of (L + 25 m)

Double-buoy mooring

(L + 50 m) ´ L/2 rectangle

Note: L; length overall of target vessel (m)

(b) Mooring with two anchors (a) Swinging mooring

(d) Double-buoy mooring (c) Single-buoy mooring Fig. C- 4.2.1 Basic Concept of the Area of Basin (per Vessel)

[Technical Notes] (1) In the determination of anchorage method and the area of a basin for stormy weather, books by Iwai and by Honda may be used as references. In the determination of the area of a basin for ultralarge vessels, which are 100,000GT or larger, publications by the Japan Marine Disasters Prevention Association may be used as a reference. (2) For the consideration of the area of a basin with respect to anchorage during stormy weather, the model suggested by Suzuki 1) may be used as a reference.

4.2.3 Area of Basin Used for Ship Maneuvering [1] Turning Basin In the determination of the area of a basin used for bow turning, due considerations shall be given to the method of bow turning, the vessel’s bow turning performance, the layout of mooring facilities and navigation channels, and the meteorological and marine conditions. [Commentary] (1) It is recommended that turning basins be located appropriately in front of mooring facilities in consideration of the layout of other navigation channels and basins. (2) The standard area of turning basin is as follows: (a) Bow turning without assistance of tugboats: Circle having a diameter of 3L (b) Bow turning using tugboats: Circle having a diameter of 2L Thrusters with a sufficient power may be considered as equivalent to a tugboat. (3) As for small ships, when the above standard area cannot be provided due to topographic conditions, the area of turning basin may be reduced to the following level by using mooring anchors, winds, or tidal currents: (a) Bow turning without assistance of tugboat: (b) Bow turning using tugboat:

Circle having a diameter of 2L Circle having a diameter of 1.5L

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(4) In those cases where the standard size cannot be provided due to topographical constraints but a water area adjacent to the basin can be used in case of emergency, the basin area smaller than the standard size may be used as long as the smaller area is considered sufficient to meet the safety requirements.

[2] Mooring / Unmooring Basin In the determination of the area of a mooring / unmooring basin, considerations shall be given to whether or not tugboats are used, effects of wind and tidal currents, and easiness of ship maneuvering. [Technical Notes] In the determination of the area of a mooring / unmooring basin, books by Iwai and by Honda may be used as references. For cases in which tugboats are used, books by Nakajima and Yamagata may be used as references.

4.3 Depth of Basin (1) The depth of basin below the datum level shall be determined by adding an appropriate keel clearance to the maximum draft expected (such as full draft). (2) When the full-load draft etc. of the target vessel are unknown, the depth of basin shall be determined appropriately. [Technical Notes] (1) Where the seasonal variation in the mean sea level is larger than the tide level variation due to astronomical tide and the mean sea level frequently becomes lower than the datum level, or where waves or swell of appreciable heights enter the basin, it is necessary to consider the effects of these phenomena. (2) When the dimensions of the target vessel cannot be specified in advance such as in the case of a public port, the water depth of berth as specified in Part Ⅷ , 2.1 Length and Water Depth of Berths may be used as the standard depth for the basin.

4.4 Calmness of Basin (Notification Article 29, Clause 2) For basins that are located in front of mooring facilities and used for accommodating or mooring vessels, the calmness of a specified level shall be achieved for 97.5% or more of the days of the year, except for those cases where the use of the mooring facilities or the area in front of the mooring facilities is categorized as a special use. [Commentary] (1) The threshold wave heights for cargo handling for basins in front of mooring facilities should be determined appropriately in consideration of the type, size, and cargo handling characteristics of the vessels. For this purpose, the values listed in Table C- 4.4.1 may be adopted. Calmness of basin is usually evaluated by the wave height in the basin, but it is desirable to consider as necessary the effects of wave direction and period, which affect the motions of moored vessels as well. Table C- 4.4.1 Theshold Wave Height for Cargo Handling Ship size

Threshold wave height for cargo handling (H1/3)

Small-sized ships

0.3 m

Medium- and largesized vessels

0.5 m

Very large vessels

0.7 ～ 1.5 m

Note: Small-sized ships are vessels smaller than about 500 GT that mainly use the basins for small crafts, and very large ships are vessels larger than about 50,000 GT that mainly use large dolphins and offshore berths. Medium- and large-sized ships are vessels that do not belong to the small-sized and very large ship categories.

(2) It is recommended that the threshold wave height and wind speed for refuge in a harbor be determined appropriately in consideration of whether the harbor is facing the open sea or the inland sea, the type and size of vessels, and the mooring method employed (e.g., wharf mooring, buoy mooring, anchorage). [Technical Notes] Table T- 4.4.1 gives as references the allowable amplitudes of vessel motions for different types of cargo handling that have been suggested by Ueda and Shiraishi 2) and PIANC 3).

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TECHNICAL STANDARDS AND COMMENTARIES FOR PORT AND HARBOUR FACILITIES IN JAPAN Table T- 4.4.1 Allowable Amplitudes of Vessel Motions for Different Types of Vessels 2) Vessel type

Surging

Swaying

Heaving

Rolling

Pitching

Yawing

（m）

（m）

（m）

（º ）

（º ）

（º ）

General cargo ships

± 1.0

± 0.75

± 0.5

± 2.5

± 1.0

± 1.5

Grain carriers

± 1.0

± 0.5

± 0.5

± 1.0

± 1.0

± 1.0

Ore carriers

± 1.0

± 1.0

± 0.5

± 3.0

± 1.0

± 1.0

Oil tankers (foreign route)

± 1.5

± 0.75

± 0.5

± 4.0

± 2.0

± 2.0

Oil tankers (coastal route)

± 1.0

± 0.75

± 0.5

± 3.0

± 1.5

± 1.5

Container ships (LO/LO)

± 0.5

± 0.3

± 0.3

± 1.5*

± 0.5*

± 0.5*

Ferries, container ships (RO/RO) and pure car carriers

± 0.3*

± 0.6*

± 0.3*

± 1.0*

± 0.5*

± 0.5*

Note: the values marked “ * ” are based on reference 3).

4.5 Timber Sorting Pond Timber sorting pond shall be equipped with facilities to prevent timber drift. The facilities to prevent timber drift include timber mooring piles, timber drift prevention fences, and other facilities. When a protective facility is provided around a timber sorting pond, the protective facility may be used as the facility to prevent timber drift. [Technical Notes] The structures and composition of such facilities to prevent timber drift are as specified in PartⅦ Ⅶ, 2.10 Breakwaters for Timber Handling Facilities and Part Ⅸ , 2.5 Timber Sorting Areas. [References] 1) Yasumasa SUZUKI: “Study on the design of single buoy mooring”, Tech. Note of PHRI, No.829, 1996 (in Japanese). 2) Shigeru UEDA, Satoru SHIRAISHI: “The allowable ship motions for cargo handling at wharves”, Rept. of PHRI, Vol.27, No.4, 1988, pp.3-61. 3) “Criteria for Movements of Moored Ships in Harbors: A Practical Guide”: Report of Working Group No. 24, Supplement to Bulletin No. 88, Permanent International Association of Navigation, 1995.

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Chapter 5 Small Craft Basins In planning and design of small craft basins, considerations shall be given to the safety in mooring, the easiness of vessel maneuvering, the meteorological and marine conditions, and the conformity with related facilities.

Chapter 6 Maintenance of Navigation Channels and Basins 6.1 General (Notification Article 30) In principle, navigation channels and basins shall be maintained appropriately for safe and smooth vessel operations in accordance with relevant standards, in consideration of the natural conditions and the status of the use. [Commentary] (1) When a navigation channel or basin is to be built in an estuary port or a seashore area where a significant level of littoral drift is expected, the volume of sediment transport under flood conditions or the rate of longshore sediment transport due to waves and tidal currents should be estimated and the amount of maintenance dredging required for the future should be predicted. (2) For a navigation channel or basin which might be silted up during a storm, periodical checks should be made in an appropriate manner and emergency checks should be made as necessary. When the required water depth or width may not be secured as found by the result of such checks, appropriate measures should be taken to remedy the situation. (3) Periodical checks are those conducted at predetermined intervals for selected areas, and emergency checks are those conducted when there is a risk that a navigation channel or basin may be silted up due to an abnormal meteorological condition. (4) The surrounding environments (topographic conditions, bottom material, marine conditions, etc.) of individual navigation channels and basins are different. Therefore, the intervals of checks, the criteria for assessing the status / conditions, and the measures for decreasing the risk of hazard in the navigation channel / basin should be determined appropriately in consideration of the surrounding environments.

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Part VII Protective Facilities for Harbors

PART VII PROTECTIVE FACILITIES FOR HARBORS

Part VII Protective Facilities for Harbors Chapter 1 General 1.1 General Consideration In the design of protective facilities such as breakwaters, jetties, seawalls, training jetties, sluices, locks, revetments, levees, groins, and parapets, the following consideration shall be made: (1) Interrelations with navigation channels, basins, mooring facilities, and other facilities. (2) Their influences exerted on the nearby water area, facilities, topography, currents, and other environment after their construction. (3) Room for future development of the port. [Commentary] (1) “Protective facilities” refers to breakwaters, jetties, seawalls, training jetties, sluices, locks, revetments, levees, groins, parapets, etc. (2) The functions of protective facilities for harbors include ensuring harbor calmness, maintaining water depth, preventing beach erosion, controlling the rise of water level in the areas behind levees during storm surges, and diminishing invading tsunami waves, as well as protecting harbor facilities and land areas from waves, storm surges, and tsunami. In recent years, protective facilities have also been required to provide facilities for water frontage amenities such that people can enjoy contact with the nature or appreciate the harbor environment. In many circumstances, protective facilities are expected to provide several of these functions. In these situations, therefore, careful consideration is required when designing the protective facilities to adequately fulfill these various functions. (3) When constructing protective facilities for harbors, their layout and structural type must be decided after giving careful consideration to the influences that will be exerted on the nearby navigation channels and basins, mooring facilities, topography, currents, and other environments. The influences caused by protective harbor facilities are as follows: (a) When protective facilities are constructed on a coast of sandy beach, they may cause various morphological changes to the surrounding area such as beach accretion or erosion. (b) Construction of breakwaters may increase the wave height at the outside of protective facilities because of reflected waves. (c) In the inside of a harbor, the calmness of water area may be disturbed because of multiple wave reflections triggered by construction of new protective facilities or harbor oscillations due to the changes of harbor shape. (d) Construction of protective facilities may bring about changes in the surrounding tidal currents or flow conditions of a river mouth, thus inviting localized changes of water quality. (4) Because of the fact that protective facilities also provide a habitat for marine organisms such as fish, marine plants, and plankton, the biological and ecological environments must also be taken into consideration when planning a facility layout and making structural design. (5) When placing protective facilities adjacent to the areas such as natural park zones or cultural facilities, it is preferable to consider not only the functions of the facilities themselves but also external appearance such as shape and color. In addition, in situations where water frontage amenity functions will be added to protective facilities, convenience and safety of people must also be taken into consideration. (6) Because the failure of protective facilities would have an enormous impact on the safety of vessels in the harbor, mooring facilities, and land areas, a thorough investigation of facility safety during the design, construction, and maintenance phases is required.

1.2 Maintenance (Notification Article 54) Protective facilities for harbors shall be maintained in good conditions based upon appropriate standards and guidelines, considering the natural conditions and construction characteristics in question, so that they will fully perform their functions. [Technical Notes] Refer to Part Ⅰ, Chapter 3 Maintenance.

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Chapter 2 Breakwaters 2.1 General When designing breakwaters, the following matters shall be investigated: (1) Layout of breakwaters (2) Influence on the surrounding topography (3) Harmony with the surrounding environments (4) Design conditions (5) Structural types of breakwaters (6) Possibility of multiple use of breakwater (7) Design method (8) Construction method (9) Economic aspects [Technical Notes] In recent years, the water frontage amenity functions such as facilities for recreational fishing have been added to breakwaters, and situations calling for multiple use continue to increase. When designing this type of breakwater with water frontage amenity, designers can refer to the “Technology Manual for Port and Harbour Environmental Enhancement Facilities”.

2.2 Layout of Breakwaters (Notification Article 55) Breakwaters shall be appropriately located so that the conditions specified in Part Ⅵ, 2.6 Calmness of Navigation Channel and Part Ⅵ, 4.4 Calmness of Basin will be satisfied. [Commentary] (1) Breakwaters are built to maintain the calmness in harbors, facilitate cargo loading and unloading, ensure the safety of ships during navigation or anchorage, and protect harbor facilities. To fulfil these requirements, the following objectives should be met: (a) Breakwaters should be so placed that the harbor entrance is at the location not facing the direction of the most frequent waves or the direction of the design storm waves in order to suppress the energy of waves entering the harbor. (b) Breakwaters should be so laid out that they will effectively shelter the harbor against the most frequent waves and the largest storm waves. (c) The harbor entrance should have a sufficient effective width so that it will not present an obstacle to ship navigation, and it should orient the navigation channel in a direction that makes navigation easy. (d) Tidal currents around the harbor entrance should be as slow as possible. (e) The influences of reflected waves, Mach-stem waves, and wave concentration on the navigation channels and mooring basins should be minimized. (f) Breakwaters should enclose a sufficiently large water area that is needed for ship berthing, cargo loading and unloading, and ship anchorage. These objectives are also mutually contradictory goals, however. A narrow harbor entrance width, for example, is best in order to achieve the calmness in a harbor but is inconvenient for navigation. The direction of most frequent waves and the direction of the strongest waves are not necessarily the same. In this kind of situation the breakwater layout must be determined through a comprehensive investigation of all the factors such as conditions of vessel utilization, construction cost, construction works, and easiness or difficulty of maintenance. For the estimation procedures for tranquillity within a harbor, refer to Part II, 4.5 Transformations of Waves and Part II, 4.6.3 Wave Transmission. Also, refer to Part Ⅵ , Chapter 2 Navigation Channels regarding the harbor entrance width and direction, and Part Ⅵ , Chapter 4 Basins with regard to water surface area within a harbor. (2) In situations where concerns for deterioration of water quality exist, consideration must be given to the exchangeability of seawater with the outside sea so that seawater within the harbor does not stagnate. (3) When building breakwaters, the natural conditions and construction conditions must be considered, and economic aspects must also be investigated. Especially, the followings should be taken into consideration: (a) Layouts that cause wave concentration should be avoided.

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(b) Locations with the especially soft and weak ground should be avoided, and breakwaters should be located at the places where construction works can be easily carried out. (c) Topographic features such as headlands and islands should be utilized as the wave shelter as much as possible. (d) On sandy beach coasts, breakwaters should be so laid out to minimize the sedimentation of littoral drift in the harbor. (e) Adequate consideration should be given to the influence of breakwater construction on the adjacent areas. Refer to Part II, 4.5.4 [3] Transformation of Waves at Concave Corners, near the Heads of Breakwaters, and around Detached Breakwaters with regard to the wave concentration. For breakwaters that will be built on sandy beach coasts, refer to Part II, Chapter 10 Littoral Drift and Chapter 5 Facilities to Prevent Shoaling and Siltation. (4) Breakwaters should be so laid out that they do not form an obstacle to the future development of the port and harbor. (5) The “effective harbor entrance width” means the width of the channel at the specified depth of water, not merely the width across the water surface of harbor entrance. The speed of the tidal currents cutting across the harbor entrance ideally should be less than 2 to 3 knots under the normal conditions. (6) In the areas surrounding shoals, the wave height often increases owing to wave refraction. In some case, impact wave forces will act on the breakwater built on a seabed with steep slope. It should be noted that a very large structure may be required when a breakwater is placed over or directly behind a shoal. (7) With detached breakwaters, it should be noted that, if the length of the breakwater is less than a few times the length of incident waves, the distribution of wave heights behind the breakwater will take a considerable variation due to the effect of wave diffraction around both ends of the breakwater and the breakwater stability will be affected accordingly. With regard to the effect of wave diffraction, refer to Part II, 4.5.3 Wave Diffraction and Partt II, 4.5.4 [3] Transformation of Waves at Concave Corners, near the Heads of Breakwaters and around Detached Breakwaters. (8) The level of calmness required within the harbor should be investigated from the standpoints of cargo handling operations and the limiting wave height for safe anchorage. For the calmness levels within basinss, refer to Part Ⅵ , 4.4 Calmness of Basin.

2.3 Design Conditions of Breakwaters The following shall be taken as the design conditions of breakwaters: (1) Calmness within the harbour (2) Winds (3) Tidal levels (4) Waves (5) Water depth and geotechnical conditions of seabed (6) Other factors

2.4 Selection of Structural Types Selection of the breakwater structural type shall be made by taking into consideration the characteristics of each structural type and making a decision after a comparative investigation of the following matters: (1) Layout of breakwaters (2) Environmental conditions (3) Service conditions (4) Construction conditions (5) Economic aspects (6) Construction period (7) Importance of breakwaters (8) Availability of construction materials (9) Maintenance

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[Technical Notes] (1) Breakwaters are normally classified according to structural type as follows (refer to Fig. T- 2.4.1(a) ～ (i) 1)): Sloping breakwaters

Rubble mound type sloping breakwaters Concrete block type sloping breakwater Caisson type upright breakwater Concrete block type upright breakwater

Upright breakwaters

Cellular concrete block type upright breakwater Mass concrete block type upright breakwater

Breakwaters

Caisson type composite breakwater Cellular concrete block type composite breakwater

Composite breakwaters

Concrete block type composite breakwater Mass concrete block type composite breakwater

Caisson breakwaters covered with wave-dissipating concrete blocks Other types of breakwaters

(2) For the breakwaters other than sloping, upright, and composite breakwaters, refer to Chapter 3 Other Types of Breakwaters.

Seaward Side

Harbor Side

ete

cr cons d rme ock efo bl

HWL

Def

Concrete crown

D Armor stone Scour matting

orm bloed co cks ncr ete

Rubble

Armor stone

(a) Rubble Mound Type Sloping Breakwater

Harbor Side Rubble Precast concrete

Seaward Side

ete

cr cokns d e rm loc efo b

HWL

D

Scour matting

con Defo cret rm e bl ed ock s

Cubic concrete blocks

Rubble (b) Concrete Block Type Sloping Breakwater

Seaward Side HWL

Harbor Side

Concrete crown

Concrete cap Foot protection concrete blocks Deformed concrete blocks

Caisson

Foot protection concrete blocks

Rubble (c) Caisson Type Upright Breakwater Fig. T- 2.4.1 (a-c) Examples of Breakwaters -360-


Seaward Side

Harbor Side

Concrete crown

HWL Concrete blocks Foot protection concrete bloc Armor stone

ot protection concrete blocks Armor stone Rubble

(d) Concrete Block Type Upright Breakwater

Harbor Side

Seaward Side Concrete crown

HWL


Concrete cap

Caisson

Deformed concrete blocks

Foot protection concrete blocks Armor stone

Rubble (e) Caisson Type Composite Breakwater

Seaward Side

Harbor Side

Concrete crown

HWL Foot protection concrete blocks Armor stone

Caisson Rubble

Concrete cap Foot protection concrete blocks Armor stone

Replacement sand (f) Caisson Type Composite Breakwater (in case of soft silt sea bed)

Seaward Side

Harbor Side Concrete crown

HWL

Cellular concrete block

Armor stone

Armor stone

Rubble

(g) Cellular Concrete Block Type Composite Breakwater

Seaward Side HWL Armor stone

Concrete crown

Harbor Side Concrete blocks Foot protection concrete blocks Armor stone

Foot protection concrete blocks Rubble

(h) Concrete Block Type Composite Breakwater Fig. T- 2.4.1 (d-h) Examples of Breakwaters (continued) -361-


Seaward Side

Harbor Side

Concrete crown

HWL Wave dissipating blocks Riprap block

Concrete cap

Caisson

Armor stone

Rubble Rubble HWL

Seaward Side

Concrete crown

Wave dissipating blocks Foot protection concrete blocks

Foot protection concrete blocksRubble Harbor Side Concrete cap

Caisson

Armor stone

Rubble Armor stone


(i) Caisson Breakwater Covered with Wave-dissipating Concrete Blocks Fig. T- 2.4.1 (i) Examples of Breakwaters (continued)

2.5 Determination of Cross Section 2.5.1 Upright Breakwater (Notification Article 46, Clause 1) The crest elevation of a breakwater shall be appropriately determined at 0.6 times the design significant wave height or greater above the mean monthly-highest water level. In this case, the appropriate height shall be established with consideration given to the factors such as the calmness of a basin and the protection of port and harbor facilities behind the breakwater. [Commentary] (1) For ports and harbors where the influence of storm surges should be considered, the datum sea level for determining the crest elevation of the breakwaters should preferably be taken at the tidal level obtained by adding an appropriate deviation of storm tide based on the past records to the mean monthly-highest water level. (2) For determination of the crest elevation of a breakwater with an added amenity purpose that will be used by the general public, a separate investigation of splash or overtopping rate is required from the standpoint of public usage and safety. [Technical Notes] (1) The crest elevation of a breakwater should be not less than 0.6 times the design significant wave height above the mean monthly-highest water level (HWL). (2) In most of existing breakwaters, the crest elevations have been determined as in the following: (a) In a harbor of large ships’ calling, where the water area behind the breakwater is so wide that wave overtopping is allowed to some extent, the crest elevation is set at 0.6H1/3 above the mean monthly-highest water level in situations where it is not necessary to consider the influence of storm surge. (b) In a harbor where the water area behind the breakwater is small and is used for small ships, overtopping waves should be prevented as much as possible. Hence the crest elevation is set at 1.25H1/3 above the mean monthlyhighest water level. (3) In the region where large storm waves close to the design waves attack frequently with long duration, even the activity of a harbor of large ships’ calling with a wide water area behind the breakwaters may be limited by the influence of waves overtopping the breakwaters, even if the breakwater has the crest elevation of 0.6H1/3 above the mean monthly highest water level. Accordingly, in such a harbor, the crest elevation is preferably set higher than 0.6H1/3 above the mean monthly-highest water level. (4) The thickness of the concrete crown should be 1.0 m or more in situations where the design significant wave height is 2 m or greater, and at least 50 cm when the wave height is less than 2 m, to avoid its destruction by overtopped waves. -362-


(5) Because the works of caisson placement, sand filling, and placing of the concrete lid and concrete crown will be constrained with the tide if the top elevation of the caisson is low, the top of the caisson is normally set higher than the mean monthly-highest water level. (6) To make placement of crown concrete easy, the top elevation of uppermost blocks of concrete block type upright breakwaters and cellular concrete block type breakwaters should not be lower than the mean sea water level (MSL). It should be above the mean monthly-highest water level, if possible.

2.5.2 Composite Breakwater The crest elevation of upright section shall follow 2.5.1 Upright Breakwater. But when the ground is soft and there is the possibility of settlement, the crest elevation shall be set higher, or the breakwater shall be designed as a structure that enables easy raising of its superstructure afterward. [Commentary] (1) The concrete crown thickness, the installation height of upright section, and the crest elevation of the composite breakwater should be based on those for upright breakwater. (2) The top elevation of the rubble mound foundation should be set as deep as possible, to prevent the generation of impulsive breaking wave forces. (3) To increase the sliding resistance of upright section, a high rubble backing is effective. In this situation, rubble stones must be tightly placed so as to have firm grips each other; otherwise overtopping waves may scatter rubbles easily. If necessary, the surface of rubble backing should be covered with concrete blocks of large size. (4) If it is expected that the soft ground settles and rubbles sink substantially, countermeasures to the settlement and sinking, such as soil improvement or use of mattresses beneath the rubble section to spread the weight of the caisson, must be implemented. [Technical Notes] (1) A rubble mound foundation is effective to broadly spread the weight of the upright section, to provide a level ground where the upright section is placed, and to prevent scouring by waves. To achieve these functions, the thickness of rubble mound is required to be 1.5 m or greater. (2) The berm width of the rubble mound foundation should be determined to ensure the required degree of stability against the circular slip failure caused by the eccentric and inclined load on the breakwater. In the locations with rough waves, the berm width should normally be 5 m or wider on the seaward side and about 2/3 that of seaward side on the harbor side. (3) The slope gradient of the rubble mound foundation is determined based upon the calculation of stability. In many cases, the seaward side of the breakwater normally has a gradient between 1: 2 to 1: 3, and the harbor side has a gradient between 1: 1.5 to 1: 2, depending upon wave conditions.

2.5.3 Sloping Breakwater (1) The crest elevation shall follow 2.5.1 Upright Breakwater. (2) The crown width shall be determined from the results of appropriate hydraulic model tests. [Commentary] (1) Because sloping breakwaters have waves permeating through them, the wave height inside the harbor will be larger than that of upright breakwater even if the crest elevation is the same. Refer to Part II, 4.6 Wave Runup, Overtopping, and Transmission with regard to wave overtopping and transmission. (2) When waves overtop heavily, a sufficiently broad crown width is required because the armor units on the top of the breakwater will become unstable. (3) For breakwaters extending from the shore and built by the end-on system from the shore, the crown width of the breakwaters should be determined based on not only stability calculation but also consideration for convenience of construction. (4) The slope gradient should be appropriately determined based upon the stability calculation. (5) For breakwaters on the soft ground, the crest elevation and construction method should be set based on 2.5.2 Composite Breakwater.

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[Technical Notes] (1) If the crest of breakwater armored with concrete blocks is set at an elevation of 0.6H1/3 above the mean monthlyhighest water level, the crown width should be equivalent to that of three or more blocks as shown in Fig. T2.5.1. Because the stability of the breakwater top section will depend upon the characteristics of the armor units and wave conditions, however, it is desirable to determine the width based upon appropriate hydraulic model experiments. (2) There are many cases where the slope gradient for rubble mound type sloping breakwaters is about 1: 2 on the seaward side of the breakwater and about 1: 1.5 on the harbor side, and about 1: 1.3 to 1: 1.5 in the case of breakwaters armored with wave-dissipating concrete blocks. When the gradient of the slope and the mass of the armor units are different between the upper and lower portions of the slope on the seaward side of the breakwater, the point at which the gradient and the mass of armor units change should be deeper than 1.5H1/3 below the design water level. 0.6

H

1/3

or more 3 blocks or more

Design water level

Concrete blocks

The number of pieces listed above are the number of hatched blocks in the upper layer of the crown.

Fig. T- 2.5.1 Sloping Breakwater Crown Width

2.5.4 Caisson Type Breakwater Covered with Wave-Dissipating Concrete Blocks (1) The crest elevation of the upright section shall be determined by referring to 2.5.1 Upright Breakwater. (2) The crest elevation of the wave-dissipating works preferably shall be set the same as that of the upright section. (3) The thickness of the crown concrete and the installation height of caissons shall be determined by referring to those of upright breakwaters. The thickness of the rubble mound foundation shall be determined by referring to that of composite breakwaters. [Commentary] (1) For a caisson type breakwater covered with wave-dissipating blocks, the overtopping waves and transmitting waves will be smaller than those for a vertical breakwater or composite breakwater. Refer to Part II, 4.6 Wave Runup, Overtopping and Transmission with regard to overtopping waves and transmitted waves. (2) Wave-dissipating works have the functions of decreasing the wave pressure, overtopping waves, transmitting waves and reflecting waves. Accurate evaluation of these functions should preferably be made based upon hydraulic model experiments. (3) When the crest elevation of wave-dissipating works is lower than that of the upright section, the impulsive breaking wave force is likely to act on the upright section. In the opposite case where the former crest elevation is higher than the latter, blocks at the crown will become unstable. (4) If the vertical faces of the upright section are not fully covered with wave-dissipating blocks at the tip of breakwater extension, large wave forces are likely to act on these vertical faces. [Technical Notes] In order to achieve a sufficient wave-dissipating effect, the crown width of the wave-dissipating works must have the width equivalent to two or more units of wave dissipating concrete blocks 2).

2.6 External Forces for Stability Calculation 2.6.1 General When calculating stability of breakwaters, external forces such as the wave force, hydrostatic pressure, buoyancy, and the breakwater deadweight shall be considered. -364-


[Commentary] (1) In addition to the above, wind pressure, seismic force, the collision force of drifting objects, and earth pressure must also be considered as external forces, if necessary. When a lighthouse is placed on the breakwater head, the wind pressure and seismic force acting on the lighthouse must be taken into consideration. (2) The leaning load from wave-dissipating concrete blocks is not considered in normal situations. When it is to be taken into consideration, refer to Part II, 5.2.4 Wave Force on Upright Wall Covered with Wave-Dissipating Concrete Blocks.

2.6.2 Wave Forces (1) Calculation of wave force shall be based on Part II, Chapter 5 Wave Force. (2) The tidal level used for calculating wave force shall be the tidal level that makes the structure most unstable. This tidal level shall be determined according to Part II, Chapter 6 Tides and Abnormal Water Levels. 2.6.3 Hydrostatic Pressure When there is a difference between the still water levels of the harbor and seaward sides of a breakwater, the differential hydrostatic pressure due to the water level difference shall be considered. 2.6.4 Buoyancy The buoyancy acting on the upright section below the still water level must be taken into consideration. When there is a difference between the still water levels of the harbor and seaward sides of a breakwater, the buoyancy must be calculated for the section below the plane that connects the water levels on the both sides of breakwater. 2.6.5 Deadweight The deadweight of breakwater can be calculated using the densities of materials used to make the breakwater. [Commentary] Refer to Part II, 15.2 Deadweight and Surcharge concerning material densities.

2.6.6 Stability during Earthuakes Examination of stability of breakwaters during an earthuakes can be omitted except for special cases. [Commentary] (1) For breakwater whose cross section is determined based on sliding stability against the waves, the calculation of stability during an earthquake can be omitted. (2) In cases where a breakwater has a slender cross section because the design wave height is small and the water depth is deep, the examination of stability against overturning caused by dynamic water pressure during an earthquake is necessary. In addition, in the case of tsunami breakwaters, the stability during an earthquake is extremely important because a tsunami is likely to arrive after the earthquake. A calculation of stability against sliding and overturning caused by dynamic water pressure during an earthquake is therefore necessary. Refer to Part II, 14.4.2 Dynamic Water Pressure during Earthquake with regard to dynamic water pressure during an earthuake.

2.7 Stability Calculation 2.7.1 Stability Calculation of Upright Section (Notification Article 48, Clauses 1 to 3) (1) In principal, examination of the stability of the upright section of gravity type breakwater shall be based on the design procedures using the safety factors against failures. (2) When the design procedures based on the safety factors are used, it shall be standard to examine the stability of the upright breakwater, the upright section of the composite breakwater, the superstructure of the sloping breakwater, and the upright section of the breakwater covered with wave-dissipating blocks, by using the items listed in (a) through (c) below. (a) Examination of stability against sliding is made using equation (2.7.1). In this examination, an -365-


appropriate value of safety factor that corresponds to the characteristics of the structure shall be used.

m ( W0 U ) Fs ≦ -------------------------P In this equation, Fs, m, W0, U and P represent the following: Fs： safety factor against sliding of the upright section m： friction coefficient between the upright section and rubble mound foundation W0： weight of the upright section in still water (kN/m) U： uplift force acting on the upright section (kN/m) P： horizontal wave force acting on the upright section (kN/m)

(2.7.1)

(b) Examination of stability against overturning is made using equation (2.7.2). In this examination, an appropriate value of safety factor that corresponds to the characteristics of the structure shall be used. W0t MU (2.7.2) Fs ≦ ----------------------MP In this equation, Fs, t, MU, and MP represent the following respective quantities, and W0 shall have the same value as in equation (2.7.1). Fs： safety factor against overturning of the upright section t： horizontal distance between the center of gravity and the heel of the upright section (m). MU： moment due to the uplift force around the heel of the upright section (kN·m/m) MP： moment due to the horizontal wave force around the heel of the upright section (kN·m/m) (c) Examination of stability against bearing capacity of the foundation at the bottom of the structure is made according to the provisions in Part V, 2.5 Bearing Capacity for zEccentric and Inclined Loads. (3) In cases where the reliability design method is used, necessary examination of the structure stability shall be carried out by establishing the appropriate threshold values, for instance, the expected sliding distance, that correspond to the functions demanded of the facility and the characteristics of the structure. [Commentary] (1) Part II, 5.2 Wave Force Acting on Upright Wall should be referred to when estimating wave forces, and the appropriate values of safety factor should be established by referring to (a) through (c) below. (a) It should be standard for the safety factor against sliding to be 1.2 or greater for wave actions. When the stability of the breakwater has been confirmed by hydraulic model experiments, however, it is acceptable for the safety factor to be less than 1.2, but it should not be less than 1.0. (b) It should be standard for the safety factor against overturning to be 1.2 or greater for wave actions and 1.1 or greater for earthquakes. (c) It should be standard for the safety factor against the bearing capacity to be 1.0 or greater for wave actions. (2) In some cases, a friction-increasing mat is laid down underneath the bottom of upright section in order to increase the friction coefficient between the upright section and the rubble mound foundation. Refer to Part II, Chapter 16 Coefficient of Friction regarding to friction-increasing mats. (3) With regard to examination of the breakwater stability, it is also possible to adopt the reliability design method that probabilistically calculates the actions of all waves which the breakwater will encounter during its lifetime and examines the stability against these waves 3),4),5),6). (4) For examination of the bearing capacity of the foundation, the expected settlement can be calculated using the reliability design method as well. [Technical Notes] (1) In the design methods based upon the safety factors against failures, as a general rule, the breakwater is so designed that it should not slide, and this is the reason that the sliding safety factor against design waves is given a margin of 0.2 or greater. Nevertheless, even for a breakwater designed with this method, the probability that sliding will not occur is not zero from the standpoint of statistical uncertainty of design factors. Consequently, even though the same safety factor is used for design, the encounter probability of sliding or the expected sliding distance will vary depending upon design conditions. The reliability design method can statistically evaluate the sliding stability of the breakwater, by using either -366-

PART VII PROTECTIVE FACILITIES FOR HARBORS one of the following; the method that calculates the encounter probability of sliding during the lifetime 3),4), or the method that calculates the expected sliding distance based on the consensus that sliding of breakwater up to a certain threshold value is permitted so long as the functions of breakwater are not impaired 5),6). In the evaluation, the encounter probability of sliding or expected sliding distance should be less than some appropriate allowable limit. For the cross section designed by the reliability design method, however, the safety factor against sliding ought to maintain the value 1.0 or greater. The expected sliding distance always means the average value, and so it is necessary to remember that some breakwater possibly slides over a distance longer than the calculated mean value. By adopting the reliability design method and permitting a deformation of breakwater to some extent, the breakwater can be designed more economically than by using the methods based on the safety factor against failure. Several items that should be considered when using the reliability design method are listed below. For details, refer to the reference 6). (a) Occurrence probability of deepwater wave height Calculate the deepwater wave height with the extreme distribution function for storm wave heights at the site, and apply the normal distribution to the calculated wave height to take into account the uncertainty due to the sample variability of extreme wave data. (b) Occurrence probability of tidal level Calculate the tidal level variation using the four principal tidal constituents and establish the cumulative distribution of tidal levels over the lifetime. Also take into consideration any storm surge anomaly to the extent as judged necessary. (c) Wave deformation calculation Calculate the design wave height in front of structures from the deepwater wave height, using Part II, 4.5 Transformations of Waves. For the wave height for the actual design, however, take into consideration its statistical variation that is assumed to be represented by the normal distribution. (d) Calculation of wave force Calculate the wave force using Part II, Chapter 5 Wave Force. However, take into consideration its statistical variation that is assumed to be represented by the normal distribution. (e) Deadweight Take into consideration its statistical variation that is assumed to be represented by the normal distribution. (f) Friction coefficient Set the value of friction coefficient based on Part II, Chapter 16 Coefficient of Friction, taking into consideration its statistical variation that is assumed to be represented by the normal distribution. Although it is believed that the coefficient depends on the number of years after the structure was built or whether sliding failure has occurred, these need not to be taken into consideration here because of many uncertainties regarding these effects. (g) Determination of sliding Assume that the breakwater will slide when the sliding safety factor falls below 1, and carry out the numerical integration or simulation of the motion of upright section, taking into consideration the statitical variations in all of the elements (a) through (f). Then calculate the encounter probability of sliding under an extreme storm event and the encounter probability of sliding during the lifetime of breakwater. In addition, use the sliding model of Shimosako et al. 6) to calculate the expected sliding distance as well. Shimosako et al. 6) carried out these calculations using the Monte Carlo simulations. For example, in the case where the reliability design method is used with the threshold value of 30 cm or 50 cm for the allowable limit for expected sliding distance, the breakwater width and cross section are often found smaller and more economical than those calculated by the conventional design method with a sliding safety factor of 1.2. However, the degree of cross-sectional reduction greatly depends upon the design conditions. It is probable that the cross section may become much larger on the contrary. In general, in the breaker zone where the upper limit of wave heights is determined by the water depth, there is a large economical benefit of using the reliability design method. When it is applied to actual designs, it is necessary to take appropriate consideration to the statistical variations of the various items mentioned above. In addition, it is also necessary to appropriately establish the threshold values for the allowable encounter probability of sliding and the allowable expected sliding distance corresponding to the importance or the lifetime length of the breakwater. (2) With regard to the bearing capacity of the foundation, Tsuchida et al. 7) calculated the expected settlement during the lifetime of breakwater built on the foundation with an insufficient bearing capacity by using the reliability design method, for the breakwater designed by the simplified Bishop method with the safety factor of 1.0. By using this method, it is also possible to design a breakwater based on the concept of expected settlement. (3) When the harbor side of the upright section is backed up with a mound of rubble stones and/or concrete blocks, careful attention must be paid to the following matters:

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(a) Investigations should be made on the possibility of hindrance to anchorage, mooring, and navigation of vessels within the harbor. (b) The safety factors against sliding and overturning by wave forces exerted upon the upright section must have the value of 1.0 or greater without any backing behind the breakwater. A small safety factor may cause violent rocking of the upright section, increase in the heel pressure, and sliding or overturning of the upright section into the seaward side during wave troughs. (c) Adequate armoring must be provided so that the backing will not be damaged by overtopped waves. (d) The backing height h should preferably be 1/3 or greater of the height of the upright section, and the width b should be the same as or greater than the height h. (e) In the case of concrete block backing, construction should be made to ensure that there are no voids between the concrete blocks and the upright section. (4) The sliding resistance when the harbor side of the upright section is backed up with a mound of rubble stones or concrete blocks muzst satisfy equation (2.7.3). m1 W 1 + R ----------------------- ≧ 1.2 (2.7.3) P where P： horizontal wave force acting on the upright section (kN/m) W1： effective weight of the upright section in still water after subtraction of uplift force (kN/m) m1： friction coefficient between the upright section and the rubble mound R： sliding resistance force of the backing of rubble stones or concrete blocks (kN/m) The sliding resistance force R can be found using the following procedure: (a) Sliding resistance force of rubble stones (refer to Fig. T- 2.7.1) (2.7.4) R = Ws tan (θ+φ) where Ws： total weight of the rubble stones in still water above the sliding surface, excluding the weight of those in the uppermost layer (KN/m) θ： angle of the sliding surface (º) φ： φ= tan-1 m2 ( m2 is the friction coefficient between rubble stones, m2 = 0.8) (º) Equation (2.7.4) can be obtained by balancing the forces exerted on the rubble section, and the slide resistance force can be calculated as the smallest value of R while changingθ by try and error. (b) Sliding resistance force of concrete blocks (refer to Fig. T- 2.7.2) (2.7.5) R= m2Ws where Ws： total weight of concrete blocks in still water, excluding the weight of those in the uppermost layer (kN/m) m2： friction coefficient between the concrete blocks and the rubble mound foundation

Fig. T- 2.7.1 Sliding Resistance Force of Rubble Stone

Fig. T- 2.7.2 Sliding Resistance Force of Concrete Blocks

(c) Takeda et al. experimentally showed that the resistance force R is a function of the ratio of wave height to water depth at the breakwater, as equation (2.7.6) demonstrates (refer to Fig. T- 2.7.3). (2.7.6) R = aWs a = 0.9 ＋ 0.2 (H / h' － 0.5): in the case of rubble stones a = 0.4 ＋ 0.2 (H / h' － 0.5): in the case of concrete blocks where Ws： a： H： h'：

b H

h'

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Ws

µ1

Fig. T- 2.7.3 Sliding Resistance Surface of a Reinforcing Body

weight in still water of the rubble stones or concrete blocks (kN/m) coefficient of resistance wave height (m) water depth on top of rubble mound (m)

When H / h’ ≦ 0.5, however, H / h’ should be set as equal to 0.5.

W1


2.7.2 Stability Calculation of Sloping Section The rubble mounds of composite breakwaters and sloping breakwaters shall be examined against the slip failure and the safety of armor units. (1) Slip Failure Examination shall be made against the slip failure caused by the eccentric and inclined loads on composite breakwaters. (2) Required Mass and Layer Thickness of Armoring Material In addition to having the adequate stable mass against wave force for each armor unit, the armor layer shall have a thickness that is capable of preventing the inner materials from being sucked out. [Commentary] (1) With regard to the slip failure caused by eccentric and inclined loads, refer to Part Ⅴ, 2.5 Bearing Capacity for Eccentric and Inclined Loads. (2) The required mass for armor units for sloping breakwaters and caisson breakwaters covered with wavedissipating blocks should be calculated according to Part II, 5.3.1 Armor Units on Slope. When the armor units are not randomly but uniformly placed or stonepitched, the required mass can be determined based on the judgment of the responsible engineer. It should be standard for the armor layer to have the thickness of two units when random placing is used. (3) The required mass for armor units of the rubble mound foundation of composite breakwater should be calculated according to Part II, 5.3.2 Armor Units on Foundation Mound of Composite Breakwater. [Technical Notes] (1) As a general principal, the required mass of rubbles or blocks below the armor layer should be approximately 1/20 of the mass of armor units or greater, although in the case of a sloping breakwater, the mass should preferably be 1/10 to 1/15 or greater. The mass of stones below the filter layer, as a general principal, should be 1/20 of the mass of filter units or greater. (2) For situations where a face of slope is covered with sand mastic asphalt, careful stability calculation is required, giving full consideration to the past construction projects and research results.

2.7.3 Stability Calculation of Whole Section (Notification Article 48, Clause 4) For gravity type breakwaters, an investigation of the stability against circular slip as prescribed in Part Ⅴ, 6.2 Stability Analysis and an investigation of settlement with due consideration to the characteristics of the ground and structure shall be undertaken to ensure the stability of the breakwater section as a whole. [Commentary] (1) See Part Ⅴ, 6.2.1 Stability Analysis Using Circular Slip Surface Method with regard to circular slip. For settlement, see Part Ⅴ, Chapter 5 Settlement of Foundations. Refer to Part Ⅴ, Chapter 7 Soil Improvement Methods with regard to soil improvement methods. (2) The tide level that is used for the calculation on stability for the whole section should be the tide level that provides the smallest factor of safety. Refer to Part II, 6.1 Design Water Level concerning determination of tide level.

2.7.4 Stability Calculation for Head and Corner of Breakwater (1) It shall be standard to set the mass of the armor stones or blocks at the breakwater head being larger than that of the trunk section of the breakwater. (2) In the case of soft ground, the slip in the direction parallel to the alignment of the breakwater shall also be examined. In this instance, the side friction resistance can be taken into account. (3) When a lighthouse is built on the breakwater head, the stability calculation shall be made by taking into consideration the seismic force, wave force, and wind pressure acting on the lighthouse. (4) Design consideration shall be given to the increase of wave height at the concave corner of breakwater. [Commentary] (1) Because there are uncertainties concerning the stability at the head of an upright breakwater or composite breakwater when compared with that at the breakwater trunk with respect to scouring of the foundation mound -369-


or external forces acting on armoring units, the mass of the armor units should preferably be set larger than that for the breakwater trunk. Refer to Part II, 5.3.2 Armor Units on Foundation Mound of Composite Breakwater with regard to the calculation of the armor unit mass. In particular, the armor units used for the breakwater head of a sloping breakwater or caisson breakwater covered with wave-dissipating blocks should preferably have a mass of 1.5 times or greater than that at the breakwater trunk, and should be installed in a form of circular mound. For the calculation of the mass of the armor units of sloping breakwaters and caisson breakwaters covered with wave-dissipating blocks, refer to Part II, 5.3.1 Armour Units on Slope. (2) Refer to Part Ⅴ, 6.2.1 Stability Analysis Using Circular Slip Surface Method concerning a slip in the direction parrallel to the alignment of a breakwater. (3) A breakwater head at which a lighthouse is built must be designed to have a sufficient stability considering additional loads caused by the lighthouse. In addition, provision of the necessary ancillary facilities to maintain the lighthouse functions must be considered. Refer to Part II, 3.3 Wind Pressure concerning wind and wind pressure acting on a lighthouse. (4) Because there are examples of damage on the shoreside end of the breakwater that was built with structures of simple construction type, adequate care must be given. (5) Around a concave corner in the alignment of upright breakwaters and composite breakwaters, an increase of wave height can occur because of superposition of reflected waves from two wings of the breakwaters in addition to the wave concentration at the concave corner itself. Because several cases of damage caused by such the wave height increase are reported, this aspect should be carefully investigated based on Part II, 4.5.4 [3] Transformation of Waves at Concave Corners, near the Heads of Breakwaters, and around Detached Breakwaters and Part II, 5.2.5 Effect of Alignment of Breakwater on Wave Force when deciding a breakwater layout and calculating the stability.

2.8 Details of Structures 2.8.1 Upright Breakwater (1) The thickness of a concrete lid of the caisson type and the cellular concrete block type breakwaters shall be carefully determined in consideration of wave conditions and construction conditions.

(3) The blocks of concrete block type breakwater shall be made as large as possible. In particular, the lowest layer shall preferably be made of one concrete block without combining multiple blocks.

Damaged cases

Thickness of concrete cap (m)

(2) The crown concrete shall be so placed to ensure solidity with the main body of breakwater. In addition, longitudinal joints shall be provided at an appropriate interval or at the junctions of caissons, when caissons are used.

Constructed cases

(4) Provision of footing in the lowest layer block of cellular concrete block type breakwater is preferable for enhancing stability.

Fig. T- 2.8.1 Examples of Concrete Lid Thickness

[Technical Notes] (1) The thickness of the concrete lid should normally be 30 cm or greater, and should be 50 cm or greater in the area of rough sea condition. There are also examples of thickness of 1.0 m or greater in the cases where wave conditions are severe and the concrete lid are left without placement of cown concrete for a long time. (refer to Fig. T- 2.8.1) (2) Longitudinal joints should be provided at an interval of 10 to 20 m on the crown concrete of monolithic breakwaters with in-situ placement concrete.

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2.8.2 Composite Breakwater Unless the top of the rubble mound foundation is set unusually deep or the waves are small and the rubble stones are large enough for stability, it is preferable to place foot protection blocks to prevent scouring of the rubble mound. The foot protection blocks shall be placed in contact with the upright section. [Commentary] (1) The rubble mound foundation of composite breakwater is extremely important to ensure the stability of the upright section. Particularly if the rubble mound beneath the upright section is scoured or washed out, the upright section will lean or easily experience sliding failure, and then the upright structure will be destroyed at worst. It is therefore necessary to protect the rubble mound beneath the upright section with foot protection blocks and prevent damage from scouring or washout due to the action of waves or currents. (2) Provision of vertical holes in the foot protection blocks reduces the uplift pressure acting on the blocks and greatly increases their stability against the waves. [Technical Notes] (1) It is recommended that two or more rows of foot protection blocks be placed on the seaward side of the upright section and one or more rows on the harbor side. (2) Because the study by Tanimoto et al. 8) shows that too large holes in foot protection blocks reduce the effect of preventing scouring and washing out, the opening ratio of about 10% is optimal. (3) It is best to search for the necessary thickness of the foot protection blocks using equation (2.8.1). (2.8.1) t ¤ H 1 ¤ 3 = d f ( h ¢ ¤ h ) 0.787 where t： necessary thickness of the foot protection blocks (m) d f ： 0.18 for the breakwater trunk, 0.21 for the breakwater head h： water depth of breakwater construction (m) h'： water depth at the top of rubble mound foundation (not including the armor blocks and foot protection blocks) (m) The application range should be h¢/ h = 0.4 ～ 1.0. For the calculation of the foot protection block dimensions, the necessary thickness can be determined using equation (2.8.1) and the dimensions listed in Table T- 2.8.1. Examples of the block shapes and dimensions are shown in Fig. T- 2.8.2. Table T- 2.8.1 Required Thickness and Dimensions of Foot Protection Blocks Required thickness of foot protection blocks t (m)

Dimensions l (m) ´ b (m) ´ t (m)

0.8 or less 1.0 or less 1.2 or less 1.4 or less 1.6 or less 1.8 or less 2.0 or less 2.2 or less

2.5 ´ 1.5 ´ 0.8 3.0 ´ 2.5 ´ 1.0 4.0 ´ 2.5 ´ 1.2 5.0 ´ 2.5 ´ 1.4 5.0 ´ 2.5 ´ 1.6 5.0 ´ 2.5 ´ 1.8 5.0 ´ 2.5 ´ 2.0 5.0 ´ 2.5 ´ 2.2

Mass (t/unit) Block with openings 6.23 15.64 24.84 37.03 42.32 47.61 52.90 58.19

Block without openings 6.90 17.25 27.60 40.25 46.00 51.75 57.50 63.25

(4) Foot protection blocks at the harbor side must be designed by taking into consideration the influence of waves inside the harbor, waves during construction, and overtopping waves. The number of the cases of failure of foot protection blocks inside a harbor is quite small until now, and it is acceptable to use a mass that is lighter than the mass of the foot protection blocks of seaward side. In the past designs there were many cases where the mass was one-half of that at the seaward side. The mass must not be smaller than the mass required by the waves inside the harbor or the waves during construction. Especially, the mass should be carefully determined for location where the offshore end of a breakwater under construction remains as a temporary head during the offwork season of each year.

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Fig. T- 2.8.2 Shapes and Dimensions of Foot Protection Blocks

(5) In situations where there are concerns about scouring or sucking-out of rubble stones, preventive countermeasures should be undertaken. Methods used for scour prevention at the toe of slope are the provision of a berm of rubbles at the end of slope, and the placement of concrete blocks, mattress work, asphalt mats, or composite resin mats. For prevention of settlement of the rubble mound due to washout, mattress works and other methods including the spreading of canvas sheets are employed.

2.8.3 Sloping Breakwater (1) Protection works against scouring at the toe of slope and sucking-out of sand beneath the breakwater shall be incorporated into the foundations of sloping breakwaters, to the extent as needed. (2) In general, when placing a superstructure on concrete block type or rubble mound type sloping breakwaters, the base of the superstructure shall be filled up with stones or small concrete blocks. (3) In a place affected by littoral drift, preventive works for blocking sand passing through the breakwater shall be provided, because suspended sediment carried by waves may cause shoaling of the harbor basin. 2.8.4 Caisson Type Breakwater Covered with Wave-Dissipating Concrete Blocks To the extent as required, preventive works against scouring and sand sucking shall be provided near the toe of the mound of wave-dissipating concrete blocks.

2.9 Detailed Design of Upright Section Design of the components of caissons, cellular concrete blocks, and L-shaped concrete blocks shall follow Part Ⅳ Precast Concrete Units.

2.10 Breakwaters for Timber-Handling Facilities 2.10.1 Breakwaters for Timber Storage Ponds and Timber Sorting Ponds (Notification Article 46, Clause 3, and Article 47, Clause 3) (1) When building breakwaters for timber-handling facilities, the crown shall be built high enough to prevent timbers from drifting out during the time of unusually high sea level but to be compatiple with the breakwater structure and the use of the pond.

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(2) When building breakwaters for timber-handling facilities, the breakwater shall be designed in accordance with the provisions of this chapter and Chapter 3 Other Types of Breakwaters, and it shall ensure the structural stability against the collision force of the timber to the extent as necessary. 2.10.2 Fences to Prevent Timber Drifting (1) The crown height and the interval between piles of the drift fence shall be designed as capable of preventing timbers from drifting away, and a coping is to be provided to the extent as necessary. (2) The drift fences and timber mooring piles shall be of the structures stable enough to withstand the collision force and tractive force of timbers.

2.11 Storm Surge Protection Breakwater (Notification Article 46, Clause 2, and Article 47, Clause 2) (1) When building breakwaters having the main function of acting as a countermeasure against storm surges, the alignment and the crest elevation of the breakwater shall be determined appropriately considering the effectiveness of breakwaters on storm surge reduction and so on. (2) When building the storm surge protection breakwaters, the stability shall be maintained against various physical characteristics of storm surges, in addition to structural stability against normal external forces such as wave forces. [Technical Notes] (1) The crest elevation of the storm surge protection breakwater should be determined at the level necessary against the sum of the design storm tide level and the design wave hight. (2) The stablility of the whole section of breakwater should be examined by assuming the suinultaneous occurrence of the design storm tide level and the peak of design waves. The rise of the water level on the harbor side due to water inflow by storm surges should also be considered. During an earthquake, the inertia force and dynamic water pressure that will exert on the breakwater as external forces should be taken into consideration. (3) It should be noted that, when the foundation of a breakwater for storm surge protection is highly ypermeable, water flows in through the foundation and the storm surge reduction function is lessened. To the extent as required, a seep-proof screen etc. are to be provided. (4) In some situations, water flows through the rubble mound foundation of the breakwater because of the tide level difference between the inside and outside of the breakwater, causing scouring of the foundation ground. In this kind of circumstance, it is necessary to consider countermeasures such as spreading layers of cobbles and stones or mattress work. See Part II, 6.6 Groundwater Level and Permeation [Technical Notes] (2) (d), regarding infiltration through rubbles.

2.12 Tsunami Protection Breakwater (Notification Article 46, Clause 2, and Article 47, Clause 2) (1) When building breakwaters having the main function of acting as a countermeasure against tsunami, the alignment and the crest elevation of the breakwater shall be determined appropriately, considering the effectiveness of breakwater on tsunami height reduction and so on. (2) When building the tsunami protection breakwaters, the stability shall be maintained against tsunami attack considering various physical characteristics of tsunamis, in addition to the structural stability against normal external forces such as wave forces. [Commentary] (1) For tsunami protection breakwaters, the same matters as for storm surge protection breakwaters should be investigated, such as tsunami height reduction effectiveness and the flow velocity through a harbor entrance. The crest elevation of tsunami protection breakwaters should be set at a higher one of the crown height needed against allowable overtopping for the design wave or that needed against allowable tsunami overtopping at HWL. Refer to Part II, 6.4 Tsunami for information concerning tsunamis. (2) Both the hydrostatic pressure due to the water level difference between the inside and outside of the breakwater and the wave force of the tsunami act as the external forces on the breakwater. Furthermore, careful attention must be paid to the fact that the water level behind the breakwater is not always the same as the still water level because of the tsunami inflow and outflow. Because there are many uncertainties regarding the wave force exerted by tsunami, the magnitude of wave force must be verified through some appropriate methods such as hydraulic model experiments. -373-

TECHNICAL STANDARDS AND COMMENTARIES FOR PORT AND HARBOUR FACILITIES IN JAPAN (3) An experimental study by Tanimoto et al. 9) has confirmed that in the situation where a tsunami flows in through a narrow harbor entrance, the flow velocity will increase and there are produced strong vortices that exert a substantial influence on the stability of the riprap material of the submerged mound section of breakwater. Tsunami also exercises strong tractive forces on the bed, which are said to be even greater than those by storm surges. Attention, thefore, must be paid in particular to the reinforcement for stability of the breakwater section at a harbor entrance and to scour prevention works for the foundation ground. (4) As tsunami protction breakwaters are often constructed in the locations where the water is deep, the upright section of breadwater needs to be tall and the stability becomes vulnerable during an earthuake. Consequently, the earthquake resistance should be examined by calculation of seismic response taking into account the nonlinearity of the rubble mound material. Furthermore, because greater seismic forces generally act on the higher part of a tall structure, the design procedure is sometimes taken by varying the seismic coeffecient according to the elevation from the bottom of the structure. In addition, the stability of the slope surface of rubble mound during an earthuake should preferably be examined. (5) Because the rubble mound becomes thicker as the water becomes deeper, it is necessary to pay careful attention to the stability of the rubble mound against wave forces and wave transformation on the slope surface of rubble mound. It will also be necessary to make extra-banking for the rubble mound against large shrinkage of the rubble mound by its own weight. (6) The effectiveness of breakwaters for tsunami height reduction was confirmed through the analysis of the habor resonance based on the tide records that were obtained in Ofunato Bay at the time of the Tokachi-Oki earthquake tsunami (May 1968) in comparison with the observation data taken before construction of the tsunami breakwater. As shown in Fig. C- 2.12.1, the wave height ratio M (bay-end amplitude to incident amplitude) with the breakwater was quite small for the range of periods longer than 25 minutes, compared with the case without the breakwater, thus demonstrating the tsunami reduction effectiveness. This effectiveness has also been proven by Ito et al. 10) through numerical calculations.

Water level observation station Ofunato Before breakwater was built Water level observation station Tsunami protection breakwater After breakwater was built

Pacific Ocean

Fig. C- 2.12.1 Effectiveness of Tsunami Protection Breakwater (Case of Ofunato Bay)

[Technical Notes] With regard to the wave force by tsunami, the change in wave force in the transition zone from short-period waves to long-period waves has been measured for three types of rubble mound heights in model tests. The results have shown that, if the wave period becomes remarkably long, the effect of rubble mound height can be ignored and that the wave force approaches the value calculated by Goda’s wave pressure formula (see Part II, 5.2.2. Wave Forces of Standing and BreakingWaves) with setting of a2 = 0. From this information, the wave force by long-period waves in the case of the Kamaishi Harbor entrance breakwater was determined as described in Part II, 6.4 Tsunami. In Fig. T- 2.12.1, the pressure p is given by 1.1 r0gH, where r0 is the density of water, g the acceleration of gravity, and H the tunami wave height.

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Fig. T- 2.12.1 Wave Force of Long-Period Waves

[References] 1) 2)

Yoshiyuki ITO: “History of discussions on breakwater design”, Tech. Note of PHRI, No. 69,1969,78p. (in Japanese). Michio MORIHIRA, Shusaku KAKIZAKI, Toru KIKUYA: “Experimental study on wave force damping effects due to deformed artificial blocks”, Rept of PHRI, Vol. 6, No. 4, 1967, pp. 1-31 (in Japanese). 3) Tomotsuka TAKAYAMA, Yasumasa SUZUKI, Hiroyasu KAWAI, Hideyoshi FUJISAKU: “Approach to probabilistic design for a breakwater”, Tech. Note of PHRI, No. 785, 1994, 36p. (in Japanese). 4) Takashi NAGAO, Noboru OKUBO, Susumu KAWASAKI, Yukio HAYASHI: “Safety factor of breakwater total system under the reliability design method (3rd report) - summary of the applicability of the levels 1 and 2 method -”, Rept of PHRI, Vol. 37, No. 2, 1998, pp. 131-176 (in Japanese). 5) Hiroyasu KAWAI, Tomotsuka TAKAYAMA, Yasumasa SUZUKI, Tetsuya HIRAISHI: “Failure probability of breakwater caisson in consideration of tidal level variation”, Rept of PHRI, Vol. 36, No. 4, 1997, pp. 3-41 (in Japanese). 6) Ken-ichiro SHIMOSAKO, Shigeo TAKAHASHI: “Reliability design method of composite breakwater using expected sliding distance”, Rept of PHRI, Vol. 37, No. 3, 1998, pp. 3-30 (in Japanese). 7) Takashi TSUCHIDA, Tang YiXin: “The optimum safety factor for stability analyses of harbor structures by use of the circular arc slip method”, Rept of PHRI, Vol. 35, No. 1, 1996, pp. 117-146 (in Japanese). 8) Katsutoshi TANIMOTO, Tadahiko YAGYU, Tsutomu MURANAGA, Kozo SHIBATA, Yoshimi GODA: “Stability of armor units for foundation mounds of composite breakwaters by irregular wave tests”, Rept of PHRI, Vol. 21, No. 3, 1982, pp. 3-42 (in Japanese). 9) Katsutoshi TANIMOTO, Katsutoshi KIMURA, Keiji MIYAZAKI: “Study on stability of dike at the opening section of tsunami protection breakwaters”, Rept of PHRI, Vol. 27, No. 4, 1988, pp. 93-121 (in Japanese). 10) Yoshiyuki ITO, Katsutoshi TANIMOTO, Tsutomu KIHARA: “Digital computation on the effect of breakwaters against longperiod waves (4th report)”, Rept of PHRI, Vol. 7, No. 4, 1968, pp. 55-83 (in Japanese).

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Chapter 3 Other Types of Breakwaters 3.1 Selection of Structural Type When employing a breakwater type other than those covered in Chapter 2 Breakwaters, the most appropriate structural type shall be selected according to the design conditions (wave conditions, water exchangeability, need for suppression of wave reflection, required calmness inside the harbour, ground conditions, etc.) and shall be designed based on the most appropriate design method or hydraulic model experiments. [Commentary] In addition to the topics covered in Chapter 2 Breakwaters, there is a diverse variety of breakwater types including shape, function, and materials 1). Many of these types respond to the functional requirements such as low wave reflectivity or water exchangeability, the site-specitic characteristics of design conditions such as the ground, water depth and waves, and particular utilization characteristics such as temporary usage. An appropriate type that satisfies the required functions must be selected, based on extensive examination of the characteristics of each type. [Technical Notes] (1) This chapter deals with special breakwaters with structures different from the breakwaters discussed in Chapter 2 Breakwaters. Although the names of these particular types of breakwaters and their classification have not been established yet, they can be classified at the present time, as follows:

Gravity type special breakwater

Deformed concrete caisson type breakwater

Upright breakwater with wave absorbing work (curved slit caisson breakwater, dual cylinder caisson breakwater, semi-circular caisson breakwater, etc.)

Upright waveabsorbing block breakwater

Slope-shaped breakwater (trapezoidal caisson breakwater, sloping-top caisson breakwater, sloping-top caisson breakwater covered with wave-dissipating blocks, etc.) Breakwater filled with wavedissipating blocks in the caisson

Other types of breakwater

Other gravity type breakwater Pile-supported type breakwater (curtain wall breakwater, steel pile-supported breakwater, etc.) Breakwater with wide footing on soft ground Floating breakwater Pneumatic type breakwater Other non-gravity type of breakwater

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Other deformed concrete type breakwater


(a) Gravity type special breakwaters are the structures that resist wave force with its own weight, and they have higher wave-dissipating performance or higher stability against wave action by improving the upright sections of conventional composite breakwaters. This type can be devided into the permeable type that allows water exchange and the impermeable type that does not: the latter is more common. The breakwaters other than gravity type have the structures that resist wave actions mainly with forces other than gravity such as pile resistance force, clay ground surface cohesion, or mooring rope tension, and they are normally permeable ones. (b) Gravity type special breakwaters can be broadly classified into the deformed concrete caisson breakwaters that use specially-shaped caissons, and the upright wave-absorbing block breakwaters that are made of perpendicular stacks of specially-shaped concrete blocks. Deformed concrete caisson breakwaters can be further classified into the wave-absorbing caisson breakwaters, the sloping caisson breakwaters, and the breakwater filled with wave-dissipating blocks in the caisson. There are various kinds of structures in upright wave-absorbing blocks, and they are normally used in places where waves are small such as in the inside of harbors. (c) Although wave-absorbing caisson breakwaters are generally used in locations where waves are comparatively small, dual cylinder caisson breakwaters with perforated-wall can be used in deep water and rough sea areas. (d) Non-gravity type breakwaters have advantages at the special conditions such as soft ground, but they also have restrictions on their application in terms of performance and stability against waves, and are generally applied for places where the wave height is comparatively small such as in inner bays or the inside of harbors. (e) Pile-supported type breakwaters have the structures that resist wave forces with pile resistance, and they include curtain wall or steel pile-supported types. (f) Breakwaters with wide footing on soft ground have the structures that resist horizontal wave force with the cohesion between the bottom of the structure and the clayey ground surface. If necessary, resistance is added by using piles. These structures were developed in order to build breakwaters over the soft clayey ground and are economical in such a condition, because the soil improvement works become unnecessary. (g) Floating breakwaters are the structures held in place by mooring lines fixed to anchors to act as a breakwater. Although there are a variety of shapes for the floating body, many are pontoon type. These structures must be appropriately anchored in order to prevent them from breaking loose during storm weather conditions and causing secondary accidents. (h) Pneumatic type breakwaters attenuate waves with air bubbles discharged from underwater air piping. While having special advantages such as not obstructing vessel navigation, they are effective only for short-period waves. (2) Because many of these breakwaters mentioned above also have a certain restriction on application by the conditions in which the breakwater can be employed, careful investigation and selection of the appropriate type corresponding to the site conditions is required. (3) Because the types of breakwaters mentioned above have the shapes different from conventional types, their hydraulic characteristics such as the wave forces, wave transmission, wave reflection and water exchangeability are still not well known. Thus, the design must be carried out based upon hydraulic model experiments.

3.2 Gravity Type Special Breakwaters 3.2.1 General Design of gravity type special breakwaters shall be made by referring to Chapter 2 Breakwaters, and their structural dimensions shall appropriately be determined considering the shapes and required functions. [Technical Notes] (1) Determination of Design Conditions Determination of design conditions should be based on 2.3 Design Conditions of Breakwaters. For waveabsorbing breakwaters, however, the target waves to be dissipated may be determined for the intended purpose and wave conditions separately from the design wave to be used for examinations of stability and calculations of structural members. (2) Determination of Structural Dimensions Related to Functions For upright wave-absorbing breakwaters, the structural dimensions related to the wave absorption function must be investigated and appropriately determined. In addition, for breakwaters that use deformed concrete caissons, the required dimensions are influenced by the respective characteristics of shapes and they must be appropriately determined. Because the wave transmission coefficient, in particular, depends upon the shape, the crest elevation must be determined appropriately in consideration of the transmission characteristics of the structure in question, and so are the dimensions of the opening set up in the caisson walls for water exchange.

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(3) Basic Cross Section Design The design of the basic cross section except for the structural dimensions related to the function should be based on 2.5 Determination of Cross Section. (4) Calculation of External Forces The calculation of external forces should be based on 2.6 External Forces for Stability Calculation. Because the wave force greatly depends on the structural shape, however, an appropriate calculation method corresponding to the structure in question or hydraulic model experiments corresponding to the conditions must be employed. For a breakwater with complicated structure, in particular, it is necessary to undertake a thorough investigation with regard to the local wave pressure acting on the structural members, in addition to the total wave force that is examined for stability calculation of the whole upright section. (5) Investigation of Stability Investigation of the stability should be based on 2.7 Stability Calculation. (6) Calculation of Structural Members The design calculation of structural members should be made based on Part Ⅳ , Chapter 4 Upright WaveAbsorbing Caissons.

3.2.2 Upright Wave-Absorbing Block Breakwater [1] General The upright wave-absorbing block breakwater shall be designed by thoroughly investigating the wave absorption function, selecting the appropriate concrete block, and conducting hydraulic model experiments if necessary. [Commentary] (1) An upright wave-absorbing block breakwater is a block-type upright or composite breakwater made by vertically stacking special blocks (upright wave-absorbing blocks). Various shapes of concrete blocks have been developed as upright wave-absorbing blocks, but the wave absorption functions of each type should be adequately examined and the appropriate type of blocks should be selected. (2) The wave reflection coefficient of upright wave-absorbing blocks greatly depends upon the wave period. When determining the reflection coefficient, it is best to carefully consider the influence of wave period based on hydraulic model experiments corresponding to the design conditions. It is also acceptable to estimate it by referring to the data from past experiments. (3) With the exception of large-scale blocks to be used as a single block structure, upright wave-absorbing block breakwaters are generally used in inner bays or the inside of harbors where wave heights are relatively small.

[2] Crest Elevation The crest elevation of upright wave-absorbing block breakwaters shall be determined by referring to 2.5.1 Upright Breakwater, taking into consideration both the crest elevation of the crown to fulfill the function of stopping wave overtopping and that of the wave-absorbing section. The wave-absorbing effectiveness shall be taken into consideration when deciding the crest elevation of the wave-absorbing section. [Commentary] Overtopping and transBreakwater crest elevation mitted waves of an Crest elevation of upright upright wave-absorbing Concrete crown wave-absorbing block block breakwater are Upright waveabsorbing block small compared with HWL those of a composite LWL breakwater, but they are larger than those of a Bottom level of upright caisson breakwater fully wave-absorbing block covered with wave-dissipating blocks. So, the crest elevation should Rubble mound be determined after giving adequate consideration to the utilization Fig. T- 3.2.1 Crown Height of an Upright Wave-absorbing Block Breakwater conditions in the area behind the breakwater. In addition, when determining the crest elevation, the minimum thickness of the concrete crown for construction should be ensured. -378-


[Technical Notes] (1) The wave-absorption effectiveness of an upright wave-absorbing block breakwater depends upon the crest and bottom elevations of the upright wave-absorbing blocks. (2) The crest elevation hc¢ should be set at least 0.5 times higher than the target significant wave height above the mean monthly-highest water level. The bottom elevation hu should preferably be two times the target significant wave height or larger below the mean monthly-lowest water level (refer to Fig. T- 3.2.1). Note that the target significant wave height may not be the same as the design significant wave height.

[3] Wave Force The wave force acting on an upright wave-absorbing block breakwater shall be calculated based on Part II, 5.2.8 Wave Force on Upright Wave-Absorbing Caisson. [Technical Notes] (1) The wave force acting on an upright wave-absorbing block breakwater should be taken as shown in Fig. T3.2.2, and the buoyancy should be considered for the displaced volume below the still water level. The uplift pressure acting on each layer of blocks or the concrete crown, when considering the stability of the section above each layer or the crown, is assumed to be of a triangular distribution with the value equivalent to the wave pressure of the front wall at which the uplift is calculated and is 0 at the rear toe. (2) The wave force reduction effectiveness for oblique incident waves should not be taken into consideration unless ascertained by model experiments.

SWL

Buoyancy

Uplift pressure

Fig. T- 3.2.2 Wave Pressure Distribution Used to Examine Stability

3.2.3 Wave-Absorbing Caisson Breakwater [1] General The appropriate type of structure shall be selected for a wave-absorbing caisson breakwater based on an adequate examination on wave-absorption performance, and then the structure shall be designed by conducting hydraulic model experiments as necessary. [Commentary] A wave-absorbing caisson breakwater will have the following merits compared with the conventional upright breakwater. (1) It can reduce wave reflection. (2) It can reduce overtopping and transmission. (3) It can reduce wave forces. In particular, when a rubble mound is relatively high, the ware force acting on a waveabsorbing caisson breakwater does not increase so remarkably, whereas a conventional caisson breakwater may generate large impulsive breaking wave forces with the same mound height. Other examples of its merits, such as aeration effects by generating air bubbles and an additional function of wave chamber as fish reefs, can also be cited 3).

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[Technical Notes] (1) “Wave-absorbing caisson” is a caisson that has a permeable wall and a wave Top slab (permeable or non-permeable) There are also many cases without an top slab chamber on its front section, and it can Concrete crown dissipate wave energy with these strucCaisson body tures 4). Figure T-3.2.3 shows the strucPerforated wall tural features of a wave-absorbing Rear wall Wave chamber Porous wall In general, in many cases Intermediate caisson in schematic form. Depending (round holes, square holes) there are no intermediate horizontal permeable wall permeable walls upon the shape and combination of the Slit wall Intermediate (vertical, horizontal) various elements, various structures, such vertical permeable wall Other as a vertical slit caisson, a horizontal slit caisson, a curved slit caisson, or a porous Bottom slab caisson, are possible. When selecting a wave-absorbing caisson among various Fig. T- 3.2.3 Structural Features of Wave-Absorbing Caissons structural types, it is necessary to select the appropriate structural type by taking into consideration the design conditions, utilization conditions, and cost, after thoroughly examining its characteristics such as wave-absorption performance and the stability against waves. (2) With regard to the various types of wave-absorbing caisson breakwater structures and characteristics, refer to the “New Type Breakwater Technology Manual”.

[2] Determination of Target Waves to Be Absorbed The target waves can be determined by taking into consideration the intended purpose of wave absorption and the local wave climate separately from the design waves employed for examination of structural stability and calculation for the structural members. [3] Determination of Dimensions for Wave-Absorbing Section The structure and dimensions of the wave-absorbing section consisting of the perforated wall and the wave chamber shall be determined appropriately, giving consideration to the wave-absorption characteristics of the structure in question and the tidal level variations, so that the wave reflection coefficient for absorption-target waves becomes lower than the target value. [Technical Notes] With regard to the reflection characteristics of vertical slit caissons that do not have a top slab, refer to the study by Tanimoto and Yoshimoto 5).

[4] Wave Force for Examination of Structural Stability The wave force used to investigate stability shall be based on Part II, 5.2.8 Wave Force on Upright Wave-Absorbing Caisson. [5] Wave Force for Design of Structural Members For the design calculation of structual members, the severest ware force applicable for each member shall be used. [Commentary] Refer to Part II, 5.2.8 Wave Force on Upright Wave-Absorbing Caisson and Part Ⅳ, 4.2 External Forces Acting on Members regarding wave forces acting on the members of the wave-absorbing caisson.

3.2.4 Sloping-Top Caisson Breakwater [1] General The appropriate structure for a sloping-top caisson breakwater shall be selected after thorough examination of its wave transmission characteristics, and preferably it shall be designed by conducting hydraulic model experiments. [Commentary] A sloping-top caisson breakwater is a breakwater that makes the horizontal wave force smaller while simultaneously making use of the vertical wave force acting on the sloping-top section for increasing the stability of the caisson. -380-


Because wave transmission is larger than that of an ordinary upright breakwater, the crest elevation should be determined after considering calmness within harbor basins more carefully.

Wave transmission coefficien,

(1) As the gradient of sloping surface becomes steeper for a sloping-top caisson breakwater, wave transimission is suppressed but wave pressure becomes larger. Then, the advantage of wave pressure reduction lessens. Based upon experiments conducted for the various slope gradient, which shows no remarkable changes in the wave transmission coefficient with gradient change, the slope gradient is recommended to be set at 45º for the wave pressure reduction effectiveness and conveniences of construction works. Other structural details should follow 2.8 Details of Structures.

KT

[Technical Notes]

Symbol Sloping-top caisson breakwater Upright breakwater

(2) As shown in Fig. T-3.2.4, transmitted wave height of the sloping-top caisson breakwater is at a level about twice that of a upright breakwater with the same crest elevation. Fig. T- 3.2.4 Wave Transmission Coefficient and When the crest elevation is set as equal to Relative Crown Height the design significant wave height H1/3 above the design water level, it is possible to lower transmitted wave height to the same level as the case of an upright breakwater having the crest elevation of 0.6 times the design significant wave height. (3) Normally the sloping surface of a sloping-top caisson breakwater is set to begin at the still water level. However, with a semi-submerged shape in which the toe end of the sloping surface is set below the still water level, further reduction of wave forces is possible. (4) With provision of a mound of wave-dissipating blocks in front of an upright part of a sloping-top caisson breakwater, not only the reflected waves are further reduced but also the entire structure becomes more stable than an ordinary sloping-top caisson breakwater. Depending upon the crest elevation of the wave-dissipating concrete blocks, however, impulsive breaking wave force may possibly be generated. In addition, careful attention must be paid to the stability of concrete blocks, because the wave-dissipating blocks that are installed around and below the still water level are exposed to strong downflow from the sloping section.

[2] Wave Force The design wave force shall be determined according to hydraulic model experiments or an appropriate calculation method. [Technical Notes] (1) As a general principle, the wave force acting on a sloping-top caisson breakwater should be obtained based upon hydraulic model experiments. But, if not feasible, it can be calculated using the following equations (refer to Fig. T- 3.2.5.): F X = F SH + F V = l SL ¢F 1 sin 2 a + l V F 2

(3.2.1)

F Z = F SV + F U = l SL ¢F 1 sin a cos a + 0.5 p U B

(3.2.2)

l SL ¢ = min [ m a x { 1.0, 23 ( H ¤ L ) ¤ tan 2 a + 0.46 ¤ tan 2 a + 1 ¤ sin 2 a }, 1 ¤ sin 2 a ]

(3.2.3)

l V = min [ 1.0, max { 1.1, 1.1 + 11d c ¤ L } 5.0 ( H ¤ L ) ] (3.2.4) where FX： total horizontal wave force acting on the sloping-top caisson breakwater (kN/m) FZ： total vertical wave force acting on the sloping-top caisson breakwater (kN/m) FSH： horizontal component of the wave force acting on the sloping section of a sloping-top caisson breakwater (kN/m) FSV： vertical component of the wave force acting on the sloping section of a sloping-top caisson breakwater; the axis extending downward is defined as positive (kN/m) FV： wave force acting on the upright section of a sloping-top caisson breakwater (kN/m) FU： uplift acting on the bottom of a sloping-top caisson breakwater (kN/m) F1： component of the horizontal wave force acting on the vertical wall corresponding to the sloping section -381-


of a sloping-top caisson breakwater, as calculated with Goda's wave pressure formula (kN/m) F2： component of the horizontal wave force acting on the vertical wall corresponding to the upright section of a sloping-top caisson breakwater, as calculated with Goda's wave pressure formula (kN/m) l SL ¢： modification coefficient of the wave force acting on the sloping section l V： modification coefficient of the wave force acting on the upright section a： gradient of the sloping section (º) p ： uplift pressure at the front toe of the ordinary caisson as calculated with Goda's wave pressure formula U (kN/m2) B： caisson width of the sloping-top caisson breakwater (m) H： wave height (m) L： wavelength (m) d c： height from the still water level to the bottom level of the sloping section (When the bottom level is above the still water level, d c is defined as positive) (m) The coefficient l SL ¢ is defined as below in the following three domains： (a)When H/L is comparatively small： l SL ¢ = sin 2 a ; thus, F SH = F 1 , F SV = F 1 tan - 1 a (b)When H/L is large： l SL ¢ = 1.0 ; thus, F SH = F 1 sin2 a , F SV = F 1 sin a cos a (c)When H/L is between 1) and 2)： lSL¢ becomes smaller as H/L increases. With regard to l V , when H/L is comparatively small, l V = 1.0, and as H/L grows larger, l V decreases. It shoud be noted however that this wave force calculation method is meant for the open sea area where the water depth is comparatively large and the design wave period is long. Before this calculation method was proposed, the wave force had been calculated assuming l SL ¢ = l V =1.0, as a simple calculation method. That calculation produced a value somewhat on the safer side, except when H/L is comparatively small. B

F1

hc

FSH α

FV

F2

h

FU

h'

dc

FSV

pU

Rubble mound

Fig. T- 3.2.5 Wave Force Acting on Sloping Top Caisson Breakwater

(2) There is a study by Sato et al. on the wave force acting on a sloping-top caisson breakwater covered with wavedissipating blocks.

3.3 Non-Gravity Type Breakwaters (Notification Article 49) For breakwaters having structural forms other than those of a gravity type breakwater, the stability shall be appropriately ensured based on the provisions of Chapter 2 Breakwaters and 3.2 Gravity Type Special Breakwaters as well as other provisions concerning structural stability. When using unique structural forms, however, the stability of the structure shall be ensured based upon appropriate hydraulic model experiments or appropriate analytical procedures corresponding to the characteristics of structures. 3.3.1 Curtain Wall Breakwater [1] General The appropriate structure for a curtain wall breakwater shall be selected by taking into consideration the wave reflection and transmission coefficients, and when required, it shall preferably be designed by conducting hydraulic model experiments. -382-


[Commentary] A curtain wall breakwater is a pile-supported structure of permeable type and has been developed for areas such as inner bays where the wave height is relatively small or where the sea bottom soils are weak and soft. This breakwater type is roughly classified into the single curtain type and the double curtain type, depending upon the configuration of the concrete curtain panel, either in a single row or in double rows (see Fig. C- 3.3.1). They can be further classified into various types depending upon the type of piles supporting the curtain panels or the shapes of the slits in the curtain panels. In addition, a breakwater without the curtain pannel, which withstands wave actions with steel piles, is called a steel pile breakwater. The design concepts for curtain wall breakwaters can be applied to steel pile breakwaters as well.

Curtain

Pile

(a) Single curtain wall breakwater

[Technical Notes] (b) Double curtain wall breakwater

Fig. C- 3.3.1 Classification of Curtain Wall Breakwaters

Wave Transmission coefficient, KT = HT/HI

(1) To study curtain wall breakwaters, Morihira et al. 6) conducted model experiments that focused on a single, solid curtain type. Based on the results, the depth of the bottom of curtain wall can be found from Fig. T- 3.3.1 when the wave transmission coefficient is prescribed, and the crown height of the curtain can be found from Fig. T- 3.3.2. The curtain crown height in Fig. T- 3.3.2 has been derived through modification of experimental data by designating R/H =1.25 for d/h = 1.0; it does not represent a crown height that can completely prevent overtopping. In the figures, d is the depth of the curtain wall bottom, h is the water depth, L is the wavelength, R is the curtain crown height, and H is the wave height. The relationship between the wave reflection coefficient and the relative curtain depth for the single curtain type is shown in Fig. T- 3.3.3.

Wave reflection coefficient, KR (%)

Fig. T- 3.3.1 Relationship between Wave Transmission Coefficient and d/h (Single Curtain)

Fig. T- 3.3.2 Crown Height Calculation Curve (Single Curtain)

Fig. T- 3.3.3 Relationship between Wave Reflection Coefficient and d/h (Single Curtain)

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/ =

Experimental values Hayashi, etal.

γ

(2) With a pile breakwater, it is possible to drive the steel piles into the ground while leaving a certain space between them so as to form a permeable-type breakwater. Based upon the study by Hayashi et al., the relationship between wave transmission coefficient KT and the ratio of pile spacing to pile diameter b/D is shown in Fig. T-3.3.4. The moment generated by the wave force is reduced as the spacing between the piles is broader, but this effect is seen only up to the level b/D = 0.1. For this form of breakwater, careful attention must be paid to scouring of the ground between the piles.

KT HT HI


Wiegel

Theoretical values Hayashi, et al Wiegel

water depth

Fig. T- 3.3.4 Relationship between Wave Transmission Coefficient and the Ratio of Pile Spacing to Pile Diameter

[2] Wave Force The wave force acting on a curtain wall breakwater of single curtain type differs from that of double curtain type, and it also varies depending upon the shapes and sizes of the slits in the curtain pannel. Because of these factors, the force shall be calculated based upon hydraulic model experiments or by an appropriate calculation method. [3] Design of Piles Calculation of the stress intensity in piles of a curtain wall breakwater shall be based on Part Ⅷ , 9.5 Design of Piles. The pile penetration depth for curtain wall breakwaters shall be calculated according to Part Ⅴ, Chapter 4 Bearing Capacity of Pile Foundations. 3.3.2 Floating Breakwater [1] General The appropriate structure for a floating breakwater shall be selected by taking into consideration the wave transmission characteristics and the stability, and then shall be designed by conducting hydraulic model experiments as necessary. [Commentary] (1) Floating breakwaters are composed of several floating bodies aligned on water to prevent wave propagation, and they have the merits that they neither hinder sea water flows or littoral drifts, nor depend on the tidal range or ground conditions, and have a mobility. At the same time, they have such shortcomings that they allow relatively large transmitted waves, their sheltering effectiveness greatly depends upon wave characteristics, they can be used only for small waves because of their limited strength against wave forces, and their resistance mechanism of anchor system against cyclic shock loads has not been well known. In addition, there is the concern that the floating bodies may drift and inflict secondary damage if their mooring lines are severed. (2) The layout and shape of floating breakwaters should be determined so that the required calmness is achieved. When determining these factors, hydraulic model experiments should preferably be conducted and the wave transmission coefficient should be measured. With regard to analytical procedures, there are the approximate calculation method of two-dimensional rectangular floating body motions based on Ito et al. 7) and the Ijima theory concerning freely floating bodies.

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[Technical Notes] Floating breakwater design can best be carried out by following the block diagram shown in Fig. T- 3.3.5.

Selection of the design conditions Required performance: wave dissipating performance: movement of floating breakwater Calculation of design external force Determination of major components : width, length Establishment of structure and cross section (Including draft and freeboard) Calculation of weight, etc. Evaluation of stability of the floating body (at static water, during towing, and installation) (Calculation of motions of a free-floating body) Calculation of motions and wave transmission coefficient

Assumption of mooring constant

Size and length of chain

Necessity of sinker and its volume

Evaluation of safety of mooring system Calculation of design load Basic design of floating body Design of fixed anchors and mooring lines (deck, bottom, siding, inclined face wall, etc.) Check for stability Detailed design Fig. T- 3.3.5 Design Sequence for Floating Type Breakwaters

[2] Selection of Design Conditions When designing floating breakwaters, the following factors shall be considered: (1) Factors Pertaining to the Function of Floating Type Breakwaters (a) Waves to be controlled (height, period, and direction) (b) Tide level and water depth (c) Required degree of calmness (2) Factors Pertaining to the Stability of Floating Type Breakwaters (a) (b) (c) (d) (e) (f) (g)

Design wave (height, period, and direction) Tide level and water depth Currents (current velocity and direction) Winds (wind velocity and direction) Ground conditions Allowable amplitudes of motions Others (vessel collisions, etc.)

(3) Design External Forces The external forces used in design shall follow Part II, Chapter 8 External Forces Acting on Floating Body and its Motions. [3] Design of Mooring System For the design related to mooring, refer to Part II, 8.3 Motions of Floating Body and Mooring Force and examine the safety of mooring system. -385-


[Commentary] (1) Mooring-related design can be divided into two stages: (a) First stage in which the tensions that will be exerted on mooring lines and sinkers are determined through static and dynamic analyses by assuming various conditions concerning mooring-related matters such as the mooring method and line length. (b) Second stage in which detailed design of the actual mooring lines and sinkers is carried out and the stability is confirmed, based on the tensions and other findings in the first stage above. (2) A mooring system should be selected by taking into consideration the matters such as the forces that will act on the system, water depth, tidal range, sea bottom conditions, waves, soil conditions, and length of the mooring lines. (3) The mooring anchor should be so designed that both its horizontal and vertical resistances are greater than the horizontal force and vertical force acting on it, respectively. (4) Dynamic analysis of the mooring lines consists of determining the fluctuating tension and displacement that arise from the motions of floating body. This analysis can be classiffied into the following two procedures: (a) Methods to analyze these factors based on the static mooring characteristics. (b) Methods to analyze these factors based on the dynamic response characteristics of mooring lines. (5) Forces acting on the mooring anchor are described in Part Ⅷ, 12.4.3 Design of Mooring Anchor.

[4] Design of Floating Body Structure The floating body shall have the sufficient strength for the safety of structure as a whole as well as for every structural member. [Commentary] (1) When selecting the material quality of the structural members of floating breakwater, it is necessary to adequately consider the characteristics of the materials and the costs. (2) The loads that should be considered from the standpoint of structural strength for floating breakwaters can be divided into static load and dynamic load, as shown below. Static load

deadweight (including weight of materials on board) hydrostatic pressure

Dynamic load

wave load (load acting as an external pressure to produce sectional force) wind load load from tidal currents and ocean currents mooring load other loads (load from towing, load accompanying installation operations, collision force, etc.)

For floating breakwaters, the following is the normal loads to be considered from the standpoint of strength analysis: Longitudinal strength. . .stresses resulting in the static water or under wave actions (longitudinal bending moment, shearing force, torsional moment). Lateral strength . . . . . . .shearing forces that occur when exposed to wave actions (bending moment, shearing force) in a direction perpendicular to the longitudinal axis of floating body. Sectional strength . . . . .stresses (same as above) that occur in wall pannels and girders. [Technical Notes] (1) Longitudinal strength calculation methods are devided into two categories, one of which considers floating body motions, and the other that does not consider. Among calculation methods that do not consider floating body motions, the Muller equation, “the Prestressed Concrete Barge Standards”, and “the Veritus Rule” are frequently used. On the other hand, the Ueda equation 8) is used as a calculation method that does take into account the floating body motions. A comparison of the methods of both categories is cited in the references 8), which can be referred to when applying the calculations. (2) Refer to Part Ⅷ , 12.3 Design of Pontoon concerning the stability of floating bodies. [References] 1)

Katsutoshi TANIMOTO, Suketo HARANAKA, Shigeo TAKAHASHI, Kazuhiro KOMATSU, Masahoko TODOROKI, Mutsuo OSATO: “An experimental investigation of wave reflection, overtopping and wave forces for several types of breakwaters and sea walls”, Tech. Note of PHRI, No. 246, 1976, 38p. (in Japanese). -386-


2) 3) 4) 5) 6) 7) 8)

Tadahiko YAGYU, Miyuki YUZA: “A compilation of the existing data of upright breakwater with wave dissipating capacity”, Tech. Note of PHRI, No. 358, 1980, 314p. (in Japanese). Yasushi HOSOKAWA, Eiichi MIYOSHI, Osamu KIKUCHI: “Experiments on hydraulic characteristics and aeration capacity of the slit caisson type seawall”, Tech. Note of PHRI, No. 312, 1979, 23p. (in Japanese). Jarlan, G. E.: “A perforated vertical wall breakwater”, The Dock and Harbour Authority, Vol. 41 No. 488, 1961, pp. 394-398. Katsutoshi TANIMOTO, Yasutoshi YOSHIMOTO: “Theoretical and experimental study of reflection coefficient for wave dissipating caisson with a permeable front wall”, Rept of PHRI, Vol. 21, No. 3, 1982, pp. 43-77 (in Japanese). Michio MORIHIRA, Shusaku KAKIZAKI, Yoshimi GODA: “Experimental investigation of a curtain-wall breakwater”, Rept of PHRI, Vol. 3, No. 1, 1964, 27p. (in Japanese). Yoshiyuki ITO, Shigeru CHIBA: “An approximate theory of floating breakwaters”, Rept of PHRI, Vol. 11, No. 2, 1972, pp. 43-77 (in Japanese). Shigeru UEDA, Satoru SHIRAISHI, Kazuo KAI: “Calculation method of shear force and bending moment induced on pontoon type floating structures in random sea”, Tech. Note of PHRI, No. 505, 1984, 27p. (in Japanese).

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Chapter 4 Locks 4.1 Selection of Location (Notification Article 58) In order that lock operations can be carried out safely and smoothly during vessel’s ingress and egress, the lock location shall be selected appropriately in response to the natural conditions at the site and the principal dimensions and number of vessels to accommondate. [Commentary] (1) Because vessel’s ingress and egress may become difficult owing to the factors such as winds, waves, tidal currents, and littoral drift, it is optimal to choose a calm water area for the lock location. In cases where the water is not calm, breakwaters should be constructed, or training jetties or guiding jetties are extended to make the water zone calm in the vicinity of the lock. (2) The size and number of vessels that will pass through the lock are also the factor in the selection of the location. That is, the lock must be located at the site where a sufficiently wide area of water can be secured for anchorage and turning basin for use by waiting vessels. (3) In addition to the above, the lock’s location must be selected with adequate consideration given to the conditions of land usage or traffic conditions of the inland area.

4.2 Size and Layout of Lock (Notification Article 52, Clause 1 and Clause 2) (1) The lock shall have a layout suitable for the natural conditions at the site and the principal dimensions and number of vessels to accommondate, in order that lock operations can be carried out safely and smoothly during vessel’s ingress and egress. (2) The lock chamber dimensions shall be determined appropriately based on the following standard dimension calculation equations. When using the following equations, appropriate values shall be employed for the depth clearance, and the allowances for width and length by giving due consideration to the motions of vessels in the lock. (a) Effective water depth = draft of vessel to accommodate + keel clearance (b) Effective width = width of vessel ´ number of vessels in tranversal direction + width allowance (c) Effective length = length of vessel ´ number of vessels in one longitudinal line + length allowance [Technical Notes] (1) The names of various parts of a lock are shown in Fig. T- 4.2.1.

Front gate chamber

Lock chamber

Rear gate chamber

Effective width of lock chamber Effective width of gate chamber

Gate recess

Effective length of lock chamber

Plane view

Lock gate

Sill height

Side view Fig. T- 4.2.1 Names of Respective Parts of Lock

(2) Generally, the clearances for the various dimensions for locks depend upon the vessel size. Fukuda, however, has proposed the following values for locks used by small vessels: Allowance for effective water depth: Allowance for effective width: Allowance for effective length:

0.2 ～ 1.0 m 0.2 ～ 1.2 m 3 ～ 10 m

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4.3 Selection of Structural Type 4.3.1 Gate (Notification Article 52, Clause 3) Lock gates shall have a structure that is safe for the size of the lock, operation time, water level difference, and external forces such as waves, and shall fulfill the requirements given below. (1) Gate’s main structure, gate supports, and surrounding fixed sections shall be safe against the loads under the normal conditions and during an earthuakes, and these loads shall be securely supported by the foundation structure. (2) Gates shall be watertight. (3) Gates shall be operated surely and smoothly. (4) Machinery parts and other moving parts shall be easy to inspect. [Technical Notes] The gates used for locks should have preventive measures against abrasion and corrosion of structural members.

4.3.2 Lock Chamber (Notification Article 52, Clause 4) The lock chamber shall have a structure appropriate to meet the conditions such as the foundation condition, water level difference between inside and outside the lock chamber, the dimensions and number of vessels to accommodate, and the quantity of water pumped into and drained out of the lock chamber.

4.4 External Forces and Loads Acting on Lock (Notification Article 52, Clause 5) The gate chambers and lock chamber of a lock shall be of structures that are safe against the foundation reaction force, the deadweights of the side walls and floor slab, the weight of the gate, and the impact force from vessels. [Technical Notes] When designing lock gate chambers and lock chambers, the deadweight of the ancillary facilities such as the operation room and the machinery rooms should be taken into consideration to the extent as necessary.

4.5 Pumping and Drainage System (Nortification Article 52, Clause 6) The pumping and drainage system for adjusting the water level in the lock chamber shall be such that its operation would not induce excessive motions of vessels inside the lock chamber, and it shall be capable of pumping and draining water in the chamber quickly.

4.6 Auxiliary Facilities (Nortification Article 52, Clause 7) Locks shall have the following auxiliary facilities installed to the extent as necessary. (1) Emergency facilities (2) Lighting facilities (3) Electric power facilities (4) Control and instrumentation facilities (5) Maintenance facilities

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Chapter 5 Facilities to Prevent Shoaling and Siltation 5.1 General In the locations where shoaling of harbors and navigation channels is expected, the mode and rate of shoaling shall be grasped through detailed investigations on the phenomena that cause shoaling. Appropriate countermeasures shall be taken after giving consideration to the various effects that shoaling prevention works will cause and to their construction and operational costs. [Commentary] Shoaling refers to the phenomenon of shallowing of basins and channels, when littoral drift, wind-blown sand, and river discharged sediment invade the inside of harbors such as navigation channels and anchorage areas, and then obstruct harbor functions by reducing water depth through sediment subsidence and deposition. There is also the case that the depth of navigation channel becomes locally shallower than the required depth with no increase nor reduction of net earth volume, as caused by the formation of sand waves 1) or by the failure of slope shoulders of a dredged navigation channel. Siltation refers to the shoaling caused by deposition of silt or other fine sediment. Shoaling occurs by the causes listed below. (a) (b) (c) (d) (e)

Invasion and deposition of littoral drift (caused by waves and/or currents) Subsidence and deposition of river discharged sediment Settlement and deposition of wind-blown sand Sediment movement within the area considered or the change in location of sediment deposition Sediment movement caused by wave agitation within a harbor, by failure of slope shoulders in navigation channel, by formation of sand waves, etc.

5.2 Jetty 5.2.1 Layout of Jetty (Notification Article 56) (1) Jetties shall be appropriately located by considering the characteristics of sediment transport, so as to exercise the expected function of longshore transport control. (2) In principal, a jetty at the updrift side shall be placed perpendicular to the shoreline within the surf zone. In the offshore zone, a jetty shall be so placed that sediment transport shall be diverted to the updrift side of harbor entrance. (3) For the case that a jetty is placed at the downdrift side of the harbor to prevent sediment from being dragged inside from the downdrift shore, the jetty shall be placed perpendicular to the shoreline, and shall have an appropriate length in consideration of the direction and deformation of incident waves. When a jetty is designed to function as a breakwater as well, it shall be placed in the appropriate location in consideration of its required function as the breakwater. (4) If a jetty is required in places such as the vicinity of a navigation channel inside a harbor, it shall be constructed in an appropriate location in consideration of the natural conditions. [Commentary] Breakwaters of a harbor located at sandy coast serve the dual functions of jetties and breakwaters, and both the functions cannot be separated. In this section the common practice is followed and refer to such structure as breakwaters, except when its function as jetties is prominent. [Technical Notes] (1) Layout of Updrift Side Breakwaters The updrift side breakwater should be extended beyond the surf zone in the direction perpendicular to the shoreline, in order to cause deposition of littoral drift at the updrift side of the breakwater (refer to Fig. T- 5.2.1). When this extension part is short or slanted towards the downdrift side from the shoreline, the efficiency of sediment catchment at the updrift side is reduced and sediment can easily move along the breakwater towards the harbor entrance. When this section is extended with a slant angle towards the downdrift side from the shoreline, it can easily become the cause of local scouring at the updrift side. In the area deeper than the breaker line, the breakwater should be slanted so that it simultaneously stops waves and disperses littoral drift toward the updrift side of the harbor entrance with the aid of reflected waves or Mach-stem waves (refer to Fig. T- 5.2.1) 2).

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(2) Position of the Downdrift Side Breakwater and Construction Time When the updrift side breakwater is extended and crosses the extension line of the downdrift side breakwater, deposition will start at the downdrift side of the latter breakwater. Sand bar will then be formed from the shore toward the harbor entrance, and it will cause beach erosion at the far downdrift shore. 3) On the other hand, if the downdrift side breakwater is extended during construction of the updrift breakwater and the slant section of the latter is not extended enough, remarkable local erosion may be caused at the harbor side of the downdrift breakwater, as shown Fig. T- 5.2.2 (a). Conversery, if the extension of the downdrift breakwater is delayed, it may cause deposition in the harbor and erosion at the downdrift shore as shown in Fig. (T) 5.2.2 (b). Very carefull attention should therefore be paid to the extension speed of both the updrift and downdrift side breakwaters, and care must be taken to maintain the appropriate balance of extensions. Contour lines

Waves

Deposition

breakwater

Longshore sediment transport

breakwater

Breaker line

Breakwater (jetty)

Reflected waves


Fig. T- 5.2.1 Conceptual Layout of Breakwater (Jetty)

Deposition

Accretion

Erosion

Erosion (a) Case with rapid extension of the downdrift-side breakwater

(b) Case with slow extension of the downdrift-side breakwater

Fig. T- 5.2.2 Construction Time of Downdrift Side Breakwater

(3) Length of Breakwater and Water Depth at Its Tip Because longshore sediment transport occurs mainly in the surf zone, breakwaters should be extended offshore beyond the surf zone. In small harbors where the water depth at the end of the breakwater remains within the surf zone during storms, it is difficult to completely prevent littoral drift from entering the harbors. At major ports in Japan, it is common that the water depth at the end of the updrift side breakwater is almost equivalent to the maximum depth of the navigation channel, which sometimes exceeds 20 m.

5.2.2 Details of Jetty (Notification Article 50) A jetty shall have an appropriate crown height to ensure their required function of sediment control in consideration of the suspended sediment transport into the inside of a harbor by wave overtopping, and shall retain structural stability in light of the influence of external forces and loads exerted on the jetty. [Technical Notes] (1) Structural Forms of Jetty Because the required function of a jetty is to stop sediment transport firmly, in principal a jetty should have a nonpermeable structure. Where rubble stones or concrete blocks are used to build a jetty around the shoreline, the core is to be filled with quarry run or small stones of up to 100 to 200 kg; there are also cases where the harbor side of the jetty is covered with nonpermeable materials such as sand mastic asphalt. In the following situations, it is preferable to adopt the structure of wave absorbing types. (a) When there is a large concern about scouring by currents. (b) When there are concerns of shoaling caused by reflected waves or of causing obstruction to the navigation of vessels. (2) Crown Height of Jetty Although it is preferable for jetties not to allow overtopping of waves to prevent the inflow of suspended sediment, there are also cases where overtopping is permitted due to structural restrictions or by reasons of construction costs. The crown height should be determined by taking the following into considerations: -391-


(a) Section around the shoreline A runup wave should not be allowed to overtop the jetty. Because sand carried by runup waves may overtop the jetty crown if it is low, the shore connection part should have a crown that is sufficiently high. It is preferable to execute a raise of crown and a landward extention of jetty, examing the situation after construction. (b) Section that is located between the shoreline and the breaker line It should be standard to use 0.6H1/3 above the mean monthly-highest water level (HWL), where H1/3 should be the significant wave height around the tip of jetty. (c) Section that is located outside the breaker line It should be standard to use a height that is obtained by adding some freeboard to the mean monthly-highest water level. In the water deeper than the breaker zone, the suspended sediment is concentrated near the seabed and overtopping water contains almost no sediment, and therefore overtopping can be permitted.

5.3 Group of Groins [Technical Notes] The length, interval, and structure of groups of groins that are constructed on the updrift side of harbors to prevent shoaling should be determined in accordance with the “Manual for Construction of Coastal Protection Facilities”.

5.4 Training Jetties 5.4.1 Layout of Training Jetties (Notification Article 57, Clause 1) To fulfill their function, training jetties shall be appropriately located in consideration of the characteristics of the longshore sediment transport at site and the tractive forces of rivers during flood and drought periods. [Technical Notes] (1) Multiplicity of Functions of River Mouth Training Jetties River mouth training jetties should have a groin-like function to stop longshore sediment transport, in additions to the function to stabilize a river’s waterway by guiding the river flow and to maintain water depth by increasing the river flow tractive force, while simultaneously having the ability to release and not to hinder the discharge during flood period. (2) Layout of Training Jetties Examples of the layout of training jetties in relation to the direction of longshore sediment transport are shown in Fig. T5.4.1 4). The most preferable one for maintaining the water depth of river mouth is to extend two parallel training jetties, because a single training jetty alone is not effective. Where two training jetties of different lengths are put in place, usually it is effective to make the training jetty on the downdrift side longer. Bending the updrift training jetty towards the downdrift side will prevent sediment moving into the river’s waterway and make the sediment transported alongshore pass smoothly to the downdrift side. For actual examples of river mouth improvement, refer to the reference 5).

A narrow but deep gut is preserved Grows shallow Longshore sediment transport

Longshore sediment transport River mouth will move towards the downdrift side

A narrow but deep gut is preserved Grows shallow Longshore sediment transport

Longshore sediment transport

Longshore sediment transport Fig. T- 5.4.1 Varieties of Training Jetty Layout 4)

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5.4.2 Water Depth at Tip of Training Jetty (Notification Article 57, Clause 2, Clause 3) (1) The water depth at the tip of training jetty shall be equal to or deeper than the water depth of the navigation channel in the vicinity of the training jetty. (2) The tip of training jetty shall be located in the offshore zone of the breaker line. 5.4.3 Structure of Training Jetty (Notification Article 51) Training jetties shall be of impermeable structures in principal, and they shall be so built that the structural stability is maintained by taking into consideration the influence of scouring by waves and the river flow acting on them. [Commentary] Because the training jetty is generally longer than groins and is exposed to intensive wave actions, it is necessary to consider scouring at the head and sides of a jetty. In addition, it should be considered that the river side of the training jetty will be subject to scouring action by the river flow.

5.5 Facilities to Trap Littoral Transport and Sediment Flowing out of Rivers [Technical Notes] When it is aimed to prevent shoaling due to littoral drift by means of maintenance dredging, an appropriate facility to trap the sediment should be built at a proper location, at which the facility can prevent sediment from invading to navigation channels or basins. The facility should be able to reduce the wave actions around it and increase the dredging efficiency. The type and layout of these sand trap facilities must be determined by taking into consideration their capability to trap the sediment, the dredging conditions, and the construction and operational costs, based on adequate investigations and researches. (1) Facilities to Trap the Sediment Transport As the method to trap the sediment, provision of a calm water area for sediment subsidence and deposition is commonly employed in various countries, by means of building a detached breakwater or partially reducing the crown heigh of updrift breakwater. There are also sediment trap facilities such as pocket dredging executed in the navigation channel crossing a large sand bar in the sea floor of straits, which is gradually restored by the natural process after dredging. Pocket dredging is also done on the river bed, where shoaling occurs by river discharged sediment. (2) Proper Location of Facilities to Trap the Sediment Transport There are two methods to place the facility. The one is to locate it at the area where deposition occurs easily under the natural condition as shown in Fig. T- 5.5.1 (a), (b), and (c). The other is to create the artificial situation that leads to settlement of sediment within the zone of intensive sediment transport as shown in Fig. T5.5.1 (d), (e), and (f). To identify the specific locations suitable for trapping sediment in the most efficient manner, understanding of the conditions and mechanism of sediment transport is indispensable. In addition to the efficiency of trapping sediment, it is also the important factor to select the place of the facility that the conditions are good for dredging operations of deposited sediment; i.e., it is easy to keep sufficient depth for navigations of dredgers, and the dredging area should be calm enough to allow easy dredging works.

Sea floor sand bar

Pocket dredging River mouth port

Pocket dredging Navigation channel

(b)

(a)

Bre

(c)

ake

(d)

rs

Submerged breakwater

(e)

(f)

Fig. T- 5.5.1 Locations of Facilities for Sediment Trap -393-


5.6 Countermeasures against Wind-Blown Sand 5.6.1 General In situations where wind-blown sand causes a problem of shoaling of harbors and navigation channels or where there is a need to protect the surrounding regional environment from wind-blown sand, appropriate prevention measures against wind-blown sand shall be taken as necessary. [Commentary] Wind-blown sand, i.e., sand that is moved by winds, is carried into harbors or navigation channels where it settles and deposits, and cause shoaling there. In some cases it also accumulates on road surfaces and is dispersed into residential areas, disrupting the daily living of the resident. In particular, there are many instances that open digging of dune or land reclamation cause problems related to wind-blown sand, and thorough countermeasures must be prepared in advance.

5.6.2 Selection of Countermeasures Countermeasures against wind-blown sand shall be determined with thorough understanding of the characteristics of each countermeasure in consideration of the present conditions and forecast of future conditions of wind-blown sand. [Technical Notes] The following methods are commonly used as prevention methods against wind-blown sand: (1) Sand Fences and Windbreaks It is a conventional method to install sand fences with the height of about 1 m in multiple rows to arrest wind-blown sand, and to expect an artificial dune growing up as a sand protection work. In recent years, there have been observed many cases to install high windbreaks sorrounding grain storage areas or in the vicinity of reclamation areas. (2) Multiple Low Fences This is a method to suppress the movement of surface sand by installing multiple low fences that increase the roughness of land surface and reduce the wind shearing forces. (3) Covering Works This is a method to cover the sand surface with some artificial material and thus restrict sand movement.

First stage: Creation of an artificial dune Sedimentation fence, step 1

Sedimentation fence, step 2

Sedimentation fence, step 3

Second stage: Covering of dune surface

Temporary covering works

(Multiple low fences)

(Transplanting of plants) (Plantation works) Tree cover

Third stage: Afforestation (Trees and shrubs)

Fig. T- 5.6.1 Steps of Artificial Dune Creation Works

(4) Plantation Works This is s method to grow plants on sand surface with which sand surface is covered. This is thought to be one of covering works. (5) Afforestation Works This is a method in which trees are planted at the downwind side of a wind-blown sand deposition, thereby serving to prevent wind-blown sand. Prevention measurs against wind-blown sand that are applied in view of so-called coastal erosion control, with the purpose to stabilize the shape of sand beaches, are to be carried out in combination of various methods. The sequences and methods are generally shown in Fig. T-5.6.1. For details, see the reference 6). In addition, refer to the “Civil Engineering Handbook” 7) concerning trees and shrubs highly suitable for coasts. [References] 1)

Hiroaki OZASA: “Field investigation of submarine sand banks and large sand waves”, Rept of PHRI, Vol. 14, No. 2, 1975, pp. 3-46. (in Japanese). -394-


2) 3) 4) 5) 6) 7)

Norio TANAKA: “Change in seabed and beach near by sand beach port”, PHRI, Seminar Proceeding, 1974, pp. 1-46. (in Japanese). Shoji SATO, NorioTANAKA, Katsuhiro SASAKI: “Change in seabed configuration during construction of Kashima Port (case study)”, Rept of PHRI, Vol.13,No.4, 1974, pp. 3-78. (in Japanese). Japan Soc. Civil Engrs.: “Civil Engineering Handbook”, 1974, pp. 2268-2270 (in Japanese). Taka-aki UDA, Akira TAKAHASHI, Hideaki MATSUDA: “Charactaristic of land configuration or river mouth and investigation on prevension measures to deformation in Japan”, PWRI Report, No.3281, 1994, 123p. (in Japanese). Kazuo TANAKA, Yuki NAKAJIMA, Jiro ENDO, Eiji KANEUCHI: “Erosion Control of Shore”, Erosion Control Series Ⅲ 9, The Japan Society of Erosion Control Engineering, Ishibashi Books, 1985 (in Japanese). Japan Soc. Civil Engrs.: “Civil Engineering Handbook”, 1974, pp. 2718-2720 (in Japanese).

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Chapter 6 Revetments 6.1 Principle of Design The following matters shall be examined when designing revetments: (1) Adequate crown height that can protect reclained land from waves and storm surges while not hindering land utilization. (2) Structure that is stable against external forces such as wave forces and earth pressure. (3) Structure that prevents landfill leakage. (4) Consideration to influences on surrounding water areas, including prevention of flowing out of muddy water during reclamation works. (5) Safe and comfortable structure for users in the case of amenity-oriented seawall. [Technical Notes] In green park areas, revetments should be preferably designed by taking into consideration the water frontage amenity aspects 1). (1) Amenity-oriented seawall should be designed with cross sections which prevents users from falling into the sea. In addition, supplemental facilities such as handrails to keep users from falling off must be appropriately installed. (2) For facilities where high waves are expected to cause wave overtopping and to rush up water to the place of user’s normal stroll, appropriate measures such as warning signs should be provided to alert users to the danger. (3) Refer to the manual 2) for designing amenity-oriented seawalls. (4) The width and gradient of the passageway along the revetment should preferably be designed to ensure the safe movement by wheelchairs used by elderly and/or handicapped people 3).

6.2 Design Conditions The following design conditions are usually taken into consideration: (1) Waves, tidal level, tidal currents, water depth (2) Subsoil conditions. (3) Earthquakes and seismic force (4) Dynamic water pressure during an earthuakes (5) Properties of soil for landfill (6) Utilization circumstances of reclaimed land and revetments (7) Allowable quantity of wave overtopping (8) Surrounding water area conditions (9) Construction method (particularly the method of closure) (10) Reclamation method [Technical Notes] (1) When determining the crown height and structure of revetment, the allowable overtopping rate is the critical factor. Allowable overtopping rate should be appropriately set in consideration of the conditions behind the revetment. For more details, refer to Part II, ,4.6.2 Wave Overtopping. (2) Refer to Part II, , Chapter 11 Subsoil with regard to landfill soil conditions. (3) Refer to Part II, , Chapter 12 Earthquakes and Seismic Force with regard to earthquakes and seismic force. (4) Refer to Part II, , Chapter 14 Earth Pressure and Water Pressure with regard to dynamic water pressure during an earthquake. (5) In the case of reclamation where a large water area is enclosed by revetments, the opening becomes smaller with the progress of revetment construction, and the considerable rapid flow occurs at closing points that is caused by difference of water levels between the inside and outside of revetments. Therefore, careful consideration is required for structure of revetments at the final closing point, which should have enough stability against the calculated flow speed. (6) As for the water level inside the enclosed reclamation area, two kinds of water levels are to be considered, i.e., the water level inside reclamation and the residual water level after landfill. The water level inside reclamation is used in calculation of seepage flow or in design of a waste water treatment plant. Residual water level is the -396-


water level immediately behind the revetment, which is used for stability calculation of the revetment. But, when the residual water level is used in the calculation of circular slip failure, the calculation may underestimate the danger of circular slip failure if the inside water level near the revetment is higher than the residual water level. In the such case, stability of revetment should be calculated using the water level inside the reclamation. (a) Water level inside reclamation The water level inside reclaimation should be established by considering the stability of revetment both during the consruction and after completion, and the influence on the surrounding water. (b) Residual water level ① For reclamation revetments, the structures with low permeability are often used to reduce pollution due to

seepage through revetments. For this reason, the residual water level behind them is generally higher than that behind quaywalls or ordinary revetments. ② According to the past design examples of reclamation revetments of gravity type, there are more cases to decrease permability by increasing the thickness of backfilling material than the cases to provide the revetment itself with structures of low permability. Thus in the former type of revetments, the residual water level to be used in design may be set equivalent to the water level behind gravity type revetments, because the water level immediately behind such the redamation revetments structure show the behavior similar to that of gravity type revetments. ③ For reclamation revetments using a sheet pile, there are examples where grout material is poured into the

sheet pile joint or a double sheet pile structure is used to increase the waterproofness. For these cases, the residual water level for design use should be carefully determined in consideration of permeability of the revetment as well as the construction procedures. (7) In cases where a reclamation revetment is built adjoining to the existing land area, construction of the revetment may cause the groundwater level to rise or may result in deterioration of groundwater quality. Adequate attention should be given to these aspects when studying the reclamation layout plan and revetment structure. It is preferable to investigate the conditions of the groundwater in the land area in advance. In addition, in cases where it is thought that reclamation revetment construction will cause deterioration of the groundwater quality, countermeasures such as construction of a cutoff wall must be considered in order to insulate the groundwater of the land from the reclaimed area. (8) The seepage analysis shall be made in order to estimate the rate of seepage that will flow to the sea area through the reclamation revetment. Darcy’s law should be applied to the seepage analysis. (a) Permeability of steel sheet pile structures The permeability of steel sheet pile structures cannot be derived from Darcy’s law. However, it can be applied by using an appropriate equivalent width and the equivalent coefficient of permeability for that width. In addition, because it cannot be assured that a laboratory test could reproduce the joint conditions of the protorype structure in proper scale, it is preferable to use the values measured in situ. ① Reference 4) is available concerning the permeability of steel sheet pile type structures. It describes the

result of analyses taking into account the in-situ measurements on residual water levels at five project sites. In the analyses, it was assumed that the sheet pile wall below the seabed are impermeable and the part of wall above the seabed is equivalent to the permeable layer of 1 m thick to which Darcy’s law can be applied. The results obtained for the coefficient of permeability (equivalent coefficient of permeability) were in the range of 1 ´ 10-5 ～ 3 ´ 10-5 cm/s. The results of the similar analysis carried out for two examples of steel pipe pile type quaywall (diameter of approximately 80 cm) yielded a value of 6 ´ 10-5 cm/s. It should be mentioned that the coefficient of permeability for backfilling material of the foregoing surveys was in the range of 10-2 ～ 10-3 cm/s. ② The permeability of sheet pile joint has the following characteristics:

In cases without backfilling material, the sheet pile joint has the similar nature as a narrow orifice of abrupt sectional reduction, and can be expressed in equation (6.2.1) with the constant n = 0.5 5),6). n

q = K×h where: q： flow rate per unit joint length (cm3/s/cm) h： difference in the water level between the front and the rear of the sheet pile (cm) K, n： constant

(6.2.1)

In cases with backfilling material, the property of the backfilling material greatly affects the quantity of seepage through the joint. In the vicinity of the backfilling material behind the sheet pile joint, there are spots at which Darcy’s law cannot be applied. There has been an effort to evaluate the permeability as a composite joint that includes a certain thickness of backfill and sheet pile joint. This idea is effective for seepage analysis. Shoji et al. 7) proposed an empirical equation based on the experiments considering both the difference in the degree of tensile force in the joint and conditions with or without sand filling. From the result of experiment, for the case that there is backfilling and joints are filled with sand, it was found that the -397-


constant n could be given an approximate value of 1.0 and the K value representing the result of experiment was derived. ③ In cases where a waterproof measure is applied at a sheet pile joint, the extent of permeability reduction will

differ depending upon the type of sealing material and its method of application and should be judged with reliable experimental results that take into consideration the construction conditions at site. Some examples of field test have shown that the rate of seepage was reduced to the level of 20% to 40% of the rate without water sealing material 8). (b) Permability of foundation ground ① Permeability of natural ground

Regarding the permability of natural ground, the coefficient of permability for each soil layer which composes the natural ground should be calculated by referring Part II, 11.2.3 Coefficient of Permability of Soil. ② Permeability of soil improvement sections

In situations where soil improvement is applied when constructing a reclamation revetment, it is necessary to investigate the change in permeability caused by the soil improvement in addition to evaluating the permeability of natural ground. ③ In case that the foundation is made of rocks, careful investigations and consideration of permability should be required, because the rock foundation may contain cracks or fault, which govern the rate of seepage 9).

6.3 Structural Stability (Notification Article 53, Clause 2) The calculation on stability for revetments shall be carried out according to Part Ⅷ , Chapter 4 Gravity Type Quaywalls and other provisions concerning the stability of structures, corresponding to the revetment structural types.

6.4 Determination of Cross Section (Notification Article 53, Clause 1) Revetments shall have an appropriate crown height with due consideration to the factors such as the overtopping rate and tidal level during storm surge, so that preservation of the land area behind the revetment shall be secured and the use of the revetment and the land behind it shall not be hindered. [Commentary] The crown height of revetments should have the required height against the design wave at the design storm tide level with addition of a freeboard, and should be decided considering the allowable overtopping rate (refer to Part II, 4.6.2 Wave Overtopping). [Technical Notes] In case of reclamation by pump dredgers, suspended soft sediment is concentrated behind a revetment and there are instances in which the earth pressure greater than anticipated pressure may act on the revetment or the hydrostatic pressure behind the revetment may act to the elevation as high as the revetment crown. These must be given adequate consideration when calculating the stability.

6.5 Details (1) Depending upon the wave conditions, scour-prevention works and apron works on and/or behind the revetments shall be provided. (2) Appropriate protection against seepage shall be made in consideration of the nature of reclamation material, structure of revetment, and residual water level, etc. (3) Ancillary facilities such as stairs shall be installed if necessary. [References] 1) 2) 3) 4) 5) 6)

Japan Soc. Civil Engrs.: “Landscape Design of Port”, Gihodo, December 1991 (in Japanese). Coastal Development Institute of Technology: “Port Environment Improvement Facilities Manual”, 1991 (in Japanese). Institute for Transport Policy Studies: “Guideline for Facility Building for Elderly and Handicapped Peoples in Public Passenger Terminals”, 1994 (in Japanese). Mitsuaki FURUDOI, Takeshi KATAYAMA: “Investigation of residual water levels”, Report of PHRI, No.115, 1971 (in Japanese). Koichi KUBO, Mamoru MURAKAMI: “Experiment on watertightening of sheetpile walls”, Soil and Foundation, Vol.11. No.2, 1963 (in Japanese). Kazuya YAMAMURA, Tadahiko FUJIYAMA, Masao INUSOKU, Kenjiro NIBA: “Experiment on watertightening of sheetpile walls”, Public Works Res. Inst. Report, Vol.123, No.3, 1964 (in Japanese). -398-


7) 8) 9)

Yoshihiro SHOJI, Seiji KOMEDA, Yukiharu TOMITA: “Experiment on permability of sheetpile walls”, Report of PHRI, Vol.21, No.4, 1982, pp.41-82 (in Japanese). Nippon Steel Co.,: “Report on Watertightening Test of Sheetpile Walls”, 1969 (in Japanese). “Rock Engineering for Civil Engineers”, Gihodo, 1975, pp.238-254 (in Japanese).

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Part VIII Mooring Facilities

PART VIII MOORING FACILITIES

Part VIII Mooring Facilities Chapter 1 General 1.1 General Consideration In the design of mooring facilities, careful consideration shall be given to their relationship with navigation channels, basins, and protective facilities, their influence on topography, harbor and other facilities, and the environment in the neighboring locations, as well as the space for the future development of the port. [Commentary] (1) The types of mooring facilities are specified in Article 2, Clause 5 of the Port and Harbour Law. These are quaywalls, mooring buoys, mooring piles, open-type wharves, floating piers, small craft quays, and slipways. (2) In determination of the dimension of a mooring facility, the cargo and passenger trend, the ship size trend, and the change of transport system in future should be carefully examined. (3) Layout of mooring facilities should be determined so that the ship maneuvering be easy for berthing and unberthing by carefully considering marine and atmospheric conditions, topography and subsoil conditions. The land transport network and the neighboring land utilization should also be carefully studied to determine the layout. In particular, when locating the following types of mooring facilities, the requirements specified in the respective paragraphs should be satisfied: (a) Mooring facilities used by passenger ships should be isolated from the areas where hazardous cargoes are handled, and there should be secured a sufficient area of land in the vicinity of the facilities for waiting room and parking lots. (b) Mooring facilities used by vessels loaded with hazardous cargoes should be located in accordance with the following conditions: ① They are isolated from such facilities as housings, schools, and hospitals. ② The required safety distance from other mooring facilities and sailing vessels is secured. ③ Countermeasures against spills of hazardous materials are easily mobilized. (c) Mooring facilities at which considerable amount of noise may be generated by vessels or cargo handling equipment should be isolated from such facilities as housings, schools, and hospitals to preserve the good environment for daily living. (d) Mooring facilities at which conspicuous dust and offensive odours may be generated during cargo handling works should be isolated from such facilities as housings, schools, and hospitals to preserve the good environment for daily living. (e) Offshore mooring facilities should not hinder the navigation or anchorage of vessels.

1.2 Maintenance of Mooring Facilities (Notification Article 77) In principle, mooring facilities shall be maintained in accordance with appropriate standards in consideration of the structural characteristics of the facility so that the facilities can perform the required function. [Technical Notes] For the inspection and/or assessment of damage, and repair of mooring facilities, refer to PartⅠ Ⅰ, Chapter 3 Maintenance.

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Chapter 2 Dimensions of Mooring Facilities 2.1 Length and Water Depth of Berths (Notification Article 64, Clause 2 and 3) (1) When the target vessel can be identified, the length and water depth of berth shall be determined as follows: (a) The length of a berth shall be set in principle by adding the lengths of the bow and stern mooring ropes to the length overall of the target vessel. (b) The water depth of a berth shall be set in principle by adding an appropriate keel clearance to the maximum draft. (2) When the target vessel cannot be identified, appropriate length and water depth shall be used. (3) The datum level to measure the water depth of mooring facilities shall be the datum level for construction work stipulated in PartⅠ Ⅰ, Chapter 2 Datum Level for Construction Work. [Technical Notes] (1) The standard length of a berth is obtained by adding the required lengths of the bow and stern mooring ropes to the length overall of the target vessel considering that the berth serves a single target vessel. (2) When a vessel is moored parallel to a quay, the mooring ropes shown in Fig. T- 2.1.1 are required. The bow and stern ropes are usually set at an angle of 30 to 45º with the quay face, because these ropes are used to prevent both the longitudinal movement (in the bow and stern directions) and lateral movement (in the onshore and offshore directions) of the vessel. (3) The standard depth of a berth is obtained by adding an appropriate keel clearance to the maximum draft of target vessel. For the keel clearance, a value equal to about 10% of the maximum draft is used as a standard. For a mooring facility used by vessel under storm weather conditions, margins for the ship motions by winds and waves should be added to the keel clearance. (4) When the target vessel cannot be identified in advance, such as in the case of a port facility for public use, the dimensions of the berth may be determined using Table T- 2.1.1 as reference. The premise here is that the vessel is moored parallel to the quay. For ferryboats, the dimensions of berths for bow or stern mooring are also listed. (5) Table T- 2.1.1 lists the water depth of berth based on Table T- 2.1.1 of PartⅡ Ⅱ, 2.1 Dimensions of Target Vessel using a minimum unit of 50 cm. In designing a berth, it is necessary to consider the draft and length of the vessel and other factors in an appropriate manner. (6) With regard to ferryboats, the dimensions of berths for “long-distance ferryboats” (sailing distance of 300 km and more) and those for “short-distance ferryboats” (sailing distance of less then 300 km) are given in different tables, because the vessel dimensions of two types differ eath other. (7) With regard to passenger ships, the dimensions of Japanese passenger ships and those of foreign passenger ships are given in different tables, because the dimensions of two types differ each other. (8) In case of a berth where flammable dangerous cargoes are handled, it is necessary to secure a distance of 30 m or more from oil tanks, boilers, and working areas that use open fire, to the cargo handling area and the mooring vessel at the berth. However, when there is no risk of catching fire of the cargo in the event of leakage, because of the surrounding topography or structure of facilities of the berth, the distance may be shortened to around 15 m. (9) In case of a berth where flammable dangerous cargoes are handled, it is necessary to secure a distance of 30 m or more from other anchored vessels and also to secure a distance of 30 m or more from other vessels navigating nearby to provide a room for their maneuvering. However, the distance may be increased or decreased as necessary in consideration of the size of cargo carrying vessel, the type and size of vessels anchored or navigating nearby, and the status of ship congestions.

bow line stern line spring lines breast lines

(10) When the target vessel is a small cargo ship of Japan and its dimensions cannot be known in advance, the berth dimensions listed in Table T2.1.2 may be used as a reference. However, care should be exercised when using these figures, as the dimensions of small cargo ships of Japan vary widely.

Fig.T- 2.1.1 Arrangement of Mooring Ropes

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(11) The main dimensions of berths for container ships and ferryboats in Table T- 2.1.1 are also listed in Part Ⅹ, 1.2.1 Length and Water Depth of Berths and Part Ⅹ, 2.2.1 Length and Water Depth of Berths. For details, refer to the respective sections. Table T- 2.1.1 Standard Main Dimensions of Berths for Cases Where the Target Vessels Cannot Be Identified

1. Cargo Ships Deadweight tons (DWT)

Length of berth

Water depth of berth 4.5 m 5.5 6.5 7.5 9.0 10.0 11.0 12.0 13.0 14.0 15.0 16.0 17.0 19.0

80 m 100 110 130 160 170 190 240 260 280 300 320 330 370

1,000 2,000 3,000 5,000 10,000 12,000 18,000 30,000 40,000 55,000 70,000 90,000 100,000 150,000

2. Container Ships Deadweight tons (DWT)

Length of berth

Water depth of berth 12.0 m 13.0 14.0 15.0

250 m 300 330 350

30,000 40,000 50,000 60,000

3. Japanese Ferries 3-A Short- and Medium-Distance Japanese Ferries (sailing distance less than 300 km) Wharf for ferries with bow and/or stern ramps Gross tons (GT)

Quay length of bow and stern side

Length of berth

3.5 m 4.0 4.5 5.5 6.5 7.0

60 m 80 90 130 160 180

20 m 20 25 25 30 30

400 700 1,000 2,500 5,000 10,000

Water depth of berth

3-B Long-Distance Japanese Ferries (sailing distance 300 km or more)

Gross tons (GT)

Wharf for ferries with side ramps Length of berth

6,000 10,000 13,000 16,000 20,000 23,000

190 m 220 240 250 250 260

Wharf for ferries with bow and/or stern ramps Quay length of bow and stern side 30 m 30 35 40 40 40

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Length of berth 170 m 200 220 230 230 240

Water depth of berth 7.0 m 7.5 8.0 8.0 8.0 8.5


4. Japanese Roll-on Roll-off Ships Deadweight tons (DWT)

Length of berth 100 m 130 150 180 200 230

700 1,500 2,500 4,000 6,000 10,000


5. Passenger Ships 5-A Passenger Ships that Belong to Japan (passenger ships with Japanese flag) Gross tons (GT)

Length of berth 110 m 140 160 190 240 280

2,000 4,000 7,000 10,000 20,000 30,000


5-B Passenger Ships that Belong to Other Countries Gross tons (GT) 20,000 30,000 50,000 70,000

Length of berth 220 m 260 310 340

Water depth of berth 9.0 m 9.0 9.0 9.0

6. Japanese Pure Car Carriers Gross tons (GT) 500 1,500 3,000 5,000 12,000 18,000 25,000

Length of berth 90 m 120 150 170 210 240 260

Water depth of berth 4.5 m 5.5 6.5 7.5 9.0 10.0 11.0

7. Tankers Deadweight tons (DWT) 1,000 2,000 3,000 5,000 10,000 15,000 20,000 30,000 50,000 70,000 90,000

Length of berth 80 m 100 110 130 170 190 210 230 270 300 300

Water depth of berth 4.5 m 5.5 6.5 7.5 9.0 10.0 11.0 12.0 140. 16.0 17.0

Table T- 2.1.2 Standard Main Dimensions of Mooring Facilities for Cases Where the Target Vessels Cannot Be Identified (for Japanese Small Cargo Ships) Deadweight tons (DWT) 500 700

Length of berth 60 m 70

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Water depth of berth 4.0 m 4.0


2.2 Crown Heights of Mooring Facilities (Notification Article 64, Clause 1) The crown of a mooring facility shall be set at an appropriate height that is suitable for the main dimensions of the target vessel and the natural conditions of the surrounding area. The tidal level that is used as the datum in the determination of the crown height shall be the mean monthly-highest water level. [Technical Notes] (1) In cases where the target vessels cannot be identified, the values listed in Table T- 2.2.1 can be used as a standard. Table T- 2.2.1 Typical Crown Heights of Mooring Facilities above the High Water Level When the tidal range is 3.0 m or more

When the tidal range is less than 3.0 m

Mooring facilities for large vessel (with a water depth of 4.5 m or more)

＋ 0.5 ～ 1.5 m

＋ 1.0 ～ 2.0 m

Mooring facilities for small vessel (with a water depth of less than 4.5 m)

＋ 0.3 ～ 1.0 m

＋ 0.5 ～ 1.5 m

2.3 Ship Clearance for Mooring Facilities The wall and front toe of the quaywall shall not contact with a berthing vessel. [Technical Notes] (1) In the cross sections of a vessel, the bottom corner sections are slightly rounded and have bilge keels projecting from them. In many cases, the radius of curvature of the corner sections and the height of the bilge keels are 1.0 to 1.5 m and 30 to 40 cm, respectively. Therefore the envelope of corner sections may be assumed nearly 90º including the bilge keels. The planned water depths of berths are generally deeper than the full-load draft of the target vessel by 0.3 m or more. (2) Figure T- 2.3.1 shows the ship clearance for mooring facilities that is set by taking into consideration the above facts and past examples 1), 2). The ship clearance of mooring facilities may be determined using this figure as reference. However, care will normally be exercised in using the ship clearance shown in the figure, because the rolling, pitching and heaving motions of vessel at berth have not been taken into consideration in the figure.

height of fender from the face line of quaywall when it is compressed

Proposed water depth

Face line of quaywall

Construction limit line

Fig. T- 2.3.1 Ship Clearance for Mooring Facilities

2.4 Design Water Depth A design water depth of a mooring facility shall be determined by considering its planned water depth as well as the structural type, the original sea bottom depth, the method and accuracy of execution of work, and the scouring susceptibility in front of the mooring facility. [Commentary] (1) In general, the design water depth is not equal to the planned water depth. The design water depth is ordinarily obtained by adding a margin to the planned water depth in order to guarantee the required stability of the mooring facility. As this margin varies according to the structural type, the water depth of the site, the construction method and accuracy, and the scouring susceptibility, it is important to determine the design water depth carefully in consideration of these factors.

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(2) When it is difficult to determine the depth of scouring due to berthing vessels or by currents, it is advisable to provide scour prevention measures as described in 2.5 Protection against Scouring.

2.5 Protection against Scouring When large scouring is anticipated in front of a mooring facility due to currents or turbulence, the front of the mooring facility shall be protected with armour stones, concrete blocks, or other materials against scouring.

2.6 Ancillary Facilities Refer to Part Ⅷ , Chapter 19 Ancillary Facilities. [References] 1) Japan Port and Harbour Association, “Port Construction Work Design Handbook”, 1959, p.215 (in Japanese). 2) Shigekazu MIYAZAKI, “A study on wharf construction limits”, Jounal of the Japan Society of Civil Engineers, Vol. 36, 8, 1951, pp.26-27 (in Japanese).

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Chapter 3 Structural Types of Mooring Facilities The structured types of mooring facilities shall be determined by considering their structual characteristics and examining the following matters: (1) Natural conditions (2) Usage conditions (3) Conditions of construction works (4) Construction and maintenance costs (5) Employment of different structural types in a port [Technical Notes] Mooring facilities are normally classified according to their structural types as in the following: (1) Gravity type quaywalls (2) Sheet pile quaywalls (3) Sheet pile quaywalls with relieving platform (4) Cellular-bulkhead type quaywalls (a) Steel sheet pile cellular-bulkhead quaywall (b) Steel plate cellular-bulkhead quaywall (5) Open-type wharves (a) Open-type wharves on vertical piles (b) Open-type wharves on coupled raking piles (6) Detached piers (7) Floating piers (8) Dolphins (9) Slipways and shallow draft quays (10) Air-cushion vehicle landing facilities (11) Mooring buoys and mooring posts

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Chapter 4 Gravity Type Quaywalls 4.1 Principle of Design Gravity type quaywalls are classified into caisson type quaywalls, L-shaped block type quaywalls, rectangular concrete block type quaywalls, cellular concrete block type quaywalls, and cast-in-place concrete type quaywalls. The provisions in this chapter shall apply to all of these gravity type quaywalls. [Technical Notes] (1) Figure T- 4.1.1 shows an example of gravity type quaywalls.

HWL LWL

Filling sand

Sand erosion prevention plate Backfilling stones Sand erosion prevention sheet nd

g grou

Existin

Proposed water depth Current water depth

Fig. T- 4.1.1 Example of Cross Section of Gravity Type Quaywall

(2) A wedged bottom caisson quaywall whose bottom is slanted downward in the onshore direction has an advantage over the normal quaywall, because a large resistance force can be achieved against the horizontal load (such as earthquake forces). On the other hand, it is necessary to give careful consideration to the large bottom reaction force and rocking motion during an earthquake in the design process. For design methods, see references 1) and 2).

4.2 External Forces and Loads Acting on Walls As the external forces and loads acting on the wall of a gravity type quaywall, the following shall be considered: (1) Surcharge (2) Deadweight of the wall (3) Earth pressure and residual water pressure (4) Buoyancy (5) Seismic forces (6) Dynamic water pressure during an earthquake (7) Tractive forces of vessels Of the external forces acting on the qwaywall, those forces which have a low probability of simultaneous occurrence with seismic forces may be neglected or can be reduced. [Technical Notes] (1) As shown in Fig. T- 4.2.1, the wall body of a gravity type quaywall can be taken as the portion between the face line of quaywall and the vertical plane passing through the rear toe of the quaywall. Normally a backfill is provided at the rear of the quaywall. In many types of gravity quaywalls, some part of this backfill acts as selfweight of the quaywall, and the portion of the backfill can be considered as a part of the quaywall body. But it is difficult to apply this concept to all cases unconditionally, because the extent of backfill considered as a part of the quaywall body varies depending on the shape of the quaywall body and the mode of failure. However, it is common practice to define the extent of backfill considered as a part of the quaywall body as shown by hatching in Fig. T- 4.2.1 to simplity the design calculation, because modest changes in the location of the quaywall body boundary plane do not affect the stability of the quaywall body significantly. -408-


(a) Wall of L-shaped concrete block type

(b) Wall of concrete block type

(c) Wall of cellular concrete block type

(d) Wall of caisson type

Fig. T- 4.2.1 Determination of Quaywall Body

(2) In the case of structure like a concrete block type quaywall for which stability must be examined for each horizontal layer, the virtual wall body should be considered as follows (usually, keys are formed between blocks for better interlocking, but in this examination, their effect may be ignored.):

Wall body portion

(a) Examination of sliding As shown in Fig. T- 4.2.2, the portion in front of the vertical plane passing through the rear toe at the examining level should be regarded as the wall body.

Wall body portion Horizontal plane to be assessed for stability

Horizontal plane to be assessed for stability Fig. T- 4.2.2 Determination of Wall Body Portion for Stability of Sliding at Horizontal Joints

Fig. T- 4.2.3 Determination of Wall Body Portion for Stability of Overturning

(b) Examination of overturning For the examination of overturning, when there are two blocks at the examining level, the portion in front of the vertical plane passing through the rear toe of the upper block on the sea side block can be regarded as a part of the wall body. For example, as in case of Fig.T- 4.2.3, it is assumed that the weight of block B and the weight of backfill A above the block B do not contribute to the resisting force against overturning. (c) Examination of bearing capacity If the same virtual wall body as that used against examination of overturning is employed for calculation of the safety factor for bearing capacity, it becomes quite small. However, when the weight of the wall body is locally concentrated on the ground, settlement occur in that portion. Therefore, the load is actually expected to be distributed over a wide area without being overly concentrated. Results of examination on the stability of existing structures show that the portion in front of the vertical plane passing through the rear toe of the quaywall can be considered as a virtual wall body. However, it is preferable to use one solid block of the bottom to ensure enough bearing capacity. (3) The residual water level should be set at the elevation with the height equivalent to one third of the tidal range above the mean monthly-lowest water level (LWL). The design water level is determined in accordance with Part Ⅱ, 6.1 Design Water Level. (4) In general, the range of residual water level difference becomes large as the tidal range increases and as the permeability of the wall body material becomes small. As the water behind the wall body permeates through voids in wall joints, foundation mound, and backfill, the residual water level difference can be reduced by increasing the permeability of these materials. However, care is necessary, as this approach may cause leakage of the backfilling material. The standard residual water difference (1/3 of the tidal range) is that for cases where a certain level of permeability can be established after a long period. In those cases where permeability is low from the beginning -409-


or reduction in permeability is expected in the long term, it is desirable to assume a large residual water level difference in consideration of those conditions. When the wave trough acts on the front face of the wall body, it is considered that a residual water level difference increases. However, in ordinary quaywall design, the increase in the residual water level difference due to the waves needs not be considered 3). (5) For the wall friction angle, it is desirable to use an angle of 15º for all types of structures. (6) The surcharge is determined in accordance with Part Ⅱ, Chapter 15 Loads. (7) As shown in Fig. T- 4.2.4, the buoyancy is calculated on the assumption that the part of the wall body below the residual water level is submerged in the water.

The part below this line is subjected to buoyancy


Front water level Wall body

The part subjected to buoyancy

Fig. T- 4.2.4 Assumption for Calculating Buoyancy

(8) For seismic forces, the values and descriptions given in Part Ⅱ, Chapter 12 Earthquakes and Seismic Force can be used as reference. The loads that can be excluded in stability calculations during an earthquake because of no simultaneous occurrence are as follows: (a) Tractive forces of vessels (b) Reaction forces of cargo handling equipment in operation or during storm conditions (c) The sidewalk live load and snow load (however, snow load should be considered based on the past data in heavy snow region) (9) For the dynamic water pressure during an earthquake, refer to Part Ⅱ, Chapter 14 Earth Pressure and Water Pressure. (10) In many cases, the fender reaction force is not taken into consideration in quaywall design, because the deadweight of the coping and the earth pressure of the material behind the quaywall work as the resistance forces. In the design of coping, however, the fender reaction force is taken into consideration.

4.3 Stability Calculations 4.3.1 Items to Be Considered in Stability Calculations In the stability calculations of a gravity type quaywall, the following items should be examined in general: (1) Sliding of the wall (2) Bearing capacity of the foundation (3) Overturning of the wall (4) Circular slip (5) Settlement 4.3.2 Examination against Sliding of Wall (Notification Article 66, Clause 1-1) The safety factor against sliding of a gravity type quaywall shall be calculated using equation (4.3.1). In this case the value of safety factor shall be set appropriately in accordance with the characteristics of the structure. fW F s ≦ ------(4.3.1) P where W： resultant vertical force acting on the wall (kN/m) P： resultant horizontal force acting on the wall (kN/m) f： coefficient of friction between the bottom of the wall body and the foundation F s： safety factor

-410-


[Technical Notes] (1) The safety factor should be 1.2 or more in ordinary conditions and 1.0 or more in extraordinary conditions. (2) The resultant vertical force should be the weight of the virtual wall body with subtraction of buoyancy and without a surcharge on the virtual wall body. The vertical component of earth pressure acting on the virtual plane should also be added. (3) The resultant horizontal force should include the following: ① Horizontal component of the earth pressure acting on the rear plane of the virtual wall body with a surcharge applied. ② Residual water pressure ③ In the stability calculations during an earthquake, the seismic force acting on the mass of the wall body with no buoyancy subtracted should be included in the resultant horizontal force in addition to the above ① and ② . The earth pressure should be the horizontal component of earth pressure during the earthquake. Furthermore, if there is a cargo handling equipment on the wall, the horizontal force transmitted through its legs should be considered. (4) With regard to the coefficient of friction, see Part Ⅱ, Chapter 16 Coefficient of Friction.

4.3.3 Examination Concerning Bearing Capacity of Foundation (Notification Article 66, Clause 1-3) Examination concerning the bearing capacity of the foundation shall be made appropriately in accordance with PartⅤ Ⅴ, 2.5 Bearing Capacity for Eccentric and Inclined Loads. [Technical Notes] (1) In the case of examination as a shallow foundation, the force acting on the bottom of the quaywall is the resultant force of vertical and horizontal loads. The examination should be made according to Part Ⅴ, 2.5 Bearing Capacity for Eccentric and Inclined Loads. (2) In general, the assessment of reaction force onto the bottom of quaywall is made for cases where no surcharge is applied on the quaywall. When a surcharge is applied on the quaywall, the distance of eccentricity decreases, but the bottom reaction may increase as the vertical component of the load increases. Thus there may be cases where assessment needs to be made for cases in which a surcharge is applied. (3) The thickness of a foundation mound is determined by examining the bearing capacity of the foundation, the flatness of the mound surface for installing the wall body, and the degree of alleviation of partial stress concentration in the ground. The minimum thickness should be as follows: ① 0.5 m or more, and 3 times the diameter of the rubbles or more in case of a quaywall with a water depth of less than 4.5 m. ② 1.0 m or more, and 3 times the diameter of the rubbles or more in case of a quaywall with a water depth of 4.5 m or more.

4.3.4 Examination Concerning Overturning of Wall (Notification Article 66, Clause 1-2) The safety factor against the overturning of gravity type quaywall shall satisfy equation (4.3.2). Wt F s ≦ ------Ph

(4.3.2)

where F s： safety factor t： distance between the line of application of the resultant vertical forces acting on the quaywall and the front toe of the quaywall (m) h： height of the application line of the resultant horizontal forces acting on the quaywall, above the bottom of the quaywall (m) The quantities W and P in the above equation have the same values as those in equation (4.3.1). [Technical Notes] The safety factor should be 1.2 or more in ordinary conditions and 1.1 or more in extraordinary conditions.

4.3.5 Examination on Soft Foundation (Notification Article 66, Clause 2) For gravity type quaywalls, the required stability of the structure shall be secured against the circular slip and the settlement due to the consolidation of the ground as described in Part Ⅴ, 6.2 Stability Analysis, in consideration of the characteristics of the ground and the structure. -411-


4.4 Stability Calculations of Cellular Concrete Blocks In the case of a cellular concrete block type quaywall (with no bottom slab), an examination shall be made against the overturning of the quaywall taking into consideration the reductionn of resistance force due to the separation of the filling from the cellular blocks. [Commentary] Cellular concrete blocks that do not have a bottom slab are different from other types of gravity quaywalls in that they maintain their integrity with the help of the filling. For this type of quaywall, therefore, it is necessary to carry out stability calculations with respect to overturning that take into consideration the risk that blocks may detach themselves from the filling, in addition to the stability assessments required for other types of gravity quaywalls. [Technical Notes] (1) The safety factor of overturning should satisfy equation (4.4.1). Wt + M f F t ≦ ------------------(4.4.1) Ph where W： resultant vertical force on the quaywall with the weight of filling excluded (kN/m) P： resultant horizontal force acting on the quaywall (kN/m) t： distance between the line of application of the resultant vertical force acting on the quaywall with the weight of filling excluded and the front toe of the quaywall (m) h： height of the line of application of the resultant horizontal forces acting on the quaywall above the bottom of the quaywall (m) M f： resisting moment provided by the friction between cellular blocks and filling (kN•m/m) F t： safety factor The safety factor should be 1.2 or more in ordinary conditions and 1.1 or more in special conditions. (2) If F t < 1, the overturning moment due to external forces becomes larger than the resisting moment caused by the resultant vertical force on the cellular concrete blocks and the frictional forces between blocks and filling. The cellular concrete blocks thus will overturn leaving the filling. In such a case, countermeasures against overturning should be taken, such as the increase of the weight of cellular concrete blocks or the provision of partition walls. (3) The resisting moment Mf provided by the friction between cellular blocks and filling is obtained as follows: In Fig. T- 4.4.1, the moment around the point A is l1F1 + l2F2. The thrusts P1 and P2 are the resultant forces of earth pressures due to filling acting on the front and rear inner walls of cellular blocks respectively, and F = fP, where f is the coefficient of friction between the filling material and the inner walls of cellular blocks. With regard to the earth pressure of filling acting on the inner walls, see Part Ⅳ, 3.3.2 Earth Pressure of Filling and Residual Water Pressure. Similarly, the friction force acting on the partition walls of cellular blocks should also be considered.

q: earth pressure due to the vertical load transmitted to filling H: H = b p: earth pressure due to filling p = KHg' K: coefficient of earth pressure g': unit weight of filling material P1, P2: resultant force of earth pressure

Fig. T- 4.4.1 Determination of Frictional Resistance

(4) The coefficient of friction used for the examination of the sliding of cellular concrete blocks with no bottom slab should be 0.6 for the part of reinforced concrete and 0.8 for the part of filling stones, but for the sake of convenience, 0.7 can be used on average.

-412-


4.5 Effects of Backfill When a backfill of good quality is used for a gravity type quaywall, the quaywall can be designed by considering the effect of the backfill. [Commentary] Backfill is expected to bring the following effects: (1) Reduction in the earth pressure due to an increase in the angle of internal friction (2) Drop in the residual water level due to an increase in permeability (3) Prevention of the outflow of the landfill material [Technical Notes] (1) Effect of Earth Pressure Reduction by Backfill The effect of earth pressure reduction by backfill can be calculated by using an analytical method that takes into consideration the composition and strength of the soil layers behind the quaywall 4). In ordinary gravity type quaywalls, rubble or cobble stones are used as backfilling material. In this case, the effect of earth pressure reduction can be evaluated using the following simplified method 5): (a) When the cross section of backfill is triangular: When backfill is made in a triangular shape from the point of intersection of the vertical line passing through the rear toe of quaywall and the ground surface with an angle of slope less than the angle of repose a of the backfilling material as shown in Fig. T- 4.5.1 (a), it may be assumed that the whole part of the rear of wall is filled with the backfilling material. When the reclaiming material is cohesive soil slurry, application of filling-up or filter sheets to the surface of the backfill should be considered in order that the cohesive soil slurry cannot pass through the void of the backfill and to reach the quaywall. (b) When the cross section of backfill is rectangular: In the case of triangular shape backfill with a slope steeper than the angle of repose of the backfilling material or any other irregular shape of backfill, the effect should be considered as in the case of rectangular shape backfill which has an area equivalent to the backfill in question. The effect of the rectangular backfill shown in Fig. T- 4.5.1 (b) should be considered as follows: When the width b of the rectangle shape backfill is larger than the height of the wall, consideration can be made as in the case of triangular backfill of Fig. T- 4.5.1 (a). When the width b is equal to 1/2 of the height, it should be assumed that the earth pressure equivalent to the mean of that due to the backfill and that due to the reclaimed soil acts on the quaywall. When the width b is 1/5 or less of the height of the wall, the earth pressure reduction effect due to the backfill should not be considered.

(a) Triangular shape backfill

(b) Rectangular shape backfill

Fig. T- 4.5.1 Shape of Backfill

(2) It is recommended to provide filling-up or filter sheets on the back face of the backfill in order to protect landfill material from piping out through the backfill caused by the change of the residual water level and to stop settlement of the base course for the apron. (3) No method has been developed yet to define the extent of increase in the earth pressure due to the infiltration of landfill material, because it varies depending on the grain size of landfill material and the size of backfilling material. Therefore it is common practice to take preventive measures such as the provision of filter sheets or filling-up on the slope of the backfill when infiltration is expected to produce a significant adverse effect. (4) When there is a risk that outflow of backfilling material may occur due to site conditions such as settlement, seal plates should be installed for the joints of the quaywall. -413-


4.6 Detailed Design The detailed design of a gravity type quaywall shall be made for the following matters: (1) Strength of structural members (2) Protection against the lealcage of reclaimed soil (3) Shape and dimensions of keys for concrete blocks (4) Structure of coping (5) Ancillary works [References] 1) Katsumi KISHIYA, Yasuhiro KUNISHIGE, Satoshi HIRANO, Masato YAMASHITA: “Characteristics and design methods for wedged bottom caisson quaywalls”, Abstracts of the 53rd Annual Meeting of the Japan Society of Civil Engineers, 1998 (in Japanese). 2) Toshikazu MORITA, Gen KIMURA, Katsuyuki SHIROMIZU, Hidenori TANAKA: “A study on the behavior of wedged bottom caisson quaywalls during earthquake”, Abstracts of the 53rd Annual Meeting of the Japan Society of Civil Engineers, 1998 (in Japanese). 3) Teruaki FURUDOI, Takeo KATAYAMA: “Field observation of residual water level”, Tech. Note of PHRI, No.115,1971 (in Japanese). 4) Takashi TSUCHIDA, Yoshiaki KIKUCHI, Tetsuo FUKUHARA, Takeo WAKO, Kazuhiro YAMAMURA: “Slice method for earth pressure analysis and its application to light-weight fill”, Tech Note of PHRI, No.924, 1999 (in Japanese). 5) Shôichi KITAJIMA, Hiroshi SAKAMOTO, Shôhei KISHI, Takuji NAKANO, Syusaku KAKIZAKI: “On some problems being conserved with preparation for the design standards on port and harbour structures” Tech. Note of PHRI, No.30,1967, pp.32-43 (in Japanese).

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Chapter 5 Sheet Pile Quaywalls 5.1 General (Notification to Article 67, Clause 1) In principle, the examination for structural stability for a steel sheet pile quaywall equipped with anchorage shall be made by checking the stabilities of both the sheet pile wall and anchorage work. [Technical Notes] (1) Fig. T- 5.1.1 shows an example of sheet pile quaywall design.

Thickness of concrete pavement = 0.20

Thickness of base course = 0.30

HWL

Tie rod

LWL

Backfilling stones

Steel pipe pile

Wooden pile (pine) Rubble stones

Sandy soil

Steel sheet pile

(Units: m)

Fig. T- 5.1.1 Example of Sheet Pile Quaywall Design (Vertical Pile Anchorage Type)

(2) The sheet pile wall design methods currently in use are as follows: (a) Design methods based on classic earth pressure theories ①Free earth support method ②Fixed earth support method (b) Design method that evaluates the embedded section as an elastic support P.W.Rowe’s method

5.2 External Forces Acting on Sheet Pile Wall 5.2.1 External Forces to Be Considered The following forces shall be considered as the external forces acting on a sheet pile wall: (1) Earth pressure (a) Active earth pressure behind the sheet pile wall (b) Subgrade reaction in front of the embedded portion of sheet piles (2) Residual water pressure (3) Earthquake forces that act on (1) (4) Dynamic water pressure during an earthquake (5) Tractive forces of vessels (6) Berthing forces of vessels

-415-


[Commentary] The forces listed above are the ones whose effects are to be rather large. In cases where the sheet pile coping is large, it may also become necessary to take into consideration earthquake forces that act on the coping. Any other external force that is considered to produce a significant effect on the sheet pile wall should be taken into consideration as appropriate. [Technical Notes] (1) The active earth pressure is normally used as the earth pressure that acts on the sheet pile wall from the backside. For the frontside reaction that acts on the embedded part of the sheet pile, it is necessary to use an appropriate value such as the passive earth pressure or a subgrade reaction that corresponds to the deflection of the wall. (2) When the “free earth support method” and the “equivalent beam method” are used in a stability calculation for a sheet pile wall, it should basically be assumed that the earth pressure and residual water pressure act as shown in Fig. T- 5.2.1, and the pressure values should be calculated in accordance with Part Ⅱ, Chapter 14 Earth Pressure and Water Pressure. The wall friction angle used for calculation of the earth pressure acting on the sheet pile wall may usually be taken at 15º for the active earth pressure and -15º for the passive earth pressure, respectively (when the ground is sandy soil layer).


Residual water level Residual water pressure

sure

Passive earth pressure

rth pres

Residual water pressure

ressure

Active earth p


LWL

ea Active

LWL

On cohesive soil

On sandy soil

Fig. T 5.2.1 Earth Pressure and Residual Water Pressure to Be Considered for Design of Sheet Pile Wall

(3) Since the earth pressure changes in response to displacement of the sheet pile wall, the actual earth pressure that acts on the sheet pile wall varies depending on the following: (a) The construction method (i.e., whether backfill is executed or the ground in front of the sheet piles is dredged to the required depth after the sheet piles have been driven in) (b) The lateral displacement of the sheet pile at the tie rod setting point (c) The length of the embedded part of the sheet pile (d) The relationship between the rigidity of the sheet pile and the characteristics of the sea bottom ground. (4) When Rowe’s method (elastic beam analysis method) is used in a sheet pile stability calculation, it is assumed that the earth pressure and residual water pressure act as shown in Fig. T- 5.2.2 and a reaction earth pressure that corresponds to the coefficient of subgrade reaction and the earth pressure at rest act on the front surface of the sheet pile. (5) When there is some cargo handling equipment (e.g., crane) on the quaywall, it is necessary to take into consideration the earth pressuredue due to the deadweight and the surchage of the equipment.

LWL

GL Tie rod


Active earth pressure + residual water pressure

Subgrade reaction

(6) In the determination of the reaction force of earth pressure that acts on the front surface of the embedded part of the sheet pile, it is necessary to assume Earth pressure at rest that dredging of the sea bottom will be excecuted to a certain depth below the planned depth, in considFig. T- 5.2.2 Earth Pressure and Residual Water Pressure eration of the accuracy of dredging work. (7) In the case of an earth retaining wall of an open-type wharf, the sea bottom in front of the sheet pile wall -416-

to Be Considered for Design of Sheet Pile Walls which Uses P.W. Rowe's Method


has a composite shape of horizontal and sloped surfaces. In such a case, the passive earth pressure may be calculated using Coulomb’s method in which the design passive earth pressure is trially calculated with several failure planes of different angles. The smallest value among them is adopted as the passive earth pressure 1). However, it is necessary to consider the empirical evidence by experiments that the behavior of the ground in front of the sheet pile wall can be well predicted under the assumption of the ground being an elastic body. (8) The residual water level to be used in the determination of the residual water pressure needs to be estimated appropriately in consideration of the structure of the sheet pile wall and the soil conditions. The residual water level varies depending on the characteristics of the subsoil and the conditions of sheet pile joints etc., but in many cases the elevation with the height equvalent to two thirds of the tidal range above the mean monthly-lowest water level (LWL) is used for sheet pile walls. In the case of a steel sheet pile wall driven into cohesive soil ground, however, care should be exercised in the determination of the residual water level, because it is sometimes nearly the same as the high water level. When sheet piles made of other materials are to be used, it is desirable to determine the residual water level based on the result of investigations of similar structures. (9) For the dynamic water pressure during an earthquake, refer to Part Ⅱ, Chapter 14 Earth Pressure and Water Pressure. (10) The berthing forces of vessels may be considered only for the design of the coping in general. The tractive force of vessel is not considered when the foundation for bollards is to be constructed separately. However, when bollards are to be installed on the coping of the sheet pile wall, it is necessary to take into consideration the tractive force of vessel in the design of the coping, tie rod, and wale.

5.3 Design of Sheet Pile Wall 5.3.1 Setting Level of Tie Rod The setting level of tie rod shall be determined by considering the convenience in execution and construction cost. [Commentary] The design cross sections of the sheet piles and the tie rod vary widely depending on the tie rod setting position. In the determination of the tie rod setting level, therefore, it is necessary to consider the convenience in execution and construction cost.

5.3.2 Embedded Length of Sheet Piles (Notification Article 67, Clause 2-1) In principle, the embedded length of sheet piles shall be determined in such a way that the bottom end of the sheet piles is firmly fixed in the ground and an appropriate level of safety factor is attained. [Commentary] The mechanical behavior of sheet pile wall changes according to the embedded length. The sheet pile wall behaves as a free earth support wall when the embedded length is short and as a fixed earth support wall when the embedded length is long. To realize a safe sheet pile wall design, it is essential to make sure that the bottom end of the sheet pile is fixed securely, which means that the wall must be designed as a fixed earth support wall. Traditionally the “free earth support method” based on a classic earth pressure theory has been used to determine the embedded length. Takahashi et al. 2) have demonstrated that embedded lengths obtained by this method frequently correspond to the fixed earth support conditions when an appropriate value of safety factor is used. The equivalent beam method used to determine the sheet pile cross section assumes that the sheet piles are in the fixed earth support condition. [Technical Notes] The mechanical behavior of sheet pile wall equipped with an anchorage work is strongly affected by the rigidity of the sheet piles, characteristics of the ground, and embedded length. In particular, the mechanical behavior varies according to the embedded length. The design method described in this chapter is based on the premise that the bottom end of the sheet pile wall is fixed. The embedded length at which the bottom end of sheet pile wall is assumed fixed varies according to the rigidity of the sheet piles and the characteristics of the ground. The embedded length determination method employed in the “free earth support method” based on earth pressure theory has some shortcomings: e.g., ① the rigidity of the sheet piles are not considered, and ② when the bottom end of the sheet pile wall is fixed, the assumed passive earth pressure distribution does not coincide with Coulomb’s triangle distribution. However, the embedded length thus determined secures the state of fixed earth support under certain conditions. The sheet pile embedded length determination method used in the free earth support method are described in the following: -417-


(1) When calculating the embedded length of a sheet pile using the free earth support method, equation (5.3.1) should be satisfied for the moments caused by the earth pressure and residual water pressure (see Fig. T- 5.2.1 of 5.2.1 External Forces to Be Considered) with respect to the tie rod setting point. (5.3.1) MP = FMA where MP： moment at the tie rod setting point by the passive earth pressure (kN•m/m) MA： moment at the tie rod setting point by the active earth pressure and residual water pressure (kN•m/m) F： safety factor The safety factor may be set at 1.5 or more in ordinary conditions and 1.2 or more in extraordinary conditions in case of a sheet pile wall driven in sandy soil, and 1.2 or more both in ordinary and extraordinary conditions in hard cohesive soil. (2) In the case of cohesive soil ground, the stability of embedded part is not established unless equation (5.3.2) is satisfied. 4c > w + Sg h + g w hw where c： cohesion of soil at sea bottom (kN/m2) w： surcharge (kN/m2) g ： unit weight of soil (kN/m3) (submerged unit weight for the part below the residual water level) h： thickness of soil layer above the sea bottom (m) g w：unit weight of sea water (kN/m3) hw： difference between the residual water level and the water level in front of the wall (m)

(5.3.2)

When the sea bottom ground is soft and equation (5.3.2) is not satisfied, it is necessary to improve the sea bottom ground with an appropriate method or take appropriate measures such as the use of relieving platforms.

5.3.3 Bending Moment of Sheet Piles and Reaction at Tie Rod Setting Point The maximum bending moment of sheet piles and reaction at the tie rod setting point shall be calculated with an appropriate method that takes into consideration the rigidity and embedded length of the sheet piles and the hardness of the ground. [Commentary] It has long been known that the mechanical behavior of sheet pile wall is affected by the rigidity and embedded length of the sheet pile and the hardness of the ground. In the past, design water depths for quaywalls were relatively shallow, the sheet piles used were those having relatively low rigidity such as U-section steel sheet piles, and the bending moment inflection point was located at the sea bottom or a shallower point when used in sandy soil ground with a medium or high degree of compaction. Therefore, a simplified method called the “equivalent beam method” had been used. However, as a result of the increase in the number of quaywalls with deeper design water depths in recent years, steel sheet piles with relatively high rigidity, such as steel pipe sheet piles, are now being used frequently. Therefore care should be exercised in using the traditional “equivalent beam method”, because the bending moment inflection point may be located below the sea bottom and the section forces may be underestimated. [Technical Notes]

(2) Equivalent Beam Method The equivalent beam method calculates the maximum bending moment and reaction force at the tie rod setting point of the sheet piles by assuming a simple beam supported at the tie rod setting point and the sea bottom with the earth pressure and residual water pressure acting as the load above the sea bottom (see Fig. T- 5.3.1).

Reaction at the tie rod point (Ap)

Residual water level Tie rod Active earth pressure

(1) The maximum bending moment and reaction force at the tie rod setting point of sheet piles may be determined using the equivalent beam method described below or Rowe’s method. However, care should be exercised when using the equivalent beam method, because the section forces may be under estimated when the rigidity of the sheet piles is high.


Fig. T- 5.3.1 Equivalent Beam for Obtaining Bending Moment

-418-


5.3.4 Cross Section of Sheet Piles The cross section of a sheet pile shall be determined to ensure that the stress calculated in accordance with 5.3.3 Bending Moment of Sheet Piles and Reaction of the Tie Rod Setting Point does not exceed the allowable stress for the material. 5.3.5 Consideration of the Effect of Section Rigidity of Sheet Piles (Notification to Article 67, Clause 2-2) In principle, the cross section of the sheet pile shall be an appropriate one that has been selected in consideration of the cross-sectional rigidity of the sheet pile. [Commentary] As described in 5.3.2 Embedded Length of Sheet Piles [Technical Notes], the behavior of sheet pile wall equipped with the anchorage work is strongly affected by the rigidity and embedded length of the sheet piles and the characteristics of the ground. In particular, the rigidity of the sheet piles strongly affects the determination of the embedded length. Therefore it is essential to take into consideration the effect of the cross-sectional rigidity of the sheet pile in the selection of the sheet piles in design. [Technical Notes] (1) The approach that uses the free earth support method and the equivalent beam method in combination has been popularly used because of its proven track record and ease of use, but it is not a design method that takes into consideration the cross-sectional rigidity of the sheet pile. (2) The analysis method described below, which is a modified Rowe’s method, examines the embedded part of sheet piles as a beam set on an elastic media. (a) Elastic beam analysis method of sheet piles This elastic beam analysis method is applied to the sheet pile wall with the theoretical equation for beams on elastic media, introducing an elastic coefficient of subgrade reaction for the ground into which the sheet pile wall is driven. The basic equation for the embedded part is as follows: EI d 4 y/dx4 = p (x) = pA0 － (lh/D) x y (5.3.3) where E： Young’s modulus of sheet pile (MN/m2) I： geometrical moment of inertia of sheet pile wall per unit width (m4/m) pA0： load intensity at the sea bottom generated by the active earth pressure and residual water pressure (MN/m2/m). lh： coefficient of subgrade reaction to the sheet pile wall (MN/m3) D： penetration depth of sheet pile (m)

GL RWL

WL

A

x y

AD

x

x

As there is no general solution to a differential equation of this form, a special technique is required to solve equation (5.3.3). Bloom and A0 Rowe proposed a method to obtain the coefficient of each term in a numerical analysis by assuming a Fig. T- 5.3.2 Earth Pressure Distribution for power series as the solution. the Analysis of Sheet Pile Wall Based on Rowe’s method 3), Takahashi and Ishiguro have published details of a method to derive a solution of the deflection curve equation of sheet pile wall and a computer-based numerical calculation method 4). Takahashi et al. have amended this method to better reflect the behavioral characteristics of actual sheet pile walls as follows (see Fig. T- 5.3.2): (5.3.4) EI d 4y/dx4 = p (x) = pA0 + KADg x － K0g x － [lh / (DF rf)] x y where KAD： coefficient of active earth pressure in the embedded part of the sheet pile wall g： unit weight of soil (MN/m3) K0： coefficient of earth pressure at rest. DF： converged embedded length of sheet pile wall (m) rf： ratio of the exerting depth of the primary positive reaction earth pressure acting on the front surface of the embedded part of the sheet pile to DF -419-


(b) Characteristic embedded length of sheet pile wall Use of the analysis method described above reveals that the behavioral characteristics of sheet pile wall change according to its embedded length. That is, the stability of the sheet pile wall is not established until the embedded length of the sheet pile reaches a certain value. The embedded length at which this critical stability state is established is called the critical embedded length DC. When the embedded length is further extended and made longer than the critical embedded length, the bending moment of the sheet pile wall reaches the maximum peak bending moment MP of free earth support state. The embedded length that attains this condition is called the transitional embedded length DP . If the embedded length is further extended, the bending moment reaches the converged maximum bending moment MF of fixed earth support state. The minimum embedded length that attains this condition is called the converged embedded length DF. (c) Flexibility number As a measure to indicate the rigidity of a sheet pile wall as a structure, the following “flexibility number” proposed by Rowe is used: r = H 4/EI where r： flexibility number (m3/MN) H： total length of sheet pile (m)

(5.3.5)

Rowe uses the sum of total height H of the sheet pile wall (from the sea bottom to the top of the sheet pile wall) and the embedded length D of fixed earth support state (that is, H + D) as the total length of sheet pile. Takahashi et al. suggest a new index called the similarity number that is derived by using the flexibility number and ground characteristics. By using the height HT from the sea bottom to the tie rod setting point for the length H in equation (5.3.5), the similarity number is defined as follows: (5.3.6) w = rlh where w： similarity number lh： coefficient of subgrade reaction of sheet pile (MN/m3) By expressing the mechanical characteristics of a sheet pile wall with a similarity number, the effect of the rigidity of the sheet piles can be estimated quantitatively. (d) Coefficient of subgrade reaction of sheet piles There are a very few reference data that gives measured or suggested values of coefficient of subgrade reaction of the sheet pile lh. Therefore it is desirable to obtain these values by means of model experiments and/or field measurements. The proposed values that have traditionally been used include the values proposed by Terzaghi and the ones proposed by Takahashi et al., which have been obtained by modifying Terzaghi’s values. The research conducted by Takahashi et al. shows that the effect of errors in the coefficient of subgrade reaction is not fatal for practical use 2). Thus the values proposed by Takahashi et al. may normally be used as the coefficient of subgrade reaction of sheet pile wall. ① Values proposed by Terzaghi 5) The values proposed by Terzaghi are as listed in Table T- 5.3.1. Table T- 5.3.1 Coefficient of Subgrade Reaction for Sheet Pile Wall in Sandy Ground (lh) (Units: MN/m3) Relative density of sand

Loose

Medium

Dense

24

38

58

Coefficient of subgrade reaction (lh)

② Values proposed by Takahashi et al. 2) Takahashi et al. confirmed that the result of Tschebotarioff’s model experiment of sheet pile wall 6) does not contradict with the values proposed by Terzaghi. They related the coefficient of subgrade reaction listed in Table T-5.3.1 with the N-value, using the Terzaghi’s relationship between the coefficient of subgrade reaction and the relative density, and another Terzaghi’s relationship 7) between the N-value and the relative density. Then they adopted the smaller value of coefficients of subgrade reaction for the safety and connected the resultant values with a smooth line as shown in Fig. T-5.3.3. They also related the coefficients of subgrade reaction with the angle of interval friction as shown in Fig. T-5.3.4, using one equation (5.3.7) of Dauham’s equation that gives a smaller angle of internal friction for a given N-value. f =

12N + 15 (º)

(5.3.7)

However, it should be noted that Fig. T- 5.3.4 is a slightly expedient one, as Danham’s equation include cases that give the larger angle of internal friction depending on the grain shape of sandy soil. Figures T- 5.3.3 and T- 5.3.4 also exhibit the values proposed by Terzaghi in addition to the values proposed by Takahashi et al. -420-

Coefficient of subgrade reaction lh (MN/m3)


Values proposed by Terzaghi

Values proposed by Takahashi et al

N-value

Coefficient of subgrade reaction lh (MN/m3)

Fig. T- 5.3.3 Relationship between Coefficient of Subgrade Reaction (lh) and N-Value

Values proposed by Terzaghi

Values proposed by Takahashi et al. Angle of internal friction (o)

Fig. T- 5.3.4 Relationship between Coefficient of Subgrade Reaction (lh) and Angle of Internal Friction (f)

(e) Design embedded length In the determination of the embedded length of sheet piles using Rowe’s method, a value that satisfies equation (5.3.8) can be used. (5.3.8) d s = D F ¤ H T ≧ 5.0916 ´ w 0.2 0.2591 where d s： ratio of the embedded length of sheet pile to the height of the tie rod setting point above the sea bottom DF： embedded length of sheet pile (m) HT： height of the tie rod setting point above the sea bottom (m) w： similarity number (= r lh) r： flexibility number (= HT4 / EI) (m3/MN) E： Young’s modulus of sheet pile (MN/m2) I： geometrical moment of inertia of sheet pile wall per unit width (m4/m) lh： coefficient of subgrade reaction to sheet pile wall (MN/m3) The embedded length calculated with this equation is the converged embedded length. According to the study conducted by Takahashi et al., an increase of just a 2%-plus in the maximum bending moment occurs when an embedded length corresponding to 70% of the converged embedded length is employed. Therefore the use of the converged embedded length as the design embedded length provides a safety factor in itself, and there is no need to consider a safety factor here. (f) Maximum bending moment of sheet pile and reaction force of the tie rod setting point In the determination of the maximum bending moment of sheet piles and the tie member fixing point reaction force while considering the effects of the coefficient of subgrade reaction and the bending moment of sheet piles, correction factors shown in Figs. T- 5.3.5 and T- 5.3.6 are to be used. The maximum bending moment -421-


-422-

Fig. T- 5.3.5 Relationship between Correction Factor (μ）and Similarity Namber (ω）

ordinary conditions seismic conditions

and the tie member fixing point reaction force calculated by means of the equivalent beam method are multiplied by these correction factors to obtain the corrected values. The seismic coefficient of 0.2 has been used in preparing Figs. T- 5.3.5 and T- 5.3.6. Values obtained from these figures may be used for seismic design unless a very detailed design is required. The symbol m in Fig. T- 5.3.5 denotes the ratio of MF to MT, where MF denotes the maximum bending moment that is generated when the embedded length is the converged embedded length DF in the deflection curve analysis, while MT denotes the maximum bending moment obtained by the equivalent beam method that uses the tie rod setting point and the sea bottom as the supporting points.


-423-

Fig. T- 5.3.6 Relationship between Correction Factor (τ) and Similarity Number (ω)

ordinary conditions seismic conditions

4

The symbol t in Fig. T- 5.3.6 denotes the ratio of TF to TT, where TF denotes the tie member fixing point reaction force that is generated when the embedded length is the converged embedded length DF in the deflection curve analysis, while TT denotes the tie member fixing point reaction force calculated with the equivalent beam method.


5.4 Design of Tie Rods 5.4.1 Tension of Tie Rod The tension acting on a tie member shall be calculated in accordance with 5.3.3 Bending Moment of Sheet Pile and Reaction at Tie Rod Setting Point. [Commentary] The tie rod tension that is calculated in accordance with 5.3.3 Bending Moment of Sheet Pile and Reaction at Tie Rod Setting Point is a tension per meter of quaywall length, but tie rods are usually installed at fixed intervals, and in some cases, tie rods may be attached at a certain angle with the line perpendicular to the sheet pile wall to avoid an existing structure located behind the wall. Therefore, it is necessary to calculate the tie rod tension considering these site conditions. [Technical Notes] (1) The tension that acts on a tie rod is calculated by equation (5.4.1). (5.4.1) T = A p l sec q where T： tension of tie rod (kN) Ap： reaction at the tie rod setting point according to 5.3.3 Bending Moment of Sheet Pile and Reaction at the Tie Rod Setting Point (kN/m) l： tie rod setting interval (m) θ： inclination angle of tie rod to the line perpendicular to the sheet pile wall (°) (2) In some cases, bollards are installed on the coping of a sheet pile wall and the tractive forces of vessels acting on the bollards are transmitted to tie rods. Usually, the coping is assumed a beam with the tie rods as elastic supports and the tie rod tension may be calculated using equation (5.4.2), assuming that the tractive force is evenly shared by four (4) tie rods near a bollard. When the tractive force of vessel is sustained by the tie rods, the allowable stress of the tie rods may be evaluated as that of extraordinary condition. T = ( A p l + P ¤ 4 ) sec q where P： horizontal component of the tractive force that acting on a bollard (kN)

(5.4.2)

The tractive force of vessel should be in accordance with PartⅡ, 2.2.4 Tractive Force Acting on Mooring Post and Bollard.

5.4.2 Cross Section of Tie Rod (Notification Article 67, Clause 3) The cross section of a tie rod shall be determined to ensure that the tensile stress calculated from the reaction at the tie rod setting point is within an acceptable range. In principle, the reaction at the tie rod setting point shall be calculated in consideration of the cross-sectional rigidity of the sheet pile. [Commentary] In principle, the cross section of tie rod is determined by means of the allowable stress method. [Technical Notes] (1) The allowable tensile stress of tie rod should be equal to or less than 40% of the guaranteed yield stress in ordinary conditions and 60% or less in extraordinary conditions. (2) The allowable tensile stress of tie wire should be set with the safety factor of 3.8 or larger against breaking strength in ordinary conditions and 2.5 or larger in extraordinary conditions. (3) The descriptions in (1) and (2) above have taken into consideration the following matters: (a) The case study on the collapse of sheet pile walls during the Niigata Earthquake in 1964 8) and the result of a model shaking experiment on a sheet pile wall conducted by the Port and Harbour Research Institute have shown that a tension larger than that calculated in accordance with 5.4.1 Tension of Tie Rod may act on the tie rod during an earthquake 9). (b) The tie rod may be bent and a stress larger than the calculated value may act on the tie rod if settlement occurs in the landfill material at the rear of the sheet pile wall in ordinary conditions.

-424-


For tie rod materials, the ratio of the allowable tensile stress to the yield stress is reduced to 2/3 of that for ordinary steel structures and the values of safety factor against breaking strength are selected as listed in Table T- 5.4.1. Table T- 5.4.1 Properties of Tie Rod Materials

Type

SS400

SS490

Breaking strength (N/mm2)

Allowable stress

(N/mm2)

Yield stress (N/mm2)

235 or more (for diameter of 40 mm or less) 402 or more 215 or more (for diameter of more than 40 mm) 275 or more (for diameter of 40 mm or less) 490 or more 255 or more (for diameter of more than 40 mm)

Yield stress Elongation divided by (%) breaking strength

Ordinary conditions

Seismic conditions

94

141

24 or more

86

129

110

Safety factor (breaking strength divided by allowable stress) Ordinary conditions

Seismic conditions

0.58

4.27

2.85

24 or more

0.53

4.67

3.12

165

21 or more

0.56

4.45

2.97

102

153

21 or more

0.52

4.80

3.20

High tension steel 490

490 or more

325 or more

130

195

24 or more

0.66

3.77

2.51


590 or more

390 or more

156

234

22 or more

0.66

3.78

2.52


690 or more

440 or more

176

264

20 or more

0.64

3.92

2.61


740 or more

540 or more

216

324

18 or more

0.73

3.43

2.28

5.5 Design of Wale The maximum bending moment of a wale shall be calculated with an appropriate calculation method. [Commentary] Many equations have been proposed to calculate the maximum bending moment of wale. The maximum bending moment should be determined in such a way that a safe and economical cross section is obtained in consideration of site conditions. [Technical Notes] (1) The maximum bending moment of wale may be calculated using equation (5.5.1). M = Tl / 10 where M： maximum bending moment of wale (kN•m) T： tension of a tie rod calculated in accordance with 5.4.1 Tension of Tie Rod (kN) l： tie rod setting interval (m)

(5.5.1)

Equation (5.5.1) is obtained by analyzing a three-span continuous beam supported at the tie rod setting points and subjected to the reaction at the tie rod point (Ap) as an uniformly distributed load. (2) When bollards are to be installed on the coping, it is necessary to design the wale near one of the bollards using a tie rod tension that takes into consideration the tractive force of vessel in accordance with [Technical Notes] (2) of 5.4.1 Tension of Tie Rod. In this case, an allowable stress 1.5 times as large as that for ordinary conditions may be used. However, when the wale is embedded into the coping, the effect of the tractive force may be ignored.

5.6 Examination for Circular Slip (Notification Article 67, Clause 5) Sheet pile quaywalls shall be designed with an appropriate safety factor against circular slip having a slip surface that extends below the bottom end of the sheet pile in accordance with Part Ⅴ, 6.2 Stability Analysis. -425-


5.7 Design of Anchorage Work 5.7.1 Selection of Structural Type of Anchorage Work The structual type of anchorage work shall be selected by considering the construction cost, construction period, construction method, and ground height before execution of work. 5.7.2 Location of Anchorage Work (Notification Article 67, Clause 4-1) In principle, the location of the anchorage work shall be set at an appropriate distance from the sheet pile wall to ensure the structural stability of the quaywall, depending on the characteristics of the anchorage work. [Commentary] The location of the anchorage work should be determined appropriately in consideration of the structural type of the anchorage work, because (a) the stability of the anchorage work itself is affected by its position and (b) the location at which the stability is achieved varies depending on the structural type. There should be a sufficient distance between the anchorage work and sheet pile wall so that the stability of both the sheet pile wall and anchorage work is secured. [Technical Notes] (1) The location of concrete wall anchorage should be so determined to ensure that the active failure plane starting from the intersection of sea bottom and sheet pile and the passive failure plane of the concrete wall anchorage drawn from the bottom of the anchorage do not intersect below the ground surface as shown in Fig. T- 5.7.1.

Tie rod Residual lane water re p level ilu e fa tc iv A

Concrete wall anchorage e failu re plan e

WL

Passiv

Fig. T- 5.7.1 Location of Concrete Wall Anchorage Works

Sheet pile

Sheet pile

WL

Tie rod Residual ane Pas water e pl failusive level ilur planere a f e tiv Ac

Vertical pile anchorage

(2) The location of vertical pile anchorage should be so determined to ensure that the passive failure plane from the point of lm1/3 below the tie rod setting point of the anchorage and the active failure plane from the intersection of sea bottom and sheet piles do not intersect at the level below the horizontal surface containing the tie rod setting point at the anchorage as shown in Fig. T- 5.7.2. The value of lm1 is the depth of the first zero point of bending moment for a free-head pile below the tie rod setting point, while the horizontal surface containing the setting point of tie rod at the anchorage is assumed as the ground surface.

Fig. T- 5.7.2 Location of Vertical Pile Anchorage

(3) The location of sheet pile anchorage may be determined in accordance with the method described in (2) when the sheet piles can be regarded as long piles. When the sheet piles cannot be regarded as long piles, the location of anchorage may be determined by ignoring the part deeper than the level lm1/2 below the tie rod setting point at the sheet pile anchorage and then applying the method described in (1).

Residual level

pl

fa

anchorage

v ti

c

e ur

il

e

A

e

an

water

Coupled-pile

(5) For ordinary sheet pile quaywalls whose tie rods run horizontally, an angle of -15º may be used as the wall friction angle in the determination of the passive failure plane that is drawn from the vertical pile anchorage or sheet pile anchorage.

Tie rod

WL

Sheet pile

(4) For the method to obtain the first zero point of the bending moment of the vertical pile anchorage and sheet pile anchorage and the method to determine whether a sheet pile anchorage can be considered as a long pile, refer to Port and Harbour Research Institute’s method described in Part Ⅴ, 4.3.4 Estimation of Pile Behavior Using Analytical Methods.

Fig. T- 5.7.3 Position of Coupled-Pile Anchorage

(6) The location of coupled-pile anchorage should be behind the active failure plane of the sheet pile wall drawn from the sea bottom when it is assumed that the tension of the tie rod is resisted only by the axial bearing capacity of the piles (see Fig. T- 5.7.3). When the tension of the tie rod is evaluated to be resisted by both the axial and lateral bearing capacity in consideration of the bending resistance of the piles, it is necessary to locate the anchorage in accordance with the method described in (2). -426-


5.7.3 Design of Anchorage Work (Notification Article 67, Clause 4-2) In principle, anchorage work shall be designed in consideration of the structural characteristics of the sheet pile quaywall and the anchorage work using an appropriate method, so that the stability of the anchorage work is secured. [Technical Notes] (1) Design of Concrete Wall Anchorage (a) Stability of concrete wall anchorage The height and placing depth of concrete wall anchorage may be determined to satisfy equation (5.7.1), on the assumption that the tie rod tension and the active earth pressure behind the concrete wall anchorage are resisted by the passive earth pressure in front of the concrete wall anchorage (see Fig. T- 5.7.4). F = EP / (AP + EA) (5.7.1) where F： safety factor AP: reaction at the tie rod setting point according to 5.3.3 Bending Moment of Sheet Pile and Reaction at Tie Rod Setting Point (N/m) EA： resultant active earth pressure acting on concrete wall anchorage (N/m) EP： resultant passive earth pressure acting on concrete wall anchorage (N/m) In the calculation of the earth pressure acting on concrete wall anchorage, the surcharge should be assumed to act as shown in Fig. T- 5.7.4 (i.e., the surcharge is only considered in the calculation of active earth pressure, and not of passive earth pressure). In this calculation, the safety factors should be 2.5 or larger in ordinary conditions and 2.0 or larger in extraordinary conditions.


Fig. T- 5.7.4 External Forces Acting on Concrete Wall Anchorage

(b) When the active failure plane of the sheet pile and the passive failure plane of the concrete wall anchorage drawn in accordance with 5.7.2 Location of Anchorage Work intersect below the ground surface level, it is desirable to consider the fact that the passive earth pressure ΔEP acting on the vertical surface above the intersection point does not function as a resistance force (see Fig. T- 5.7.5); it should be subtracted from the value of EP of equation (5.7.1). When the intersection point is located above the residual water level, the active earth pressure to be subtracted may be calculated using equation (5.7.2).

lu

re

iv

e

ct

A

i fa

e

n

la

p

Pa

ssi

ve

fai

lur

ep la

ne

Passive earth pressure to be subtracted (

E p)

Fig. T- 5.7.5 Earth Pressure to Be Subtracted from the Passive Earth Pressure that Acts on Anchorage Wall When the Active Failure Plane of Sheet Pile Wall and the Passive Failure Plane of Concrete Wall Anchorage Intersect -427-


ΔEP = KPg hf2 / 2 where KP： coefficient of passive earth pressure g： unit weight of soil (N/m3) hf ： depth from the ground surface to the intersection of the failure planes (m)

(5.7.2)

(c) Cross section of concrete wall anchorage Concrete wall anchorage should be designed safely against the bending moment caused by the earth pressure and the tie rod tension. In general, the maximum bending moment may be calculated by assuming that ① the earth pressure is approximated to an equally distributed load and ② the concrete wall anchorage is a continuous slab in the horizontal direction and a cantilever slab fixed at the tie rod setting point in the vertical direction, and then using equation (5.7.3). MH = Tl / 12 MV = Th / 8l where MH： horizontal maximum bending moment (N•m) MV： vertical maximum bending moment per meter in length (N•m/m) T： tie rod tension according to 5.4.1 Tension of Tie Rod (N) l： tie rod interval (m) h： height of concrete wall anchorage (m)

(5.7.3)

The layout of the reinforcing bars for MH may be determined on the assumption that the effective width of the concrete wall anchorage is 2b with the tie rod setting point as the center, where b is the thickness of the concrete wall anchorage at the tie rod setting point. (2) Design of Vertical Pile Anchorage Vertical pile anchorage may be designed as vertical piles with an external force due to tie rod tension. (3) Design of Coupled-Pile Anchorage Coupled-pile anchorage may be designed as a coupled piles with an external force due to tie rod tension. (4) Design of Sheet Pile Anchorage (a) When the sheet pile anchorage below the tie rod setting point is long enough to be regarded as a long pile, the cross section of the sheet pile anchorage may be determined in accordance with (2) in the above. (b) Sheet piles anchorage that cannot be regarded as a long pile may be designed in accordance with (1) in the above on the assumption that the earth pressure acts on a range down to lml/2 point below the tie rod setting point, as shown in Fig. T- 5.7.6. The length lm1 is the vertical distance from the tie rod setting point to the first zero point of the bending moment of sheet piles assuming that the sheet pile anchorage is a long pile.

Fig. T- 5.7.6 Virtual Earth Pressure for Short Sheet Pile Anchorage

5.8 Detailed Design 5.8.1 Coping Coping shall be designed to be safe against the earth pressure behind the wall, the tractive forces of vessels, and the berthing forces of vessels. [Technical Notes] Coping may be designed as a cantilever beam that is fixed at the top of the sheet pile and subjected to the earth pressure as a load. However, it is necessary to consider ① the tractive forces of vessels and the active earth pressure behind the wall for the parts on which bollards are installed and ② the berthing forces of vessels and the passive earth pressure behind the wall for the parts on which fenders are installed. The only factor that should be considered with regard to extraordinary conditions is the active earth pressure during an earthquake.

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5.8.2 Fitting of Tie Rods and Wale to Sheet Piles The joint between the sheet piles and tie rods and the joint between the sheet piles and the wale shall be designed in such a way that the horizontal force acting on the sheet pile wall is equally and safely transmitted to each tie rod through the wale. 5.8.3 Tie Rod Tie rods shall be designed to ensure that the tie rod tension obtained in 5.4.1 Tension of Tie Rod is transmitted safely to the anchorage work. When the bending stress caused by the settlement of backfill soil is anticipated, the design shall be made by taking it into consideration. 5.8.4 Fitting of Tie Rods to Anchorage Work The fitting between anchorage work and tie rods shall be designed to ensure that the tie rod tension obtained in 5.4.1 Tension of Tie Rod is transmitted safely and equally to the anchorage work. [Technical Notes] A continuous beam along the face line of quaywall is usually attached on top of the anchorage piles, and the tie rods are attached to the beam. This beam may be designed as a continuous beam subjected to the tie rod tension and the reaction force of the piles.

5.9 Special Notes for Design of Sheet Pile Wall on Soft Ground [Technical Notes] (1) The design of a sheet pile wall on soft ground should be made by comparing and examining the methods according to 5.3 Design of Sheet Pile Wall, 5.4 Design of Tie Rods, and 5.7 Design of Anchorage Work, as well as other design methods. Unexpected large deformation may occur in sheet piles constructed on soft ground due to lateral flows that are caused by the settlement of the ground behind the sheet pile wall. Such a sheet pile wall should be designed to be safe against this type of deformation by using a suitable method of lateral flow prediction (several methods have been proposed by researchers) 10). (2) Care should be exercised in using the design methods for sheet pile quaywall described in the previous sections, because many of these design methods assume that a steel sheet pile wall is driven mainly into sandy soil ground or hard clayey soil ground. For soft ground, it is desirable to perform soil improvement work. When it is not possible to perform soil improvement work because of site conditions, it is necessary to consider to use other design methods in addition to the design methods described in the previous sections so that the quaywall will be designed appropriately. (3) The term “soft ground” here refers to the cases in which the sea bottom ground is weak alluvial clay or similar type of soil. (4) Other than the methods described in 5.3.2 Embedded Length of Sheet Piles, the “deflection curve method” may be used in the design of the embedded length of sheet piles driven into soft ground. This is a kind of fixed earth support method based on a classic earth pressure theory on sheet piles with long embedded length. In this method, elastic equations are solved under the external force conditions shown in Fig. T- 5.9.1 with the conditions that the displacement and deflection angle is zero at the bottom end of the embedded part and the displacement at the tie rod setting point is zero. (5) In the determination of the bending moment of sheet piles and the tension of tie rod, it is desirable to use the method described in 5.3.3 Bending Moment of Sheet Pile and Reaction at Tie Rod Setting Point as well as the elastic line method described in (4) and to examine the results in a comprehensive manner.

p

Deflection curve

d

Fig. T- 5.9.1 Earth Pressure and Deflection Curve

(6) It must be remembered that the determination of the embedded length and the cross section of the sheet piles in a sheet pile wall design should be done using a single method and not using a combination of muliple methods. [References] 1) Genpei SAWADA: “Calculation method of passive earth pressure for sheet pile wall embedded to inclined sea bottom” Tech. Note of PHRI, No.9, 1964 (in Japanese).

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2) Kunio TAKAHASHI, Yoshiaki KIKUCHI, Yuji ASAKI: “Analysis of flexural behavior of anchored sheet pile walls”, Tech. Note of PHRI, No. 756, 1993 (in Japanese). 3) P. W. Rowe: “A theoretical and experimental analysis of sheet-pile walls”, Proc. I. C. E., Vol. 4, Pt. 1, 1955. 4) Kunio TAKAHASHI and Ken ISHIGURO: “Vertical beam analysis of pile and sheet pile under lateral load”, Sankaido Press, 1992, pp. 177-183 (in Japanese). 5) K.Terzaghi: “Evaluation of Coefficients of Subgrade Reaction”, Geotechnique, Vol. 5, 1955, pp. 297-326. 6) Tschbotarioff: “Large Scale Earth Pressure Tests with Model Flexible Bulkheads”, Princeton Univ. 1949. 7) Terzaghi and Peck (translated by Hoshino et al.): “Soil Mechanics-Applications”, Maruzen, 1970, p. 192 (in Japanese). 8) Port and Harbour Bureau, 1st District Port Construction Bureau, and Port and Harbour Research Institute, Ministry of Transport: “Report on damage to port facilities by Niigata Earthquake Part 1”, 1964, p. 101 (in Japanese). 9) ditto, Part 2, 1965, p. 213 (in Japanese). 10) Yasuo KASUGAI, Kenichiro MINAMI and Hiroyuki TANAKA: “Estimation of deformation for port facilities by lateral flow of soft ground”, Tech. Note of PHRI, No.276, 1992 (in Japanese).

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Chapter 6 Sheet Pile Quaywalls with Relieving Platform 6.1 Scope of Application The provisions in this chapter shall be applied to the design of quaywall with relieving platform that comprises a relieving platform, a sheet pile wall in front of the relieving platform, and relieving platform piles. [Commentary] Sheet pile quaywall with a relieving platform normally comprise a platform, a sheet pile earth retaining wall in front of the relieving platform, and relieving platform piles. This chapter gives provisions on quaywalls with relieving platforms. The relieving platform is in many cases constructed as an L-shaped structure of cast-in-place reinforced concrete and is usually buried under landfill material, but sometimes a box shape platform is used to reduce the weight of the platform and the earthquake forces that act on it (see Figs. C- 6.1.1 and C- 6.1.2). WL

WL

Platform

Hollow Platform

Platform piles Sheet pile wall

Fig. C- 6.1.1 Structure of Quaywall with Relieving Platform (L-Shaped Platform)

Fig. C-T- 6.1.2 Structure of Quaywall with Relieving Platform (Box Shape Platform)

6.2 Principles of Design (1) The design of a quaywall with a relieving platform shall be made separately for the sheet piles, the relieving platform, and the relieving platform piles. (2) In the design of a quaywall with a relieving platform, the stability of the structure as a whole shall be examined against sliding, overturning, and circular slip. [Commentary] (1) The sheet piles should be designed in accordance with Chapter 5 Sheet Pile Quaywalls, and the relieving platform and its supporting piles should be designed in consideration the reaction force acting on the upper part of sheet piles, earth pressure, surcharge, deadweight, and earthquake forces as the external forces and in accordance with PartⅤ Ⅴ, Chapter 4 Bearing Capacity of Pile Foundations. (2) As for the stability analysis of the structure as a whole, the provisions given in Chapter 4 Gravity Type Quaywalls should be observed with respect to sliding and overturning, and the provisions given in PartⅤ Ⅴ, Chapter 6 Stability of Slopes should be observed with respect to circular slip. Examination of stability with respect to circular slip is required for soft ground, but generally not required for relatively hard ground such as sand and sandy soil.

6.3 Determination of Height and Width of Relieving Platform The installation level and shape of the relieving platform shall be appropriately determined in consideration of the external force conditions, cost, easiness of construction work, etc. [Technical Notes] The design of the installation level and shape of the relieving platform should be made by taking careful consideration to the following points: (1) When the height of relieving platform is large and its bottom is located at a low elevation, the earth pressure acting on the sheet piles can be lessened, therefore making the cross section and embedded length of the sheet piles smaller. But this generally requires a heavier weight of relieving platform, so accordingly larger seismic forces against it. Therefore this results in requiring a larger quantity and longer length of relieving platform piles. A search should be made for the solution with the minimum cost. -431-


(2) It is recommended to lower the bottom of the relieving platform down to the residual water level to prevent the corrosion of relieving platform piles, because the ground below the relieving platform bottom may settle and produce void below it, which enhances corrosion with the presence of air. (3) The width of relieving platform is normally determined in such a way that the relieving platform and the active failure plane of the sheet piles drawn from the sea bottom intersect so that the earth pressure acting on the sheet pile wall is reduced. In doing this, it is necessary to check that a sufficient width is secured that accomodates the required number of relieving platform piles to be deployed in an appropriate manner.

6.4 Earth Pressure and Residual Water Pressure Acting on Sheet Piles The earth pressure and residual water pressure acting on sheet piles shall be calculated appropriately in consideration of the structural characteristics of the quaywall with a relieving platform. [Technical Notes] (1) When the active failure plane of sheet piles from the intersection between the sea bottom and sheet piles intersects the relieving platform, the active earth pressure acting on the sheet pile wall should be calculated on the assumption that the bottom of the relieving platform is the virtual ground surface as shown in Fig. T- 6.4.1. (2) The residual water pressure acting on the sheet pile wall should be considered the same as that of the case without a relieving platform. But the force to be adopted should be the residual water pressure acting on the range below the bottom level of relieving platform (see Fig. T- 6.4.1). (3) The passive earth pressure in front of the sheet pile should be determined in accordance with Chapter 5 Sheet Pile Quaywalls. Design water level

Active earth pressure




(LWL)

Fig. T- 6.4.1 Earth Pressure and Residual Water Pressure Acting on Sheet Pile Wall

6.5 Design of Sheet Pile Wall The embedded length of sheet piles shall be such that the bottom end of sheet piles is secured sufficiently against the external forces that act on the sheet pile wall. The cross section of the sheet piles shall be such that the stress calculated by a method suitable for the structural characteristics of the quaywall with a relieving platform does not exceed the allowable strength of the material. 6.5.1 Embedded Length of Sheet Piles [Technical Notes] The embedded length of sheet piles of a sheet pile wall with a relieving platform should be basically designed in accordance with 5.3 Design of Sheet Pile Wall and 5.9 Special Notes for Design of Sheet Pile Wall on Soft Ground by assuming the joint between the sheet pile wall and relieving platform as a hinger support and replacing the tie rod setting point with the bottom elevation of relieving platform.

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6.5.2 Cross Section of Sheet Piles [Technical Notes] (1) The cross section of sheet piles should be basically designed in accordance with 5.3 Design of Sheet Pile Wall and 5.9 Special Notes for Design of Sheet Pile Wall on Soft Ground by replacing the tie rod setting point with the bottom elevation of the relieving platform. (2) In addition to the bending moment due to earth pressure, the bending moment and vertical force transmitted from the relieving platform act on the sheet piles of a sheet pile wall. Normally the bending moment transmitted from the relieving platform is not taken into consideration, because it usually acts in a direction opposite to that of the maximum bending moment that acts on the sheet piles (and thus reduces the maximum bending moment). (3) Normally the vertical force transmitted from the relieving platform to the sheet pile wall is not taken into consideration because the front row of relieving platform piles are usually driven in as close to the sheet pile wall as possible and this dramatically reduces the vertical force acting on the sheet piles.

6.6 Design of Relieving Platform and Relieving Platform Piles 6.6.1 External Forces Acting on Relieving Platform The following shall be considered as the external forces acting on the relieving platform: (1) Horizontal force transmitted from the sheet pile wall (2) Earth pressure and residual water pressure acting on the rear of the relieving platform (3) Deadweight of the relieving platform, weight of soil on the relieving platform, and surcharge (4) Seismic force acting on the above (3) (5) Dynamic water pressure during an earthquake [Technical Notes] (1) The horizontal force transmitted from the sheet pile wall should be calculated with the method same as that for the reaction at the tie rod setting point obtained in accordance with 5.4 Design of Tie Rods by regarding the bottom elevation of relieving platform as a tie rod setting point. (2) It should be assumed that (a) the earth pressure acting on the rear of the relieving platform acts on the vertical plane that contains the rear face of the relieving platform, and (b) the part above the bottom of relieving platform is subjected to the active earth pressure and the part below the bottom is subjected to both the active and passive earth pressure up to the point where the sum of the two becomes 0, as shown in Fig. T- 6.6.1.

Earth pressure

Acting on sheet pile reaction force Design water level (LWL)

Pp

Pa

Pp



Pa Pp

Fig. T- 6.6.1 External Forces Acting on Relieving Platform

(3) The tractive force of vessels and fender reaction force also act on the relieving platform, but usually these external forces need not be considered. (4) The external forces transmitted from the sheet pile wall to the relieving platform include the horizontal force and bending moment. However, the transmission of the bending moment is ignored for the sake of safety, because the attachment of the sheet piles to the relieving platform may not be secure enough. (5) The earth pressure and residual water pressure acting on the back of the relieving platform should be calculated in accordance with Part Ⅱ, Chapter 14 Earth Pressure and Water Pressure. In the calculation of earth pressure, surcharge should be taken into consideration. In the part below the bottom of relieving platform, there exist the passive earth pressure acting on the front of the vertical plane at the rear of platform and the action earth pressure acting on the rear of the vertical plane. The difference between them acts as the active earth pressure down to the depth where the two pressures are balanced.

6.6.2 Design of Relieving Platform Relieving platform shall be designed appropriately in consideration of its structural shape.

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[Technical Notes] (1) A relieving platform should be designed as a continuous beam supported by the head of the relieving platform piles. When the relieving platform is an L-shaped structure, the upright section should be designed as a cantilever beam supported at the slab section.

w +wd

(2) A relieving platform should be designed as a continuous beam for both the direction of quaywall alignment and the direction perpendicular to the alignment (see Fig. T- 6.6.2). Loads should not be distributed in the two directions.

Bending moment

Coupled piles

w + wd

Bending moment due to surcharge

M0 : maximum bending moment of the upright section Ap : reaction force of sheet pile wall at the upper supporting point w : surcharge wd : load due to soil and deadweight of relieving platform

Bending moment transmitted from the upright section

Tension Fig. T- 6.6.2 Continuous Beam Assumed in Design of Relieving Platform

6.6.3 Design of Piles Relieving platform piles shall be designed appropriately in consideration of the ground conditions, assignment of external forces and loads to piles, easiness of construction work, cost, etc. [Technical Notes] (1) Piles should be designed in accordance with Part Ⅴ, Chapter 4 Bearing Capacity of Pile Foundations as a pile foundation that is subjected to the external forces described in 6.6.1 External Forces Acting on Relieving Platform. (2) In principle, relieving platform piles should consist of a combination of coupled piles and vertical piles. The horizontal external force may be assigned to the coupled piles only, and the vertical external force may be assigned to the vertical piles only. It may be assumed that each of the coupled piles shares the horizontal force equally. (3) In the design of relieving platform piles, assessment should be made for the most dangerous state of each pile by varying the surcharge, direction of earthquake forces, and sea level within the design condition ranges.

6.7 Examination of Stability as Gravity Type Wall The examination of the stability of a quaywall with relieving platform as a whole shall be made by assuming that the quaywall with relieving platform is a kind of gravity type wall.

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[Technical Notes] A quaywall with relieving platform may be considered as a rectangular gravity wall defined by a vertical plane containing the rear face of the relieving platform and a horizontal plane containing the bottom ends of the front side batter piles of the coupled piles, as shown in Fig. T- 6.7.1.

6.8 Examination of Stability against Circular Slip When a sheet pile quaywall with a relieving platform is to be constructed on soft ground, an examination should be made on the stability against the circular slip passing the ground Fig. T- 6.7.1 Virtual Wall as Gravity Type Wall below the lower end of the sheet pile wall.

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Chapter 7 Steel Sheet Pile Cellular-Bulkhead Quaywalls 7.1 Principle of Design The provisions in this chapter shall be applied to the design of quaywalls with steel sheet pile cellularbulkhead structure. [Commentary] The design method described in this chapter is based on the results of cellular-bulkhead model tests 1), 2) conducted on a sandy soil ground with an embedded length ratio of 0 to 1.5 and a “equivalent wall width / wall height” ratio of 1 to 2.5. For the cases where (1) the embedded length ratio is very small (less than 1/8), (2) the equivalent wall width is very small relative to the wall height, or (3) the quaywall is to be constructed on a clayey soil ground or ground improved by the sand compaction piles, etc., further examinations should be made as required in addition to the examination using the design method described in this section, because these cases involve factors that cannot be fully clarified with the method described here. [Technical Notes] (1) The design of a steel sheet pile cellular-bulkhead quaywall is preferably made in accordance with the sequence depicted in Fig. T-7.1.1.

Determination of design conditions Assumption of wall dimensions Re-assumption of wall body dimensions

Calculation of external forces Examination against shear deformation Estimation of coefficient of subgrade reaction Calculation of subgrade reaction and wall body displacement Examination of bearing capacity of the subsoil Examination against sliding of wall Examination of displacement of wall top Examination of stability against circular slip Layout of cells and arcs

Consideration of soil improvement

Calculation of sheet pile tension Detailed design Fig. T- 7.1.1 Sequence of Design of Steel Sheet Pile Cellular-Bulkhead Quaywall

(2) It is recommended that the hollow space in the cells be filled to a sufficient density with sand or gravel of good quality. It is not desirable to use a clayey soil as the filling material. When clayey soil is to remain in the cells, it is necessary to make separate examination because the deformation of the cells may become significantly large. (3) When a foundation for a crane, quay shed, or warehouse is to be built within a cell, it is desirable to use bearing piles to transmit the load to the bearing stratum. (4) An example of steel sheet pile cellular-bulkhead quaywall is shown in Fig. T- 7.1.2. -436-


Face line

HWL LWL

Cell D

PC pile L

Fig. T- 7.1.2 Example of Steel Sheet Pile Cellular-Bulkhead Quaywall

7.2 External Forces Acting on Steel Sheet Pile Cellular-Bulkhead Quaywall As the external forces acting on the steel sheet pile cellular-bulkhead, the following shall be considered: (1) External Forces Acting on the Cellular-Bulkhead Itself. (a) Earth pressure ①Earth pressure behind the wall ②Earth pressure in front of the wall ③Earth pressure of the filling (b) (c) (d) (e)

Residual water pressure Seismic force and dynamic water pressure that act on the wall Weight of the wall Surcharge

(2) External Forces Acting on the Coping (a) Horizontal force ①Horizontal component of earth pressure ②Residual water pressure ③Seismic forces acting on the coping, the soil above the coping, and surcharge (b) Vertical forces ①Weight of the coping and the soil above the coping and surcharge ②Vertical component of earth pressure [Technical Notes] (1) Calculation of external forces are based on Part Ⅱ, Chapter 12 Earthquake and Seismic Force, Part Ⅱ, Chapter 14 Earth Pressure and Water Pressure, and PartⅡ Ⅱ, Chapter 15 Loads. (2) The rear of the wall is subjected to active earth pressure above the sea bottom. It should be assumed that the part of the wall below the sea bottom is subjected to (a) active earth pressure in the examination of shear deformation and (b) earth pressure generated by surcharge pressure such as the weight of backfilling and surcharge load in the examination of stability of the structure as a whole (see Fig. T- 7.2.1). According to the model tests, it can be said that the part of wall below the sea bottom is subjected to the loads corresponding to the earth pressure at rest, because the deformation of the embedded section of the cell is small. According to the results of vibration experiments, the earth pressure acting on this part works as a resisting force against overturning of the wall. In the design of quaywall, therefore, the earth pressure that is generated by backfilling and surcharge under ordinary conditions should be used as this earth pressure, as shown in equation (7.2.1) (see Fig. T- 7.2.2). -437-


p ac = 0.5 ( Sg h + w ) i i where pac： earth pressure acting on the rear of wall below the sea bottom (kN/m2) gi： unit weight of each layer of backfilling (kN/m3) hi： thickness of each layer of backfilling (m) w： surcharge (kN/m2)

(7.2.1)

(3) In principle, the residual water level of the backfilling should be taken at the elevation with the height equivalent to two thirds of the tidal range above the mean monthly-lowest water level (LWL). However, when using a backfilling with low permeability, the residual water level may become higher than this and thus it is desirable to determine the residual water level based on results of investigations of similar structures. The residual water level in the filling material used to fill the cells may be set to the same level as that of the backfilling for the wall body. (4) For the sea bottom and above, the seismic coefficient to be used in the calculation of the seismic force that acts on the filling material is the design seismic coefficient. For the part below the sea bottom, this value is reduced linearly in such a way that it becomes zero at 10 m below the sea bottom. In principle, the seismic coefficient is not considered for the part deeper than that level (see Fig. T- 7.2.3).

Surcharge RWL

LWL

backfilling

Wall body


Sea bottom

Active earth pressure Fig. T- 7.2.1 Earth Pressure Acting on the Back of Wall Body (for Examination of Shear Deformation) Seismic coefficient Surcharge

RWL

LWL

Wall body

backfilling

Sea bottom Active earth pressure

Sea bottom Earth pressure that acts on the part below the sea bottom (according to equation (7.2.1))

Fig. T- 7.2.2 Earth Pressure Acting on the Back of Wall Body (for Examination of the Stability as Gravity Wall)

Fig. T- 7.2.3 Design Seismic Coefficient for Filling

7.3 Examination of Wall Width against Shear Deformation 7.3.1 General The examination of the strength of the wall body against shear deformation shall basically be made for the loads acting on the wall under ordinary conditions. [Commentary] The cell shell and filling of the steel sheet pile cellular-bulkhead usually act as an integrated structure, because the filling is constrained in the cell shell. Therefore the deformation of the cell wall body may be ignored relative to its displacement and the overall behavior of the cell wall body may be considered the same as that of a rigid body. This has been verified by model tests in which cell wall bodies did not show significant deformation under the loads much larger than the external forces that are expected to act on the cell wall body both under ordinary conditions and during an earthquake. In general, therefore, it can be said that shear failure does not occur in the filling. However, when the diameter of the cell is very small or the strength of the filling material is extremely low, it may not be possible to consider the cell wall body as a rigid body. Therefore it is necessary to make examination of the strength of the filling against shear deformation due to the loads under ordinary conditions in order to reduce deformation of the cell wall body to a negligible level. -438-


7.3.2 Equivalent Width of Wall For designing the width of wall, equivalent width of wall may be used. The equivalent width of wall shall be the width of a rectangular virtual wall whose cross-sectional area is the same as the combination of cell and arc section. [Commentary] The equivalent width of wall is the width of a rectangular virtual wall body that is used in place of the wall body combined with cells and arc sections to simplify design calculations (see Fig. C- 7.3.1). The virtual wall is defined in such a way that the area of the horizontal cross section of the virtual wall body becomes the same as that of the combined cells and arc sections.

B= S/L B: equivalent wall width (m) L: effective length of one set of cell (m) S: area of set of cell (m2)

(a) Circular cells

(c) Clover Leaf Type Cells (b) Diaphragm Type Cells

Fig. C- 7.3.1 Plan View of Cellular-Bulkhead Structure and Equivalent Width B

[Technical Notes] The equivalent width of wall should be such that the equations (7.3.1) and (7.3.2) for the shear deformation of wall body are satisfied. (7.3.1) Md F ≦ Mr M d ¢F ≦ M r ¢ where M d： deformation moment with respect to wall bottom (kN•m/m) M d ¢： deformation moment with respect to sea bottom (kN•m/m) M r： resisting moment with respect to wall bottom (kN•m/m) M r ¢： resisting moment of filling material of cell with respect to sea bottom (kN•m/m) F： safety factor

(7.3.2)

In principle, the safety factor against the shear deformation should be 1.2 or greater.

7.3.3 Calculation of Deformation Moment The deformation moment shall be an overturning moment at the sea bottom or the wall bottom due to external forces such as active and passive earth pressures above the sea bottom or the wall bottom and residual water pressure. [Commentary] In the calculation of deformation moment, earth pressure is considered only in terms of the horizontal component. The vertical component is not taken into consideration. The vertical force of the surcharge is not taken into consideration in the calculation of deformation moment. However, the surcharge is taken into consideration in the calculation of active earth pressure (see Fig. C- 7.3.2).

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Surcharge

Md

LWL

R.W.L. backfilling soil Residual water Active earth

pressure

pressure

Sea bottom (surface)

Active earth


pressure

Fig. C- 7.3.2 Loads and Resisting Forces to Be Taken into Consideration in the Examination of Shear Deformation

7.3.4 Calculation of Resisting Moment The resisting moment of wall shall be calculated appropriately in consideration of the structural characteristics of the cellular-bulkhead wall and deformation of the wall. [Technical Notes] (1) The result of model tests 1) shows that the resisting moment with respect to the wall bottom may be increased by increasing the embedded length ratio D/H (see Fig. T- 7.3.1). This can be calculated using equation (7.3.3).

R

values of individual cases.

R

Deformation moment obtained by experiment

Shear resistance moment according to the modified formula of Kitajima

Note: Plotted values are mean

R Case No.

Group A Group B Group C Group D Group E

Embedded length ratio (

D /H )

Fig. T- 7.3.1 Relationship between Resisting Moment and Embedded Length Ratio

(7.3.3) Mr = (1 + a・ D/H) (Mro + Mrs) where Mr： resisting moment with respect to wall bottom (kN•m/m) Mro： resisting moment of the filling with respect to wall bottom (kN•m/m) Mrs： resisting moment due to the friction force of sheet pile joints, with respect to wall bottom (kN•m/m) D： embedded length (m) H： height from wall bottom to wall top (m) (see Fig. T- 7.3.2.) a： increment factor for the embedded length ratio (D/H) For the increment factor a, it is recommended to use 1.0, which is close to the lowest value found in the experiment results shown in Fig. T- 7.3.1, because the equation given above has been derived based on experiments and not fully clarified theoretically. -440-


644474448

(a) Equation for calculating the resisting moment of filling In the determination of the resisting moment of filling at the wall bottom, it is assumed that ① an active failure plane is generated from the bottom of the front of the wall and a passive failure plane is generated from the bottom of the rear of the wall and ② the active and passive earth pressures act on the respective failure planes, as shown in Fig. T- 7.3.2. The active and passive failure angles as well as the active and passive earth pressures are calculated using the following Rankine’s equations: p f : active failure plane z a = --- + --4 2 p f : passive failure plane z p = --- --4 2 (7.3.4) 1 sin f p a = K a · g h , K a = ------------------- : active earth pressure 1 + sin f + sin f p p = K p · gh , K p = 1------------------- : passive earth pressure 1 sin f where f： internal friction angle of filling (º) The moment caused by the earth pressure acting on the shear plane may be calculated by using equation (7.3.5) referring to Fig. T- 7.3.2. M ro =

d

ò0 ( pp pa ) ( d x )

·

2 --- tan q dx 3

(7.3.5)

B

LWL

H x

pa

pp

d D

Fig. T- 7.3.2 Assumed Shear Planes of Filling

When the geotechnical constants of the ground and those of the filling differ, equation (7.3.5) becomes complex as the failure angle and the earth pressure level vary from one soil layer to another. However, when there is no significant difference in the internal friction angle between the ground and filling, or when the embedded length ratio is large and the failure planes do not reach the filling, the following simplified equation may be used: 1 3 (7.3.6) M ro = --- g 0 · R 0 · H 0 6 2 2 R 0 = --- n 0 ( 3 n 0 cos f ) tan f sin f 3 where g 0： equivalent unit weight of filling (unit weight of the filling which assumes that the unit weight is uniform throughout the filling; normally g 0 = 10 kN/m3 is used.) n 0 = B/H0 B： equivalent wall width (m) H0： equivalent wall height measured from wall bottom. The equivalent wall height is employed to calculate the resisting moment due to the filling by using the equivalent unit weight of the filling. It is evaluated by means of equation (7.3.7). 1 H 0 = ---- Sg i · h i g0

(7.3.7)

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where g i： unit weight of the i-th layer of filling (kN/m3) hi： thickness of the i-th layer (from wall bottom to wall top) (m) (b) Equation for calculating resisting moment due to friction force of the joints of sheet piles The resisting moment due to friction force of joint is calculated as follows: 1 3 M rs = --- g 0 · R s · H s 6 3 R s = --- ns f tan f 2 where g 0： equivalent unit weight of filling (kN/m3) φ： internal friction angle of filling (º) n s = B/Hs B： equivalent wall width (m) f： coefficient of friction between sheet pile joints; usually 0.3 is used. Hs： equivalent wall height measured from the wall bottom.

(7.3.8) (7.3.9)

The equivalent wall height Hs is employed to calculate the resisting moment due to the friction force between the sheet pile joints when the equivalent unit weight of the filling is used. It is evaluated using equation (7.3.10) so that the resultant force of the distributed earth pressure, ignoring surcharge, between the wall bottom and wall top (see Fig. T-7.3.3 (a)) becomes equal to that of the equivalent earth pressure distribution (see Fig. T- 7.3.3 (b)). In this calculation, 1 ¤ 2 tan f is used as the coefficient of earth pressure of the filling. Note that this is different from the coefficient of earth pressure used in the calculation of hoop tension. SP i H s = 2 ---------------g 0 tan f where Pi： resultant earth pressure of the i-th layer of filling (kN/m)

(a) Earth pressure distribution

(7.3.10)

(b) Equivalent earth pressure distribution

Fig. T- 7.3.3 Equivalent Wall Height

(2) In the calculation of the resisting moment of the filling with respect to the sea bottom, equations (7.3.11) and (7.3.12) may be used. 1 M r ¢ = --- g 0 · R 0 ¢ · H 0 ¢ 3 6 R0¢ = n 0 ¢2 (3 － n 0 ¢cos f ¢) sin f ¢ where n 0 ¢=B/H0¢ M r ¢： resisting moment of sheet pile cell with respect to sea bottom (kN•m/m) f ¢： internal friction angle of the filling above sea bottom (º) H 0 ¢： equivalent wall height measured from sea bottom.

(7.3.11) (7.3.12)

The equivalent wall height H 0 ¢ is employed to calculate the resisting moment due to the filling by using the equivalent unit weight of the filling. It is evaluated by means of equation (7.3.13). 1 H 0 ¢ = ----Sg i ¢ · h i ¢ (7.3.13) g0 where g¢ i： unit weight of the filling of the i-th layer above sea bottom (kN/m3) hi¢： thickness of the i-th layer above sea bottom (m) -442-


(3) Increasing the strength of the filling heightens the rigidity of the cellular-bulkhead wall. Therefore, improvement work of filling is effective in increasing the stability of the cellular-bulkhead wall.

7.4 Examination of Stability of Wall Body as a Whole 7.4.1 General In the examination of stability of wall body as a whole, the wall shall be considered as a gravity wall supported by the ground. [Technical Notes] (1) In the examination of the stability of the wall as a whole, the subgrade reaction generated against the load and the displacement of the wall are calculated by considering the wall as a rigid body elastically supported by the ground. (2) Within the elastic range of the ground, the subgrade reaction force is calculated as the product of the modulus of subgrade reaction and the displacement. Here it is considered that the stability of the wall as a gravity wall is obtained when the subgrade reaction force and the displacement of the wall do not exceed the respective allowable limits.

7.4.2 Modulus of Subgrade Reaction In principle, the modulus of subgrade reaction used in the examination of the stability of wall as a gravity wall shall be determined based on the results of in-situ investigation. [Commentary] The modulus of subgrade reaction includes the modulus of horizontal subgrade reaction, the modulus of vertical subgrade reaction, and the horizontal shear modulus at the wall bottom. [Technical Notes] The modulus of subgrade reaction may be calculated as below, based on the results of in-situ investigation: (1) Modulus of Horizontal Subgrade Reaction Modulus of horizontal subgrade reaction may be calculated by referring to Yokoyama’s diagram 3) shown in Part Ⅴ, 4.3.4 Estimation of Pile Behavior Using Analytical Method [Technical Notes] (3) as below. KH = 2N where KH： modulus of horizontal subgrade reaction coefficient (N/cm3) N： N-value

(7.4.1)

When the ground consists of the strata of different characteristics, the modulus of horizontal subgrade reaction should be calculated for each stratum. (2) Modulus of Vertical Subgrade Reaction For the modulus of vertical subgrade reaction at the wall bottom, the same value as the modulus of horizontal subgrade reaction at the wall bottom should be used. When the ground consists of the strata of different characteristics, the modulus of vertical subgrade reaction should correspond to the stratum at the wall bottom. However, when there is an extremely soft stratum below the wall bottom, it is necessary to give careful consideration to its effects. (3) Horizontal Shear Modulus The horizontal shear modulus at the wall bottom may be calculated by means of equation (7.4.2) using the modulus of vertical subgrade reaction. KS = l KV (7.4.2) where KS： horizontal shear modulus (N/cm3) l： ratio of the horizontal shear modulus to the modulus of vertical subgrade reaction KV： modulus of vertical subgrade reaction (N/cm3) Past studies suggest the use of l values in the range of 1/2 to 1/5 4), 5). In the case of steel sheet pile cellularbulkhead however, it is considered that the value of l may be set as about 1/3.

7.4.3 Calculation of Subgrade Reaction and Wall Displacement The subgrade reaction and wall displacement shall be calculated on the assumption that the wall under the load is supported by the subgrade reaction force and the vertical frictional force along the wall surface. -443-


[Commentary] The subgrade reaction acting on the embedded part of steel sheet pile cellular-bulkhead and the wall displacement should be calculated on the assumption that the wall subject to the external forces is supported by the horizontal subgrade reaction, vertical subgrade reaction and horizontal shear reaction at the bottom of wall, and vertical frictional force along the front and rear of the wall. [Technical Notes] (1) Subgrade Reaction (a) Horizontal subgrade reaction Horizontal subgrade reaction may be calculated by means of equation (7.4.3) shown in (5), but this should not exceed the passive earth pressure intensity calculated in accordance with PartⅡ, Chapter 14 Earth Pressure and Water Pressure to prevent the yielding of the ground. The angle of wall friction used to calculate passive earth pressure should basically be taken at -15º. Figure T- 7.4.1 illustrates the distribution of subgrade reaction of a sample case in which the subgrade reaction reaches the passive earth pressure up to a certain depth. (b) Vertical subgrade reaction The vertical subgrade reaction at the wall bottom acts in a trapezoidal or triangular distribution. It should be assumed that no tensile stress is generated. Cell

Portion where subgrade reaction reaches the passive earth pressure intensity

M

Sea bottom Passive earth pressure intensity

V

Backfilling soil

H

Cell embedment portion

Portion where subgrade reaction force does not reach the passive earth pressure intensity

Horizontal subgrade reaction due to the displacement of the cell

Fig. T- 7.4.1 Example of Distribution of Horizontal Subgrade Reaction

(2) Vertical Frictional Force It should be assumed that vertical frictional force acts on the front and rear of the wall and is calculated as the product of the horizontal earth pressure or subgrade reaction force and tan d , where d denotes the angle of wall friction. (3) Distribution of External Forces Figure T- 7.4.2 shows standard distribution patterns of the external forces acting on steel sheet pile cellularbulkhead quaywall such as loads and subgrade reaction forces, etc.

tan reaction

subgrade

reaction

Horizontal

Horizontal subgrade

acting on the part

Sea bottom

Seismic forces

Dynamic water pressure

Earth pressure

Dead weight

LWL

below the ground Horizontal component of active earth pressure tan surface tan

Surcharge

RWL



Earth pressure that acts on the part below the ground surface

Shear reaction at the bottom surface

(Trapezoidal distribution)

Vertical subgrade reaction force

(Triangular distribution)

Fig. T- 7.4.2 Distribution Patterns of External Forces Acting on Steel Sheet Pile Cellular-Bulkhead Quaywall -444-


Center axis

Center axis

(4) Displacement Modes of Wall As shown in Fig. T- 7.4.3, it is assumed that the cell wall rotates around its center of rotation O, which is horizontally away from the center axis of the cell by the distance e and vertically away from the sea bottom by the depth h. When the center of rotation is located inside the cell, the horizontal subgrade reaction is generated in the rear of the wall for the part below the center of rotation.

O O

(a) When the center of rotation is located outside the wall body

(b) When the center of rotation is located inside the wall body

Fig. T- 7.4.3 Displacement Modes of Wall

(5) Equation for Calculating Subgrade Reaction and Wall Displacement Figure T- 7.4.4 shows a calculation model for a case in which (a) horizontal force, vertical force, and moment act at the intersection of the ground surface and the center axis of the cell wall and (b) the ground comprises n layers of soil. Equations for calculating the subgrade reaction and cell wall displacement of the model shown in Fig. T- 7.4.4 are as follows: Cell

Backfilling soil

Ground surface Layer 1 Layer 2

z

Layer 3 Layer

i

Horizontal subgrade reaction

di dn

Layer

n

Shear reaction

q q

2

(Trapezoidal distribution)

1

Vertical subgrade reaction

(Triangular distribution)

q

1

Fig. T- 7.4.4 Calculation Model for Subgrade Reaction and Wall Displacement

-445-


(a) When the vertical subgrade reaction acts in a trapezoidal distribution ① Horizontal subgrade reaction (kN/m) p 21 = K H2 ( h d 1 )q p 22 = K H2 ( h d 1 d 2 )q … æ i1 ö p i1 = K Hi ç h å d j÷ q ç ÷ j=1 ø è i æ ö p i2 = K Hi ç h å d j÷ q ç ÷ j=1 ø è

… n1 ö æ p n1 = K Hn ç h å d j÷ q ç ÷ j=1 ø è

p n2

n æ ö ç = K Hn h å d j÷ q ç ÷ j=1 ø è

64444444444744444444448

p 12 = K H1 ( h d 1 )q

(7.4.3)

② Vertical subgrade reaction (kN/m3) q 1 = K V ( e + B ¤ 2 )q q 2 = K V ( e B ¤ 2 )q

｝

(7.4.4)

③ Shear reaction force that acts at the wall bottom (kN/m) Q = K S ( h D )q A

(7.4.5)

④ Horizontal displacement of the wall (m) d z = ( h z )q

(7.4.6)

⑤ Angle of wall rotation (º) MK 1 + HK 3 q = -------------------------------K 1 K4 K 2 K 3

(7.4.7)

⑥ Depth of the center of wall rotation (m) MK 2 + HK 4 h = ----------------------------MK 1 + HK 3

(7.4.8)

⑦ Distance from the wall center axis to the center of rotation of the wall (m) ì ü n n æi1 d iö ï 1 ïV ç ÷ e = ----------- í --- h å K Hi d i tan d i + å K Hi d i å d j + ---- tan d i ý ç ÷ KV A ï q 2 ï i=1 i=1 èj = 1 ø î þ where n

K1 =

å K Hi di + K S A

i=1

ì ü æi1 d iö ï ï ç ÷ K 2 = å í K Hi d i å d j + ---- ý + K S AD ç 2÷ ï i = 1ï èj = 1 ø î þ n

-446-

(7.4.9)


ì æi1 öü di B ï ï ç K 3 = å í K Hi d i å d j + ---- + --- tan d i÷ ý + K S AD ç ÷ 2 2 i = 1ï èj = 1 øï î þ n

ì i i1 æd 2 öü æi1 d iö ï ï 1 Bç i ç ÷ K 4 = å í K Hi d i ----- + å d j å d j + --- å d j + ---- tan d i÷ ý + K S AD 2 + ------ K V A 3 ç3 ÷ ç ÷ 12 2 2 i = 1ï j=1 j=1 è øï èj = 1 ø î þ n

The angle of wall friction d is negative for strata whose horizontal subgrade reaction force acts on the front of the wall, and positive for strata whose horizontal subgrade reaction force acts on the rear of the wall. (b) When the vertical subgrade reaction acts in a triangular distribution The horizontal subgrade reaction, horizontal wall displacement, angle of rotation, and depth of the center of rotation are expressed in the same form as those in (a). ① Vertical subgrade reaction (kN/m2) B q 1 = K V æ e + ---ö q è 2ø

(7.4.10)

②Shear reaction that acts at the wall bottom (kN/m) Q = K S ( h D )qA¢ where B A¢ = e + --2

(7.4.11)

③Distance between the wall center axis and the center of rotation of the wall (m)

e =

ì ü n n æi1 d iö ï B 2 ïV ç ÷ ------ í --- h å K Hi d i tan d i + å K Hi d i å d j + ---- tan d i ý --ç ÷ KV ï q 2 ï 2 i=1 i=1 èj = 1 ø î þ

(7.4.12)

where n

K1 =

å K Hi di + K S A¢

i=1

ì ü æi1 d iö ï ï ç ÷ K 2 = å í K Hi d i å d j + ---- ý + K S A¢D ç 2÷ ï i = 1ï èj = 1 ø î þ n

ì æi1 öü di B ï ï ç K 3 = å í K Hi d i å d j + ---- + --- tan d i÷ ý + K S A¢D ç ÷ 2 2 i = 1ï èj = 1 øï î þ n

ì i i1 æd 2 öü æi1 d iö ï ï 1 Bç i ç ÷ K 4 = å í K Hi d i ----- + å d j å d j + --- å d j + ---- tan d i÷ ý + K S A¢D 2 + --- K V A¢ 2 ( B e ) ç3 ÷ ç ÷ 2 6 2 i = 1ï j=1 j=1 è øï èj = 1 ø î þ n

The angle of wall friction d should be negative for strata whose horizontal subgrade reaction acts on the front of the wall, and positive for strata whose horizontal subgrade reaction acts on the rear of the wall. The notations used in equations in (a) and (b) are as follows: V： vertical force acting on the wall (kN/m) H： horizontal force acting on the wall (kN/m) M： moment acting on the center of the wall at the level of ground surface (kN•m/m) (external forces that act on the wall are those for the unit length in the direction along the face line of wall) D： embedded length (m) di： thickness of each soil layer of the ground between sea bottom and wall bottom (m) B： equivalent width (m) -447-


KHi： modulus of horizontal subgrade reaction of each layer of the strata between sea bottom and wall bottom (kN/m3) KV： modulus of vertical subgrade reaction at cell bottom (kN/m3) KS： horizontal shear modulus (at bottom) (kN/m3) A： area of wall bottom per unit length of the wall in the direction face line (m2/m) A¢： area of wall bottom per unit length of the wall in the direction of face line (when the value of vertical subgrade reaction is positive)(m2/m)

7.5 Examination of Bearing Capacity of the Ground The subgrade reaction acting on the wall bottom shall not exceed the allowable bearing capacity for the ground.

7.6 Examination against Sliding of Wall For the examination of wall stability against sliding, the shearing stress between the wall bottom and the ground shall not exceed an allowable shearing stress. [Technical Notes] (1) The allowable shear resistance force acting between the wall bottom and the ground is calculated in accordance with equation (7.6.1). 1 Q a = ----- (Vertical resultant force of subgrade reaction) ´ tan f Fs where Qa： allowable shear resistance force (kN/m) φ： internal friction angle of the soil at wall bottom (º) Fs： safety factor; 1.2 or greater for ordinary conditions and 1.0 or greater during an earthquake

(7.6.1)

(2) The vertical components of the earth pressure acting on the front and rear of the wall that should be taken into consideration include (a) the vertical component of the active earth pressure, (b) the friction force due to the earth pressure below the ground surface, and (c) the vertical component of the passive earth pressure. The vertical component of earth pressure is considered a positive force when it acts in the same direction as that of the wall weight. (3) When the internal friction angle of the soil above the wall bottom is different from that below the wall bottom, it is recommended to use the smaller value as the internal friction angle at the wall bottom.

7.7 Examination of Displacement of Wall Top In principle, the horizontal displacement of the wall top shall not exceed the allowable limit of displacement. [Technical Notes] (1) In general, the allowable displacement is determined in consideration of the structural stability and usage of the structure. Because an excessive displacement of the wall top may hamper functional operation of mooring facilitis even when the subgrade reaction force etc. are lower than the respective allowable limits, it is desirable to define an allowable displacement. (2) In the determination of an allowable displacement for a mooring facility, the following factors are taken into consideration 6): (a) Structural stability (b) Function (c) Construction and maintenance cost (3) In the determination of allowable displacement, it is usually convenient to use the relationship between the top displacement of gravity type or sheet pile structures and the levels of damage that has been obtained from past earthquake damage reports 7). This relationship is shown in Fig. T- 7.7.1. The figure shows that the ratio of the top displacement to the wall height (d/H) is larger than 1.5% for most of the structures that have been categorized into “Damage Level II ”. When performing design calculations in accordance with the method described in this section, therefore, it is desirable to make the allowable horizontal displacement for the wall top equal to or less than 1.5% of the wall height.

-448-


Table T- 7.7.1 Damage Levels Damage level

Description No damage.

I

No damage has been found in the main body, but destruction and/or deformation has been found in ancillary structures.

II

Significant deformation has occurred in the main body.

III

The structure still holds its original shape, but the main body is considered to have been wrecked.

IV

The structure has been demolished and collapsed.

Top displacement of structure / height of structure measured from sea bottom to wall top (%)

0

Sheet pile structures Concrete block structures Concrete caisson structures Other gravity-type concrete structures

Damage level

Fig. T- 7.7.1 Damage Level and “Top Displacement to Wall Height Ratio”

7.8 Examination of Stability against Circular Slip When the ground is not so hard, examination of stability against circular slip shall be made as necessary. In principle, the circular slip surface shall not go through the inside of the wall in this analysis. [Commentary] When the internal friction angle of the soil behind the wall and the ground is 30º or larger, the examination of stability against circular slip is often omitted. In the case of steel sheet pile cellular-bulkhead quaywalls, it may be assumed that the wall is a rigid body and thus the circular slip surface does not go through the inside of the wall.

7.9 Layout of Cells and Arcs The cells and arcs shall be arranged to make the area of the cellular-bulkhead to be equal to the area of the virtual wall with the equivalent width obtained in 7.3 Examination of Wall Width against Shear Deformation or 7.4 Examination of Stability of Wall Body as a Whole. [Technical Notes] (1) Cells should be arranged evenly along the total length of the face line of the quaywall wherever possible. In general, it is advisable to set the cell center interval being 10 to 15% larger than the cell diameter. (2) Arcs should be arranged in such a way that they are connected perpendicularly to the wall of cell shell. The radius of the arc should be made smaller than that of the cell. (3) In general, front tips of arcs tend to shift forward during and/or after the filling work. Therefore it is advisable to arrange arcs in such a way that their front tips are located about 100 to 150 cm inside the front face line of cell walls. It is also advisable to arrange cells in such a way that their front face line is located about 30 cm inside the design face line of the quaywall.

7.10 Calculation of Hoop Tension Hoop tension of sheet pile of the cells and arcs shall not exceed the allowable hoop tension.

-449-


[Technical Notes] (1) Sheet pile tension is usually calculated at the sea bottom. In principle, an allowable hoop tension value of 2,000 kN/m is used for a straight sheet pile having joint tensile strength of 5,900 kN/m, and an allowable sheet pile tension value of 1,500 kN/m is used for a straight sheet pile having joint tensile strength of 3,900 kN/m. (2) Hoop tension should be calculated in accordance with equation (7.10.1). T = [(g0H0¢ + q) Ki + gwhw]R where T： tension of sheet pile (kN/m) Ki： coefficient of earth pressure of the filling g0： equivalent unit weight of the filling (kN/m3) gw： unit weight of sea water (kN/m3) H0¢： equivalent wall height (m) hw： difference between water level inside cell and that in front of cell (m) R： radius of cell (m) q： surcharge (kN/m2)

(7.10.1)

(3) The equivalent wall height H0¢ should be calculated using equation (7.3.13) in 7.3.4 Calculation of Resisting Moment [Technical Notes] . (4) When no compaction is performed, the coefficient of earth pressure can be set as 0.6 for the filling material. When the filling is to be compacted, tanφ should be used as the coefficient of earth pressure of the filling, because the internal pressure of the cell and the internal friction angle of the filling become larger. The coefficient of filling earth pressure of the filling for the arc sections should be 1/2 tanφ.

7.11 Design of T-Shaped Sheet Pile 7.11.1 General Cells and arcs shall be connected by using T-shaped sheet piles. [Technical Notes] T-shaped sheet pile is a sheet pile with a special cross section to join the cell to arcs (see Fig. T- 7.11.1).

Fig. T- 7.11.1 T-Shaped Sheet Pile

7.11.2 Structure of T-Shaped Sheet Pile The structure of T-shaped sheet pile shall have sufficient safety against the tension acting on the sheet pile of cells and arcs. [Technical Notes] (1) The standard structures of T-shaped sheet pile are shown in Figs. T- 7.11.2 and T- 7.11.3.

(Material equivalent to SM-490A)

Rivet 25 (SV-400) Rivet interval: 85mm Straight sheet pile (SY-295) = 12.7mm t

PL 270 x 14 (material equivalent to SM-490A)

(Units: mm) Fig. T- 7.11.2 Standard Cross Section of T-Shaped Sheet Pile (for rivet connection with rivet intervals of 85 mm)

-450-


PL - 200 x 12 (SM-490A)

Straight sheet pile (SY-295) = 12.7mm t

(Units: mm)

Fig. T- 7.11.3 Standard Cross Section of T-Shaped Sheet Pile (for welding connection)

(2) Strength of the cross sections shown in Figs. T- 7.11.2 and T- 7.11.3 has been confirmed by a breaking test where the tensile strength of the joint of the sheet pile in a cell is 3,900 kN/m (allowable sheet pile tension = 1,500 kN/m) and the arc diameter is 2/3 or less of the cell (tensile strength = 2,600 kN/m, allowable sheet pile tension = 980 kN/m). The rivet and welding joints for tests were made in a factory.

7.12 Detailed Design 7.12.1 Design of Pile to Support Coping Piles to support coping shall be designed against the vertical force and horizontal force, and/or bending moment. 7.12.2 Design of Coping The reinforcement arrangement calculation of coping shall be made appropriately on the following parts: (1) Upright part (2) Bottom slab [References] 1) Kunio TAKAHASHI, Setsuo NODA, Katsumi KANDA, Satoshi MIURA, Taisaku MIZUTANI, Shigeki TERASAKI: “Horizontal loading tests on models of steel pile cellular-bulkhead-Part 1 Static behavior”, Tech. Note of PHRI, No. 638, 1989 (in Japanese). 2) Setsuo NODA, Kunio TAKAHASHI, Katsumi KANDA, Shigeki TERASAKI, Satoshi MIURA, Taisaku MIZUTANI,: “Horizontal loading tests on models of steel pile cellular-bulkhead-Part 2 Dynamic behavior”, Tech. Note of PHRI, No. 639, 1989 (in Japanese). 3) Yukimitsu YOKOYAMA: “Design and Execution of Steel Pile”, Sankaido, 1963, pp.95-96 (in Japanese). 4) Iwao YOSHIDA, Ryunoshin YOSHINAKA: “Engineering properties of Akashi and Kobe Strata”, Rept. of Public Works Research Inst., Vol. 129, 1966 (in Japanese). 5) Iwao YOSHIDA, Yoshio ADACHI: “Experimental studies of static horizontal resistance of caisson foundation”, Rept. of PWRI, Vol. 139, 1970, pp. 24-25. 6) Setsuo NODA, Sosuke KITAZAWA, Takeshi IIDA, Nobuo MORI, Hiroshi TABUCHI: “An experimental study on the earthquake resistance of steel plate cellular-bulkheads with embedment”, Rept of PHRI, Vol. 21, No. 3, 1982 (in Japanese). 7) Yukio SAIMURA, Kiyohiro MORIMOTO, Yukinori TAKASE: “Results of field measurement on the earth pressure of fill of the embedded steel pipe pile cellular-bulkhead”, Proc. 36th Annual Meeting of JSCE, Part 3, 1981 (in Japanese).

-451-


Chapter 8 Steel Plate Cellular-Bulkhead Quaywalls 8.1 Scope of Application The provisions in this chapter shall be applied to the design of quaywalls with placement-type and embedded-type steel plate cellular-bulkhead structures. [Commentary] Placement-type steel plate cellular-bulkhead quaywalls are steel plate cellular-bulkhead quaywalls without an embedded section. In many cases these quaywalls are constructed on a strong foundation subsoil whose bearing capacity is considered sufficiently large or on the subsoil that has been improved to have sufficient bearing capacity. Embedded-type steel plate cellular-bulkhead quaywalls are steel plate cellular-buklhead quaywalls whose steel plate cells are partly embedded into sandy subsoil to improve the stability of the structure.

8.2 Placement-Type Steel Plate Cellular-Bulkhead Quaywalls 8.2.1 Principle of Design The provisions in this section shall be applied to the design of steel plate cellular-bulkhead quaywalls without an embedded section. [Technical Notes] (1) The design of placement-type steel plate cellular-bulkhead quaywalls is usually made according to the procedure depicted in Fig. T- 8.2.1.

Determination of design conditions Assumption of wall dimension Calculation of external forces No

Examination against shear deformation Yes Examination of stability of the wall body as a whole

No

Yes Examination of bearing capacity of the ground

No

Yes Examination of stability against circular slip

No

Consideration of subsoil improvement

Yes Determination of cell plate thickness Layout of cells and arcs Detailed design Fig. T- 8.2.1 Sequence of Design of Placement-Type Steel Plate Cellular-Bulkhead Quaywall

(2) Figure T - 8.2.2 shows an example of placement-type steel plate cellular-bulkhead quaywalls. (3) In general, (a) the examination of wall deformation is required for design in ordinary conditions only and (b) the examination of wall stability against overturning is required for seismic design only. (4) For the cell filling material and the backfill material, well-compacted sand of good quality or a material that does not liquefy, such as gravel, should be used in order to eliminate the risk of liquefaction during an earthquake.

-452-


Face line

V-shaped rubber fender HWL LWL

Steel plate cell Filling material

Steel pipe pile

Frontal protection sand Replaced sand

Steel pipe pile (Units: m)

Steel pipe pile

Fig. T- 8.2.2 Example of Steel Plate Cellular-Bulkhead Quaywall

8.2.2 External Forces Acting on Steel Plate Cellular-Bulkhead The external forces acting on steel plate cellular-bulkhead shall be in accordance with 7.2 External Forces Acting on Steel Sheet Pile Cellular-Bulkhead Quaywall. 8.2.3 Examination of Wall Width against Shear Deformation Examination of the shear deformation of the wall body shall be made for the loads acting under ordinary conditions. The equivalent wall width of the wall body shall be determined in accordance with 7.3.2 Equivalent Width of Wall. Calculation of the deformation moment shall be performed in accordance with 7.3.3 Calculation of Deformation Moment. The resisting moment shall be calculated appropriately in consideration of the structural characteristics of the cellular-bulkhead and the deformation of the wall. [Commentary] The term “deformation moment” in the above paragraph refers to the moment at the sea bottom due to external forces acting on the wall body above the sea bottom, including active earth pressure and residual water pressure. [Technical Notes] (1) The safety factor against shear deformation should basically be 1.2 or greater for ordinary conditions (i.e., nonearthquake environments). (2) When the deformation of the wall body is not to be allowed (i.e., when the horizontal displacement of the cell top is approximately less than 0.5% of the cell height), the resisting moment against deformation is calculated using equations (8.2.1) and (8.2.2). Mrd = (1/6) g 0 Hd¢3 Rd Rd = where Mrd ： Hd¢： Rd： g 0： n d： B： f：

nd

2(3

(8.2.1)

－ n d cos f ) sin f

(8.2.2)

resisting moment against deformation of steel plate cellular-bulkhead (kN•m/m) equivalent wall height used in the examination of deformation of cellular-bulkhead (m) deformation resistance coefficient equivalent unit weight of filling (kN/m3) equivalent “wall width / wall height” ratio used in calculating resisting moment, n d = B/Hd' equivalent wall width (m) internal friction angle of filling material (º) -453-


(3) In the calculation of resisting moment, the equivalent wall height of the wall Hd ¢ is calculated by means of equation (8.2.3). The height Hd ¢ is that above the sea bottom. 1 Hd ¢ = ----S g i h i g0 where g i： unite weight of the i-th layer of filling (KN/m3) h i： thickness of the i-th layer (from sea bottom to wall top) (m) g 0： equivalent unit weight of filling (kN/m3) ; normally, g 0 = 10 kN/m3

(8.2.3)

In the calculation of the equivalent wall height Hd ¢, surcharge may be ignored as discussed in 7.3.4 Calculation of Resisting Moment. (4) When the filling material can be regarded as uniform, the height Hd of the quaywall top above the sea bottom can be used in place of the equivalent wall height Hd ¢ of equation (8.2.1).

8.2.4 Examination of Stability of Wall Body as a Whole Examination of the stability of the wall body as a whole against sliding shall be performed in accordance with 4.3.2 Examination against Sliding of Wall. In the calculations to examine the stability of a steel plate cellular-bulkhead against overturning, the stability of cellular-bulkhead shall be examined against the external forces acting above the wall bottom, including earth pressure, residual water pressure, and earthquake forces. [Technical Notes] (1) The safety factor against overturning for extraordinary conditions should be 1.1 or greater. (2) The resisting moment of steel plate cellular-bulkhead against overturning is calculated by means of equations (8.2.4) and (8.2.5). Mro = (1/6) g0 H ¢3 Rt 2

(8.2.4) (a2

b2)

+ 6n b (8.2.5) － Rt = n ¢ (3 － n ¢ cos f ) sin f + 3 where Mro： resisting moment of steel plate cellular-bulkhead against overturning (kN•m/m) a = Ka tan d b = Ka tan d + ( n ¢/2) (4 - n ¢ cos f ) tan f tan d n ¢ = n － (a + b ) H ¢：： equivalent wall height of the cellular-bulkhead to obtain the resisting moment against overturning (m) Rt： overturning resistance coefficient n： equivalent “wall width /wall height” ratio of the cellular-bulkhead, n = B/H ¢ B：： equivalent wall width of the cellular-bulkhead (m) d： wall friction angle of filling material (º) ; normally, d = 15°is used. Ka： active earth pressure coefficient of filling material For other symbols, refer to those used in equations (8.2.1) and (8.2.2). (3) The equivalent wall height H ¢ used to calculate the resisting moment against overturning should be calculated using equation (8.2.6). 1 H¢ = g----S g i h i (8.2.6) 0 where, g i ： unite weight of the i-th layer of filling (kN/ m3) h i： thickness of the i-th layer (from wall bottom to wall top) (m) (4) Because the equation to calculate the overturning resistance coefficient Rt is complex, the relationship between Rt, f , and n is given in Fig. T- 8.2.3. The figure assumes that d is 15º. (5) In general, the filling of a steel plate cellular-bulkhead used as a quaywall is not uniform because the major portion of such filling is located under the water surface and thus subjected to buoyancy. Therefore, the equivalent wall height is used here as in the calcula-454-

Fig. T- 8.2.3 Relationship between Overturning Resistance Coefficient Rt , n , and f


tion of the resisting moment of the steel sheet pile cellular-bulkhead against deformation. When the filling material can be considered as uniform, the total wall height of the cellular-bulkhead H may be used in the same calculation in place of the equivalent wall height H ¢ of equation (8.2.4). Figure T- 8.2.3 may be used in this calculation as well. (6) The overturning moment is the moment at the bottom of cellular-bulkhead due to the external forces acting above the bottom. The equivalent wall height of the cellular-bulkhead H ¢ used in the calculation of the resisting moment should be a height above the cell bottom.

8.2.5 Examination of Bearing Capacity of the Ground The maximum subsoil reaction force generated at the front toe of the cellular-bulkhead shall be calculated appropriately in consideration of the effect of the filling material acting on the front wall of the cellularbulkhead . [Technical Notes] (1) The maximum subsoil reaction force generated at the front toe of the cellular-bulkhead should be examined by means of equation (8.2.7). 1 V t = --- gH 2 tan 2 f 2 where Vt： maximum reaction force generated at the front toe of cellular-bulkhead (kN/m) g： unit weight of filling (kN/m3) H： total wall height of the cellular-bulkhead (m) f： internal friction angle of filling material (º)

(8.2.7)

Equation (8.2.7) represents the weight of the filling material mass which is hanging on the front wall under the assumption that the product of the earth pressure coefficient of the filling material and the wall friction coefficient is equal to tan 2 f . When the filling material is not uniform, therefore, calculation may be performed in the same manner as that for earth pressure calculation. (2) The wall height H should normally be considered as the height of the wall top above the wall bottom. However, when the superstructure of the cellular-bulkhead is supported by foundation piles, it may be considered as the height of the bottom of superstructure above the wall bottom. (3) Equation (8.2.7) represents the cell front toe reaction force when the overturning moment is roughly equal to the overturning resistance moment of equation (8.2.4). Without occurrence of overturning, the reaction force is smaller than the value obtained from equation (8.2.7). According to a model experiment, the maximum front toe reaction force Vt is nearly proportional to the overturning moment. Therefore reaction force without occurrence of overturning should be calculated using equation (8.2.8). (8.2.8) V = Vt (M / Mro) where V： front toe reaction force of the cellular-bulkhead corresponding to overturning moment M (kN/m) M： overturning moment (kN•m/m) Mro： resisting moment against overturning (kN•m/m) Hence, use of larger cell radius makes the cellular-bulkhead safer against overturning by increasing the resisting moment Mro, while reducing the front toe reaction force V. (4) The examination of the bearing capacity of the ground should be calculated referring to Part Ⅴ,　Chapter 2 Bearing Capacity of Shallow Foundations.

8.2.6 Examination of Stability against Circular Slip The examination of the sability of the wall against circular slip shall be made in accordance with 7.8 Examination of Stability against Circular Slip. 8.2.7 Determination of Thickness of Steel Plate of Cell Shell The thickness of steel plate of the cell shell shall be determined by calculating appropriately the maximum horizontal tension acting on the cell shell. [Technical Notes] (1) The maximum horizontal tension acting on the cell shell should be calculated by means of equation (8.2.9). T = [(g0Hd ¢ + q) tan f + gw hw] R

(8.2.9) -455-


where T： R： Hd ¢： hw： g0： gw： f： q：

horizontal tension (kN/m) radius of cell (m) equivalent wall height above the sea bottom surface (m) (refer to equation (8.2.3)) difference between the water level in front of the cell and that of the inside of the cell (m) equivalent unit weight of filling (kN/m3) unit weight of sea water (kN/m3) internal friction angle of filling material (º) surcharge (kN/m2)

When the filling material is uniform, the height of the cell top above the sea bottom Hd can be used in place of the equivalent wall height Hd ¢. (2) The thickness of the cell plate should be such that the stress due to the tension of equation (8.2.9) becomes equal to or less than the allowable stress in accordance with PartⅢ Ⅲ, 2.3 Allowable Stress. Considering corrosion and rigidity of the cell shell, a minimum cell plate thickness of 6 mm should be used.

8.2.8 Layout of Cells and Arcs The layout of cells and arcs shall be determined in accordance with 7.9 Layout of Cells and Arcs 8.2.9 Detailed Design The detail design of placement-type steel plate cellular-bulkhead quaywall shall be made in accordance with 7.12 Detailed Design.

8.3 Embedded-Type Steel Plate Cellular-Bulkhead Quaywalls 8.3.1 Principle of Design The provisions in this section shall be applied to the design of steel plate cellular-bulkhead quaywalls with an embedded section. [Technical Notes] (1) The designing of embedded-type steel plate cellular-bulkhead quaywall is usually made in accordance with the procedure depicted in Fig. T- 8.3.1. (2) Embedded-type steel plate cellular-bulkhead structures are structures whose stability has been improved by means of embedment of their steel plate cell shell in sandy subsoil. The design method described in this section is based on the results of model experiments on embedded steel plate cell shell mainly driven into sandy subsoil and applies to embedded-type steel plate cellular-bulkhead structures whose embedded length is about 1/8 or more of the wall height 1), 2). For cases where (a) the sobsoil is hard and it is not possible to achieve an embedded length of 1/8 or more of the wall height, (b) the ratio of the equivalent wall width to the wall height is significantly different from those of ordinary cells, which are about 1/2 to 2, and/or (c) the structure is to be constructed on clayey subsoil or subsoil improved by means of sand compaction piles etc., further examination should be made in addition to the design method described in this section, as these cases involve unsolved problems.

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Determination of design conditions Assumption of wall dimensions Re-assumption of wall dimensions or consideration of soil improvement

Calculation of external forces

Examination against shear deformation Estimation of coefficient of subgrade reaction Calculation of subgrade reaction and wall displacement Examination of bearing capacity of subsoil Examination against sliding of wall Examination of wall top displacement Examination of stability against circular slip

Consideration of soil improvement

Layout of cells and arcs Determination of plate thickness of cells and arcs Detailed design

Fig. T- 8.3.1 Sequence of Design Embedded-Type Steel Plate Cellular-Bulkhead Quaywall

(3) Figure T- 8.3.2 shows an example of embedded-type steel plate cellular-bulkhead quaywalls.

LWL Filling sand

Steel plate cell

Silty clay

Replacement sand Compaction area

Fig. T- 8.3.2 Example of Embedded-Type Steel Plate Cellular-Bulkhead Quaywall

8.3.2 External Forces Acting on Embedded-Type Steel Plate Celluler-Bulkhead The external forces acting on embedded-type steel plate cellular-bulkheads shall be calculated in accordance with 7.2 External Forces Acting on Steel Sheet Pile Cellular-Bulkhead Quaywall. 8.3.3 Examination of Wall Width against Shear Deformation Examination of the strength of the wall body against shear deformation shall be made in accordance with 7.3 Examination of Wall Width against Shear Deformation. -457-


8.3.4 Examination of Stability of Wall Body as a Whole Examination of the stability of embedded-type steel plate cellular-bulkhead as a whole shall be made in accordance with 7.4 Examination of Stability of Wall Body as a Whole. 8.3.5 Examination of Bearing Capacity of the Ground Examination of the bearing capacity of the ground shall be made in accordance with 7.5 Examination of Bearing Capacity of the Ground. 8.3.6 Examination against Sliding of Wall Examination of the stability of the cellular-bulkhead against sliding shall be made in accordance with 4.3.2 Examination against Sliding of Wall and 7.6 Examination against Sliding of Wall. 8.3.7 Examination of Displacement of Wall Top Examination of the horizontal displacement at the top of the cellular-bulkhead shall be made in accordance with 7.7 Examination of Displacement of Wall Top. 8.3.8 Examination of Stability against Circular Slip Examination of the stability of the cellular-bulkhead against circular slip shall be made in accordance with 7.8 Examination of Stability against Circular Slip. 8.3.9 Layout of Cells and Arcs The layout of cells and arcs shall be determined in accordance with 7.9 Layout of Cells and Arcs. 8.3.10 Determination of Plate Thickness of Cell Shell and Arc Section The maximum horizontal tensile stress generated in the cell shell and arc sections shall not exceed the allowable stress. [Technical Notes] (1) The maximum horizontal tensile stress generated in the cell shell and arc section should be calculated using equation (8.3.1). (8.3.1) s=T/t where s： maximum horizontal tensile stress generated in the cell shell or arc section (kN/mm2) t： plate thickness of the cell shell and arc section (mm) T： maximum horizontal tension acting on the cell shell or arc section under ordinary conditions (kN/mm). In addition to the horizontal tensile stress due to the filling, the cell shell is subjected to vertical stress and shearing stress. According to model experiments and in-situ measurements, however, the horizontal tensile stress is dominant among the stresses generated in the cell shell. Therefore, examination of the strength needs to be made only against the horizontal tension. (2) The maximum horizontal tension acting on the cell shell and arc section should be calculated in accordance with 8.2.6 Determination of Thickness of Steel Plate of Cell Shell. (3) As a method to determine the cell plate thickness against the stress generated during the driving of the cell shell, equation (8.3.2) has been suggested based on results of cylindrical cell buckling experiments and past construction data. t ≧ 0.032 ( RND¢ / E)0.5 where t： plate thickness of cell shell (mm) E： Young’s modulus of steel material (kN/mm2) R： radius of cell shell (cm) N： average N-value of the subsoil into which the cell shell is to be driven D¢： driven depth of cell shell (cm)

(8.3.2)

It is desirable to make the plate thickness larger than about 8 mm, because 8 mm is the minimum thickness that has been proven to be safe through past construction experiences.

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8.3.11 Joints and Stiffeners The joints between the cell shells and arc sections shall be structurally safe against the maximum horizontal tension acting on the arcs. They shall also be structurally safe against the stresses generated during manufacturing, transportation, and installation work. 8.3.12 Detailed Design Detailed design of embedded-type steel plate cellular-bulkhead quaywall shall be conducted in accordance with 7.12 Detailed Design. [References] 1) Shoichi KITAJIMA, Setsuo NODA, Tanekiyo NAKAYAMA: “An experimental study on the static stability of the steel plate cellular-bulkheads with embedment”, Tech. Note of PHRI, No. 375, 1981 (in Japanese). 2) Setsuo NODA, Sosuke KITAZAWA, Takeshi IIDA, Nobuo MORI, Hiroshi TABUCHI: “An experimental study on the earthquake resistance of steel plate cellular-bulkheads with embedment”, Rept of PHRI, Vol. 21, No. 3, 1982 (in Japanese).

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Chapter 9 Open-Type Wharves on Vertical Piles 9.1 Principle of Design (1) The provisions in this chapter shall be applied to the design of open-type wharves on vertical piles. (2) Open-type wharves on vertical piles shall be designed to satisfy the performance requirements against the loads due to berthing and mooring of ships, earthquakes, etc. [Commentary] (1) The provisions in this chapter are basically applied to open-type wharves on vertical steel pipe piles or shaped steel piles, and may also be used for similar structures provided that the dynamic characteristics of the structures be taken into consideration. (2) Open-type wharves on vertical piles should be designed in accordance with the method described in this chapter. Examination of general earthquake-resistant characteristics other than those described in this chapter should be made in accordance with PartⅡ Ⅱ, Chapter 12 Earthquaks and Seismic Forces. (3) In the design of open-type wharves on vertical piles, it is necessary to give consideration to their dynamic behaviors during an earthquake, because these structures are generally more flexible than other types of structures including those of gravity type. Examination of earthquake-resistant performance considering the dynamic behavior of the structure should be made in accordance with 9.6 Examination of EarthquakeResistant Performance. (4) The design of open-type wharves described here is based on the assumption that no major deformation should occur in the foundation or the ground behind the structure due to liquefaction etc. When there is a risk of major deformation of the foundation or nearby ground, therefore, it is necessary to take appropriate preventive measures. [Technical Notes] (1) Design of open-type wharves on vertical piles should normally be made in accordane with the procedure depicted in Fig. T- 9.1.1. In the design flow shown in the figure, (a) a cross section of open-type wharf is designed at first to resist nonearthquake forces, (b) the cross section is examined against earthquake force, and (c) the design proceeds to detailed design stage if the earthquake-resistance performance of the cross section is satisfactory. If not, the cross section should be modified and examined again. That is, the steel pipe piles are designed in accordance with the allowable stress method for non-earthquake forces in the first step of the design flow, and the load carrying capacity design method is used in the second step against earthquake forces, which assumes that the deformation of steel pipe piles enters the plastic zone during an earthquake in absorbing seismic force energy. (2) In principle, the “performance requirement” should mean (a) that the stress generated in the steel pipe pile should be smaller than the allowable stress against non-earthquake forces, and (b) that the settlement, damage level, etc. are acceptable in terms of the allowable values defined based on the importance of the facility or other relevant criteria against earthquake forces. The bearing capacity of the pile and overall stability of the structure may also be used as the measures of performance requirement. (3) Figure T- 9.1.2 shows an example of open-type wharves on vertical piles. As shown in the figure, an open-type wharf comprises a pile-supported section and an earth retaining section. (4) When the wharf is to be constructed on the ground where consolidation or compression of soil is expected to occur, it is desirable to separate the earth-retaining section from the pile-supported section, because the earthretaining section usually settles more than the pile-supported section. It is advisable to give careful consideration in the design and/or construction work to the fact that dragged settlement or lateral movement of subsoil may occur in part or all of the pile-supported section due to the settlement of the earth-retaining or landfill section even when the pile-supported section has been separated from the earth-retaining section. (5) It has been reported that pile-supported sections were damaged due to the liquefaction of their foundation and/or the ground behind the earth-retaining revetment in the Hyogoken Nanbu Earthquake (1995). The typical damage was (a) the buckling of steel pipe piles near the boundaries between different soil strata due to the liquefaction of the ground and (b) the buckling of steel pipe pile head and/or coping cracking due to the loads generated when the revetment was pushed seaward. Because it is not clear yet how to incorporate these phenomena into the design, it is stated in [Commentary] (4) of this section that sufficient measures should be taken to prevent the occurrence of these phenomena. However, as it is difficult to totally prevent the deformation of the ground during an earthquake, it is desirable to examine in advance the expected degree of ground deformation and the effect of the deformation on facililies through numerical analysis or another appropriate method for particularly important facilities.

-460-


Design conditions

Determination of the size of each block and the cross section and arrangement of piles

Analysis of slope stability

Assumption of superstructure dimensions

Stability calculation for the earth-retaining section

Arrangement of mooring posts and fenders

Calculation of the deadweight of superstructure etc.

Calculation of design external forces Surcharge Vertical forces

Wind load (crane, etc.)

Live loads (crane, etc.) Uplift (when the wharf is expected to be subjected

Horizontal forces

Reaction force of fenders Tractive force

to significant wave actions)

Assumptions about the sea bottom subsoil

Calculation of the section forces acting on each pile

Examination of the stresses generated in each pile

Examination of the bearing capacity of each pile Axial direction Tranaverse direction

Examination of the displacement of wharf as a whole

Cross section to examine the earthquake-resistant perfotmance

Examination of earthquake-resistant performance Level 1 ground motion Level 2 ground motion (in the case of a wharf strengthened against strong earthquake)

Detailed design

Fig. T- 9.1.1 Flowchart of the Designing of Open-Type Wharf on Vertical Piles

(6) When cargo handling equipment, such as container cranes, is to be installed on an open-type wharf on vertical piles, it is desirable to install it in such a way that all of its feet are positioned on either the pile-supported section or earth-retaining section. If, for example, one foot of a cargo handling equipment is positioned on the pile-supported section and another is positioned on the earth-retaining section, the equipment becomes susceptible to adverse effects such as differential settlement and unsychronized seismic motions, due to the difference in the response characteristics of the two sections. When it is unavoidable to position one foot of a cargo handling equipment on the pile-supported section and another on the earth-retaining section, it is recommended to provide a sufficient foundation work (such as foundation piles) to prevent differential settlement due to the settlement of the shore side foot. In this case, it is common practice not to locate a fixed foot of cargo handling equipment (such as portal crane) on the pile-supported section. (7) When cargo handling equipment, such as container cranes, is to be installed on an open-type wharf on vertical piles, it is recommended to conduct an earthquake response analysis in consideration of the combined vibration of the cargo handling equipment and the pile-supported section, because the response characteristics of the pilesupported section during an earthquake may vary widely depending on the mass ratio and the natural period ratio between the equipment and the pile-supported section. For details, refer to Part Ⅸ , 2.4 Cargo Handling Equipment. (8) It is recommended that the calculation of the bending moment of piles be performed for both the parallel and transverse directions to the wharf face line. (9) When it is considered necessary to examine the effect of the rotation of the deck block in the calculation of member forces acting on the piles, the design calculation should be made by taking this effect into account. (10) At the areas where significant wave actions are expected, it is recommended to examine the following resistance characteristics against the uplift forces acting on the superstructure and access bridge: (a) Stability of the access bridge against the uplift and the pulling resistance of the piles (b) Strength of the members of the superstructure and access bridge against the uplift -461-


For details on the uplift by waves, refer to PartⅡ Ⅱ, 5.5.1 Uplift Acting on Horizontal Plate near the Still Water Level. Shoreside

(Apron)

Seaside

Bollard 250kN Connection plate Curbing

Concrete pavement

HWL Backfilling sandstone

Fender H = 600 LWL

Mortar lining

Foundation rubble stones (30 to 50 kg)

Steel pipe pile SKK 490

= 1000


= 1000


= 1000

Fig. T- 9.1.2 Example of Open-type Wharf on Vertical Piles

9.2 Layout and Dimensions 9.2.1 Size of Deck Block and Layout of Piles The size of a deck block and the distances between piles and pile rows shall be determined appropriately in consideration of the following: (1) Apron width (2) Location of the quay shed (3) Sea bottom ground (especially slope stability) (4) Existing revetments (5) Matters related to construction work such as the concrete casting capacity (6) Static loads and live loads (especially crane sizes) 9.2.2 Dimensions of Superstructure The dimensions of the superstructure of pile-supported section shall be determined appropriately in consideration of the following: (1) Distances between piles and pile rows, and the shape and dimensions of piles (2) Static loads and live loads (3) Tide level (4) Easiness of form work and timbering (5) Ground conditions -462-


(6) Arrangement of bollards (7) Arrangement, shape, and dimensions of fenders 9.2.3 Arrangement of Fenders and Bollards Fenders and bollards shall be so arranged that exertion of eccentric external forces to a single deck block be minimized as much as possible.

9.3 External Forces Acting on Open-Type Wharf 9.3.1 Design External Forces (1) The main external forces acting on the pile-supported section of an open-type wharf shall be as listed in Table 9.3.1. Table 9.3.1 Design External Forces Vertical forces Deadweight of the superstructure Static loads Live loads Train load Vehicle load Cargo handling equipment load Sidewalk live load Tractive force of vessel Uplift

Horizontal forces Earthquake forces acting on the superstructure Earthquake forces acting on the static loads Earthquake forces acting on the live loads Wind load acting on the live loads Reaction force of the fender Tractive force of vessel

(2) The earthquake forces acting on the superstructure, static loads, and live loads shall be considered in the examination of the earthquake-resistant performance of the pile-supporting section. (3) The external forces acting on the earth-retaining section shall be determined by taking the relevant forces described in the respective chapters of this part according to its structural type and then adding the reaction force of the access bridge to it. [Technical Notes] (1) Among the external forces listed in Table 9.3.1, the earthquake forces, the wind load and the tractive force of vessel should normally be considered as the loads under extraordinary conditions. (2) It is a common practice that the deadweight of the superstructure and the earthquake forces acting on the superstructure shown in the table include the deadweight of the piles and the earthquake forces acting on the piles, respectively. (3) The static loads should be determined in accordance with Part Ⅱ, 15.3 Static Load. The earthquake forces due to static loads may normally be considered to act on the upper surface of the deck slab. However, when the center of gravity of the static loads is located at an especially high level, it is important to take the height of the center of gravity as the point of application of the horizontal force. (4) Live loads should be determined in accordance with Part Ⅱ, 15.4 Live Load. The earthquake force due to a railtype traveling crane (see in 9.1 Principle of Design [Technical Notes] (7)) should be calculated by multiplying its deadweight by the seismic coefficient, and the force can be considered to be transmitted from the wheels to the pile-supported section. The wind load acting on crane etc. should be determined in accordance with Part Ⅱ, 3.3 Wind Pressure. (5) The fender reaction force should normally be calculated in accordance with 19.4 Fender System using the external forces calculated in accordance with PartⅡ Ⅱ, 2.2.2 Berthing and PartⅡ Ⅱ, 2.2.3 Moored Vessels, and the reaction force should be considered as a load under ordinary or extraordinary conditions depending on the approaching velocity. When an approaching speed of about 10 cm/s or more is assumed for an general cargo vessel, the reaction force may be considered as a load under extraordinary conditions. For approaching speeds at which vessels berth frequently, it is recommended to consider the reaction force as a load under ordinary conditions. The impact force generated by the oscillatory motions of moored vessels should also be considered as a load under ordinary or extraordinary conditions depending to the meteorological and marine conditions of the site. For general berths, this force may be considered as a load under extraordinary conditions when a vessel is moored there for refuge mooring in an unusual environment such as typhoon. (6) The tractive force of vessels should be determined in accordance with Part Ⅱ, 2.2.4 Tractive Force Acting on Mooring Post and Bollard. In many cases one bollard is installed to one deck block. -463-


9.3.2 Calculation of Fender Reaction Force The calculation of fender reaction force shall be made in accordance with PartⅡ Ⅱ, 2.2 External Forces Generated by Vessels. [Technical Notes] (1) When rubber fenders are installed as a damper on an ordinary large wharf with deck blocks of 20 to 30 m in length, a common practice is to provide two rubber fenders on one block. In many cases, fender intervals of 8 to 13 m are used. A measurement of the berthing behavior of various sizes of vessels has been made by installing 1.5-meter-long rubber fenders on an ordinary large wharf. The measurement has revealed that it is appropriate to calculate the berthing force on the assumption that the vessel’s approaching energy is absorbed by one fender. Therefore, the reaction force may basically be calculated on the assumption that the berthing energy is absorbed by one fender when using rubber fenders as a damper. However, this does not apply when fenders are deployed continuously along the face line of a wharf.

cu e rc fo n

e

tio

rv

ac

cu

en

t-

re

gy er

en

em

t-

en

ac

m

pl is D

Energy or reaction force

(3) Figure T- 9.3.1 shows an example of the pair of “displacement - energy curve” and “displacement - reaction force curve” of a rubber fender. If a single fender absorbs an berthing energy of E1, the corresponding fender deformation d1 is obtained. Then, using the other curve, the corresponding reaction force acting on the pier is obtained as H1(d1 → C → H1). However, if fenders are deployed too close to each other and the berthing energy is absorbed by two fenders, the approaching energy acting on one fender becomes E2 = E1/2 and this causes the corresponding fender deformation d2 . As can be seen from the figure ( d2 → D → H2), the reaction force acting on the pier is almost the same as that generated in the single fender case, because of the characteristics of rubber fender. Thus the horizontal reaction force acting on the pier becomes 2H2≒ 2H1, which means that the horizontal force to be used in the design is twofold. When using fenders that have such characteristics, therefore, it is desirable to give careful consideration to this behavior of reaction force in the designing and locating of fenders.

rv e

(2) The berthing energy is also absorbed by the displacement of the pile-supported section. However, it is a common practice not to take this into consideration because in many cases the energy absorbed by the pile-supported section accounts for less than 10% of the total berthing energy.

pl

is

D

e ac

Displacement

Fig. T- 9.3.1 Rubber Fender Characteristics Curve

9.4 Assumptions Concerning Sea Bottom Ground 9.4.1 Determination of Slope Inclination When an earth-retaining structure is to be constructed behind a slope beneath the superstructure of opentype wharf, the structure shall be located appropriately in consideration of the stability of the slope. [Technical Notes] For soft subsoil, it is recommended that slope stability be examined against a n circular slip as well. However, a slope natio incli e beneath the superstructure of an open-type p o gn sl wharf is rarely comprised of a soft clay Desi Design seabed layer, and are usually comprised of a sandy soil layer or rubble stones. When an Fig. T- 9.4.1 Position of Earth-Retaining Structure behind Slope earth-retaining structure is to be constructed behind a slope, therefore, it is desirable to locate it in the area behind an imaginary slope obtained by drawing a straight line from the foot of the actual slope at the angle a with the horizontal line given by equation (9.4.1) (see Fig. T- 9.4.1). a=f—e (9.4.1) where a： inclination angle of imaginary slope measured from the horizontal line (º) f： internal friction angle of the main material of actual slope (º) e： tan-1kh¢ kh¢： horizontal seismic coefficient in the water -464-


Equation (9.4.1) cannot apply to hard mudstone or bedrock slopes.

9.4.2 Virtual Ground Surface In calculation of lateral resistance and bearing capacity of piles, a virtual ground surface shall be assumed at an appropriate level for each pile. [Technical Notes] When the inclination of the slope is considerably steep, the virtual ground surface for each pile to be used in the calculation of lateral resistance or bearing capacity may be set at an elevation that corresponds to 1/2 of the vertical distance between the surface of the slope at the pile axis and the sea bottom in front of the slope (see Fig. T- 9.4.2).

Fig. T- 9.4.2 Virtual Ground Surface

9.5 Design of Piles 9.5.1 General (Notification Article 69, Clause 1-1) It shall be standard to calculate the lateral resistance, bending moment, and axial force of piles, and the bending moment and shear force of the superstructure with an appropriate method selected based on the structural characteristics of the open-type wharf on vertical piles. [Commentary] (1) Piles used for open-type wharves on vertical piles are usually deployed in the form of groups of piles, and the connection between the piles and coping often takes the form of a rigid joint. In such a case, structural analysis of the pile-supported section can be made by replacing it with a rigid frame or a frame structure (in which the ground is evaluated as an elastic spring), under the assumption that each pile is embedded up to an appropriate depth. (2) For analysis of a single pile subject to lateral forces, the PHRI method or Chang’s method can be used. According to PartⅤ Ⅴ, 4.3.4 Estimation of Pile Behavior Using Analytical Methods, it is recommended to use the former method. However, for the behavior of piles with free ends (such as those used for a pier structure) under the actions of design loads, the “virtual fixed point method based on Chang’s method” may be used because there is almost no difference between the results obtained from the PHRI method and that obtained from Chang’s method.

9.5.2 Coefficient of Horizontal Subgrade Reaction In the calculation of the lateral resistance of piles, it is desirable to obtain the coefficient of horizontal subgrade reaction of the subsoil through lateral loading tests of piles in situ. In case that no tests are conducted, it may be estimated by means of analytical methods. [Commentary] There are some measured data available on the coefficient of horizontal subgrade reaction obtained by the tests in which the lateral loads were applied to piles until they nearly reached to their yield points (as in the case of piles of open-type wharves). Although some of these data have been related to the N-value, the coefficient of horizontal subgrade reaction cannot be estimated accurately from the N-value. Thus, it is desirable to estimate it by means of lateral loading tests in situ (refer to Part Ⅴ, 4.3.3 Estimation of Pile Behavior Using Loading Tests). [Technical Notes] When lateral loading tests of piles are not carried out, the coefficient of horizontal subgrade reaction of the subsoil may be estimated with the method described in Part Ⅴ, 4.3.4 Estimation of Pile Behavior Using Analytical Methods. When using Chang’s method, equation (9.5.1) may be utilized for this purpose. However, some in situ measurement data indicate that the coefficient value of horizontal subgrade reaction of rubble stones is smaller than -465-


the estimate by means of equation (9.5.1) with Chang’s method. Thus, careful examination and judgement are necessary. (9.5.1) kh = 1.5N where kh： coefficient of horizontal subgrade reaction (N/cm3) N： average N-value of the ground down to a depth of about 1/b b： refer to 9.5.3 Virtual Fixed Point

9.5.3 Virtual Fixed Point When analyzing the open-type wharf as a rigid frame structure by assuming a fixed point located below the sea bottom, the virtual fixed points of the piles shall be defined appropriately. [Technical Notes] When performing a rigid frame calculation for an open-type wharf on vertical piles, the virtual fixed points of the piles may be considered to be located at a depth of 1/b below the virtual ground surface. The value of b is calculated by means of equation (9.5.2). 4 kh D (cm-1) (9.5.2) b = --------4EI where kh： horizontal subgrade reaction coefficient (N/cm3) D： diameter or width of the pile (cm) EI： flexural rigidity of the pile (N•cm2)

9.5.4 Member Forces Acting on Individual Piles (Notification Article 69, Clause 1-2 ) It shall be standard to calculate appropriately the horizontal forces acting on the heads of individual piles, pile intervals, displacements of the wharf as a whole and individual piles, pile head moments of individual piles, and axial forces of individual piles by considering the rotation of each block of the pile-supported section and other factors. [Technical Notes] (1) The horizontal force acting on the heads of individual piles may be calculated using the following equations: (a) When it is not necessary to consider the rotation of the deck block: K Hi H i = ---------------H å K Hi

(9.5.3)

i

(b) When it is necessary to consider the rotation of the deck block: When the symmetry axis of the deck block is perpendicular to the face line of the wharf and the direction of action of the horizontal force is parallel to the symmetry axis, the horizontal force may be calculated using equation (9.5.4). K Hi x i K Hi (9.5.4) H i = ---------------H + --------------------- eH 2 hi K K x å Hi å Hi i i

1

βi

Face line

i

where Hi： horizontal force on the i-th pile (kN) KHi： horizontal spring constant of the i-th pile (kN/m) 12EIi K Hi = -----------------------3 1ö æ h + --è i b iø

Center of gravity of

the pile group hi： vertical distance between the i-th pile head and the virtual ground surface (m) Symmetry ax e bi： see 9.5.3 Virtual Fixed Point (m-1) i H EIi： flexural rigidity of the i-th pile (kN•m2) i H： horizontal force acting on the block (kN) e： distance between the block’s symmetry axis and the i-th pile line of action of the horizontal force (m) xi： distance between the block’s symmetry axis and the Fig. T- 9.5.1 Distance between the Center of Gravity of the Pile Group and i-th pile (m)

x

-466-

Individual Piles

x


(2) The displacements of the wharf as a whole and individual piles may be calculated using equations (9.5.5) through (9.5.7). The definitions of the symbols are the same as those described in (1). (a) Horizontal displacement of the pile-supported section d (m) H d = --------------å K Hi

(9.5.5)

i

(b) Rotational angle of the pile-supported section a (º) eH a ≒ --------------------å KHi x i2

(9.5.6)

i

(c) Horizontal displacements of the heads of individual piles Δi (m)

Δ i ≒ d + a xi

　　　　　　(9.5.7)

(3) The pile head moments of individual piles may be calculated using equation (9.5.8). 1 1 M i = --- æ h i + ----ö H i 2è b iø

(9.5.8)

where Mi： pile head moment of the i-th pile (kN•m) (4) The axial force of individual piles may be calculated using equation (9.5.9). (9.5.9) Pi = hPi + vPi where Pi ： axial force of the i-th pile (kN) hPi： axial force of the i-th pile due to horizontal force (kN) vPi： axial force of the i-th pile due to vertical load (kN) M i 1 , i + M i, i 1 M i, i +1 M i +1 , i hPi = S i, i 1 + S i, i +1 = ------------------------------------------------------------------------------------l S i, i 1 ： shear force acting on the head of the i-th pile due to the horizontal force of the side beam of the (i－1)-th pile (kN) S i, i +1 ： shear force acting on the head of the i-th pile due to the horizontal force of the side beam of the (i＋1)-th pile (kN) M i 1 , i： bending moment acting on the head of the (i－1)-th pile due to the horizontal force of the side beam of the i-th pile (kN•m) M i, i 1 ： bending moment acting on the head of the i-th pile due to the horizontal force of the side beam of the (i－1)-th pile (kN•m) M i, i +1 ： bending moment acting on the head of the i-th pile due to the horizontal force of the side beam of the (i＋1)-th pile (kN•m) M i +1 , i： bending moment acting on the head of the (i＋1)-th pile due to the horizontal force of the side beam of the i-th pile (kN•m) l： beam span (pile interval) (m)

678

The values of M i, i 1 and M i, i +1 should be evaluated by equation (9.5.10) for the intermediate supports, and they should be taken as equal to the pile head moment Mi for the end supports. (see Fig. T- 9.5.2). M i, i 1 = aM i (9.5.10) M i, i +1 = bM i The values of a and b may be taken as follows: 2-span beam: a = 0.6, b = 0.6 3-span beam: 2nd pile head; a = 0.5, b = 0.7 3rd pile head; a = 0.7, b = 0.5

Fig. T- 9.5.2 Pile Head Moments -467-


9.5.5 Cross-Sectional Stresses of Piles Appropriate methods shall be used in the calculation of the cross-sectional stresses of vertical piles supporting the deck. [Technical Notes] (1) The cross-sectional stresses of vertical piles should be calculated using equation (9.5.11), and the allowable stresses for piles should be calculated as those of the members subjected to the axial force and bending moment by meanes of Table T- 2.3.2 of Part Ⅲ, 2.3.3 Steel Piles and Steel Pipe Sheet Piles. P M s c = --- , s b = ----A Z where s c： stress due to the axial force of pile (N/mm2) s b： stress due to the bending moment of pile (N/mm2) A： cross-sectional area of pile (mm2) P： axial force of pile (N) Z： section modulus of pile (mm3) M： bending moment of pile (N•mm)

(9.5.11)

(2) Shear stress is normally negligible in vertical piles for open-type wharf. Therefore, the examination of shear stress may be omitted for the piles, unless the loading conditions are special.

9.5.6 Examination of Embedded Length for Bearing Capacity (Notification Article 69, Clause 1-3 ) It shall be standard to calculate appropriately the embedded length of each vertical pile by considering the characteristics of the ground and the pile lateral resistance in accordance with Part Ⅴ, 4.1 Allowable Axial Bearing Capacity of Piles. In this case, the bearing capacity of a pile driven into a sloped ground can be calculated with the assumption that the soil layers below the virtual ground surface are the effective supporting layers. 9.5.7 Examination of Embedded Length for Lateral Resistance The embedded length of each vertical pile shall be determined appropriately in accordance with the method of analysis of the pile lateral resistance. [Technical Notes] The embedded lengths of vertical piles are generally taken at 3/b below the virtual ground surface based on the results of pile lateral resistance analyses. The value of b should be calculated in accordance with 9.5.3 Virtual Fixed Point.

9.5.8 Examination of Pile Joints (1) When a pile joint is to be used in a pile, the pile shall be designed to be safe against the impact stress generated in the joint during driving, in addition to the external forces described in 9.3.1 Design External Forces. (2) The location of pile joint shall be determined carefully in consideration of construction conditions so that no excessive stress is generated in the pile. [Technical Notes] For the method to join piles, refer to Part Ⅴ, 4.5.3 Joints.

9.5.9 Change of Plate Thickness or Material of Steel Pipe Pile Any change relating to plate thickness or material of a steel pipe pile shall be made in accordance with Part Ⅴ，4.5.4 Change of Plate Thickness or Material of Steel Pipe Piles. [Technical Notes] (1) In many cases, the plate thickness and/or material of steel pipe piles of ordinary open-type wharves are changed at a depth of 2/b to 3/b below the virtual ground surface. (2) The strengths of joints and sections of thickness change should be examined carefully because there are some examples in which piles of open-type wharves buckled at these sections due to ground deformation and others during an earthquake in a deep ground where no bending stresses are generated under normal load conditions. -468-


9.6 Examination of Earthquake-Resistant Performance Earthquake resistance of open-type wharves shall be examined to ensure that they satisfy the performance requirement in the expected seismic motion environment. [Commentary] (1) As described in 9.1 Principle of Design, it should be checked and confirmed that the pile-supported section of an open-type wharf will satisfy the performance requirement during an earthquake using an appropriate method. In this examination, the performance of open-type wharves which are to be constructed as ordinary mooring facilities should be examined against the Level 1 ground motion as described in Part Ⅱ , Chapter 12 Earthquakes and Seismic Force, and the performance of open-type wharves to be constructed as high seismic resistant facilities should be examined against both the Level 1 and the Level 2 ground motions. (2) In those cases where the natural period of a pile-supported section and that of the ground are close to each other or the natural period of the seismic motion is close to these two, a kind of resonance may occur and cause a larger response of pile-supported section than that expected to occur under ordinary conditions. In such a case, it is necessary to take appropriate measures, such as to adopt a different structural type of the pile-supported section for shifting its natural period and to performe a subsoil improvement work for changing the natural period of the ground. (3) The structure of an access bridge connecting the pile-supported section and the earth-retaining revetment should be determined in consideration of the following: (a) When an access bridge is securely fixed, an unexpectable excessive horizontal force may act on the superstructure of pile-supported section in the mode of earthquake motions in which the distance between the pile-supported section and the earth-retaining section is shortened. (b) The structure of an access bridge should be such that the access bridge will be able to adequately accommodate the expected displacement of the pile-supported section and prevent itself from falling off in the mode of eathquake motions in which the distance between the pile-supported section and the earth-retaining section is elongated. [Technical Notes] (1) It is recommended that seismic design of open-type wharves be basically conducted in accordance with the sequence depicted in Fig. T- 9.6.1. Selection of cross section for examination of earthquake-resistant performance

Determination of seismic coefficient to be used in performance examination against Level 1 seismic ground motion

Simplified method Elasto-plastic analysis Dynamic response analysis

Examination of the earthquake-resistant performance against Level 1 ground motion Type of failure Load carrying capacity Maximum and residual displacement Bearing capacity of pile

Heigh seismic resistant wharf

Ordinary wharf

Determination of the seismic coefficient to be used in performance examination against Level 2 ground motion

Simplified method Elasto-plastic analysis Dynamic response analysis

Examination of the earthquake-resistant performance against Level 2 ground motion Type of failure Load carrying capacity Maximum and residual displacement Bearing capacity of pile

End

Fig. T- 9.6.1 Flowchart of Examination of Earthquake-Resistant Performance of Open-Type Wharf -469-


(2) When the cargo handling equipment such as container cranes is to be installed, it is desirable to examine the overall earthquake-resistant performance that takes into consideration the combined vibrations of the pilesupported section and the equipment as described in 9.1 Principle of Design [Technical Notes] (7), in addition to performing the examination of the earthquake-resistant performance of the pile-supported section itsef. (3) The earthquake-resistant performance levels required of pile-supported sections differ depending on the importance of facility and other criteria. However, the damage level, maximum displacement, residual displacement, etc. may be used as typical performance indexes. In determining the performance level, it is necessary to give careful consideration to the easiness of the repair and/or restoration work of damaged pilesupported section. (4) The earthquake-resistant performance examination method described in this section is used for open-type wharves of ordinary dimensions on vertical steel pipe piles. Thus, it may be applied to structures which exhibit similar behaviors during earthquakes. However, when the structure is to be constructed on the horizontal ground without a slope or when batter piles are to be used, it is necessary to carefully consider its applicability.

9.6.1 Assumption of Cross Section for Earthquake-Resistant Performance Examination The cross section of an open-type wharf for examination of the earthquake-resistant performance shall be selected in an appropriate manner. [Technical Notes] (1) In principle, a cross section designed for external loads other than earthquake loads should be used for examination of the earthquake-resistant performance of pile-supported sections as described in 9.1 Principle of Design [Technical Notes] (1). In the areas in which relatively strong earthquakes may occur however, it is desirable to use data from similar existing structures and facilities in the neighborhood as reference data for the selection of cross section, because the final cross section is likely to be determined by earthquake-resistant performance. (2) When the result of earthquake-resistant performance examination shows that the selected cross section could not achieve the required performance during an earthquake, modification of the cross section should be performed.

9.6.2 Examination Method of Earthquake-Resistant Performance Examination of the earthquake-resistant performance of an open-type wharf shall be performed for the earthquake forces considered in the design by appropriately modeling the ground and the pile-supported section. [Commentary] (1) Because examination of the earthquake-resistant performance of an open-type wharf is made in terms of the type of failure and/or damage that occur during an earthquake, the load carrying capacity, the maximum and residual displacement, and the bearing capacity of piles, it is necessary to use the analysis models of the ground and the pile-supported section that enable appropriate evaluation of these characteristics. In particular, when a strong earthquake force such as the Level 2 ground motion is expected to act on the pier, it is necessary to model the nonlinearity of the structural elements appropriately. (2) In examination of earthquake-resistant performance, the earthquake forces may be considered in the horizontal direction only, unless the wharf requires special examination to the vertical direction for the structural reasons. [Technical Notes] (1) Although many studies are being conducted on the dynamic behavior and earthquake-resistant performance of open-type wharves during earthquakes, there are still many characteristics that are yet to be made clear. In general, open-type wharves are considered to have large load carrying capacity against strong earthquakes (such as those with the Level 2 ground motion) because they can withstand large deformations. However, because the ground motion and deformation of the structural members of a wharf may exceed their elastic limits under strong seismic motions and they may be in the respective plastic zones, it is necessary to bear in mind that the design of wharf using the virtual fixed point method described in 9.5 Design of Piles may result in the dangerous side of safety. It is recommended that the examination of the earthquake-resistant performance of open-type wharf be performed using either (a) simplified method, (b) elasto-plastic analysis method, or (c) nonlinear dynamic analysis method described below. (a) Simplified method This method evaluates the load carrying capacity of open-type wharf by summing up the strength of the steel pipe piles, while assuming that the wharf superstructure is a rigid body. This method is applicable to wharves on vertical piles in which the rigidity ratios between the pile rows do not differ significantly from each other. -470-


Examination is performed ① to confirm that the inertial force is smaller than the load carring capacity and the bearing force of the pile is lower than the ultimate bearing capacity and ② to check the levels of maximum and residual displacement, by considering an earthquake energy absorption capacity corresponding to the allowable displacement ductility factor for the pile. (b) Elasto-plastic analysis method This method evaluates the load carrying capacity of open-type wharf by regarding the pile-supported section as a “pile - rigid frame” structure model and by taking into consideration the lateral resistance of the subsoil and the elasto-plastic behavior of the steel pipe piles and superstructure. It is thought that the influence of the section force redistribution effect is significant when the structure of the wharf is complex or batter piles are used together with vertical piles. In such cases, detailed analysis using the elasto-plastic method or other relevant method needs to be employed, because the simplified method may overestimate the load carrying capacity. In the elasto-plastic method, the type of failure of the wharf is first analyzed based on examination of the damage events that will occur in the wharf (e.g., generation of plastic hinges, yielding of the ground spring, damage to the superstructure, etc.), and then earthquake-resistant performance is examined accordingly. In addition, the magnitudes of the maximum and residual displacements are examined. (c) Nonlinear dynamic analysis method This method examines the strength of open-type wharf by modeling the whole structure and performing nonlinear dynamic analysis using the finite element analysis, etc. When the overall structure including the revetment structure behind the pile-support section is complex, or when large deformation of the ground is expected, examination is performed by modeling the entire system including the wharf and ground. For nonlinear dynamic analysis, the same technique used in elasto-plastic analysis may be mobilized. (2) Examination by means of the load carrying capacity is a method to allow occurrence of a certain damage and/or failure in some parts of the wharf. Therefore, it is necessary to specify the range and degree of tolerable damage according to the functions required of the wharf after an earthquake, importance of the wharf, easiness of repair and restoration work, etc.

9.6.3 Determination of Seismic Motion for Examination of Earthquake-Resistant Performance In principle, the seismic motion that is used in the examination of earthquake-resistant performance shall be obtained through dynamic analysis of the ground. [Technical Notes] (1) When the nonlinear dynamic analysis described in 9.6.2 Examination Method of Earthquake-Resistant Performance is not performed, the earthquake forces to be inputted to the pile-supported section should be determined by means of the modified seismic coefficient method. The seismic coefficient should be determined based on the natural period of the pile-supported section using the linear acceleration response spectrum derived from the response acceleration at a depth of 1/b below the virtual ground surface that is obtained from dynamic analysis of the ground at the construction site. Figure T- 9.6.2 shows a flowchart of the determination of seismic coefficient for examination of earthquake-resistant performance.

Dynamic analysis of ground Determination of maximum base acceleration

Natural period of the pier

Preparation of ground response analysis model

Cross section for earthquake-resistant performance examination

Calculation of the natural period of the ground Ground response analysis (SHAKE etc.) Response acceleration at a depth of 1/

"Pile-Rigid Frame" model etc. below the sea bottom

Calculation of acceleration response spectrum

Calculation of the natural period of the pier

Determination of the seismic coefficient for earthquake-resistant performance examination

Fig. T- 9.6.2 Flowchart of Determination of Seismic Coefficient to Be Used in the Examination of Earthquake-Resistant Performance

-471-


(2) In the determination of the strength level of the Level 1 ground motion, the common practice is to assume a seismic motion with a return period of 75 years. The expected value of bedrock acceleration is that of ground motion listed in Table T- 12.4.1 of PartⅡ Ⅱ, 12.4 Design Seismic Coefficient according to the regional classification of the pier construction site. The design seismic coefficient for examination of earthquakeresistant performance is obtained by dividing the response acceleration of the ground corresponding to the natural period of pile-supported section by the acceleration of gravity. The response acceleration of the ground is derived from the acceleration response spectrum at a depth of 1/b below the sea bottom (b is calculated using equation (9.5.2) of 9.5.3 Virtual Fixed Point) that is computed from the input of the expected value of bedrock acceleration by means of the multiple reflection model “SHAKE” described in Part Ⅱ, 12.5 Seismic Response Analysis. For the dissipation damping constant to be employed in the calculation of the acceleration response spectrum, the values listed in Table T- 9.6.1 may be used according to the type of ground. Table T- 9.6.1 Dissipation Damping Constant Natural period of the ground (Tg)

Damping constant

Shorter than 0.1 s

0.1

0.1 s or longer and shorter than 0.5 s

0.2

0.5 s or longer

0.3

(3) When the dynamic analysis of ground described in (4) is not performed, the standard seismic coefficient values shown in Fig. T- 9.6.3 (that were obtained from results of analysis 1) of open-typed wharves having standard dimensions) may be used. The seismic coefficient can be obtained from the figure for different regional classifications (region categories A to E based on the regional seismic coefficients defined in Part Ⅱ , 12.4 Design Seismic Coefficient) and the natural periods of the ground and the pile-supported section. (4) The seismic coefficient for the Level 2 ground motion is obtained by determining the expected value of maximum bedrock acceleration using the method described in Part Ⅱ, 12.4 Design Seismic Coefficient and using the technique described in (2) above. In principle, the seismic coefficient of the Level 2 ground motion should be obtained by performing dynamic analysis of the ground, because no reliable methods for determining the standard seismic coefficient for the Level 2 ground motion have yet to be developed. (5) The natural period of ground described in (1) and (2) may be calculated by means of equation (9.6.1) using the shear wave speed and the layer thickness of each layer for the soil layers between the upper surface of armor stone and the bedrock surface for engineering analysis directly below the pile-supported section. Hi T g = 4S ------V si where Tg： natural period of the ground (s) Hi： thickness of the i-th layer (m) Vsi： shear wave velocity in the i-th layer (m/s)

(9.6.1)

(6) The natural period of the pile-supported section may be calculated with equation (9.6.2) using (a) the horizontal spring constant of the pile-supported section as a rigid frame structure whose virtual fixed point (see 9.5.3 Virtual Fixed Point) is located at a depth of 1/b below the sea bottom and (b) the sum of the deadweight of the pile-supported section and the surcharge acting on it during an earthquake. W (9.6.2) T s = 2p --------------g SK Hi where Ts： natural period of pile-supported section (s) W： sum of deadweight acting on one pile row and surcharge during an earthquake (kN) g： acceleration of gravity (m/s2) KHi： horizontal spring constant of the i-th pile (kN/m) (see 9.5.4 Member Forces Acting on Individual Piles) To obtain the value of KHi, it is necessary to evaluate b by appropriately assuming the horizontal subgrade reaction coefficient of the subsoil. Equation (9.5.1) described in 9.5.2 Cofficient of Horizontal Subgrade Reaction is an equation to calculate a static subgrade reaction coefficient, and it is not appropriate to apply this equation to dynamic analysis. As the number of data and research results on dynamic subgrade reaction coefficient is small, it is recommended to use a value about twice as large as the coefficient obtained by equation (9.5.1) to calculate the natural period of a pile-supported section in accordance with the “Specifications for Highway Bridges”. There is also a report which says that the natural period of a pile-supported section can be calculated more accurately when the virtual ground surface is not considered. 1)

-472-


When the bending moment generated in the pile exceeds the fully plastic state moment (i.e., generation of plastic hinge), the fully plastic state rigidity of the pile (Mp /φp in Fig. T- 9.6.6) should be used for the flexural rigidity of a pile. 1.00

Region A

1.00

Region B

Seismic coefficient

Seismic coefficient

2

3 1

0.10

0.01 0.1

1.0

2

3 1

0.10

0.01 0.1

10

1.0

Period (units: s) Region C

1.00

Seismic coefficient

Seismic coefficient

1.00

2 3 1

0.10

0.01 0.1

1.0

10

Region D

1

2 3

0.10

0.01 0.1

1.0

Period (units: s)

Seismic coefficient

1.00

10

Period (units: s)

10

Period (units: s)

Region E

1

2 3

0.10

Legend

0.01 0.1

Tg

0.1

0.1

S

0.5

S

S

Tg Tg

0.5

S

Tg ; natural period of the ground calculated 1.0

with equation (9.6.1) (units: s)

10

Period (units: s)

Fig. T- 9.6.3 Standard Seismic Coefficient for the Level 1 Ground Motion

9.6.4 Examination of Load Carrying Capacity Using Simplified Method In the examination of the load carrying capacity of an open-type wharf using the simplified method, the pile-supported section shall be represented with a frame structure model and the horizontal displacement ductility factor of the pile-supported section shall be used. [Technical Notes] (1) In the simplified method described in 9.6.2 Examination Method of Earthquake-Resistant Performance, the examination of load carrying capacity is fundamentally carried out by making assumptions based on the -473-


“Newmark’s Equal-Energy Principle” shown in Fig. T- 9.6.4 and using an equivalent linearized system with one degree of freedom having the same rigidity as the first rigidity of the “wharf - ground” structural system. The basic equation for examination is as follows: Ra ≧ khW (9.6.3) where Ra： load carrying capacity of pile-supported section during an earthquake (kN) kh： seismic coefficient derived from linear response (using the method described in 9.6.3 Determination of Seismic Motion for Examination of Earthquake-Resistant Performance) W： deadweight of pile-supported section and surcharge acting during an earthquake (kN) Horizontal force P

Pe: horizontal elastic response force Py: horizontal force corresponding to the elastic limit

d p: horizontal elasto-plastic response displacement d e: horizontal elastic response displacement d y: horizontal displacement corresponding to the elastic limit

Horizontal displacement

Fig. T- 9.6.4 Elasto-Plastic Horizontal Response Displacement of Pier Superstructure

The load carrying capacity of the pile-supported section during an earthquake, Ra is calculated by equation (9.6.4): (9.6.4) R a = 2m a 1 + q ( m a 1 ) 2 P y where ma： allowable displacement ductlily factor q： ratio of the secondary gradient (gradient of the line CD in Fig. T- 9.6.4) to the primary gradient (gradient of the line OC) in the load - horizontal displacement graph. However, when the allowable displacement ductility factor is determined by means of the method described in (3), q is assigned the value of 0. Py： horizontal force corresponding to the elastic limit (kN) (2) The allowable displacement ductility factor used in (1) is expressed as the ratio of the allowable horizontal displacement to the horizontal displacement at the elastic limit of superstructure. The elastic limit is a point at which the horizontal displacement starts to increase sharply in the horizontal load - horizontal superstructure displacement relationship of the wharf. This may normally be considered as the state that the pile head bending moment reaches the fully plastic state moment (see equation (9.6.7)) in a majority of the piles in the rows normal to the face line of the wharf. (3) The allowable displacement ductility factor should essentially be determined in consideration of the progress of the failure and/or damage of the wharf. The displacement ductilily capacity of an open-type wharf as a whole is governed by the displacement ductility capacity of the steel pipe piles. The standard allowable displacement ductility factor (that is based on the Py of equation (9.6.4)) may normally be determined using the values listed in Table T- 9.6.2 for the Level 1 seismic motion and those obtained from equation (9.6.5) for the Level 2 seismic motions. The recommendations are based on the result of examination of the load - deformation relationship by a frame structure model of the steel pipe piles normally used for the open-type wharf and that of a laboratory loading test on a structural model. The importance levels listed in Table T- 9.6.2 are those defined in Table - 12.4.4 of Part Ⅱ, 12.4 Design Seismic Coefficient. ma = 1.25 + 62.5 (t / D) ≦ 2.5 where ma： allowable displacement ductility factor for the Level 2 seismic motion t： thickness of steel pipe pile (mm) D： diameter of steel pipe pile (mm)

-474-

(9.6.5)


Table T- 9.6.2 Standard Allowable Displacement Ductility Factor for the Level 1 Seismic Motion Importance level

Allowable displacement ductility factor (ma)

Special

1.0

Class A

1.3

Class B

1.6

Class C

2.3

(4) The horizontal force Py corresponding to the elastic limit can be calculated by assuming that all the pile head displacements are equal in the “pile - rigid frame” structure model. As this involves complex calculations however, Py may normally be calculated in the simplified method by using equation (9.6.6) which was derived based on the result of the examination of the load - deformation relationship in the “pile - rigid frame” model using the dimensions of the steel pipe piles generally used in the open-type wharf. Py= 0.82 Puall P uall =

2M

pi S --------------------

(9.6.6)

1ö æ h + --è i b iø

where Mpi： fully plastic state moment of the i-th pile calculated using equation (9.6.7) to appear later (kN•m) hi, bi： refer to 9.5.4 Member Forces Acting on Individual Piles [Technical Notes] (1). The term Puall indicates the horizontal load level at which the bending moments of all the piles of the wharf reach the fully plastic state moments both at the pile heads and underground virtual fixed points. This is the upper limit load that the wharf can withstand from the viewpoint of structural analysis. The equation that Py equals to 82% of Puall is based on the results of parametric analysis performed on open-type wharves of ordinary sizes. (5) The residual displacement can be obtained by drawing a straight line with the inclination of the initial gradient (initial rigidity) from the point of the maximum response displacement in the horizontal load - displacement relationship and by reading off the displacement at its point of crossing with the abscissa.

9.6.5 Examination of Load Carrying Capacity Using Elasto-Plastic Analysis In the examination of the load carrying capacity of an open-type wharf using elasto-plastic analysis, the pile-supported section shall be represented by a frame structure model, and the nonlinearity of steel pipe piles, concrete superstructure, and the ground shall be taken into consideration. [Technical Notes] (1) In the examination of the earthquake-resistant performance of an open-type wharf as an ordinary mooring equipment, it is recommended to define a state before the generation of a plastic hinge in the ground (i.e., fully plastic state moment) as the ultimate state, because it is difficult to discover buckling of a steel pipe pile in the underground and to repair it. In addition, examination of maximum displacement and residual displacement should also be made by using the allowable displacement ductility factor listed in Table T- 9.6.2 as a reference. The section forces generated in the reinforced concrete superstructure and the pile bearing capacity should also be examined. (2) In the examination with respect to the Level 1 ground motion of open-type wharves constructed as high seismic resistant facilities, examination should be made as to whether they are within the elastic behavior limits. The examination of this type of wharves with respect to the Level 2 seismic motion may be made in the same manner as that of (1). (3) The load carrying capacity during an earthquake is calculated by means of equation (9.6.4) using the horizontal displacement of the superstructure that corresponds to the tolerable damage level and the elastic limit load obtained from the elasto-plastic analysis. In this calculation, the ratio q of the secondary to the primary gradients in the load - horizontal displacement relationship and the displacement ductility factor μ corresponding to the displacement of the tolerable damage level may be defined as shown in Fig. T- 9.6.5.

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Load

Tolerable damage level -

Displacement

Fig. T- 9.6.5 Examination of Load Carrying Capacity Using Elasto-Plastic Method

Bilinear model

Trilinear model

Curvature Fig. T- 9.6.6 Bending Moment - Curvature Relationship of Steel Pipe Pile

(4) Analysis using the elasto-plastic method is usually made in the following manner:

N M y = æ f y ----ö Z e è Aø

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(a) Analysis is performed by reducing the flexural rigidity of a steel pipe pile depending on the magnitudes of the axial force and bending moment that are generated in the pile. The relationship between the bending moment and curvature of a steel pipe pile is represented by a trilinear model that has an upper limit value equal to the fully plastic state moment (see Fig. T- 9.6.6). A bilinear model (shown with the broken line in Fig. T- 9.6.6), which makes the calculation simpler, may be used in place of the trilinear model in the cases of open-type wharves of ordinary sizes, as there is no significant difference in calculation results between the bilinear and trilinear models for those wharves. The fully plastic state moment Mp, yielding moment My, yield curvature φy, and fully plastic state curvature φp may be calculated by means of equation (9.6.7). ap M p = M p0 cos æ -------ö è 2ø (9.6.7)

Mp My f y = ------- , f p = ------- f y My EI where Mp0： fully plastic state moment of steel pipe pile when no axial force is acting (N•mm); Mp0 = Zp fy Zp： plastic sectional modulus of steel pipe pile (mm3); Z p = ( 4 ¤ 3 ) [ r 3 ( r t ) 3 ] r： radius of steel pipe pile (mm) t： thickness of steel pipe pile (mm) a： ratio of the acting axial force N to the yield axial force N0 (N0 = Afy) when no bending moment is acting A： cross-sectional area of steel pipe pile (mm2) Ze： sectional modulus of steel pipe pile in the elastic domain (mm3) EI： flexural rigidity of steel pipe pile (N•mm2) fy： design yield strength of steel pipe pile (N/mm2) (b) The element characteristics of reinforced concrete or prestressed concrete superstructure should be obtained from an elasto-plastic model that is defined by the bending moment - curvature relationship shown in Fig. T9.6.7.

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The moments Mc, My, and Mu in the figure represent the cracking moment, reinfocing bar yield moment, and ultimate bending moment, respectively. As a standard procedure, the ultimate bending moment etc. should be calculated using the limit state design method described in Part Ⅲ, Chapter 3 Concrete. Since the load carrying capacity decreases sharply when a shear failure occurs in the concrete structure, it is necessary to design it in such a way that a bending failure occurs before a shearing failure and a sufficient degree of safety is secured against the shear failure.

Curvature Fig. T- 9.6.7 “Bending Moment - Curvature” Relationship of Concrete Superstructure

(c) The pushing-in bearing capacity of pile should be calculated by means of a bilinear model whose initial gradient and upper limit are equal to the axial subgrade reaction coefficient of the pile and the pile pushing-in upper limit PNU, respectively. The latter is calculated by equation (9.6.8). (9.6.8) PNU = min (Ru, Rpu) where Ru： ultimate bearing capacity of the pile that is calculated from the ground conditions (kN) (refer to Part Ⅴ, 4.1.5 Estimation of Ultimate Axial Bearing Capacity by Static Bearing Capacity Formulas.) Rpu： upper limit of pushing-in bearing capacity that is calculated from the axial compressive load corresponding to the yield point of steel pipe pile (kN) (d) The pulling-out bearing capacity of pile should be evaluated with a bilinear-type elasto-plastic model. The upper limit of pulling-out bearing capacity should be calculated in accordance with Part Ⅴ, 4.2.3 Maximum Pulling Resistance of Single Pile. (e) The horizontal subgrade reaction may normally be represented by a bilinear-type elasto-plastic model. (5) In many cases, the ultimate state of flexural deformation performance of a steel pipe pile is determined by the plastic buckling and/or long column buckling of the steel pipe. The strain (emax) at local buckling in a steel pipe pile is given by the following equation:

emax = 0.44 t / D where t： thickness of steel pipe pile (mm) D： diameter of steel pipe pile (mm)

(9.6.9)

When it is necessary to prevent the local buckling of steel pipe piles, it is necessary to perform examination by assuming that generation of the strain calculated by equation (9.6.9) is the ultimate state. The magnitude of the bending moment that produces the strain at local buckling is almost equal to the value of the fully plastic state bending moment calculated by equation (9.6.7) in steel pipe piles of ordinary dimensions that are used for opentype wharves.

9.7 Design of Earth-Retaining Section (Notification Article 69, Clause 2 ) The examination of the structural stability of the earth-retaining section of open-type wharf on vertical piles shall be made in accordance with the relevant structural stability guidelines, such as Chapter 4 Gravity Type Quaywalls and Chapter 5 Sheet Pile Quaywalls in consideration of its structural type. [Technical Notes] Whenever possible, the superstructure and the earth-retaining section of an open-type wharf should be connected by means of a simply supported slab having clearances or buffer material provided on the both sides of slab, in order to prevent the forces acting on the earth-retaining section from being transmitted to the superstructure and to be able to accommodate the differential settlements between the superstructure and the eath-retaining section. The superstructure and the earth-retaining section are usually constructed in such a way that the minimum clearance roughly equal to the maximum horizontal displacement of the open-type wharf be secured between them.

9.8 Examination of Stability against Circular Slip In principle, the stability of the earth-retaining section of open-type wharf on vertical piles against circular slip shall be examined in accordance with Part Ⅴ, 6.2.1 Stability Analysis Using Circular Slip Surface Method.

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9.9 Detailed Design 9.9.1 Load Combinations for Superstructure Design The loads to be used for superstructure design shall be an appropriate combination of the deadweights and surcharges acting on the deck slab, access bridges, and beams, in order to secure the safety of the opentype wharf. [Technical Notes] In the design of superstructure, the following load combinations should be used and the superstructure should be safe against their actions. (1) Deck Slab and Access Bridge ① deadweight and static load ② deadweight and live load ③ deadweight and uplift (2) Beam ① deadweight and static load ② deadweight and live load ③ deadweight, moment generated at pile head due to the action of horizontal force, and static load when the horizontal force is acting ④ load due to elastic settlement of the piles

9.9.2 Calculation of Reinforcing Bar Arrangement of Superstructure In principle, calculation of reinforcing bar arrangement of reinforced or prestressed concrete superstructures shall be made based on the limit state design method. [Commentary] (1) Sufficient reinforcement should be provided against shear failure of the superstructure so that shear failure will not precede bending failure. (2) Since the superstructure is exposed to severe deterioration environment than other structures, it is necessary to take appropriate measures to satisfy the requirements for structural details such as the covering depth and the reinforcement required for the control of crack width under the serviceability limit state, so that a sufficient level of durability is retained.

9.9.3 Design of Pile Head Pile heads shall be designed to be safe against the various stresses generated there. [Commentary] (1) The design of pile heads should be performed in accordance with PartⅤ Ⅴ, 4.5.2 Design of Joints between Piles and Superstructure. (2) Because the premise of the seismic design of open-type wharf is such that the joints between the superstructure and piles function as fully rigid joints, detailed design should be made to satisfy this condition during an earthquake as well. (3) When welding is applied in construction, examination should be made to confirm that the weld connections between reinforcing bars and plates as well as those between those steel piles and plates are strong enough. [Technical Notes] (1) It is recommended that the method of joining the pile heads and the superstructure and the embedment length of piles in the superstructure be determined through loading tests or other relevant measures, because the bond strength characteristics of a joint between a steel pipe and concrete have not fully been clarified yet. When it is not possible to conduct loading tests, the embedment length may be determined using the methods described in paragraphs (2) and (3) below. It should be borne in mind that when the fixed condition (fully rigid joint) defined in the design or earthquake-resistant performance examination is not satisfied in the pile head section, the magnitude of horizontal displacement of the superstructure tends to become larger than that estimated in the design.

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(2) When the axial forces are to be transmitted from the beams to the piles through the bonding between the circumference of pile heads and the concrete, the embedment length may be determined by means of the following methods in consideration of the cross-sectional shape of piles. Because the thickness of the concrete above the pile heads is usually small in open-type wharf, the embedment length of pile is calculated in accordance with equation (9.9.1) on the assumption that there is no punching shear of concrete and the loads are transmitted from the beams to piles only through the bonding between the pile circumference and concrete without assistance of shear strengh of concrete against punching. P (9.9.1) l ≧ -------------g b y f bod where l： embedment length (mm) P： calculated vasue of axial force acting on pile in design (N) 　 y： outer perimeter of the cross section of pile (mm) fbod： design bond strength between the pile and concrete (N/mm2) g b： element factor (it may be assumed as g b = 1.0) When a design value fbod for the bond strength between the pile and concrete cannot be obtained from experiments etc., the equation fbod = 0.11 fck' 2/3 /gc may be used, because the main reinforcing bars of the beam are usually fixed to the pile heads and this makes the bond strength between the pile and concrete roughly equal to that between the round bars and concrete. The terms fck' and gc represent the characteristic value of the compressive strength of concrete (N/mm2) and the material factor of concrete (it may be assumed that gc = 1.3), respectively. (3) The embedment length required for the transmission of the pile head moment should be determined in accordance with Part Ⅴ, 4.5.2 Design of Joints between Piles and Superstructure. This determination should theoretically be made against the combined stress of the bearing stress of the pile owing to the horizontal force and that owing to the bending moment, but the bearing stress owing to the horizontal force is normally negligibly small and thus it may be ignored. It is advisable not to make the concrete slab width B of the part adjoining the pile too small (see Fig. T- 9.9.1). When this width is insufficient, concrete will scale off during the punching shear state and concrete will be unable to bear the bearing stress.

Fig. T- 9.9.1 Bearing Stress Generated in Pile Head

[Reference] 1) Hiroshi YOKOTA, Naoto TAKEHANA, Kenichiro MINAMI, Kunio TAKAHASHI, Noriyuki KAWABATA: “Consideration of design seismic coefficients of an open type wharf based on dynamic response analyses ”, Rept. of PHRI, Vol. 37, No. 2, 1998, pp. 75-130 (in Japanese).

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Chapter 10 Open-Type Wharves on Coupled Raking Piles 10.1 Principle of Design The provisions in this chapter shall be applied to open-type wharves with coupled raking piles that bear the horizontal force acting on the wharf. [Commentary] An open-type wharf on coupled raking piles is a wharf structure that resists the horizontal force acting on the wharf (including the earthquake force, fender reaction force, and tractive force) with coupled raking piles. Therefore, this type of wharves must be constructed on the ground that yields the sufficient bearing capacity for coupled raking piles. Because the coupled raking piles are so layed out to resist the horizontal forces in the direction normal to the face line of the wharf, the horizontal displacement in that direction is smaller than that of open-type wharves on vertical piles. Coupled raking piles are seldom layed out to resist the horizontal forces in the direction of wharf face line. Therefore, it is desirable to examine the strength of the wharf against the horizontal force parrallel to the face line in the manner same as the examination for open-type wharves on vertical piles. The examination of the strength of wharf during an earthquake should be made using the seismic coefficient method, but when earthquake-resistant performance needs to be examined, the nonlinear dynamic analysis method or the elasto-plastic analysis method described in 9.6.2 Examination Method of Earthquake-Resistant Performance [Technical Notes] (1) should be used. Otherwise open-type wharves on coupled raking piles may basically be considered the same as open-type wharves on vertical piles. [Technical Notes] (1) Figure T- 10.1.1 shows an example of open-type wharf on coupled raking piles.

Prestressed Concrete girder Water supply pipe

LWL Stone fill

s

ne

sto or

m

Ar

on

ti da

les

bb

ru

un

Fo

Replaced quarry refuse

Steel pipe pile

Steel pipe pile

Steel pipe pile

(Units: m)

Fig. T- 10.1.1 Example of Open-Type Wharf on Coupled Raking Piles

(2) When coupled raking piles are not layed out in a manner to resist the forces in the direction of wharf face line, the wharf resists them with the lateral resistance of piles. In this case, the calculation of the embedded length of piles must be made by examining their lateral resistance. (3) The uplift forces acting on the superstructure and access bridges should be calculated in accordance with PartⅡ, 5.5.1 Uplift Acting on Horizontal Plate near the Still Water Level.

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10.2 Layout and Dimensions 10.2.1 Size of Deck Block and Layout of Piles The size of a deck block and the layout of piles of open-type wharf on coupled raking piles shall be determined in accordance with 9.2.1 Size of Deck Block and Layout of Piles. [Technical Notes] The distance between piles or between centers of coupled raking piles is usually set to 4 to 6 m in consideration of loading conditions and construction work. It is desirable to use a large raking angle of coupled piles from the viewpoint of securing resistance against horizontal force, but in many cases an inclination of 1: 0.33 to 1: 0.2 is used because of constraints of keeping minimum distances from other piles and construction work-related constraints such as the capacity of pile driving equipment available.

10.2.2 Dimensions of Supersutructure The dimensions of the superstructure of open-type wharf on coupled raking piles shall be determined in accordance with 9.2.2 Dimensions of Superstructure. [Technical Notes] The beam width of the superstructure of open-type wharf on coupled raking piles varies depending on the method of joining the heads of coupled piles, but it ordinarily becomes wider than that on vertical piles.

10.2.3 Arrangement of Fenders and Bollards The arrangement of fenders and bollards shall be determined in accordance with 9.2.3 Arrangement of Fenders and Bollards.

10.3 External Forces Acting on Open-Type Wharf on Coupled Raking Piles 10.3.1 Design External Forces The external forces acting on open-type wharf on coupled raking piles shall be determined in accordance with 9.3.1 Design External Forces. [Technical Notes] When designing piles for an open-type wharf on coupled raking piles, a deadweight of 21 kN/m2 may be used as the weight of reinforced concrete superstructure.

10.3.2 Calculation of Fender Reaction Force The calculation of fender reaction force shall be made in accordance with 9.3.2 Calculation of Fender Reaction Force.

10.4 Assumptions Concerning Sea Bottom Ground 10.4.1 Determination of Slope Inclination The determination of the inclination of slope beneath the superstructure shall be made in accordance with 9.4.1 Determination of Slope Inclination. 10.4.2 Virtual Ground Surface Refer to 9.4.2 Virtual Ground Surface.

10.5 Determination of Forces Acting on Piles and Cross Sections of Piles 10.5.1 Horizontal Force Transmitted to Heads of Coupled Raking Piles (Notification Article 70, Clause 1) The horizontal forces that are transmitted to the heads of the coupled raking piles of open-type wharf on coupled raking piles shall be calculated appropriately in consideration of the cross sections, raking angles, and lengths of coupled piles. In this calculation, it can be assumed that the horizontal forces are totally borne by the coupled raking piles. -481-


[Technical Notes] (1) When the cross section and/or the raking angle differ among the piles, the horizontal force to be transmitted to the head of each pair of coupled raking piles should be calculated using equation (10.5.1) or (10.5.2) (see Fig. T10.5.1). (a) When the piles can be fully considered as bearing piles ①When it is not necessary to consider the rotation of the deck block: Ci H i = ---------- H S Ci

(10.5.1)

i

②When it is necessary to consider the rotation of the deck block: Ci Ci xi - eH H i = ---------- H + ---------------S Ci S Ci x i2 i

(10.5.2)

i

where sin 2 ( q i1 + q i2 ) C i = ---------------------------------------------------------------------- ( N ¤ m ) l i1 l i2 --------------- cos 2 q i2 --------------- cos 2 q i1 A i1 E i1 A i2 E i2 H： horizontal force acting on the deck block (N/m) horizontal force to be transmitted to the i-th pair of coupled raking piles (N/m) H： i e： distance between the centerline of coupled raking piles group and the point at which the horizontal force acts (m) xi： distance between the i-th pair of coupled raking piles and the centerline of coupled raking piles group (m) li： total length of the pile (m); when there is a pulling force, the length l for friction pile is used. Ai： cross-sectional area of pile of the i-th pair of raking piles (m2) Ei： elastic modulus of pile of the i-th pair of raking piles (N/m2) qi1,qi2： angle between piles of the i-th pair of raking piles and vertical plane (º) In the above, the subscripts 1 and 2 indicate pile numbers within a pair of coupled raking piles. The coordinate of the centerline of the coupled raking piles group can be calculated using the equation SCi ξi/ SCi.The valuable ξi indicates the coordinate of each pair of coupled raking piles in the direction of wharf face line relative to a given coordinate origin. (b) When the piles can be fully considered as friction piles: ① Sandy soil Equations (10.5.1) and (10.5.2) are used with li being replaced with ( 2l i + l i ) ¤ 3 . ② Cohesive soil Equations (10.5.1) and (10.5.2) are used with li being replaced with ( l i + li ) ¤ 2 . where l i： li：

length of the portion of the i-th pile along which skin friction does not act effectively (m) total length of the i-th pile (m)

Horizontal force

Center line of coupled raking piles group Vertical pile

Coupled raking piles Fig. T- 10.5.1 Distance between the Centerline of the Coupled Piles Group and Each Pair of Coupled Raking Piles -482-


(2) When the cross section, raking angle and length are equal among all raking piles, the horizontal force transmitted to each pair of coupled raking piles is calculated by means of equations (10.5.3) and (10.5.4). (a) When it is not neccesary to consider the rotation of the deck block: 1 H i = --- H n where n： number of couples of piles

(10.5.3)

(b) When it is neccesary to consider the rotation of the deck block: xi 1 - eH H i = --- H + --------n S x 2i

(10.5.4)

i

10.5.2 Vertical Load Transmitted to Heads of Coupled Raking Piles (Notification Article 70, Clause 2) The vertical load transmitted to each pair of coupled raking piles shall be calculated appropriately in consideration of the layout and structure of the piles as well as the external forces acting on the wharf. [Technical Notes] For the pushing-in force acting on the piles due to each vertical load, the reaction force calculated by assuming the superstructure as a simply supported beam at the locations of piles may be used.

10.5.3 Pushing-In and Pulling-Out Forces of Coupled Raking Piles (Notification Article 70, Clause 3) The pushing-in and pulling-out forces of each pair of coupled raking piles shall be calculated appropriately based on the vertical and horizontal loads defined in consideration of the use status of the wharf. [Technical Notes] (1) The pushing-in and pulling-out forces of each raking pile may be calculated in accordance with Part Ⅴ, 4.3.6 Lateral Bearing Capacity of Coupled Piles using (a) the horizontal force transmitted to each pair of coupled raking piles that is calculated in accordance with 10.5.1 Horizontal Force Transmitted to Heads of Coupled Raking Piles and (b) the vertical load transmitted to the head of a pair of coupled raking piles that is calculated in accordance with 10.5.2 Vertical Load Transmitted to Heads of Coupled Raking Piles. (2) The pushing-in force of each vertical pile within a open-type wharf on coupled raking piles should be evaluated as the reaction force calculated in accordance with 10.5.2 Vertical Load Transmitted to Heads of Coupled Raking Piles.

10.5.4 Cross-Sectional Stresses of Piles Cross-sectional stresses of piles shall be calculated appropriately by considering each pile as a pile subjected to the axial force and bending moment. [Technical Notes] (1) The cross-sectional stresses of vertical piles should be calculated as a pile subject to the axial load, and those of raking piles should be calculated as a pile subject to either the axial load alone or both the axial load and bending moment (refer to 9.5.5 Cross-Sectional Stresses of Piles). (2) The bearing capacity of raking piles can be calculated by referring to the provisions of [Technical Notes] (1) or (2) in Part Ⅴ, 4.3.6 Lateral Bearing Capacity of Coupled Piles. When using the method prescribed in (1), it is recommended to reduce the allowable stresses of piles by 20% to 30% from those prescribed in 9.5.5 CrossSectional Stresses of Piles [Technical Notes] (3) in or order to neutralize the effects of bending moments and secondary stresses that are neglected in this design method. (3) It is usual that the coupled raking piles have their inclinations in the direction perpendicular to the face line of an open-type wharf. Therefore, it is normal to design the piles as those subject to the axial load against external forces acting perpendicular to the face line. (refer to 10.6 Examination of Strength of Wharf in the Direction of Its Face Line). (4) Open-type wharves on coupled raking piles are normally constructed on the ground with sufficiently large bearing capacity. Therefore, the piles should be carefully designed by examining the effects of impact stresses, buckling, and others during pile driving (refer to Part Ⅴ, 4.5.1 Examinations of Loads during Construction).

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10.6 Examination of Strength of Wharf in the Direction of Its Face Line (1) When coupled raking piles are so layed out to resist the forces acting on the wharf in the direction of its face line, the strength of wharf in the direction of its face line shall be examined in accordance with the method described in 10.5 Determination of Forces Acting on Piles and Cross Sections of Piles, in the manner same as that for examination of strength with respect to the direction normal to the wharf face line. (2) When there is no coupled raking pile layed out to resist forces acting on the wharf in the direction of its face line, examination of strength shall basically be made in accordance with 9.5. Design of Piles in the manner same as that for open-type wharves on vertical piles.

10.7 Embedded Length of Raking Pile The bearing capacity of raking pile shall be examined in accordance with 9.5.6 Examination of Embedded Length for Bearing Capacity and 9.5.7 Examination of Embedded Length for Lateral Resistance.

10.8 Design of Earth-Retaining Section The design of earth-retaining section shall be performed in accordance with 9.7 Design of EarthRetaining Section.

10.9 Examination of Stability against Circular Slip The stability against circular slip shall be examined in accordance with 9.8 Examination of Stability against Circular Slip.

10.10Detailed Design The detailed design of the coping and connection plate shall be performed in accordance with 9.9 Detailed Design.

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Chapter 11 Detached Pier 11.1 Scope of Application The provisions in this chapter shall be applied to the design of a detached pier, which comprises a detached pier section and an earth-retaining section behind the pier. [Technical Notes]

Pine pile

Figure T- 11.1.1 shows an example of a cross section of detached pier.

Steel sheet pile type

Replaced sand

Cohesive soil

Cohesive soil Sandy soil

(Units:m)

Fig. T- 11.1.1 Example of Cross Section of Detached Pier

11.2 Principle of Design A detached pier shall be designed to be stable against all the loads acting on the piers and girders. In addition, the type and dimensions of bridge crane, the traveling characteristics of crane, and the settlement of rails after installation shall be duly considered in determination of the structure of detached pier.

11.3 Design of Detached Pier 11.3.1 Layout and Dimensions The distance between the piers and the land, the gauge of rails for crane, and the span between piers in the direction of the face line shall be determined from the viewpoint of cost and easiness of construction work, by considering the dimensions of crane, the ground conditions of sea bottom, and others. 11.3.2 External Forces and Loads In principle, external forces and loads to be considered in the design of a detached pier shall be the following: (1) Vertical Forces (a) (b) (c) (d)

Wheel loads of cargo handling equipment Tractive forces of vessels Deadweight of superstructure Deadweight of piers

(2) Horizontal Forces (a) (b) (c) (d) (e) (f)

Tractive forces of vessels Berthing force of vessels Wind load acting on cargo handling equipment Seismic force acting on cargo handling equipment Seismic force acting on superstructure Seismic force acting on piers -485-


[Commentary] (1) In the design of a detached pier, the wave force and uplift, and the wind load acting on the superstructure are taken into consideration as appropriate, in addition to the forces listed above. (2) The braking force of the cargo handling equipment is always included in the horizontal forces for the design of girders, but it is taken into account as the necessity arises in the design of piers.

11.3.3 Design of Piers Piers shall be designed appropriately in consideration of their structural type. [Commentary] Design of piers should be carried out in accordance with Chapter 9 Open-Type Wharves on Vertical Piles or Chapter 10 Open-Type Wharves on Coupled Raking Piles. However, design of piers of gravity type or steel sheet pile cellular type should be made by referring to Chapter 4 Gravity Type Quaywalls or Chapter 7 Steel Sheet Pile Cellular-Bulkhead Quaywalls, respectively.

11.3.4 Design of Girder Girders shall be designed to be safe against the vertical as well as horizontal forces and loads. [Commentary] Structural elements having the sufficient strength against the expected vertical and horizontal forces should be used for the girders of a detached pier, because the rails for a crane are directly installed on the girders. In the examination of vertical loads, the increase in the wheel loads due to the wind load or earthquake force acting on the bridge crane should be taken into account. [Technical Notes] (1) Girders are often designed as a simple beam to cope with the differential settlement of the piers. (2) When both legs of the bridge crane are fixed ones, the horizontal load acting on each leg is determined by apportioning the total horizontal load to each leg based on the proportion of the wheel load. When the bridge crane has a fixed leg and a suspended leg, the total horizontal load should be given to the fixed leg for making the design on the safer side. At the same time, however, the horizontal force being one-half of the force acting on one fixed leg in the case of the both legs being fixed should be applied to the suspended leg.

11.4 Ancillary Equipment Ancillary equipment such as fenders, mooring posts and gangways shall be provided with a detached pier. [Technical Notes] For fenders and mooring posts, Chapter 19 Ancillary Facilities can be referred to. For gangways and slabs, a sidewalk live load is set as approximately 5.0 kN/m2.

11.5 Detailed Design 11.5.1 Superstructure The cross section of the superstructure shall be determined in accordance with 9.9.1 Load Combinations for Superstructure Design. 11.5.2 Gangways Gangways shall be designed in accordance with the “Specifications for Highway Bridges and Commentary” (Japan Road Association) and the “Technical Standards for Installation of Grade Separation Facilities and Commentary” (Japan Road Association).

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Chapter 12 Floating Piers 12.1 Scope of Application (1) The provisions in this chapter shall be applied to floating piers having floating bodies (hereinafter referred to as “pontoons”) that are moored by means of mooring chains, etc. (2) The design methods given in this chapter shalls be applied to floating piers installed in places where the actions from waves, tidal currents, and winds are relatively weak. [Commentary] (1) Figures C-12.1.1 and C-12.1.2 show the main components of a floating pier and the structure of a pontoon. A floating pier comprises pontoons, an access bridge that connects the pontoons with land, gangways that interconnect pontoons, mooring chains that moor pontoons, mooring anchors, and other elements. (2) Design of various floating structures to be utilized by unspecified groups of many people should also be performed in accordance with the guidelines given in this chapter.

Mooring anchor Mooring chain Pontoon Gangways

Access bridge

Fig. C-12.1.1 Notation of Respective Parts of Floating Pier

chain hole Protective steel sheet

Fender

Manhole

Partition wall

Fairleader

Chain post Mooring post

Partition wall Supporting beam

Slab

Side wall Bottom slab

Fig. C-12.1.2 Notation of Respective Parts of Pontoon

[Technical Notes] When the site conditions are outside the coverage of this chapter, design work is to be conducted in accordance with the “Technical Manual for Floating Body Structures” and by referring to Part Ⅱ, 5.5 Wave Force Acting on Structure Located near the Still Water Level, Part Ⅱ , Chapter 8 External Forces Acting on Floating Body and Its Motions, and Part Ⅶ , 3.3.2 Floating Breakwater as necessary. -487-


12.2 Principle of Design Floating piers shall be so designed that they are stable and safe during the handling and loading / unloading of cargoes and the embarking / disembarking of passengers and vehicles and that they have sufficient durability. The mooring system including mooring chains and anchors shall be designed to have sufficient strength against the anticipated external forces.

12.3 Design of Pontoon 12.3.1 Dimensions of Pontoon (Notification Article 72, Clause 3) A pontoon shall have a surface area and freeboard appropriate for its purpose of utilization. Dimensions of a pontoon shall be appropriate to make it stable against the external forces and loads acting on it. 12.3.2 External Forces and Loads Acting on Pontoon The external forces and loads that shall be considered in the design of a pontoon are as follows: (1) Static load and live load (2) Reaction forces of gangways and access bridge (3) Hydrostatic pressure (4) Deadweight (5) Counterweight [Commentary] The fender reaction force, wave force, current force, and dynamic water pressure need not be considered unless required to do so because of special site conditions etc. However, when there is a reasonable risk that the pontoon may be subjected to wave actions, it is necessary to consider the following forces: the wave forces exerted upon the stationary pontoon that is assumed to be rigidly fixed in position and the fluid forces due to the motions of the pontoon 1) (refer to Part Ⅱ , Chapter 8 External Forces Acting on Floating Body and Its Motions). In this case, the mooring force is to be calculated by taking into consideration the motions of the pontoon. [Technical Notes] A live load of 5.0 kN/m2 for passengers is commonly used for floating piers, which are mainly used for people boarding and unboarding passenger ships.

12.3.3 Stability of Pontoon (Notification Article 72, Clause 4) Pontoons shall be designed appropriately in such a way that the structural stability levels required for their purposes of utilization are secured. [Technical Notes] (1) In the examination of the stability of a pontoon, the following requirements should be satisfied: (a) The pontoon must satisfy the stability condition of a floating body and have the required freeboard, even with actions of the reaction force from the access bridge and full surcharge on the deck and with presence of some water inside the pontoon owing to leakage. (b) Even when the full surcharge on the deck is loaded on only one side of deck divided by the longitudinal symmetrical axis of the pontoon and the reaction force from an access bridge is applied on this side (if the bridge is attached there), the pontoon must satisfy the stability condition of a floating body and the inclination of the deck should be equal to or less than 1:10 with the smallest freeboard of 0 or more. (2) The height of the water accumulated inside the pontoon by leakeage to be considered in the examination of pontoon stability is usually taken at 10% of the height of pontoon. The freeboard to be maintained in this case is mostly about 0.5 m. (3) When the pontoon is subjected to a uniformly distributed load, the pontoon may be regarded as stable when equation (12.3.1) is satisfied. gwI (12.3.1) -------- CG > 0 W where I： geometrical moment of inertia of the cross-sectional area at the still water level with respect to the longitudinal axis (m4) W： weight of pontoon and uniformly distributed load (kN) -488-

PART VIII MOORING FACILITIES gw： specific weight of seawater (kN/m3) CG： distance between the center of buoyancy and that of gravivity When the pontoon is partially filled with water by leakage, the pontoon may be regarded as stable when equation (12.3.2) is satisfied. The symbols W, I, C, and G of the equation refer to those at the state with water inside. gw ----- ( I Si ) CG > 0 (12.3.2) W where i： geometrical moment of inertia of the water surface inside each chamber with respect to its central axis parallel to the rotation axis of the pontoon (m4) When the pontoon is subjected to an eccentric load, the pontoon may be regarded as stable when the value of tanα obtained by solving equation (12.3.3) satisfies equation (12.3.4) (see Fig. T-12.3.1). b2 d b 2 tan a æ -------- tan 2 a + c ---ö tan a P [ a + ( h c ) tan a ] = 0 ( W 1 + P ) -----------------------2ø 12d cos 2 a è 24d 2( h d) tan a < -------------------b

｝

1 tan a < -----10 where W1： weight of pontoon (kN) P： total force of eccentric load (kN) b： width of pontoon (m) h： height of pontoon (m) d： draft of pontoon when P is applied to the center of pontoon (m) c： height of the center of gravity of the pontoon measured from the bottom (m) a： deviation of P from the center axis of pontoon (m) a： inclination angle of pontoon (º)

(12.3.3)

(12.3.4)

Fig. T-12.3.1 Stability of Pontoon Subjected to Eccentric Load

12.3.4 Design of Individual Parts of Pontoon (Notification Article 72, Clause 5) Stresses generated in the individual parts that comprise the pontoon shall be examined by using an appropriate method selected in consideration of the use of the pontoon, external forces and loads acting on the respective parts, and their structural characteristics. [Technical Notes] (1) Floor slab A floor slab is normally designed as a two-way slab fixed on four sides with supporting beams and side walls against the loads that yield the largest stress out of the following combinations of loads.: (a) When only static load acts on a pontoon Static load and deadweight (b) When live load acts on a pontoon Live load and deadweight (c) When the supporting point of an access bridge is set on a pontoon (without adjustment tower) Supporting point reaction force of an access bridge and deadweight

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(2) Side Walls A side wall is normally designed as a two-way slab fixed on four sides with a floor slab, a bottom slab, and side walls or supporting beams, against hydrostatic pressure acting when the pontoon submerges by 0.5 m above the deck. (3) Bottom Slab A bottom slab is normally designed as a two-way slab fixed on four sides with side walls or supporting beams, against hydrostatic pressure acting when the pontoon submerges by 0.5 m above the deck. (4) Partition Walls A partition wall is designed as a slab fixed on four sides. (5) Supporting Beams and Supports The supporting beams of the floor slab, bottom slab and sidewalls and the center support are normally designed as a rigid frame box under the condition that the maximum load is acting on the floor slab of the pontoon and the hydrostatic pressure for the draft of the pontoon being equal to its height is applied. (6) When the wave actions are to be taken into consideration, calculations of section forces are made using Muller’s equation 2), the method of “Prestressed Concrete Barge”, or DnⅤ Rule. When it is necessary to consider effects of the motions of the floating body, waves height and period, and water depth, the method with cross-sectional division by Ueda et al 1), 3), 4), 5) may be used.

12.4 Design of Mooring System 12.4.1 Mooring Method (Notification Article 72, Clause 1) The method to moor the pontoons of a floating pier shall be selected appropriately in consideration of the natural conditions of the installation area so that the floating pier can serve its purposes adequately. 12.4.2 Design of Mooring Chain [1] Design External Forces (Notification Article 72, Clause 6) The structure of mooring chains shall be examined by using an appropriate method in such a way that the chains can hold securely a pontoon in position under the action of the largest force. [Technical Notes] (1) The guidelines given in this section applies for the pontoons moored by chains, which have been used most frequently in floating pier design. (2) With regard to external forces, the combination among the following that gives the largest total force may be used. (a) Fender reaction force and tidal current force (b) Tractive force of ship’s mooring lines and tidal current force (c) Wave force and tidal current force

[2] Setting of Chain Each chain shall be set to one of the chain posts provided at the four corners of a pontoon through a chain hole, and fixed to the sea bottom by an anchor. [3] Diameter of Chain The diameter of the chain shall be determined not to be broken during the high tide by the external forces specified in [1] Design External Forces. [Technical Notes] (1) The maximum tension that acts on each chain is ideally determined through dynamic analysis of the chain and the pontoon, but as this is very difficult, static analysis may be used as a second-best method. A chain is normally designed on the condition that only one chain must resist against all the external forces (see Fig. T12.4.1). Assuming that the chain forms a catenary line, the maximum tension that acts on the chain is given by equation (12.4.1). T = P sec q2

(12.4.1)

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The horizontal force that acts on the mooring anchor is the same as the horizontal force acting on the pontoon, and the vertical force acting on the anchor is given by equation (12.4.2). Va = P tanq1

(12.4.2)

The vertical force that acts on the joint between the chain and the pontoon is given by equation (12.4.3). Vb = P tan q2

(12.4.3)

The angles q1 and q2 are calculated by solving equation (12.4.4) with an assumed chain length l and an assumed chain weight w per unit length of the chain. P (12.4.4) l = ---- ( tan q 2 tan q 1 ) w P h = ---- ( sec q 2 sec q 1 ) w

Fig. T-12.4.1 Notations Related with Mooring Chain

The horizontal distance between a mooring anchor and the pontoon when a horizontal force is acting on the pontoon is given by equation (12.4.5), and thus the amount of horizontal shift of the pontoon from its stationary position under no horizontal force can be easily evaluated. P (12.4.5) K h = ---- [ sinh 1 ( tan q 2 ) sinh 1 ( tan q 1 ) ] w Because the catenary line of the chain of normal diameter can be approximately represented with a straight line, it can be assumed in equations (12.4.1) to (12.4.3) that q 2 = q 1 = sin 1 ( h ¤ l ) and K h = l 2 h 2 . where T： maximum tension that acts on the chain (kN) P： horizontal external force (kN) Va： vertical force that acts on the mooring anchor (kN) Vb： vertical force that acts on the joint between the chain and pontoon(kN) q1： angle that the chain makes with the horizontal plane at the joint between the mooring anchor and chain (º) q2： angle that the chain makes with the horizontal plane at the joint between the mooring chain and pontoon (º) l ： length of the chain (m) w： submerged weight per unit length of the chain (kN/m) h： water depth beneath the bottom of pontoon (m) Kh： horizontal distance between the mooring anchor and the joint between the pontoon and chain (m) (2) In the determination of the diameter of the chain, careful consideration should be given to the abrasion, corrosion, and biofouling of chain. In addition, appropriate maintenance work needs to be carried out on the chain, including periodical checks on the chain and its replacement on an as-necessary basis. (3) When determining the chain diameter using numerical dynamic simulation, the characteristics of displacement restoration force relationship of the mooring system need to be determined using an appropriate method such as the catenary theory 6).

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12.4.3 Design of Mooring Anchor [1] Design External Forces In the design of mooring anchors, the maximum tension that acts on the mooring chain as calculated in 12.4.2 Design of Mooring Chain shall be used as the design external force. [Technical Notes] The following values calculated in 12.4.2 Design of Mooring Chain may be used as external forces: (a) Vertical force : Va = P tan q1 (b) Horizontal force : P For Va, P, and q1, refer to Fig. T-12.4.1.

[2] Design of Mooring Anchor (Notification Article 72, Clause 7) A mooring anchor shall be capable of providing the resistance forces required to keep the pontoon stable against the maximum tension that acts on the mooring chain and shall be designed with an appropriate factor of safety. [Technical Notes] (1) Normally the following forces are considered as the resistance forces of a mooring anchor, but it is recommended that in-situ stability tests be made for a mooring anchor: (a) In the case of concrete block: ① On clay: Horizontal resistance force: Cohesion of the surfaces of bottom and sides, difference between the passive and active earth pressures Vertical resistance force: Submerged weight, effective overburden weight ② On sand: Horizontal resistance force: Bottom friction force, difference between the passive and active earth pressures Vertical resistance force: Submerged weight, effective overburden weight The vertical force employed in the calculation of the bottom friction force is the difference between the submerged weight of the block and the vertical component of the chain tension that acts on the block. (b) In the case of steel mooring anchor: Horizontal resistance force: Holding power Vertical resistance force: Submerged weight The holding power of a steel mooring anchor is calculated by means of equation (12.4.6).

64748

Soft mud: TA = 17 WA2/3 Hard mud: TA = 10 WA2/3 Sand : TA = 3 WA Flat rock: TA = 0.4 WA where TA： holding power of the mooring anchor (kN) WA： submerged weight of the mooring anchor (kN)

(12.4.6)

(2) For mooring anchor stability, a safety factor of 1.2 or more may be used.

12.5 Design of Access Bridge and Gangway 12.5.1 Dimensions and Inclination (Notification Article 72, Clause 2) An access bridge and gangways of a floating pier shall have appropriate dimensions and inclinations which ensure that the floating pier provides the required handling performance for passengers, vehicles, and cargoes.

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12.5.2 Design of Access Bridge and Gangway Access bridges and gangways shall be designed in accordance with the “Specifications for Highway Bridges with Commentary” (Japan Road Association) and the “Technical Standard with Commentary for Installation of Grade Separation Facilities” (Japan Road Association). 12.5.3 Adjusting Tower An adjusting tower shall be designed to be sufficiently safe against the reaction forces of the access bridge and seismic forces. [Technical Notes] Figure T-12.5.1 shows an example of adjusting tower.

Fig. T-12.5.1 Adjusting Tower

[References] 1) Shigeru UEDA, Satoru SHIRAISHI, Kazuo KAI: “Calculation method of shear force and bending moment induced on pontoon type floating structures in random sea”, Tech. Note of PHRI, No. 505, 1984 (in Japanese). 2) Jean Muller: “Structural Consideration Configurations Ⅱ”, University of California Extension Berkeley, Seminar on Concrete and Vessels, Sept. 1965. 3) Shigeru UEDA, Satoru SHIRAISHI, Takashi ISHISAKI: “Calculation method of forces and moments induced on pontoon type floating structures in waves”, Rept. of PHRI, Vol. 31, No. 2, 1992 (in Japanese). 4) Shigeru UEDA, Satoru SHIRAISHI, Takashi ISHISAKI: “Example of calculation of forces and moments induced on pontoon type floating structures with figures and tables of radiation forces”, Tech. Note of PHRI, No. 731, 1992 (in Japanese). 5) Shigeru UEDA: “Analytical method of ship motions moored to quaywalls and the applications”, Tech. Note of PHRI, No. 504, 1984 (in Japanese). 6) Shigeru UEDA, Satoru SHIRAISHI: “Determination of optimum mooring chain and design charts using catenary theory”, Tech. Note of PHRI, No. 379, 1981 (in Japanese).

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Chapter 13 Dolphins 13.1 Principle of Design The provisions in this chapter shall be applied to the design of berthing facilities of dolphin-structure such as pile type, steel cellular-bulkhead type, and caisson type. [Technical Notes] (1) Figure T-13.1.2 shows an example of a pile type dolphin Breasting dolphin Mooring dolphin Mooring dolphin Cargo handling platform

Cargo handling platform Mooring post

Bitt

Mooring post

Bitt

Mooring post

HWOST LWOST

original sea botton

Bearing subsoil

(Units: mm)

Fig. T-13.1.2 Example of Pile Type Dolphin

13.2 Layout The layout of a dolphin-berth shall be determined appropriately in consideration of the dimensions of the target vessels that are to use the berth, water depth, wind direction, wave direction, and tidal currents, as well as no adverse effects on the navigation and anchorage of other vessels. [Technical Notes] (1) In the determination of the layout of breasting dolphins, the following items need to be examined: (a) Dimensions of the target vessel The broadside of large vessels that use dolphins is usually composed of a straight line defining the outline of the central part (which accounts for about 3/4 of the length overall (L) of vessel) and two curves defining the outline of the bow and stern parts (each of which accounts for about 1/8 of the length overall (L) of vessel). It is recommended that the breasting dolphins be layed out in such a way that the vessels can be berthed to them with the straight line part. When a special cargo handling equipment is required for dolphins such as in the case of oil dolphins, a cargo handling platform is installed midway between the breasting dolphins. In this case, it is desirable to locate the cargo handling platform with its seaside front slightly backward from that of the breasting dolphins, in order to prevent the ship berthing force acting directly on the cargo handling platform. (b) Water depth, wind direction, wave direction, and tidal currents It is desirable to layout dolphins in a way that the longitudinal axis of dolphins becomes parallel to the prevailing directions of winds, waves, and tidal currents. This helps to ease vessel maneuvering during berthing and undocking and to reduce external forces that act on the dolphins when the vessel is moored. (2) The normal layout of mooring dolphins is such that they are set at the positions located on the lines drawn with the angle of 45º from the bitts on ship’s bow and stern and having a certain setback from the front face of the breasting dolphins. (3) The distance between breasting dolphins is closely related to the length overall (L) of the target vessels. Figure T-13.2.1 gives the relationship between the breasting dolphin interval and the water depth derived from past construction data for reference.

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Distance between breasting dolphins (m)


Pile type Steel sheet pile cellar cofferdam type Caisson type

Water depth (m)

Fig. T-13.2.1 Distance between Breasting Dolphins

13.3 External Forces Acting on Dolphins The following shall be considered as the external forces acting on dolphins: (1) Berthing force (2) Tractive force from vessel’s mooring lines (3) Vertical forces due to deadweight and live load. (4) Seismic force on structures and cargo handling equipment (5) Dynamic water pressure during an earthquake (6) Wind pressure on cargo handling equipment

13.4 Pile Type Dolphins In the basic design of pile type dolphins, the following items shall be examined as a general rule: (1) Stresses of the piles (stresses due to the horizontal force, vertical force, and torsion) (2) Embedded lengths of the piles (3) Deflection of the dolphin [Technical Notes] (1) The items listed above should be examined according to Chapter 9 Open-Type Wharves on Vertical Piles and Chapter 10 Open-Type Wharves on Coupled Raking Piles. (2) When pile type dolphins are employed, the berthing energy may normally be calculated on the assumption that it is absorbed by the deformations of the fenders and the piles. (3) Large tankers are usually berthed at a slant angle with the dolphin alignment line. As the characteristics of fenders vary depending on the berthing angle, it is recommended in such a case to use the characteristics curve appropriate to the berthing angle. In addition, a slanting berthing entails the risk that some of the fenders attached to a breasting dolphin may not absorb the berthing energy effectively. Therefore, it is desirable to examine carefully which fenders will come in contact with the hull of vessel in consideration of the berthing angle.

13.5 Steel Cellular-Bulkhead Type Dolphins In the basic design of steel cellular-bulkhead type dolphins, the following items shall be examined as a general rule: (1) Equivalent wall width to shear deformation (2) Embedded length of cellular-bulkhead (3) Stability of the wall body as a whole (sliding, overturning, bearing capacity) (4) Tension of cell shell (5) Foundations for cargo handling equipment and mooring post.

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[Technical Notes] (1) Examination of the items (1) to (4) above should be made in accordance with Chapter 7 Steel Sheet Pile Cellular-Bulkhead Quaywalls and Chapter 8 Steel Plate Cellular-Bulkhead Quaywalls, and examination of the item (5) is made in accordance with Part Ⅴ , Chapter 4 Bearing Capacity of Pile Foundations and Chapter 21 Foundation for Cargo Handling Equipment. (2) In the case of a cylindrical cellular type, the equivalent wall width can be calculated using equation (13.5.1). B = 3R where B： equivalent wall width (m) R： radius of cylindrical cell (m)

(13.5.1)

13.6 Caisson Type Dolphins In the basic design of caisson type dolphins, the following items shall be examined as a general rule: (1) Sliding of caisson (2) Overturning of caisson (3) Subsoil bearing capacity (4) Rotation of caisson (5) Strength of structural elements [Commentary] Rotation of a caisson occurs when an eccentric external force acts on a dolphin. Examination of stability against rotation must be made even when the results of the examination of the items (1) to (3) listed above are satisfactory, because the confirmation of the stability with respect to the items (1) to (3) does not guarantee that the caisson is safe against rotation. In the examination of stability against rotation, it is assumed that the resisting force is the friction force along the caisson bottom which is proportional to the bottom reaction to be calculated in accordance with Part Ⅳ, 1.4.5 External Forces after Construction. The safety factor is determined in accordance with 4.3.2 Examination against Sliding of Wall.

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Chapter 14 Slipways and Shallow Draft Quays 14.1 Slipways 14.1.1 Principle of Design Slipways shall be designed in such a way that no operational or structural problem would arise. [Commentary] (1) A slipway is a facility used to take ships up to the shore and down to the sea for such purposes as repair, protection from storm waves and storm surges, and land storage of ships during winter. (2) In many cases, rails or cradles are employed in the lifting and lowering of ships of 30 tons or larger in gross tonnage, but the provisions in this section applies to facilities used to lift and lower ships smaller than 30 tons in gross tonnage directly over the slope of slipway. (3) Figure C-14.1.1 defines the notations of various parts of a slipway.

MSL

Concrete block covering Foundation

Height of front wall Armouring Sea bottom in front of the slipway

Slope

Front wall

Intermediate retaining wall

Ship storing yard Concrete pavement Foundation

retaining wall

Fig. C-14.1.1 Slipway

14.1.2 Location of Slipway Locations of slipways shall be determined in such a way that the following requirements are satisfied: (1) The frontal water area is calm. (2) The frontal water area is free from siltation or scouring. (3) Navigation and anchorage of other ships are not hindered. (4) There is an adequate space in the background for the work for ship lifting and lowering as well as for ship storage. 14.1.3 Dimensions of Individual Parts The elevations of individual sections and the length of the slipway, the area of the yard behind the slipway, the depth of the water in front of the slipway, the inclination of the slipway, and the area of the anchorage in front of the slipway shall be designed in consideration of the types and sizes of ships that are to untilize the slipway as well as the ease of usage. [1] Elevations of Individual Parts [Technical Notes] (1) It is recommended that the crown of the front wall of the landing section be located at a level lower than the mean monthly-lowest water level (LWL) by the amount of the draft of ships under consideration. It is owing to the requirement that it is possible to lift ships even at the low water of neaps. The draft of the ship should be the light draft for the case of repair, refuge, and wintertime storage, and should be the full-load draft for the case of lifting small fishing boats filled with catches. For slipways that are to be constructed in the areas where tidal ranges are small or for the slipways that are to be used even at the low water springs with high waves, it is necessary to lower the crown elevation of the front wall further. (2) The ground elevation of the ship storage area should be determined in accordance with 2.2 Crown Heights of Mooring Facilities. However, when the ship storage area is located adjacent to a quaywall, the ground elevation of the ship storage area may be set equal to the crown elevation of the quaywall to facilitate ease of usage. In cases where waves are high in the basin in front of the ship storage area, the wave runup height should be taken into consideration.

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[2] Slipway Length and Background Space (Notification Article 76, Clause 1) The slope length of the slipway and the area of the background space shall be adequate for the handling of the ships being considered. [Technical Notes] (1) The area of the background space refers to the yard area in which lifted ships are stored. This ordinarily means the area of the flat portion of land, but may include a part of the upper slope when the ground elevation of the ship storage area is high. It is desirable to make the length of this part of the slope about 5 m longer than the length overall of the largest ship is to be stored. (2) The distance between lifted ships for storage should be 2 m or more in the direction of the ship’s bow, and 1 m or more in the direction of the ship’s beam.

[3] Water Depth [Technical Notes] The reference depth of water in front of the slipway is the sum of the draft of the ship being considered and a margin of 0.5 m

[4] Gradient of Slipway (Notification Article 76, Clause 3) The gradient of slipway shall be determined appropriately in consideration of the shape of ships that are to use the slipway, the characteristics of foundation, and the tidal range, so that the lifting of ships can be performed smoothly. [Technical Notes] (1) When the slope is to be utilized by small ships, it is desirable to design the slope as having a single-gradient. Single-gradient slopes are frequently used in slipways for human power-based lifting of ships that are constructed in shallow waters. For this type of slipways, a slope inclination of 1:6 to 1:12 may be used as a reference. (2) When the water in front of the slipway is deep or the area of the construction site is limited, the slope may be built with two or more gradients. When this is the case, a two-gradient slope may be employed when the crown elevation of the front wall is about -2.0 m, and a three-gradient slope may be employed when the crown elevation of the front wall is lower than -2.0 m. The following values may be used as reference gradients: When the slope has two gradients: Front slope: 1:6 to 1:8 Rear slope: 1:8 to 1:12 When the slope has three gradients: Front slope: an inclination steeper than 1:6 Central slope: 1:6 to 1:8 Rear slope: 1:8 to 1:12

[5] Basin Area (Notification Article 76, Clause 4) The basin in front of a slipway shall have an appropriate area that allows a smooth operation of ship lifting and/or lowering without damage to the ships, and a safe and smooth navigation of nearby ships. [Technical Notes] (1) When the ship is lowered to the sea by free fall over the slope, the ship runs over a certain distance after hitting the water with the speed it gains during the fall. This distance is more than about five times the ship’s length overall, although it varies depending on the slope inclination, slope friction, and fall distance. However, because the ship becomes maneuverable after running a distance about 4 to 6 times its length, it is sufficient to secure a distance about five times the ship’s length overall from the waterfront line of the slope to the other end of the basin. When strong tidal currents exist, it is necessary to add an appropriate margin. (2) When the ship is lowered to the sea gently by means of wire ropes and other means, a distance of about three times the ship’s length overall will suffice to define the required basin area.

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14.1.4 Front Wall and Pavement [1] Front Wall (Notification Article 76, Clause 2 and 5) (1) The structure of the front wall of slipway shall be designed appropriately in consideration of the dimensions of ships that are to use the slipway, the crown elevation of the front wall, and the ship lifting method. (2) The frontwall of slipway shall be designed appropriately in accordance with applicable guidelines given in Chapter 4 Gravity Type Quaywalls through Chapter 10 Open-Type Wharves on Coupled Raking Piles and Chapter 17 Other Types of Mooring Facilities in consideration of the structure of the wall, so that the stability of the wall is secured. [2] Pavement The pavement shall be designed in such a way that the following requirements are satisfied: (1) The pavement will not be broken by the anticipated loads of ship weights etc. (2) The pavement will not be broken by waves actions. (3) The pavement will not be broken by differential settlement of the foundation. [Technical Notes] In most cases, the pavement is constructed using cement concrete. The average dimensions are as follows: Cast-in-place concrete: Slab thickness : 20 to 35 cm Joint interval : 5 to 10 m Precast concrete block: Slab size :2m´2m Slab thickness : About 30 cm However, when waves are expected to become high or in the case of restoration work after a natural disaster, the block thickness may be determined in accordance with Fig. T-14.1.11. As for the base course, it is common practice to use a design thickness of about 30 cm and to make thorough compaction in order to avoid differential settlement.

Wave height

H

1/3

(m)

t : thickness (cm)

Period

T

(s)

Fig. T-14.1.1 Required Thickness for Precast Concrete Block and Height of Waves in Front of Slipway

14.2 Shallow Draft Quay (Notification Article 75) Shallow draft quays shall be designed appropriately in accordance with applicable guidelines given in Chapter 4 Gravity Type Quaywalls through Chapter 10 Open-type Wharves on Coupled Raking Piles as well as Chapter 12 Floating Piers and Chapter 17 Other Types of Mooring Facilities according to the characteristics and use of the facility, so that the structural safety is secured.

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Chapter 15 Air-Cushion Vehicle Landing Facilities 15.1 Principle of Design Air-cushion vehicle landing facilities shall be located at an appropriate position and have an appropriate structure for the safe boarding of passengers and safe and smooth landing of the vehicle. [Technical Notes] (1) Air-cushion vehicle landing facilities are normally constructed on the shore. These facilities usually use slopes similar to those of slipways as described in Chapter 14 Slipways and Shallow Draft Quays to land and glide down air-cushion vehicles. (2) Figure T-15.1.1 illustrates an air-cushion vehicle. (3) Figure T-15.1.2 shows an example of the layout of air-cushion vehicle landing facilities. Fan for floating

Propeller

Antenna

Air rudder Propeller

Steering seat

Flexible skirt

Leg

Submarine (Placed in water and used as rudder)

Fig. T-15.1.1 Air-Cushion Vehicle (MV-PP5)

.

Fueling

Garage

facility

Parking lot River

Slip

Apron

Sound-proof wall

Passenger Rotary

rk

a ip

x Ta

shed

in g

Park

ea

ar

Fig. T-15.1.2 Example of Layout of Air-Cushion Vehicle Station

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15.2 Location In determining the location, the following requirements shall be taken into consideration: (1) The water basin in front of the facility is calm. (2) The effects of strong winds and beam winds on the vehicle are minimum. (3) Operation of the vehicles does not hinder navigation and mooring of other ships. (4) Influences of noise and water spray from the operation of vehicles upon other navigation ships and neighboring area are minimum. [Commentary] (1) Air-cushion vehicles are excellent of stability in high-speed operation, but they are susceptible to influences of winds during low-speed operation such as required to approach and leave a landing facility. In the determination of the location of an air-cushion vehicle station, therefore, it is necessary to give careful consideration to the level of calmness of the basin in front of the facilities and the direction of the prevailing wind. (2) As noises from an air-cushion vehicle may be as high as 100 dB (at a distance of 50 m from the vehicle), it is necessary to locate air-cushion vehicle landing facilities far away enough from hospitals, schools and housing areas, or to shut off the noises by surrounding the facilities with sound-proof walls.

15.3 Air-Cushion Vehicle Landing Facilities The air-cushion landing facilities shall be provided with a slipway, apron, and passenger boarding facility. In addition, lighting facilities, a hanger, sound-proof walls, an oil supply facility, a repair facility and others shall be provided as necessary.

15.4 Dimensions of Individual Parts [Technical Notes] (1) Slope (a) The structure of the slope should be determined in accordance with Chapter 14 Slipways and Shallow Draft Quays. (b) The width of the slope should be determined in consideration of the lateral movement of the air-cushion vehicle during the landing or gliding-down operation due to beam winds. Usually a width about three times the width of the vehicle is provided. (c) The inclination of the slope should be determined in consideration of its psychological effect on passengers, performance of the air-cushion vehicle, and site-related limitations. Usually an inclination of 1:10 or gentler is adopted. (2) Apron In many cases the apron width is the same as that of the slope and the apron length is about two times the length of the air-cushion vehicle. In cases where two or more air-cushion vehicles utilize the landing facility simultaneously, a parking space should be provided alongside the apron. (3) Hangar When a hangar is to be constructed, it should be located adjacent to the apron to facilitate the servicing and maintenance of air-cushion vehicle and to provide the refuge space of air-cushion vehicle in rough weather. The dimensions of the hangar should be as follows: Width: 1.5 times the width of the air-cushion vehicle (per vehicle) Length: 1.2 times the length of the air-cushion vehicle (per vehicle) Height: There should be a clearance of about 0.5 m from the ceiling to the top of the air-cushion vehicle when the vehicle is lifted afloat.

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Chapter 16 Mooring Buoys and Mooring Posts 16.1 Mooring Buoys 16.1.1 Principle of Design (Notification Article 73, Clause 1) The mooring buoy shall have an appropriate structure and the stability in consideration of the mooring method used, the natural conditions at the site, and the dimensions of the target vessel to be moored at the buoy. [Technical Notes] (1) Mooring buoys are structurally categorized into three types. These are sinker type, anchor chain type, and sinker and anchor chain type. The sinker type mooring buoy comprises a floating body, main chain, and sinker. It does not have a mooring anchor, as shown in Fig. T-16.1.1(a). The anchor chain type mooring buoy comprises a floating body, anchor chain, and mooring anchor. It does not have a sinker, as shown in Fig. T-16.1.1(b). Although the construction cost is lower as compared with the other types, it is not suitable for cases where the area of the mooring basin is limited, because the radius of vessel’s swinging motion is large. The sinker and anchor chain type mooring buoy comprises a floating body, main chain, ground chain, mooring anchor, and sinker, as shown in Fig. T-16.1.1(c). The sinker and anchor chain type mooring buoys are being used widely in ports and harbors. This type of buoys can be used even when the area of the mooring basin is small, because the radius of vessel’s swinging motion can be reduced by increasing the weight of the sinker. Floating body

Floating body

Floating body

Main chain Main chain Anchor chain

Ground chain Sinker

Mooring anchor

Mooring anchor

Sinker chain Sinker

(a) Sinker Type

(b) Anchor Chain Type

(c) Sinker and Anchor Chain Type

Fig. T-16.1.1 Types of Mooring Buoys

(2) In general, it is recommended that the design of mooring buoy be performed in the procedure depicted in Fig. T16.1.2. Determination of design conditions ↓ Calculation of the tractive force ↓ Determination of the weight and shape of the mooring anchor ↓ Assumption of the weight of the sinker and the diameter and length of the sinker chain ↓ Calculation of the diameter and length of the ground chain ↓ Determination of the diameter and length of the main chain ↓ Calculation of the buoyancy of the floating body ↓ Detailed design Fig. T-16.1.2 Flowchart of Mooring Buoy Design

(3) Figure T-16.1.3 shows a typical makeup of mooring buoy.

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Mooring chain

Fender

Lift chain

A: Harp shackle (mooring ring) or quick-release hook B: Anchor shackle C: Swivel piece D: Joining shackle E: Mooring piece F: Long ring G: Joining shackle H: Anchor shackle I: Joining shackle J: Anchor shackle K: Chain (inside the buoy) L: Main chain M: Sinker chain N: Chain or ground chain P: Sinker or pan Q: Anchor or screw anchor R: Buoy

1 1 1 1 1 2 2 1 2 2 1 1 1 4 1 1 1

Fig. T-16.1.3 Typical Makeup of Mooring Buoy

(4) This chapter provides guidelines for the design of sinker and anchor chain type mooring buoys, because the sinker type and anchor chain type buoys are simplified versions of the sinker and anchor chain type buoy. (5) In the design of a mooring buoy, the following items should be considered: (a) Mooring conditions: ① Mooring method (single-buoy mooring or two-point mooring) ② Allowable range of the vessel’s motion (allowable swinging radius) (b) Natural conditions: ① Winds ② Tidal currents ③ Waves ④ Soil condition of mooring basin ⑤ Water depth ⑥ Tide level (c) Dimensions of vessel to be moored at the buoy ① Length overall ② Draft ③ Molded breadth ④ Height of the hawser hole as measured from the water surface ⑤ Front and side projected areas above the water surface ⑥ Front and side projected areas below the water surface

16.1.2 Tractive Force Acting on Mooring Buoy (Notification Article 73, Clause 2) In principle, the tractive force that acts on a mooring buoy through the hawser of a vessel shall be calculated in consideration of structural characteristics of the mooring buoy, based on the external forces determined in accordance with the provisions in Part Ⅱ　 Design Conditions and the existing data on the tractive forces of vessels moored at buoys. [Technical Notes] (1) It is recommended that the tractive force which acts on a mooring buoy be determined in consideration of the loads that exert upon moored vessels (including those due to winds, tidal currents, and waves) and using the existing tractive force data on the buoys of the same type as reference 1), 2). (2) Wind and wave forces are dynamic loads, and thus there are many uncertainties with regard to their relationships with the tractive forces of vessels. However, the tractive force may be calculated by referring to 2.2.3 Moored Vessels. -503-


(3) When the motions of buoy due to wave actions cannot be ignored, their effect should be included in the calculation of the wave force and the wave-making resistance. (4) In a dynamic analysis of a moored floating body, the response characteristics of the floating body vary widely depending on the wave period. If the analysis is made using monochromatic waves only, either underestimated or overestimated results are obtained. When performing a dynamic analysis of the motions of a floating body, therefore, it is necessary to employ irregular waves having a directional spectral characteristics. (5) Table T-16.1.1 lists examples of design conditions and loads for mooring buoys built in the past. Table T-16.1.1 Some Examples of Design Conditions and Loads for Mooring Buoys Vessel to use the mooring buoy (DWT)

Tractive force (kN)

Mooring method

Wind speed (m/s)

Tidal currents (m/s)

Wave height (m)

1,000

Single-buoy mooring

50

0.5

2.0

185

3,000

〃

50

0.5

4.0

409

15,000

〃

15

0.51

0.7

245

20.000

〃

20

1.0

-

589

130,000

〃

60

0.67

10.0

1,370

260,000

〃

25

0.51

3.0

1,840

30,000

Two-point mooring

15

-

-

1,490

100,000

Six-point mooring

20

-

1.5

1,470

16.1.3 Design of Individual Parts of Mooring Buoy [1] Mooring Anchor (Notification Article 73, Clause 3) The sizes and required strengths of individual parts of a mooring buoy, including the mooring anchor, sinker, sinker chain, ground chain, main chain, and floating body, shall be determined appropriately in accordance with the relevant provisions in Chapter 12 Floating Piers and in consideration of the tractive forces of vessels, the structure of the mooring buoy, and the mooring method in use, etc. [Technical Notes] (1) Normally three mooring anchors are used for a mooring buoy. In the design of a mooring buoy, however, it should be assumed that only one of the three anchors resists the horizontal force. The arrangement of the mooring anchors should be made in such a way that the buoy will not capsize even when one of the anchor chains is cut. (2) In principle, it should be assumed that the horizontal force that acts on the mooring buoy is resisted by the mooring anchors’ resistance force only. The holding power of the mooring anchors should be calculated in accordance with 12.4.3 Design of Mooring Anchor.

[2] Sinker and Sinker Chain [Technical Notes] (1) Normally a sinker chain of 3 to 4 m in length is used for a mooring buoy. It is recommended not to use an excessively long sinker chain, because it makes the range of the upward movement of the sinker larger and increases the risk of the tangling of the sinker chain and thus the risk of abrasion and accidental breaking of the chain. The sinker chain should be of the same diameter as that of the main chain. (2) The vertical and horizontal forces that act on the sinker should be calculated based on the chain tension of floating body and the distance of horizontal movement of the floating body as calculated 3) in accordance with [4] Main Chain.

678

PV = TA sinq1 = (TC - wl) sinq1 PH = TA cosq1 = (TC - wl) cosq1

(16.1.1)

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where PV , PH： q1： TA： TC： w： l：

vertical and horizontal forces that act on the sinker, respectively (kN) angle that main chain makes with the horizontal plane at the sinker attachment point (º) tension of main chain at the sinker attachment point (kN) tension of main chain at the floating body attachment point (kN) submerged weight of the main chain per unit length (t/m) length of main chain (m)

64748

The angle q1 is obtained by solving the following equations: T A cos q 1 l = --------------------- ( tan q 2 tan q 1 ) w T A cos q 1 1 1 DK = --------------------- [ sinh ( tan q 2 ) sinh ( tan q 1 ) ] w

(16.1.2)

where DK： distance of horizontal movement of the floating body (m) q 2： angle that main chain makes with the horizontal plane at the floating body attachment point (º) When a tractive force equal to the design force is applied, the alignment of the floating body chain becomes approximately a straight line and thus the following approximation can be used: DK (16.1.3) q 2 ≒q 1 = cos -1 -------l (3) Weight of Sinker Currently a sinker weight of about 50 kN is most frequently used for 5,000 GT vessels, and a sinker weight of about 80 kN tons is most frequently used for 10,000 GT vessels. Determination of the sinker weight should be made using these values as references (the values mentioned above indicate submerged weight). Sinkers may be of any shape and material as long as they satisfy the weight requirement, but in Japan the common practice is to use disk-shaped cast iron sinkers and concrete is seldom used. It is said that use of disk-shaped cast iron sinkers having a slightly concaved bottom surface improves the adhesion of the sinker to the soft sea bottom ground significantly. (4) The role of the sinker is to absorb the impact force that acts on the chain and to make the main chain shorter. When the main chain is to be shortened to reduce the range of ship movement, therefore, the weight of the sinker must be increased accordingly. (5) In certain cases, buried anchors may be used instead of sinkers.

[3] Ground Chain [Technical Notes] (1) Because chains are manufactured with a lot length of 25 m, a ground chain of 50 m long is usually used. The angle with which the chain makes with the sea bottom at the mooring anchor attachment point should be smaller than 3º, because the holding power of the mooring anchor drops sharply as the angle increases beyond 3º. In many cases, the weight of the ground chain is designed in such a way that the ground chain satisfies the abovementioned condition when the anticipated tractive force acts on the buoy. When the tractive force is large, the attachment angle that the mooring anchor makes with the ground chain may be made smaller by using a ground chain longer than the above-mentioned value. The inclination angle q1 of the ground chain at the mooring anchor attachment point can be calculated by using equation (12.4.4) described in 12.4.2 [3] Diameter of Chain. The symbols in equation (12.4.4) are redefined as follows (see Fig. T-16.1.4): l： length of the ground chain (lg in Fig. T-16.1.4) (m) h： vertical distance between the upper end of the ground chain and the sea bottom (i.e., the sum of the length of the sinker chain, height of the sinker, and allowance (hg in Fig. T-16.1.4) (m) P： horizontal component of the tractive force that acts on the floating body (kN) w： submerged weight of the ground chain per unit length (kN/m) q 2： inclination angle of the ground chain at the upper end of the chain (º) In this calculation, the value of q 1 should be calculated by assuming the values of lg, w, and hg ; q1 should be kept at 3º or less. (2) The maximum tension Tg of the ground chain should be calculated using equation (12.4.1) described in 12.4.2 [3] Diameter of Chain. Here P represents the horizontal component of the tractive force of vessel that acts on the buoy, and q 2 represents the inclination of the ground chain at the upper end of the chain. The diameter of the ground chain should be such that the chain can withstand this tension.

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(3) The allowable tensile force of chain should be calculated in accordance with 12.4.2 Design of Mooring Chain. In the case of mooring buoys, however, the diameter of chain is usually determined not on the basis of strength only, but on the basis of total analysis that takes into consideration any measures taken to reduce forces that act on the chain such as the use of a heavier chain to absorb the energy of impact forces, and the use of a shorter chain to reduce the radius of the vessel’s swinging. In general, the chain diameter is selected in such a way that the maximum tension to be exerted upon the chain will be 1/5 to 1/8 of the breaking strength.

f

Fig. T-16.1.4 Notation for Sinker and Anchor Chain Type Mooring Buoy

[4] Main Chain [Technical Notes] (1) The length lf of the main chain should be designed in such a way to lessen the tension acting on both the main chain and the mooring hawser as well as to reduce the raduis of the vessel’s swinging. The ratio of the main chain length to the water depth may affect the degree of abrasion of the main chain, but their relationship has not been clarified yet. (2) It is recommended that the tension acting on the main chain and the displacement of the floating body be derived by means of a simulation, but the data of past tests conducted under similar conditions may also be used to determine the tension and displacement. Or these may be calculated using the method described below. (3) The submerged weight of the main chain per unit length wf (t/m) can be calculated using equation (12.4.4) described in 12.4.2 [3] Diameter of Chain. Here, l and h of this equation represent the length of the main chain (lf in Fig. T-16.1.4) (m) and the vertical distance between the upper and lower ends of the main chain (hf in Fig. T-16.1.4) (m), respectively. In other words, h is the vertical distance between the floating body attachment point and the upper end of the sinker chain with the sinker being lifted up to the point where the bottom of the sinker is completely detached from the sea bottom surface. The force P represents the horizontal component (kN) of the tractive force that acts on the buoy, and q 2 and q 1 represent the inclination angles (º) of the main chain at the upper and lower ends, respectively ( q 2 ¢ and q 1 ¢ in Fig. T-16.1.4). The inclination angle q 1 ¢ of the main chain at the lower end of the chain can be calculated as exhibited in Fig. T-16.1.5 from the conditions of balance among the main chain lower end tension Tfv , the ground chain upper end tension Tg , and the sinker chain upper end tension Tsv , where Tsv is the same as the submerged weight of the sinker and sinker chain. The tension Tg and its direction are calculated in accordance with [3] Ground Chain.

Fig. T-16.1.5 Tensions of Ground Chain

(4) The tension of the main chain at its upper end should be calculated using equation (12.4.4) described in 12.4.2 [3] Diameter of Chain. Here the horizontal component of the tractive force is used as the horizontal external force. The angle q 2 ¢ that the floating body chain makes with the horizontal plane at the floating body attachment point can be calculated by means of equation (12.4.4) described in 12.4.2 [3] Diameter of Chain -506-


using the previously calculated submerged weight of the main chain per unit length. This tension is used to examine the strength of the main chain. (5) The horizontal displacement DK of the floating body can be calculated by means of equation (12.4.5) described in 12.4.2 [3] Diameter of Chain. Here q1 and q2 of the equation are defined below. q 1 : angle that the main chain makes with the horizontal plane at its lower end ( q 1 ¢ in Fig. T-16.1.4) (º) q 2 : angle that the main chain makes with the horizontal plane at its upper end ( q 2 ¢ in Fig. T-16.1.4) (º) The resultant value of displacement should be examined in comparison with the size of the mooring basin. If it is found too large, the main chain should be shortened, the weight of the sinker should be increased, or the unit length weight of the main chain should be increased.

[5] Floating Body [Technical Notes] The floating body should be designed in such a way that it does not submerge when the mooring buoy is subjected to the design tractive force. When no vessel is moored, the floating body should be afloat with the main chain (and part of the ground chain and sinker chain, as the case may be) suspended beneath it with the condition that a freeboard equal to 1/2 to 1/3 of its height must be maintained above the water surface. The buoyancy should be designed to meet these two requirements. The floating body buoyancy required to meet the requirement under the action of tractive force is calculated by means of equation (16.1.4). P (16.1.4) F = V a -------------------------lc 2 æ ----ö 1 è dø where F： required buoyancy of the floating body (kN) V=： vertical force that acts on the floating body (kN) (this is calculated by means of equation (12.4.2) described in 12.4.2 [3] Diameter of Chain). P： design tractive force (kN) lc： length of the mooring hawser (m) d： vertical distance between the vessel’s hawser hole and the water surface (m) However, the total buoyancy that is actually required is the sum of the buoyancy required to resist the tractive force and the deadweight of the floating body.

16.2 Mooring Posts (Notification Article 74) Mooring posts shall be designed in accordance with the relevant provisions given in Chapter 9 OpenType Wharves on Vertical Piles and Chapter 10 Open-Type Wharves on Coupled Raking Piles in consideration of their structural characteristics, so that the required level of structural stability and safety and smooth use of the facility will be achieved. [References] 1) Yasumasa SUZUKI: “Study on the design of single point buoy mooring”, Tech. Note of PHRI, No. 829, 1996 (in Japanese). 2) Tetsuya HIRAISHI, Yasuhiro TOMITA: “Model test on countermeasure to impulsive tension of mooring buoy”, Tech. Note of PHRI, No. 816, 1995, 18p. (in Japanese). 3) Dep. of the Navy Bureau of Yards & Docks: “Mooring Guide”, Vol. 1, 1954, 61p.

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Chapter 17 Other Types of Mooring Facilities 17.1 Quaywall of Wave-Absorbing Type 17.1.1 Principle of Design The quaywall of wave-absorbing type structure shall be designed to have the required capability of wave energy absorption and shall be located at strategetic positions in enhancing the calmness in a harbour. [Commentary] Waves within a harbor are the result of superposition of the waves entering the harbor through the breakwater openings, the waves transmitted over the breakwaters, the wind waves generated within the harbor, and the waves reflected inside the harbor. By using quaywalls of wave-absorbing type, the reflection coefficient can be reduced to 0.3 to 0.6 from that of 0.7 to 1.0 of solid quaywalls. To improve the harbor calmness, it is important to design the alignments of quaywalls and openings of breakwaters in a careful manner. The suppression of reflected waves through the provision of wave energy dissipating structures within the harbor is also an effective means of improving the calmness.

17.1.2 Determination of Structural Form Quaywalls of wave-absorbing type are structurally categorized into wave-absorbing block type and upright wave-absorbing caisson type. An appropriate structural type shall be selected in consideration of the dimansions of the quaywall and the design wave conditions. [Commentary] (1) Quaywalls of wave-absorbing block type are constructed by stacking layers of concrete blocks of complicated shapes that create hollows in between. This type is normally used to build relatively small quaywalls. The shape of blocks used varies according to design requirements. The quaywall width is determined by stability calculation as a gravity type quaywall. (2) Upright wave-absorbing caisson type quaywalls include slit-wall caisson type and perforated-wall caisson type. This type is normally used to build large quaywalls. The wave-absorption efficiency can be enhanced by optimizing the aperture rate of the front wall, the water chamber width, and others for the given wave conditions. [Technical Notes] (1) The quaywall of wave-absorbing type is generally composed of a perforated (slit) front wall, a water chamber and an impermeable rear wall. It aims to reduce the reflection coefficient through the dissipation of wave energy owing to the horizontal jet flow passing through slits or perforations in the front wall and the roughness resistance inside the structure. Thus, while extreme storm waves are used for the structural design, ordinary or frequent storm waves are employed for examination of the wave dissipating capability. (2) The reflection coefficient should be determined by means of a hydraulic model experiment whenever possible, but it may be determined in accordance with Part VII, 3.2.2 Upright Wave-Dissipating Block Breakwater and Part VII, 3.2.3 Wave-Absorbing Caisson Breakwater. Figures T-17.1.1 through T-17.1.3 show examples taken from the results of model experiments 1), 2) on slit-wall caissons and perforated-wall caissons. (3) It is recommended that the crown elevation of the wave absorption section of a wave-absorbing block type quaywall is set as high as 0.5 times the significant wave height or more above HWL (mean monthly-highest water level) and that the bottom elevation of the wave absorption section is set as deep as 2 times the significant wave height or more below LWL. (4) The water chamber width of the upright wave-absorbing caisson type quaywall may be determined in the manner same as that of the wave-absorbing block type quaywall. It is recommended however that the effects of the ceiling slab height and the air holes of the slab on the reflection coefficient are to be examined by means of a hydraulic model experiment.

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Side section

Top view

Front view

Fig. T-17.1.1 Slit-Wall Caisson Type Quaywall for Model Experiment

HWL LWL

Period

Slit bottom

WL HWL LWL

Reflection coefficient

Legend

Deep water wave steepness Fig. T-17.1.2 Relationship between Reflection Coefficient and Slit Length of Slit-Wall Caisson Type Quaywall (without filling blocks)

Fig. T-17.1.3 Reflection Coefficient of Perforated Caisson under Action of Irregular Waves 1)

17.2 Cantilever Sheet Pile Quaywall 17.2.1 Principle of Design (1) The provisions in this section shall be applied to the design of quaywalls with a cantilever sheet pile wall that retains the earth behind it. (2) The design method described in this section shall be applied for the sheet piles driven in sandy soil. [Commentary] The design method described in this section assumes that the ground comprises a sandy soil, and therefore cannot be used in the case of a clayey soil ground. It is advisable to avoid constructing a cantilever sheet pile wall on a clayey -509-


soil ground whenever possible, because there are still many unknown factors about the design of cantilever sheet pile walls on such a ground. It is also not desirable to build cantilever sheet pile walls on a clayey soil ground from engineering standpoints, because there may be adverse effects such as the deformation owing to creep. [Technical Notes] (1) Cantilever sheet pile walls resist the earth and water pressure that acts on the back of the sheet piles with the horizontal subgrade reaction force of the embedded part. The calculation of the bending moment generated in the sheet pile wall should be made in accordance with Part V, 4.3.4 Estimation of Pile Behavior Using Analytical Methods. (2) Figure T-17.2.1 shows an example of cantilever sheet pile quaywall. Mooring Bitt

Apron (Concrete pavement)

Curbing Crown height

V-type rubber fender

HWL

Pavement holding concrete work

(Crown height of steel pipe sheet pile)

Reclaimed material

LWL

Original ground surface

Backfilling cobble

Design water depth

Steel pipe type sheet pile

(Unit: m)

Fig. T-17.2.1 Example of Cantilever Sheet Pile Quaywall

17.2.2 External Forces Acting on Sheet Pile Wall The external forces acting on a cantilever sheet pile wall shall be determined in accordance with 5.2 External Forces Acting on Sheet Pile Wall. [Technical Notes]

h art ee +

re

n + ee nd w sure )a re es r e p su c ) s n e h e e t pr ssur er ear iff er e D tive at pr w c rth al (a a u id ve e s re ssi a (p t be

re

ssu

pre

re

ss

Pa

ssu

rth

ea ive

ssu

pre

Sea bottom

pre

ter

a al w

idu

tiv

res

(2) The earth pressure and residual water pressure should be calculated in accordance with 5.2.1 External Forces to Be Considered.


LWL

Ac

(1) Where the sea bottom ground is of sandy soil, a virtual bottom surface is assumed at the elevation where the sum of the active earth pressure and residual water pressure is equal to the passive earth pressure. It is assumed that the earth pressure and residual water pressure will act on the part of cantilever sheet pile wall above such the virtual bottom surface, as illustrated in Fig. T-17.2.2.

Virtual bottom surface

Fig. T-17.2.2 Determination of Virtual Surface Bottom

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17.2.3 Determination of Cross Section of Sheet Piles The maximum bending moment in a sheet pile wall shall be calculated appropriately by using an analysis method corresponding to the mechanical behavior characteristics of the wall. [Technical Notes] (1) The maximum bending moment in a sheet pile wall is normally calculated by the PHRI method concerning the lateral resistance of piles. (2) The lateral resistance of pile should be calculated in accordance with Part V, 4.3.4 Estimation of Pile Behavior Using Analytical Methods. (3) When steel pipes are used as sheet piles, the secondary stress often develops in steel pipes of a sheet pile wall due to the deformation of the steel pipe cross section (i.e., a circular cross section is deformed into an elliptic one) that is caused by the earth and residual water pressure. Cantilever sheet pile walls are the structures characterized by large displacement, and there is a risk about such walls that a relatively high secondary stress may develop in the areas around the point where the bending moment becomes maximum. The larger the diameter of the steel pipe, the higher the level of secondary stress becomes. In such a case, therefore, it is desirable to perform examination of strength against the secondary stress. The secondary stress of a steel pipe is calculated using equation (17.2.1). D 2 s = p æ ----ö ´ 2.1 ´ 10 4 è tø where s： secondary stress (N/mm2) p： earth pressure and residual water pressure acting on the sheet pile wall (kN/m2) D： diameter of pipe (mm) t： plate thickness of pipe (mm)

(17.2.1)

17.2.4 Determination of Embedded Length of Sheet Piles The embedded length of sheet piles shall be equal to or longer than the effective length of piles that is calculated in accordance with Part V, 4.3 Allowable Lateral Bearing Capacity of Piles. [Technical Notes] Because a cantilever sheet pile wall retains the earth behind the wall in the mechanism same as piles do, the embedded length of the sheet pile may be calculated in the same way as in the case of a pile. In the PHRI method for the lateral resistance of piles, the required embedded length is calculated as 1.5 lm1, where lm1 represents the depth of first zero point of the bending moment of cantilever pile. It should be noted that the embedded length calculated here is that measured not from the sea bottom surface, but that measured from the virtual bottom surface.

17.2.5 Examination of Displacement of Sheet Pile Crown A cantilever sheet pile quaywall shall be designed in such a way that the displacement of the sheet pile crown does not hinder safe and effective use of the quaywall. [Technical Notes] (1) As there is no regulation specifying an allowable displacement for the sheet pile crown of cantilever sheet pile quaywalls, the quaywall is designed in such a way that the crown displacement does not hinder the safe and effective use of the quaywall. Therefore, it is necessary to determine the acceptable level of displacement appropriately in consideration of the utilization conditions. In many cases, the upper limit values of 5 cm and 10 cm are adopted for normal and earthquake environments, respectively. (2) The sheet pile crown displacement d is expressed as the sum of the following three quantities (see Fig. T-17.2.3):

Sea bottom Virtual bottom surface

Fig. T-17.2.3 Displacement of Pile Crown

(a) Deflection of the sheet pile at the virtual bottom surface ( d 1 ) (b) Deflection of the sheet pile in the part above the virtual bottom surface ( d 2 ) -511-


(c) Sheet pile top displacement generated by the inclination of the sheet pile above the virtual bottom surface with the deflection angle of the sheet pile at the virtual bottom ( d 3 ) The deflection d 1 and displacement d 3 are generally calculated by means of the PHRI method described in Part V, 4.3.4 Estimation of Pile Behavior Using Analytical Methods. The deflection d 2 is generally considered as equal to the deflection calculated by assuming that the wall is a cantilever subjected to the earth pressure behind the wall and other forces.

17.2.6 External Forces during Construction Cantilever sheet pile quaywalls shall be designed to have sufficient safety against the external forces that act on them during construction work. 17.2.7 Detailed Design Detailed design of a cantilever sheet pile quaywall shall be carried out by referring to the provisions in 5.8 Detailed Design.

17.3 Sheet Pile Quaywall with Batter Anchor Piles 17.3.1 Principle of Design The provisions in this section shall be applied to the design of sheet pile type quaywalls that are built by coupling the sheet pile heads with the batter anchor piles driven in the ground behind the sheet piles that retain the earth in the back. [Technical Notes] (1) The design of sheet pile quaywalls with batter anchor piles is preferably made according to the procedure depicted in Fig. T-17.3.1. (2) Figure T-17.3.2 shows an example of a cross section of a sheet pile quaywall with batter anchor piles. Determination of design conditions

Provisional determination of structural dimensions (Cross section and embedded length of the sheet pile; height of the connection points of sheet pile with anchor pile; cross section and embedded length, interval,tilting angle, etc. of batter anchor pile)

Calculation of earth pressure and residual water pressure

Calculation of the acting forces (horizontal and vertical) on the connecting point between sheet pile wall and batter anchor pile. Examination of the cross sections of sheet pile and batter anchor pile Examination of embedding lengths of sheet pile wall and batter anchor pile

Examination of circular slip

Design of soil improvement works

Detailed design Fig. T-17.3.1 Flowchart of Design of Sheet Pile Quaywall with Batter Anchor Piles -512-


LWL

Sheet pile Batter anchor pile (Units: m) Fig. T-17.3.2 Example of Sheet Pile Quaywall with Batter Anchor Pile

17.3.2 External Forces Acting on Sheet Pile Wall with Batter Anchor Piles The external forces acting on a sheet pile wall with batter anchor piles shall be determined in accordance with 5.2 External Forces Acting on Sheet Pile Wall. 17.3.3 Calculation of Horizontal and Vertical Forces Acting on Connecting Point The calculation of the forces acting on the connecting point between a sheet pile and a batter anchor pile shall be made appropriately in consideration of the structural characteristics of connecting point. [Technical Notes] The horizontal and vertical forces acting on the connecting point between a sheet pile and a batter anchor pile can be calculated by assuming that the connecton is a pin structure.

17.3.4 Determination of Cross Sections of Sheet Pile and Batter Anchor Pile The maximum bending moments generated in a sheet pile and in a batter anchor pile shall be calculated by assuming the both piles as the beams that bear the concentrated loads of the horizontal and vertical forces acting on the connecting point as well as the distributed loads of the earth pressure and residual water pressure. The cross sections of the sheet pile and the batter anchor pile shall be so designed that the stresses yielded by the maximum moments in these piles do not exceed the allowable strengths. 17.3.5 Determination of Embedded Lengths of Sheet Pile and Batter Anchor Pile The embedded length of the sheet pile or batter anchor pile that is required to resist the forces acting in the axial direction as well as the direction perpendicular to the axis shall be calculated in accordance with Part V, Chapter 4 Bearing Capacity of Pile Foundations. However, the bearing capacity in the axial direction of the sheet pile and that of the batter anchor pile shall be examined through the loading and pulling tests in situ. 17.3.6 Detailed Design Detailed design of the sheet pile quaywalls with batter anchor piles shall be made by referring to the provisions in 5.8 Detailed Design and Part V, 4.5.2 Design of Joint between Piles and Superstructure.

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17.4 Sheet Pile Quaywall with Batter Piles in Front 17.4.1 Principle of Design The provisions in this section shall be applied to the design of sheet pile quaywalls that are built by coupling the sheet pile heads with the batter anchor piles driven in the ground in front of the sheet piles that retain the earth in the back. [Commentary] (1) Sheet pile quaywalls with batter piles in front are normally constructed with an open-type superstructure built in front of the sheet pile wall. There are two cases that the open-type superstructure may or may not be integrated into the sheet pile wall, but this section provides guidelines for the cases in which the superstructure and sheet pile wall are integrated. For the cases in which the open-type superstructure is not integrated into the sheet pile wall, refer to Chapter 5 Sheet Pile Quaywalls, Chapter 9 Open-Type Wharves on Vertical Piles, and Chapter 10 Open-Type Wharves on Coupled Raking Piles. (2) The design method described in this section is based on the sheet pile design with the equivalent beam method. Therefore, the structural types covered by this section are steel sheet pile walls driven into a sandy soil ground or a hard clayey soil ground. [Technical Notes] (1) The design of the sheet pile quaywall with batter piles in front is preferably made in the procedure depicted in Fig. T-17.4.1.

Determination of design conditions Provisional determination of the size of one block of the superstructure, piles, cross section of sheet piles, and layout Provisional determination of the superstrusture Calculation of the earth pressure, residual water pressure, and other external forces on the sheet pile Determination of the embedded length of sheet pile Examination of circular slip Design of soil improvement work Calculation of the support reaction and bending moment of the sheet pile Determination of sheet pile cross section

64748

Calculation of design external forces Vertical component of earth pressure Vertical forces Deadweight Static loads Live loads (truck, crane, etc.)

Layout of mooring posts and fenders Support reaction of sheet pile due to Earth pressure and residual water pressure force Horizontal forces Seismic Wind load (crane, etc.) Berthing force Tractive force

6447448

Calculation of the deadweight of superstructure, etc.

Assumptions concerning the sea bottom subsoil (gradient of slope, virtual bottom surface, kh value) Calculation of distribution of forces to each pile head Distribution of horizontal forces to batter piles Distribution of the vertical forces Calculation of axial force on each pile Calculation of bending moment at connection of sheet pile and batter pile Examination of the stresses of pile

678

Calculation of the embedded length of the pile Vertical bearing capacity Horizontal resistance Examination of the stability to the uplift on the superstructure Detailed design Fig. T-17.4.1 Flowchart for Design of Sheet Pile Quaywall with Batter Piles in Front -514-


(2) Figure T-17.4.2 shows an example of a cross section of a sheet pile quaywall with batter anchor piles in front. (Apron width) Mooring bitt Rubber fender

HWL

LWL Rubble stones

(Design sea water depth)

Steel pipe pile

Steel pipe pile

Steel pipe pile

(Units: m)

Fig. T-17.4.2 Example of Sheet Pile Quaywall with Batter Anchor Piles in Front

17.4.2 Layout and Dimensions (1) The size of one block of the superstructure and layout of piles shall be determined in accordance with 9.2.1 Size of Deck Block and Layout of Piles. (2) Dimensions of the superstructure shall be determined in accordance with 9.2.2 Dimensions of Superstructure. [Technical Notes] It is recommended that layout and inclination of the batter piles be determined in consideration of their positional relationship with other piles and construction work-related constraints such as those concerning the capacity of pile driving equipment. A pile inclination of about 20º is normally used for batter piles.

17.4.3 Design of Sheet Pile Wall Design of the sheet pile wall shall be made in accordance with Chapter 5 Sheet Pile Quaywalls. [Technical Notes] The embedded length, supporting point reaction force, and bending moment of the sheet pile may be calculated in accordance with Chapter 5 Sheet Pile Quaywalls by assuming that the joint between the batter pile and sheet pile is one of the supporting points.

17.4.4 Design of Open-Type Superstructure Design of the open-type superstructure shall be made in accordance with Chapter 9 Open-Type Wharves on Vertical Piles and Chapter 10 Open-Type Wharves on Coupled Raking Piles. [Technical Notes] (1) The deadweight of reinforced concrete of the open-type superstructure can be calculated with a unit weight of 21 kN/m2 in the design of the vertical and batter piles and sheet piles. -515-


(2) The external forces acting on the open-type superstructure should be calculated in accordance with 9.3.1 Design External Forces. The external force of the earth pressure and residual water pressure acting on the sheet pile wall may be considered as the supporting point reaction force at the connecting point between the batter pile and sheet pile. (3) The fender reaction force should be calculated in accordance with 9.3.2 Calculation of Fender Reaction Force. (4) Assumptions concerning the sea bottom should be made as specified in 9.4 Assumptions Concerning Sea Bottom Ground. Estimation of the lateral resistance and other behavior of piles can be made by Chang’s method. (5) Horizontal forces on the quaywall are assumed to be borne entirely by the batter piles. The vertical loads acting on the pile heads can be calculated as the support reaction forces under the assumption that the open-type superstructure is a simple beam supported at the positions of pile heads. The axial forces on the batter pile and sheet pile should be calculated according to equation (4.3.21) in Part V, 4.3.6 Lateral Bearing Capacity of Coupled Piles, using the horizontal force on the quaywall and the vertical load distributed to pile heads. For the axial force of a vertical pile, the vertical load distributed to the pile head may be used. (6) The bending moment at the connection of the batter pile and the sheet pile may be calculated as the moment due to the earth pressure, residual water pressure and other horizontal forces, by assuming that the batter and sheet piles consitute a rigid frame fixed at the virtual fixed point. The virtual fixed point of piles should be defined as specified in 9.5.3 Virtual Fixed Point. (7) The stresses of the piles should be calculated by means of equation (9.5.11) described in 9.5.5 Cross-Sectional Stresses of Piles, by assuming that individual piles are subjected to either the axial force only or the combination of axial force and bending moment.

17.4.5 Embedded Length The embedded lengths of the vertical, batter, and sheet piles shall be determined through the examination of the bearing capacity and lateral resistance of the piles. [Technical Notes] Examination of the embedded length with respect to the bearing capacity should be made in accordance with 9.5.6 Examination of Embedded Length for Bearing Capacity, and examination of the embedded length with respect to the lateral resistance should be made in accordance with 9.5.7 Examination of Embedded Length for Lateral Resistance.

17.4.6 Detailed Design (1) The detailed design of sheet pile quaywalls with batter piles in front shall be made by referring to the provisions in 5.8 Detailed Design and 9.9 Detailed Design. (2) The connecting point of the sheet pile wall and batter anchor pile shall be so designed that the transmission of load is made satisfactorily. (3) The open-type superstructure shall be so designed that it fully withstands the bending moment conveyed from the sheet pile wall. [Commentary] The connecting point between the sheet pile wall and batter pile should be designed by providing a sufficient margin of safety with some reinforcement, because the breakage or damage at the connecting point could lead to the collapse of the entire quaywall. The bending moment generated in the head of the sheet pile is transmitted to the open-type superstructure. Therefore, it is recommended that this bending moment be taken into consideration in the design of the superstructure.

17.5 Double Sheet Pile Quaywall 17.5.1 Principle of Design The provisions in this section shall be applied to the design of double sheet pile quaywalls. [Commentary] Double sheet pile quaywalls are earth retaining walls that consist of two rows of sheet pile walls, jointed by tie members etc., and filled in the space between the walls with sandy soil.

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[Technical Notes] (1) Figure T-17.5.1 depicts the fundamental procedure of the conventional design method of double sheet pile quaywall.

Determination of design conditions

Assumption of dimensions of the wall (wall height, wall width, embedded length, etc.)

Calculation of external forces (earth pressure, wave force, water pressure, earthquake force, etc.)

Calculation of the shear resistance of the filling material

Determination of the embedded length of the sheet pile (as a single sheet pile wall having ordinary anchorage work)

Calculation of the stresses of the sheet pile (as a single sheet pile wall having ordinary anchorage work)

Calculation of the tension of the tie member

Calculation of the bending stress of the waling

Calculation of the total stability of the wall (sliding, circular slip, etc.)

Calculation of the displacement of the wall

Detailed design (coping, partition wall, etc.)

Fig. T-17.5.1 Flowchart for Design by Conventional Method

(2) In the design of double sheet pile quaywalls, the design methods for steel sheet pile cellular-bulkhead quaywalls or sheet pile quaywalls with sheet pile anchorage have conventionally been applied. Therefore, when designing a double sheet pile quaywall with the conditions that are similar to those used in existing quaywalls, these conventional methods may be used. In the conventional methods, the displacement of double sheet pile wall is not examined. However, it is important to examine the amount of displacement when employing a double sheet pile structure for a large-sized permanent structure. As a solution to this problem, Sawaguchi 3) proposed a design method to calculate the displacement of a double sheet pile structure. Later, Ohori et al. 4) proposed a design method that examines the behavior of double sheet pile walls in a comprehensive manner, based on Sawaguchi’s method.

17.5.2 External Forces Acting on Double Sheet Pile Quaywall The external forces that act on a double sheet pile wall may be calculated in accordance with 7.2 External Forces Acting on Steel Sheet Pile Cullular-Bulkhead Quaywall. 17.5.3 Design of Double Sheet Pile Quaywall Double sheet pile quaywalls shall be designed using an appropriate method in consideration of the design conditions and the importance of the quaywall. [Technical Notes] (1) When designing a double sheet pile quaywall with the design conditions that are similar to those used for existing quaywalls, the following traditional methods may be used as appropriate: (a) The calculation to determine the width between two sheet pile walls to achieve the required strength against shear deformation is made in accordance with 7.3.2 Equivalent Width of Wall. (b) The calculation of the deformation moment is made in accordance with 7.3.3 Calculation of Deformation Moment. (c) The calculation of the resisting moment is made in accordance with 7.3.4 Calculation of Resisting Moment. However, the resisting moment due to the frictions at the joints between sheet piles of the partition walls is not considered normally. -517-


(d) The embedded length of sheet piles is determined as the longer one of either that calculated by the method for sheet piles having ordinary anchorage (refer to 5.3.2 Embedded Length of Sheet Pile) or that satisfying the allowable wall top horizontal displacement requirement (refer to 7.4 Examination of Stability of Wall Body as a Whole and 7.7 Examination of Displacement of Wall Top). In the case of an important structure, it is desirable to perform a behavior analysis beforehand, using the method by Ohori et al. or other suitable methods. (e) The bending stress of the sheet pile may be analyzed, in consideration of past construction data etc., by assuming that the sheet pile wall behaves just like an ordinary sheet pile wall with anchorage subjected to the active earth pressure by the filling material, irrespectively of the distance between the two sheet pile walls. (f) The calculation of the tension of the tie member should be made in accordance with 5.4.1 Tension of Tie Rods. (g) The design of the waling should be made in accordance with 5.5 Design of Wale. (h) A double sheet pile quaywall can be considered as a kind of gravity wall. Thus it is necessary to make examination of the stability against sliding of the quaywall and the slope stability including the quaywall, as in the case of a cellular-bulkhead type quaywall (refer to 4.3 Stability Calculations). The stability against sliding must be examined at the virtual bottom surface, which is taken at either the sea bottom or the horizontal plane at the toe of the sheet pile wall that yields the smaller factor of safety. In the former case, the resistance of the sheet pile wall below the sea bottom should be ignored. In the examination of the slope stability including the double sheet pile quaywall, the embedded length of the double sheet pile quaywall must be compared with the required embedded length calculated for a corresponding single sheet pile quaywall with anchorage. If the former is found longer than the latter, the resistance of the portion of sheet piles below the calculated toe of the latter sheet piles should be ignored against the circular slip plane passing the level below the toe. (i) The slab and upright section of the superstructure are designed in accordance with 6.6.2 Design of Relieving Platform. Foundation piles are sometimes driven into the filling material to support the superstructure. These piles should have sufficient safety against the horizontal and vertical forces transmitted from the superstructure. Here it is assumed that the vertical force transmitted from the superstructure is entirely borne by the piles, and the vertical bearing capacity of the pile is calculated by ignoring the skin friction between the pile and the filling material. The horizontal force that acts on the superstructure is transmitted to the double sheet pile quaywall partly through the piles and partly through the sheet piles. Therefore it is necessary to determine what percentages of the horizontal force are borne by the two. (2) When a double sheet pile quaywall is to be used as an important structure, it is desirable to examine its mechanical behavior by using Sawaguchi’s method 3) or the method by Ohori et al. 4), without relying on the conventional methods. Sawaguchi’s method is an approximation method that calculates displacement by assuming that the filling material is an incompressible elastic body and by examining the structure as a hybrid structure that consists of the sheet pile walls and filling material. The method by Ohhori et al. is a calculation method that makes a shear deformation analysis of the filling material by dividing the filling material into layers and applies Chang’s method for the analysis of the embedded section of the sheet pile, based on Sawaguchi’s concept. This is a computer-based numerical calculation method. The elasto-plasticity of the ground is evaluated with the shear modulus of the filling material and the subgrade reaction coefficient of the embedded section of the sheet piles, both of which are represented as the functions of deformation. [References] 1) Katsutoshi TANIMOTO, Suketo HARANAKA, Shigeo TAKAHASHI, Kazuhiko KOMATSU, Masahiko TODOROKI, Mutsuo OSATO: “An experimental investigation of wave reflection, overtopping and wave forces for several types of breakwaters and sea walls ”, Tech. Note of PHRI, No. 246, 1976, 38 p. (in Japanese). 2) Yoshimi GODA, Yasuharu KISHIRA: “Experiments on irregular wave overtopping characteristics of seawalls of low crest types”, Tech. Note of PHRI, No. 242, 1976, 28 p. (in Japanese). 3) Sawaguchi, M.: “Lateral behavior of a double sheet pile wall structure”, Soils and Foundations, Vol. 14, No. 1, 1974, pp. 45-59. 4) Kouichi OHORI, Yoshihiro SHOJI, Kunio TAKAHASHI, Hiroshi UEDA, Michihiko HARA, Yutaka KAWAI, Keisuke SHIOTA: “Static behavior of double sheet pile wall structure”, Rept. of PHRI, Vol. 23, No. 1, 1984, pp. 103-151 (in Japanese).

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PART I MOORING FACILITIES

Chapter 18 Transitional Parts of Quaywalls 18.1 Principle of Design In the design of the transitional part of quaywalls the following matters shall be taken into consideration: (1) Natural conditions in the vicinity of the transitional part (2) Conditions of the structures adjacent to the transitional part (3) Sequence of execution and availavility of construction machineries

18.2 Transitional Part Where Frontal Water Depth Varies The design conditions for the transitional part between the quaywalls at different water depths shall be the same as those for the quaywall at the deeper side, and the design method and safety factor shall be taken as specified for the structural type of respective transitional parts.

18.3 Transitional Part Where Quaywalls of Different Type Are Connected At the location where two quaywalls of different types are connected, the two structures are connected with each other either directly or with insertion of a transitional section. The design conditions for the structure of transitional section shall be selected from those of the two connecting quaywalls that are more unfavorable for the stability of the transitional section.

18.4 Outward Projecting Corner For the structure of transitional section at an outward projecting corner, the design conditions shall be selected from those for the two connecting quaywalls that are more unfavorable for the transitional section. [Commentary] Careful consideration should be made in designing the structure of a transitional section that forms an outward projecting corner, because the an outward corner may sometimes pose stability problems. It is advisable to avoid making a sharp angle corner, because it would demand a complex structural design and make construction work more difficult.

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Chapter 19 Ancillary Facilities 19.1 General Port facilities shall be provided with ancillary facilities as necessary. In particular, mooring facilities for vessels that transport dangerous goods shall have the following ancillary facilities: (1) Provision against Leakage and Spillage of Dangerous Goods Mooring facilities for vessels that transport dangerous goods shall be of such structure that the facility will not be damaged or excessively corroded in the event of a leakage accident during the handling of dangerous goods at the facility. When it is necessary to prevent direct spillage of dangerous goods into the sea, drainage facilities such as side ditches and holding basins / tanks shall be provided. Provision of equipment for preventing oil spills is subject to Article39, Clause 3 of the “Law Relating to the Prevention of Marine Pollution and Maritime Disaster” (Law No. 136 of 1970) and the associated orders. (2) Railings, Signs, and Notices Entrances to a mooring facility for vessels that transport dangerous goods shall be provided with railings or fences to prevent unauthorized people from entering the mooring facility, or with signs or notices to notify that entry into the mooring facility is restricted to authorized people. (3) Fire Fighting Equipment and Alarm Systems Mooring facilities for vessels that transport dangerous goods shall be provided with fire fighting equipment such as fire extinguishing foam ejector, fire extinguishers, and sprinklers. They shall also have alarm systems that are capable of activating audible alarm upon detection of a fire or a sign of fire and a communicating system that enables to immediately notify the people who can take appropriate measures of the fire fighting. [Commentary] (1) Ancillary facilities include: (a) (b) (c) (d) (e) (f) (g) (h)

Mooring posts, bollards, and mooring rings Fender system Curbing Stairways and ladders Water supply facilities Drainage facilities Fueling and electric power supply facilities Other facilities ① Skirt guard ② Passenger embarkation / disembarkation facilities ③ Vehicle embarkation / disembarkation facilities ④ Lifesaving facilities ⑤ Railings, ropes, etc. ⑥ Fire fighting equipment and alarm systems ⑦ Lighting facilities ⑧ Lavatories and telephones ⑨ Marker lamps and signs ⑩ Others

19.2 Mooring Equipment The mooring equipment for a mooring facility, excluding the slipways, shall have sufficient strength for the mooring of vessels in consideration of the natural conditions at the site such as the topography, weather and marine phenomena as well as the size of vessels such as the length overall, molded breadth, molded draft, and gross tonnage.

19.3 Mooring Posts, Bollards, and Mooring Rings 19.3.1 General Mooring facilities shall be equipped with mooring posts, bollards, and/or mooring rings.

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19.3.2 Arrangement of Mooring Posts, Bollards and Mooring Rings (Notification Article 80) The arrangement of mooring equipment shall be such that the mooring of vessels and the handling of cargoes can be done in a smooth and safe manner. The mooring equipment such as mooring posts, bollards, and mooring rings shall be arranged appropriately in consideration of the positions of bitts aboard the vessels. [Technical Notes] (1) Mooring posts that are used in stormy weather should be installed in the proximity of both ends of the wharf at the distance as far away as possible from the waterfront line. Bollards should be installed on the wharf and near the waterfront line, so that they can be used for mooring vessels under ordinary weather conditions and for berthing / leaving of vessels. In the determination of the bollard interval and the minimum number of bollards per berth, the values listed in Table T-19.3.1 may be used as reference. Table T-19.3.1 Intervals and Minimum Required Number of Bollards (per Berth) Gross tonnage of vessel

Maximum interval of bollards (m)

Minimum number of bollards per berth (unit)

Less than 2,000 GT

10 ～ 15

4

2,000 GT or greater, and less than 5,000 GT

20

6


25

6


35

8


45

8

(2) When a quay is a small one for small ships and there is no risk of upward pulling of mooring ropes by ships, deployment of bollards at intervals of 10 to 20 m along the waterfront line will suffice and there is no need to deploy mooring posts. For small quays, mooring rings having a strength equivalent to that of the bollards may be installed at intervals of 5 to 10 m instead of the bollards. When small quays are provided with mooring rings, they should be installed at an appropriate level in consideration of the tide level. In many cases, the mooring of small ships to mooring rings is made with one mooring rope connected to the bow and another to the stern of the ship, and therefore the mooring ring interval should be 5 to 10 m. (3) Mooring posts should normally be deployed to suit the use of the vessels being considered. Mooring posts are usually deployed in such a way that the angle between the vessel’s axis and mooring rope becomes as close to 90º as possible so that the mooring posts provide a sufficient resistance against the external forces that act on the vessel in the direction perpendicular to the vessel’s axis. The number of mooring posts installed on a berth is usually two. The bow and stern ropes are stretched to suppress the surging motion of a vessel. For this purpose, the angles between these ropes and the vessel’s axis are made small, but the angles should not be smaller than 25°to 30° . Installment of bollards becomes necessary to satisty this requirement. Figure T-19.3.1 shows typical arrangements of mooring posts and bollards. Stern rope

Bow rope

Bollard

Mooring post (a) When the angles at which the mooring ropes from the mooring posts make with the vessels axis are 90

Bollard

Mooring post (b) When the angles at which the mooring ropes from the mooring posts make with the vessels axis are 45

Fig. T-19.3.1 Typical Arrangements of Mooring Posts and Bollards

19.3.3 Tractive Force of Vessel The tractive forces of vessels shall be calculated appropriately in consideration of the berthing and mooring conditions of vessels.

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[Technical Notes] The tractive forces of vessels should be calculated in accordance with Part II, 2.2.4 Tractive Force Acting on Mooring Post and Bollard.

19.3.4 Structure (Notification Article 78) Mooring post and bollards shall be structurally safe against the vessel’s tractive forces that act on them. [Commentary] The tractive forces of vessels that act on mooring posts and bollards should be determined in accordance with Part II, 2.2.4 Tractive Force Acting on Mooring Post and Bollard. [Technical Notes] The structure of mooring post may be determined in accordance with the “Standard Design of Mooring Post” of “Standard Design of Port Structures (Issue 1)” published by the Japan Port Association.

19.4 Fender System 19.4.1 General The layout, structure and others of the fender system of a mooring facility shall be designed appropriately in consideration of the berthing forces of vessels and the forces generated by the motions of moored vessels. [Technical Notes] (1) When a vessel is berthed to a wharf or when a moored vessel is put into motion owing to wind and wave forces, the berthing force and friction force are generated between the vessel and the mooring facility. To prevent damages to the vessel’s hull and mooring facility due to these forces, a fender system is installed on the mooring facility. However, when the vessel has a fender system such as ship fenders or tires (e.g., small ships, certain types of ferries) and the maneuvering of such a vessel is done very carefully in consideration of the fender system’s energy absorption capacity, the mooring facility does not necessarily have to be equipped with a fender system, because the berthing force is relatively low.

Determination of target vessel Layout of fenders (During berthing) Determination of the coefficient of eccentricity and approaching velocity

(When a vessel is moored) Determination of the layout and type (characteristics) of mooring ropes

Calculation of the berthing energy

Determination of the wave conditions, wind speed, current velocity etc.

Assumption of the type and shape of fender

Assumption of the type and shape of fender

Calculation of the absorbed energy, reaction force, and deformation of the fender

Calculation of the vessel motion, fender deformation and reaction force Is the fender deformation within the allowable deformation? Isnt the fender deformation too small ?

(2) For fender systems, rubber and pneumatic fenders are popularly used. Other types such as foam types, water pressure types, oil pressure types, suspended weight types, pile types, and timber types are also used 2).

Determination of the fender to be used Fig. T-19.4.1 Flowchart of Fender Design

(3) Fender systems should be provided with contact panels etc. as necessary to reduce the load per unit area (bearing or surface pressure) in order to prevent the damage to the vessel’s hull. To reduce the shearing force that acts on the fender systems, it is recommended to attach epoxy resin plates etc. to the surface of the contact panel. (4) Design of rubber fenders, pneumatic fenders, and pile type fenders should basically be conducted in the procedure shown in Fig. T-19.4.1. -522-


19.4.2 Arrangement of Fenders (Notification Article 81) Fenders shall be arranged in such a way that the fenders can absorb the required portion of the berthing energy of the target vessel and the vessel will not come in direct contact with the mooring facility 3 ) ,4 ) . 19.4.3 Berthing Energy of Vessel (Notification Article 82) In principle, the berthing energy of vessel that is used in the design of fenders shall be calculated in accordance with Part II, 2.2 External Forces Generated by Vessels. 19.4.4 Selection of Fender (Notification Article 83) In principle, fenders shall be selected appropriately in consideration of the following: (1) The characteristics of the structure of the mooring facility and the vessels to berth there. (2) In the case of a mooring facility being exposed to wave actions, the motions of moored vessels and the berthing conditions of vessels including the berthing angles. (3) The exertion of the fender reaction force generated during vessel’s berthing upon the structural members of the mooring facility. (4) The ranges of variation in the physical characteristics of fenders due to the deviations of the manufactured products from the standard, the response to dynamic loading, and the thermal response to the air temparature, etc. [Technical Notes]

Absorbed energy

(1) The berthing force is calculated from the energy absorption curve that is obtained by adding the energy absorbed by the deformation of fenders Ef1 and that absorbed by the deformation of the berth structure Ef2, as shown in Fig. T-19.4.2.

(Berth Structure) + (Fenders) Fenders

Berth Structure

Reaction force (load)

Fig. T-19.4.2 Calculation of Berthing Force

(2) Gravity type quaywalls, sheet pile quaywalls, and quaywall with relieving platform have sufficient levels of resistance capacity against normal berthing forces. However, flexible structures such as open-type wharves, detached piers, and dolphins have relatively low horizontal resistance capacity (especially in the case of a vertical pile type mooring facility). Therefore, care should be taken to ensure that the berthing force becomes smaller than the allowable horizontal load when using these types of structures. (3) A fender system should also be safe against the shearing force due to the friction between the fender and the vessel’s hull generated by oblique berthing of vessels. This force is normally calculated by means of the equation suggested by Vasco Costa 5). When a vessel is berthed to the wharf at an angle of 6°to 14°with the berth’s front line, this force becomes 10% to 25% of the fender reaction force. (4) In mooring facilities that are exposed to wave actions, vessels move in both the horizontal and vertical directions. The vessel’s motions may cause excessive shear deformation in fenders in addition to the normal compressive deformation, which sometimes leads to breakage of fenders. If the shearing force is assumed to be the friction force, the force is estimated as about 30% to 40% of the fender reaction force. (5) The single pile structure (SPS) is a type of structure that absorbs the berthing energy by the deformation of piles made of high tensile strength steel. In the design of dolphins that use SPS, it is necessary to calculate the amount of energy absorption by considering the residual deformation of the piles due to repeated berthing. As shown in Fig. T-19.4.3, the amount of energy absorbed by piles is calculated by using the displacement obtained by subtracting the residual displacement from the loading point displacement. -523-


Load

The loading point displacement that takes the residual displacement into consideration is calculated by means of equation (19.4.1). (19.4.1) Ph 3 y top = A 1 y 0 + A 2 i 0 h + --------3EI where ytop： displacement of the pile at loading point, considering residual displacement (m) y0： pile displacement at sea bottom at the time of initial Absorbed energy loading (m) i0： pile deflection angle at sea bottom at the time of initial loading (rad) P： horizontal load (N) h： height of loading point (m) Residual EI： flexual rigidity of pile (N・ m2) displacement displacement A1, A2： influence coefficient due to repeated loading Note: The term “initial loading” refers to the first application of a load that is the largest among the past loading. Table T-19.4.1 lists the values of influence coefficient due to repeated loading that have been suggested on the basis of the result of an on-site full-scale loading experiment and the result of a model experiment 7).

Fig. T-19.4.3 Energy Absorbed by Deformation of Piles

Table T-19.4.1 Values of Influence Coefficient Due to Repeated Loading For obtaining the maximum displacement

For obtaining the energy absorbed by the deformation of piles

For obtaining the residual displacement

A1

1.4

0.4

0.8

A2

1.2

0.6

0.5

(6) Energy Absorption by Fenders There are various types of rubber fenders such as V-shaped rubber fenders, circular hollow rubber fenders, and rectangular hollow rubber fenders. Each of these types differs from others in terms of the relationship between the reaction force and deformation as well as the energy absorption rate. Manufacturers’ catalogs show diagrams of the amount of energy absorption versus deformation, and those of the reaction force versus deformation for each type of fenders. Constant-reaction force fenders such as V-shaped fenders are characterized with low-reaction forces and high energy absorption rates. It should be borne in mind however that the total reaction force to the mooring facility may become large when a vessel comes in contact with two to three fenders simultaneously. This is because of the fact that the reaction force level rises nearly to the maximum value when the energy absorption rate reaches to 1/3 of the design capasity. (7) Variation in Characteristics of Rubber Fenders Factors that cause variations in characteristics of fenders include the product deviations from the standards, aging in quality, dynamic characteristics (velocity-dependent characteristics), creep characteristics, repetition characteristics (compression frequency-dependent characteristics), oblique compression characteristics, and thermal characteristics. In the design of fenders used to moor floating structures, these factors are important in the evaluation of the safety of the mooring system. In the design of fenders for wharves, it is also appropriate to give consideration to the product deviations, dynamic characteristics (velocity-dependent characteristics), oblique compression characteristics, thermal characteristics, etc. For example, when the product deviation (tolerance) of the fender is ±10%, it is desirable to employ the energy absorption characteristics lowered by 10% from the catalog value and to use the reaction force characteristics raised by 10% from the catalog value in the design of the fenders and mooring facility. With regard to dynamic characteristics, it is recommended to confirm that the reaction force of the fender at the time of berthing will not exceed the standard value shown in the catalog in consideration of the berthing velocity of vessels. It should also be borne in mind that the fender reaction force becomes higher in a low-temperature environment than in the standard temperature environment. It has been recommended by a working group of the Permanent International Association Navigation Congress (PIANC) to perform correction on the energy absorption and reaction force in the selection of fender, in order to reflect changes in characteristics due to the environment in which the fender is used, such as the vessel’s approaching velocity, the temperature, etc8). Actual values of these correction coefficients should be checked with the manufacturer, as they vary depending on the approaching velocity, temperature, and type of rubber used for the fender. It should also be borne in mind that the reaction force conveyed to the berth structure may become larger when a small ship is berthed at a high approaching velocity than when a large vessel is berthed at a low approaching velocity 6), 9). -524-


19.5 Safety Facilities 19.5.1 General The mooring facilities shall be provided with a skirt guard or other safety equipment to hold small ships from moving under piled piers. Safety equipment other than the skirt guard includes fences, ropes or similar equipment, and signs showing safety passage for pedestrians. Installation of such equipment shall be made as specified below. 19.5.2 Skirt Guard (Notification Article 84) When there is a risk that small ships may accidentally underrun a mooring facility such as piled piers and dolphins, an appropriate skirt guard shall be provided. [Commentary] At a piled pier, open-type wharf, dolphins or any other mooring facilities where small ships might go below the deck, a skirt guard should be provided.

19.5.3 Fence and Rope (Notification Article 85) Mooring facilities for passenger ships shall be provided with fences or ropes as necessary, to ensure the safety of passengers, to secure the pathways for passengers, and to prevent vehicles from entering the facility. [Technical Notes] Usually a standard height of 0.7 m is used for fences and ropes. For areas with a risk that passengers may fall off, fences having a height of 1.1 m or more should be provided.

19.5.4 Signs or Notices Refer to 19.11 Signs, Notices, and Protective Fences. 19.5.5 Curbing Refer to 19.9 Curbing. 19.5.6 Fire Fighting Equipment and Alarm Systems The provision of fire fighting equipment should be in conformity with the “Fire Service Law”, the “Orders on the Implementation of the Fire Service Law”, and the “Implementation Rules for the Fire Service Law”.

19.6 Service Facilities 19.6.1 General The mooring facilities shall be provided with lighting appliances and other equipment required for facilitating their usage. Here, “other equipments required” includes facilities for embarking and desembarking passengers and vehicles, water supply facilities, drainage facilities, fueling facilities, electric power supply facilities, marks and signs, lavatories, telephones, and others. Installation of these facilities shall be made as specified in the subsections hereinafter. 19.6.2 Lighting Facilities Refer to 19.12 Lighting Facilities. 19.6.3 Facilities for Passenger Embarkation and Disembarkation Refer to Part Ⅸ , Chapter 5 Facilities for Passenger and Part Ⅹ , Chapter 2 Ferry Terminals.

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19.6.4 Vehicle Ramp (Notification Article 93) The requirements for vehicle ramps are as follows: (1) The width of a vehicle ramp shall be an appropriate value that is equal to or greater than the value listed in Table 19.6.1. In the case of a movable bridge, the structural characteristics of the bridge shall be taken into consideration in an appropriate manner. (2) The inclination of a vehicle ramp shall be an appropriate value that is equal to or smaller than the value listed in Table 19.6.1. (3) The standard length of the horizontal part shall be 7 m. In the case of a small-scale facility, a length of 4 m may be used as a standard. (4) The radius of the centerline of a curved section shall be an appropriate value equal to or larger than 15 m. (5) Signs and notices shall be provided appropriately in consideration of the structural characteristics and use of the facility. Table 19.6.1 Width and Inclination of Vehicle Ramp (Notification Article 93) Number of lanes

Width (m)

Ramp mainly used by vehicles with a width of 1.7 m or less (small-scale facility)

1

3

2

5

Ramp mainly used by vehicles with a width of 2.5 m or less

1

3.75

2

6.5

1

4

2

7

Type of facility

Ramp frequently used by container trailers

Inclination (%) Fixed section

Movable section

12

17

10

12

－

－

Note: The inclination of a facility that is to be frequently used as a ramp for container trailers shall be determined appropriately in consideration of the safety of container trailers during embarkation / disembarkation and the state of their traffic on the facility.

[Commentary] The term “small-scale facility” refers to a vehicle ramp dedicated for small-sized vehicles and light vehicles. [Technical Notes] In the design of small-scale and standard-size facilities, the range of vertical movement at the tip of the movable part is often set as equal to the spring tidal range (the difference between the mean monthly-highest water level and the mean monthly-lowest water level) with an addition of 1 m.

19.6.5 Water Supply Facilities (Notification Article 88) A mooring facility shall be provided with a water supply facility as necessary in consideration of the size and use of the mooring facility. The provisions specified in Part Ⅸ , 4.2 Water Supply Facilities shall apply as appropriate to the provision of the water supply facility for the mooring facility. 19.6.6 Drainage Facilities (Notification Article 87) A mooring facility shall be provided with a drainage facility (such as drainage ditches and water drain work) as necessary in consideration of the quality of the waste water, the structural characteristics and use of the mooring facility. 19.6.7 Fueling and Electric Power Supply Facilities (Notification Article 89) A mooring facility shall be provided, as necessary, with fueling and/or electric power supply facilities that enable safe and efficient fueling and power feeding, in consideration of the size of vessels to moor at the facility, type and scale of the cargo handling, and structural characteristics of the mooring facility.

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19.6.8 Signs or Notices Refer to 19.11 Signs, Notices, and Protective Fences.

19.7 Stairways and Ladders (Notification Article 90) The requirements for stairways, ladders, and other similar facilities toward the water surface shall be as follows: (1) They shall be located in such a way that they can be used easily in the event of an emergency. (2) They shall have a structure that enables safe usage. [Commentary] It is recommended that mooring facilities to be used for the embarkation and disembarkation of passengers of ferries and passenger ships are to be provided with one or more stairways, ladders, or other similar facility per berth. [Technical Notes] Ladders should be designed with a design load of 1 kN per meter of ladder length for both horizontal and vertical directions. Metal fixtures should have especially high strength. Provision should be made to enable easy repair of the ladder in the event of its breakage, corrosion, etc.

19.8 Lifesaving Facilities (Notification Article 91) Mooring facilities that are to be used by a passenger ship of 500 tons or more in gross tonnage shall be provided with appropriate lifesaving facilities as necessary.

19.9 Curbing (Notification Article 92) The structure, shape, layout, and material of curbing shall be designed appropriately in such a way that the safety of users in vehicles will be ensured and cargo handling work will not hindered, in consideration of the use and structural characteristics of the mooring facility. [Technical Notes] The designing of curbing can be conducted by refering to the “Curving Design Manual”.

19.10 Vehicle Ramp A vehicle ramp shall be designed in accordance with Part Ⅹ , 2.3 Design of Vehicle Ramp.

19.11 Signs, Notices and Protective Fences 19.11.1 General Mooring facilities shall be provided with the following signs and notices or protective fences specified in the following subsections as appropriate. 19.11.2 Provision of Signs In order to ensure the safety of port users and to make for convenient use of ports, signs shall be provided in the following cases: (1) When it is necessary to ensure that port users can arrive at their destinations in a speedy and safe manner and to provide guideboards of the location of port facilities. (2) When it is necessary to warn port users about dangers associated with the use of facilities and cargo handling works. (3) When it is necessary to provide instructions to port users about methods to use facilities and guide them to ensure safe and smooth use of facilities. (4) When it is necessary to regulate the behavior of port users to ensure their safety and smooth activities, to prevent disasters such as fire and falling accident, and to prevent environmental pollution by littering.

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19.11.3 Types and Location of Signs (1) Signs shall basically be the same in type as those used for ordinary roads and be designed appropriately in consideration of their purposes in such a way that port users can easily recognize them. (2) Signs shall be installed at locations where they are easily recognized by port users and they do not hinder the traffic and the cargo handling works. [Technical Notes]

6.0

h

An example of a sign is shown in Fig T-19.11.8.

h: Character size n: Number of characters

n

h

Fig. T-19.11.8 Example of Sign

19.11.4 Position of Sign Signs shall be installed in accordance with the following guidelines: (1) Horizontal Position In principle, signs installed on roads shall be positioned in such a way that one edge of the sign plate is aligned with the horizontal boundary of road clearance limits. (2) Height Signs shall be installed at an appropriate height at which users can easily recognize them (as in the case of signs on ordinary roads) in consideration of their purposes. [Technical Notes] (1) Figure T-19.11.9 shows an example of the road clearance limits. (2) Figure T-19.11.10 shows the height requirements for road signs.

Road clearance limits

Side walk clearance limit 0.25m

4.5m

2.5m

Road facility strip

Median strip

Fig. T-19.11.9 Clearance Limits for Roads

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2.8 m 1.0

5.0 m


Single-post type

1.0 (b) Warning

1.0 m or more

Dual-post type (a) Information

1.0 m or more

2.1 m

Overhang type

(c) Instruction

(d) Regulation

Fig. T-19.11.10 Height of Signs

19.11.5 Structure of Sign Signs shall be designed in such a way that the post(s), foundation work, and sign plate have sufficient structural strengths in consideration of the following: (1) The external forces that act on the sign shall be the deadweight, wind load, and earth pressure. (2) The calculation of the wind load and the design of the foundation work shall be made in accordance with the “Guidelines for Construction of Foundation of Facilities Ancillary to Roads” (Public Notice from the Chief of the Planning Section, Road Construction Burean, Ministry of Construction; Road Construction Planning Notice No.52, July 15, 1975). (3) The posts shall be designed in such a way that they have sufficient strength and are aesthetically acceptable, in consideration of the size of the sign plate and the place at which the sign is installed. (4) The thickness of the sign plate shall be designed appropriately in consideration of the size of the plate so that the sign will be kept discernible for a long time free from excessive distortion or deformation. The sign plate shall be provided with appropriate stiffeners. (5) Movable signs shall be designed in such a way that they can be moved easily and have a sufficiently stable structure. [Technical Notes] (1) The wind load is calculated by means of the following equation: 1 P = --- raV 2 CA 2 where P： wind load (kN) V： design wind speed (m/s) A： area subjected to wind (effective projected area) (m2) C： drag coefficient The following values should be used as standard: post： 0.7 plate：1.2 ra : density of air, r a = 1.23 × 10-3 (t/m3) For the design wind speed, the following values should be used as standard: -529-

(19.11.1)


Single-post type small signs：40 m/s Suspended large signs： 50 m/s The wind load should be treated as a short-term load. (2) Thickness of Sign Plate Recently aluminum alloys are frequently used for the base board of sign plates. Normally the following thickness are used for the base board according to the size of the sign plate. (a) Small sign plates (smaller than 1.0 m2)： 1.2 mm (b) Large sign plates (1.0 m2 or larger)： 2.0 mm

19.11.6 Materials (1) In principle, signs shall be made using the following materials: (a) Sign plate: aluminum alloy plate or synthetic resin plate (b) Post: steel pipe or aluminum pipe (c) Metal fixtures: aluminum alloy of extruded shapes or steel (2) All steel products being used shall be provided with adequate corrosion protection treatment on their surface. (3) When a reflector is to be used, a reflector sheet shall be used. (4) All materials shall meet the applicable JIS requirements, or they shall be of equivalent quality. [Commentary] The JIS requirements for aluminum alloy plates, steel products, etc. are as follows: Aluminum alloy plates: JIS H 4000 “Aluminum and Aluminum Alloy Plates” Aluminum pipes: JIS H 4100 “Aluminum and Aluminum Alloy of Extruded Shapes” Steel pipes: JIS G 3444 “Carbon Steel Pipes for General Structural Purposes” Reflector sheets: JIS Z 9117 “Reflector Sheets and Tapes Used for Safety Purposes” [Technical Notes] There are a variety of reflector sheet products available, but the most commonly used ones are agglutinative materialencapsulated lens-type reflector sheets and air-encapsulated lens-type reflector sheets. Figure T-19.11.11 shows the structure of these types. Air-encapsulated lens-type

Agglutinative material-encapsulated lens-type

Transparent plastic film Agglutinative material

Transparent plastic film Air layer

Agglutinative material (spacer)

(cushion coat)

Spherical lenses

Reflection film Adhesive layer Liner

Fig. T-19.11.11 Major Types of Reflector Sheet and Their Structure

19.11.7 Maintenance and Management Signs shall be maintained and managed in such a way that they are always kept in good condition. 19.11.8 Protective Fences Protective fences shall be provided in such a way that they can sufficiently ensure the safety of facility users, in consideration of the conditions of the area in question. The design, layout and others of protective fences may be determined in accordance with the “Guidelines for the Provision of Protective Fences”.

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19.11.9 Barricades Barricades shall be of such structure that they can be easily shifted and are resistant to damage during shifting.

19.12 Lighting Facilities 19.12.1 General (Notification Article 86, Clause 1) Appropriate lighting facilities shall be provided at the wharves and related areas where cargo handling works such as loading, unloading and transfer, berthing / leaving of vessels, and use by passangers and others are made at night, in consideration of the state of usage of the mooring facility. [Commentary] (1) The provisions in this section apply to the installation, improvement, and maintenance of lighting facilities for wharves that are used by passengers and other users at night or wharves at which cargo handling work, berthing / leaving of vessels, or other work is performed at night. (2) For water frontage amenity facilities and areas for recreational purposes such as amenity-oriented revetments, marinas, parks, and promenades, provision of appropriate lighting facilities should be considered as necessary for the proper function(s) of the facility. [Technical Notes] In response to recent social demands concerning improvement of lighting facilities for port facilities, there have been introduced many lighting facilities that are designed to highlight night scenery of buildings, parks, and waterfronts, especially in the areas in and around cities and sightseeing spots. These lighting facilities are designed to provide not only illumination, but also a good light color and color rendering performance that create serene, relaxing, and enjoyable atmosphere in the place. On the other hand, needs have arisen to reconsider the policy for the provision of lighting facilities from the viewpoints of preventing adverse effects of artificial light on environments of nearby areas and promoting energy conservation as the number of illuminated facilities / areas has increased. Thus lighting facilities for marinas, parks, etc. should be planned and designed appropriately on a case-by-case basis in consideration of these needs as well as the required lighting functions.

19.12.2 Standard Intensity of Illumination [1] Definition The term “standard intensity of illumination” is defined as the average intensity of illumination at a specified standard horizontal surface, and it refers to the minimum intensity of illumination that is required to use the facility in question in a safe and efficient manner. [Commentary] (1) The most common measure of brightness in the design of lighting facilities is the intensity of illumination. The term “average intensity of illumination at a specified standard horizontal surface” refers to the average value of the intensity of illumination as measured on the floor or ground surface. (2) Because this standard intensity of illumination designates the minimum value required, a higher intensity may be used as necessary. (3) Standards on intensity of illumination are being developed by the International Committee on Illumination (CIE), which has already published “A Guide to Design of Illumination for Outdoor Workplaces”.

[2] Standard Intensity of Illumination for Outdoor Lighting (Notification Article 86, Clause 2) The intensity of illumination for outdoor lighting facilities shall be designed appropriately in accordance with the requirements specified in the JIS “Requirements for Intensity of Illumination” or other relevant requirements in consideration of the type and style of the work, so that safe and smooth use of the facility is ensured. [Commentary] For intensity of illumination for outdoor facilities, the values listed in Table C-19.12.1 should be used as standard, in consideration of the type of the facility.

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Table C-19.12.1 Standard Intensity of Illumination for Outdoor Lighting Facility

Apron

Wharf

Yard

Passageway Safety and security Roads Road and park

Parking space Greens zone

Standard intensity of illumination (lx)

Mooring facilities for passenger or vehicles, general cargo berths, and container berths

50

Slopes for pleasure-boats, and aprons where dangerous cargo is handled by pipelines

30

Aprons in which simple cargo handling work is performed using pipelines and belt conveyers

20

Yards used to store, load and unload, and transport containers and general cargoes

20

Boarding bridge for passengers and vehicles

75

Passageways for passengers and vehicles

50

Other passageways

20

All facilities

1~5

Major roads

20

Other roads

10

Parking area for ferries

20

Other parking area

10

Passage in green zone

3

[3] Standard Intensity of Illumination for Indoor Lighting (Notification Article 86, Clause 2) The intensity of illumination for indoor lighting facilities shall be designed appropriately in accordance with the requirements specified in the JIS “Requirements for Intensity of Illumination” or other relevant requirements in consideration of the type and style of the work, so that safe and smooth use of the facility is ensured. [Commentary] For intensity of illumination for indoor facilities, the values listed in Table C-19.12.2 should be used as standard, in consideration of the type of the facility. Table C-19.12.2 Standard Intensity of Illumination for Indoor Lighting Facility Passenger terminal

Shed and warehouse

Intensity of illumination (lx)

Waiting rooms

300

Passengers’ passage and boarding bridge.

100

Fish sorting sheds at berths for fishing boats.

200

Container freight stations and transit sheds used by vehicles only.

100

Sheds and warehouses used for simple cargo handling work

70

Other sheds and warehouses

50

19.12.3 Selection of Light Source (1) Light source for wharf lighting shall be selected in such a way that the following requirements are met: (a) (b) (c) (d)

The light source shall be of a high efficiency and long service life. The light source shall be stable against the variations of ambient temperature. The light source shall provide a good light color and good color rendering performance. The time of the stabilization of the light after turning-on shall be short.

(2) Any light source other than a light bulb shall be used together with an appropriate stabilizer.

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[Commentary] (1) Types of Light Source Light sources are classified by type as shown in Fig. T- 19.12.1.

Incandescent lighting

Light source

Electrical discharge lighting

Light bulbs Tungsten halogen lamps Low-pressure discharge lamps High-pressure discharge lamps

HID lamps (High-luminance lamps)

Fluorescent lamps Low-pressure sodium vapor lamps Mercury lamps (transparent-type, fluorescence-type) Metal halide lamps High-pressure sodium vapor lamps

Fig. T-19.12.1 Types of Light Source

[Technical Notes] (1) Table T-19.12.1 lists the characteristics of various lamps. Table T-19.12.1 Characteristics of Lamps Characteristics Lamp efficiency (lm/w)

Light color (k)

Color rendering performance (Ra)

Service life (hours)

Type of lamp

Stability against influences of ambient temperature

Activation

Reactivation

Dimming

Incandescent lamp

Low 15 to 20

Whitish orange 2,800

Good 100

Short Stable 1,000 to 2,000

Instantaneous

Instantaneous

Easy

Halogen lamp

Low 17 to 22

Whitish Good orange 100 3,000 to 3,200

Short Stable 1,000 to 2,000

Instantaneous

Instantaneous

Easy

Fluorescent lamp (white)

Intermediate 80 to 100

Relatively White good 3,000 to 4,000 50 to 95

Long 6,000 to 12,000

Influenced

Quick Quick Possible 2 to 3 seconds 2 to 3 seconds

Orange 1,700

Intermediate 9,000

Stable

20 minutes

Relatively quick 10 seconds

Difficult

Stable

8 minutes in normal temperature environment

Relatively slow 10 minutes or less

Possible up to 50%

Difficult

Low-pressure sodium Highest lamp 100 to 180

Bad

Mercury vapor lamp

Relatively low (Bluish) white Intermediate 40 to 60 3,500 to 4,000 40 to 50

Long 9,000 to 12,000

Metal halide lamp


White Good 4,000 to 6,500 70 to 90

Influenced to Intermediate a certain 6,000 to 9,000 extent

5 minutes in normal temperature environment

Relatively slow 10 minutes or less

Whitish orange 2,100

Long 9,000 to 12,000

5 to 10 minutes

Relatively Possible up to quick 50% 1 to 5 minutes

High-pressure sodium Intermediate vapor lamp 60 to 120


Stable

Note: The figures for lamp efficiency, light color (k), average color rendering performance evaluation number (Ra), etc. were determined based on “A Guide to Design of Illumination for Indoor Workplaces (Table A.3.1)” published by the Japan Illumination Association.

(2) Color Temperature and Warm / Cool Sensation Color temperature (K) is a representation of light color in the form of numerical figure and directly related to human being’s warm / cool sensation. Light color becomes more reddish as the color temperature becomes lower, and more palish as the color temperature becomes higher. Table T-19.12.2 lists the relationship between the color temperature and warm / cool sensation. Table T-19.12.2 Color Temperature and Warm / Cool Sensation Color temperature (K)

Warm/cool sensation

3,300 or less

Warm

3,300 to 5,300

Intermediate

5,300 or more

Cool

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(3) Color Rendering Performance and Average Color Rendering Performance Evaluation Number As a main index of color rendering performance, an “average color rendering performance evaluation number” (Ra) is used. The number Ra is the average of the color rendering performance evaluation numbers obtained for the eight designated test colors. The CIE (International Commission on Illumination) has defined limits of application of various types of lamps for outdoor workplaces based on Ra. Table T-19.12.3 lists the color rendering performance classes together with the corresponding average color rendering performance evaluation numbers and lamp types, as well as the recommended use of each type. Table T-19.12.3 Color Rendering Performance Classification for Outdoor Lighting Color rendering performance class

Average color rendering performance evaluation number (Ra)

1

80 ≦ Ra Very good

2

60 ≦ Ra < 80 Good

3

40 ≦ Ra < 60 Relatively good

4

20 ≦ Ra < 40 Acceptable

High-pressure sodium vapor lamp

※

Ra < 20

Low-pressure sodium vapor lamp

Type of lamp

Use

Incandescent lamp Fluorescent lamp Metal halide lamp High color rendering performance-type high-pressure sodium vapor lamp

Work that involves selection of hues

Mercury lamp Ordinary work

Not suitable for work that involves selection of colors

※ Not recommended

19.12.4 Selection of Lighting Equipment [1] Outdoor Lighting Lighting for outdoor illumination shall be selected in consideration of the following requirements: (1) Lighting equipment shall be rainproof. When a large amount of flammable dangerous goods is to be handled in the proximity of the lighting equipment, lighting equipment shall be explosion-proof. (2) Materials for the lamp, reflector surface, and illumination cover shall be of good quality and have high durability and good resistance against deterioration and corrosion. (3) Sockets shall be of appropriate type for the respective light source. (4) Stabilizers and the internal wiring shall be capable of withstanding the expected increase in the temperature of the equipment. (5) Lighting equipment shall be of high-efficiency type. (6) Luminous intensity distribution shall be controlled appropriately in consideration of the use of the equipment. [2] Indoor Lighting Lighting for indoor illumination shall be selected in consideration of the following requirements: (1) Luminous intensity distribution shall be controlled appropriately in consideration of the use of the equipment. (2) Sockets shall be of appropriate type for the respective light source. (3) Stabilizers and the internal wiring shall be capable of withstanding the expected increase in the temperature of the equipment. (4) Lighting equipment shall be of high-efficiency type.

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19.12.5 Design of Lighting In the designing of lighting, the layout of lighting equipment shall be determined by considering the items listed below for the lighting method, light source, and equipment selected, in consideration of the characteristics of the area where the equipment is to be installed. Those equipment whose influence area extends to the sea shall be deployed in such a way that they do not hinder the navigation of nearby ships. (1) Standard intensity of illumination (2) Distribution of intensity of illumination (3) Glare (4) Adverse effects of light and energy conservation considerations (5) Light color and color rendering performance [Commentary] (1) Layout of Lighting Equipment In the design of lighting, the lighting equipment should be layed out in such a way that the standard illumination intensity requirements described in this 19.12.2 Standard Intensity of Illumination are sufficiently met after examining the lighting method, distribution of intensity of illumination, glare from the lighting equipment, and light color and color rendering performance of the light source. (2) Distribution of Intensity of Illumination Improper distribution of intensity of illumination at the irradiated surface not only makes passengers and workers uncomfortable, but also creates dark areas in which people cannot see objects or other people clearly. This may lead to reduction in work efficiency or accidents. In the design of lighting, therefore, it is necessary to give careful consideration to the following: (a) In the determination of the layout of lighting equipment, care should be taken to achieve an appropriate ratio between the installation interval and height of the lighting equipment so that proper distribution of intensity of illumination will be achieved. (b) When an area is likely to be shadowed by plants or cargoes, ancillary lighting equipment should be provided. (c) The Illumination Guide issued by the CIE lists the mean values that should be maintained for the intensity of illumination at the floor or ground as well as the recommended values for degree of evenness. The degree of evenness here is defined as the ratio of the minimum intensity of illumination to the average intensity of illumination. (3) Glare The term “glare” refers to an excessively bright light or light with excessively uneven brightness that causes temporary discomfort or a reduction in eyesight to people. Glares caused by lighting equipment are categorized into “glares that affect ships” and “glares that affect passengers and workers”. (a) Glares that affect ships When a ship crew member or pilot is affected by a glare, his ability to recognize navigational aids and anchored ships is impaired. This may lead to ship maneuvering errors, which in turn may cause accidents such as collision with other ship or a quaywall. Therefore, the layout and luminous intensity distribution of lighting equipment should be carefully examined to ensure safe navigation of ships. (b) Glares that affect workers and passengers When a worker or passenger is affected by a glare, his ability to recognize cargoes, tags, and obstacles is impaired. This may lead to reduction in work efficiency and unnecessary fatigue. Therefore, arrangements should be made to prevent direct exposure of people’s eyes to light in consideration of the height of the eyes of workers and passengers and the position of the lighting equipment. The CIE has recently introduced the upper limits for glares in their lighting standards for outdoor workplaces as recommendations to ensure the efficient visual works of workers and the safety of traffic. (4) Adverse Effects of Light and Energy Conservation Considerations Leakage of light from outdoor lighting facility produces adverse influences that obstruct astronomical observation, cause glares hindering visual recognition of objects, and deteriorate ecosystems of animals and plants. In addition, leakage of light results in a loss of energy. As these are problems that have social implications, it is important to give careful consideration to these problems in the design of lighting. (5) Light Color and Color Rendering Performance Light sources for recreational facilities such as amenity-oriented revetments, promenades, and parks should be selected in consideration of the suitability of light colors for the facility. In the case of a facility with high intensity of illumination, it is desirable to give careful consideration to color rendering performance in the selection of the light source. -535-


(6) Illumination Intensity Calculation Method There are two illumination intensity calculation methods, namely, the flux method and point-by-point method. The flux method employs a relatively simple equation and is used to calculate the required number of lighting equipment. The point-by-point method can accurately calculates the intensity of illumination of a given point. The point-by-point method enables to examine the distribution of the intensity of illumination, the average value of which is calculated by the flux method. Thus the former is used to examine the evenness of the intensity of illumination. [Technical Notes] (1) Calculation by Flux Method By using the flux method, the average intensity of illumination of the irradiated area can be calculated as follows: N F UM E = -----------------A where E： average intensity of illumination (lx) N： number of lighting equipment (unit) F： total flux of one light source (lm) U： utilization factor (see (3) below) M： maintenance factor A： area of the irradiated surface (m2)

(19.12.1)

(2) Calculation by Point-by-Point Method To determine the evenness of the intensity of illumination in the design of lighting, it is necessary to use the point-by-point method. The point-by-point method first divides the area to be illuminated into a grid of rectangular blocks, and then calculates the intensity of illumination at the center of each block. The intensity of direct horizontal plane illumination Eh at point P on a given surface by a light source L is given by equation (19.12.2) (see Fig. T-19.12.2). I qa cos q (19.12.2) E h = ------------------2 l

where Eh： intensity of direct horizontal plane illumination at P (lx) I q： luminous intensity in the θ direction (cd) l： distance between the light source L and P (m) q： incident angle (º)

l

It is recommended that the shape of grid blocks be made as close as possible to a square. The length ratio of one side to the other should be 0.5 to 2.0. The maximum size p (i.e., length of the longer side) of grid blocks should be the smaller of the two values obtained by equation (19.12.3).

678

p = 0.2 × 5log d p = 10

Fig. T-19.12.2 Horizontal Intensity of Illumination at Point P

(19.12.3)

where d： width of the area in which intensity of illumination is to be calculated (m) p： maximum length of the longer side of grid block (m) Value of evenness (Eev) is the ratio of the minimum intensity of illumination to the average intensity of illumination, and the latter is calculated by equation (19.12.4) (see Fig. T- 19.12.3). E1 + E2 ¼ + En Fig. T-19.12.3 Division of Illuminated Area into Grid Blocks (19.12.4) E ev = ------------------------------------n where n： number of grid blocks in which intensity of illumination is to be calculated -536-


En： intensity of illumination at the center of the n-th grid block (3) The utilization factor gives the ratio of the flux that reaches the irradiated surface to the total flux of the light source of the lighting equipment. As the value of utilization factor of indoor lighting depends on the efficiency of the lighting equipment, the area of the irradiated surface, conditions of the room, and the difference among reflection coefficients of various parts of the room, it is important to take these factors into consideration in lighting calculations. Utilization factor of outdoor lighting can be calculated using the efficiency of the lighting equipment and the area of the irradiated surface. Actual values range from 0.2 to 0.5, and normally a utilization factor of 0.4 is used.

19.12.6 Maintenance and Management [1] Inspections (1) Periodical inspections shall be conducted for the following matters: (a) Light condition (b) Conditions of stain and damage of the equipment (c) Any peel of the coating (2) Illumination intensity measurement shall be conducted periodically for several measurement points per each facility. The measurement points shall be selected in such a way that the data taken represents the overall intensity of illumination of the facility. For measurement of the intensity of illuminations, several typical points shall be selected for the respective facilities, and the measurement be made periodically. [Technical Notes]

) of light source (

Ml

Design luminous flux maintenance factor

As the flux of any light source wanes with the increase in the cumulative lighting time, it is necessary to replace the lamp when the flux has become lower than the design flux. Figure T-19.12.4 shows the relationship between the cumulative lighting time of lamp and the luminous flux maintenance factor Ml (ratio of flux after the passing of the cumulative lighting time to initial flux (100-hour value)) for various lamp types.

Incandescent lamp

Cumulative lighting time (replacement interval) NH : High-pressure sodium vapor lamp (External starter type and built-in starter type) HF : Fluorescent mercury lamp M

: Metal halide lamp

ML : Metal halide lamp (Low starting voltage type) FL : Fluorescent lamp LW : Incandescent lamp for general lighting purposes

Fig. T-19.12.4 Design Luminous Flux Maintenance Factor of Light Source

-537-


[2] Cleaning and Repair (1) Cleaning As accumulation of dirt on the inside and outside surfaces of lighting equipment degrades the intensity of illumination at the yard or road surface, cleaning shall be performed based on the result of visual inspection or illumination intensity measurement. (2) Repair When an inspection has revealed a failed or damaged part, that part shall be repaired quickly. [Technical Notes] The flux of lighting equipment lessens with time after the start of use owing to accumulation of dirt on the lighting equipment. Therefore, it is necessary to clean the lighting equipment regularly so that the flux will not become lower than the design flux. Figure T-19.12.5 shows the rate of reduction of design flux due to accumulation of dirt on the lighting equipment (design luminous flux maintenance factor Md), and Table T-19.12.5 lists the grade of maintenance factor curves in Fig. T-19.12.5 for various combinations of lighting equipment type and surrounding environment.

Type of lighting equipment

Surrounding environment

Table T-19.12.4 Grade of Maintenance Factor Curves for Combinations of Lighting Equipment Type and Surrounding Environment

Exposed type Indoor

Outdoor

Bottom opening type Simplified sealing type (with bottom cover) Indoor Outdoor Indoor Outdoor

Sealing type Indoor Outdoor

(Fluorescent lamp)

(Light bulb)

Good

Intermediate

Bad

Design luminous flux maintenance factor of lighting equipment (Md)

(Notes) (1) For locations where significant accumulation of dirt is expected (such as those inside a tunnel), it is recommended to use the F to H curves. (2) The illustrations of lighting equipment shown in the table are typical examples.

Elapsed time (cleaning interval)

Fig. T-19.12.5 Design Luminous Flux of Lighting Equipment -538-

[Month]


[References] 1) Hirofumi INAGAKI, Koichi YAMAGUCHI, Takeo KATAYAMA: “Standardization of bollards and storm bitts for wharf”, Tech. Note of PHRI, No. 102, 1970 (in Japanese). 2) Sigeru UEDA, Eijiro OOI: “On the design of fending systems for mooring facilities in a port”, Tech. Note of PHRI, No. 596, 1987, pp. 122-127 (in Japanese). 3) Shoichi KITAJIMA, Hiroshi SAKAMOTO, Shohei KISHI, Takuji NAKANO, Syusaku KAKIZAKI: “On some problems being concerned with preparation for the design standards on port and harbour structures”, Tech. Note of PHRI, No. 30, 1967 (in Japanese). 4) loc. cit. 2), pp. 59-76. 5) Vasco Costa: “The berthing ship”, The Dock & Harbour Authority, Vol. XLV, May ～ July, 1964. 6) Sigeru UEDA, Satoru SHIRAISHI: “On the design of fenders based on the ship oscillations moored to quay walls”, Tech. Note of PHRI, No. 729, 1992 (in Japanese). 7) Yoshiaki KIKUCHI, Kunio TAKAHASHI, Misao SUZUKI: “Lateral resistance of single piles under large repeated loads”, Rept. of PHRI, Vol. 31, No. 4, 1992, pp. 33-60 (in Japanese). 8) PIANC: Report of PIANC Working Group No. 33 “Guidelines for the Design of Fenders”, Supplement to Bulletin, 2000. 9) PIANC: Report of the International Commission for Improving of Design of Fender Systems, Supplement to Bulletine, No. 45, 1984.

-539-


Chapter 20 Aprons 20.1 Principle of Design (Notification Article 95) (1) The apron shall be provided between the face line of wharf and the sheds or the open storage yards in order to ensure safe and smooth operations of temporary stacking of loading / unloading cargoes, cargo handling work, entry and exit of cargoes, and traffic of cargo handling vehicles. (2) The apron shall have an adequate area ensuring safe and smooth cargo handling.

20.2 Type of Apron 20.2.1 Width (Notification Article 96) The width of apron shall be designed appropriately in consideration of the size and use of the wharf, and the structure and use of the sheds or warehouses behind the wharf, so that safe and smooth cargo handling is ensured. [Technical Notes] (1) Currently the values listed in Table T- 20.2.1 are being used as the standard apron width for ordinary wharves. Table T- 20.2.1 Standard Apron Width Water depth of berth (m)

Apron width (m)

Less than 4.5

10

4.5 or more and less than 7.5

15

7.5 or more

20

(2) For general cargo wharves, provision of the space for cranes, temporary stacking space, cargo handling space, and access road should normally be considered. It is desirable to have an apron width of 15 to 20 m when there is sheds behind the wharf and forklift trucks are used, and an apron width of 10 to 15 m when the back of the wharf faces a road or open storage yards where trucks are to enter the apron for direct loading to and unloading from ships.

20.2.2 Gradient (Notification Article 97, Clause 1) The gradient of the apron shall be designed appropriately in consideration of the rainfall intensity and the use of the area behind the apron so that cargo handling will be performed in a smooth manner. [Technical Notes] Ordinarily a gradient of 1% to 2% toward the sea is used, but a steeper gradient may be used for small quays. In general, a relatively steep gradient is used for areas with a certain snowfall to facilitate the removal of snow. There are also cases in which a reverse gradient is employed in consideration of the use of the apron and environmental protection.

20.2.3 Type of Pavement The type of pavement for the apron shall be comprehensively determined in consideration of the soil condition beneath the subgrade, difficulty of construction, conditions of pavement in the neighborhood, type of cargo handling, cost and maintenance, etc.

20.3 Countermeasures against Settlement of Apron (Notification Article 97, Clause 2) For aprons, appropriate countermeasures shall be taken to prevent excessive settlement (due to sand outflow or consolidation of the lower landfill material) that would hinder cargo handling work and the traffic of vehicles. [Commentary] In general, the material below the subgrade of apron pavement is subject to settlement due to consolidation. There is also a risk of settlement due to outflow of the landfill material used as part of the layers below the subgrade through quaywall joints, or compression of the backfilling material behind the quaywall. There are many cases of the failure of pavement that are thought to be attributable mainly to these types of settlement. Therefore, measures will normally be taken to prevent these types of settlement such as the provision of countermeasure against sand outflow and the compaction of the backfilling material behind the quaywall. -540-


20.4 Load Conditions (Notification Article 98) In the design of the pavement structure, the design load shall be selected in such a way that the calculated pavement thickness would become largest, in consideration of the loads and their ground contact areas of the trucks, truck cranes, forklift trucks, and straddle carriers to be mobilized, as well as the types of cargoes and the mode of cargo handling. [Technical Notes] Table T- 20.4.1 lists the reference values of loads which are applied to the apron pavement. Table T- 20.4.1 Load Conditions Maximum load per outrigger or wheel (kN)

Type of Load

Truck crane

Truck Tractor-trailer

Contact area per outrigger or wheel (cm2)

Contact pressure (N/cm2)

Type20

200

1,250

160

Type25

250

1,500

167

Type30

290

1,700

171

Type40

390

2,000

195

Type50

490

2,300

213

Type80

740

2,850

260

Type100

900

3,400

265

Type120

1,060

3,850

275

Type150

1,310

4,500

291

T-25

100

1,000

100

For 20-ft

50

1,000

50

For 40-ft

50

1,000

50

For 40-ft yard use only

70

1,000

70

2t

25

300

83

3.5t

45

500

90

Forklift truck

Transfer crane

6t

70

800

88

10t

110

1,400

79

15t

170

2,300

74

20t

240

3,150

76

25t

300

3,800

79

35t

440

5,000

88

20t

200

2,400

83

25t

250

2,850

88

30t

290

3,250

89

110

1,200

92

Straddle carrier

20.5 Design of Concrete Pavement 20.5.1 Design Conditions As for the design conditions for concrete pavement, the following shall be considered: (1) Design load (2) Traffic volume (3) Bearing capacity of subgrade (4) Materials (5) Others [Commentary] (1) The design load should be the one that makes the concrete slab thickness largest among those calculated with the candidate loads. (2) The design bearing capacity coefficient K30 should be determined for the subgrade of the pavement area under consideration. -541-


[Technical Notes] (1) The standard design bearing capacity coefficient K30 for the subgrade should be 200 N/cm3. (2) The standard 28-day bending strength of concrete should be 450 N/cm2. The bending strength of concrete should be determined in accordance with the JIS A 1132 “Method of Making and Curing Concrete Specimens” and the JIS A 1106 “Method of Test for Flexural Strength of Concrete” (3) In the design of concrete pavement, the “Cement Concrete Pavement Guideline” 1) issued by the Japan Road Association and the “Standard Specifications of Concrete (pavement)” 2) issued by the Japan Society of Civil Engineers may be used as references. (4) The bearing capacity coefficient should be determined by means of plate load test specified in JIS A 1215. The bearing capacity coefficient K30 used for calculation should correspond to a settlement of 0.125 cm. (5) The bearing capacity coefficient to be used in design should be calculated by equation (20.5.1) using the measured values at site from at least three points in a block where the subgrade is made of the same material. Before applying the equation, the extreme values in the measurement should be removed.

(

Design bearing capacity coefficient

) ( ＝

Average of the bearing capacity coefficient values of the test points

) (

Maximum value of bearing capacity coefficient

－

) ( －

Minimum value of bearing capacity coefficient

)

(20.5.1)

d2

where d2 is the value listed in Table T- 20.5.1. The plate load test will normally be performed at the rate of one or two points for every 50 m of the wharf in the direction of extension line. Table T- 20.5.1 Values of d2 Number of points (n)

3

4

5

6

7

8

9

10 or more

d2

1.91

2.24

2.48

2.67

2.83

2.96

3.08

3.18

(6) When the subgrade has already been constructed, the bearing capacity coefficient should be obtained by performing a plate load test on the subgrade at the condition of maximum moisture content. When it is not possible to conduct a plate loading test in such condition, the bearing capacity coefficient should be obtained by correcting the value using equation (20.5.2). The CBR values in the equation should be obtained from undisturbed soil samples.

(

Corrected bearing capacity coefficient

)

＝

(

Measured bearing capacity coefficient

)

×

CBR (four-day submersion) CBR (natural water content)

(20.5.2)

20.5.2 Composition of Pavement (Notification Article 99) The composition of the apron pavement shall be determined in consideration of the pavement type, the structure of base course, the characteristics of the materials used, etc., based on an appropriately defined subgrade bearing capacity. [Commentary] (1) Concrete pavement normally comprises a concrete slab and a base course, which is constructed on the subgrade. (2) The composition of pavement should be determined appropriately in consideration of the load conditions, traffic conditions, subgrade conditions, meteorological conditions, and construction and maintenance costs. (3) It is recommended to determine the design of the base course in such a way that the bearing capacity coefficient of the base course becomes equal to 200 N/cm3 by preparing a test base course and making prototype tests. When it is difficult to make the prototype tests, the thickness of the base course may be determined by using design curves shown in Fig. C- 20.5.1. The minimum thickness of base course is 15 cm. (4) The “Cement Concrete Pavement Guideline” 3) issued by the Japan Road Association may be used to assess the grading and characteristics of base course material.

-542-

t

en

m

Gr ad

ed

ag

Cru

she r-ru n, e tc. gre ga te

Base course thickness (cm)


n

by

ce

io

at

liz

i ab

il

st

So

(Bearing capacity coefficient of base course) K1 = (Design bearing capacity coefficient of subgrade) K2 K1: bearing capacity coefficient K30 of base course (200 N/cm3) K2: bearing capacity coefficient K30 of subgrade

Fig. C- 20.5.1 Design Curves for Base Course Thickness

[Technical Notes] (1) The composition of concrete pavement may be determined by using the values listed in Tables T- 20.5.2 and T20.5.3 as the reference. Table T- 20.5.2 Base Course Thickness of Concrete Pavement Design condition

Base course thickness (cm)

Design bearing capacity coefficient K30 of subgrade (N/cm3) Equal to or above 50 and less than 70

Base course

Grading adjusted material

－ 20

40 －

－ 20

25

－ 20

－ 15

－ 15

20

－ 15

15

－ 20 －

Cement stabilization

－ Equal to or above 70 and less than 100

Equal to or above 100

Subbase course


－

－ 15

Crusher run etc.

Total base course thickness

20

60

－ 30

40

－ 20

35

－ 15

30

－

－

20

－

－

15

－

Table T- 20.5.3 Concrete Slab Thickness Design load classification

Slab thickness (cm)

CP1

20

CP2

25

CP3

30

CP4

35

On the deck slab of open-type wharf

10

-543-

55 40 30


(2) The design load may be determined by using the values listed in Table T- 20.5.4 as the reference. Table T- 20.5.4 Design Loads by Classification Classification of design load CP1

CP2

CP3

CP4

Type of load

Load (kN)

Radius of contact area (cm)

Forklift truck

(2t)

25

9.8

Tractor-trailer

(for 20-ft and 40-ft)

50

17.8

Forklift truck

(3.5t)

45

12.6

Tractor-trailer

(for 40-ft yard use only)

70

17.8

Forklift truck

(6t)

70

16.0

Truck

(T-25)

100

17.8

Forklift truck

(10t)

110

21.1

110

19.5

Straddle carrier Forklift truck

(15t)

170

27.1

Transfer crane

(20t)

200

17.6

Truck crane

(25 Type)

200

20.0

Forklift truck

(20t)

240

31.7

Truck crane

(25 Type)

250

21.6

(3) In the design of pavement in the cold region where the pavement is subject to freezing and thawing, the thickness of pavement should be made greater than the depth of frost penetration. When the pavement thickness is smaller than the depth of frost penetration, a frost blanket or a layer preventing frost penetration should be provided. (4) The concrete slab should be reinforced with wire mesh. (5) When the bearing capacity coefficient of the subgrade is less than 50 N/cm3, the pavement should be designed using Fig. C- 20.5.1 and Fig. T- 20.5.1 or equation (20.5.3). (6) Figure T- 20.5.1 shows the relationship between the concrete slab thickness and bending stress. The bending stress value in the ordinate has been calculated using equation (20.5.3), which is generally referred to as the Picket equation or Arlington equation. a æ ö --÷ 10CP ç l - ç1 ---------------------------------- ÷ s = ------------2 a h ç 0.925 + 0.22 --- ÷ è lø where

s： C： P： h： a：

(20.5.3)

maximum stress of concrete slab at the right angle corner (N/mm2) coefficient; when slip bars are used, C = 3.36 may be used. load (kN) thickness of concrete slab (cm) radius of ground contact area (cm) 4

Eh 3 ---------------------------------12 ( 1 n 2 )K 75 E = 3,500,000 N/cm2

l： radius of relative stiffness of the pavement (cm),

l =

E： elastic modulus of concrete (N/cm2), n： Poisson’s ratio of concrete n = 0.15 K75： design bearing capacity coefficient of subgrade (N/cm3) Under the assumption that K30 / K75 = 2.8, the coefficient K75 = K30 / 2.8 = 200/2.8 ≒ 70 (N/cm3) can be used in equation (20.5.3).

-544-


0.45

u Tr

ck cr

an e

(5

0 p Ty

cm2

cm .1 27

kN

,a N 0k 49 cm .2 25

Bending stress of concrete

P

0.20

cm .9 39

0.25

,a N 0k 39

0.30

a P , N ) cm pe 0k .6 Ty 44 23 0 (4 P e ,a an t) N cr 5 k 0k (3 uc k 29 c Tr ru P tt lif ) rk pe Fo Ty 0 cm .6 (3 e cm 21 cm an .0 8 . cr m ,a 20 k 34 .1c N uc 0k Tr 30 ,a ,a 25 a kN kN P , 0 00 N 30 2 0k cm P P 25 .7 ) ) 31 P 5t pe (2 e Ty ,a ck cran 0 ru (2 kN er tt e sf 40 lif an 2 an rk cr cm Tr k Fo P .6 uc cm 27 Tr ) .1 0t 27 (2 ,a N ck 0k ru ,a tt 20 kN lif 70 rk P 1 Fo ne P ra ) cm rc 5t .1 sfe (1 21 an ck Tr ru ,a tt lif kN 0 rk 11 Fo cm P .5 ) 19 0t (1 a , ck N tru 0k ift 11 rkl cm Fo P .8 17 er rri a ca N, dle 0k ad 10 Str P m ) 25 cm .0c T16 .8 k( 17 a uc , Tr a kN , 70 kN P 70 cm cm 6t) .8 6 P k( 2. ) 17 1 ruc nly ft t o kli ,a ,a rd For kN ya kN 45 -ft 50 40 P or t) P r (f 3.5 aile k( ft) r-tr 0ruc t 4 t cto nd klif ) Tra 2t For ft a ( cm ck (20 er tru 9.8 ail i ft r-tr ,a r kl cto kN Fo 25 Tra P

0.35

e)

0.40

0.15 15

20

25

30

35

CP1

CP2

CP3

CP4

40

45

50

55

Thickness of concrete (cm)

Fig. T- 20.5.1 Relationship between Concrete Slab Thickness and Bending Stress

20.5.3 Joints (Notification Article 100) (1) In principle, the concrete pavement of apron shall be provided with joints. (2) Joints of the concrete pavement of apron shall be arranged appropriately, in consideration of the size of apron, structure of the wharf, and the type of joint. In addition, joints shall have a structure that is appropriate for the type of joint. [Commentary] (1) Joints of concrete slabs are provided for the purpose of reducing stresses of concrete slab by allowing the expansion, shrinkage, and warping of the slab to some extent. The interval of joints should be designed appropriately in consideration of the ground conditions, the load conditions, and the structure of the base of the apron. (2) Joints are structurally weak parts of concrete pavement and also require construction work more complex than other sections. The larger the number of joints used, the higher the costs of construction and maintenance become. Therefore, it is advisable to use as small a number of joints as possible. [Technical Notes] (1) Longitudinal Joint Longitudinal construction joints should be butt joints with tie-bars. When the concrete pavement is constructed on a deck slab of open-type wharf, no tie-bars are used. Longitudinal construction joints adjacent to the quaywall coping or the shed should be of the structure that uses joint filler and joint plates. The interval between longitudinal construction joints is determined in consideration of the type of paving equipment used, total width of pavement, wheel distance of truck cranes, etc. Usually an interval of 3.5 to 5.0 m is used. -545-


Longitudinal construction joints are provided to facilitate construction work. It is known from experience that the concrete slab is subject to longitudinal cracks when longitudinal construction joints are made at intervals of 5 m or longer, and subject to transverse cracks when longitudinal construction joints are made at intervals of 2.5 m or less. Nowadays it is possible to make the maximum interval up to 7.5 m using large concrete finishers. Such a large interval is convenient in terms of workability and construction cost, but it is recognized that a longitudinal contraction joint must be provided along the center of each slab when the thickness of concrete slab is 30 cm or less and the interval is 5.1 m or longer. In terms of structural type, longitudinal contraction joints should be the dummy joints. It is recommended that longitudinal construction joints are to be located along the shoulder line of the backfill, above joints of the quaywall structure, and above the centerline of anchor sheet piles, in order to avoid the effect of the change in bearing capacity of ground below the base course or the effect of the presence of quaywall joints on the concrete slab of pavement. (2) Transverse Joint (a) Transverse contraction joint Transverse contraction joints are dummy joints with slip-bars. When the concrete pavement is constructed on the deck slab of open-type wharf, however, no slip-bars are used. The interval of transverse contraction joints should be determined by referring to the data on past construction projects in the respective area. Under normal meteorological conditions, a transverse contraction joint interval of about 5 m is often used. It is advisable that contraction joints are located on the joints of the quaywall structure. (b) Transverse construction joint Transverse construction joints are butt joints with slip-bars. When the concrete pavement is constructed on the deck slab of open-type wharf, however, no slip-bars are used. Transverse construction joints are provided when some unavoidable reasons arises, such as to divide a daily construction area or as to stop concrete placing work due to sudden rain or troubles of plant equipment. The location of transverse construction joints should be at the positions assigned to the predetermined transverse contraction joints. (c) Transverse expansion joint A transverse expansion joint has the structure that comprises a joint-filler at surface, a joint plate at bottom, and slip-bars. When the concrete pavement is constructed on the deck slab of open-type wharf, however, no slip-bars are used. The interval between transverse expansion joints should be determined in consideration of empirical data on past construction projects in the respective area. In many cases, a transverse expansion joint interval of 100 to 200 m is used when the construction is conducted during a hot season, and a transverse expansion joint interval of 50 to 100 m is used when the construction is conducted during a cold season. As expansion joints constitute the serious weak point of pavement, the number of such joints should be made as small as possible. (3) Tie-bars are provided to prevent adjoining slabs from separating, and sinking / rising of either slab at joints. Tiebars also serve as a reinforcement to transfer the sectional force. Because the apron pavement has a relatively small width and is physically constrained by the main structure of the quaywall or sheds, separation of apron concrete slabs at joints rarely occurs. However, it is necessary to provide tie-bars at longitudinal construction joints to prevent sinking / rising of either slab at joints due to differential settlement of layers below the base course, and to accommodate a wide variety in the directions of traffic load that is not observed on ordinary roads. (4) Slip-bars transmit loads from one slab to another and prevent sinking / rising of either slab at joints. Transverse contraction joints, transverse construction joints, and transverse expansion joints should all be provided with slip-bars to ensure effective transmission of loads. (5) Figures T- 20.5.2 through T- 20.5.5 exhibit standard structures of various types of joints.

Joint-filler

Joint-filler

Tie-bar

Chair

Slip-bar

Chair

(This side is coated with a paint

Wooden base

and grease or with two coats of a bituminous material)

Fig. T- 20.5.2 Longitudinal Construction Joint

Fig. T- 20.5.3 Transverse Contraction Joint

-546-

Slip-bar


Joint-filler

Joint-filler

Cap Slip-bar

Chair

Chair

(This side is coated with a paint

Joint plate

Slip-bar (This side is coated with a paint

and grease or with two coats of

and grease or with two coats of

a bituminous material)

a bituminous material)

Fig. T- 20.5.4 Transverse Construction Joint

Fig. T- 20.5.5 Transverse Expansion Joint

20.5.4 Tie-Bar and Slip-Bar Specifications and intervals of tie-bars and slip-bars shall be determined appropriately in consideration of the situations that the vehicle traveling load acts on the apron pavement in both the longitudinal and transverse directions. [Technical Notes] (1) In the determination of dimensions and intervals of tie-bars and slip-bars, the values listed in Table T- 20.5.5 may be used as the reference. Table T- 20.5.5 Dimensions and Intervals of Tie-Bars and Slip-Bars Classification of design load

Slab thickness (cm)

Tie-Bar SD295A Diameter (cm)

Length (cm)

Slip-Bar SS400 Interval (cm)

Diameter (cm)

Length (cm)

Interval (cm)

CP1

20

25

80

45

25

50

45

CP2

25

25

100

45

25

50

45

CP3

30

32

100

40

32

60

40

CP4

35

32

100

40

32

60

40

(2) For tie-bars, SD295A (deformed bar) as specified in the JIS G 3112 “Steel Bars for Reinforced Concrete” is used. (3) For slip-bars, SS400 (round bar) as specified in the JIS G 3101 “Rolled Steel for General Structures” is used.

20.5.5 End Protection An end protection work along the landward side of pavement shall be provided at a location where there is a risk of destruction of the base course due to infiltration of rain water or destruction of the concrete slab and base course due to heavy loading.

20.6 Design of Asphalt Pavement 20.6.1 Design Conditions The following shall be considered as the design conditions for asphalt pavement: (1) Design load (2) Traffic volume (3) Bearing capacity of subgrade (4) Materials (5) Others [Commentary] (1) The design load should be the one that makes the pavement thickness largest among those calculated with the candidate loads. (2) The design CBR should be determined for the subgrade of the pavement area under consideration.

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[Technical Notes] The CBR test can be carried out as specified in the JIS A 1211 “Test Methods for the California Bearing Ratio (CBR) of Soils in Laboratory”. The subgrade soil should be compacted under the natural water content and submerged in water for four days before obtaining the CBR value. Aggregates of 40 mm or larger in diameter should be removed from the sample subgrade soil. The sample soil should be placed into a mould in three layers of equal thickness; each layer should be compacted 67 times. When the subgrade has already been constructed, the in-situ CBR test should be carried out at the site at the time when the subgrade soil is most saturated with water. However the value of the in-situ CBR test may be corrected by equation (20.6.1) if the in-situ CBR cannot be carried out at such the condition. CBR (four-days submersion) CBR (natural water content)

Modified CBR ＝ In-situ CBR ×

(20.6.1)

The design CBR can be obtained from such CBR values by equation (20.6.2) after removing extremely diviated values. (max.CBR value - min.CBR value) Design CBR ＝ (average value of CBR at different points) － (20.6.2) d2 where d2 is of a value listed in Table T- 20.5.1. Sampling should be made at one or two places for every 50 m in the direction of the face line of the wharf and at a depth greater than 50 cm from the completed subgrade or the exposed surface of a burrow pit for subgrade soil.

20.6.2 Composition of Pavement (Notification Article 99) The composition of apron pavement shall be determined based on an appropriately designed subgrade bearing capacity, in consideration of the pavement type, the structure of base course, the characteristics of the materials used, etc. [Commentary] (1) Asphalt pavement normally comprises a surface course, a binder course, and a base course, which is constructed on the subgrade. The base course is normally constructed in two stages (base course and subbase course). (2) The composition of pavement should be designed appropriately in consideration of the load conditions, traffic conditions, subgrade conditions, meteorological conditions, and construction and maintenance costs. (3) The “Manual for Asphalt Pavement” 4) issued by the Japan Road Association should be applied for evaluating the material characteristics and gradation of material. (4) The layer equivalency factor to be employed for the equivalent pavement thickness TA is listed in Table C20.6.1. Table C- 20.6.1 Layer Equivalency Factor Layer

Construction method / material

Requirements

Layer equivalency factor

Surface and binder courses

Hot asphalt mixture for surface and binder courses

-

1.00

AC I~AC Ⅳ

Marshall stability level 3.43 kN or greater

0.80

A-treated material II

Marshall stability level 2.45 to 3.43 KN

0.55

A-treated material I

Grading adjustment

Corrected CBR 80 or greater

0.35


Crusher-run, slag, sand, etc.

Corrected CBR 30 or greater

0.25

Corrected CBR 20 to 30

0.20

Bituminous stabilization Base course

Subbase course

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Remark

Grain material


[Technical Notes] (1) Table T- 20.6.1 lists the standard composition of asphalt pavement. Table T- 20.6.1 Standard Composition of Asphalt Pavement Design conditions Design load

Surface course Type

h1 (cm)

Type

h2 (cm)

Type

h3 (cm)

h4 (cm)

H

TA (cm)

Equal to or above 3 and less than 5

AC I

5

AC Ⅲ

5

Grading adjusted materiall

25

35

70

25.8

AC I

5

-

-


25

35

65

25.8

20

25

55

22.0

Equal to or above 8 and less than 12 Equal to or above 12 and less than 20

Base course

Total thickness

AC I

5

AC Ⅲ

5


AC I

5

-

-


20

30

55

22.0

AC I

5

AC Ⅲ

5


15

20

45

19.3

AC I

5

-

-


15

30

50

19.3

15

15

40

18.3

AC I

5

AC Ⅲ

5


AC I

5

-

-


15

20

40

17.3

15

15

40

18.3

AC I

5

AC Ⅲ

5


AC I

5

-

-


15

15

35

16.3

AC I

5

AC Ⅲ

4 or greater

-

-

-

9 or greater

-


AC II

5

AC Ⅳ

5


25

35

70

25.8

AC II

5

-

-


25

35

65

25.8


AC II

5

AC Ⅳ

5


20

25

55

22.0

AC II

5

-

-


20

30

55

22.0


AC II

5

AC Ⅳ

5


15

20

45

19.3

AC II

5

-

-


15

30

50

19.3

15

15

40

18.3

Equal to or above 20 On the deck slab of opentype wharf

AP2 50 kN Contact pressure 70 N/cm2 or greater

Binder course

Subbase course

Design CBR of subgrade (%)

Equal to or above 5 and less than 8 AP1 50 kN Contact pressure less than 70 N/cm2

Composition of pavement


AC II

5

AC Ⅳ

5


AC II

5

-

-


15

20

40

17.3

Equal to or above 20

AC II

5

AC Ⅳ

5


15

15

40

18.3

AC II

5

-

-


15

15

35

16.3

AC II

5

AC Ⅳ

4 or greater

-

-

-

9 or greater

-

On the deck slab of opentype wharf

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Table T- 20.6.1 Standard Composition of Asphalt Pavement (continued) Design conditions Design load


Binder course

Subbase course

Base course

Total thickness

Design CBR of subgrade (%)

Surface course Type

h1 (cm)

Type

h2 (cm)

Type

h3 (cm)

h4 (cm)

H

TA (cm)


AC II

5

AC Ⅳ

15


30

45

95

40.0

AC II

5

AC Ⅳ

10


20

45

80

40.0

25

30

75

34.8


AC II

5

AC Ⅳ

15


AC II

5

AC Ⅳ

10


20

20

55

35.0


AC II

5

AC Ⅳ

15


15

20

55

29.3

AC II

5

AC Ⅳ

10


15

15

45

30.0


AC II

5

AC Ⅳ

15


15

15

50

28.3

AC II

5

AC Ⅳ

10


15

15

45

30.0

15

15

50

28.3

Equal to or above 20 On the deck slab of opentype wharf Equal to or above 3 and less than 5



AC II

5

AC Ⅳ

15

Grading adjusted materials

AC II

5

AC Ⅳ

10


15

15

45

30.0

AC II

5

AC Ⅳ

4 or greater

-

-

-

9 or greater

-

AC II

5

AC Ⅳ

15


40

60

120

46.0

AC II

5

AC Ⅳ

10


20

70

105

45.0

30

45

95

39.5


AC II

5

AC Ⅳ

15


AC II

5

AC Ⅳ

10


20

40

75

39.0


AC II

5

AC Ⅳ

15


25

30

75

34.8

AC II

5

AC Ⅳ

10


15

35

65

34.0

15

25

60

30.3

Equal to or above 12 and less than 20 Equal to or above 20 On the deck slab of opentype wharf

AC II

5

AC Ⅳ

15


AC II

5

AC Ⅳ

10


15

15

45

30.0

15

15

50

28.3

AC II

5

AC Ⅳ

15


AC II

5

AC Ⅳ

10


15

15

45

30.0

AC II

5

AC Ⅳ

4 or greater

-

-

-

9 or greater

-

Note that H and TA in Table T- 20.6.1 are the total thickness of pavement and equivalent pavement thickness, respectively. In case of the deck slab of open-type wharf, the boxes of the binder course in Table T- 20.6.1 refer to the value for the total of filling material and binder course. This course does not necessarily have to be asphalt concrete. When the design CBR of the subgrade is equal to or above 2 and less than 3, the subgrade should be replaced with a good material or a cutoff layer should be provided. When the design CBR is less than 2, the subgrade should be replaced with a good material and the pavement thickness should be designed separately.

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(2) The design load is classified as listed in Table T- 20.6.2. Table T- 20.6.2 Classification of Design Load AP1

Tractor-trailer

For 20-ft, 40-ft, and 40-ft yard use only

AP2

Forklift truck

2t, 3.5t, 6t

AP3

Forklift truck Truck Transfer crane Straddle carrier Truck crane

10t, 15t T-25 20t

Truck crane Transfer crane

25Type 25t

AP4

20 Type

(3) The type and material of asphalt concrete should be as listed in Table T- 20.6.3 Table T- 20.6.3 Type and Quality of Asphalt Concrete Type Use Number of blows for Marshall stability test Marshall stability (kN) Flow value (1/100cm) Porosity (%) Degree of saturation (%)

AC Ⅰ

AC Ⅱ

AC Ⅲ

For surface course

AC Ⅳ

For binder course

50

75

50

75

4.9 or greater 20 ～ 40 3～5 75 ～ 85

8.8 or greater 20 ～ 40 2～5 75 ～ 85

4.9 or greater 15 ～ 40 3～6 65 ～ 80

8.8 or greater 15 ～ 40 3～6 65 ～ 85

Note: The columns of “number of blows 75 times” apply to cases where the tire ground contact pressure of the design load is 70 N/cm2 or greater, or where the traffic of large vehicles is heavy and rutting is expected.

(4) In the design of pavement for the cold region where the pavement is subjected to freezing and thawing, the thickness of pavement should be made greater than the depth of frost penetration. When the pavement thickness is smaller than the frost penetration depth, a layer preventing frost penetration should be provided.

20.6.3 End Protection See 20.5.5 End Protection.

20.7 Design of Concrete Block Pavement 20.7.1 Design Conditions The following shall be considered as the design conditions for concrete block pavement: (1) Design load (2) Traffic volume (3) Bearing capacity of subgrade (4) Materials (5) Others [Commentary] (1) The design load should be the one that makes the thickness of the concrete block largest among those calculated with the candidate loads. (2) The design bearing capacity coefficient K30 should be determined for the subgrade of the pavement area under consideration. [Technical Notes] (1) The bearing capacity coefficient K30 of the subgrade should normally be 200 N/cm3. (2) The 28-day bending strength of concrete should normally be 450 N/cm2. (3) The bearing capacity should be determined by means of the plate load test specified in JIS A 1215. The settlement used to obtain the bearing capacity coefficient is 0.25 cm. -551-


20.7.2 Composition of Pavement (Notification Article 99) The composition of apron pavement shall be designed based on an appropriately designed subgrade bearing capacity, in consideration of the pavement type, the structure of base course, the characteristics of the materials used, etc. [Commentary] (1) Concrete block pavement normally comprises a concrete block layer and a base course, which is constructed on the subgrade. (2) The composition of pavement should be designed appropriately in consideration of the load conditions, traffic conditions, subgrade conditions, meteorological conditions, and construction and maintenance costs. [Technical Notes] (1) Table T- 20.7.1 lists the standard composition of concrete block pavement. The blocks here are of square shape. Table T- 20.7.1 Standard Composition of Block Pavement Design conditions

Design load

BP1

BP2

BP3

BP4


Side length of Design bearing capacity coefficient block (m) 3 of subgrade K30 (N/cm )

Base course Thickness of block (cm)

Subbase course

Total Grading Grain material thickness (cm) adjusted (cm) material (cm)

Equal to or above 50 and less than 70 Equal to or above 70 and less than 100 Equal to or above 100

0.9

25

20

30

75

0.9

25

20

15

60

0.9

25

20

-

45


0.8

25

20

30

75

0.8

25

20

15

60

0.8

25

20

-

45


1.1

30

20

30

80

1.1

30

20

15

65

1.1

30

20

-

50


1.2

35

20

30

85

1.2

35

20

15

70

1.2

35

20

-

55

(2) The design load is classified as listed in Table T- 20.7.2. Table T- 20.7.2 Classification of Design Load BP1

Forklift truck Tractor-trailer

2t, 3.5t, 6t 20-ft, 40-ft

BP2

Forklift truck Tractor-trailer Straddle carrier

10t For 40 ft yard use only -

BP3

Truck Forklift truck Truck crane Transfer crane Transfer crane

T-25 15t 20-Type 20t 25t

BP4

Truck crane

25-Type

(3) The base course should be designed in accordance with 20.5.2 Composition of Pavement [Commentary](3).

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20.7.3 Joints Joints of concrete block pavement shall be of such a structure that its function will not be lost due to differential settlement of blocks. [Technical Notes] Figure T- 20.7.1 exhibits the structure of standard block pavement joint. Joint-filler

Concrete block

Concrete block

Sand

Fig. T- 20.7.1 Structure of Joint

[References] 1) 2) 3) 4)

Japan Road Association: “Cement Concrete Pavement Guideline”, Maruzen, April 1997 (in Japanese). JSCE: “Standard Specifications of Concrete (Pavement)”, Maruzen, March 1996 (in Japanese). loc. cit. 1) Japan Road Association: “Manual for Asphalt Pavement”, Maruzen, February 1998 (in Japanese).

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Chapter 21 Foundation for Cargo Handling Equipment 21.1 Principle of Design The foundation for rail-type traveling cargo handling equipment shall be designed appropriately in consideration of the external forces that act on the foundation, allowable displacement for the foundation, degree of difficulty of maintenance, effects on the wharf structure, construction and maintenance costs, etc. [Technical Notes] The displacement of rails is small at the time of completion of construction, but it increases with the lapse of time. Therefore, it is common practice to make construction errors as small as possible. Tolerance differs somewhat among manufacturers of equipment. Table T- 21.1.1 1) lists the installation and maintenance requirements that are commonly employed. Table T- 21.1.1 Requirements for Installation of Rails for Traveling of Equipment Item

Installation requirements

Maintenance requirements (upper limits for operation)

Rail span

±10 mm or less for the entire rail length


Lateral and vertical warps of rail

5 mm or less per 10 m of rails

10 mm or less per 10m of rails

Elevation difference between seaward and landward rails

1/1000 of rail span or less

1/500 of rail span

Gradient in the travelling direction

1/500 or less

1/250 or less

Straightness



Vertical and lateral differences: 0.5 mm or less

Vertical and lateral differences: 1mm or less

Gap: 5 mm or less

Gap: 5 mm or less

―

10% or less of the original dimension

Rail joints Wear of the head of rail

21.2 External Forces Acting on Foundation External forces that act on the foundation for cargo handling equipment shall be designed appropriately in due consideration of the type, size, and operation conditions of the cargo handling equipment. [Technical Notes] (1) External forces that act on the foundation of cargo handling equipment should be calculated in accordance with the JIS B 8821 “Specification for the Design of Crane Structures” or the “Crane Structure Standards” (Labor Ministry Notification). For the wheel loads of container cranes installed on the high seismic resistant quaywalls, refer to Part Ⅸ , 2.4 Cargo Handling Equipment. The maximum wheel loads are to be examined in the crane condifions as specified in a form of Table T- 21.2.1 2). (2) The wheel loads are assumed to act to the sections of the entire length of rails when the crane is operational or during earthquakes. During storms, the wheel loads act to the section where the crane is stationed. (3) For the wheel loads that act on the rails when the crane is operational, a traveling load that is equal to 120% of the maximum static wheel load of the crane should be considered. However, this should be 110% of the maximum static wheel pressure of the crane when the traveling speed is less than 60 m/min 3).

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Table T- 21.2.1 Specification From for Maximum Wheel Load (Units: kN/wheel) Seaward

Landward

○○○

○○○

Perpendicular to rails

○○

○○

Parallel to rails

○○

○○

○○○

○○○

○○

○○

Vertical In operation

Horizontal

Vertical During storm


Horizontal

Vertical During an earthquake

Horizontal

○○○

○○○


○○

○○

Parallel to rails

○○

○○

Items to be listed • Crane capacity (t/h) • Deadweight of the crane (kN) • Wheel base (m) • Rail span (m) Seaward (wheels) • Number of wheels Landward (wheels) • Wheel interval (m) • Travelling speed of the crane (m/min) • Fixed leg / hinged leg

678

Legs

Crane conditions

21.3 Design of Foundation with Piles 21.3.1 Concrete Beam (1) Concrete beams constructed on pile foundation shall be assumed to be a continuous beam supported at the locations of pile heads. Effects of ground reactions shall be ignored. (2) Concrete beams constructed on pile foundation shall be designed against the bearing pressure between the rail and concrete, and against the loads transmited from the rail. [Technical Notes] The rail stress is usually calculated by assuming that the rail is an infinite continuous beam supported by elastic foundation. This method is often used for the cases where the wheel loads are spread over the beam by means of inserting an elastic material (such as rubber pads) between the rail and the concrete beam to prevent crushing of concrete. (a) The rail stress and the bearing pressure between the rail and concrete can be calculated using the method described in 21.4.2 Concrete Beam [Technical Notes]. In this case, the symbols Ec, Ic, and K of equation (21.4.1) should be read as follows: Ec： elastic modulus of the rail Ic： moment of inertia of the rail K： elastic modulus of the material placed under the rail (when tie pads are used, use the elastic modulus of the tie pad) When the bearing stress is too high, it should be reduced by means of stress spreading through insertion of elastic plates under the rail. (b) The fastening force between the rail and the foundation can be calculated by using the beam theory on elastic foundation 4), but it is necessary to have a sufficient allowance for the effect of impact. For calculation of the fastening force for the cases where the double elastic fastening method is employed, refer to Minemura’s study 5). In many cases, bolts with a diameter of about 22 mm are used at intervals of approximately 50 cm.

21.3.2 Bearing Capacity of Piles Piles shall have the strength that is safe against the loads of the cargo handling equipment and its foundation. [Technical Notes] (1) The force that acts on the piles should be the reaction force calculated in accordance with 21.3.1 Concrete Beam. (2) The bearing capacity of piles should be calculated in accordance with Part Ⅴ, Chapter 4 Bearing Capacity of Pile Foundations. (3) When piles are under the influence of the active earth pressure failure plane, the method described in 6.6.3 Design of Piles should be used. -555-


(4) When piles are under the influence of the active earth pressure failure plane, the required embedment length differs between the seaward piles and landward piles. However, it is common practice to use foundation piles of the same length for both the seaward and landward, to prevent a differential settlement of the foundation. When the piles on both sides are driven into the bearing stratum, there is no need to use the same embedded length.

21.4 Design of Foundation without Piles 21.4.1 Examination of Effects on Wharf When no pile is used to support the foundation for cargo handling equipment, the effect of the loads of the cargo handling equipment and its foundation on the main structure of wharf shall be examined. [Technical Notes] (1) Application of a load on the area behind a gravity type structure increases the earth pressure and may cause forward sliding of the quaywall 6). The influence of a concentrated load on the earth pressure is large in the zone at the levels immediately below the loading point. But the influence becomes smaller as the depth increases. When the quaywall height is small and the quaywall length (in the direction of extension) is short, design should be conducted carefully because of strong influence of concentrated load. When the load is applied directly on a quaywall, the subsoil reaction force increases. In particular, when the load is applied on the quaywall at its front end, the subsoil reaction force at the front toe becomes significantly large. In a quaywall of small width and short length, this tendency of reaction force increase is amplified and thus design should be conducted carefully. (2) In ordinary sheet pile quaywalls, the maximum stress occurs at a level midway between the tie member fixing point and the sea bottom. However, when a concentrated load is expected to act on the area behind the sheet pile wall, design should be conducted carefully because the maximum stress may occur at the level near the tie member fixing point, although the concentrated load rarely produces an adverse effect on the embedded part of the sheet pile. It is recommended that a sufficient overburden (earth covering) thickness for the tie members be provided to avoid adverse effects on the tie members.

21.4.2 Concrete Beam The reinforced concrete beam that is to be constructed directly on a rubble stone foundation built on the original ground shall be designed in such a way that the beam is sufficiently safe against the bending moment, shearing force, and deflection, and that the settlement of the beam is smaller than the allowable amount of settlement. [Technical Notes] (1) The bending moment, shearing force, and deflection of a reinforced concrete beam that is constructed on a rubble stone foundation can be calculated by means of equations (21.4.1) through (21.4.6). (a) When the loading point is located around the center of the beam: 4

M =

Ec Ic ---------- SW i ebxi ( cos bx i sin bx i ) 64K

(21.4.1)

1 S = --- SWi e bxi cos bx i 2 y =

4

(21.4.2)

1 ----------------------3- SW i ebxi ( cos bx i + sin bx i ) 64E c I c K

(21.4.3)

(b) When the loading point is located at the edge or joint of the beam: Wi M = S ------ eb xi sin bx i b

(21.4.4)

S = SW i ebx i ( sin bx i cos bx i )

(21.4.5)

2Wi b y = ------------- e bxi cos bx i K where

(21.4.6)

K ------------4E c I c M： bending moment of the beam section under examination (N• mm) S： shearing force of the beam section under examination (N) y： deflection of the beam section under examination (mm) b =

4

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Ec： Wi： Ic： K： C： b： xi：

modulus of elasticity of concrete (N/mm2) wheel load (N) moment of inertia of the concrete foundation (mm4) modulus of elasticity of the ground; K = Cb pressure required to cause a unit amount of settlement on a unit area of ground (N/mm3) width of the concrete beam bottom (mm) distance between the location of the wheel load and the beam section under examination (mm)

[References] 1) Society of Mechanization of Port Cargo Handling Equipment: “Investigation report for standardization of ancillary facilities (6th report)”, 1988 (in Japanese). 2) Society for Mechanization of Port Cargo Handling Equipment: “Report of committee on researching of container handling equipment”, 1993 (in Japanese). 3) Architectural Institute of Japan: “Design Standard for Steel Structures”, 1994, p.4 (in Japanese). 4) Japan Society of Mechanical Engineers: “Mechanical Engineering Course, Cargo Handling Equipment”, 1959, 239 p. 5) Yoshio MINEMURA: “Method to fix rail”, Text of Track Maintenance Course, Track Maintenance Institute, 1958. p.4 (in Japanese). 6) Shoichi KITAJIMA, Osami HORII: “Mobile cargo handling equipment (mobile crane),” Technical Note of PHRI No.29, 1967, pp.4-62 (in Japanese).

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-558-

Part IX Other Port Facilities

PART IX OTHER PORT FACILITIES

Part IX Other Port Facilities Chapter 1 Port Traffic Facilities 1.1 General 1.1.1 Scope of Application The provisions in this chapter shall be applied to the design of facilities for land traffic in port zones. 1.1.2 Operation and Maintenance of Facilities for Land Traffic (Notification Article 103) The port land traffic facilities shall be operated and maintained in accordance with appropriate standards corresponding to their structural characteristics, so that vehicles, trains, and others can utilize them safely and smoothly.

1.2 Roads 1.2.1 General (Notification Articles 104, 109 and 110) (1) The road structure shall be determined appropriately in consideration of actual conditions of road use such as traffic generation, design traffic volume, topography of the area, smooth connection with other roads, etc. (2) The structure of roads used mainly by pedestrians and bicycles shall be determined in consideration of the state of utilization of the port facilities located nearby. (3) For the roads that are connected to the high seismic resistant facilities or connected to the open spaces that are designated as the place of temporary storage of relief supplies after strong earthquakes, the road structures shall be determined appropriately as to ensure the required functions at the occasion of earthquakes. (4) Unless otherwise specified in this section, matters concerning road structures, locations, and the related facilities shall be determined in accordance with the “Enforcement Regulations for Road Structures” (Government Ordinance No.320 of 1970) in light of the characteristics of traffic generated in ports and harbors. 1.2.2 Design Vehicles (Notification Article 105) In case that heavy traffic of semitrailer trucks is expected, such vehicles may be defined as the design vehicles. 1.2.3 Roadways and Lanes (Notification Article 107) (1) The number and dimensions of road lanes shall be determined appropriately so that the flow of traffic generated in the port shall not be congested. (2) When the total traffic volume employed in the design of roadway is less than the design standard traffic volume (defined by the maximum allowable total traffic volume per hour) for each type of road listed in Table 1.2.1, the number of road lanes (excluding ascent lanes, turning lanes or speed change lanes) shall be two. Table 1.2.1 Design Standard Traffic Volume for Different Kinds of Roads (Notification Article 107, Appendix Table 14) Type of road

Design standard traffic volume for double lanes (vehicles/hour)

Road connecting port and major road such as national highway

650

Other road

500

(3) In roads other than those specified in (2), there shall be four lanes or more (a multiple of two unless otherwise required by traffic conditions). The exact number of lanes shall be determined in

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accordance with the ratio of the design traffic volume of the road to the design standard traffic volume per lane for the same type of road listed in Table 1.2.2. Table 1.2.2 Design Standard Traffic Volume per Lane for Determination of Number of Lanes (Notification Article 107, Appendix Table 15) Type of road

Design standard traffic volume (vehicles/hour/lane)

Road connecting port and major road such as national highway

600

Other road

350

(4) The width of a lane shall be 3.25 m or 3.5 m in principle. The value of 3.5 m shall be the standard width for roads frequently used by large-size vehicles. The width may be reduced to 3.0 m as required by topographical constraints. (5) A stopping lane may be planned on the left side of the roadway as necessary to ensure safe and smooth traffic of vehicles. [Commentary] In general, the characteristics of peak traffic volume of the roads in ports can be estimated using the past traffic data in ports, because such traffic volume is generated mainly by the transport and distribution of cargo handled in the port. Then such characteristics of peak traffic volume can be considered in determination of design traffic volume. Therefore the standard design traffic volume in Tables 1.2.1 and 1.2.2 is defined as maximum allowable vehicle traffic volume per hour in order to determine the number of lanes. The “design traffic volume” means the sum of both-way automobile traffic volume per hour defined in light of future development in the surrounding area and the future conditions of motor vehicle traffic to be used as the basis in designing the road. In this context, consideration should be given to the traffic conditions generated in the port zone as well as to the characteristics of peak traffic. [Technical Notes] Matters to be considered in calculating the design traffic volume in the target year include the cargo generated from the wharves, the land use in the surrounding area, the concentration of vehicle traffic, and the connections with other roads. The estimate of design traffic volume generally begins with the trip generation and attraction, then proceed to the trip distribution, and is completed with the assignment of traffic volume to various routes. An example of methods to calculate the generated concentrated traffic volume is listed in Table T- 1.2.1. Table T- 1.2.1 Calculation of Trip Generation and Attraction Type of traffic

Calculation method

Traffic related to port cargo transportation

A method to use actual values of the modal split ratio of motor vehicles, load per truck, percentage of loaded trip, and related vehicle ratio, based on the volume of cargo handled at the port.

Traffic generated from industries located in the port zone

A method to use the number of trips generated per unit site area and the number of employees.

Traffic generated from green area or public attraction facilities

A method to use the data of similar facilities on the number of users and the percentage of motor vehicle use.

1.2.4 Clearance Limit (Notification Article 108) If traffic of special vehicles is expected, such as semitrailer trucks loaded with tall cubic containers and mobile cranes, the clearance limit shall be defined appropriately to ensure safe passage of such vehicles. [Technical Notes] The height of semitrailer trucks loaded with marine containers is listed in Table T- 1.2.2.

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Table T- 1.2.2 Size of International Marine Containers and Height of Semitrailer Truck when Loaded with Container Size

Size of international marine container

Type of container

Length (mm)

Width (mm)

Height (mm)

Height of semitrailer truck when loaded with container (m)

Remarks

ISO Type 1C 20¢ × 8¢ × 8¢

6,058

2,438

2,438

3.748

Flat-bed chassis

ISO Type 1CC 20¢ × 8¢ × 8¢6¢¢

6,058

2,438

2,591

3.796

Low-bed chassis

ISO Type 1A 40¢ × 8¢ × 8¢

12,192

2,438

2,438

3.643

Low-bed chassis

ISO Type 1AA 40¢ × 8¢ × 8¢6¢¢

12,192

2,438

2,591

3.796

Low-bed chassis

ISO type 1AAA 40¢ × 8¢ × 9¢6¢¢

12,192

2,438

2,896

4.101

Low-bed chassis

Note:

Floor height of chassis above the ground for international marine containers may differ slightly by manufacturers. Here, the height is considered to be 1205 mm for low-bed chassis and 1310 mm for flat-bed chassis.

1.2.5 Widening of Roads at Bends Lanes shall be widened at curves according to the design vehicle and the radius of the curve, unless topographical or other conditions require otherwise. [Technical Notes] The widening at bends listed in Table T- 1.2.3 is applicable to roads in the port zone. Table T- 1.2.3 Widening at Bends Design vehicle

Ordinary motor vehicle

Semitrailer truck

Widening margin per lane (m)

Radius of curve (m)

90 or over and less than 160 60 90 45 60 32 45 26 32 21 26 19 21 16 19 15 16

150 or overand less than180 100 150 70 100 60 70 50 60 40 50 32 40 28 32 25 28

0.25 0.50 0.75 1.00 1.25 1.50 1.75 2.00 2.25

1.2.6 Longitudinal Slope The longitudinal slope of a roadway shall be determined appropriately according to the design speed and design vehicle for the road under design. [Technical Notes] (1) The longitudinal slope of a roadway should be set within the standard values as listed in Table T- 1.2.4 according to the design speed and design vehicle for the road under design. When required by topographical conditions or other special reasons, the slope may be set within 2% to 3% above the values listed in the table. Table T- 1.2.4 Limit of Longitudinal Slope Design speed (km/h)

Design vehicle Ordinary motor vehicle

Semitrailer truck

2% 3% 4% 5% 6% 7% 8% 9%

1% 1% 2% 3% 4% 5% 6% 7%

120 100 80 60 50 40 30 20

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(2) When a slope extends only for a certain length, the special limit of longitudinal slope as listed in Table T- 1.2.5 may be referred to. Table T- 1.2.5 Special Limit of Longitudinal Slope for Limited Slope Length Design vehicle Ordinary motor vehicle

Design speed (km/h)

Semitrailer truck

Slope (%)

Limit of slope length (m)

Slope (%)

Limit of slope length (m)

120

3 4 5

800 500 400

2 3 4

1,000 500 400

100

4 5 6

700 500 400

2 3 4

1,600 800 500

80

5 6 7

600 500 400

3 4 5

1,300 700 500

60

6 7 8

500 400 300

4 5 6

600 400 300

50

7 8 9

200 150 100

5 6 7

250 150 100

40

8 9 10

100 100 100

6 7 8

100 100 100

1.2.7 Level Crossings The design of level crossings of roads in ports shall conform to the provisions of the “Enforcement Regulations for Road Structures”, while taking into consideration the traffic volume at the level crossing, the design vehicle, the conditions of surrounding road networks, the land use along the road, etc. [Technical Notes] The width of any additional lane (turning lane or speed change lane) at a level crossing may be determined with reference to Table T- 1.2.6. Table T- 1.2.6 Width of Lanes at Level Crossings (units in m) Width of lanes in basic roadway segment

Width of through lanes in creating additional lanes

3.50

3.50 (3.25)

3.25 (3.00)

3.25

3.25 (3.00)

3.00

Width of additional lanes

Note: The values in the parentheses are the reduced values when constrained by topographic conditions.

1.2.8 Pavement (Notification Article 106) The structure of pavement shall be determined according to the wheel load of running vehicles considering the traffic volume of special vehicles such as semitrailer trucks and mobile cranes. [Commentary] In general, roads are planned and designed according to the “Enforcement Regulations for Road Structures”. Pavement is designed based on the “Cement Concrete Pavement Guideline” 1), the “Manual for Asphalt Pavement” 2) and “Standard Specifications of Concrete (Pavement)” 3) established according to the “Enforcement Regulations for Road Structure”. Roads in port zones should therefore conform to these standards. [Technical Notes] Cement concrete pavement or asphalt pavement is used for roads in port zones in general. -562-


(1) The design of cement concrete pavement should follow the procedure depicted in Fig. T- 1.2.1. Determination of design conditions (design traffic volume in unit road section, traffic of semitrailor trucks or mobile cranes, meteorological conditions, etc.) ↓ Estimate of wheel load distribution of vehicles ↓ Determination of bearing capacity factor at subgrade ↓ Determination of bearing capacity factor at base course ↓ Calculation of base course thickness ↓ Calculation of concrete slab thickness ↓ Calculation of frost penetration depth ↓ Determination of total pavement thickness ↓ Determination of joint Fig. T- 1.2.1 Design Procedure for Cement Concrete Pavement

(2) The design of asphalt pavement should follow the procedure depicted in Fig. T- 1.2.2. Determination of design conditions (design traffic vol in unit road section, traffic of semitrailer trucks or mobile cranes, etc.) ↓ Determination of cumulative number of equivalent 5-ton wheel load ↓ Determination of design CBR ↓ Calculation of total pavement thickness ↓ Calculation of frost penetration depth ↓ Determination of total pavement thickness ↓ Determination of the thickness of each pavement part ↓ Determination of overall composition Fig. T- 1.2.2 Design Procedure for Asphalt Pavement

(3) The “Cement Concrete Pavement Guideline” 1) and the “Asphalt Pavement Guideline” 2) describe two methods of determining the traffic volume for the design of pavement structure: (a) method using large commercial vehicles per day and (b) method using the wheel load of vehicles.

1.2.9 Signs (Notification Article 102) (1) Port traffic facilities shall be equipped with appropriate signs according to the type and use of the facilities. (2) The style of signs shall be determined appropriately in light of the visibility to port users and the usage condition of traffic facilicities. -563-


(3) Signs shall be so placed that they may be clearly visible to port users and pose no obstacles to traffic or cargo handling.

1.3 Car Parks 1.3.1 General Car parks in ports (referring to the car park designated as port facilities) shall be in accordance with the standards for structure and equipment provided in the “Enforcement Order for the Car Park” (Government Ordinance No.340 of 1957). Other required matters shall be examined as necessnary. [Commentary] The basic law on the installation of car parks is the “Car Park Law” (Law No.106 of 1957). Standards for the structure and equipment of parking lots are specified in its “Enforcement Order”. Car parks in ports should also be in accordance with those standards. Special considerations for car parks in ports are to be specified separately. Standards provided in the “Enforcement Regulations for the Car Park Law” for the structure and equipment of car parks outside the roads are applicable to a car park with the parking area of 500 m2 or larger. Even if the car park is smaller, it is preferable to comply with them as much as possible when constructing a car park.

1.3.2 Size and Location (Notification Articles 114 and 115) (1) The size and location of a car park shall be determined so as not to pose any obstacle to the use of port facilities and the smooth road traffic, in consideration of the traffic generated in a port and the condition of roads in the vicinity. (2) A car park shall be located off the roads. If topographical conditions or other reasons necessitate a car park to be located on the roads, the size and location of a car park shall satisfy the following conditions: (a) Not to be located on a road connecting a port and a major inland highway. (b) Not to be located at a place which may hinder the vehicles from going to and from the cargo handling area or sheds. (c) Not to be located at a place adjacent to a dangerous cargo handling area, unless required otherwise by topographical conditions or other reasons. (3) The width of roadways in the car park, the size of parking lots, and the width of roadways for driving backward and turning cars into parking lots shall be determined appropriately according to the type of cars using the car park, the parking angle, and the parking method. [Technical Notes] (1) Practice of design may be defined with reference to the “Parking Lots Manual” 4), the “Standards of Car Park Design and Constraction with Commentary” 5) and “Commentary and Practice of Enforcement Regulations for Road Standards” 6). (2) Standard values for parking lot specifications and the configuration of parking lots may be determined with reference to Table T- 1.3.1 and Fig. T- 1.3.1. (3) See Part Ⅹ, 2.5.3 Car Parks for parking lots at ferry terminals.

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Table T- 1.3.1 Standard Values for Parking Lot Specifications

Special large vehicles

Large vehicles

Small vehicles

Type of vehicle

Roadway width Parking width at Aw (m) upper line right angles with road Aw1 lower line Sd (m) Aw2

Parking width in parallel with road Sw (m)

Unit parking width W (m)

Parking area required per vehicle A (m2)

Corresponding subfigure numbers in Fig. T- 1.3.1

Parking angle (º)

Parking method

30

Forward parking

4.00

4.50

4.50

6.50

29.3

(a)

45

Forward parking

4.00

5.10

3.20

7.10

22.8

(b)

45 cross parking

Forward parking

4.00

4.30

3.20

6.30

20.2

(c), (d)

60

Forward parking

5.00

5.45

2.60

7.95

20.7

(e)

60

Backward parking

4.50

5.45

2.60

7.70

20.0

(f)

90

Forward parking

9.50

5.00

2.25

9.75

21.9

(g)

90

Backward parking

6.00

5.00

2.25

8.00

18.0

(h)

30*

Forward parking Forward starting

4.00 6.00

9.30 9.30

6.50 6.50

19.30 19.30

125.5 125.5

(i) (i)

45*


7.00 6.50

11.50 11.50

4.60 4.60

25.00 25.00

115.0 115.0

(j) (j)

60*


11.00 7.50

12.90 12.90

3.75 3.75

31.40 31.40

117.8 117.8

(k) (k)

90*


19.00 11.00

13.00 13.00

3.25 3.25

43.00 43.00

139.8 139.8

(l) (l)

Parallel

Backward parking Forward starting

6.00 6.00

3.25 3.25

19.00 19.00

6.25 6.25

118.8 118.8

(m) (m)

30*


6.50 10.50

11.50 11.50

7.00 7.00

28.50 28.50

199.5 199.5

(n) (n)

45*


10.50 13.00

14.50 14.50

5.00 5.00

38.00 38.00

190.0 190.0

(o) (o)

60*


14.00 15.00

16.50 16.50

4.00 4.00

45.50 45.50

182.0 182.0

(p) (p)

90*


22.00 13.00

17.00 17.00

3.50 3.50

52.00 52.00

182.0 182.0

(q) (q)

Parallel

Backward parking Forward starting

6.00 6.00

3.50 3.50

25.00 25.00

6.50 6.50

162.5 162.5

(r) (r)

Note: * indicates that parking lots are placed in a single row in the direction of parking.

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(a) 30º forward parking (small vehicles)

(c) 45º cross parking type A (small vehicles)

(e) 60º forward parking (small vehicles)

(b) 45º forward parking (small vehicles) (i) 30º parking (large vehicles)

(j) 45º parking (large vehicles)

(k) 60º parking (large vehicles)

(l) 90º parking (large vehicles)

(d) 45º cross parking type B (small vehicles)

(f) 60º backward parking (small vehicles)

(m) Parallel parking (large vehicles) (g) 90º forward parking (small vehicles)

(h) 90º backward parking (small vehicles)

(n) 30º parking (special large vehicles)

(p) 60º parking (special large vehicles)

(q) 90º parking (special large vehicles)

(r) Parallel parking (special large vehicles)

(o) 45º parking (special large vehicles)

(Units in m)

Fig. T- 1.3.1 Configuration of Parking Lots

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1.4 Railways According to their classifications, railways in a port shall conform to the “Structural Rule of Common Railway” (Ministry of Transport Ordinance No.14 of 1987) or the “Structural Rule of Special Railway” (Ministry of Transport Ordinance No.19 of 1987). Tracks shall conform to the “Track Construction Regulation” (Ministry of the Interior and Railways Regulation of 1923).

1.5 Heliports Heliports shall conform to the “Civil Air Law” (Law No.231 of 1952).

1.6 Tunnels 1.6.1 General Tunnels as port traffic facilities shall conform to the provision in Article 12 (2) of the Technical Standards for Port and Harbour Facilities (the Ministrial Ordinance). The detailed specifications of tunnels shall conform to the following principles: (1) Notwithstanding the provision on ventilation in Article 12 (2) - 4 of the Ministrial Ordinance, installation of ventilators may be omitted in a railway tunnel as ventilation is ensured by passage of trains. (2) The provisins in this section shall be applied to the immersed road tunnels in a port. Tunnels for other usage or those of other types may comply with the relevant standards and specifications. [Commentary] (1) Tunnels as port facilities include road tunnels, railway tunnels, etc. From the viewpoint of construction techniques, mountain tunnels, cut and cover tunnels, shield tunnels, and immersed tunnels are included. The provisions in this section are applied specifically to road tunnels using the immersed tunnel method. (2) Structural components in this section are defined in Fig. C- 1.6.1. Ventilation towers should be installed as necessary. Ventilation tower

Open cut

On-ground section

Ventilation tower

Immersed tunnel element submerged section

Access road

On-ground section

Open cut

Access road Immersed tunnel section

Fig. C- 1.6.1 Definition of Components of Immersed Tunnels

[Technical Notes] For the design of immersed tunnel for port road, the “Engineering Manual of Immersed Tunnel” 7) may be referred to. For seismic design, the reference 8) is used.

1.6.2 Principle of Planning and Design (Notification Article 111) The location, alignment, and cross-sectional profile of a tunnel constructed with immersed box elements (hereinafter referred to as “immersed tunnel”) shall be determined appropriately according to the condition of tunnel use and the natural conditions. [Commentary] (1) The cross section of an immersed tunnel should be determined by taking into consideration the traffic volume of vehicles, the ratio of large vehicles in total traffic, the need for a sidewalk, the need for a bicycle track, the types of cables and pipes in a common duct, the allowance of hazardous goods transport, the existence of toll gates, and connections with other roads. Sufficient consideration should also be given to the future development plans of any other related facilities including those concerning the probability of deepening of a waterway above the immersed tunnel. Future utilization of the immersed tunnel should be carefully studied because any expansion of its function such as -567-


widening is difficult once it has been completed. (2) If pedestrian and bicycle tracks are to be installed, due consideration should be given to the use by the elderly and the physically handicapped persons. (3) As the safety facilities in the event of accident or disaster, evacuation paths and emergency telephones should be installed as necessary. Safety equipment and evacuation paths for fire should be provided, and the tunnel itself should be a fireproof structure. (4) The longitudinal slope of an immersed tunnel may be made as steep as possible within the restriction of design speed for the road to enable reduction of construction cost in general. However, consideration should be given to the fact that smoke concentration in the exhaust gas of vehicles increases rapidly as the slope becomes steeper, thus raising the cost of ventilation equipment. [Technical Notes] (1) According to the structure of immersed tunnel box elements, fireproof coating may be required. In such cases, the thickness of fireproof coating should be considered in determining the dimensions of inner section. (2) Under the provision of the “Road Law” (Article 46), the port management body has the right to prohibit or restrict the traffic of vehicles loaded with volatile or inflammable substances or other hazardous goods.

1.6.3 Depth of Immersion (Notification Article 112, Clause 2 and 3) (1) The top surface of immersed tunnel box elements shall be covered with appropriate materials of necessary thickness to ensure structural safety of the elements considering the penetration depth of ship’s anchors dropped and/or dragged, the frequency of anchor drops and dragging, the buoyancy of the tunnel, and scouring due to waves and currents, etc. (2) The depth of immersion shall be designed appropriately in consideration of any future plan of dredging in and around the area of the tunnel. [Technical Notes] The thickness of the coverage should preferably be 1.5 m or greater including the thickness of the protective concrete coverage on top of the upper slab.

1.6.4 Structure and Length of Immersed Tunnel Elements The structural type and the length of an immersed tunnel element shall be determined in consideration of the sectional stress, construction method, construction cost, fabrication yard, etc. [Commentary] (1) The structural types of immersed tunnel elements are classified into steel shell type, concrete (reinforced concrete, prestressed concrete) type, and composite (hybrid) type. The most appropriate structure should be selected in light of their individual characteristics. (2) The length of an immersed tunnel element should be determined in light of the sectional forces derived from design computation including seismic analysis, joint structures, the size of fabrication yard, execution methods of immersion and connection, and cost of construction including joint connection. [Technical Notes] Composite type elements, particularly that of steel-concrete sandwich structure, may be designed and constructed according to the reference 9).

1.6.5 Ventilation Towers (Notification Article 112, Clause 6) The structure of ventilation towers shall be designed with the appropriate method corresponding to the characteristics of the structure and the ground against the external forces and loads acting on it. [Commentary] (1) Ventilation machines, electric power equipment, and control equipment should be installed functionally in a ventilation tower. Its structure should be equipped with inlet and outlet ports for efficient ventilation as well as with ducts connected with the immersed tunnel itself. (2) Sufficient space should be provided inside a ventilation tower so that monitoring, inspection, and brief repair of the installed equipment can be conducted smoothly. In particular, large components such as ventilation machines should be so designed that their transportation in and out of the tower is easily executed. -568-


(3) The location and structure of inlet ports should be such that the intake of exhaust gas from the outlet or from the entrance of the tunnel is kept as little as possible. (4) The location of outlet ports should ensure that concentration of exhaust gas at the ground level remains under the permissible level.

1.6.6 Access Roads The structure of access roads shall be designed with due consideration to the traffic under planning, natural conditions, social conditions, construction methods and construction cost. 1.6.7 Calculation of Stability of Immersed Tunnel Section (Notification Article 112, Clause 4) It shall be standard to examine the structural stability of the immersed tunnel section in both the longitudinal and transverse directions of the tunnel. [Commentary] An immersed tunnel is considerably long in the longitudinal direction. Thus it requires examination of structural stability in the longitudinal direction as well as the transverse direction. The design loads to be considered are as follows: (1) Deadweight (2) Live load (3) Earth pressure (4) Hydrostatic pressure (5) Buoyancy and uplift (6) Subgrade reaction (7) Influence of settlement of the foundation (8) Influence of temperature variation (9) Load during construction (10) Influence of creep and shrinkage of concrete (11) Prestressing forces (12) Earthquake (13) Others (tsunami, anchor dropping and dragging, ship sinking, etc.) [Technical Notes] (1) Calculation in the transverse direction may regard an immersed tunnel itself as a rigid frame structure. (2) Calculation in the longitudinal direction of an immersed tunnel should consider the effects of earthquake ground motion, temperature variation, and settlement, by representing the tunnel as a beam supported on elastic springs of the ground. (3) It should be examined whether the foundation has the sufficient capacity to support the weight of the immersed tunnel including the soil on its top. Due consideration should also be given to the settlement of foundation. (4) The earthquake ground motion may be propagated to an immersed tunnel from every direction. However, the maximum bending moment and shearing force appear in the transverse direction, while the axial force becomes maximum in the longitudinal direction of the tunnel. For design purposes, examination should be made in these two directions. (5) An immersed tunnel may be constructed in a soft ground. In such cases, it should be confirmed that a slip failure will not occur in the surrounding area due to an earthquake. Examination against liquefaction should also be made when the tunnel is surrounded by loose, saturated sandy soil. (6) Appropriate materials should be selected for filling in light of the safety against subsidence and surfacing, liquefaction due to earthquake, and maintenance dredging of navigation channel. (7) An immersed tunnel is a structure under the seabed and is often constructed in soft ground. Sufficient examination should be made lest its functions should be lost due to water seepage from cracks or joints. (8) There are several examples of the tunnel design that introduced the earthquake load in design calculation; i.e., Dears Island Tunnel, Bart Tunnel, Osaka Sakishima Tunnel, Kobe Port Minatojima Tunnel, etc.

1.6.8 Design of Immersed Tunnel Elements (Notification Article 112, Clause 1) Immersed tunnel elements shall be designed by taking into consideration the following factors: (1) Watertightness (2) Cracks in concrete -569-


(3) Stability during towing and loading for immersion operation. [Commentary] The following factors should also be considered to ensure the structural safety of immersed tunnel elements: (1) Load applied during and after construction (2) Construction method (3) Surfacing of the element due to buoyancy after immersion (4) Ventilation and disaster prevention functions (5) Any facilities attached to the tunnel [Technical Notes] (1) Design and construction should be made to avoide the generation of cracks in concrete as much as possible. In addition, it is preferable to apply waterproof coating on the circumference of the element to ensure complete watertightness. (2) Cathodic protection should be applied to the outer steel shell plates.

1.6.9 Joints (Notification Article 112, Clause 5) Joints of immersed tunnel elements shall be of the structure ensuring safety against the stress and displacement during earthquakes as specified in Part Ⅱ, Chapter 12 Earthquakes and Seismic Force. [Commentary] The location and structure of joints of immersed tunnel elements are normally determined in consideration of the size of fabrication yard, shifting of navigation channels, capacity of construction machines, differential settlement of the foundation after completion, and influence of temperature variation. However, the location and structure of joints are also important factors in assessing the earthquake resistance of an immersed tunnel. Thus, the earthquake resistance needs to be adequately examined when determining the joint location and structure.

1.6.10 Control and Operation Facilities (Notification Article 113) Immersed tunnels shall be equipped with the following facilities as necessary: (1) Ventilation facilities (2) Emergency facilities (3) Lighting equipment (4) Electric power facilities (5) Security and instrumentation equipment (6) Monitoring and control facilities (7) Drainage facilities [Commentary] Ventilation is essential to immersed tunnels for preventing the adverse effect of exhaust gas from motor vehicles on the air inside tunnels. Although natural ventilation may be sufficient for short tunnels, ventilation facilities should be installed in principle for immersed tunnels for roads in port zones.

1.7 Bridges 1.7.1 General Article 12 (3) of the Ministrial Ordinance provides the rules concerning bridges. Bridges shall conform to the “Enforcement Regulations for Road Structures” (Government Ordinance No.320 of 1970) for road bridges, and the “Structural Rule of Common Railway” (Ministry of Transport Ordinance No.14 of 1987) or the “Structural Rule of Special Railway” (Ministry of Transport Ordinance No.19 of 1987) for railway bridges. 1.7.2 Design Requirements (Notification Articles 116 and 117) (1) A bridge crossing over the space above any navigation channel or basin shall meet the following requirements: (a) The bridge girders shall be constructed at an appropriate elevation above the highest high water level to ensure safe navigation of vessels. -570-


(b) The bridge piers shall not be located near the waterway nor obstruct the navigation of vessels, unless the safety of navigation is ensured otherwise. (c) Signs shall be posted as necessary to prevent vessels from colliding with the bridge girders or piers. (2) A bridge crossing over the space above any mooring equipments or cargo handling facilities shall meet the following requirements: (a) The location of piers and the elevation of girders shall be set appropriately so that they may not obstruct the safe and smooth use of the mooring or cargo handling facilities. (b) Signs shall be posted as necessary to prevent cargo handling equipment and vehicles from colliding with the bridge piers or girders. [Technical Notes] (1) The design of a bridge should take into consideration the future situation of the activities in the area when there is any areal development plan. (2) The decision on the clearance of a bridge over a navigation channel should be made by considering the following factors: (a) (b) (c) (d) (e)

Maximum mast height of the vessels entering the port Tide Trim of vessel Wave height Psychological impact on ship captains and mates

1.7.3 Structural Durability (Notification Article 118) (1) Design of bridge structure and selection of structural materials shall be made appropriately in consideration of the natural conditions surrounding the bridge. (2) Application of painting or other measures shall be made for steel bridges to prevent or reduce the corrosion of steel as necessary. (3) The following measures shall be taken for concrete bridges to prevent the deterioration of superstructure and substructure due to salt damage as necessary. (a) Retain the required thickness of protective covering. (b) Minimize the crack width on the surface of reinforced concrete. (c) Take the following countermeasures particularly under a harsh condition or when maintenance is difficult. ① Surface treatment of reinforced concrete ② Corrosion protection of reinforcement bars and prestressing steel [Technical Notes] The thickness of covering over reinforcement bars of a concrete bridge should be equal to or greater than the values listed in Table T- 1.7.1 10). Table T- 1.7.1 Minimum Thickness of Covering on Bars (Units in cm) Superstructure

Substructure Girder

Lower surface of slab, curb, and hand railing 5.0

Precast prestressed concrete girder made by pretensioning system

Other girders

5.0

7.0

Column

7.0

1.7.4 Fender System Fenders shall be installed on bridge piers as necessary to protect them against the impact of vessel collision as well as to reduce the damage to the colliding vessel.

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The fender requires appropriate maintenance that depends on its structure etc. to retain its functions in good conditions. [References] 1) 2) 3) 4) 5) 6) 7) 8)

Japan Road Association: “Cement Concrete Pavement Guideline”, April 1997 (in Japanese). Japan Road Association: “Manual for Asphalt Pavement”, February 1998 (in Japanese). JSCE: “Standard Specifications of Concrete (Pavement)”, 1996 (in Japanese). Japan Parking Engineering Committee: “Parking Lots Manual”, 1981, Additional Description,1990 (in Japanese). Japan Road Association: “Standards of Car Park Design and Construction with Commentary”, 1992 (in Japanese). Japan Road Association: “Commentary and Practice of Enforcement Regulations for Road Structures” (in Japanese). CDIT: “Engineering Manual of Immersed Tunnel”, April 1994 (in Japanese). Subcommittee on Seismic Dessign of Tunnels, Earthquake Engineering Committee, JSCE: “Basic Problems and Policy to Design of Seismic Design of Tunnels”, March 1998 (in Japanese). 9) CDIT: “Design of Steel-Concrete Sandwich Structure and Construction of High Fluidity Concrete”, November 1996 (in Japanese). 10) Japan Road Association: “Manual of Measure to Damage on Road Bridges from Salt Affect and Commentary”, 1984 (in Japanese).

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Chapter 2 Cargo Sorting Facilities 2.1 General The provisions in this chapter shall be applied to the design of cargo sorting facilities. [Technical Notes] References 1), 2), and 3) may be consulted for design.

2.2 Cargo Sorting Areas (Notification Articles 121 and 122) (1) The size of a cargo sorting area shall be designed appropriately by considering the kinds and quantities of cargoes handled and their handling conditions. (2) Cargo sorting areas shall be paved appropriately according to the purpose and condition of their use. The structure of pavement shall conform to the provisions in Part Ⅷ, 20.5 Design of Concrete Pavement and Part Ⅷ, 20.6 Design of Asphalt Pavement. (3) The widths and curves of passageways shall allow safe and smooth operation of vehicles and cargo handling equipment. (4) A cargo sorting area shall be provided with drainage facilities such as drain ditches. (5) The provisions of Part Ⅷ, 19.12 Lighting Facilities shall be applied to the lighting equipment in the cargo sorting areas that are operated at night. (6) A cargo sorting area that is dangerous for the public shall prohibit the entrance of public with provisions of signs, boards, and fences. (7) In a cargo sorting area where the contents of cargo may be blown and scattered by wind, walls or other proper measures shall be provided around its perihery.

2.3 Quay Sheds (Notification Articles 125 and 126) (1) Quay sheds shall meet the following requirements to ensure smooth cargo handling before the entrance and after the departure of vessels. (a) The size of a shed shall be determined appropriately by considering the kinds and quantities of cargoes and their handling conditions. (b) The widths and curves of passageways in a shed shall be determined to allow safe and smooth operation of cargo handling equipment. (2) Where dust is raised by cargo handling works, proper ventilating equipment shall be provided as necessary. (3) Quay sheds handling cargoes at night shall be equipped with appropriate lighting facilities as provided in Part Ⅷ, 19.12 Lighting Facilities. (4) When there is danger of inundation due to a storm surge, a tide gate or any other equipment shall be installed for preventing the intrusion of sea water. (5) Quay sheds shall be equipped with the provisions of appropriate signs to ensure safe and smooth use as necessary. [Commentary] The structure of quay sheds should comply with the provisions of the “Building Standard Law” (Law No.201 of 1950).

2.4 Cargo Handling Equipment 2.4.1 General Cargo handling equipment installed in cargo sorting facilities or mooring equipments shall have the required structure and capacity, and be located at the position most appropriate to the purpose of the facilities. It shall be guaranteed of the structural safety, be provided with the anti-pollution systems against dusts and noise, and ensure smooth and safe operation in cargo handling activities. -573-


[Commentary] (1) The purposes of cargo handling equipment in ports are to reduce labor works, to speed up cargo handling activities, and to ensure safety. The selection of the type, structure and capacity of cargo handling equipment should be made by considering the vessels to be served, the type, shape, volume, and particular conditions of handled cargoes, as well as the relations with yard facilities behind, and the mode of secondary transportation. (2) Sufficient space should be secured around cargo handling equipment for its various operations by clearing the space of obstacles such as buildings and electric wires. Cargo handling equipment should be designed and located so that it should not come into contact with vessels at berthing and leaving or while mooring. (3) Countermeasures against noise and dusts are required for equipment handling bulk cargoes because their handling is likely to produce noise and dusts. In particular, inflammable dusts require countermeasures against explosion. [Technical Notes] At the time of an earthquake, the behavior of cargo handling equipment including cranes may be affected by the ground conditions on which it is located 4), 5), 6). Container cranes are usually designed according to the crane structural specifications with no consideration of dynamic interaction with the quay structures. As is the case with high seismic resistant quaywalls, however, the impact of dynamic interaction between the quay and the cranes placed thereon during a larger-scale earthquake should be amply examined in design. Particularly in the case of an open-type wharf (i.e., pier), the impact of dynamic interaction between cranes and the pier may be significant under certain conditions. The impact of dynamic interaction between container cranes and the pier may differ according to their mass ratio, the natural period of container cranes, the natural period of the pier, and the input earthquake ground motion. Such impact should be examined in designing container cranes through the seismic response analysis of container cranes and the pier loaded with them, or by other appropriate method allowing sufficient consideration of dynamic interaction between container cranes and the pier 7). Detailed design should be referred to “Guideline for Seismic Design of Container Crane” 8).

2.4.2 Oil Handling Equipment (Notification Article 127) (1) Loading arms shall be of a structure ensuring safety against the weight and pressure of oil within them, their deadweight, and the loads generated by wind pressure and seismic force. (2) The distance between loading arms and the face line of the mooring facilities shall be determined appropriately in light of the length of the arms and the height of fenders so that they may pose no obstacles to oil handling. 2.4.3 Operation and Maintenance of Cargo Handling Equipment Cargo handling equipment shall be appropriately maintained to retain its operational performance in good condition as well as to enable to take preventive measures against accidents 9), 10), 11).

2.5 Timber Sorting Areas (Notification Articles 123 and 124) (1) The appropriate location and size of the land or water surface shall be secured for a timber sorting area to enable safe and smooth timber handling. (2) In timber sorting areas, measures allowing incineration shall be taken as necessary, including the installation of special plants for bark incineration. [Commentary] (1) Security measures against drift wood and sunken wood should be taken as well as environmental preservation measures against bark scattering. (2) Timber sorting ponds should be located in calm water area. (3) Timber sorting ponds should be equipped with facilities preventing the drifting of timbers. (4) Timber-handling water area should be so located that the waterway connecting the mooring basin for timber rafts and the sorting pond or timber yard will not obstruct the navigation of vessels by crossing major navigation channels or mooring basins. (5) See Part Ⅶ , 2.10 Breakwaters for Timber Handling Facilities. [Technical Notes] See references 12) and 13) for the configuration of timber sorting facilities. -574-


2.6 Sorting Facilities for Marine Products Sorting facilities for marine products shall be designed to ensure synergistic relations among their functions. [Commentary] Marine product sorting facilities consist of marine product sorting areas and related ancillary facilities. Marine product sorting areas deal with all activities ranging from the water-washing to carrying-out of the catch, and are covered with roofs for the maintenance of freshness, the prevention of pollution, and the preservation of working conditions. Ancillary facilities include loading area, icehouse, bidding room, and the other facilities, which contribute to enhancement of the functions of marine product sorting areas, as well as wastewater treatment facilities and freezer facilities. [Technical Notes] See references 14), 15), 16) and 17) for the surface area required for marine product sorting and the planning of ancillary facilities.

2.7 Sorting Facilities for Hazardous Cargo (1) Sorting facilities for hazardous cargo shall conform to the “Fire Service Law” (Law No.186 of 1948), the “High Pressure Gas Security Law” (Law No.204 of 1951) and regulations based on these laws. (2) If the handled cargo itself does not constitute hazardous goods but is subject to fumigation under the “Imported Plant Quarantine Regulation” (Ministry of Agriculture and Forestry Notification No.206 of 1950), it is preferable to post signs around the cargo sorting area or shed conducting fumigation clearly indicating that they are off limits to the public due to danger. [References] 1) Shingo FUJINO, Yoshikazu KAWASAKI: “Port Planning”, New Civil Eng. Series No.81, Gihodo, 1981, pp.135-138 (in Japanese). 2) Shingo NAKAYAMA: “Port Engineering”, Sankaido, 1985, pp.36-37 (in Japanese). 3) Japan Soc. Civil Engrs., “Civil Engineering Handbook (4th Edition)”, Gihodo, 1989, pp.1620-1621 (in Japanese). 4) Suketo TANAKA, Takamasa INATOMI, Eiju IIZUKA: “Seismic response analyses of gantry cranes of container berth and the time of Hyougo-South Earthquake”, Proc. 23th Conf. Earthquake Eng., 1995, pp.213-216 (in Japanese). 5) Takamasa INATOMI, Suketo TANAKA: “Seismic characterristics of port cargo handling equipment”, Proc. 4th Conf. on Cargo Transportation and Distribution, Japan Soc.Mechanical Engrs., 1995, pp.62-67 (in Japanese). 6) Suketo TANAKA, Takamasa INATOMI, Eiju IIZUKA: “Basic research for seismic response analyses of gantry cranes of container berth at the time of Hyougo-South Earthquake”, Jour. Structural Concrete, Vol.42A, 1996, pp.1229-1234 (in Japanese). 7) Masashi MIYATA, Takahiro SUGANO, Junji HAMADA, Toshio TANABE, Susumu NAKAJIMA, Takeshi TANAKA: “Research study on seismic response interaction between piled pier and container cranes”, Proc. 10th Japan Earthquake Eng. Symp., 1998, pp.635-640 (in Japanese). 8) Yoshiharu NAKAZONO: “Commentary to guideline for seismic design of container crane”, Jour. Port Cargo Handling, Vol.43, No.6, 1998, pp.635-640 (in Japanese). 9) Society of Mechanization of Port Cargo Handling Equipment: “Code of Maintenance of Pulling Type Unloader Crane”, 1979 (in Japanese). 10) Society of Mechanization of Port Cargo Handling Equipment: “Code of Maintenance of Container Crane”, 1980 (in Japanese). 11) Society of Mechanization of Port Cargo Handling Equipment: “Tables of Maintenance of Container Crane”, 1983 (in Japanese). 12) Ibid. 1), pp.184-189 (in Japanese). 13) Ibid. 3), p.1605 (in Japanese). 14) National Fishing Port Association, “Manual of Fising Port Planning (’92 Edition)”: 1992, pp.109-135 (in Japanese). 15) Toshio SATO, “Planning of land use for fishing port facilities (4),” Jour. Fishing Port, Vol.13, No.4, 1971, p.51 (in Japanese). 16) Toshio SATO, “Planning of land use for fishing port facilities (5),” Jour. Fishing Port, Vol.14, No.1, 1972, p.40, p.42 (in Japanese). 17) Toshio SATO, “Planning of land use for fishing port facilities (11),” Jour. Fishing Port, Vol.16, No.2, 1974, p.28-33 (in Japanese).

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Chapter 3 Storage Facilities 3.1 General The provisions in this chapter shall be applied to the design of storage facilities.

3.2 Yards for Dangerous Cargo and Oil Storage Facilities (Notification Article 129) (1) Yards for dangerous cargo and oil storage facilities shall be located in a close proximity, unless topographical conditions or other reasons require otherwise. (2) The open space with an appropriate width shall be reserved around the yards for dangerous cargo and oil storage facilities in correspondence to the type of dangerous cargo and the structure of the facilities.

3.3 Other Storage Facilities (Notification Article 130) Other storage facilities shall conform to the provisions in Chapter 2 Cargo Sorting Facilities and this chapter to ensure the functions required for facilities. As necessary, the structure and location of these facilities shall be determined appropriately according to other related provisions. [Commentary] Other storage facilities include warehouse, open storage yard, timber yard, and coal storage yard. (1) Warehouse Warehouses should ensure safe and smooth storage in line with the standards for warehouses provided in Article 3 of the “Regulation on Execution of Warehousing Business Law” (Ministry of Transport Regulation No.59 of 1956). (2) Open Storage Yard Open storage yards should conform to the provisions in 2.2 Cargo Sorting Areas. The surface area of an open storage yard should be calculated in consultation with reference 1). (3) Timber Yard Timber yards should conform to the provisions in 2.5 Timber Sorting Areas. (4) Coal Storage Yards [Reference] 1) Japan Soc. Civil Engrs.: Civil Engineering Hanbook (4th Edition)”, Gihodo, 1989, p.1606 (in Japanese).

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Chapter 4 Facilities for Ship Services 4.1 General The provisions in this chapter shall be applied to the design of facilities for ship services. [Commentary] Ship service facilities include the water, oil and coal supply facilities for ships (excluding water boats, bunker boats, tank trucks, and other mobile facilities), as well as ship repairing facility and ship storage facility.

4.2 Water Supply Facilities (Notification Article 132) (1) The layout and capacity of hydrants shall be determined appropriately according to the type of the vessels. (2) Water supply facilities shall meet the following sanitation requirements: (a) Hydrants shall be of a structure enabling the prevention of water pollution. (b) Periodic and occasional water examinations shall be conducted, and hydrants shall be kept clean. [Commentary] (1) Intakes of hydrants should be located in a manner that facilitates the attachment of hoses and should have a structure allowing the prevention of water pollution. In particular, drainage should be provided when they are installed beneath the apron floor. (2) Intakes should be equipped with caps. (3) Periodic and occasional water examinations should be conducted according to Article 15 of the “Regulation on Execution of Water Works Law” (Ministry of Health and Welfare Regulation No.45 of 1957). Periodic inspections should also be conducted to ensure that hydrants are kept clean.

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Chapter 5 Facilities for Passenger 5.1 Facilities for Passenger Boarding 5.1.1 General The provisions in this section shall be applied to the design of fixed and movable passenger boarding facilities (hereinafter referred to as “passenger boarding facilities”). [Commentary] (1) Passenger boarding facilities should have the safe and efficient functions for use by passengers. They should be separated from boarding facilities for vehicles. (2) Passenger boarding facilities should not cause passengers to feel danger. They should also have a stable structure against ship motions and winds.

5.1.2 Structural Types (Notification Articles 135 and 136) (1) The structure of passenger boarding facilities shall meet the following requirements. (a) The width of a passage shall be set at 75 cm or more considering the conditions of use. (b) A passageway shall have ancillary provisions such as the sidewalls and hand rails on both sides, and a skid proof agent shall be applied or a non-slippery material shall be used on the surface of passageway. (c) The rise of step of stairway shall be set in light of safety. A landing shall be set on the stairway as necessary. (d) The boarding facility shall not be used both for passengers and vehicles. If passengers can be separated from vehicles, however, the facility may be used for both passengers and vehicles. (e) The gradient of a slope shall be determined appropriately in light of safety. (2) The allowable vertical displacement at the seaward end of a movable bridge for passenger boarding shall be determined in light of the tidal range, the change in the molded draft, and the pitching and rolling of ships. [Technical Notes] (1) Although the width of a passage requires a minimum of 75 cm, it is preferable to set at 1.2 m or more in light of the use by the elderly and the physically handicapped persons. (2) The gradient of a slope is normally be 12% or less, but preferably be 5% to 8% or less in light of the use by the elderly and the physically handicapped persons. (3) Sufficient care should be taken to enable safe passage of the elderly and the physically handicapped persons in wheelchairs 1).

5.1.3 Design of Facilities for Passenger Boarding Passenger boarding facilities shall be of safe structures and designed according to the “Specifications for Highway Bridge and Commentary” (Japan Road Association) 2) and the “Technical Standards for Installation of Grade Separation Facilities” (Japan Road Association) 3). Loads and external forces to be considered in design are the deadweight, live load, wind load, load by an earthquake, thermal stress by temperature variations, etc. 5.1.4 Ancillary Facilities Ancillary facilities shall be equipped appropriately in light of the safety of passengers. [Technical Notes] (1) Handrails not only represent a preventive measure against falling but also work to ensure smooth passage by alleviating passengers’ feel of danger. The height of handrails should be 1.1 m or more, which may prevent an average-sized adult from jumping over the handrail. However, the handrails may not work effectively by itself for infants, children, and users in wheelchairs. Therefore, stays, crosspieces, and/or netting should be installed.

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(2) Fences, ropes, and chains should be provided at the gateways, in order to guiding passengers safely to boarding facilities. Fences should be at least 70 cm high to ensure safety. Ropes and chains should be extended tightly and should not be loosened unnecessarily. (3) When a roof is provided, the height should be 2.1 m or more. (4) Emergency exits should be provided when the length of the passageway exceeds 60 m. The distance between entrance / exit and emergency exit or that between emergency exits should be 60 m or less. Marking signs or signs indicating an emergency exit should be provided in the passageway.

5.2 Passenger Building 5.2.1 General [Commentary] Passenger building should be equipped with the following facilities as necessary: (a) (b) (c) (d) (e) (f)

Ticket window Waiting room CIQ facilities Baggage room, toilet, telephone, and other facilities serving passengers Appropriate lighting equipment Alarm device and other communication equipment

[Technical Notes] The passenger building and ancillary equipment should be designed to allow safe movement of the elderly and the physically handicapped persons in wheelchairs.

5.2.2 Design of Passenger Buildings The design of passenger buildings shall consider the conditions of location, the characteristics of the facility, and the number of passengers. [Technical Notes] The calculation of the floor area of a passenger building should consider the type of ships, the number of sailings per day, the number of passengers, the concentration rate, the quantity of baggage handling, and the number of workers. An appropriate size of the floor area should be calculated for ticket window, waiting room, baggage room, CIQ facilities, office, worker station, toilet, telephone, and other facilities serving passengers.

5.2.3 Ancillary Facilities Passenger buildings shall be equipped with required facilities for the safety, sanitation, and amenity of passengers. [Technical Notes] Required facilities include the following. (1) Evacuation facilities (2) Ventilation facilities (3) Lighting equipment (4) Fire fighting facilities and alarm device (5) Fences, ropes, and other similar facilities (6) Signs (7) Toilet (8) Telephones

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Boarding gate Sound guidance system

Handrail Gangway

Shop

Public phone Telephone with facsimile Toilet for physically handicapped

Escalator Elevator Handrail

Water fountain

Guidance display for departure and arrival (super LED)

Sound guidance system

Flickering guiding light with sound guidance equipment

Counter Parking lots for physically handicapped

Office

Restaurant

Telephone with facsimile Public phone Toilet for physically handicapped Billboard Sound detecting guide map Automatic door Bench Sound guidance system Low counter Guidance display Sound detecting guide map Handrail Slope

Fig. T- 5.2.1 Model Passenger Building

[References] 1) Japan Transport Economics Research Center: “Guideline for Facility Building for Elderly and Handicapped Peoples in Public Passenger Terminals”, 1994 (in Japanese). 2) Japan Road Association: “Specifications of Highway Bridges”, 1996 (in Japanese). 3) Japan Road Association: “Technical Standard for Installation of Grade Separation Facilities”, 1979 (in Japanese).

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Part X Special Purpose Wharves

PART X SPECIAL PURPOSE WHARVES

Part X Special Purpose Wharves Chapter 1 Container Terminals 1.1 Principle of Design A container terminal serves as a focal point of marine container transport. It shall be planned and designed with due considerations for the factors such as the trend of container transport, the economic activities of its hinterland, the conditions of connected transportation, and the space of land available. Because the smooth and efficient operations of container loading / unloading and storage of containers and container cargo must be assured, a container terminal needs various facilities that are installed with proper arrangement for their full functioning, and occupies a sufficiently wide area. [Technical Notes] (1) The size of a container terminal and the arrangement of facilities should be carefully determined in consideration of the following requirements: (a) It assures efficient cargo handling. (b) It enhances the total system economic efficiency of land and sea container transportation: in particular, smooth and efficient connection with land transportation systems is assured. (c) It deals flexibly with future expansion of the terminal and innovations in container transportation and handling. (2) A basic plan for container terminal facilities should be formulated by considering the following factors: (a) (b) (c) (d) (e) (f) (g) (h) (i) (j)

Design volume of containers to be handled Cargo movement characteristics (export and import ratio, transit ratio, etc.) Calling interval and types of container ships Operation and management method of the terminal Method of container handling at quays and yards Area and shape of available land space Conditions of storage facilities in direct hinterland Transportation mode to hinterland and the traffic conditions Land use and ship navigation conditions in the surrounding area Activities of neighboring container terminals

(3) For planning and design of an efficient container terminal, detailed analysis should be conducted on the arrival and departure of container ships, the movement of container cargo and containers in the terminal as well as the conditions of transportation to and from hinterland. Such analysis should consider the following factors: (a) System characteristics of the container terminal ① Service time of the terminal (annual and daily service time of gates and container yards) ② Container ship arrivals (distribution of arrival times) ③ Distribution of loading ratio and number of loading / unloading containers of each ship ④ Types of handled containers and conditions of cargo lot (the ratio of LCL cargo etc.) ⑤ Container collection and delivery (distribution of collections and deliveries) ⑥ Container staying time in the terminal ⑦ Distribution of empty containers (b) Characteristics of container handling plans including the plan for storage in yard and handling at quay, as well as the number of available cargo handling equipment and its work efficiency (c) Specifications of equipment on quay and facilities in the container terminal (d) Development costs of the container terminal, cargo handling equipment and related facilities, as well as the total management cost Fig. T- 1.1.1 shows an example of facility layout plan of an ocean-line container terminal.

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Transformer Substation

Maintenance Shop

Fuel Station Washing Facility Container Freight Station Receptacle Container Reefer

Transformer Substaition

Container Yard Fender

Crane Rail

Container Ship

350m

Container Crane

Green Belt

Road

Office

Gate

Lighting Tower 350m

Fig. T- 1.1.1 Example of Facility Layout Plan of Ocean-line Container Terminal (Kobe Port)

1.2 Design of Mooring Facilities 1.2.1 Length and Water Depth of Berths The length and water depth of a berth for mooring a container ship shall be determined to allow safe and smooth berthing of the target vessel. [Technical Notes] (1) Ships dealing with container transport include LO/LO (lift-on/lift-off) ships, RO/RO (roll-on/roll-off) ships, and semi-container ships, each having a characteristic shape. Ships of the same type may also differ in shape depending on the shipping company. Therefore, if the type and size of ship using the berth is specified, the length and water depth of the berth should accommodate that ship. However, if the dimensions of the ship are unknown at the stage of design, the berth length and depth can be determined as listed in Table T- 1.2.1. Table T- 1.2.1 Berth Length and Depth of Container Wharf Size of ship (DWT)

Berth length (m)

Berth depth (m)

30,000 40,000 50,000 60,000

250 300 330 350

12.0 13.0 14.0 15.0

(2) As in Part Ⅷ , 2.1 Length and Water Depth of Berths, standard specifications of berths in Table T- 1.2.1 are based on Part Ⅱ, 2.1 Dimensions of Target Vessel.

1.2.2 Mooring Equipment Mooring equipment shall be installed by considering the size and type of the container ships using the berth and in compliance with Part Ⅷ , 19.3 Mooring Posts, Bollards and Mooring Rings. [Commentary] In comparison with general cargo ships using a berth with the same length and water depth, a container ship has a larger ratio of gross tonnage to deadweight tonnage and a larger projective area above the water surface, because containers are loaded even on the deck. Therefore, the structure of mooring equipment should be designed by considering the fact that container ships generally have large projected areas above the water surface, which are subject to the wind force.

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1.2.3 Fender System Fender system shall be installed by considering the size and type of the container ships using the berth and in compliance with Part Ⅷ , 19.4 Fender System. [Commentary] A container ship has larger displacement tonnage corresponding to the deadweight tonnage than that of a general cargo ship using a berth with the same length and water depth. Specifications of fender systems should therefore be determined in this light.

1.3 Design of Land Facilities 1.3.1 Apron The size and facility design of apron shall be determined to allow the safe and smooth operation of temporary placing of containers and hatch covers of container ship, and that of vehicles and equipment for container handling. 1.3.2 Container Cranes Container cranes shall have appropriate capacities by considering the size of the container ships to be served, the size and type of containers, the number of containers to be handled, quaywall structure, the method of cargo handling in the yard, as well as the type of cargo handling facilities and equipment in the yard. 1.3.3 Container Yard The size and layout of a container yard shall be determined appropriately to ensure smooth storing, marshalling, receipt, and dispatch of containers or chassis. Facilities in the yard shall be designed accordingly. 1.3.4 Container Freight Station A container freight station shall be installed as necessary for the handling of small lots of cargo in the yard. Its location shall be determined in considering traffic stream lines in the yard. The size of container freight station and the cargo handling equipment to be used shall be determined appropriately to ensure safe and smooth handling and temporary storage of cargo. The container freight station shall then be designed accordingly. 1.3.5 Maintenance Shop The location and size of a maintenance shop shall be determined to allow the smooth inspection and repair of containers and the inspection, maintenance, and repair of vehicles and container handling equipment. 1.3.6 Administration Building The location and size of an administration building shall be determined to allow the smooth management and operation of the container terminal. 1.3.7 Gates The location and size of gates shall be determined to allow the efficient check and weighing of containers delivered to and from the container terminal as well as the smooth transfer of documents. 1.3.8 Ancillary Facilities In a container terminal, if necessary, ancillary facilities such as a washing station, sewage treatment plant, oil station, transformer substation, parking lots, and chassis yards shall be provided.

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Chapter 2 Ferry Terminals 2.1 Principle of Design A ferry terminal shall be designed with particular attention paid to both structural stability and safety of users, because it is usual that a ferry terminal is utilized simultaneously by both passengers and vehicles, day and night, [Commentary] (1) The location of ferry terminals should be selected by considering its interrelationship with other port facilities. Considerations should also be made for the characteristics of ferry lanes, the size and use of the terminal, as well as road traffic conditions in the hinterland. (2) The configuration and location of facilities in a ferry terminal should be determined so that the navigation of ferryboats, the rolling-on and rolling-off of vehicles, the embarking and disembarking of passengers, and the queuing or parking of vehicles are ensured in a safe and smooth manner. [Technical Notes] (1) In addition to navigation channels, basins and mooring equipments, a ferry terminal should be equipped with the following facilities (see Fig. T- 2.1.1). (a) (b) (c) (d) (e) (f)

Facilities for vehicle boarding Facilities for passenger boarding Roads Car parks Passenger terminal building Safety facilities

(2) In addition to normal mooring equipment, a ferry terminal should be equipped with a shift berth as necessary. A shift berth serves ferryboats put on hold or receiving oil and water supply so as to improve the efficiency of mooring equipment utilization. (3) External forces to be considered in designing ferry terminals include the wave force, earth pressure, wind pressure, seismic force, fender reactions, tractive force of ferryboats, and traffic loads. For safety purposes, particular attention should be paid to the mode of exertions of these external forces.

5m )

Legend Quay wall (-7.5m) Pier (-7.5m)

uay

wa

ll (

-5.

Water supply facility

Mooring basin (-7.5m)

bli cq

Car park

Seawall

Pu

Car park

Vehicle boarding facility Car park

Passenger boarding facility Passenger terminal Road No.11

1

2

Berth No.

Berth No.

5m) Wharf (-7. = 430m 5m 21 @ 2 berths

Seawall 257m

At

tac hm

ent

51 .4

3m

Car park

Mooring basin

Car park

3

Berth No.

Pier 190m

Viaduct

Berth No.4

Mooring basin

Shallow draft quay (-3.5m) Bridge

Fig. T- 2.1.1 Example of Ferry Terminal Plan

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2.2 Design of Mooring Facilities 2.2.1 Length and Water Depth of Berths The length and water depth of berths for mooring ferryboats shall be determined appropriately to ensure the safe and smooth use by the target ferryboat. [Commentary] The length and water depth of a berth can be determined with reference to Part Ⅷ , 2.1 Length and Water Depth of Berths. Consideration should also be given to the location of vehicle ramps and the method of ship maneuvering including the use of side thrusters. [Technical Notes] The length and water depth can be determined with reference to the values listed in Table T- 2.2.1 if the target ferryboat cannot be identified in advance. Table T- 2.2.1 Length and Water Depth of Ferry Terminal Berths for Japanese Ferries

Short- and medium-distance Japanese ferries (sailing distance less than 300km) Target ferry gross tonnage (GT)

Wharf for ferries with bow and/or stern ramps Quay length of bow and stern side (m)

Length of berth (m)

Water depth of berth (m)

60 80 90 130 160 180

3.5 4.0 4.5 5.5 6.5 7.0

20 20 25 25 30 30

400 700 1,000 2.500 5,000 10,000

Long-distance Japanese ferries (sailing distance 300km or more) Target ferry gross tonnage (GT)

Wharf for ferries with side ramps Length of berth (m)

6,000 10,000 13,000 16,000 20,000 23,000

Wharf for ferries with bow and/or stern ramps Quay length of bow and stern side (m)

Length of berth (m)

Water depth of berth (m)

30 30 35 40 40 40

170 200 220 230 230 240

7.0 7.5 8.0 8.0 8.0 8.5

190 220 240 250 250 260

2.2.2 Mooring Equipment Mooring facilities shall be provided with mooring equipment. The structure and arrangement of mooring equipment shall be determined by considering the layout of mooring facilities and the size of ferryboat as well as the wind force acting on the ferryboat. [Commentary] (1) The design of mooring equipment should be carried out by taking into account the fact that ferryboats are generally vulnerable to wind force because of their large exposed areas. The wind force and tractive force should be calculated according to Part Ⅱ, 2.2.3 [3] Wind Load Acting on Vessel and Part Ⅱ, 2.2.4 Tractive Force Acting on Mooring Post and Bollard, respectively. (2) Standard locations of mooring posts are as follows: (a) For mooring facilities for ferryboats with bow and stern ramps, one or more mooring posts around the tip of the berth and two or more mooring posts around the base of the berth. (b) For mooring facilities for ferryboats with side ramps, one or more mooring posts around both ends of the berth. (3) See Part Ⅷ , 19.3 Mooring Posts, Bollards and Mooring Rings for bollards and other mooring equipment.

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2.2.3 Fender System Mooring facilities shall be provided with a fender system to protect the ferryboat’s hull from damage at loading and berthing, and damage due to tide level changes. [Commentary] (1) Fender systems should be designed in accordance with Part Ⅱ, 2.2 External Forces Generated by Vessels. (2) Fender systems should be installed considering the vertical range of the hull movement of ferryboat.

2.2.4 Protection Works against Scouring For a ferry terminal, considerations shall be given for protection works against scouring in front of mooring facilities. [Commentary] Most ferryboats are equipped with side thrusters to facilitate berthing and unberthing operations. The design of mooring facilities should take into account the fact that use of side thrusters tends to cause more scouring than the case without side thrusters. (see Part Ⅷ , 2.5 Protection against Scouring.)

2.3 Design of Vehicle Ramp The design of vehicle ramp shall be made by considering the tidal range at the berth, the draft of the ferryboats to be served, the location of vehicle ramp on ferryboats, and the ferryboat motions at the berth, so as to ensure safe and smooth roll-on and roll-off operations of vehicles. 2.3.1 Width, Length, Gradient, and Radius of Curvature The width, length, gradient, and radius of curvature of vehicle ramp shall be set in accordance with PartⅧ Ⅷ, 19.6.4 Vehicle Ramp. 2.3.2 Ancillary Facilities and Signs (Notification Article 93, Clauses 1 and 5) Signs and notices shall be posted appropriately according to the structural characteristics and utilization conditions of the facilities. 2.3.3 Design of Moving Parts (1) The structure of a moving part shall be designed by examining the stresses, deformations, and stability so that sufficient safety is ensured during the operation of the ramp. (2) As the external forces acting on the bridge of the moving parts, the following shall be considered: (a) Deadweight, surcharge (vehicle load and sidewalk live load), and impact load as the primary loads. (b) Wind load, thermal stress, and seismic force as the secondary loads. (c) If necessary, the other loads such as snow load, braking load, and wave uplift shall be considered. (3) A lift and its ancillary equipment shall be so designed that they would function on the safer side in the events of the power shutdown and the excessive vibrations. [Technical Notes] (1) See Part Ⅱ, 15.4 Live Load for traffic load. (2) The impact load should be defined as 40% of traffic load. (3) The design of moving parts should be made as necessary against either the T traffic load or the L traffic load defined in the “Highway Bridge Specification and Commentary” whichever exerts more adverse stress.

2.4 Facilities for Passenger Boarding The design of passenger boarding facilities shall be made by considering the tidal range at the berth, the draft of the ferryboats to be served, the location of boarding facilities on ferryboats, and the ferryboat motions at the berth, so as to ensure safe and easy embarking and disembarking of passengers.

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[Commentary] (1) Passenger boarding facilities should be installed separately from vehicle boarding facilities as a general rule. (2) The following requirements should be satisfied when the same boarding facilities have to be utilized for both passengers and vehicles. (a) Passengers should be separated from vehicles by fences or ropes. (b) Sufficient safety measures should be taken, including the placement of guides to separate the time of use between passengers and vehicles.

2.4.1 Width, Length, Gradient, and Ancillary Facilities The width and other aspects of passenger boarding facilities shall be set in accordance with Part Ⅸ, 5.1 Facilities for Passenger Boarding. 2.4.2 Design of Moving Parts The design of moving parts shall be made in accordance with 2.3.3 Design of Moving Parts.

2.5 Design of Other Facilities 2.5.1 Roads The design of roads shall be made in accordance with Part Ⅸ, 1.2 Roads. 2.5.2 Passageways A passageway used exclusively for passengers ideally shall extend from the passenger waiting room to the boarding facilities. 2.5.3 Car Parks Car parks at ferry terminals shall have sufficient areas for avoiding any deterioration of traffic conditions in their vicinity in consideration of the number of boarding vehicles per ferryboat, utilization rate, and concentration rate of arriving vehicles. [Commentary] Many factors should be considered in calculating the area of car parks, including the following: (1) Number of berths (2) Number of vehicles loaded on a ferryboat (separately for passenger cars and trucks) (3) Frequency of arrival and departure of ferryboats and embarking / disembarking time (4) Arrival pattern of vehicles (separately for passenger cars and trucks) (5) Operation method of the car park [Technical Notes] The area of car parks in a ferry terminal may be calculated by multiplying the unit parking area necessary for an eight-ton truck (50 m2) with the maximum number of vehicles loaded on a ferryboat using the terminal, in an equivalent unit of eight-ton truck. The shares of unaccompanied vehicles and trailers should also be considered. The number of vehicles loaded on an ordinary ferryboat is listed in Table T- 2.5.1. One passenger vehicle is equivalent to 0.5 eight-ton truck and one large bus corresponds to 1.5 eight-ton truck. Table T- 2.5.1 Standard Number of Loaded Vehicles Target ferry (gross tonnage) 100 200 300 400 500 600

Maximum number of loaded 8-ton trucks 11 14 16 19 22 24

Target ferry (gross tonnage)


1,000 2,000 3,000 4,000 5,000 6,000

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Target ferry (gross tonnage) 7,000 8,000 9,000 10,000 11,000 12,000



2.5.4 Passenger Terminals Passenger terminals shall be so designed to provide safe and efficient services for passengers. [Commentary] (1) Passenger terminals should be designed according to Part Ⅸ , 5.2 Passenger Building. (2) Passenger terminals constitute a specific building defined in Articles of the “Law for Promoting Construction of Specific Buildings Available for Smooth Utilization by the Elderly and the Physically Handicapped”. Therefore, due consideration should be given to related laws and regulations. (3) Any development of passenger terminal facilities should be conducted with reference to the “Guideline on Development of Facilities for the Elderly and the Physically Handicapped in Public Transportation Terminals”.

2.5.5 Safety Equipment A ferry terminal shall be provided with equipment such as fences, crossing gate, fire-extinguishing equipment, safety indications, safety lamps, and/or lifesaving equipment, as necessary.

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Part XI Marinas

PART XI MARINAS

Part XI Marinas Chapter 1 Introduction The location of a marina shall be selected in light of the size and capacity as well as the natural conditions, the social conditions, and the economical advantage. Harbor facilities in a marina shall be arranged in consideration of the behaviors of boats and users, and mutual relations among facilities based on the designed size of each facility, so as to ensure the safety, convenience and efficiency of the marina as a whole. Due consideration shall also be given to preservation of natural environment and landscape. [Commentary] (1) Marinas refer to harbors consisting of navigation channels, basins for mooring and storing pleasure boats, protective facilities, and service facilities such as club houses, car parks, boat yards, training facilities, and greenery areas. (2) The location of a marina should be determined through the assessment of the relevance of the site in terms of geographical location, activities of pleasure boats, and feasibility of marina facilities construction in light of natural conditions, social conditions, and economic advantage. (3) Thorough investigations on the behaviors of boats and users and the safety of pleasure boat users are necessary for the process of designing the configuration of facilities. Prospects for future development should also be considered adequately. (4) Facility plans for marinas should be made in consideration of the preservation of marine and land environment. (5) Marina facilities should be harmonized with the surrounding landscape. (6) It is preferable that marine facilities are to be arranged taking into account the utilization by disadvantaged people including the elderly and the physically handicapped persons. [Technical Notes] (1) Harbor facilities in marinas are listed in Table T- 1.1.1 Table T- 1.1.1 Harbor Facilities in Marinas (1) Navigation channels and mooring basins (2) Protective facilities (breakwaters, seawalls) (3) Mooring facilities (quays, open-type wharves, floating open-type wharves, mooring dolphins, mooring buoys, boat lifting/ lowering facilities (slipways, rail ramps, boat lifters)) (4) Facilities for ship services (water supply facilities, power supply facilities, repairing facilities, land-based storage facilities (boat yards, boat houses, racks), club houses) (5) Land traffic facilities (roads, car parks)

(2) In addition to the facilities shown in Table T-1.1.1, marina facilities include administration and operation facilities (information facilities, communication facilities, lifesaving facilities, control facilities, etc.).

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Chapter 2 Main Dimensions of Target Boats The principal dimensions of the target boats shall be determined appropriately in light of the present condition and future trend of pleasure boats in the area around the marina. [Commentary] Sufficient investigations on the maneuverability of the pleasure boats and their motion characteristics in the winds, waves, and tidal currents should be made so as to ensure safe and smooth entry, departure, anchorage and mooring of all the pleasure boats not only under normal weather conditions but also under emergency evacuation conditions due to stormy weather. [Technical Notes] The principal dimensions of the target boats may be determined referring to Table T- 2.1.1, if it is not possible to identify particular target boats. Table T- 2.1.1 Standard Size of Pleasure Boats Type

Length overall (m)

Extreme breadth (m)

Draft (m)

Mass (kg)

Cruiser

7.0 7.5 8.0 8.5 9.0 9.5 10.0 10.5 11.0 11.5 12.0 12.5 13.0 13.5 14.0 15.0 16.0

2.8 2.9 3.0 3.1 3.2 3.4 3.5 3.6 3.7 3.8 3.9 4.1 4.2 4.4 4.5 4.8 5.1

1.5 1.6 1.6 1.7 1.8 1.8 1.9 1.9 2.0 2.0 2.1 2.2 2.3 2.3 2.4 2.6 2.8

2,300 2,600 2,900 3,200 3,600 4,100 6,700 7,200 7,800 8,400 9,100 9,800 10,700 11,500 12,500 14,800 17,500

Dinghy

3.5 4.0 4.5 5.0 5.5 6.0 6.5 7.0 7.5

1.6 1.7 1.8 1.9 2.0 2.1 2.2 2.3 2.5

0.9 1.0 1.1 1.2 1.3 1.3 1.4 1.6 1.7

60 80 110 150 250 330 440 600 820

Motorboat

6.0 7.0 8.0 9.0 10.0 11.0 12.0 13.0 14.0 15.0 16.0 17.0 18.0

2.6 2.8 3.0 3.6 3.8 4.0 4.1 4.3 4.7 4.9 4.9 5.1 5.4

0.6 0.7 0.7 1.1 1.1 1.1 1.1 1.1 1.1 1.2 1.2 1.2 1.2

1,800 2,100 2,800 7,600 8,700 10,000 11,600 13,400 15,600 18,300 21,500 25,600 29,800

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PART XI MARINAS

Chapter 3 Navigation Channels and Basins 3.1 General Design of navigation channels shall conform to Part Ⅵ Navigation Channels and Basins as well as to the provisions hereinafter.

3.2 Navigation Channels Navigation channels shall be arranged in the calm water with a sufficient width and depth in light of the direction of prevailing wind, to ensure safe navigation of pleasure boats. [Commentary] (1) The width of navigation channels should be determined so as to ensure safe and smooth navigation of all the pleasure boats using the marina. (2) The depth of navigation channels should be determined appropriately in light of (a) the draft of the pleasure boats, (b) the increase of draft due to the loads for cruising, (c) the boat’s trim, (d) the motion amplitudes of pleasure boats, and (e) a margin for the keel clearance. [Technical Notes] (1) In some marinas, the width of navigation channel measures more than twice the length of pleasure boats with engines and more than five times the length of pleasure boats without engines. In particular, considerations should be given for simultaneous navigation of numerous pleasure boats if a yacht race is to be held. (2) In some cases, as for the depth of navigation channel, a margin of 0.6 m to 1.0 m has been added to the draft of the pleasure boats when the magnitudes of other factors were not identified. (3) The influence of prevailing wind direction may be disregarded if sail boats are not allowed. (4) The centerline of a curving section of navigation channels should be determined appropriately in light of the turning capacity of the pleasure boats.

3.3 Mooring Basins Mooring basins shall be located in the calm water with the depth and area sufficient to ensure safe mooring and smooth maneuvering of pleasure boats. [Commentary] (1) The depth of mooring basins should be the same as the depth provided in 3.2 Navigation Channels, unless a mooring area is specified for each single type of the pleasure boats. (2) The calmness of mooring basins should be defined in compliance with Part Ⅵ , 4.4 Calmness of Basin and in light of weather and usage conditions. In this regard, sufficient consideration should be given to waves penetrating through a harbor entrance, transmitted waves over breakwaters, reflected waves, ship waves, long period waves, and seiche. [Technical Notes] (1) In most cases, allowable wave height for the use of mooring basins (H1/3) is set at 0.3 m or less, and the wave height in the basins under stormy weather (H1/3) is planned to remain at 0.5 m or less. Nonetheless, it is preferable to envisage a lower wave height under stormy weather according to the types and sizes of boats and the condition of use, unless land-based storage is possible for all those boats. (2) In ensuring calmness, the layout, length and crown height of breakwaters should be planned appropriately. It is also preferable to install wave-absorbing seawalls, permeable breakwaters, and slipways as well as to utilize the wave-absorbing functions of beaches effectively. [Reference] 1) “Planning and Design Guidelines for Small Craft Harbors”, American Society of Civil Engineers, 1994, 68p.

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Chapter 4 Protective Facilities The harbor entrance and the crown height and structural types of protective facilities shall conform to Part Ⅶ Protective Facilities of Harbors and the following provisions shall also be taken into consideration: (1) Harbor Entrance The direction, location, water depth, and width of harbor entrance shall be arranged adequately in consideration of the prevailing wind, wind waves, tidal currents, ships navigating nearby areas, and littoral transport so as to ensure safe entry and departure of pleasure boats. (2) Crown Height The crown height of breakwaters and seawalls shall be designed appropriately in view of ensuring required calmness. (3) Structural Type Consideration shall be given to the structural types of breakwaters and seawalls to ensure required calmness. [Technical Notes] (1) Layout of protective facilities should be determined to facilitate safe return entry and evacuation of many pleasure boats in case of sudden change in weather, and to provide a marina with a sufficient basin area. (2) The direction of harbor entrance should be arranged so as to prevent direct penetration of wind waves and tidal currents from outside and to ensure that the entrance will not be blocked by littoral sand drift. Care should be taken to protect pleasure boats from transverse waves around the entrance so as to ensure their safe and easy entry and departure. Considering that dinghy yachts cannot move straight against the wind, as shown in Fig. T- 4.1.1, the direction of harbor entrance should be kept at the angle between 45°and 90° to the prevailing wind, as in Fig. T4.1.2.

Close hold

Wind direction

Close hold

Close reach

Close reach

Board tack

Board tack Prevailing wind

Wind abeam

Direction of harbor entrance

Wind abeam

Broad reach

Broad reach

Quarter lee

Quarter lee

Fig. T- 4.1.1 Precluded Directions

Fig. T- 4.1.2 Direction of Harbor Entrance and Prevailing Wind

(3) The location of harbor entrance should be arranged in consideration of the safety of ships navigating nearby. It should also be ensured that pleasure boats will be free from the influence of wave concentration, wave breaking, and tidal currents. Sufficient water depth and width should be secured at the entrance in light of the dimensions for navigation channels. (4) The crown height of breakwaters should ensure required calmness even against storm waves. Nonetheless, consideration should also be given to the elevation of sighting on pleasure boats to ensure a clear vision for safe navigation. Care should further be taken not to cut off the winds necessary for dinghy yachts sailing in the harbor. (5) As necessary, the seawalls of sloping, wave-absorbing, permeable types should be considered for adoption to ensure calmness and cleanliness in the harbor. (6) It is preferable to pay attention to the aesthetic view of breakwaters and seawalls.

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PART XI MARINAS

Chapter 5 Mooring Facilities 5.1 General The type and size of mooring facilities shall be determined appropriately in accordance with Part Ⅷ Mooring Facilities and in light of the purpose of mooring, boat size, and tidal range. [Technical Notes] Specifications of mooring facilities may be determined with reference to Fig. T- 5.1.2. The distance between two adjacent piers should be determined appropriately according to the number of pleasure boats to be moored between the two piers so as to prevent any contact of pleasure boats with auxiliary piers or any contact between pleasure boats themselves. Avoidance of wind-induced touching of the masts of cruiser yachts should be considered in the arrangement of piers. B2

B1

B1

S

B1

b'

b

b

W1 W2

B2

B2

S

W3

Width of piers b ¢: b

Length overall of pleasure boat

Length of berth

Slip

~

~

Distance between piers (minimum) W

= (ship breadth) + b + (0.3 ~ 0.6 m)

W

= 2× (ship breadth) + b + (1.5 ~ 2.0 m)

W

= (ship breadth) + (1.0 ~ 2.0 m)

1 2

~

main piers 1.5 ~ 3 m

: auxiliary piers 1.0 ~ 1.5 m

3

Note: The distance between the piers should decided in consideration of wind-induced touching of masts for cruiser yachts.

Fig. T- 5.1.2 Examples of Mooring Facilities Specifications

5.2 Design Conditions for Mooring Facilities Design conditions for mooring facilities shall be defined appropriately in accordance with to Part Ⅱ Design Conditions and in light of their use as marinas. [Technical Notes] (1) Safety of floating piers should be examined in consideration of the following loads as necessary: (a) Support load of any connection bridge on the pier (b) Snow load in heavy snow areas (c) Footway live load (see Figs. T-5.2.1 ~ T.5.2.3 for the methods of loading on floating piers.) Auxiliary pier

Main pier

Auxiliary pier

Main pier

One-side loading

Fig. T- 5.2.1 Example of Loading in Examining Stability of Main Floating Pier

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One-side loading

Loading position

Fig. T- 5.2.2 Example of Loading in Examining Stability of Auxiliary Floating Pier

Fig. T- 5.2.3 Example of Loading in Examining Sinkage of Auxiliary Floating Pier

(2) Loads in examining the safety of floating piers against the sectional failure may be defined refering to Part Ⅱ, Chapter 8 External Forces Acting on Floating Body and Its Motions. Normally, waves, winds, tidal currents, and wave drift forces are to be considered as acting on floating bodies. In addition to those forces, wave making resistance is also to be considered as an external force on the motions of floating bodies. (a) Methods of simple calculation of wave forces acting on floating piers themselves and moored pleasure boats include the following: ① Assuming that the hydrostatic pressure acts on a floating body, the wave force can be calculated according to equation (5.2.1), by taking the difference between the water pressures in the front and rear walls as shown in Fig. T- 5.2.4. P = r0 gHLpd where P r0 g H Lp d

(5.2.1)

：wave force acting on floating body (kN) ：density of seawater (t/m3) (1.03 t/m3) ：gravitational acceleration (m/s2) (9.8 m/s2) ：wave height (Hmax) (m) ：length of floating body (m) ：draft of floating body (m)

Floating body

Wave Mean water level

Water pressure in rear

Water pressure in rear Difference of water pressure Water pressure in front

Fig. T- 5.2.4 Wave Force Calculation Assuming Hydrostatic Pressure Acting on Floating Body

② When a large inertia force of the floating body or any action of breaking wave pressure is anticipated, the wave force may be calculated according to equation (5.2.2), by considering the wave pressure calculated by the Goda formula as lateral load, as shown in Fig. T- 5.2.5. As Fig. T- 5.2.5 applies to the action of wave crest, consideration should also be given to the action of wave trough, as shown in Fig. T- 5.2.6. A triangular distribution should be applied to the uplift, by taking the pressure p3 at the front end and zero at the rear end of the floating body. When the breadth of the floating body B exceeds L/4 (L: wave length), uplift should be distributed in a L/4-wide triangular form.

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PART XI MARINAS

2

1 ì 4p ¤ L ü a 1 = --- í ----------------------------- ý 2 î sinh ( 4h ¤ L ) þ

6444474448

h* = 0.75 (1 + cosb)l1H p1 = 0.5 (1 + cosb)a1l1r0 gH p3 =a3 p1

(5.2.2)

ü dì 1 a 3 = 1 --- í --------------------------------------- ý h î 1 cosh ( 2h ¤ L ) þ where H： wave height (Hmax) (m) h： water depth (m) r0： density of seawater (t/m3) (1.03 t/m3) g： gravitational acceleration (m/s2) (9.81 m/s2) b： angle between the line normal to the longitudinal axis of floating body and the principal wave direction (with a deflection toward the former by up to 15° ) l1： modification coefficient of wave pressure (normally 1.0) Wave pressure according to Goda formula




Fig. T- 5.2.6 Pressure at Wave Trough

Fig. T- 5.2.5 Pressure at Wave Crest

(3) In calculating the wind force, the area of exposure should be defined appropriately in light not only of mooring facilities but also of the mooring condition of pleasure boats and the layout of mooring facilities and surrounding structures. (4) External forces caused by the motions of moored pleasure boats and linked floating piers act on a floating body, as well as the forces by waves, winds, and currents exerted on the floating body itself. Such external forces should be calculated according to appropriate analytical methods or hydraulic model experiments. (5) It is important to consider the tractive force of moored pleasure boats due to waves, winds, and currents under stormy weather as well as the impact of collision at berthing of pleasure boats to the pier.

5.3 Floating Piers 5.3.1 General Floating piers as mooring facilities in marinas shall conform to Part Ⅷ , Chapter 12 Floating Piers. 5.3.2 Structure Main structures of floating piers and connecting mechanisms between main structures shall be durable and able to resist the external forces calculated in 5.2 Design Conditions for Mooring Facilities. 5.3.3 Examination of Safety The safety of floating piers shall be examined in accordance with Part Ⅷ , 12.3.3 Stability of Pontoon and in light of mutual impact of floating bodies constituting the pier. When the surcharge calculated in 5.2 Design Conditions for Mooring Facilities are exerted, floating piers shall satisfy the stability condition of a floating body. The slope of the deck, the sinking of the floating bodies, and the freeboard shall create no obstacles to their use. In this case, the freeboard of floating piers shall be determined in light of the dimensions of the pleasure boats to be served and the wave conditions to ensure smooth boarding and alighting of pleasure boat users.

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[Commentary] The slope and sinking of the deck and the freeboard should create no obstacles to the use of the floating pier when it receives a uniform or concentrated load or an eccentric load corresponding to the surcharge calculated in light of the dimensions of the floating pier, the condition of use, and the natural conditions. [Technical Notes] The examination of the stability of a floating pier should satisfy the following conditions: (1) When the loads (a) ~ (c) in [Technical Notes] of 5.2 Design Conditions for Mooring Facilities (here referred to as the “Notes”) act on the deck, the stability condition of a floating body should be satisfied and the necessary freeboard should be maintained. The freeboard of a floating pier is usually set around 30 ~ 50 cm to ensure that any user falling into the water can climb back up easily. (2) When the footway live load under the condition of the load (c) of the Notes sketched in Fig. T-5.2.1 acts on one side of the main pier along with the loads (a) and (b) as necessary, the slope of the deck should be 1:10 or less and the smallest freeboard should be zero or greater. (3) When the footway live load under the condition of the load (c) of the Notes sketched in Fig. T-5.2.2 acts on one side of the auxiliary pier along with the loads (a) and (b) as necessary, the slope of the deck should be 1:10 or less and the smallest freeboard should be zero or greater. (4) When the footway live load under the condition of the load (c) of the Notes sketched in Fig. T-5.2.3 acts on the auxiliary pier up to 1 m from the top along with the load (b) as necessary, the sinking of the deck should not be excessive.

5.3.4 Structural Design Based on 5.2 Design Conditions for Mooring Facilities, the structural members of a floating pier shall be safe against the external forces acting on the pier under stormy weather, and during mooring and berthing of pleasure boats. [Technical Notes] The structural members of a floating pier should be examined on their safety against the bending and shear around the longitudinal and lateral axes. The stresses resulting from the motions of floating bodies or pleasure boats should also be considered as necessary. (1) As shown in Figs. T- 5.3.1 and 5.3.2, the bending moment or shearing force around the longitudinal axis should be examined appropriately according to the structural type of the floating body, based on the concept of simple beam or the Muller method. (2) Bending moment or shearing force around the lateral axis should be examined appropriately based on the concept of simple beam. Dead load

Dead load

Buoyancy

Buoyancy

Fig. T- 5.3.1 Simple Beam (Separate Type)

Fig. T- 5.3.2 Muller Method (Monocoque Type)

5.3.5 Mooring Method The mooring method of a floating pier shall be selected through the consideration of fulfilling the function of fixing the pier on the water surface, ensuring the resistance against external forces including waves, winds, and currents, and retaining the durability. 5.3.6 Access Bridges Dimensions of an access bridge shall be determined appropriately so as not to undermine the safety of users.

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PART XI MARINAS

[Technical Notes] (1) According to structural types, some bridges may move up and down in response to the change in water level, while others may further move left and right in response to the motions of the pier. (2) The breadth of an access bridge should be 75 cm or greater. In particular, an appropriate breadth should be secured if consideration is required for utilization by disadvantaged people. (3) It is preferable that the slope of an access bridge is not steeper than 1:4. (4) It is preferable that access bridges are equipped with antiskid devices.

5.4 Ancillary Facilities Ancillary facilities shall be installed as necessary for the purpose for prevention of danger, environmental control, and efficient operation of the marina.

5.5 Lifting / Lowering Frame Facilities The structural type and size of boat lifting / lowering frame facilities shall be determined appropriately in light of the type, size and number of pleasure boats to be served as well as the handling capacity of the facilities. [References] 1) Sigeru UEDA, Satoru SHIRAISHI, Takashi ISHISAKI: “Calculation method of forces and moments induced on pontoon type floating structures in waves”, Rept. of PHRI, Vol. 31, No. 2, 1992 (in Japanese). 2) Satoru SHIRAISHI, Haruo YONEYAMA, Kazuyuki YOKOI: “Calculation method of forces and moments induced on combshaped floating structures in waves”, Tech. Note of PHRI, No. 828, 1996 (in Japanese). 3) Sigeru UEDA, Satoru SHIRAISHI, Kazuo KAI: “Calculation method of shear and bending moment induced on pontoon type floating structures in random sea”, Tech. Note of PHRI, No. 505, 1984 (in Japanese). 4) Sigeru UEDA, Satoru SHIRAISHI: “Determination of optimum mooring chain and design charts using catenary theory”, Tech. Note of PHRI, No. 379, 1981 (in Japanese).

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Chapter 6 Facilities for Ship Services 6.1 General The type and size of ship service facilities shall be determined appropriately in accordance with Part Ⅸ , Chapter 4 Facilities for Ship Services to ensure smooth ship services in the marina. [Commentary] Ship service facilities include water supplying facilities, oil bunkering facilities, power supplying facilities, boat washing facilities, lighting equipment, boat repair facilities, wastewater and refuse treatment facilities, club house, and land-based storage facilities.

6.2 Land Storage Facilities The type and size of land storage facilities shall be determined appropriately in light of the type, size, and number of the pleasure boats to be served. [Commentary] Land storage facilities include boat yards, boat houses, boat racks, and multi-storied storage facilities. Their type and size should be determined appropriately in light of the type of the pleasure boats. Dimensions of land storage facilities should be determined in light of the operation area required by the equipment for boat shifting, in addition to the size of the boats to be served. [Technical Notes] (1) Dimensions of land storage facilities can be determined with reference to Fig. T- 6.2.1. (2) In the area frequently affected by stormy weather including typhoons or the areas with a long off-season, the manner of storage should be determined with due consideration of local features.

Mark length

Mark width

L: ship

Road width When tractors are not used:

S=A

length

B: ship

When traveling vehicles

breadth

(forklift, trailer) are used: The value of S is defined in light of the traveling vehicles radius of gyration.

Fig. T- 6.2.1 Dimensions of Land Storage Facilities

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PART XI MARINAS

Chapter 7 Land Traffic Facilities Land traffic facilities shall be determined in accordance with Part Ⅸ , Chapter 1 Port Traffic Facilities and in light of the utilization of the marina and the concentration of vehicles. [Technical Notes] The structure and size of roads and car parks related to marinas may be determined with reference to Table T- 7.1.1. Table T- 7.1.1 Considerations in Designing Land Traffic Facilities Facilities

Concept of development

Roads

Examination is required concerning the breadth and radius of gyration of vehicles towing trailers loaded with pleasure boats.

Car parks

The area of car parks required is 1.5 times the area for one vehicle per boat. A vehicle attached with a trailer needs about twice the area for ordinary vehicles. It is preferable that car parks can be used as landing sites for pleasure boats in emergencies.

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-600-

INDEX

INDEX

INDEX

apparent seismic coefficient 187, 202, 204, 205, 335 approaching energy 464 approaching speeds 463 approximate lowest water level 4 apron 89, 184, 207, 208, 226, 229, 413, 462, 501, 532, 540, 541, 542, 545, 546, 547, 548, 552, 577, 583 apron works 398 arcs 449, 450, 456, 458, 459 area of improvement 329, 331 armor layer 84, 112, 113, 117, 118, 119, 369 armor stones 100, 112, 113, 114, 116, 117, 118, 119, 125, 140, 369 Arthur 149, 153 artificial dune 394 asphalt concrete for pavement 228 asphalt mats 228, 272, 372 asphalt pavement 547, 548, 549, 550, 553, 562, 563, 572, 573 asphalt paving guidelines 229 asphalt stabilization 228 assignment of traffic volume to various routes 560 astronomical tides 127 atmospheric pressure 128, 129, 130, 133, 138 availability of construction materials 359 average color rendering performance evaluation number 533, 534 average degree of consolidation 171, 172, 310, 340 axial bearing capacity of piles 284, 288, 290, 302, 468 axial compressive stress 212, 213, 214 axial spring constant of pile head 304 axial tensile stress 212, 213, 214

Symbols B = 0 method 173

A abnormal water levels 127, 365 abrasion resistance 26 absorbing caissons 267, 378, 380 access bridge 123, 124, 461, 463, 469, 478, 480, 487, 488, 489, 492, 493, 596, 597 accidental load 221, 222 active earth pressure 200, 201, 202, 204, 262, 266, 280, 325, 327, 329, 332, 333, 334, 415, 416, 418, 419, 427, 428, 432, 433, 437, 439, 441, 448, 453, 454, 492, 510, 518, 555, 556 active load 221 actual mooring lines 386 additional lane 562 adjusting tower 493 administration and operation facilities 589 AE (air entrained) concrete 225 afforestation works 394 aforementioned currents 138 air-cushion vehicle 407, 500, 501 air-cushion vehicle landing facilities 407, 500, 501 alarm systems 520, 525 alignment of breakwater 110, 370 alkali-aggregate reaction 224, 225 allowable axial bearing capacity of piles 284, 288, 290, 468 allowable axial compressive stress 213 allowable bearing capacity 274, 275, 276, 277, 280, 285, 288, 289, 294, 295, 307, 448 allowable bending compressive stress 213, 224, 305 allowable bond stress 224 allowable displacement ductility factor 471, 474, 475 allowable displacement for the sheet pile crown 511 allowable limit for expected sliding distance 367 allowable overtopping 373, 396, 398 allowable pulling resistance of piles 291, 292, 293 allowable shear resistance force 448 allowable shearing stress 224, 448 allowable stress method 211, 221, 224, 424, 460 allowable stresses 212, 213, 214, 215, 223, 224, 233, 305, 468, 483 allowable tensile stress 213, 224, 291, 293, 324, 424, 425 allowable upward displacement 293 aluminum 207, 218, 239, 530 aluminum alloy anodes 218 aluminum, titanium 234 amenity aspects 396 amenity-oriented seawall 396 amplitude amplification factor 134 anchor chain type mooring buoy 502, 506 anchorage work 415, 417, 419, 426, 427, 429 ancillary facilities 234, 370, 389, 398, 406, 486, 520, 557, 575, 578, 579, 583, 586, 587, 597 ancillary works 414 angle of friction between backfilling material and backface wall 201, 202 angle of friction of the wall 257 angle of internal friction 174, 175, 176, 200, 201, 202, 231, 232, 238, 239, 240, 315, 318, 329, 334, 335, 413, 420, 421 angle of repose 314, 413 apparent cohesion 173, 277, 278, 279, 314, 315, 316

B backfill 200, 201, 202, 205, 217, 231, 234, 235, 262, 325, 326, 328, 329, 333, 336, 343, 397, 408, 409, 413, 416, 429, 437, 438, 540, 546 backfilling 201, 202, 231, 232, 234, 235, 238, 257, 325, 326, 397, 409, 413, 438, 540 backfilling materials 232 backfilling stones 257 backshore 154 bar type beach 155 barricades 531 base course material 232, 238, 542 basic cross section 378 basins 1, 32, 133, 150, 151, 345, 346, 348, 351, 352, 353, 355, 357, 358, 381, 390, 393, 401, 520, 574, 584, 589, 591 bathymetry measurement 150 bay characteristics 129 bay entrance correction factor 133, 134 beach deformation 154, 156, 163, 165, 166 beach erosion 126, 154, 165, 357, 391 bead 192, 234, 326, 327 beam method 418, 422 bearing capacity 167, 168, 174, 177, 232, 253, 254, 273, 274, 275, 276, 277, 278, 279, 280, 281, 282, 283, 284, 285, 286, 287, 288, 289, 290, 291, 292, 293, 294, 295, 301, 302, 303, 304, 305, 306, 307, 309, 312, 313, 314, 315, 316, 326, 336, 366, 367, 369, 384, 409, 410, 411, 426, 431, 434, 448, 452, 455, 458, 460, 465, 468, 470, 471, 475, 477, 480, 483, 484, 495, 496, 513, 516, 518, 541, 542, 544, 546, 547, 548, 551, 552, 555, 563 bearing capacity coefficient 542, 543 bearing capacity factors 274, 275 bearing piles 284, 285, 290, 304, 436, 482 bearing stress 212, 214, 224, 479, 555 bedload 148, 155, 156, 166 bedrock acceleration 188, 189, 191, 472 -I-1-


bending compressive stress 212, 213, 214, 224, 305 bending failure 304, 477, 478 bending tensile stress 212, 213, 214, 224, 305 berm 100, 105, 106, 107, 113, 117, 118, 363, 372 berm width of the mound 100, 108 berth configuration factor 16, 17 berthing energy 16, 17, 22, 464, 495, 523 berthing force 16, 415, 417, 428, 464, 485, 494, 495, 522, 523 berthing velocity 16, 17, 18, 19, 27, 524 berths 17, 19, 21, 26, 351, 353, 402, 403, 405, 532, 582, 585, 587 bilge keels 405 bitts 494, 521, 539 bituminous materials 228, 229 blast furnace 238, 239 blast furnace slag 238, 239, 328, 329, 333, 334, 343 block coefficient 20, 21, 22, 23 block failure 289 block type improvement 323 Bloom and Rowe 419 blown asphalt 228, 229 boat houses 589, 598 boat racks 598 boat yards 589, 598 bollard 22, 25, 26, 27, 417, 424, 425, 428, 463, 481, 520, 521, 522, 539, 582, 585 bore 132, 167, 177, 180 bottom reaction 251, 252, 253, 257, 260, 262, 266, 268, 269, 408, 411, 496 bottom slab 222, 242, 243, 244, 247, 248, 249, 250, 251, 252, 253, 254, 255, 256, 257, 258, 259, 260, 261, 266, 269, 412, 451, 490 bottom slope 75, 76, 77, 78, 79, 80, 81, 82, 84, 85, 86, 87, 88, 91, 92, 100, 103, 104, 106, 113, 116, 132, 148, 154, 160 Boussinesq’s equation 69, 93, 99, 340 Boussinesq’s solution 310 bow and stern side 403, 585 box shear test 175, 180 breaker 159 breaker index 78 breaker line 81, 390, 392, 393 breaker zone 36, 37, 75, 76, 77, 78, 79, 91, 109, 159, 162, 367, 392 breaking 32, 33, 36, 37, 41, 43, 49, 69, 78, 79, 80, 91, 93, 94, 97, 100, 101, 102, 103, 104, 105, 106, 107, 108, 109, 110, 111, 113, 120, 121, 124, 125, 149, 153, 159, 230, 231, 319, 377, 451, 504, 594 breaking point 36, 75, 82, 83, 91, 107, 159 breaking strength 26, 424, 425, 506 breaking wave force 100, 101, 104, 107, 108, 109, 121, 364, 381 breaking wave height 78, 79, 80, 91, 110 breakwater 52, 53, 54, 55, 56, 57, 58, 59, 60, 61, 62, 63, 64, 65, 66, 67, 68, 69, 70, 71, 72, 74, 90, 91, 94, 99, 102, 103, 104, 105, 106, 107, 108, 109, 110, 112, 113, 114, 115, 116, 117, 118, 119, 124, 125, 221, 222, 223, 231, 244, 250, 251, 253, 261, 263, 266, 267, 270, 272, 277, 278, 322, 339, 348, 349, 350, 354, 357, 358, 359, 360, 361, 362, 363, 364, 365, 369, 370, 372, 373, 374, 375, 376, 377, 378, 379, 380, 381, 382, 383, 384, 385, 386, 387, 388, 390, 391, 393, 508, 518, 574, 589, 591, 592 breakwater alignment 106, 110, 116, 119 breakwater caisson covered with a mound of wave-dissipating concrete blocks 250 breakwater head 116, 118, 119, 162, 365, 369, 370, 371 breakwater trunk 113, 117, 118, 119, 369, 370, 371 breakwaters with wide footing 377 Brebner-Donnelly formula 118 Bretschneider method 46 Bretschneider-Mitsuyasu spectrum 38, 39

buckling 213, 308, 309, 458, 460, 475, 477, 483 buoyancy 145, 187, 243, 249, 251, 252, 253, 274, 275, 277, 289, 292, 335, 364, 365, 408, 410, 411, 454, 489, 502, 507, 568, 569, 570 buoys 142, 234, 351, 502, 503

C caisson breakwater fully covered with wave-dissipating blocks 378 caisson breakwaters 124, 125, 222, 269, 360, 369, 370, 377 caisson type 193, 194, 222, 227, 322, 360, 361, 364, 370, 372, 494, 508, 509 caisson type composite breakwater 360, 361 caisson type dolphins 496 caisson type quaywalls 221, 227, 408 caisson type upright breakwater 360 calculation of deformation moment 439, 453, 517 calculation of time-settlement relationship 310 calmness 32, 94, 348, 349, 351, 353, 357, 358, 359, 362, 376, 381, 384, 385, 501, 508, 591, 592 cantilever sheet pile wall 509, 510, 511 canvas sheets 372 car parks 564, 584, 587, 589, 599 cargo handling 28, 32, 70, 93, 94, 133, 187, 207, 208, 209, 223, 227, 345, 351, 353, 354, 359, 401, 402, 410, 411, 416, 461, 470, 485, 486, 494, 495, 496, 526, 527, 528, 531, 532, 540, 541, 554, 555, 556, 557, 564, 571, 573, 574, 575, 581, 583 cargo handling equipment 207, 209, 401, 410, 411, 416, 461, 463, 470, 485, 486, 494, 495, 496, 554, 555, 556, 557, 571, 573, 574, 575, 581, 583 cargo handling equipment load 207, 209 cargo ship 2, 9, 10, 12, 17, 18, 19, 21, 354, 402, 403, 404, 582, 583 cargo sorting area 573, 575, 576 cast steel 212, 214 cast-in-place concrete 305, 408, 499 cast-in-place concrete piles with outer casing 305 catenary line 490, 491 catenary mooring 142, 143 catenary theory 491, 493, 597 cathodic protection method 217, 218, 219, 220, 226, 272 CBR test 548 CD condition 173, 174 ceiling slab 111, 112, 125, 268, 269, 508 celerity 95, 149, 159 cellular 459 cellular blocks 210, 261, 262, 263, 264, 265, 266, 412 cellular concrete block type composite breakwater 360, 361 cellular concrete block type quaywalls 408 cellular concrete block type upright breakwater 360 cellular concrete blocks 210, 372, 412 cellular-bulkhead 194, 329, 407, 436, 437, 438, 439, 440, 443, 444, 449, 451, 452, 453, 454, 455, 456, 457, 458, 459, 486, 494, 495, 496, 517, 518 cellular-bulkhead type quaywalls 407 cement concrete pavement 542, 553, 562, 563, 572 cement-based hardeners 327 cement-mixed soils 333 center of buoyancy 243 center of gravity 20, 21, 243, 315, 366, 463, 466, 489 Chang’s method 295, 300, 465, 466, 516, 518 characteristic embedded length 420 characteristic values 154, 222, 225, 226, 263 chart datum level (CDL) 2, 4, 128, 131 CIQ facilities 579 circular arc analysis 194, 276, 277, 315, 316, 319 circular arc analysis of the modified Fellenius method 276 -I-2-

INDEX

circular hollow 524 circular slip 277, 315, 316, 318, 329, 331, 333, 340, 363, 369, 370, 397, 410, 411, 425, 431, 435, 449, 455, 458, 464, 477, 484, 512, 518 circular slip failure 315, 316, 318, 331, 333, 363, 397 clayey ground 174, 180, 275, 276, 277, 280, 281, 286, 289, 290, 293, 296, 310, 312, 314, 377 clearance limit 528, 560 clinker ash 239, 240 club houses 589 cnoidal wave 36 coal ash 8, 238, 239, 240 coal storage yards 576 coarse sand 169, 199, 328 coastal erosion control 394 coating materials 217, 234, 236 coating method 217, 220, 272 coefficient of consolidation 168, 172, 311, 313, 320, 321, 322, 342 coefficient of earth pressure 202, 203, 250, 258, 263, 293, 419, 442, 450 coefficient of friction 280, 293, 327, 410, 411, 412, 442 coefficient of horizontal subgrade reaction 281, 465, 466, 472 coefficient of lateral subgrade reaction 300, 304 coefficient of linear thermal elongation 212 coefficient of permeability 113, 168, 169, 170, 173, 175, 195, 328, 397 coefficient of secondary compression 312 coefficient of stress distribution 319 coefficient of volume compressibility 168, 170, 171, 311, 319, 341 cohesion of soil 201, 317, 331, 334, 418 cohesive materials 334 cohesive resistance 280 collision force 268, 365, 373, 386 collision load 221 compensation currents 138 composite (hybrid) type 568 composite breakwater 90, 91, 99, 104, 105, 108, 112, 113, 114, 117, 118, 124, 125, 193, 360, 363, 364, 365, 369, 370, 371, 375, 377, 378 composite resin mats 372 composite seismic angle 202, 204 composite slabs 271 compressibility properties 310 compression frequency-dependent characteristics 524 compression index 171, 312 compression strength 181, 222 concave corners 41, 72, 110, 134, 359, 370 concentrated corrosion 216, 217, 220 concrete (reinforced concrete, prestressed concrete) type 568 concrete aggregate 238 concrete block pavement 551, 552, 553 concrete block type composite breakwater 360, 361 concrete block type sloping breakwater 360 concrete block type upright breakwater 360, 361 concrete blocks 72, 83, 84, 89, 91, 100, 103, 106, 109, 112, 113, 114, 116, 117, 119, 125, 140, 250, 253, 360, 362, 363, 364, 365, 367, 368, 372, 377, 378, 381, 391, 406, 414, 508 concrete crown 252, 362, 363, 378, 379 concrete lid 226, 248, 250, 251, 252, 253, 262, 272, 363, 370 concrete pavement 541, 542, 543, 545, 546, 553, 562, 563, 572, 573 concrete wall anchorage 426, 427, 428 cone Parameters 178 confining pressure 174, 278, 327, 337 consolidated and drained triaxial compression tests 330

consolidation characteristics 167, 168, 180 consolidation properties 170, 171, 177, 310, 311 consolidation rate 319, 342 consolidation settlement 170, 171, 173, 290, 310, 311, 312, 318, 325 consolidation tests 170, 310, 311, 312, 341 consolidation yield stress 172, 312, 327 constant of lateral resistance of ground 295 construction conditions 7, 173, 174, 235, 327, 358, 359, 370, 398, 468 construction cost 7, 8, 167, 358, 391, 417, 426, 502, 546, 568, 569 construction joints 226, 227 construction method 8, 199, 284, 286, 336, 358, 363, 396, 405, 416, 426, 548, 568, 569, 570 construction period 8, 174, 285, 320, 341, 359, 426 construction works 8, 154, 231, 234, 238, 278, 288, 314, 318, 319, 322, 323, 326, 358, 359, 381, 407 container cranes 461, 470, 554, 574, 575, 583 container freight station 532, 583 container ships 10, 14, 17, 19, 21, 27, 354, 403, 581, 582, 583 container terminals 581 container yard 581, 583 continuity of sediment flux 165 converged embedded length 419, 420, 421, 422, 423 coping 140, 256, 262, 345, 373, 410, 414, 416, 417, 424, 425, 428, 437, 451, 460, 465, 484, 545 copper granulated blast furnace slag 238 corrected peak bedrock acceleration 191 correction factor 103, 123, 421, 422 correction factor for scattered strength 324 corrosion control 216, 217, 218, 220, 234, 272 Corrosion Control and Repair Manual for Port and Harbor Steel Structures 216 corrosion control rate 217, 218 corrosion rate 216, 217 Coulomb’s equation 304 Coulomb’s method 417 counterballast 243, 249 countermeasures against scouring 161 coupled piles 294, 301, 302, 303, 309, 434, 435, 481 coupled-pile anchorage 426, 428 covering 89, 91, 106, 109, 114, 117, 118, 119, 125, 129, 140, 223, 226, 227, 254, 259, 260, 264, 265, 266, 269, 478, 556, 571 covering of main reinforcements 254, 259, 260, 264, 265, 266, 269 covering works 394 crack widths 223 cracking 221, 223, 287, 460, 477 crashed concrete 240 creep 177, 223, 290, 312, 324, 510, 569 creep characteristics 290, 524 crest elevation 35, 36, 37, 81, 362, 363, 364, 373, 377, 378, 379, 381 cross-shore sediment transport 156 crown 396 crown concrete 363, 364, 370 crown height 83, 89, 91, 100, 104, 106, 108, 109, 113, 117, 256, 373, 378, 381, 383, 391, 396, 398, 405, 497, 591, 592 cruiser 590, 593 CU condition 173 curbing 520, 525, 527 current drag force 143, 144, 146 current efficiency 218 current force 7, 16, 22, 24, 121, 138, 144, 488, 490 current pressure coefficient 24, 25 current pressure force 24 current velocity 138, 140, 156, 385 -I-3-


currents and current force 7, 138 curtain wall breakwater 91, 376, 382, 383, 384 curved slit caisson breakwaters 125, 222 cut and cover tunnels 567 cyclic triaxial test 179, 180, 195, 198, 318 cylindrical failure surface 314 cylindrical members 119 cylindrical structures 100

design significant wave height 103, 118, 362, 379, 381 design standard traffic volume 559, 560 design tide level 252 design traffic volume 559, 560, 563 design vehicle 559, 561, 562 design water depth 405, 418 design water level 109, 127, 130, 133, 364, 369, 381, 409 design waves 41, 100, 107, 110, 116, 120, 362, 366, 373, 380 design wind velocity 29, 30 detached break-water 71 detached pier 124, 407, 485, 486, 523 detailed design 306, 372, 386, 414, 422, 428, 451, 456, 459, 460, 478, 484, 486, 502, 512, 513, 516, 574 deviation 19, 38, 39, 40, 49, 128, 130, 131, 184, 239, 240, 284, 362, 489, 523, 524 diagonal reinforcement 224 differential settlement 255, 312, 313, 461, 477, 486, 499, 546, 553, 556, 570 diffracted wave 68, 69, 94, 104, 110 diffraction 32 diffraction coefficient 33, 52, 56, 57, 58, 59, 60, 61, 62, 63, 64, 65, 66, 67, 69, 71 diffraction diagrams 52, 68, 71 diffraction force 22, 23 dimensions of target vessel 9, 402, 582 dinghy 590, 592 direction perpendicular 120, 121, 139, 301, 386, 390, 434, 483, 513, 521 directional spectrum 38, 42, 98, 150 directional spreading function 38, 39 directional spreading method 69 directional spreading parameter 39, 52 dislodging 248, 253, 254, 256, 262, 263, 264 displacement - energy curve 464 displacement tonnage 9, 16, 17, 18, 22, 583 displacements of the stabilized body 325 dissipation volume 218 distance between centers of piles 305 distance between the center of buoyancy and that of gravivity 489 distribution functions 47, 48 distribution of intensity of illumination 535 divergent waves 95, 96, 97, 98 DOL criterion 49 dolphin 16, 142, 143, 353, 407, 494, 495, 496, 523, 525, 589 dolphin mooring 142, 143 double sheet pile quaywall 516, 517, 518 double-buoy mooring 351, 352 downdrift 390, 391, 392 drag coefficient 23, 119, 120, 121, 139, 140, 143, 144, 529 drag force 119, 120, 121, 138, 139, 144 drain pile diameter 319 drain piles 319, 320, 321, 339 drainage distance 172, 320 drainage facilities 28, 89, 520, 525, 526, 570, 573 dredged soil 238, 326 dredged spoils 326 drift force coefficient 145 driven depth of cell shell 458 drying shrinkage 223 dual cylinder caisson 376, 377 dynamic modulus of deformation 178 dynamic penetration resistance 306 dynamic properties 106, 178, 186, 192, 238, 239, 240, 284, 327 dynamic water pressure 200, 205, 206, 326, 332, 365, 373, 396, 410, 417, 437, 488 dynamic water pressure during an earthquake 205, 326, 365, 396, 408, 410, 415, 417, 433, 495

D damage rate, damage level, relative damage 113 damage ratio 113, 114, 116, 117, 118, 119, 140 damping constant 178, 179, 180, 192, 472 damping factor 184, 186, 327 Darcy’s law 169, 397 datum level 402 datum level for construction work 2, 4, 128, 402 Dauham’s equation 420 Davenport 24, 27 deadweight 7, 123, 173, 184, 187, 207, 209, 221, 244, 249, 257, 258, 262, 266, 269, 277, 282, 291, 292, 364, 365, 367, 386, 389, 403, 404, 408, 410, 416, 431, 433, 463, 472, 474, 478, 481, 485, 488, 489, 495, 507, 515, 529, 555, 569, 574, 578, 586 deadweight tonnage 2, 9, 10, 11, 12, 16, 17, 19, 21, 582, 583 deadweight tons (DWT) 403, 404 deck slab 124, 463, 478, 543, 545, 546, 549, 550 deep foundations 273, 280, 281, 282 deep mixing (DM) machine 324 deep mixing method 193, 194, 322, 323, 334, 343 deepwater wave energy flux 160 deepwater wave steepness 84, 93, 106, 107 deepwater waves 32, 33, 35, 40, 41, 43, 49, 51, 94, 96, 97, 103, 149 definitions 2, 9, 10, 33, 128, 131, 323, 467 deflection 68, 223, 233, 302, 416, 422, 423, 429, 495, 511, 512, 524, 556, 595 deflection curve equation 419 deflection curve method 429 deformation level 114, 117, 192 deformation modulus 170, 177, 180, 327 deformation moment 439, 453, 517 deformation resistance coefficient 453 deformed concrete caisson type breakwater 376 degree of corrosion 220 degree of importance 5, 7, 89 density currents 138, 148 design bearing capacity coefficient 541, 542, 543, 544, 551, 552 design CBR 547, 548, 549, 550, 563 design conditions 7, 103, 104, 109, 111, 113, 118, 142, 174, 175, 220, 245, 246, 256, 261, 266, 285, 328, 358, 359, 366, 367, 376, 377, 378, 380, 385, 396, 502, 504, 512, 517, 519, 541, 547, 549, 550, 551, 552, 563, 593, 595, 596 design for compaction of loose sandly subsoil by the vibro-composer method 337 design lifetime 222 design load 245, 246, 247, 248, 249, 253, 254, 269, 465, 527, 541, 543, 544, 547, 549, 550, 551, 552, 569 design luminous flux maintenance factor 537, 538 design method 1, 2, 100, 116, 178, 182, 187, 211, 221, 226, 227, 241, 242, 243, 256, 261, 267, 270, 320, 322, 326, 329, 339, 358, 366, 367, 408, 414, 415, 417, 419, 429, 436, 456, 477, 478, 483, 487, 509, 514, 517, 519 design of lighting 531, 535, 536 design seismic coefficient 7, 182, 183, 184, 186, 187, 256, 257, 438, 472, 474, 479 -I-4-

INDEX

E

equivalent coefficient of permeability 397 equivalent crown height coefficient 89 equivalent deepwater wave height 33, 74, 75, 80, 81, 91, 93, 107, 111, 112, 114, 157 equivalent N-value 196, 197, 198, 199 equivalent relative velocity 23 equivalent wall height 441, 442, 450, 453, 454, 455, 456 equivalent wall width 436, 441, 442, 453, 454, 456, 495, 496 equivalent width of wall 439, 453, 517 equivalent-thickness method 311 erosion area of cross section 114 estuarine hydraulic phenomena 148 estuarine hydraulics 7, 148 evaluation 5, 22, 27, 42, 75, 94, 105, 127, 147, 151, 168, 174, 182, 187, 198, 227, 280, 309, 313, 326, 331, 341, 364, 367, 430, 470, 524, 533, 534 excess pore water pressure 171, 191, 310, 312, 331, 334 expected sliding distance 366, 367, 375 Extended Goda Formula 102, 103, 109, 111 extended Tanimoto formulas 118 external force generated by berthing of a vessel 16 external stability 323, 325, 327 extremely soft subsoil 319, 320

e-log p curve method 171 earth 412 earth pressure 187, 200, 201, 202, 203, 204, 221, 234, 244, 250, 253, 257, 258, 260, 262, 263, 264, 266, 267, 268, 269, 277, 280, 281, 282, 293, 304, 318, 325, 326, 327, 328, 329, 330, 331, 332, 333, 334, 335, 336, 343, 365, 396, 398, 408, 410, 411, 412, 413, 414, 415, 416, 417, 418, 419, 427, 428, 429, 430, 431, 432, 433, 437, 438, 439, 441, 442, 444, 448, 450, 451, 453, 454, 455, 492, 510, 511, 512, 513, 516, 518, 529, 555, 556, 569, 584 earth pressure against wall 329 earth pressure calculation formula 330 earth pressure of filling 250, 253, 262, 263, 266, 412 earth pressure reduction 328, 329, 413 earth retaining section 460 earth-retaining structure 464 earthquake load 221, 222, 470, 569 earthquake-resistance performance 182, 460 earthquake-resistant performance 460, 463, 469, 470, 471, 472, 473, 474, 475, 478, 480 ebb tide 160, 161 eccentric and inclined load 274, 277, 278, 279, 315, 316, 363, 366, 369, 411 eccentric distance 282 eccentricity factor 16, 20 economical design 113 effective buckling length 213 effective diameter 321, 322 effective fetch length 44 effective grain size 169, 170 effective harbor entrance width 359 effective overburden pressure 175, 176, 196, 197, 293, 311, 330 effective stress analysis and total stress analysis 191 effective surcharge pressure 279 effective voltage 218 effective weight 277, 315, 316, 317, 326, 331, 368 elastic beam analysis method 416, 419 elastic constants 170, 310 elastic solution 310 elasto-plastic method 471, 476 electrical cone test 175 electrical static cone penetration test 176, 177 embedded length 274, 275, 277, 280, 281, 282, 286, 287, 289, 293, 294, 295, 304, 417, 418, 419, 420, 421, 422, 423, 429, 431, 432, 436, 440, 447, 456, 468, 480, 484, 495, 511, 512, 513, 515, 516, 518, 556 embedded length ratio 287, 436, 440, 441 embedded-type steel plate cellular-bulkhead 452, 456, 457, 458, 459 embedded-type steel plate cellular-bulkhead quaywalls 452, 456, 457 embedment length 478, 479, 556 emi-submerged shape 381 encounter probability 8, 41, 187, 366, 367 end bearing area 287 end protection 547, 551 energy loss 43, 44, 49, 132, 134, 135 enforcement regulations for road structures 559, 562, 570, 572 environmental conditions 7, 8, 216, 217, 218, 220, 225, 234, 327, 345, 359 epicenter 131 epoxy resin coated reinforcements 226 equivalent 197 equivalent acceleration 196, 197, 198 equivalent beam method 416, 417, 418, 419, 422, 423, 514

F fabri-packed drain method 319 facilities for passenger boarding 578, 584, 586, 587 facility to trap the sediment 393 factor for effective cross-sectional area 324 fatigue failure 221, 223, 272 fatigue limit state 221, 223, 227, 242, 261 fatigue strength 224, 272 fault distance 188, 189 fences to prevent timber drifting 373 fender reaction 269, 410, 433, 463, 464, 480, 481, 488, 490, 516, 523, 524, 584 fender reaction force 410, 433, 463, 464, 480, 481, 488, 516, 523, 524 fenders 16, 17, 20, 21, 22, 25, 27, 142, 234, 428, 463, 464, 481, 486, 495, 522, 523, 524, 539, 571, 574 ferries 9, 11, 17, 18, 21, 27, 345, 354, 403, 522, 527, 532, 585 ferronickel granulated slag 238 ferry terminals 525, 564, 584, 587 fetch 42, 43, 44, 45, 46, 99 fetch length 42, 43, 44, 46, 129 fiber reinforced plastic (FRP) 234 field measurement 119, 138, 341, 342, 420, 451 field welding 215 fill 319, 438, 451 fill embankment 319 fillet welding 215 filling 221, 250, 252, 262, 263, 266, 412, 436, 437, 438, 440, 441, 442, 443, 449, 450, 454, 455, 456, 458, 518 filter sheet 234, 235, 413 final consolidation settlement 171, 310, 311 finite amplitude wave 35, 36, 106 finite amplitude wave theory 120 finite element analysis 311 finite multilayered 335, 336 finite water depth 149 fire fighting equipment 520, 525 Fisher-Tippett type II distribution 48 fixed earth support method 415, 429 fixed type 323 fixing length 224 FLAC 192 -I-5-


flexibility number 420, 421 FLIP 192 floating body 22, 142, 143, 144, 145, 146, 377, 384, 385, 386, 487, 488, 490, 502, 504, 505, 506, 507, 594, 595, 596 floating breakwater 142, 145, 147, 376, 377, 384, 385, 386, 387, 487 floating bridges 142 floating pier 401, 407, 487, 488, 490, 492, 499, 504, 593, 594, 595, 596 floating structures 7, 100, 142, 387, 487, 493, 524, 597 floating type 193, 323, 385 flocculation 150, 153 flood tide 160, 161 floor slab 123, 389, 489, 490 flow velocity parameter 119 fluid mud 150, 151, 153 fluid mud layer 151 fluorescent sand tracers 157, 159 FLUSH 192, 193 flux method 536 fly ash 225, 239, 240 foam treated soil 326 foot protection block 226, 371, 372 footing 105, 243, 251, 253, 255, 256, 257, 258, 259, 261, 264, 370, 376, 377 footway live load 593, 596 forced displacement method 318 foreshore 154, 160, 161 forged steel 214 foundation ground 273, 276, 277, 278, 279, 316, 373, 374, 398 foundations 273, 274, 276, 277, 279, 280, 281, 282, 310, 314, 317, 369, 372, 495, 518 free earth support method 415, 416, 417, 418, 419 frequency 10, 12, 13, 14, 15, 19, 22, 24, 37, 38, 39, 40, 42, 43, 50, 99, 104, 150, 178, 221, 222, 223, 568, 587 frequency spectrum 24, 38, 39, 42, 99, 104 frequency spectrum of wind velocity 24 friction coefficient 24, 210, 331, 332, 366, 367, 368, 455 friction drag 143 friction increasing mats 229 friction piles 284, 285, 304, 306, 482 friction-increasing mat 366 frictional resistance 24, 140, 210, 280, 412 front toe reaction force 455 Froude-Kriloff force 22, 23 fueling and electric power supply facilities 520, 526 Fujita’ formula 28 fully plastic state moment 473, 474, 475, 476 Fundamental Law for Countermeasures against Natural Disaster 186

Goda pressure formula 102 Goda’s Formula for Estimating Surf Beat Amplitude 92 Goda’s wave pressure formula (kN/m) 382 gradient winds 29, 42 grain size characteristics 155, 169 grain size distribution curve 155, 169, 170, 195 granulated blast furnace slag 238, 328, 329, 333, 334, 343 gravity type quaywalls 182, 183, 193, 221, 222, 277, 398, 407, 408, 411, 413, 477, 486, 499, 523 gravity type special breakwater 376, 377, 382 gravity type structure 136, 277, 322, 339, 556 Green functions 69 Green’s equation 132 grid 130, 193, 321, 322, 536, 537 groins 154, 161, 357, 392, 393 gross tonnage 2, 9, 10, 11, 16, 17, 21, 25, 26, 27, 497, 520, 521, 527, 582, 585, 587 gross tons 403, 404 groundwater level 135, 160, 161, 167, 281, 289, 340, 373, 397 group velocity 35, 43, 98, 150, 159, 166 grouting material 234, 237 Gumbel distribution 47, 48 gust factor 30

H H-shaped steel 270, 287 Hachinohe bedrock incident waves 189 Hamada 150, 227 handicapped people 396, 398, 580 handrails 234, 396, 578 harbor 359 harbor calmness 32, 69, 94, 357, 508 harbor entrance 16, 22, 69, 74, 94, 134, 348, 349, 358, 359, 373, 374, 390, 391, 592 harbor paradox 134 harmful effects of freezing and thawing 225 harmonic analysis of tidal observation data 4 haunch 255, 272 hazardous cargoes 401 Hazen 47, 169 headed studs 271 heaving 146, 354, 405 heliports 567 Hertz’s equation 212, 214 high crested upright wall 104 high seismic resistant structures 182, 183, 186, 187 high water of ordinary spring tides 128 high-density blocks 117 high-fluidity concrete 225 high-speed ferry 96 highest one-tenth wave 33, 37, 90 highest one-tenth wave height 37, 90 highest wave 33, 35, 37, 38, 77, 78, 80, 100, 102, 103, 105, 107, 108, 109, 120, 123 highest wave height 33, 35, 37, 38, 77, 78, 80, 102, 103, 105, 107, 108, 109, 120, 123 highly flowable concrete 225 Highway Bridge Specifications and the Commentary 213, 223 Hiley’s equation 306 Hino 24 hinterland 41, 89, 581, 584 Hiroi formula 103 holding powers 351 Hom-ma’s formula 132 hooks 255, 260, 261 Horikawa 107, 111, 124, 160

G galvanic anodes method 218 gate supports 389 gates 389, 567, 581, 583 generalized Hudson formula 112, 113, 118 generated electricity flux 218 geometrical moment of inertia 419, 421, 488, 489 geostrophic wind 29 geosynthetics 234 geotechnical conditions 167, 359 glare 535 global warming 8 glued laminated timber 233 Goda equation 102 Goda formula 102, 103, 104, 105, 594 -I-6-

INDEX

horizontal coefficient of consolidation 311, 313, 321, 322 horizontal force 282, 294, 301, 303, 305, 309, 316, 386, 410, 411, 412, 429, 433, 434, 437, 445, 447, 451, 463, 464, 466, 467, 469, 474, 475, 478, 479, 480, 481, 482, 483, 485, 486, 491, 492, 495, 504, 505, 516, 518 horizontal methods 257 horizontal plate 122, 123, 124, 462, 480 horizontal seismic coefficient 184, 186, 282, 464 horizontal shear modulus 443, 448 horizontal slit 267, 380 horizontal tension 455, 456, 458, 459 horizontal-plate type permeable breakwaters 91 Hudson’s formula 112, 113, 117, 118 HWL 128, 216, 217, 362, 373, 392, 508 HWOST 128 hybrid caissons 270, 271, 272 hydraulic gradient 136, 169 hydraulic model experiments 16, 33, 39, 69, 71, 79, 80, 83, 84, 90, 100, 105, 106, 109, 110, 111, 112, 113, 117, 119, 121, 122, 132, 140, 143, 145, 146, 163, 267, 268, 364, 366, 373, 376, 377, 378, 379, 380, 381, 382, 384, 595 hydraulic phenomena 148 hydraulic radius 148 hydrostatic pressure 102, 106, 108, 200, 221, 244, 248, 249, 250, 251, 252, 253, 327, 364, 365, 373, 386, 398, 488, 490, 569, 594 Hyogoken-Nanbu Earthquake 182, 183, 187, 189, 190, 192, 193, 194 hyperbolic wave 36

inshore 154 inspection 5, 117, 288, 401, 538, 568, 583 installation depth 222 integrity of concrete blocks 116 intensity of rainfall 28 intensity of wave pressure 101, 105, 107 intermediate soil 173, 174, 175, 181 internal friction angle 231, 274, 275, 277, 278, 279, 280, 314, 317, 329, 330, 334, 340, 441, 442, 448, 449, 450, 453, 455, 456, 464 internal water pressure 221, 244, 250, 253 International Hydrographic Organization 4 international marine chart datum 4 international marine container 561 IPCC 127, 136 Iribarren number 114 Iribarren-Hudson formula 113 Irie 162, 166 irregular wave 32, 39, 41, 50, 51, 52, 69, 71, 72, 74, 75, 80, 84, 99, 100, 102, 103, 107, 112, 113, 117, 119, 125, 141, 145, 150, 375, 504, 509, 518 Isbash’s constant 140 Isbash’s Equation 140 Ishiguro 419 Isozaki and Uji’s MRI model 43

J Japanese Agricultural Standards (JAS) 233 Japanese Industrial Standards (JIS) 212 Japanese unified soil classification system 169, 180 jetty 357, 390, 391, 392, 393 Johnson 149, 153 joint board 234 joint distribution 47, 94 joint sealing materials 234 joints 135, 137, 142, 226, 227, 234, 235, 271, 272, 288, 291, 293, 307, 308, 370, 397, 409, 413, 442, 451, 459, 468, 478, 479, 517, 540, 545, 546, 553, 554, 569, 570 judgment of liquefaction 195, 198

I IHO 4 illumination intensity calculation method 536 Immediate Settlement 310 immersed tunnel element 568, 569, 570 immersed tunnels 182, 183, 184, 192, 193, 567, 570 impact load 284, 586 impact velocity 123 impact wave force 122, 359 impermeable type 267, 377 importance factor 184, 186, 187 importance of breakwaters 359 improved subsoil system 323 impulsive breaking wave force 101, 104, 108, 109, 125, 363, 364, 379, 381 impulsive breaking wave force coefficient 104 impulsive pressure 102, 104, 105, 106, 107, 108, 112, 124, 268 impulsive uplift 122, 124, 268 impulsive wave breaking force 100 impulsive wave pressure 102, 107, 108, 262 in-situ CBR 548 in-situ permeability test 169, 175 in-situ tests 179, 180, 323, 325 incident wave height 33, 52, 69, 110, 123, 132, 145 incident waves 22, 23, 68, 69, 70, 71, 72, 83, 91, 102, 109, 110, 112, 120, 133, 144, 146, 163, 189, 190, 191, 349, 359, 379, 390 increase factor 223, 330, 331 indoor lighting 532, 534, 537 inertia coefficient 119, 120, 121, 122 inertia force 119, 120, 121, 178, 373, 594 infiltration 234, 320, 373, 413, 547 infiltration of sediment 235 influence coefficients on crack width 243, 245, 247 inorganic lining 220 Inoue’s model 43

K Karman’s theory 123 Katoh 159, 160, 166 KC number 120 KD value 112, 113, 116 keel clearance 17, 348, 349, 350, 353, 388, 402, 591 Kelvin’s theory 96, 98 Keulegan-Carpenter number 120 Kirby and Parker1 151

L L-shaped block type quaywalls 408 L-shaped blocks 256, 257, 258, 259 L-shaped member 139 L-type scouring 163 laboratory mixing tests 323, 330 laboratory model experiments, in situ prototype experiments 277 laboratory permeability test 169 lakes 2, 127 landfill 173, 174, 326, 328, 329, 330, 396, 460 landfill material 238, 329, 413, 424, 431, 540 lane 559, 560, 561, 562 large isolated structures 119 large-scale triaxial compression tests 231, 278 LAT 4 -I-7-


latent hydraulic property 328, 329 lateral axial spring constant of pile head 304 lateral bearing capacity 281, 282, 293, 294, 295, 301, 302, 304, 483, 511, 516 lateral displacement 303, 312, 328, 416 lateral flows 429 lateral loading tests 295, 300, 465 lateral resistance of piles 309, 465, 480, 511 lateral strength 386 layer equivalency factor 548 layout of breakwaters 351, 358, 359 length between perpendiculars 20, 21, 23 levee 81, 84, 89, 132, 246, 357 level 1 earthquake motion 183 level 2 earthquake motion 183, 187, 190, 191 level crossings 562 life cycle cost 5 lifesaving facilities 520, 527, 589 lifetime 7, 8, 41, 116, 125, 182, 183, 187, 211, 220, 221, 226, 230, 366, 367 lift coefficient 139, 140 lift force 120, 121, 125, 126, 138, 139, 140, 269 lighthouse 365, 369, 370 lighting facilities 389, 501, 520, 525, 531, 573 lightweight aggregate concrete 224 lightweight treated soil 326, 327, 328 lightweight treated soil method 326 limit state 243 limit state design method 2, 211, 221, 226, 227, 241, 242, 243, 256, 261, 267, 270, 477, 478 line load 310 liquefaction 7, 180, 181, 182, 193, 195, 196, 198, 199, 273, 284, 292, 318, 329, 330, 331, 333, 334, 336, 452, 460, 569 liquefaction countermeasure works 199 liquefaction prediction and judgment 195 littoral drift 7, 148, 150, 154, 155, 156, 157, 158, 159, 165, 166, 355, 359, 372, 388, 390, 391, 393 live load 207, 209, 462, 463, 478, 488, 489, 495, 569, 578, 586 Lloyd’s Maritime Information 9, 347 LNG carriers 2 load - settlement curve 285 load and pile head displacement curve 295 load carrying capacity 460, 470, 471, 473, 474, 475, 476, 477 load carrying capacity design method 460 load factor 221, 222, 243, 244, 245, 246, 247, 248, 257, 262, 268 load from cargo handling machinery 221 load inclination ratio 277 loading arms 574 loading tests 284, 285, 286, 287, 288, 292, 295, 296, 301, 303, 306, 478 local buckling 477 lock 388, 389 lock chamber 388, 389 lock gate chambers 389 locks 357, 388, 389 logarithmic extreme value distribution 47 long linear wave equations 93 long piles 296, 426 long walls 323, 326 long waves 35, 128, 132, 133, 149 long-distance Japanese ferries 403, 585 long-period waves 16, 22, 40, 93, 94, 99, 134, 135, 374, 375 long-term consolidation 310, 312 longitudinal bending moment 386 longitudinal construction joints 545, 546 longitudinal slope 561, 562, 568 longitudinal strength 386

longshore currents 138, 148, 165 longshore sediment transport 156, 157, 355, 391, 392 longshore sediment transport rate 154, 158, 159, 165 Longuet-Higgins 149, 153 low crested upright wall 104 low water level 2, 220 low water of ordinary spring tides 128 lowest astronomical tide 4 LPG carriers 2 luni-solar diurnal tide 4 LWL 128, 216, 217, 220, 250, 263, 409, 417, 438, 497, 508 LWOST 128

M mach-stem waves 358, 390 maintenance 1, 5, 8, 127, 150, 153, 218, 220, 233, 236, 355, 357, 358, 359, 389, 401, 407, 448, 491, 501, 531, 537, 538, 540, 542, 545, 548, 552, 554, 557, 559, 569, 571, 572, 574, 575, 583 maintenance dredging 153, 355, 393, 569 maintenance factor 536, 537, 538 maintenance Shop 583 Manning’s roughness coefficient 148 Manning’s roughness formula 132 Manual for Maintenance and Repair of Port and Harbor Structures 5 marinas 1, 531, 589, 591, 593, 595, 599 Marshall stability level 548 mass concrete block type composite breakwater 360 mass concrete block type upright breakwater 360 mast height 9, 347, 571 material factor 221, 222, 479 mattress works 372 maximum draft 353, 402 maximum horizontal tensile stress 458 maximum instantaneous wind velocity 30 maximum scouring depth 162 maximum wave 33, 47, 74, 97, 102, 103, 106, 107 mean adhesion 281, 286, 293 mean differential settlement ratios 313 mean high water level (MHWL) 128, 216 mean low water level (MLWL) 128, 216, 217, 218, 220 mean monthly-highest water level 128, 362, 363, 364, 379, 392, 405, 508, 526 mean monthly-lowest water level 128, 217, 220, 379, 409, 417, 438, 497, 526 mean period 33, 40, 91 mean sea level (MSL) 4, 127, 128, 353, 363 mean water level 36, 80, 91, 92, 120, 128, 130 mean wave 33, 37, 39, 90 mean wave height 33, 37, 39, 90 mega-float 142, 145 member factor 221, 222 mesh 42, 130, 544 metacenter 243 Meteorological Agency’s Technical Observation Notes 30 method based on the gradation and N-value 195 method of cathodic protection by power impression 218, 219 method of Koegler 310 method of least squares 48 Miche’s equation 81 Mikasa’s consolidation theory 173 mild-slope wave equation 51 Ministry of Transport Notification 1, 2 Ministry of Transport Ordinance 1, 567, 570 MIR criterion 49 -I-8-

INDEX

Mitsuyasu equation 107 Mitsuyasu type spreading function 38 mixing blades 324 model experiments 16, 21, 32, 33, 39, 69, 71, 79, 80, 83, 84, 90, 100, 104, 105, 106, 109, 110, 111, 112, 113, 114, 116, 117, 118, 119, 120, 121, 122, 132, 140, 143, 145, 146, 163, 222, 267, 268, 296, 364, 366, 373, 376, 377, 378, 379, 380, 381, 382, 383, 384, 420, 456, 458, 508, 595 moderate heat portland cement 225 modification factor due to wave breaking 114, 116 modified Fellenius method 276, 277, 315, 316, 340 modified Rowe’s method 419 modulus of elasticty 212 modulus of horizontal subgrade reaction 443, 448 modulus of subgrade reaction 443 modulus of vertical subgrade reaction 443, 448 Mononobe-Okabe principle 334 Monte Carlo simulations 367 moored vessels 22, 28, 29, 133, 353, 463, 503, 522, 523 mooring / unmooring basin 353 mooring anchor 142, 351, 352, 386, 487, 491, 492, 502, 504, 505 mooring basin 7, 153, 502, 503, 507, 574 mooring buoy 142, 145, 146, 401, 407, 502, 503, 504, 506, 507, 589 mooring by universal joint 143 mooring chains 487, 488, 490 mooring equipment 142, 146, 475, 520, 521, 571, 573, 582, 584, 585 mooring facilities 1, 16, 17, 19, 20, 21, 22, 26, 27, 221, 345, 352, 353, 357, 401, 402, 404, 405, 407, 469, 497, 499, 508, 520, 523, 525, 527, 532, 539, 574, 582, 585, 586, 589, 593, 595, 596 mooring force 142, 143, 145, 146, 385, 488 mooring lines fixed to anchors 377 mooring piles 354, 373, 401 mooring post 25, 26, 27, 226, 407, 424, 463, 486, 495, 502, 507, 520, 521, 522, 582, 585 mooring rings 26, 520, 521, 582, 585 mooring rope 25, 26, 93, 146, 377, 402, 521 mooring rope tension 377 mooring ropes 25, 26, 93, 402, 521 mooring with two anchors 351 mooring-related matters 386 Morison’s equation 119, 120 motions of floating body 145, 385, 386 motions of moored vessel 22 motorboat 2, 96, 346, 590 mound materials 278 mountain tunnels 567 moving parts 389, 586, 587 Muller equation 386 Muller method 596 multi-storied storage facilities 598 multicomponent coupling method 69 multidirectional random waves 89 multilayered ground 276 multiple low fences 394 multiple reflection model 187, 191, 192, 472 multiple-vertical cylinder breakwaters 91 multiple-wing type permeable breakwaters 91 mv method 171 Myers’ formula 28, 130

N-value 168, 174, 175, 176, 177, 185, 189, 195, 196, 197, 198, 199, 279, 286, 290, 291, 293, 296, 297, 300, 318, 330, 336, 337, 338, 420, 443, 458, 465, 466 N-value at pile center 338 narrow band 37 natural beach 72, 154, 161 natural conditions 8, 41, 230, 347, 350, 355, 357, 358, 388, 390, 405, 407, 490, 502, 503, 519, 520, 567, 569, 571, 589, 596 natural period of ground 472 natural period of the pile-supported section 471, 472 natural periods 131, 133, 134, 186, 472 nautical charts 4 navigation aids 346 navigation channel 346, 347, 348 navigation of vessels 28, 70, 150, 346, 347, 348, 368, 391, 570, 571, 574 near highest high water level (NHHWL) 128 negative skin friction 284, 289, 290, 291, 308, 325 negative uplift pressure 105 negative wave force 105, 106 negative wave pressure 105, 106 net method 234 Newmark’s equal-energy principle 474 Nihon Senpaku Meisaisho 9 Noda 187, 193, 194, 451, 459 nominal diameter 112, 114, 116 nominal strength 222 non-exceedance probability 48, 49 nonlinear interaction 42, 43, 149 nonlinear long wave theory 74 nonlinearity 22, 74, 122, 170, 178, 374, 470, 475 nonpermeable structure 391 Nonwoven cloth 235 normal distribution 47, 367 NOWPHAS 41 NS value 113 numerical calculation 52, 121, 133, 136, 150, 419, 518 numerical computation 69, 130, 131, 132, 166 numerical simulations 132, 146, 163, 165 nylon ropes 26

O oblique compression characteristics 524 obliquely incident waves 68 observation records 128, 132 occurrence probability 37, 41, 94, 130, 367 ocean currents 128, 138, 386 ocean-line container terminal 581, 582 offshore petroleum stockpiling bases 142 offshore zone 156, 390, 393 Ofunato bedrock incident waves 189 one-dimensional consolidation 170, 171, 313, 320 one-line theory 165 open circuit anode voltage 218 open storage yard 208, 540, 576 open-type wharf 123, 124, 182, 223, 401, 407, 416, 460, 461, 462, 463, 464, 465, 466, 468, 469, 470, 471, 473, 474, 475, 476, 477, 478, 479, 480, 481, 483, 484, 486, 495, 499, 507, 514, 515, 523, 525, 543, 545, 546, 549, 550, 574, 589 open-type wharves on coupled raking piles 407, 480, 483, 486, 495, 499, 507, 514, 515 open-type wharves on vertical piles 407, 460, 465, 480, 484, 486, 495, 507, 514, 515 opening ratio of ventilation holes 268 operational function of the facilities 7 ordinary waves 32

N n-th moment of the wave spectrum 40 N-type scouring 163 -I-9-


oretical generated electricity flux 218 organic lining 219, 220 original sea bottom depth 405 oscillatory motions of moored vessels 463 other non-gravity type of breakwater 376 outdoor lighting 531, 532, 534, 535, 537 overburden pressure 171, 175, 176, 177, 196, 197, 276, 293, 311, 330, 337 overconsolidated clay 311 overlapping length 224 overtopped waves 362, 368 overtopping 33, 41, 72, 80, 81, 84, 89, 90, 99, 362, 363, 364, 371, 373, 378, 379, 383, 386, 391, 392, 398, 518 overturning 112, 130, 295, 315, 323, 365, 366, 368, 409, 410, 411, 412, 437, 439, 452, 454, 455, 495, 496 overturning resistance coefficient 454 Ozasa and Brampton 165

pile joints 284, 291, 417, 440, 442, 468 pile lateral resistance 468 pile type dolphins 495 pile type fenders 522 pile-supported section 460, 461, 462, 463, 464, 465, 466, 467, 469, 470, 471, 472, 473, 474, 475 pile-supported type breakwater 376, 377 pipe type breakwaters 91 pipeline 139, 182, 192, 532 pitching 146, 354, 405, 578 placement-type steel plate cellular-bulkhead quaywalls 452 plain concrete 207, 223, 224, 225, 226, 227, 252 planar slip surface 314, 316 plantation works 394 plastic sectional modulus 476 plastic-board drain 320 plastics 234 plate load test 542, 551 pleasure boats 2, 589, 590, 591, 592, 593, 594, 595, 596, 597, 598, 599 plunging breakers 114 pneumatic fenders 522 pneumatic type breakwater 376, 377 point-by-point method 536 Poisson’s ratio 170, 179, 180, 212, 310, 327, 544 pontoon 234, 377, 386, 387, 487, 488, 489, 490, 491, 492, 493, 595, 597 porous caisson 380 Port and Harbor Construction Work Common Specifications 227 Port and Harbour Law 1, 401 Port and Harbour Law Enforcement Order 1 Port and Harbour Law Enforcement Regulations 1 Port Island bedrock incident waves 189 port traffic facilities 559, 563, 567, 599 portland blast-furnace slag cement 225 portland fly ash cement 225 pozzolan activity 239 precast wave-dissipating concrete blocks 72 preconsolidation pressure 319 prediction of liquefaction 195, 196, 198, 199 preloading method 320 premixing method 329, 331, 333 pressure coefficient 24, 25, 30, 108, 143, 144, 332, 455 pressure gradient 29 pressure moment coefficient 23 prestressed concrete 221, 476, 478, 490, 568, 571 Prestressed Concrete Barge Standards 386 Prestressed Concrete for Ports and Harbors Structure Design Manual 221 prestressed high-strength concrete piles 305 primary consolidation 171, 172, 173, 312 principal direction 38, 41, 51, 52 principal lunar diurnal tide 4 principal lunar semi-diurnal tide 4 principal solar semi-diurnal tide 4 probability density function 19, 37, 39 progressive 124 progressive waves 23, 104, 107, 110, 123, 124 protective current density 219 protective facilities 1, 47, 345, 348, 357, 401, 589, 592 protective fences 525, 527, 530 protective potential 218, 219 prototype design 113 pulling resistance of the piles 461 pulling tests 292, 513 pump dredgers 398 pumping and drainage system 389

P P-N-J method 46 parapet retreating type seawall 89 parapets 107, 357 parking lots 401, 564, 565, 566, 572, 583 partial safety factors 2, 221, 222 particle density 168, 238, 240, 330 particle size distribution 318, 337 partition slit column 269 passageways 532, 573, 587 passenger building 579, 580, 588 passenger ship 2, 527 passenger terminals 398, 580, 588 passive earth pressure 200, 201, 202, 204, 281, 282, 304, 325, 416, 417, 418, 427, 428, 429, 432, 433, 439, 441, 444, 448, 510 peak frequency 39 peak strength 327 peak waves 47, 48 penetration depth 287, 384, 419, 551, 563, 568 perforated wall 125, 267, 269, 380 perforated-wall caisson 106, 111 Permanent International Association Navigation Congress 524 permanent load 221, 222, 223, 244, 247, 248 permeability coefficient of the inner layer 114 permeable rubble-mound breakwater 135 permeable type 91, 377, 383, 592 permeable wall 380 permeation coefficient 135 permissible rate of overtopping 89 petrolatum lining 220 PHRI method 295, 296, 465, 511, 512 physical property of the facilities 8 Picket equation or Arlington equation 544 pile 291 pile breakwater 105, 383, 384 pile driven by hammer 286 pile driving formulas 306, 307 pile driving tests 307 pile foundation 167, 273, 284, 309, 384, 431, 434, 496, 513, 555 pile group 284, 288, 289, 290, 291, 293, 301, 305, 482 pile group block 289 pile head displacement 294, 295, 304, 475 pile head fixed 304 pile head hinged 304 pile head moments 466, 467 pile installation by inner excavation 287 pile installation method 287 -I-10-

INDEX

punching shear 224, 479 punching shear failure 276 pure car carriers 11, 19, 354, 404 pushing-in bearing capacity of pile 477

relieving 434 relieving platform 418, 431, 432, 433, 434, 435, 518, 523 relieving platform piles 431, 432, 433, 434 removal and replacement method, 318 repair 5, 401, 470, 471, 475, 497, 501, 527, 538, 568, 583, 598 replacement method 318, 328 required relative density 336 residual displacement 470, 471, 475, 523, 524 residual water level 167, 201, 202, 204, 205, 217, 232, 235, 258, 263, 331, 332, 396, 397, 409, 410, 413, 414, 417, 418, 427, 432, 438 residual water pressure 205, 221, 257, 260, 262, 263, 267, 269, 326, 408, 411, 412, 415, 416, 417, 418, 419, 432, 433, 437, 439, 453, 454, 510, 511, 512, 513, 516 resistance coefficient 97, 136 resisting moment 177, 278, 315, 412, 439, 440, 441, 442, 450, 453, 454, 455, 517 response analysis 22, 178, 182, 183, 186, 187, 189, 190, 191, 193, 194, 196, 198, 333, 461, 472, 574 responsible engineer 369 restoring force 143, 145, 146, 191, 192 resultant earth pressures 336 return period 8, 29, 41, 47, 48, 130, 183, 187, 188, 222, 472 revetment 89, 161, 183, 220, 261, 263, 267, 270, 272, 312, 322, 328, 329, 339, 357, 396, 397, 398, 460, 462, 469, 471, 531, 535 Reynolds number 30, 120, 121, 125, 136, 139, 140, 143, 144 rheology characteristic 151 Rice’s theory 40 rip currents 138 river mouth 149, 150, 357, 392, 395 river’s waterway 392 rivers 2, 98, 127, 128, 148, 154, 392, 393 road tunnels 567 roadway 559, 560, 561, 562, 565 rocking 368, 408 roll-on roll-off Ships 404 rolling 20, 146, 230, 349, 354, 405, 578 root mean square amplitude 93 Rowe’s method 415, 416, 418, 419, 421 rubber expansion joints 234 rubber fender characteristics curve 464 rubber fenders 464, 522, 524 rubber mat 235 rubbers 8, 234 rubble backing 363 rubble for foundation 231 rubble mound 72, 90, 91, 106, 112, 113, 114, 116, 118, 125, 136, 140, 162, 163, 230, 234, 235, 277, 279, 316, 360, 363, 364, 366, 368, 369, 371, 372, 373, 374, 379 rubble mound breakwater 112 rubble mound foundation 363, 364, 366, 368, 369, 371, 373 rubble mound type sloping breakwaters 360, 364, 372 rubble stones 112, 117, 140, 231, 326, 363, 367, 368, 371, 372, 391, 464, 465 runup 33, 80, 81, 82, 83, 131, 132, 166, 392 runup wave 392

Q qu method 174, 175 quantity of overtopping 83, 84, 89, 132 quay sheds 573 quaywalls 23, 27, 70, 133, 135, 182, 183, 192, 193, 208, 221, 222, 244, 253, 263, 267, 272, 277, 278, 312, 322, 328, 329, 339, 349, 397, 398, 401, 407, 408, 411, 412, 413, 414, 418, 425, 431, 436, 449, 452, 456, 477, 486, 493, 496, 499, 508, 509, 512, 514, 517, 519, 523, 554, 574

R radius of gyration 20, 213, 599 rail-type traveling cargo handling equipment 554 rails for traveling of equipment 554 railway tunnels 567 railways 567 random placement 113, 117 Rankine’s equations 441 rate of increase in strength 340 rate of overtopping 84, 85, 86, 87, 88, 89 ratio of strength increase 175, 319, 340 Rayleigh distribution 37, 38, 39 reaction at the tie rod setting point 418, 424, 427, 433 reaction forces 146, 410, 444, 488, 493, 516, 524 reaction from fender 268 ready-mixed concrete 225 rebound 307 rebound coefficient 307 REC criterion 49 reclamation revetments 397 recreational fishing 358 rectangular concrete block type quaywalls 408 recyclable resources 238 reduction of earth pressure 329, 332 reef 79, 80, 99, 110, 111, 113, 116, 125, 126, 379 reflected waves 23, 41, 70, 71, 94, 99, 132, 351, 357, 358, 370, 381, 390, 391, 508, 591 reflection 32, 69, 71, 72, 94, 110, 125, 132, 135, 145, 161, 187, 191, 192, 378, 380, 387, 508, 509, 537 reflector sheet 530 refraction 32, 33, 39, 41, 49, 50, 51, 52, 69, 70, 71, 74, 94, 98, 149, 150, 153, 159, 359 refraction coefficient 33, 41, 49, 50, 51, 69, 159 Regional Disaster Prevention Plan 186, 187 regional seismic coefficient 184, 185, 187, 188, 472 regular placement 117 regular waves 22, 49, 50, 51, 72, 75, 79, 80, 83, 89, 100, 107, 112, 113 reinforce 250 reinforced 211, 252 reinforced concrete 207, 211, 223, 224, 225, 226, 227, 241, 270, 271, 412, 431, 475, 476, 481, 515, 547, 556, 568, 571 reinforced concrete piles (RC piles) 305 reinforcement 5, 223, 225, 228, 229, 230, 260, 374, 451, 478, 516, 546, 571 relative damage 113, 116 relative density 168, 176, 300, 337, 420 reliability 1, 2, 42, 47, 110, 116, 175, 177, 190, 198, 217, 288, 293, 324, 366, 367, 375 reliability design method 1, 2, 116, 366, 367, 375 reliability factor of overlapping 324

S S-M-B method 43, 44, 46 safe nautical depth 150 safety factor 1, 7, 26, 187, 222, 227, 243, 249, 274, 275, 276, 277, 278, 280, 281, 282, 283, 284, 285, 289, 290, 291, 292, 294, 307, 310, 312, 314, 315, 316, 317, 318, 320, 324, 327, 331, 332, 340, 341, 365, 366, 367, 368, 375, 409, 410, 411, 412, 417, 418, 421, 424, 425, 427, 439, 448, 453, 454, 492, 496, -I-11-


519 Sainflou formula 103 Sakai 150, 153 Sakamoto-Ijima method 44, 45 sand bar 391, 393 sand compaction pile method 175, 318, 336, 339, 340, 341, 342, 343 sand fences 394 sand filling 135, 363, 397 sand mastic asphalt 228, 229, 230, 369, 391 sand mat 238, 319, 320, 339 sand ripples 156 sand volume to be supplied 336 sandy beach 135, 154, 357, 359 sandy ground 274, 275, 277, 279, 280, 286, 289, 291, 293, 296, 310, 314, 420 Sato 156 saturated copper sulfate electrode 218, 219 saturated mercurous chloride electrode 218, 219 Savage 159, 166 scour prevention 234, 372, 374, 406 scour-prevention works 398 scouring 161, 162, 163, 234, 235, 363, 369, 371, 372, 373, 384, 390, 391, 393, 405, 406, 497, 568, 586 scouring prevention mats 229 sea bottom topography 32, 80, 82, 89, 100, 130, 346 seabed gradient 222 seal materials 234 seal plates 234, 235 seal rubber gaskets for immersed tunnel units 234 seals for dredging pipe joints 234 seawalls 80, 99, 166, 173, 174, 357, 396, 518, 589, 591, 592 seawater immersion ratio 217, 218 seawater-silver chloride electrode as the reference 218 secondary consolidation 171, 172, 173, 312 section modulus 468 sectional strength 308, 386 sediment grain size 154, 157, 160 sediment sorting action 154 sedimentation 153, 166, 359 seep-proof screen 373 seepage 137, 169, 237, 396, 397, 398, 569 seepage analysis 397 seepage efficiency 237 seepage limits 237 seiche 40, 93, 94, 127, 128, 131, 132, 133, 134, 135, 591 seismic coeffecient 374 seismic coefficient 182, 184, 186, 187, 188, 189, 193, 202, 204, 205, 332, 335, 422, 438, 463, 471, 472, 474 seismic coefficient method 178, 180, 182, 183, 184, 186, 187, 190, 333, 471, 480 seismic deformation method 182, 190, 192 seismic force 7, 178, 180, 182, 244, 277, 279, 314, 315, 316, 317, 334, 335, 365, 369, 374, 396, 408, 410, 411, 431, 437, 438, 460, 469, 485, 493, 495, 570, 574, 584, 586 seismic inertia force 326 seismic response analysis 178, 182, 183, 187, 189, 190, 191, 196, 198, 472, 574 self weight of embankment 318 self-hardening coal ash 333 selfweight 249 semi-container ships 582 semi-infinite breakwater 52, 53, 54, 55, 71, 74 semi-infinite elastic body 310 semitrailer truck 559, 560, 561, 562, 563 service conditions 207, 208, 209, 359 serviceability limit state 221, 222, 223, 243, 245, 246, 248, 256,

261, 478 setting level of tie rod 417 settlement 170, 171, 172, 173, 177, 184, 230, 234, 255, 273, 274, 276, 285, 290, 291, 302, 303, 310, 311, 312, 313, 318, 319, 323, 325, 328, 341, 342, 363, 366, 367, 369, 372, 390, 393, 409, 410, 411, 413, 424, 429, 460, 461, 477, 478, 485, 486, 499, 540, 542, 546, 551, 553, 556, 557, 569, 570 settlement of mound 242 settlement of piles 284, 291 settlement ratio 313, 341, 342 SHAKE 192, 193, 472 shallow draft quays 407, 497, 499, 500, 501 shallow foundations 273, 274, 326, 455 shallow water waves 34, 35, 43, 44, 45, 96, 103, 150 shape factor of foundation 274, 275 shear deformation 310, 312, 437, 438, 439, 453, 457, 495, 517, 518, 523 shear deformation of the wall body 453 shear failure 274, 276, 310, 438, 477, 478 shear modulus 177, 178, 179, 180, 192, 212, 327, 518 shear reinforcement 222 shear resistance steel 272 shear strain amplitude 178, 179, 180 shear strength 168, 173, 174, 175, 177, 231, 275, 289, 290, 314, 315, 327, 329, 339, 340, 341 shear stress 138, 178, 179, 192, 196, 197, 212, 213, 214, 260, 271, 312, 314, 315, 324, 448, 458, 468 shear wave velocity 189, 472 shearing force 224, 271, 386, 394, 522, 523, 556, 569, 596 sheet 426 sheet flow 155, 156 sheet pile 135, 136, 137, 184, 193, 212, 213, 214, 215, 216, 217, 220, 318, 397, 398, 407, 415, 416, 417, 418, 419, 420, 421, 424, 425, 426, 427, 428, 429, 430, 431, 432, 433, 435, 436, 437, 438, 440, 442, 443, 444, 448, 450, 451, 455, 468, 477, 486, 509, 510, 511, 512, 513, 514, 515, 516, 517, 518, 523, 546, 556 sheet pile anchorage 426, 428, 517 sheet pile quaywall with (a) relieving platform 407, 431, 435 sheet pile quaywall with batter anchor piles 512, 515 sheet pile quaywall with batter piles in front 514 sheet pile quaywalls 407, 415, 425, 426, 431, 432, 477, 511, 512, 513, 514, 515, 516, 517, 523, 556 sheet pile wall 135, 136, 397, 415, 416, 417, 418, 419, 420, 421, 424, 426, 427, 429, 430, 431, 432, 433, 435, 509, 510, 511, 512, 513, 514, 515, 516, 517, 518, 556 Sherman’s formula 28 shield tunnels 567 ship dimensions 7, 27 ship navigation 358, 581 ship waves 94, 95, 96, 97, 98, 591 ship-generated waves 94, 346, 351 shoaling 32, 33, 36, 41, 49, 69, 74, 75, 78, 81, 96, 130, 144, 153, 359, 372, 390, 391, 392, 393, 394 shoaling coefficient 41, 49, 69, 74, 75, 78, 81 shoreline 36, 51, 81, 91, 92, 104, 111, 129, 130, 140, 154, 156, 160, 165, 166, 390, 391, 392 short- and medium-distance Japanese ferries 403, 585 SI units 2, 3 side resistance 274, 280, 281 side thrusters 585, 586 sidewalk live load 207, 209, 410, 463, 486, 586 significant wave 32, 33, 37, 38, 39, 40, 41, 42, 43, 44, 46, 47, 69, 70, 76, 77, 78, 79, 80, 90, 91, 93, 94, 101, 102, 103, 104, 107, 112, 114, 118, 120, 130, 162, 379, 392, 461, 508 significant wave height 33, 37, 38, 40, 41, 42, 43, 44, 46, 69, 70, 76, 77, 78, 79, 80, 90, 93, 94, 101, 102, 103, 112, 114, 118, -I-12-

INDEX

130, 162, 362, 379, 381, 392, 508 significant wave period 33, 38, 39, 43, 44, 70, 91, 102, 103, 104, 107, 114, 118, 120 signs or notices 520, 525, 527 siltation 150, 151, 152, 153, 359, 390, 497 similarity number 420, 421, 423 simplified Bishop method 277, 278, 315, 316, 367 simplified method 311, 312, 315, 327, 413, 418, 470, 471, 473, 475 single anchor leg mooring 142 single pile 285, 290, 291, 292, 293, 295, 465, 477 single pile structure 523 single-buoy mooring 351, 352 sinker and anchor chain type mooring buoys 502, 503 sinker type mooring buoy 502 sinking currents 138 Skewness parameter 155 skirt guard 520, 525 slag 232, 238, 239, 548 slenderness ratio 284, 288 slice method 315, 335, 336, 340 sliding 107, 108, 110, 112, 130, 155, 182, 194, 200, 204, 278, 329, 331, 332, 365, 366, 367, 368, 375, 409, 410, 412, 431, 448, 454, 458, 495, 496, 518, 556 sliding resistance 363, 368 sliding resistance force 368 sliding stability 365, 366 slip failure 279, 314, 315, 316, 317, 318, 329, 369, 569 slip surface 277, 278, 314, 316, 317, 477 slip-bars 546, 547 slipway 249, 401, 407, 497, 498, 499, 500, 501, 520, 589, 591 slit 89, 91, 111, 124, 267, 268, 269, 376, 380, 387, 508, 509 slit caisson 267, 268, 376, 380, 387 slope failure 314 slope gradient 106, 107, 114, 161, 232, 363, 364, 381 slope stability analysis 314, 317 slope-shaped breakwater 376 sloping breakwaters 125, 360, 363, 369, 370, 372 sloping caisson breakwaters 377 sloping-top caisson 104, 106, 376, 380, 381, 382 sluices 357 slump 225, 226 SMAC-type strong motion seismograph 189 small amplitude wave theory 33, 34, 35, 74, 120 small craft basins 150, 345, 355 small craft quays 401 smear 320 snow load 28, 207, 208, 410, 586, 593 social function of the facilities 8 soft clayey ground 310, 312, 377 softness factor 16, 17 soil improvement 175, 199, 273, 318, 319, 320, 322, 323, 338, 363, 369, 377, 398, 429, 512 soil improvement material 238 soil improvement methods 175, 273, 318, 369 soil skeleton 310, 312 solidified coal ash 333, 334 solidified geotechnical materials 333, 334, 335 solitary wave 36, 80, 98 sorting coefficient 155 sorting facilities for hazardous cargo 575 sorting facilities for marine products 575 spatially-distributed load 310 specific gravity 24, 112, 117, 140, 218, 231, 235 spectral methods 42, 43, 46 splash zone 216, 272 squeezed-out failure 276

SRC structure 270 stability 30, 32, 40, 49, 102, 104, 106, 107, 111, 112, 118, 125, 127, 141, 168, 173, 174, 178, 182, 183, 184, 193, 204, 205, 208, 210, 227, 230, 231, 242, 243, 258, 261, 273, 314, 316, 317, 318, 319, 323, 324, 325, 326, 327, 328, 329, 331, 333, 336, 339, 340, 341, 343, 351, 359, 363, 364, 365, 366, 369, 370, 371, 373, 374, 375, 377, 378, 379, 380, 381, 382, 384, 385, 386, 391, 393, 396, 397, 398, 405, 408, 409, 410, 411, 412, 415, 416, 418, 420, 425, 426, 427, 431, 434, 435, 437, 438, 443, 448, 449, 452, 454, 455, 456, 458, 459, 460, 461, 462, 464, 477, 484, 488, 489, 492, 495, 496, 499, 501, 502, 507, 508, 518, 519, 533, 551, 569, 570, 584, 586, 593, 594, 595, 596 stability against bearing capacity 366 stability against overturning 365, 366, 452 stability against sliding 329, 365, 448, 518 stability analysis 278, 314, 315, 316, 317, 318, 333, 340, 369, 370, 411, 425, 477 stability during earthuakes 365 stability number 112, 113, 114, 117, 118, 119 stability of slope 273, 314, 318, 326, 431 stabilized body 237, 323, 324, 325, 326 stabilized soil 232, 322, 323, 324, 325 stabilizer content 329, 330 stainless steel 8, 234, 238 stairways and ladders 520, 527 standard concrete strength 224 standard intensity of illumination 531, 532, 535 standard laboratory consolidation test 171 standard penetration test 167, 168, 174, 175, 176, 177, 195, 196, 296 Standard Specifications of Concrete 221 standard spectrum 93 standard tensile strength 224 standing wave force 100, 101 standing wave height 74, 110, 132 standing waves 72, 74, 81, 103, 104, 106, 122, 123, 124, 132, 162, 163, 166 static bearing capacity formulas 284, 285, 286, 292 static load 28, 124, 180, 207, 208, 229, 285, 308, 386, 462, 463, 478, 488, 489 statistical properties 37, 103 steel cellular-bulkhead type dolphins 495 steel manufacture slag 238, 239 steel piles 212, 213, 287, 288, 290, 293, 300, 305, 306, 307, 308, 383, 384, 460, 468, 478 steel plate cellular-bulkhead quaywall 407, 452, 453, 456, 457, 459, 496 steel reinforcement 222, 223, 224, 225, 226, 227 steel sheet pile cellular-bulkhead quaywall 407, 436, 437, 444, 449, 453, 457, 486, 496, 517 steel sheet pile quaywall 415 steel sheet piles 212 steel shell type 568 Steel Ship Regulations 26 steel slag 232, 238, 239 step type beach 155 stepped seawall 89 Stewart 149, 153 stiffeners 459, 529 still water level 34, 37, 101, 103, 105, 108, 113, 116, 117, 120, 122, 123, 124, 132, 205, 206, 243, 365, 373, 379, 381, 382, 462, 480, 487, 488 stirrups 260 Stokes wave 36 Stokes’ third order wave theory 123 stone 91, 100, 112, 113, 114, 116, 117, 118, 119, 125, 136, 140, -I-13-


207, 226, 228, 229, 230, 231, 232, 240, 243, 278, 279, 326, 363, 367, 368, 369, 371, 372, 373, 391, 406, 412, 413, 464, 465, 472, 556 stopping action or radiation stress 149 storm conditions 32, 410 storm surge 28, 29, 30, 127, 128, 129, 130, 132, 357, 362, 367, 373, 374, 396, 398, 497, 573 storm surge protection breakwater 373 storm tide 128, 129, 130, 362, 373, 398 storm waves 32, 99, 110, 154, 278, 358, 362, 497, 508, 592 straight asphalt 229, 230 strain dependency 170, 327 strain level 170, 178, 191, 310, 327 stream function 36, 98 stress concentration coefficient 340 stress distribution method 340 stress in soil mass 310, 311 stress intensity in piles 384 stress reduction coefficient 340, 341 stress sharing ratio 340, 342 structural analysis factor 221, 222 structural members 30, 102, 111, 119, 211, 223, 225, 226, 233, 257, 258, 267, 268, 377, 378, 380, 386, 389, 414, 470, 523, 596 structural steel 212, 213, 215 structure factor 221, 222 structure with steel plates 270 structured types 182, 407 subbase course material 238 submerged breakwater 91, 113, 140 submerged dykes 151, 153 submerged members 100, 121, 138, 140, 144 submerged unit weight 187, 201, 205, 274, 275, 277, 289, 315, 316, 317, 328, 332, 418 submerged upright breakwater 72 submerged zone 217 subsoil 7, 108, 167, 169, 171, 174, 176, 180, 182, 183, 184, 185, 186, 187, 189, 192, 195, 196, 198, 199, 235, 275, 304, 310, 311, 312, 314, 318, 319, 322, 323, 324, 326, 328, 329, 330, 331, 332, 334, 335, 336, 337, 338, 339, 340, 341, 342, 343, 396, 401, 417, 452, 455, 456, 458, 460, 464, 465, 469, 471, 472, 496, 556 subsoil reaction 455, 556 super-large vessel 2 superstructure 113, 124, 125, 221, 222, 223, 226, 227, 256, 260, 268, 276, 291, 293, 294, 322, 363, 365, 372, 455, 461, 462, 463, 464, 465, 469, 470, 471, 474, 475, 476, 477, 478, 479, 480, 481, 483, 485, 486, 513, 514, 515, 516, 518, 571 superstructure of open-type wharf 221, 223, 226, 464, 481 supply rate 337 surcharge 7, 184, 201, 204, 207, 232, 244, 248, 250, 251, 252, 253, 257, 258, 260, 262, 263, 266, 268, 269, 276, 277, 314, 315, 316, 318, 319, 326, 327, 335, 365, 408, 410, 411, 418, 427, 431, 433, 434, 437, 438, 439, 442, 450, 454, 456, 472, 474, 478, 488, 586, 595, 596 surf beat 91, 92, 93, 111, 140 surf similarity parameter 114 surf zone 99, 103, 138, 141, 156, 159, 162, 166, 390, 391 surface covering 328 surface elevation 34, 119, 132 surface layer movement 154, 157 surface roughness 113, 120 surging 93, 146, 354, 521 surging breaker 114 suspended sediment 138, 155, 156, 159, 372, 391, 392 swash zone 156, 160 swaying 146, 354

swell 39, 40, 43, 46, 47, 94, 98, 353 swinging mooring 351 synthetic fiber ropes 26

T T-shaped sheet piles 450 Takahashi 419 Takahashi et al. 104, 107, 108, 109, 111, 114, 116, 417, 419, 420, 421 Takayama’s method 69 Talbot’s formula 28 Tanaka 156, 162, 163, 166 tankers 10, 12, 15, 17, 18, 19, 21, 22, 145, 354, 404, 495 target vessel 9, 10, 16, 17, 21, 23, 346, 347, 348, 351, 352, 353, 402, 405, 494, 502, 523, 582 taut mooring 142, 143 Technical Manual for Premixing Method 329 Technical Manual for the Deep Mixing Method in Land Construction Works 323 Technical Manual for the Deep Mixing Method in Marine Construction Works 323 Technical Manual for the Lightweight Treated Soil Method in Ports, Harbors and Airports 326 temporary structures 7, 217 tensile strength 222, 224, 235, 236, 293, 450, 451, 523 tensile yield strength 222 tension of sheet pile 449, 450 tension of tie rod 424, 425, 428, 429, 518 terminology 2, 154, 155, 242, 323 Terzaghi 170, 180 Terzaghi’s consolidation theory 310, 312 theory of consolidation 170, 171, 173, 310 theory of elasticity 310 thermal characteristics 524 thickness of a concrete lid 370 thickness of steel plate 455, 458 three-dimensional experiments 140 threshold depth of sediment movement 154, 156, 157, 166 threshold wave heights for cargo handling 353 tidal currents 16, 24, 26, 138, 140, 144, 148, 242, 346, 347, 348, 350, 351, 352, 353, 355, 357, 358, 359, 386, 388, 396, 487, 494, 498, 503, 504, 590, 592, 594 tidal level 100, 359, 362, 365, 367, 375, 380, 396, 398, 405 tidal zone 216, 217, 272 tie rod setting point 416, 418, 419, 420, 421, 422, 424, 425, 426, 427, 428, 429, 432, 433 tie-bars 545, 546, 547 timber 210, 233, 305, 345, 354, 372, 373, 462, 522, 574, 576 timber sorting areas 574 timber sorting ponds 372, 574 timber storage ponds 372 timber-handling facilities 372, 373 time factor 171, 172, 310, 321, 322 time-dependent mild-slope irregular wave equations 69 time-dependent properties 312 Tokyo Peil 128 tolerable damage level 475 top displacement 448, 449, 512 torsional moment 386 total length of sheet pile 420 total movement 154, 157 total stability 329 TP 128 tractive force 16, 22, 25, 26, 27, 373, 374, 392, 415, 417, 424, 425, 428, 433, 463, 480, 485, 490, 495, 502, 503, 504, 505, 506, 507, 521, 522, 584, 585, 595 -I-14-

INDEX

V

tractive forces of vessels 25, 408, 410, 424, 428, 503, 521, 522 traffic loads 584 training jetties 154, 161, 357, 388, 392, 393 transformations of waves 41, 49, 358, 367 transitional embedded length 420 transmission 80, 90, 91, 94, 271, 358, 363, 364, 377, 379, 380, 381, 382, 383, 384, 479, 516, 546 transmitted waves 90, 91, 94, 364, 378, 384, 591 transverse contraction joint 546, 547 transverse expansion joint 546, 547 transverse waves 95, 96, 97, 98, 592 trapezoidal caisson 104, 124, 376 trapped air 122 treated soil 323, 326, 327, 328, 329, 330, 332, 333, 334, 343 treated subsoil 329, 330, 331, 332, 333, 334 trenches 151, 152 triaxial CD test 174, 175 trip distribution 560 trip generation and attraction 560 Tschebotarioff’s model 420 tsunami 50, 127, 130, 131, 132, 136, 140, 373 tsunami barrier 130 tsunami period 130, 131 tsunami protection breakwater 140, 141, 373, 374, 375 tsunami runup 130 tsunami simulation 132 tsunami wave force 132 tsunami wave height 130, 131, 132 Tsuruya 153 tunnels 567, 570, 572 turbulent eddy viscosity 138 turning 346, 350, 352, 388, 559, 562, 564, 591 turning basin 352 turning diameters 346 two-layer placement 117 type of the breakwater 222 typhoon 28, 41, 44, 128, 130, 153, 463

V-shaped rubber fender 524 vacuum consolidation method 320 values proposed by Terzaghi 420 Van der Meer’s Formula 114 vane shear test 168, 175, 176, 177, 180 variable load 221, 222, 223, 248 Vasco Costa 523, 539 vehicle load 207, 209, 463, 586 vehicle ramp 526, 527, 585, 586 velocity coefficient 148 velocity pressure 30 velocity-dependent characteristics 524 Veritus Rule 386 vertical breakwater 70, 161, 364 vertical coefficient of consolidation 311 vertical drain method 318, 319, 341 vertical force 282, 386, 410, 411, 412, 433, 437, 439, 445, 447, 451, 463, 485, 491, 492, 495, 507, 513, 518 vertical loading tests 286, 287 vertical pile anchorage 415, 426, 428 vertical-slit type seawall 89 vertical-slit wall caisson 111 very soft clay 173, 175 vessel berthing force 221 vessel pulling force 221 vibratory pile driving method (vibrohammer method) 286 vinylon ropes 26 virtual fixed point method based on Chang’s method 465 virtual fixed points 466, 475 virtual ground surface 432, 465, 466, 468, 471, 472, 481 virtual mass factor 16, 21 visco-elasto-plasticity 311, 312 vortices 134, 139, 140, 156, 159, 374

W Wagner’s theory 123 wale 417, 425, 429 wall body 408, 409, 410, 411, 438, 439, 443, 449, 453, 454, 457, 458, 495, 518 wall displacement 177, 443, 444, 445, 447 wall type improvement 322, 323, 324, 326 wall-type structure 100 warehouse 207, 208, 436, 532, 540, 576 warning signs 396 waste disposal sites 326 waste water treatment plant 396 water absorption rate 231 water exchangeability 376 water frontage amenity 84, 357, 358, 396, 531 water level 36, 41, 74, 75, 79, 80, 84, 91, 92, 93, 100, 111, 120, 123, 124, 127, 128, 129, 130, 131, 132, 133, 138, 148, 149, 154, 202, 205, 216, 217, 218, 220, 245, 246, 250, 281, 289, 327, 332, 350, 357, 362, 363, 364, 365, 373, 389, 396, 397, 398, 405, 417, 418, 450, 456, 570, 597 water particle acceleration 34, 119, 121 water particle velocity 34, 35, 36, 119, 120, 121, 126, 159 water quality 8, 219, 357, 358, 397 water supply facilities 520, 525, 526, 577, 589 water-cement ratio 225, 226 waterproofness 228, 397 wave actions 41, 102, 109, 112, 366, 377, 383, 386, 393, 461, 488, 490, 504, 523 wave breaking 75, 78, 79, 80, 81, 91, 98, 99, 103, 107, 111, 114, 116, 118, 119, 132, 141, 150, 159, 592 wave breaking point 75, 82, 83, 91, 107

U U.S. Army Corps of Engineers 159 ultimate axial bearing capacity 284, 285, 286 ultimate bearing capacity of the pile 286, 477 ultimate limit state 221, 222, 243, 245, 246, 248, 254, 256, 261 ultimate load 285, 286, 291, 295 ultrasonic propagation test 327 unconfined compressive strength 168, 174, 175, 177, 185, 189, 231, 279, 286, 287, 290, 300, 317, 324, 325, 327, 330, 331, 334 underwater concrete 210, 227 undisturbed samples 170, 174, 175, 179, 198, 287, 310 undrained shear strength 170, 173, 174, 175, 176, 177, 178, 180, 275, 276, 277, 315, 316, 319, 340 unevenly distributed load 208 uniformity coefficient 169, 170, 195, 240, 328 updrift 390, 391, 392, 393 uplift pressure 102, 104, 105, 109, 221, 244, 247, 252, 253, 267, 268, 371, 379, 382 upright breakwater 72, 91, 125, 360, 361, 362, 363, 364, 365, 369, 370, 376, 378, 379, 381, 387 upright wall 72, 81, 100, 101, 102, 103, 104, 105, 106, 107, 108, 109, 110, 111, 114, 118, 132, 263, 365, 366 upright wave-absorbing caisson 111, 267, 378, 508 upright wave-absorbing structure 72 upwelling currents 138 utilization factor 536, 537 UU condition 173, 174 -I-15-


wave chamber 111, 112, 379, 380 wave crest 36, 37, 83, 95, 98, 101, 106, 116, 120, 123, 149, 245, 248, 250, 594, 595 wave development 42, 43, 44, 46, 47 wave diffraction 33, 52, 71, 74, 144, 359 wave direction 39, 41, 47, 49, 50, 51, 68, 70, 71, 72, 74, 94, 96, 97, 100, 101, 102, 104, 106, 108, 113, 116, 118, 119, 132, 157, 165, 353, 494, 595 wave energy balance equation 51 wave energy flux 49, 157, 159, 165 wave force 16, 22, 23, 27, 32, 36, 41, 70, 74, 75, 98, 99, 100, 101, 102, 103, 104, 105, 106, 107, 108, 109, 110, 111, 112, 116, 119, 120, 121, 122, 124, 125, 126, 132, 139, 140, 145, 180, 221, 222, 244, 247, 250, 251, 252, 261, 262, 263, 267, 268, 269, 277, 314, 315, 316, 363, 364, 365, 366, 367, 368, 369, 370, 373, 374, 375, 377, 378, 379, 380, 381, 382, 384, 386, 396, 486, 487, 488, 490, 503, 504, 518, 522, 584, 594 wave force due to a broken wave 100 wave force under wave crest 101 wave force under wave trough 105 wave height 22, 32, 33, 34, 35, 36, 37, 38, 39, 40, 41, 42, 43, 44, 46, 47, 48, 49, 52, 69, 70, 71, 72, 73, 74, 75, 76, 77, 78, 79, 80, 81, 90, 91, 93, 94, 95, 96, 97, 98, 99, 100, 101, 102, 103, 104, 105, 106, 107, 108, 109, 110, 111, 112, 113, 114, 117, 118, 120, 123, 124, 130, 131, 132, 133, 146, 149, 150, 157, 159, 160, 162, 165, 222, 235, 347, 350, 353, 357, 359, 362, 363, 365, 367, 368, 369, 370, 374, 377, 379, 381, 382, 383, 392, 504, 508, 571, 591, 595 wave hindcasting 32, 41, 42, 47, 69 wave observation 32, 33, 37, 47, 94, 159 wave overtopping 81, 83, 84, 90, 99, 363, 378, 391, 396, 398, 518 wave pressure 101, 102, 103, 104, 105, 106, 107, 108, 133, 222, 230, 262, 269, 279, 315, 317, 364, 374, 378, 379, 381, 382, 594, 595 wave pressure modification factors 101, 109 wave reflection 69, 70, 71, 104, 110, 123, 135, 161, 162, 357, 376, 377, 378, 379, 380, 382, 383, 386, 518 wave refraction 33, 49, 69, 150, 359 wave runup 80, 83, 111, 117, 135, 156, 161, 166, 363, 364, 497 wave setup 79, 80, 91, 111, 130 wave shoaling 36, 49, 69, 74, 75 wave spectrum 32, 38, 39, 40, 42, 43, 51, 98, 99, 113, 146 wave steepness 35, 36, 39, 80, 91, 95, 107, 113, 114, 161 wave transformation 32, 33, 41, 71, 99, 374 wave transmission 90, 94, 358, 377, 380, 381, 383, 384 wave trough 105, 106, 245, 246, 248, 250, 263, 368, 410, 594, 595 wave velocity 34, 35, 49, 98, 180 wave-absorbing block 377, 378, 379, 508 wave-absorbing seawall with concrete block mound 89 wave-absorbing work 71, 72, 83, 106, 109 wave-absorption performance 379, 380 wave-dissipating blocks 109, 250, 364, 365, 369, 370, 376, 377, 378, 381, 382 wave-dissipating concrete blocks 72, 83, 84, 89, 91, 100, 103, 106, 109, 113, 114, 116, 117, 125, 253, 360, 362, 364, 365, 372, 381 wave-drift force 143, 144, 145, 146 wave-exciting force 143, 144, 145, 146 wave-making resistance 143, 145, 504 wavelength 32, 33, 34, 35, 36, 50, 69, 71, 72, 74, 81, 91, 93, 95, 96, 101, 102, 103, 105, 107, 108, 109, 111, 114, 118, 119, 121, 123, 133, 134, 144, 145, 149, 150, 157, 159, 160, 163, 349, 350, 382, 383 wavelength in deepwater 91 weather charts 42 Weibull distribution 19, 47, 48 weight of filling 248, 249, 250, 251, 252, 253, 412, 441, 442, 453,

454, 455, 456 welded joints 272 welded zone 214, 215 well 226, 231, 234, 238, 256, 259, 260, 268, 277, 279, 283, 290, 292, 295, 296, 307, 308 wet unit weight 180, 201, 289, 328 wheelchairs 396, 578, 579 wide mound berm 105 width of flexural cracks 223 width of sediment movement zone 166 Wilson’s method 44, 45 wind direction 29, 30, 31, 42, 44, 129, 130, 143, 494, 591 wind drag coefficient 143 wind drag force 143, 144, 146 wind drift currents 138 wind duration 42, 43, 46 wind force coefficients 23 wind load 23, 30, 386, 463, 485, 486, 529, 530, 578, 585, 586 wind pressure 7, 26, 27, 28, 30, 143, 144, 221, 365, 369, 370, 463, 495, 574, 584 wind pressure coefficient 30, 143, 144 wind setup 128, 129, 130 wind tunnel experiments 30, 143 wind velocity 23, 24, 26, 28, 29, 30, 32, 42, 43, 46, 89, 129, 130, 138, 143, 385 wind waves 39, 40, 43, 44, 46, 50, 94, 98, 131, 508, 592 wind-blown sand 154, 390, 394 wind-driven drift currents 138 windbreaks 394 Wooden Structure Design Standard 233 woven cloth 235

Y Yamaguchi and Tsuchiya’s model 43 yards for dangerous cargo 576 yawing 20, 146, 349, 350, 354 yield load 285, 286, 290, 295 yield strength 213, 222, 224, 476 Yu 149, 153

Z zero-upcrossing method 37, 40, 131 zeroth moment of the wave spectrum 39

-I-16-

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