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PROCEEDINGS OF THE INTERNATIONAL SYMPOSIUM ON SLOPE STABILITY ENGINEERING - IS-SHIKOKU’99/MATSUYAMA/SHIKOKU/JAPAN/81 1 NOVEMBER 1999
Edited by
Norio Yagi Ehime Universiq, Japan
Takuo Yamagarni & Jing-Cai Jiang University of Tokushima, Japan
VOLUME 2
A.A. BALKEMA/ROTTERDAM/BROOKFIELD/ 1999
The texts of the various papers in this volume were set individually by typists under the supervision of each of the uuthors concerned.
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0 1999 A.A. Balkema, Rotterdam Printed in the Netherlands
Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Table of contents
6 Design strength parameters Undrained flow and instability of anisotropically consolidated sand YTsukamoto, K. Ishihuru, S. Nukayumu & I:Nosuka
675
Model test on granular soil slope and determination of strength parameters under low confining stresses near slope surface H. Matsuoku, S. H. Liu & TOhushi
68 1
Determination of shear strength parameters of unsaturated sedimentary residual soils for slope stability analyses S. Muriuppun, E H.Ali & L. T Huut
687
The characteristics of landslides caused by the hydrothermal metamorphic clay H.Yumashitu, M. Suga, H. Fujitu, K.Yokotu & R.Yatuhe
693
Influence of clay minerals on strength characteristics of landslide clay in Mikabu T.Ishii, R. Yutuhe,A? Yugi & K. Yokitu
697
Strength of landslide clay from mineralogical point of view NYugi, R.Yutuhe, K.Yokotu & A? P Bhundury
70 1
Role of soil composition on collapsible behavior of natural and stabilized slopes V R. Ouhudi
705
Deformation characteristics of a compacted clay in wetting tests under isotropic and triaxial stress state S. Kuto 6;K. Kuwai
709
Development of an automatic cyclic direct shear test apparatus for landslide slope stability analysis M. Okawuru, T.Mituchi & M. Tunudu
715
Strength and deformation characteristics of clay subjected to pore water pressure increment TUmezaki, M.Suzuki & TYumamoto
72 1
Parameters for curvilineared residual strength envelope S Giho & S Nukurnui-u
727
Pore water pressure loading tests of a clay S. Ohtsuku, Y Miyatu & H. Toyotu
731
V
Shear behavior of clay subjected to change of normal stress M.Suzuki, TUmezaki & TYamamoto
735
A simple model to predict pore water pressures during shearing along undulating surfaces D.J. Petley & PTaylor
741
Modelling rapid shearing of cohesive soils along undulating shear surfaces D.J. Petley & PTaylor
745
Apparent cohesion of unsaturated soils as correlated with suction f/: Huang & K. Ishihara
75 1
Unconfined compression shear strength of an unsaturated silty soil subjected to high total suctions ?:Nishimura & D.G. Fredlund
757
Shear strength mobilization in shear box test under constant volume I. Kohayushi, A. Iizuku, H. Ohta & M. Hirata
763
Undrained shear strength of unsaturated compacted clays VSivakumur & I. G.Doran
769
Landslide at Malakasa, Greece: Investigation, analysis, remedial works R.J. Chandler & S.Schina
775
Method for determining design strength parameters for slope stability analysis I: Mitachi, M. Okuwara & I:Kawaguchi
781
Evaluation of the shear strength for stability analysis of a heavily weathered tertiaiy rock K. Tsuji, K. Suzuki & H. Hanzawa
787
Effect of degradation on the strength of rock A. Kobuynshi, K. Yamarnoto & K. Fujii
793
Some considerations of Patton model on rock joint shear strength M. Doi & S. Ohtsuku
799
Behavior of jointed model material under biaxial compression A. K.Tyugi, K. S. Rao & A. S.Gupta
805
7 Slope stability oflar,dfills and waste materials Stability of slopes of hydraulic-fill dams A. Zh. Zhusupbekov, A. S Zhakulin & M. R. Nurguzhin
811
Stability of embankment dams based on minimum-experience of safety factor Morii, K. Shimada & ?:Hasegawa
817
Stability of embankment using foam composite lightweight soil f/: Watunuhe & I: Kaino
823
Slope stability of embankment model composed of municipal bottom ash: Centrifuge model tests and FDM analysis K.Gotoh, M.Yamanaku, XIkuta & TOgawa
827
VI
Comparison of deformation of a fill with results from a new elastoplastic method 7:Hurudu, A. Mochizuki & I: Kanedu
83 1
Evaluation of slope stability incorporating pre-compression characteristics of cohesive soils M. Yanzaguchi, K. Nurita & Y Ohne
837
Earth pressure acting on the side of core block in high embankment K. Nomoto, I: Sugirnoto & T Fujiwuru
841
Case study of a liquefiable mine tailing sand deposit WWehr, I. Herle, I? Kudellu & G.Gudehus
847
Bilinear model for stability calculation of domestic waste landfills G.Ziehmann
853
The stabilization of frozen technogenic dumps VLGrehenets, S.NTitkov, A.G.-o. Kerimov & VM.Anishin
859
Stability of MSW mass: Use of an improved limit equilibrium analysis A. Bouazzu & I. B. Donuld
863
Stability of bentonite wall by the unified method of molecular dynamics and homogenization analysis f/: Ichikuwu, K. Kuwumura, M. Nukano, TSeiki & TNuttuvut
869
8 Stabilization and remedial works Model tests of a new deep pile system for landslide prevention at Kamenose landslide area K. Nishiyama, S.Tochirnoto, H. Fujitu, S. Kinoshita, S. Sukajo, M. Ohno, K. Ugui & M. Kimura
877
Stability of slope reinforced with piles FCai & K. Ugai
883
Numerical study of landslide of bridge abutment in Surabaya, Indonesia VTandjiria
889
Application of FEM as a design method for slope stability and landslide prevention pile work M.Gotoh & YOhnishi
895
Design and constructional aspects of an anchored slope and gabion revetment system M. H. Kubir & A. M.Humid
90 1
Evaluation of pull-out capacity of repeat-grouting type ground anchor by in-situ and laboratoiy tests H. Wadu, H. Ochiai,K.Omine & Y Muecla
907
Design and obseivation of the prevention works for crystalline schist slope
913
N.Shintani, K. Kawuhuru, A. Ueclu, K. O h & TYamamoto Case study on slips in soft laterite cut-slopes on BG rail link in Southern Peninsular India VK.Jain & K. Keshuv
919
Hydrodynamic seeding with the use of sewage sludge and fly-ash for slope protection M.Glaiewski & J. Kulotka
925
Investigation and stabilization of a sliding hillside J. Furkzs
93 1
VII
Stability reinforcement of the old embankment sanitary landfills for remediation works E. Kodu
937
Stabilization and remedial works on some failed slopes along the East-West highway, Malaysia A.Jamuludin & A. N. Hussein
943
Landslide controlling measures at construction sites nearby King’s palace at Narendra Nagar D. Mukherjee, K. Kishor & 0.l? Yuduv
949
Reduction of land cutting effects by the application of lightweight embankments J. Nakano, H. Miki, H. Kohashi & A. Fujii
955
Relaxation effect in retaining wall on passive mode Erizul, T.Sukai & S. Miyuuchi
959
Stabilization and geoenvironmental restoration of the main central channel in the Fucino plain, Italy - A case history G.Totani, I? Monaco, M. Leopardi, A. Furroni & A. R. Spena
965
Slope stabilization in residual soils of Peru A. Carrillo-Gil & A. Currillo-Acevedo
97 1
Case study of a cut slope failure in diatom earth A.Yashimu, H. Shigematsu, S. Okuzono & M. Nishio
977
9 Stability of reinforced slopes Centrifuge model testing of reinforced soil slopes in the perspective of Kanto Loam G.Pokharel, A. Fujii & H. Miki
985
Dynamic behavior of vertical geogrid-reinforced soil during earthquake A.Takahushi, J. Takemuru & J. Izawa
99 1
Model tests on some geosynthetics-reinforced steep earth fills XTanubushi, 7:Hirui, J. Noshimura, K.Yusuharu & K. Suyama
997
Field behavior of a reinforced steep slope with a cohesive residual soi backfill A. Kasa, F: H.Ali & Z. Chik
1003
Full-scale model test on deformation of reinforced steep slopes I: Naguyoshi, S Tuyuma, K Ogata & M. Tadu
1009
Relation between wall displacement and optimum amount of reinforc ments on the reinforced retaining wall K. Okabuyushi & M. Kawumura
1015
Stability analysis of reinforced slopes using a strain-based FEM T.Mutsui, K. C.Sun & A. Porbuhu
1021
Numerical analysis on the stability of GHD-reinforced clay embankment M. Kumon, M. Mimuru, N Tukeo & rAkai
1027
New design method of composite fabrics - Reinforced earth fill XTunubushi, iV Wukudu, K. Suyama, K.Yusuharu, T.Hirai & J. Nishimuru
1033
Vlll
Design method for steel grid reinforced earth structure considering bearing resistance TMatsui, Y Nuheshima, S.G.Zhou & NOgawa
1039
A promising approach for progressive failure analysis of reinforced slopes TYamagami, S.Yamabe, J.-C.Jiung & YA. Khan
1043
3-D stability analyses for asymmetrical and heterogeneous nailed slopes C C. Huang, C.C.Tsai & M.Tateyama
1049
Numerical analysis of reinforced soil slopes under working stress conditions B. ir:Dantas & M. Ehrlich
1055
Design method of vertical reinforced slopes under rotational failure mechanism X. Q.Yang, S.X. He & Z. D. Liu
1061
Reinforcement mechanism in soil nailing for stabilization of steep slopes 7:Nishigata & K. Nishida
1065
The study of direct shear tests of woven geotextiles with granular soils M. Matys, TAyele & S. Hric
1071
10 Probabilistic slope stability Localized probabilistic site characterization in geotechnical engineering S. Pumjan & D. S. Young
1079
A localized probabilistic approach for slope stability analysis D. S. Young & S. Pumjun
1085
Probabilistic analysis of structured rock/ soil slopes - Several methods compared D.Xu & R.Chowdhury
1089
Reliability analysis and risk evaluation of the slopes of open pit mine Q.Yung, J. Jiao, M. Luan & D. Shi
1095
Risk evaluation for slope failure based on geographical information data I!Kitazono, A. Suzuki, N Nakusone & TTeruzono
1101
Gray system evaluation for slope stability engineering H.-CWU,T.Bao, X.-B.Zhung & X.Hu
1105
Statistical variability of ring shear test results on a shea-zone in London Clay E. N Bromhead, A.J. Harris & M-L. Ihsen
1109
Overall stability of anchored retaining walls with the probabilistic method L. Belabed
1115
11 Landslide investigations Methodological study of judgement on landslide occurrence M.-B.Su, L.-CChun & G.-S.Lee
1123
The retrogressive slide at Nipigon River, Ontario, Canada K.7:Law & C.F:Lee
1129
IX
Simplified model for estimating a scale of sliding debris M. Fukudu & S. Suwu
1135
Landslide prediction using nonlinear dynamics model based on state variable friction law K.T.Chuu
1139
Characteristic weathering profiles as basic causes of shallow landslides M. Chigiru & E. Ito
1145
Long-term movements of an earthflow in tectonised clay shales L. Picurelli, C. Russo & A. Mundolini
1151
Characteristics of groundwater quality in fracture zone landslides at Shikoku area
1159
i? Nishimuru, R. Yutube,h? Yugi, K. Yokotu & I: Shibutu Use of H,O(+) for landslide investigations and mapping Ude S.Juyuwurdenu, E. Izuwa & K. Wutunube
1165
The mechanism of creep movement caused by landslide activity and underground erosion in crystalline schist, Zentoku, Shikoku, Japan G. Furuyu, K. Sussu, H. Hiuru & H. Fukuoku
1169
Mechanism of large-scale collapse at Tue Valley in the Shikoku mountainous region, Japan EOchiui, H.Sokobiki, TNoro & S. Nukuyuma
1175
Causes and mechanisms of slope instability in Dessie town, Ethiopia L.Ayulew & A. Vernier
1181
Structural deterioration of residual soils and the effect on landslides J. Suhrez
1187
Study of a huge block slide with relevance to failure mechanism I. Lazunyi, I. Kabai & B.Vizi
1193
Landslide clay behavior and countermeasures works at the fractured zone of Median Tectonic Line R.Yutube,NYiqi, K.Yokotu & N 19 Bhandary
1199
Geological and soil mechanical study of Sawatari landslide in Ehime H. Kono, M. Tuni, R. Yutube,h? Yugi & K. Yokota
1203
The general characteristics of landslide along the Median Tectonic Line due to the road construction k:Momiyumu, K. Kumuno, M. Tunuku & I: Ishii
1207
An investigation on the stability of two adjacent slope movements G.Gotturdi & L. Tonni
1211
Evaluation of stream-like landslide activity based on the monitoring results L. Petro, l? Wugner & E. PoluEinova
1217
Snow induced landslides in Japan I:It0
1223
Physical properties of clay from landslides in large fracture zones N. Ogitu, X Kito, ?:Kimizu & R.Yutube
1229
X
Investigation of landslide damage in Korea, 1998 D. Park, K. Oh & B. Park
1233
Monitoring of the Vallcebre landslide, Eastern Pyrenees, Spain J. Corominas, J. Moya, A. Ledesma, J. Rius,J.A.Gili & A. Lloret
1239
12 Landslide inventory, landslide hazard zonation and rockfall Disaster prevention and sustainable development in Central America S. Mora
1247
Preliminary landslide hazard mapping along a hill road in westein Nepal B. P Mainalee, N. Morishima & H. Fujimura
1253
Hazard evaluation of landslide in Iran G. R. Lushkaripour
1259
Zonation of areas susceptible to rain-induced embankment failure in Japan railways K. Okacla, 1:Sugiyama, H. Muraishi & 1:Noguchi
1263
An estimation of slope failures based on erosion front and weathering front H. Inagnki & TYunohara
1269
Typical case study on destabilization and genetic mechanism of urban slopes in China K Liu, E;: Niu & Z. Cheng
1275
Estimation of the slope failure using remote sensing data S. Shima & H. Yoshikuni
1281
Application of hazard and iisk maps for highway slopes management and maintenance KA. 0.Fiener & E;: H.Ali
1287
Application of hazard and risk mapping to a mountainous highway in Malaysia
1291
A.Jamaluclin, Z.Mucla, S.Alias & N M.Yusof
A landslide risk assessment in a hydropower plant area D. Paunescu & D. Deacu
1297
Applications of quantitative landslide risk assessment in Hong Kong C K. M. Wong & C.K.7:Lee
1303
Landslide risk assessment - Development of a hazard-consequence approach C K. KO,P Flentje & R. Chowdhury
1309
Data-bases and the management of landslides R. M. Faure
1317
Seismicity in the development of the geological process in the Republic of Tajikistan S. Vinnichenko
1331
Evaluating rockfall hazard from carbonate slopes in the Sele Valley, Southern Italy M. Pcrrise
1337
Effect of soil slope gradient on motion of rockfall S. Kawahara & 1:Muro
1343
XI
Study of accidents caused by rockfall in Kochi Prefecture TUshiro, YMatsumoto,NAkesaku & N.Yagi
1349
The coefficient of restitution for boulders falling onto soil slopes with various values of dry density and water content K.TChuu,J.J.Wu, R.H.CWong & C.F:Lee
1355
The May 5th 1998 landsliding event in Campania, Southern Italy: Inventory of slope movements in the Quindici area D. Calcaterra, M. Purise, B. Pulma & L. Pelella
1361
I3 Simulation and analysis of debrisjow A proposed methodology for rock avalanche analysis R.Couture, S.G. Evans, J. Locat, J. Hadjigeorgiou & PAntoine
1369
The Otari debris flow disaster occurred in December 1996 H. Kuwakami, H.Suwa, H.Marui, 0.Sato & K.Izurni
1379
Dimensional analysis of a flume design for laboratory debris flow simulation L.C.PChan & KTChuu
1385
Shear characteristics at the occurrence and motion of debris flow YYamashita,NYagi, R.Yatabe & K.Yokota
1391
Three-dimensional numerical modeling of muddy debris flows H Chen & C.F: Lee
1397
Mechanism of soil deformations during the displacernents of flow slides 0.VZerkul& V N Sokolov
1403
Author index
1409
6 Design strength parameters
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Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Undrained flow and instability of anisotropically consolidated sand YTsukamoto, K. Ishihara & S. Nakayama Department of Civil Engineering, Science University of Tokyo,Japan
Y. Nosaka Obayashi Corporation,Japan
ABSTRACT: Soils of interest for slope instability are inherently subjected to initial shear stress. In order to examine the characteristics of undrained flow behaviour of such soils, undrained triaxial compression tests are carried out on isotropically as well as anisotropically consolidated soil specimens. The degree of anisotropic The two soil materials are used in this study; Omigawa silty sand and consolidation is defined by K, (=cT~,~/cT~,~). Jamuna river silty sand. The steady state and phase transformation envelopes are found to be uniquely determined independentlyof the K, values, in the plot of effective mean stress p7against shear stress q, however, there are a series of peak stress state envelopes for different K, values. The steady state and phase transformation lines are examined in (e, p’) and (e, q) plots. The flow characteristics of anisotropically consolidated soil specimens are then examined with the initial state ratio r,. slope failures occurred at Omigawa area located at the foot of the terrain in Chiba, Japan, as shown in Figure 1, In order to examine the behaviour and instability of during the passage of Typhoon in 1971. The soil matesloping soil masses subject to rainfall, the principles of rial is a silty sand with non-plastic fines, and was resaturated soil mechanics were put into perspective in cently recovered fi-om one of the sites where rainfall the past literatures. A notable development was the induced slope failures took place. The official report of concept of drained initiation and undrained mobiliza- the disasters caused by this stormy rainfall was pubtion illustrated by Anderson & Sitar (1995), Zhu & lished by Chiba Prefecture (1972), and part of the laboAnderson (1998) and others, in which natural soil ratory study on this soil material was described by slopes, which usually exist in unsaturated conditions Tsukamoto et al. (1 998). The Jamuna is a might river and are subjected initially to deviatoric stresses, be- dividing Bangladesh into western and eastern land come saturated due to water infiltration of rainfall es- masses, originating from the Himalayas and flowing pecially at the potential sliding zones, and the mean into the Bay of Bengals. Bridging the Jamuna between effective stress hrther reduces mainly due to seepage Sirajganj and Bhuapur was a major civil engineering flow and the stress states of the soil masses critically challenge, allowing more links for transportation and approach the failure conditions, which is called a energy supply between the divided west and east, as drained initiation, whereby sufficiently greater strains shown in Figure 2, (Tappin et al. 1998). To implement are induced within the soil masses to cause undrained the river training of this braided river, the construction mobilization of the entire soil masses. In this study, of west and east guide bunds was planned and dredging undrained monotonic triaxial tests are carried out on of the work harbour basin and the reclamation of the anisotropically consolidated saturated soils, to exam- east guide bund started in October 1994. The conine the characteristics of flow and instability of sloping struction of the west guide bund resumed with dredging of nearby Jamuna river bed in 1-in-3.5 slope and soil masses. placement of the geotextile and fascine mattress on its slope in August 1995. However, a number of soil slips occurred successively in the dredged 1-in-3.5 slope, 2 EXPERIMENTAL, DETAILS and the gradient of the dredged slope had to be changed to 1-in-5/1-in-6 subsequently. The Jamuna 2.1 Soil materials river bed consists of loose to medium dense micaceous Two soil materials are used in this study, Omigawa sands. The soil material was recovered by the second silty sand and Jamuna river silty sand. A number of author from one of the sites at the west guide bund 1 INTRODUCTION
675
Table 1. Physical properties of soil materials. Omigawa sand Jamuna river sand Specific gravity ern,, %in
2.694 1.282 0.796
Atterberrr limits
NP
2.745 1.202 0.602 NP
Figure 1. Location of Omigawa area.
Figure 3. Grain size distributions of soil materials. In the (p’, q) diagrams of Figure 4(a) and Figure 5(a), where p’ = ((3 1’ + ci 3’)/2, q = (ci - ci3’)/2,ci 1,3’ = ci 1,3 - U and U is the excess pore water pressure, the consolidation processes are represented by the movement of the stress points from origin to points a and a’ for isotropic consolidation (K, = I)., and then to points b and b’ for anisotropic consolidation (K, < 1). 2.3 Undrainedcompression Upon completion of consolidation, the soil specimens are then subjected to undrained monotonic compressive straining. Figures 4(a) and 4(b) show the (p’, q) and (E,, q) diagrams, respectively, for the test results Figure 2. Location of Jamuna river bridge. on Omigawa sand which are anisotropically consolidated to the same K, value. In these diagrams, E, is the where the soil slips occurred. Table 1 summarizes the axial strain and e is the void ratio of the specimen physical properties, and Figure 3 shows the grain size achieved after consolidation. It is found that some of distributions of the two soil materials used in this the specimens experience peak stress states, then the study. shear stress q drops off to achieve quasi-steady states (states of phase transformation), and the shear stress q eventually begins to increase again to reach steady 2.2 Consolidation states. In other specimens, the shear stress q continues Soil specimens are prepared by the method of wet to increase in which the specimens experience the tamping (moist placement). The details of the soil sam- states of phase transformation to reach steady states. ple preparation methods are described by Ishihara Figures 5(a) and 5(b) show the test results for Jamuna (1993). They are then saturated and isotropically con- river sand. Noteworthy is that the states of phase solidated to designed confining stresses, 03’ (= ci 1’). transformation are not evident for this sand, which For a series of anisotropically consolidated undrained might be related to micaceous contents of this sand (ACU) compression tests, the axial stress ci 1’ is then consisting of flat-shaped aggregates, which claim a increased to achieve designed K, (= 0 3 , ’ / ci 1,’) values. weak horizontal resistance as foundation soils.
676
(b) (G, q) diagram Figure 4. ACU tests (Omigawa sand). 3 CHARACTERISTIC ENVELOPES
Four series of tests were carried out on Omigawa sand, with different K, values of 0.4, 0.5, 0.6 and 1. Figure 6 shows the (p7,q) diagram for the isotropically consolidated undrained (ICU) compression tests, (K, = 1). The peak state (P.S.) envelope can be defined on the test results which experience peaks in shear stress q. The phase transformation (P.T.) envelope and steady state (S.S.) envelope can also be drawn. The inclinations of these three characteristic envelopes on (p’, q) diagram may be called MPS,MPTand Ms, respectively. Figure 7 shows the same diagram for the test series with K, = 0.5. Noteworthy is that the inclinations ofthe steady state and phase transformation envelopes are uniquely determined independent of the K, values, however, the inclinations of the peak state envelopes are dependent upon the K, values from which undrained straining commences. Figure 8 summarizes the inclinations of these three characteristic envelopes against the K, values, in which Mc is the inclination of anisotropically consolidated states on (p’, q) diagram,
(b) (G, q) diagram Figure 5 . ACU tests (Jamuna river sand) i.e. M, = ( l-K,)/( l+K,). Figures 9 and 10 show the test results on Jamuna river sand, for isotropically consolidated specimens (K, = 1) and the specimens anistropically consolidated to K, = 0.7, respectively. As described above, the phase transformation envelope appears to be vacant for this sand. The inclinations of the other characteristic envelopes for this sand are summarized in Figure 1 1.
4 CHARACTERISTIC LINES Figure 12 shows the steady state lines for the two soil materials in the plot of void ratio e against logarithm of effective mean stress p’. For each soil, there is a unique steady state line for isotropically consolidated as well as anisotropically consolidated specimens. However, the inclinations of the steady state lines are not the same, most probably because the mineralogical sources of the two soil materials are different. For Omigawa sand which exhibits the states of phase transformation during undrained straining, the states of phase trans-
677
Figure 6. (p’, q) plot for ICU tests (Omigawa sand).
Figure 9. (p’, q) plot for ICU tests (Jamuna river sand).
Figure 7. (p’, q) plot for ACU tests (Omigawa sand).
Figure 10. (p’, q) plot for ACU tests (Jamuna river sand).
Figure 8. Characteristic envelopes (Omigawa sand). formation are summarized in (e, q) plot as shown in Figure 13. It is found that the steady state line is uniquely determined, however, a series of phase transformation lines are present for different axial stresses 0 1’. In other words, the soil specimens pass through the same phase transformation line, if the spe-
Figure 1 1. Characteristic envelopes (Jamuna river sand). cimens are consolidated to the same axial stress c)i independent of a confining stress 03’ and therefore the K, value.
678
dated specimens (K, = I), however, the dividing value of r, reduces as the K, value reduces. It implies that as the degree of anisotropic consolidation increases, the soil becomes more susceptible to undrained flow. Figure 15 shows the normalized residual shear strength against r, for Omigawa sand, where
-1
4, (= CT 1’ - 0 3 ’ ) , & and Ms are the shear stress, inter nal friction angle at states of phase transformation and the inclination of the phase transformation envelope, respectively. It is evident that the above equation also holds true for anisotropically consolidated specimens. However, the dividing value of r, cannot be deduced from this diagram, as all the results with various K, values are included in this diagram. Figure 16 shows the initial state ratio r, against K, for Jamuna river sand, and Figure 17 shows the normalized residual shear strength against r, for Jamuna river sand. The same observations can be made for Jamuna river sand.
Figure 12. Steady state lines.
Figure 13. Characteristic lines (Omigawa sand).
5 INITIAL STATE RATIO
Ishihara (1993) introduced the definition of an initial state ratio r,,; Figure 14. Initial state ratio (Omigawa sand). (1)
pc
where and are the effective mean stresses after consolidation and at states of phase transformation, respectively. Ishihara (1 993) examined the flow behaviour of isotropically consolidated sand specimens subjected to undrained triaxial compression, and characterized it into three modes, flow, flow with limited deformation (F.L.D.) and no flow. This study extends it to anisotropically consolidated soil specimens. Figure 14 shows the initial state ratio r, against K, for Omigawa sand, in which one can find that the boundary between no flow and flow is defined as the initial state ratio of about 2 for isotropically consoli-
Figure 15. Nomalized residual strength (Omigawa sand).
679
anisotropically consolidated sand continues in our group.
REFERENCES Anderson, S.A. & N. Sitar 1995. Analysis of rainfallinduced debris flows. J. Geotech. Eng., ASCE, 121(7): 544 - 552. Chiba Prefecture, Civil and River Division 1972. Report of disaster in Chiba due to autumn rain front on September 6 - 7, 1971, and Typhoon No.25. (in Japanese). Ishihara, K. 1993. Liquefaction and flow failure during earthquakes. Geotechnique, 43 (3): 35 1 - 4 15. Tappin, R.G.R., J. van Duivendijk & M. Haque 1998. The design and construction of Jamuna bridge, Bangladesh. Proc. Instn. Civ. Engrs., Civ. Engng., 126, NOV.:150 - 162. Tsukamoto, Y., K. Ishihara & Y. Nosaka 1998. On the initiation of rainfall induced soil failure. Geotechnical Hazards, Maric, Lisac & Szavits-Nossan (eds), Balkema: 883 - 890. Zhu, J.-H. & S.A. Anderson 1998. Determination of shear strength of Hawaiian residual soil subjected to rainfall-induced landslides. Geotechnique, 48( 1): 73 - 82.
Figure 16. Initial state ratio (Jamuna river sand).
Figure 17. Normalized residual strength (Jamuna river sand).
6 CONCLUSIONS Undrained triaxial compression tests were carried out on isotropically as well as anisotropically consolidated soil specimens. For the two soil materials used in this study, the steady state and phase transformation envelopes were present for Omigawa silty sand, however, there was no clear phase transformation envelope for Jamuna river silty sand. It was found that these two envelopes are uniquely determined in (p’, q) plot, independent of the degree of anisotropic consolidation. It was also found that there are a series of peak stress state envelopes, whose inclinations in (p’, q) plot are dependent on the degree of anisotropic consolidation. The steady state line and the phase transformation lines were also examined in (e, p’) and (e, q) plots. The boundary between flow and no flow for anisotropically consolidated soil specimens was examined with respect to the initial state ratio rc, and was found to depend upon the degree of anisotropic consolidation. A hrther study for a more unified approach to undrained flow of
680
Model test on granular soil slope and determination of strength parameters under low confining stresses near slope surface H. Matsuoka, S. H. Liu & T.Ohashi N q o y n Institute of Technology,Jcipa n
ABSTRACT: A series of model tests of the slope called tilting box tests are carried out on different kinds of granular materials in dry and wet states. It is found that the surface slip, not the circular slip, occurs in the dry granular soil slope without cohesive forces (cohesion c=O), whereas the rigid body-slip with some depth similar to the circular slip occurs in the wet granular soil slope with cohesive forces (cohesion 00). The failure mechanism of the surface slip in the dry granular soil slope is successfully simulated by DEM (Distinct Element Method). Based on this failure mechanism, an effective reinforcement method to stop the motion of particles near the surface of the slope is proposed. Furthermore, a simplified direct box shear test is used to determine strength parameters under low confining stresses (less than 1kPa) near the slope surface, and the measured angles of internal friction of samples in dry state agree well with the slope angles at failure obtained by the tilting box tests.
1 INTRODUCTION In studying slope problems, people usually pay their attentions to stability analysis or reinforcing methods. In this paper, the failure mechanism of granular soil slopes is studied both by model tests of the slope called tilting box tests and by numerical simulation using DEM (Distinct Element Method). The tilting box tests are carried out on 2-D model granular materials of aluminum rod mass and real granular materials of glass beads, Toyoura sand and crushed sand in dry and wet states. One of the titling box tests on aluminum rod mass is simulated by DEM. It becomes clear that the surface slip occurs in the dry granular soil slope without cohesive forces (cohesion c=O), whereas the rigid body-slip with some depth similar to the circular slip occurs in the wet granular soil slope with cohesive forces (cohesion 00). Based on this failure mechanism, an effective reinforcement method to stop the motion of particles near the surface of the slope is proposed. The another important problem in slope study is to determine strength parameters under ultra-low confining stresses (less than 1kPa) near the slope surface. It is usually considered that for the slope with dry granular materials (cohesion c=O), the slope angle at failure is equal to the angle of internal
friction of the dry granular materials, but it is difficult to confirm it quantitatively by tests. Umetsu, et al. (1997,1998) used plane strain compression tests and titling direct box shear tests to determine the angles of internal friction of dry sands, and the measured values were some less than the slope angles at failure by the tilting box tests. In this paper, a simplified direct box shear test is introduced, by which the strength parameters of granular materials under ultra-low confining stresses can be exactly determined (Matsuoka and Liu, 1998) and a series of tests are carried out on the same samples as used in titling box tests. The measured angles of internal friction of dry granular materials agree well with the slope angles at failure obtained from titling box tests.
2 TILTING BOX TESTS AND SIMPLIFIED DIRECT BOX SHEAR TESTS Figure 1 shows a sketch of the set-up of the model slope called “titling box”. The titling box is gradually inclined when a steel rod is driven upwards by an electric motor. The geometrical configuration of it allows a maximum inclination angle of 60°, and the inclination speed of the titling box can be adjusted by the motor. Figure 2 shows a
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Figure 1. Model test apparatus called “titling box”
Figure 3. Results of titling box tests
Figure 2. Simplified direct box shear test device sketch of simplified direct box shear test device, by which the normal and shear stresses on the shear plane can be exactly measured (Matsuoka and Liu, 1998). Since the upper shear box as used in the standard direct box shear test device is replaced by a loading plate and the normal load can be only the own weight of the loading plate, this device makes it possible to determine the strength parameters of soils under ultra-low confining stresses (less than 1kPa) near the slope surface. Figure 4. Results of simplified direct box shear tests
2.1 Tests on the samples in dry states between the slope angle at failure and the length of the model slope obtained from the titling box tests; and Figure 4 shows the relationship between the shear strength r and the normal stress U obtained from the simplified direct box shear tests. It is seen from Figure 3(d) that the slope angle at failure by the titling box tests is somewhat influenced by the length of the model slope, but the average of it tends to a stable value when the length of the model slope is longer than 80cm. This is due to the big size of the aluminum rods. In the simplified direct box shear test, the specimen is usually sheared along a plane that is away from the loading plate with a depth of
A series of both the titling box tests (Matsuoka, et al., 1996,1997) and the simplified direct box shear tests are performed on the following samples in dry states: 2-D model granular materials of aluminum rod mass, real granular materials of glass beads (0.355mm-0.6mm), Toyoura sand (D5,=0.2mm) and crushed sand (0.42mm-2mm). Two kinds of cylindrical aluminum rod mass are used: one is 1.6mm and 3mm in diameter, 50mm in length and 3:2 in mixing ratio by weight, the other is 5mm and 9mm in diameter, 50mm in length and 3:2 in mixing ratio by weight. Figure 3 shows the relationship 682
Photo.l. Order of slope failure of aluminum rod mass in dry state (a-+b+c-+d)
Table 2. Input parameters for DEM simulation
Table 1. Comparison between slope angles at failure and angles of internal friction of samples in dry state Sample 34.6"
35.5
Aluniinum rod mass
25 a
O
25.5
Normal stiffness kN(N/m/m) Shear stiffness k, (N/m/m) Normal damping q N (N s/m/m) Shear damping -qs (N s/m/m) Friction angle $k,(deg.) Density of articles D (krr/m3) Time increment At (sec.)
I I I I
I
Particle-article 6.7~10~ 2.0~10' 3.8~10~ 2.0X1O3 16 2700 2x10-7
of slope failure is shown in Photo.2. It is seen from Photo.2 that the rigid body-slip similar to the circular slip occurs in the slope of granular soils with cohesion c. The simplified direct box shear test is also performed on this wet aluminum rod mass. The measured angle of internal friction and apparent cohesion are 25" and 77Pa, respectively, by which the stability analysis is performed on the hypothesis of the circular slip. When the safety factor Fs is equal to 1.0, the calculated slope angle is 29.3" , very near to the slope angle at failure (29.5" ) obtained by the titling box test.
one or two layers of particles. Also because of the big size of the aluminum rods, the weight of a layer of aluminum rods is taken into account in the normal stress of Figure 4(d). The average slope angles at failure and the angles of internal friction of all samples are summarized in Fable 1. A good agreement between them can be seen from Table 1. Photo.1 shows the slow motion pictures during the slope failure of aluminum rod mass (1.6mm and 3 m m in diameter) taken by a video camera. It is clearly seen from Photo.1 that the failure of the slope starts firstly from the movement of the particles at the surface of the slope, gradually develops to the inside of the slope and finally a slip line is formed. That is to say, the surface slip, not the circular slip, occurs in the dry granular soil slope. To further study this failure mechanism from a microscopic viewpoint, one of the tests on aluminum rod mass expressed by the plot A in Figure 3(d) is simulated by DEM, which will be stated later.
3 MICROSCOPIC STUDY O N FAILURE MECHANISM O F GRANULAR SOIL SLOPE IN DRY STATE BY DEM As stated above, the surface slip, not the circular slip, occurs in the dry granular soil slope. In order to confirm this failure mechanism, one of the titling box tests on aluminum rod mass (5mm and 9mm in diameter) with a slope length of 8Ocm, corresponding to the plot A i n Figure 3(d), is simulated by DEM. The initial particle arrangement used in DEM simulation, as shown in Figure 5 , is digitized from the picture taken at the beginning of the test (see Photo.3), and the great effort has been made to make the particle arrangement in Figure 5 and in Photo.3 as coincident as possible. Table 2 gives the input parameters for DEM simulation. The calculated slope angle at failure by D E M is 24" ,
2.2 Tests on the aluminum rod mass in wet state In order to consider the influence of the apparent cohesion c on the slope failure, w e wet the aluminum rod mass (1.6mm and 3mm in diameter) with water, so that some cohesive force between particles is induced by the surface tension of water. The titling box test is performed on the wet aluminum rod mass (water content w=1.4%), and the slope angle at failure increases up to 29.5" , about 5" higher than that in the dry state. The pattern 683
Figure 5. Particle arrangement used in DEM
Photo.3. Particle arrangement taken in titling box test
Figure 6. (a) Distribution of particle displacernents on average, (b) Mobilized angles of internal friction along planes parallel to slope surface
near to that by the titling box test and also near to the angle of internal friction of aluminum rod mass by the simplified direct box shear test (see Table 1).
3.1 Distribution of particle displacements and mobilized angles of internal friction As the bottom and top of the slope are greatly influenced by its boundary, only the middle part with a slope length of 60cm is taken into consideration. The displacements of particles from the beginning of the slope titling to the failure of the slope (slope angle is 24" ) are averaged at every lOcm range along the planes parallel to the slope surface with a depth span of 9mm. The distribution of them is shown in Figure 6(a). It can be seen from Figure 6(a) that the particles of the slope deform on average with a pattern similar to simple shear in the middle part of the slope within a certain depth, namely, the particles move nearly along the planes parallel to the slope surface. Figure 6(b) shows the distribution of the average mobilized angles of internal friction along the planes parallel to the slope surface. It is seen from Figure 6(b) that, corresponding to the area with a deformation pattern similar to simple shear, the average mobilized angle of internal friction is about 22" -25" , nearly equal to the slope angle at failure.
Figure 7. Frequency distribution of contact normals and orientations of principal stresses from contact forces at slope angle of 24O Furthermore, the average stresses in this area are calculated from the interparticle contact forces using the formula: q, =;elq/v (Christoffersen, et al., 1981), where R is the calculation domain, V is the volume of the domain, ( , i s the length of vectors connecting the centers of contacting particles and F, is the contact force. It is found that the major principal stress is inclined to the plane parallel to the slope surface at an angle of about 33.5" , nearly equal to the angle between the direction of thc major principal stress and the mobilized plane (7r/4- 4 /2=33" ), as shown in Figure 7. This also means that the particles slip along the planes parallel to the slope surface within the middle part of the slope.
684
(deg.) 15r
Titling to 24"
( r =S%)
Figure 10. Distribution of change in contact normal directions on mobilized planes
3.2 Change in contact normal orientations along the planes parallel to the slope surface Since the surface slip of the dry granular soil slope is similar to the phenomenon of simple shear, the frequency distribution of contact normals and its change during the slope titling are studied. Figure 7 shows the frequency distribution of contact normals in the area with a deformation pattern similar to simple shear at a slope angle of 24" (shear strain Y =8%). The preferred direction of it agrees nearly with the major principal stress direction. Figure 8 shows the normalized frequency distribution of interparticle contact angles N( 8 )/N,,,, on the planes parallel to the slope surface in the area with a deformation pattern similar to simple shear at a slope angle of 24" (shear strain Y = 8%), where the interparticle contact angle 8 means the angle between the contact plane and the mobilized plane (the potential slip plane; in this case, the plane parallel to the slope surface). It is seen from Fig.8 that, with the titling of the slope, the distribution of N( e ) shifts to the right side, namely, the number of contacts increases in the positive zone o f e where the angle 8 is effective to resist shearing. Figure 9 shows the frequency distribution of contact normals which have newly been generated during slope titling, Ng( 8 ), and the frequency distribution of contact normals which have disappeared during slope titling, Nd( e ). It is interesting to find that Ng( 8 ) concentrates in the positive zone of 8 , while Nd( e ) concentrates in the negative zone of e . This is the reason why the distribution of N( 8 ) on the mobilized plane shifts to the positive zone of e , i.e., the effective direction to resist shearing. Figure 10 shows the change in contact normal directions < on the mobilized plane from the beginning of the slope titling to the failure of the slope (shear strain Y ~ 8 % ) . It is nearly proportional to the shear-normal stress ratio 7: ( 8 ) / u Ne( ) on the contact plane expressed by
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Photo.4. Failure pattern in aluminum rod mass slope reinforced by sticky tapes (lcm in width) in dry state
the following form (Yamamoto et a1.,1994): -z( 8 ) -
sin $,,,"COS(28 - $!", )
(8)
1 +sin$,,,,, sin( 28 - $,",, )
U,
4
REINFORCEMENT BASED ON FAILURE MECHANISM OF GRANULAR SOIL SLOPE
As stated above, the surface slip with a deformation pattern similar to simple shear occurs in the dry granular soil slope, that is, the displacements of the particles near the slope surface are the biggest. Therefore, it may be an effective reinforcement to stop the movement of the particles near the slope surface. To confirm the effectiveness of this method, the titling tests are carried out on the dry aluminum rod mass with thin sticky tapes (lcm in width) pasted on both the sides of the slope surface (forward and backward edges of rods). The length of the sticky tapes is 60cm, 3/4 times length of the full slope, and they are fastened at the upper edge of the slope to the titling box. It is surprising that the slope angle at failure increases to 36" -37" , increasing greatly than that of no reinforcement in the dry state, and the slope slips nearly in the pattern similar to the case in the wet state (see Photos.2 and 4). This can be explained that the weight of the reinforcing materials induces the increase in the confining stress U , ) within the slope, and the particles of the slope
behave as if there were a cohesion c (=o-,].tanb,). By considering the frictional forces between the reinforced part and the sliding body as shown in Figure 11, the slope stability is analyzed using Fellenius’s method and the strength parameters in the dry state. The calculated factors of safety is 0.98 when the slope angle is 37 . Therefore, the reinforcing effect of stopping the movement of the particles near the slope surface is explained quantitatively.
5
CONCLUSIONS
1. For the granular soil slope without cohesion c=O, a surface slip with a deformation pattern similar to simple shear occurs, whereas for the granular soil slope with cohesive force, a rigid body-slip with some depth similar to the circular slip occurs. 2. The simplified direct box shear test can be used to exactly determine strength parameters of the granular soils under very low confining stresses (less than 1kPa) near the slope surface. The angles of internal friction of the dry granular soils by the simplified direct box shear test are nearly equal to the slope angles at failure by the tilting box test. And, by using the strength parameters (c and@) of the aluminum rod mass in wet state determined by the simplified direct box shear test, the calculated slope angle at failure is well in agreement with that observed in the tilting box test. 3. The method to stop the movement of the particles near the slope surface is very effective, which can be well explained quantitatively by considering the frictional forces between the reinforced part and the sliding body. This method can also be understood intuitively in such a way that the sliding body behaves as if it were sandwiched in between the upper reinforced part and the lower slip plane.
ACKNOWLEDGEMENTS The authors would like to acknowledge the cooperation in the experimental work provided by Mr. Y. Sugiyama and Mr. M. Ichimura, former students of Nagoya Institute of Technology. The authors also wish to express their sincere gratitude to Dr. S. Yamamoto of Obayashi Corporation for his great help in DEM calculation.
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Figure 11. Forces applied on slip strip
REFFERENCES Christoffersen, J., Mehrabadi, M.M. & Nemat-Nasser, S. 1981: A micromechanical description of granular material behavior, J. Appl. Mech., Vol. 48,NO.2, pp .339-344. Matsuoka, H. and Liu, S.H. 1998: Simplified direct box shear test on granular materials and its application to rockfill materials. Soils and Foundations, Vo1.38, No.4, pp.275-284. Matsuoka, H., Ohashi, T., Ichimura, M. and Liu, S.H. 1997: Failure mechanism and effective reinforcement of granular material slope, Proc. of 32th Japan National Con. on SMFE, 938, pp.1879-1880 (in Japanese). Matsuoka, H. and Sugiyama, Y 1996: Failure mechanism and effective reinforcement of granular soil slope, Proc. of Int. Symp. on Earth Reinforcement, Fukuoka, Kyushu, Japan, pp.803-808. Umetsu, K. and Ishigami, A. 1997: Tilting box shear test and direct box shear test on dry sand, Proc. of 32th Japan National Con. on SMFE, 258, pp.5 17-518 (in Japanese). Umetsu, K. and Ishigami, A. 1998: Simple titling test and plane strain compression test on Gifu sand, Proc. of 33th Japan National C o n . on SMFE, 261, pp.527-528 (in Japanese). Yamamoto, S. and Matsuoka, H. 1994: A relationship between fabric changes and shear strain of granular materials under shear, Journal of Geotechnical Engineering, JSCE, ~ 0 . 5 0 j / 29,pp.219-228 (in Japanese).
Slope Stability Engineering, Yagi, Yarnagarni& Jiang 0 1999Balkerna, Rotterdam, ISBN 90 5809 079 5
Determination of shear strength parameters of unsaturated sedimentary residual soils for slope stability analyses Saravanan Mariappan, Faisal Haji Ali & Low Tian Huat Civil Engineering Department, University of Malaya, Kuala Lumpur, Malaysia
ABSTRACT: Weathered granite, sedimentary and metamorphic rocks are the main types of residual soil in Malaysia. In natural state the soil above ground water level are in unsaturated condition. Major parts of residual soils in Malaysia are in unsaturated state, therefore studies have to be done in order to understand the influence of soil suction on shear strength of these residual soils. Soil suction has important influences on water entry, structural stability, stiffness, shear strength and volume change, which are an important variables in soil engineering design. Shear strength determination was carried out on unsaturated sample using specially modified apparatus. At the same time the concept of multistage multi suction is implemented in order to eliminate soil variations. Discusion in the paper covers the modification of testing equipment, method of sample collection, details of multi stage lest procedure and test results. INTRODUCTION Residual soils are product of the in-situ weathering of igneous, sedimentary and metamorphic rocks. They occur in most countries of the world but the greater areas and depths are normally found in tropical humid areas such as Malaysia. Residual soils in Malaysia mainly consist of weathered igneous or sedimentary rock. The interest of this research is to study the shear strength of partially saturated weathered sedimentary residual soil. Various soil samples were collected from slope at locations of different soil weathering grades. Figure 1.0 shows the description given by Geological Society Engineering Group for residual weathering grade. Figure 2.0 shows the cut slope layout with soil sampling locations. Figure 3.0 indicates a map of weathering grades on the cut slope. Due to the variation in soil profiles, the focus is only on weathered sand stone material. UNSATURATED SOILS The principal and fundamental research on unsaturated soil mechanics started in 1962 by Jennings and Burland in Imperial College. At that time much interest was on Terzaghi's (1923) principle of effective stress for saturated soil which
was proposed by him in the First International Conference on Soil Mechanics in 1936. Fredlund and Morgenstern introduced the third factor of (U, U,) into the earlier equation of effective stress: 'I: = C I
+ (CT- U,) tan
+
(U, - u,)tan$b
-----
(1)
where: ct = effective cohesion 0 = total stress ua = pore -air pressure $' = effective angle of internal friction U, = pore water pressure (U, - U,) = matric suction Qb = angle indicating the rate of increase in shear strength with respect to changes in (U, - U,) when (CT - U,) is held constant. The above equation assumes a planar failure envelope, the internal friction angle $', remains essentially constant under saturated and unsaturated condition. The angle $b, which quantifies the effect of suction, is measured from the 'I: Vs (U, - U,) plot. The cohesion intercepts c1, c2 and c3 due to the applied suction (U, - U,) vary if the angle of internal friction Qt remains constant at different suction levels. Figure 4.0 shows the matric suction drawn on failure envelope.
687
Sides of the soil mass were then trimmed slowly and carefully to fit the sample box size. The box was then fitted to the specimen with the bottom cap opened. The whole soil mass with the box in place were dug and removed. The top cover was placed and sealed with paraffin to prevent moisture lost. All the boxes were carried with care to the laboratory and kept in constant temperature humidified room. The sample from the block sample was removed using specially fabricated split-mould sampler. During extrusion of sample, silicon oil was applied to the sampler to reduce friction. During sampling the sampler was pushed into the block sample by using hydraulic jack, cutting it to the required diameter. Finally the extruded sample will be cut to the required thickness. Figure 5.0 illustrates the split sampler. Four numbers of such split samplers were pushed into the sample at the same time in order to obtain 4 soil samples. The samples were used to perform two multistage multi suction tests, one multistage CIU test and one for soil water characteristics curve.
Figure 1.O : A schematic representation of tropical soil weathering profiles.
Figure 2.0 : Slope layout with sampling locations
Figure 4.0 : Matric suction drawn on failure envelope
Figure 3.0 : Geological map of the cut slope
TEST SETUP AND PROCEDURE
SOIL SAMPLING Undisturbed block samples were collected from the site in boxes made of metal plates measuring 200x200x200 mm. After choosing a suitable location, the topsoil of about 300mm was removed using lightweight shovels. Trenches were dug all around the soil mass of about 25Ox250x250mm.
688
Bishop-Wesley triaxial cell set was modified to carry out the test on suction induced soil specimens. The top cap of the triaxial cell was modified to provide inlet for air pressure applied at the top of specimen. Suction was applied by controlling the pore air and pore water pressure. The layout of the modified triaxial setup is shown in Figure 6.0. Axis
translation technique (Hilf, 1956) was used to apply soil suction to the specimens. A 15 bar high air entry disc was sealed on a modified base pedestal. This allowed the air and water pressures to be controlled during the application of deviator stress in order to maintain the constant matrix suction throughout the test. However, with time pore air may diffuse through the water in the high air entry discs and appear as air bubbles in the water compartment below the disc. Therefore the water compartment was fabricated to facilitate flushing of the diffused air bubbles on a periodic ba.cis. Figure 7.0 : Diffused air volume indicator
Figure 5.0 : Split sampler used
Diffused Air Volume Indicator (DAVI as shown in Figure 7.0) was used to measure the amount of air that diffused through the ceramic disc and accumulated under ceramic disc. The recorded volume change during testing could indicate the suction equilibrium in the specimen. Suction equilibrium of the specimen could determine when there were no infinitesimal changes of water volume during suction equilibrium stage. The diffused air volume measurement was performed once or twice a day or more frequently when high pressures were used. The measured water volume changes were adjusted in accordance with the diffused air volume. Multistage triaxial set up was adopted due to the limited specimens and to eliminate the effect of the soil variability. Multistage multi suction, shear test was chosen in which, the Qb value be determined based on known value of @’. According to the unsaturated soil mechanics theory (D.G. Fredlund, H. Rahardjo, 1993), the @’ for different matric suction is the same for a particular soil sample. A multistage CIU triaxial test was conducted to obtain the @ value. The test procedure for the multistage multi suction shear test is as follows:i. The specimens was sampled and mounted in the modified triaxial setup with filter paper at the bottom of the sample. (This is to prevent the fine clay material from blocking the fine pores in the high air entry disc). 11.
...
111.
Suction equilibrium and consolidation was carried out before the shearing process. Matric suction equilibration is generally attained in about one week or more. After consolidation the sample was sheared at a constant rate.
iv. Just before peak shear stress the axial force was immediately released until no significant shear
689
resisting force, allowing the sample to recover elastically . V. For the second stage of multi suction multistage shear test, the matric suction was increased to another higher suction value. Suction equilibrium had to be carried out first according to steps 2. vi . The matric suction was increases for every shearing stage. Figure 9.0 : Stress - strain curve for multi suction multistage test at berm 4
vii. Since the +' is assumed the same for every suction value, the failure envelope can be obtained for every stage. The Qb value was found based on the relationship between effective cohesion and the suction. ...
v111.
-
This multi stage multi suction shear test can actually reduce the number of samples used and time in order to obtain the shear strength parameter of the unsaturated soil.
The triaxial test setup used for testing was fully computerized (as shown in Figure 8.0jl This setup uses three pressure controllers for cell, back and lower chamber and a digital pressure Interface to measure and maintain pore water/air pressure respectively .
Figure 11.0 : Mohr circle plots for multistage CIU test From the CIU test results, friction angle $' of 26' and 33' were obtained for sample a! grade IV and grade 111. Using the friction angle (+, ), parallei lines are plotted to obtain effective cohesion for various suction, shown in Figure 12.0 and Figure 13.0.
TEST RESULTS AND DISCUSSIONS Two sets of test results are presented here for discussions. The sample were collected at TP5 Level 1 (weathering grade IVj and Berm 4 (weathering grade 111). Both samples were collected from the sandstone zone. A typical test results of : 1. stress-strain curve for multi suction mu1tistage, 2. plots of continuous water volume change during suction consolidation, 3. mohr circle plots for multistage CIU test results, are shown in Figure 9.0, 10.0 and 1 1 .O respectively.
Figure 12.0 : Mohr circle plots for multi suction multistage test at TP5 level 1
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hgure 13.0 : Mohr circle plots for multi suction multistage test at berm 4 The effective cohesions obtained are then plotted with matric suction to determine the value of $b (angle indicating the rate of increase in shear strength with respect to changes in (U, - U,)) in Figure 14.0 and Figure 15.0. From the above plots the contribution of suction in shear strength reduces when the suction value gets higher. In addition to the above tests, soil-water characteristic curves were also determined for both laboratory and field tests. A typical plot of soil water characteristic curve at berm 4 (soil of grade IV) is shown in Figure 161. Many more samples will be tested in the future to verify these test results. In the final part of this research work, stability analysis of the slope will be conducted at various sectional profiles to determine the changes in factor of safety caused by suction.
Figure 16.0 : Combination of field and laboratory soil-water characteristic curve for berm 4 CONCLUSION The proposed multi-stage triaxial testing procedure to evaluate the rate of increment in shear strength $b concerning matric suction is possible provided that $' is assumed to be constant at all suction level. Furthermore triaxial test on unsaturated soil specimens using multi-stage technique will greatly reduces the sample or soil variation and disturbances. REFERENCES
a,
Affendi A., (1996). Field and laboratory study on unsaturated residual soils in relation to slope stability analysis. Ph.D. Thesis. University of Malaya, Malaysia.
n (d
Affendi A, Faisal A. & Chandrasegaran S, (1994) Triaxial shear tests on partially saturated undisturbed residual soil. Geotropika, Malacca, Malaysia. D.G. Fredlund, H. Rahardjo,( 1993). Soil mechanics for unsaturated soils, John Wiley & Sons.
5
2 50 2 40
W
Low Tian Huat, Soenita Hashim, Faisal Hj. Ali. (1997). Shear strength of Undisturbed partially saturated residual soils. Geotropika, Johor, Malaysia. pp 69-8 1.
a,
.k! 30
cn
$20
2
Yong, R. Wakentin, B. P. (1975), Soil water behavior of soil,: Chapter 6, pp 127-150. "
10 O
b
50
10;
1L.o
do0
Matric Suction (kPa)
25;
Figure 15.0 : Matric suction drawn on failure envelope for sample at Berm 4 69 1
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Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
The characteristics of landslides caused by the hydrothermal metamorphic clay H.Yam ashita Shikoku Regional Bureau, Japan Highway Public Corporation, Japan
M. Saga & H. Fujita Takumutsu Engineering Office, Japan Highway Public Corporation, Jupan
K.Yokota & R.Yatabe Fuculty ojEngineering, Ehime University, Matsuyama, Jupun
ABSTRACT Various studies on landslides have long been carried out at different sites along fractured zones of Median Tectonic Line in Shikoku, Japan. The study carried out in this paper relates to the landslide sites where the soil of the slip layer is hydrothermal metamorphic clay that is a clay formed by the hydrothermal alteration of metamorphic rocks. Soil samples from three such landslide sites were collected and tested for strength and clay minerals content. Tri-axial compression and ring shear test results showed that 6 ' and 6 of hydrothermal metamorphic clay range from 19" to 45" and 7 to 37 respectively. It was also clear that 4 for hydrothermal metamorphic clay is very low; and the difference in 6 ' and 6 is higher compared to that for other clays. Results of x-ray diffraction showed mica and chlorite content in most of the soil samples. However, samples with lesser 4 I values were found to contain smectites and expansive chlorites. O
O
by the hydrothermal metamorphic clay were also studied. For this, three landslide sites at fractured zones along the MTL were chosen. Soil samples from these sites were taken and tested for strength as well as minerals content.
1 INTRODUCTION Many landslides have resulted along the fractured zones of the Median Tectonic Line (MTL) in Shikoku region of Japan due to slope cutting as a measure for slope stability during expressway construction. MTL is a first class active fault line in the country. Several intrusive rocks are distributed widely in the fractured zones of MTL. So many landslides with a slip layer of hydrothermal metamorphic clay are in the active state. Geological study makes it clear that the hydrothermal metamorphic clay is formed by the hydrothermal alteration of metamorphic rocks. It is a very weak clay with mineral content mostly of chlorites and smectites. These two minerals are the weakest clay minerals, and show some peculiar behaviors with water; as a result, making the clay mass weak in strength. It is supposed that when black and green schist at fracture zone come in contact with a very hot underground water, they are changed metamorphically into hydrothermal metamorphic clay. Hot ground water spreads all around in a plane, and whole plane of those rocks changes into the hydrothermal metamorphic clay that later becomes the slip layer of the landslide. The purpose of this study was to determine the strength characteristics and the minerals content of hydrothermal metamorphic clay. At the same time, the mechanical characteristics of landslides caused
2 STUDY AREA There were three landslide sites namely Takao, Higashimine and Shintani, chosen for the study, as shown in Figure 1.
Figure 1: Landslide sites location map.
2.1 Takao site
This site is located at Donari town of Tokushima prefecture. The slope at this site is a cut slope. Plan and profile of the landslide site have been shown in Figures 2 and 3 respectively. Slope length of the
693
sliding mass is 80m, maximum width is 80m, and maximum depth is 10m. The cut slope of the landslide mass is 34" Base rocks of this landslide soil mass are shale and sand rocks of Izumi soil group. The hydrothermal metamorphic clay is in between these two rocks.
of the sliding mass is 150m, maximum width is 140m, and maximum depth is 20m. The maximum slope of the sliding soil mass near the toe is 30" and the average slope is 25". Base rock of this landslide soil mass is green schist. The sliding soil mass is supposed to be a deposit of slope failure or landslide in ancient times.
Figure 2: Plan of Takao landslide site. Figure 4:Plan of Higashimine landslide site
Figure 3: Profile of Takao landslide site.
Figure 5: Profile of Higashimine landslide site
The problem of landslide at this site started when the soil slope was cut as a measure for slope stability during the construction of expressway. After one year of the cutting, the soil mass at this site started moving resulting to a large scale landslide. Later after some investigations, the clay layer along the slip layer of this landslide was found to be hydrothermal metamorphic clay (white in color).
The soil mass at this place started sliding resulting to large scale landslide, when a 25m deep bridge pier was inserted into the ground. After the soil investigations, the clay layer along the slip surface of this landslide site was also found to be the hydrothermal metamorphic clay. 2.3 Shintani site This site is located at Oozu city of Ehime prefecture. Plan and profile of the landslide site are shown in Figures 6 and 7 respectively. Slope length of the sliding mass is 160m, maximum width is 80m, and maximum depth is 30m. The cut slope of the sliding soil mass is 30".
2.2 Higashimine site
This site is located at Futami town of Ehime prefecture. Plan and profile of the landslide site are shown in Figures 4 and 5 respectively. Slope length
694
The base rocks of this landslide mass are black and green schists.
the landslide clays, tri-axial compression and ring shear tests were carried out. The test results showed that c' and c, were zero. A comparison between the results of 6 ' and @ of ordinary landslide clay and hydrothermal metamorphic clay is shown in Figure 8; and a comparison in relationship between I, and (6'-@ ,) of the same clays is shown in Figure 9. It is clear from the Figure 8 that @ ' for hydrothermal metamorphic clay is ranging from 19" to 45",and @ ,for the same is ranging from 7"to 37 ". Similarly in Figure 9, (6'-@ r) is ranging from 4"to 22". This shows that the angle of shearing resistance of hydrothermal metamorphic clay at residual state is very low; and the difference i n @ ' a n d @ ,is higher compared to that for other clays. This means that even if the slope is very gentle the movement of soil mass above this clay layer can easily occur. The movement of soil mass above the hydrothermal metamorphic clay is due to the very same reason.
Figure 6: Plan of Shintani landslide site.
Figure 7: Profile of Shintani landslide site. Figure 8: Results of d ' and dr values of tested landslide clays.
The problem here also came to be known when the soil slope at this site was also cut during the construction of expressway as a measure of slope stability. Just after slope cutting, the soil mass started moving resulting to a large scale landslide. The clay layer along the slip surface of this landslide was also found to be hydrothermal metamorphic clay.
3 STRENGTH CHARACTERISTICS OF LANDSLIDE CLAY Soil samples from the slip surfaces of all the three landslide sites were taken out by out crop and core boring methods. Soil samples from other ordinary landslide sites for strength comparison were also tested. All the tests were carried out with remolded samples. TO determine the strength parameters of
Figure 9: Relationship between I, and ( d '- @ J values of tested landslide clays.
695
4 X-RAY ANALYSIS FOR CLAY MINERALS
5 CONCLUSION
After the tests, when 6 ' and 6 were found to be ranging from very small to large values, the clay samples then were tested for clay mineral content by x-ray diffraction. The methods of x-ray diffraction test carried out were the powder method, ethylene glycol treatment, and 500°C heat treatment. The results of x-ray diffraction test on the clay samples from the all the landslide sites are shown in Figures 10, 11, and 12. From the result, it is clear that the clay mineral contents as a whole in all the samples are mica and chlorite, whereas those in the sample with very small 6 were found to be smectite and expansive chlorite.
From the results of strength tests and x-ray analysis, the following two points as the conclusion of this study can be written: 1. Shear strength of hydrothermal metamorphic clay in compared to that of other clays is less. It is because, it contains smectite and expansive chlorites (as shown by x-ray analysis) which have very small 6 values. 2. Landslides occur at the region of hydrothermal metamorphic clay because this clay spreads all around in a plane which later becomes a slip layer due to its weak shearing strength.
1
500°C heat trcatment method
REFERENCES 1. M. Enoki, N. Yagi and R. Yatabe: Shearing characteristics of landslide clay, Proc. of seventh ICnVL, pp.231-236, Aug.1993.
A )\
ethylene glycol method
I
2. Ryuichi Yatabe, Norio Yagi and Meiketsu Enoki: Ring shear characteristics of clays in fractured zone landslide, JSCE Journal No.436/111-16, pp.93-101, 1991.9.
I
I
'O 2 e ( c u . ~ a ) ' O
30
3. Shuji Sato, Akira Miyamoto, Norio Yagi and Masayuki Okuzono: The mechanical characteristics and countermeasures of landslides at the fractured zone on median tectonic line, JSCE Journal No.546NI-32, pp.125-132, 1996.9.
Figure 10: x-ray analysis of landslide clay from Takao.
Oo0
r
kL
ethylene glycol method
, ( I
,
powder method J '028(Cu,Ka)
30
Figure 11: x-ray analysis of landslide clay from Higashimine. 1000 dycol method ~
0
8oo[ GO0
Ioriented sediment method
I 0
I
I
I
I
10
20
30
40
2 e (cu, K
1
Figure 12: x-ray analysis of landslide clay from Shintani.
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Slope Stability Engineering, Yagi, Yamagami& Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Influence of clay minerals on strength characteristics of landslide clay in Mkabu Tomonori Ishii Mutsuyama City Ofjce, Japan
Ryuichi Yatabe, Norio Yagi & Kinitada Yokota Ehime University,Mutsuyama, Japan
ABSTRACT: Many landslides have occurred on Mikabu belt in Shikoku Island. In this Paper, the influence of the clay minerals on strength characteristics of the landslide clay in Mikabu belt whose mother rock is green rock were investigated. The green rock is very weak and easily weathered. The main clay mineral of Mikabu green rock and its weathered clay is chlorite. There are two kinds of chlorite in it. One is chlorite and the other is expansive chlorite. The other clay minerals of green rock and its weathered clay are the montmorillonite, the quartz and feldspar. The strength parameter of landslide clay containing the quartz and the feldspar at the peak and residual state was large, and that of containing the expansive chlorite or the montmorillonite was small.
1 INTRODUCTION There are complex geological and the Median tectonic line in Shikoku region (shown in Fig.1). Therefore a lot of landslides have occurred along this line. The types of landslide, the strength parameters (at peak and residual state) and the amount of weathering of landslide clay are quite different in same geological belt. It is difficult to construct reliable countermeasure work for above problem. The cause of difference in the type and in the strength parameters may be due to different clay mineral content and the amount of weathering. In order to investigate the clay mineral content, X-ray analysis of landslide clay was carried out. The shear
Fig.1 Location of the landslides and geological belts in Shikoku Region
tests to find out peak and residual strength were also carried out. The chlorite, which is the main mineral in the weathered green rock, was contained in whole of the specimens. It is clear that the strength parameters 4 , 4 I of the specimens containing the quartz and the feldspar were large where as those of the specimen containing the expansive chlorite or the montmorillonite were small.
2 EXPERIMENTAL METHOD 2.1 Sample preparation and shear tests The core of the sliding layer was obtained from bore hole. The properties of the samples are shown in Table 1. The ordinary consolidated undrained triaxial test with pore pressure measurement to find out the peak strength parameter 4,’ and the ring shear test (Yatabe,et.a1.,1991) to find out the residual strength parameter 4 were carried out. Remolded samples were used for the shear tests. Particle diameter of the sample was less than 420p m. The sliding layer soil samples of the landslide include particles larger than 420 P m diameter. 4 ,’ in terms of effective stress gives the same value for both undisturbed and remolded samples. (Yagi,et.a1.,1989). However, if undisturbed samples contain much amount of sand and gravel, GP’ , Q r should be larger than that of remolded samples. The influences of sand and gravel have been investigated by the authors (Yagi7et.al.,1994).
697
Table.1 Physical properties of landslide clay obtained from bore hole.
If sand and gravel content in the samples were less than 30%, its strength parameters Q, p' and 4 were approximately the same as those of the remolded samples.
2.2 X-ray diffraction
of 500°C has also same position (in Fig.3). But the sample in Fig.2 does not have peaks. This fact implies that the specimen in Fig.:! contains the montmollironite, and the one of the specimens in Fig.3 contains the expansive ch1orite.X-ray diffraction of the other specimens of Mikabu green
X-ray diffraction was carried out at an electricity condition of 30kv and 15mA, with Cu as the target and Ni as the filter. Scanning speed was 1" /min. At first, the powder method was carried out. Then, in order to investigate existence of the expansive chlorite and the montmorillonite, the specimens were treated in order, with hydrochloric acid, ethylene glycol and by heat of 500°C. 3. TEST RESULTS
3.1 Clay minerals in slip layer clay Fig.2 and Fig.3 show an example of X-ray pattern. The sample in Fig2 contains the montomorillonite and the sample in Fig.3 contains the expansive chlorite. There is a peak at 6A for the original clay in Fig.2 and in Fig.3. The sample treated with the echylen glicol has also a peak at the Same position. The position of peak for the sample treated by heat 698
Fig.2 X-ray diffraction patterns (containing montmorillorite)
rock were also carried out. These results showed that the main clay mineral of Mikabu green rock is the chlorite. However, there are two kinds of chlorites. One is chlorite, and another is the expansive chlorite. The feldspar, the torenolic, the quartz and the montomorillonite are also contained.
Fig.3
compared to those of the specimens not containing and quartz or feldspar. The difference between 4 4 of the specimens containing quartz or feldspar is also large.
X-ray diffraction patterns (containing chlorite)
Fig.5
;and
6
;,
6
and Ip
Fig.5 and Fig.6 show the relationship between , 4 and plasticity index Ip, and that between , 4 and clay fraction CF(0.02. Feda (1968) presents the following equation to evaluate the possible potential of collapsibility of soils: Kb={ (wo/SO)-PL]/PI>O. 85
In the above equation WO is the in situe moisture content and SO is the degree of saturation. In addition, PL and PI are the plasticity limit and plasticity index. Even though in the above mentioned equations some major geotechnical parameter are taken into account, the role of physico-chemical factor have not been considered. Yong & Ouhadi (1997), evaluate the reaction factors impacting on instability of bases on natural and lime-stabilized marls. They present a mechanistic model to explain the different aspects of the collapsible performance of soils. On the soil sample studied, they indicate that the maximum swelling for the natural sample is not significantly lesser than the washed sample over the longer term period. However, we note that the washed sample reaches the maximum swelling in at least one-fourth the time period taken by the natural sample to reach its own maximum free swell. This performance to gether with the index properties shown in Table 1 are pieces of information which reveal the reaction effects that will contribute to strength reduction and subsequent instability for the compacted collapsible clay.
Whereas one could argue that the reduction in the various salts and sulfate are considerable, and that the reductions are not reflected in comparable property changes as noted in Table 1, the impact of these changes need to be viewed in terms ofthe physico-chemical processes. Before discussing the reaction consequences via mechanistic model interpretations, we can view the XRD peak intensity for clay fiaction for the natural and washed state as presented in the Table 2. Table 2. XRD peak intensities of clay fraction for natural and washed soil samdes. Basal spacing Natural sample Washed sample
..~An.t3.st!o.!?:! 2.6 4.1 4.4 5.3 6.1 10.2
.........................................................................................................
4.6 24 8.7 6 5 30
4.8 28 8.9 9 4 42
As noted from the results shown in the Table 2, there is a significant increase in the peak intensities of the different reflection lines at the various basal spacing except at the second last basal spacing. This information lends weight to the mechanistic model developed by Yong & Ouhadi (1997). The wetted state mechanistic model which is developed by them to show the changes in the integrity of the compacted natural collapsible soils, benefits from the collective information presented in Table 1 and Table 2. In fact the wetted state can lead to collapse of the compacted soil or to dispersive behavior. The @ model of the natural compacted soil (not wetted) Natural Washed Test shows precipitated carbonate and sulfate bonds 49.6 Liquid limit % 45.8 forming the core of the cementing relationships for 24.4 30.3 P.I.% the flocculated structure. The equivalent matrix290 Sulphate, ppm 5520 osmotic pressures developed as a result of 230 Na, PPm 18230 interpenetration of the diffise ion-layers from 3 10 40 adjacent particle. The wetted-state condition which K, PPm Ca, PPm 670 20 is developed after compaction of the soil sample Mg, PPm 410 60 results in weakening of the cementation effect Qpt.yd Mg/m3 1.78 1.71 produced by the carbonates and sulfates and Opt. a,% 19 22.5 significant reduction in the salt content of the soil. Maximum free Swell, 'Yo 9.9 10.4 The destabilizing outcome of the above points occurs through the increase in the matrix-osmotic pressures because of the reduction in salt concentration. This increase in matrix-osmotic The significance of the results lies in the pressures can be predicted from diffise double-layer phenomenon of leaching of the compacted soil by theory, and should these pressures exceed the influent water, generally obtained as rainfall.
707
American Society of Civil Engmeers, 925947. Feda, J. 1968. Structural stability of salient loess soils from praha-djevice. Engineering Geology, Vol. 1.t Loloret, A. & E.E. Alonso 1980. Consolidation of unsaturated soils including swelling and collapse behaviour. Geotechnique 30, No. 4, pp 449-477. Rollins, K.M., Rollins, R.L., Smith, T.D., & G.H. Beckwith 1994. Identification and characterization of collapsible gravels. Journal of Geotechnical Engineering, Vol. 120, NO.3, 528-542. Tadepalli, R., Fredlund, D.G. & H .Rahardjo 1992. Soil collapse and matrix suction change. Proc. 7th Int. Con$ on Expansive Soils Dallas, I , 286-291. Zur, A. & Wiseman, G. 1973. A study of collapse phenomena of an undisturbed loess. Proc. Intern. Con$ on Soil Mech . and Found Engg., Vol. 2.2, 256-269. Yong, R.N., & V.R. Ouhadi 1997. Reaction factors impacting on instability of bases on natural and lime-stabilized marls. Special Lecture, Proceedings of the International Conference on Foundation Failures, 87-97.
confining stress and bonding established by the cementing bonds,swelling of the soil results, andor self detachment of particles occurs, leading thereby to dispersive soil behavior. Continued exposure to water in the wetted state will contribute to instability. The information showed in the Table 1 indicate that the structural integrity established in the bonded soil has been destroyed because of the wetted state reactions, as described. Confirmation of the dispersed structure is obtained by the XRD information shown in Table 2. The higher intensities shown by the washed samples indicate well oriented particle arrangements. We can therefore expect a dispersed structure for the wetted state, and a dispersive behavior of the system.
3 CONCLUSIONS
1. Mechanical properties of soils are not enough for evaluation of the collapsible potential of soils. Equations presented based on the mechanical properties have the same problem. 2. Physico-chemical factors including XRD analysis can be used as a safe factors to explain the general basic causes for collapsible Performance. Xray diffraction is able to provide some realistic evaluation of the collapsible performance of soils. REFERENCES Abelev, I.M. 1968. Principles of planning and execution in collapsible loess soils. Moscow. Alonso, E.E. 1993. Problematic soils, state of the art report. Proceeding of the Second International Seminar on Soil Mechanics and Foundation Eng. ofIran, 52- 100. Baker, A. A. 1964. Geology of the Quadrangle Utah Map GQ-241 . Geologrc Quadrangle Maps of the United States, US. Geological Survey, Washington, D.C. Basma, A.A. & E.R. Tuncer 1992. Evaluation and control of collapsible soils. Journal of Geotechnical Engineering, Vol. 118, No. 10, 1491-1504 Denison, N. Y. 1953. Properties of loess soils in construction. Moscow. Dudley, J . H. 1970. Review of collapsing soils. Journal of the Soil Mechanics and Foundations Division, Proceedings of the
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Slope Stability Engineering, Yagi, Yamagami& Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Deformation characteristics of a compacted clay in wetting tests under isotropic and triaxial stress state S. Kato & K. Kawai Department of Architecture and Civil Engineering, Kobe University, Japan
ABSTRAm:Defomation in collapse has been studied with a tnaxial test apparatus modified for unsaturated soil. Two kinds of wetting tests, in which the conditions of suction and stresses were known, were conducted for specimens of a compacted clay. Deformation characteristics in collapses under different stress states were studied and discussed. The relations between void ratio change and increase in water content observed in collapses under these different stress states have the same tendency. And in the case of triaxial stress state, it took several times hours till collapse occurred than that which were needed in the case of isotropic stress state. These test results mean that, after inundation, the states of the unsaturated soil is independent of the stress state, but the process of deformation and absorption during wetting process is affected by the stress state. 1 INTRODUCTION One reason for failure and deformation of a slope after rainfall and sinking of a fill in inundation is collapse that happens by saturation of the soil fiom unsaturated state. This phenomena has been studied by the inundation test to which oedometer test apparatus was used. (for example, Lawton & Fragaszy 1989) But with this apparatus, the lateral stress and suction in the specimen were unknown. s o the test results have been analyzed by some experimental method. It is difficult to grasp the essence of collapsc from these analyses. And there have been few data for collapse in which the suction and all of the stress are obvious. It is therefore necessary to accumulate the test data for collapse under known stresses and suctions. The aim of this paper is to provide data relevant to the understanding of collapse mechanism, by wetting tests with using the trkaxkal test apparatus in which suction and net strcss for the specimen were controlled. And we will show that collapses, one of which occurs under isotropic stress state and the other of which occurs under triaxial state, have the same tendency after wetting processes, but the processes of deformation and absorption during each wetting process are afkted by the stress state.
order to keep some water content, was compacted in five layers with a compaction rod. The compaction stress was about 314 kPa, and each layer was compacted 15 times. The compacted sample was trimmed to a specimen of 35mm diameter and 8Omm height. The optimum water content obtained by this compaction ll of the specimens were method was about 35%. A prepared at water content of 26%, which is the dry side for the optimum water content. The initial states of the specimens were as follows:(l) the void ratio was about 1.31;(2)the degree of saturationwasabout 53%. Fig.2 shows a schematic drawing of the triaxial test cell used for all of the tests. A ceramic disk, whose air entry value is 275 kPa,is equipped into the pedestal. The suction, which is defined as a pressure difference between pore air pressure and pore water pressure, was given for the specimen by the pressure plate method. A lateral displacement-measuring device was used, and the volume of the specimen was calculated by using the approximation that the specimen had a section of "beer bane1 shape" whose side view was a parabola decided by the measured diameter of the specimen. 2.2 Stressp a t h for wetting test Two series of tests were carried out:(l) wetting tests under isotropic stress state;(2) wetting tests under triaxial stress state in which shear stresses and mean net principle stress were kept constant. All of the processes in these tests were carried out by the stress control method with step loading under drained condition. One stress state was kept for 8 hours which was usually enough time for deformation and drainage to reach its equilibrium state. But during the wetting processes, at the suction of 0 kPa under isotropic
2 EXPERIMENTALPROCEDURE 2.1 Soil ype and test apparatus A powder clay, whose specific gravity was 2.71, was used. The liquid limit was 40% and the plastic index was 12.3. The grading of the clay is shown in Fig.l. The sample, to which the required quantity of distilled water was added in 709
Fig.3 Stress paths of wetting test
Fig.2 Schematic drawing of triaxial test cell stress state and under a constant shear stress, the stress states were kept about 24 hours and about 10 days, respectively. These times were needed for collapse to occur as shown in test results later. All the required time for one test was about from 2 to 4 weeks. In spite of such long time, drying of the specimen was limited to be about 2% reduction in water content. Fig.3(a) shows the stress paths in the wetting tests under isotropic stress state on a plane represented by mean net stress vs. suction. Tests were started from a initial stress point "A" at which the mean net stress was 20 Wa and the suction was 49 kPa. Suction was increased to 245 kPa, and specimens were compressed under the constant suction to the mean stresses of 98,196 and 392 Wa. After cornpression by applying the mean net stress, the suction was decreased to 0 Wa in steps under the constant mean net stresses. These stress path are shown as ACODO, ACClDl and ACC2D2 in Fig.3(a), respectively. In another test, from the stress point A, suction was decreased to 0 Wa under the constant mean stress of 20 kPa, and then mean net stress was increased. This stress path is shown as ADDoDlD2 in Fig.3(a). From the stress points DO, D1 and D2, triaxial compression tests were carried out under the constant mean net stresses. Fig.3(b) shows the stress paths of the wetting tests under
710
constant shear stresses. Form the initial stress point A, the specimen was compressed by suction and mean net stress with a stress path of ACCZ in Fig.3(a). Then tnaxial compression tests were carried out under the constant mean net stress and suction. At the shear stresses of 381 Wa and 421 kPa, the suction was decreased to 0 kPa in one step under the constant shear stresses and the constant mean net stress. These stress paths are shown as CZEIF1 and CZE2F2 in Fig.3(b) respectively. 3 DEFORMATIONCHARA~RTSTICSUNDER ISOTROPICAND TRLAxlALSTRESS STATE Figs.4(a) and (b) show plots of void ratio and water content against mean net stress in cornpression and wetting process under isotropic stress state. In these figures, the solid lines and black dots show the results of wetting test which traced the stress paths CCoDo, CClDl and CC2D2 in Fig.3(a), and the dotted lines and white dots show the results of compression test which traced the stress paths DDoDID2 in Fig.3(a), respectively. In all of the former test results, collapses occurred during the wetting process. And after the wetting process, the state of the former results agree with those of the later test results. Fig.5 provides plots of the difference of void ratio and degree of saturation between the specimen, which traced the stress path CC2, and the specimen, which traced the stress path DDoDlD2 against mean net stress during the wetting process. The difference of void ratio is equivalent to result of the double oedmeter test, and corresponds to the settlement that occurs in collapse. From this figure, it is
Fig.4 Void ratio and water content against mean net stress in wetting process
Fig.5 DifPerence of void ratio and degree of saturation during wetting process
Fig.6 Void ratio and water content against stress ratio in the wetting test under constant shear stresses found that the difference of void ratio shows a similar tendency with the difference of degree of saturation, and that the maximum settlement occurred at the mean net stress that gave the maximum difference of degree of saturation. In the past studies which were conducted with oedmeter test apparatus, the maximum settlement was observed around the overburden pressure which corresponds to the preconsolidation stress. From this result, it is found out that the dif€erencein degree of saturation affects on the settlement. Figsqa) and (b) show plots of void ratio and water content against stress ratio in the wetting test under constant shear stresses. In these figures, white dots show the results of initially soaked sample which traced the stress path ADD2 in Fig3(a), and the black dots show the results of soaked sampIe which traced the stress path ACCZ shown in Fig.3(a). From these figures, it is deduced that after the wetting process under constant shear stresses, the states of the soaked sample agree with that of the initially soaked sample. This is the same tendency observed in the wetting test results under isotropic stress state shown in Figs.qa) and (b). From these results, the stress path independence of the void ratio and water content is confirmed under obvious stress state. Fig.7 shows the relations between shear strain and
stress ratio, volumetric strain for the same data shown in Fig.6. In this figure,the triangles show the results of triaxial compression test for the non- soaked samples which traced the stress path ACCz in Fig.3(a), and the circles show the results of triaxial compression test for the initially soaked samples which traced the stress path ADDzin Fig.3(a). It is deduced that after the wetting processes, the shear strains and volumetric strains become bigger than those in the initially soaked sample at the same stress state. Fig.8(a) shows a plot of increase of water content against decrease of void ratio in the wetting process under isotropic stress state. In the case of p=20 kPa, almost all the decrease of void ratio occurred during the decrease of suction fiom 10 to 0 kPa. The reason why collapse did not occur during the decrease of suction from 245 to 10 kPa is considered to be that the influence of meniscus on stfiess of soil skeleton was dominant. In the other data, the decrease of void ratio increases linearly with the increase of water content. It should be noted that this tendency is almost independent of the confining stress. And in all samples, the more water was absorbed, the more compression occurzed ultimately. This result indicates that the quantity of the absorbed water has an influence on the compression. Fig.8(b) shows a plot of water content change against 711
Fig.7 Relations between shear strain and stress ratio, volumetric strain in triaxial compression tests
Fig.8 Comparison of water content change against void ratio change during wetting process under isotropic stress state void ratio change in the wetting process under constant shear stresses. In this figure, circles show the result under a constant shear stress of q=343 kPa,and triangles show the result under a constant shear stress of q=421 kPa. The broken line is the same as that shown in Fig.8(a). It should be noted that test results are around the broken line. This means that the decrease of void ratio is in proportion to the quantity of the absorbed water. This is the same phenomenon as shown in isotropic stress state. From these results, it is concluded that the collapses under isotropic stress state and under shear stress state occur in the same process. 4 DEFORMATION C X A R A ~ R I S T I C S AGAINST ELAPSED T IME Figs.9(a) and (b) show plots of void ratio and water content against elapsed time in the wetting process under isotropic stress state respectively. The change of void ratio and water content converged gradually to a particular state in each process. Figs.lO(a) and (b) show the same relations in wetting
process, in which the suction was decreased from 245 to 0 kPa in one step, under constant shear stresses. The state of the specimens changed gradually and then changed rapidly. The elapsed time needed for this rapid change is longer for the case of q=421kPa than for the case of q=343 E a . This rapid change is caused by collapse. It must be emphasized that the elapsed times when collapse occurred under constant shear stresses are more longer than that for isotropic stress state. This result is considered to be under the influence of the change of the bulk water to the meniscus water (Karube and &to, 1994) by shear deformation. Fig.11 shows the concept of the bulk water and the meniscus water. When the unsaturated soil mass was subjected to shear stress, macro pores, in some of which the bulk waters exist, deform and some water skins of the bulk waters are broken. Then the bulk waters are redktniuted to meniscus waters on the contact points of soil particles around them. Because of this reason, the area ratio of the bulk water to the cross section of the soil mass including voids decreases, and this decrease of the area ratio is concerned with the decrease in the permeability. When the suction is decreased in this state, the meniscus waters at the contact points expand by water absorption
712
Fig.9 Void ratio and water content against elapsed time in wetting under isotropic stress state
Fig.10 Void ratio and water content against elapsed time in wetting under a constant shear stress
Fig.] 1 Definition of the bulk and the meniscus waters
due to wetting gmdually. To some contact points, the expanded meniscus waters combine and change to a bulk water. Then the area ratio of the bulk water increases and the permeability of the soil mass increases. When these combined bulk water condense continuously, a water chmel is made. This water channel causes the water to enter into the voids and collapse to occur suddenly. 5 CONCLUSIONS
The deformation of a compacted clay in collapsing was
studied by triaxial test apparatus being modified for unsaturated soil. Wetting tests under isotropic stress state, wetting tests under constant shear stress were conducted. The following conclusions were derived from the results and discussions: The quantity of compression in collapsing increased linearly with the quantity of absorbed water. This means that collapse occurs in the voids into which the absorbed water enters. The relation between the void ratio change and the water content change in wetting test under isotropic stress state had the same tendency as that observed in wetting test under constant shear stresses. This result means that the collapse occurring under constant shear stress is almost the same process as that occurring under isotropic stress state. The relations between decrease of void ratio and elapsed time in collapsing under constant shear stress were different from those under isotropic stress state. The cause of this phenomena is lowering of the permeability which occurs when the sample was
713
sheared under high suction. This phenomenon is affected by the change of the bulk water to the meniscus water according to the shear deformation under high suction.
REFERENCES Karube, D. & S. &to 1994. An ideal unsaturated soil and the Bishop's soil, Proc. 13th Int. Con$ SMFE, 1,:43-46. Lawton, E.C., Fragaszy, R.J. & J.H. Hardcastle 1989. Collapse of compacted clayey sand. ASCE. :115.GT9:1252-1267.
NOTATIONS oli ;total
principal stress (i=1,2 and 3),
,U , ;pore air and water pressure, oi= oti - U , ;net principle stress (i=1,2 and 3),
U,
p = (ol+ oz+ 0 3 /)3 ;mean net principal stress,
q = o1- o3;shear stress, s = U , - U , ;suction, t'i
;principal strain (i=1,2and 3),
t'd
= 2/ 3 x
t', = cl
(zl - E ~ ;shear ) strain,
+ 2~~;volumetricstrain.
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Slope Stability Engineering, Yagi, Yarnagarni & Jiang 0 1999 Balkerna, Rotterdam, ISBN 90 5809 079 5
Development of an automatic cyclic direct shear test apparatus for landslide slope stability analysis Masafumi Okawara Department of Civil and Environmental Engineering, Faculty of Engineering, Iwate University, Morioka, Jupun
Toshiyuki Mitachi Division of Structural and Geotechnical Engineering, Graduate School of Engineering, Hokkaido University, Sapporo, Japan
Makoto Tanada Faculty of Engineering, Iwate University, Morioka, .Japan
ABSTRACT: The determination of strength parameters for stability analysis is the most important job in landslide slope stability evaluation. In actual practice in Japan, strength parameters have been almost always determined by an empirical method named as ''reverse calculation method". In this method, strength parameters are back calculated based on equilibrium conditions of sliding earth mass. In the first place, the apparent cohesion ( c ) is assumed as CO + d (kIWm2) (d:thickness of sliding mass(m)) and then the angle of shear resistance ( 4 ) is obtained by substituting the value of c in the stability equation and assuming the current safety factor FO= 1.0. Although criticisms have frequently been made to this conventional method, it is still widely used in practice. The authors proposed a method for determining strength parameters for stability calculation rationally, and also proposed a practical method in which the strength parameters for design purpose are given by combining the conventional reverse calculation method with the strength parameters obtained by laboratory shear test ( Mitachi et al. 1996 and 1999 ) . Landslide slope stability calculation by using this method requires a shear test apparatus by which the strength parameters corresponding to peak, fully softened and residual states can be evaluated. The present authors have newly developed a high-precision automatic cyclic direct shear test apparatus using a digital servomotor for vertical force loading. This paper presents the results of cyclic direct shear tests on several clay samples using the new apparatus and comparisons with the results obtained from other test methods by ring shear test apparatus and also by "one way cyclic" test using the new direct shear apparatus. Examples of calculating design strength parameters by the authors' method ( Mitachi et al., 1999) using the test results obtained by new testing apparatus on the specimens sampled from the slip surface of actual landslide sites are also presented.
1. Test Apparatus The shear test apparatus newly developed by the present authors is made up of a main unit of the apparatus, a personal computer and a controller box. Figure 1 gives an overview of the apparatus which can measure the strengths of cohesive soils corresponding to "peak" state for undisturbed samples and "fully softend" and "residual" states for remolded normally consolidated samples. Vertical loading system In order to make the structure of apparatus as simple as possible and to improve the precision of testing, a displacement controlling system ( with a digital servomotor) is used for vertical loading. The minimum control of vertical displacement is 1.525 x 10" mm. Construction of Shear box Specimens with two shapes can be tested; 6 cm 715
square pieces and 6 cm diameter circular pieces. The square shear box has small high-rigidity load cells as shown in Figure 2 for measuring the frictional forces between the upper and lower shear boxes and the specimen shear surface. The circular shear box has teflon-coated inside wall to reduce frictional forces between sample and shear box and has a structure that allows the bottom of the shear box to be moved up and down to match the landslide surface of the specimen sampled fiom the site with the level of the contact surface between two halves of the box. Measuring system of vertical and shear forces Vertical and shear forces are measured with high-rigidity load cells equipped as shown in Figurel. In order to measure correctly the vertical force applied on the shear surface of specimen, the apparatus is structured so that the vertical force on the shear surface can be measured without being influenced by the peripheral surface fiiction force exerted inside the shear box ( Shibuya et a1.,1993 ) . The shear force acting along the sliding surface of the specimen is obtained by subtracting the frictional force measured by the compact high-rigidity load cells installed in the shear box as shown in Figure 2 from the overall shear force measured by the load cell equipped in front of the pushing rod.
shear box and the specimen shear surface, direct shear tests were carried out under constant pressure condition. The material used for the test was NSF clay ( p s=2.76g/cm3, LL=54%, Ip=26) which was preconsolidated for ten days at 100 kPa, and then trimmed into 6 cm square and 2 cm high specimen. The consolidation pressure was set to 300 kPa . Shear test was started after discontinuing consolidation by "3t method" standardized by Japanese Geotechnical Society ( 1990) and the rate of shear was set as 0.02 mdmin. The opening between the two halves of the shear box was set as 0.2 mm and the maximum application of horizontal displacement during shear was set as 6 mm. ( 2 ) Cyclic direct shear test Cyclic direct shear tests were carried out using kaolin clay under consolidated constant pressure condition of 200, 300 and 400 kPa. After discontinuing consolidation by "3t method" , cyclic direst shear test was carried out. The rate of shear was kept as 0.02 mm/min. up to a horizontal displacement of 3 mm,then changed to 0.17 mm/min. After reached to the horizontal displacement of 6 mm, the lower box was moved to the reverse direction until the horizontal displacement reached to -6mm. In this test program, the kaolin clay sample preconsolidated from a slurry at 100 kPa was used by trimming them into cylindrical specimens with a diameter of 6 cm and 2 cm height.
2.2 Test Results and Discussion
Figure 2 Measuring system of vertical loads and the frictional forces between the shear boxes and the specimen shear surface.
2. Materials and Test Results 2.1 Materials and Test Method ( 1 ) Evaluation of friction between the shear box and the specimen shear surface In order to clarify the relationship between horizontal displacement and the friction between the
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( 1 ) Frictional force between shear box and test specimen shear surface Figure 3 shows the horizontal displacement ( HD) versus frictional force ( Fr between the specimen shear surface and the shear box relationship in direct shear tests under constant pressure condition. The vertical loads ( VL) acting on the end surface of shear box through the soil specimen which are measured by the load cells installed inside the shear box as shown in Fig.2 and the overall shear force (SF) measured by the load cell equipped in front of the pushing rod are also illustrated in the figure. The frictional forces ( Fr) measured by the upper and lower load cells increase with the progress of shear as shown in Figure 3 and are averaged 5.19'0 of the overall shear force (SF) when the horizontal displacement is 6 mm at which the reduction of shear surface is 10%. The frictional force and vertical force measured by the lower load cell are greater than the corresponding values measured by the upper load cell. In this test apparatus, vertical load is applied
by digital servomotor equipped as shown in Fig.1 and the upper box is fixed in placc while the lower box is movable. When the lower shear box moves with the progress of shear loading, the vertical force applied by the loading plate is transmitted to the lower shear box through the soil specimen. Therefore, the greater the vertical force acting on the surface of lower box is, the greater the frictional force (Fr) acting on the same surface. As the frictional force between shear box and test specimen shear surface is rather small comparing with the reduction of shear surface as mentioned above, no correction was done for the measured shear force, and the shear stress was calculated by using original sectional area of specimen for all the data mentioned in the following articles.
Figure 4 The relationship between shear stress (r and hor i zonta I d i sp I acement (HD) f o r the cycl i c shear t e s t under constant pressure cond i t i on.
F i g u r e 3 The horizontal displacement(HD) versus f r i c t i o n a l force (Fr) between the specimen shear surface and the shear box r e l a t i o n s h i p .
(2) Cyclic direct shear test Figure 4 shows the relationship between shear stress (T) and horizontal displacement (HD) for the cyclic shear test under constant pressure condition. Shear stress exhibits a peak at a horizontal displacement of 4-5 rnm and converges to a residual state after 2 cycles of shear. Maximum shear stress versus vertical stress relationship obtained from constant pressure test series is shown as Figure 5. The straight line representing shear stress versus vertical stress relationship for the residual state passes through the origin. ( 3 ) Comparison with other shear tests Figure 6 shows the relationship between the shear stress ( z ) and horizontal displacement ( HD) obtained by a "one-way cyclic" direct shear test, in which shear stress is always applied repeatedly in
Figure 5 Maximum shear stress versus v e r t i c a l stress r e l a t i o n s h i p obtained from constant pressure t e s t ser i es.
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the same direction, and the vertical stress is unloaded during the shear box moves in the reverse direction. This series of test was carried out by using NSF clay specimen. The value of the shear stress for the residual state obtained by "one-way cyclic" test result is almost the same as obtained ordinary cyclic test results. Table 1 shows the results of ring shear test and cyclic direct shear test on the clay specimen sampled from the sliding surface of the Yamagata Prefecture Dozangawa landslide ( Igarashi et a1.,1997) . The strength parameters obtained from
straight lines PQ and AB gives the design strength parameters, where the points A and B are plotted using the data listed in Table 2 and are corresponding to "fully softened" and "residualf' state strength parameters. By applying the authors' method of determining design strength parameters, it becomes possible to combination limit the range of changing (c', $'I along PQ line into the possible combination based on the material strength characteristics. Table 2 The results of cyclic direct shear t e s t s performed under constant vertical pressure condition on t h e clay sampled from t h e Dozangawa landslide. Sample Fully Softend Strength Residual Strength
Dozangawa Slip Surface Clay [kPa] 36.4 d J ' S [" 1 13.0 c'r [kPa] 0.0
C*S
d ' r ["
1
2.3
Figure 6 The relationship between the shear stress ( -c 1 and horizontal displacement (HD) obtained by a "one-way cyc I i c" d i rect shear test.
Ring Shear Test Cyclic Direct Shear Test
c'r CkPa] 0
0
T
O
3.2 2.3
cyclic direct shear test is even lower than the results obtained from ring shear test. 2.3 Example of Calculation of Landslide Average Strength Parameter
The following are examples of calculating the design strength parameters according to the authors' method ( Mitachi et. al., 1999) using the test data obtained by direct shear test apparatus newly developed by the present authors. ( 1) Strength parameters for Dozangawa landslide Table 2 shows the results of cyclic direct shear tests performed under constant vertical pressure condition on the clay sampled from the sliding surface of the Dozangawa landslide, which occurred in Okura Village in Mogami County, Yamagata Prefecture. From the results of stability calculation by Fellenius method on the main section of the landslide, the analytically possible combination of strength parameters (c', 4') for a current safety factor Fs of 1.0 is expressed as the linear equation of c' = 681.9tan$' + 109.1. When this relationship is plotted as the c-tan+ graph in Figure 7, a straight line PQ is obtained and CO = 18.0 (kPa) and tan40 =0.13 are given as strength parameters as the intersection of line PQ to the axes of coordinates. The intersection point C of the two 718
Figure 7 The detemining strength parameters for Dozangawa I ands I i de.
(2 ) Strength parameters for Yokote landslide Table 3 shows the results of cyclic direct shear tests on the clay sampled from a place near the site of the landslide that occurred at a road construction site in Yokote City, Akita Prefecture. From the results of stability calculations by Fellenius method on the main section for this landslide, the analytically possible combination of strength parameter (cl, 4') for a current safety factor Fs of 1.0 is expressed as the linear equation c' = - 58.7 tan+' +36.4 which is represented by the straight line PQ in Figure 8. The points A and B in Figure 8 correspond to fully softened and residual state strength parameters. The strength parameters plotted as point D which correspond to the values for peak state strength were obtained on undisturbed clay
specimen and the surfaces of the upper and lower boxes can be measured by the load cells installed inside the shear boxes. Peak strength parameters can be obtained from the monotonic loadig shear test with this apparatus and using undisturbed clay specimen sampled from the site. Strength parameters corresponding to fully softened and residual states can also be obtained from the cyclic shear test with this apparatus and using the specimen fully remolded and preconsolidated from the state of slurry. The two or three sets of strength parameters ( cp, (b ,) for peak strength state, (cs, (b for fully softened state and (cr, (b r > for residual state, respectively obtained from the tests mentioned above, can be used for the method of determination of design strength parameters proposed by the authors. Case studies for two sites of landslide proved the suitability of the strength parameters determined from the method based on the experimental data obtained by new designed cyclic direct shear test apparatus.
Table 3 The results of cyclic direct shear tests on the clay sampled from the Yokote landslide.
REFERENCES Figure 8 The detemining strength parameters for Yokote I ands I i de.
specimens sampled near the site of the landslide slip surface. According to the authors' method of determining design strength parameters, the points corresponding to peak strength of overconsolidated states for different overconsolidation ratios are plotted on the prolongation of AD line. In this case, the design strength parameters are given by the intersection point C of the prolongation line AD and the line PQ which was obtained as mentioned above.
3. Concluding Remarks The suitability of strength parameters for stability calculation is very much important in evaluating landslide slope stability. In this study, a new cyclic direct shear apparatus which can even be brought in the field was designed for the purpose of quick and rational determination of design strength parameters. The features of this apparatus are: 1) the apparatus can be brought in the field due to its compactness and light weight, 2) cyclic shear test with any cycle and amount of shear displacement can be performed automatically, 3) since normal force is controlled by high accuracy digital servo motor system, both constant pressure test and constant volume test can easily be performed, and 4) friction force between sliding surface of the 719
1) Mitachi, T. and M. Okawara 1999.Method for determining design strength parameters for landslide slope stability analysis, Proceedings of International Symposium on Slope Stability Engineering : Geotechnical and Geoenvironmental Aspects. 2)Mitachi, T., A.Sano and M. Okawara 1996.The relationships between strength parameters obtained from laboratory shear tests and those for use of stability calculation, Proc. of 35th Annual Convention of Japan Landslide Society, pp.345-348 (in Japanese) . 3) Shibuya, S., T.Mitachi, A. Kitajima and M.Takada 1993. Strength of sand as observed in a newly developed direct shear box apparatus, Bulletin of the Faculty of Engineering, Hokkaido University No. 166, pp. 1- I I . 4) Igarashi, IS. and S. Yamashina 1997. On the Dozangawa Landslide, Proceedings of Japan Landslide Society, pp.55-56 ( in Japanese) .
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Slope Stability Engineering, Yagi, Yamagami& Jiang (cj 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Strength and deformation characteristics of clay subjected to pore water pressure increment T.Umezaki Deparment of Civil Engineering, Shinshu Universit)! Nugano, Jupun
M.Suzulu & TYamamoto Department of Civil Engineering, Yurnuguchi University, Uji, .Iupan
ABSTRACT In order to clarify the shear behavior of clay subjected to an increase in pore water pressure, a triaxial slice shear test apparatus was developed and a series of tests was performed on kaolin clay. The triaxial slice shear tests can impose a large shear strain on the specimen and a pore water pressure in the specimen can be uniformly increased. After the clay specimen reaches the residual state, the shear stress decreases along the strength line at the residual state accompanying the increase in pore water pressure.
1 INTRODUCTION
Landslides occur frequently during the heavy rain and snow melting seasons. Based on the field investigations, Ogawa et al. (1987) demonstrated that the groundwater level and the pore water pressure in landslide areas increased remarkably during these seasons. They also proposed that the strength parameter of soil subjected to the increase in pore water pressure should be used to analyze the stability of a reactivated landslide. Katagiri et al. (1996) examined the deformation characteristics of clay subjected to the increase in pore water pressure using a triaxial test apparatus. However it takes a very long time to equalize the pore water pressure in a cylindrical specimen used in such a test. The conditions of stress and strain are respectively different at each part of the specimen during the equalization of pore water pressure. Eigenbrod et al. (1987) and Tokida et al. (1987) also have researched this topic. Figure 1 shows the mechanical condition of a soil element on a slip surface during a rise of groundwater level within a slope. The authors suppose that the soil element on the slip surface in a reactivated landslide, in which the stress condition on the slip surface has almost reached the residual state by a large deformation, is subjected to 721
fluctuations in the pore water pressure due to rainfall. Ring shear tests simulating this phenomenon have been performed on clay (Suzuki et al. 1999). Instead of increasing the pore water pressure in the specimen directly, the total normal stress equivalent to the increment in pore water pressure is decreased during drained shearing in the ring shear tests just described. In order to simulate a slope failure accompanying the increase in pore water pressure, a new triaxial slice shear test apparatus was developed.
Fig.1 Schematic diagram of soil element on a slip surface during a rise of groundwater level within a slope
Photo.1 Specimen in triaxial slice shear test Fig.2 Outline of triaxial slice shear test apparatus
A triaxial slice shear test is able to increase the pore water pressure in the specimen uniformly for a short time. This paper describes the shear behavior of the clay subjected to the increase in pore water pressure after reaching the residual state from the viewpoint of the effective stress. 2 TRIAXIAL SLICE SHEAR TEST
2.1 Test apparatus The triaxial slice shear test was developed in the Norwegian Geotechnical Institute. Shibata et al. (1968) introduced details of this apparatus and examined its applicability. Umezaki et al. (1992) recently used a triaxial slice shear test apparatus to examine frictional characteristics between clay and steel. The main features of the triaxial slice shear test are summarized as follows. 1) A specimen in the shape of a slice is deformed in the mode of the simple shear. 2) A large shear deformation can be given to the specimen. 3) The pore water pressure in the specimen can be controlled directly. Figure 2 schematically shows the triaxial slice shear test apparatus. A specimen and a filter paper are placed on a pedestal at an incline of 45" inside the cell (see Photo.1). The specimen was cut off a column of 50 mm in diameter and formed a slice of' 10 mm in thickness and 45" in angle. The end cap is set to move in the horizontal direction smoothly
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Fig.3
Stress and deformation conditions of specimen
through the ball bearings. The axial force is loaded to the specimen, so that the shear stress and the normal stress act on the boundary surfaces of the specimen. As the specimen is deformed in the mode of the simple shear, the end cap moves downward in the direction of 45" . The horizontal and vertical displacements are measured with a clip gauge and a dial gauge, respectively. The drainage routes of pore water from the specimen are connected with two double burettes and a transducer for measuring the volumetric change and the pore water pressure in the specimen, respectively. The pore water pressure in the specimen is increased by imposing an additional back pressure on the upper surface of the specimen. The operation is handled with an air regulator and does not interfere with the values of cell and back
Table 2 Physical property of kaolin
Table 1 Test cases and symbols used in figures
Specific gravity of soil particles, G, Liquid limit, wL Plastic Limit, wp Plasticity index, I, Clay Fraction, F, ( < 2 M m)
“1 :Isotoropic consolidation stress
“2:Effeclive confining stress
*3:OCR=O’,
10’3
‘4:Undrained shear test
“5:Pore water pressure increment test
A transmission of pore water pressure On the bottom surface of the specimen is measured with time. 2.2 Stress and deformation conditions of specimen Figure 3 schematically shows the conditions of stress and deformation of the specimen in the shear process. The isotropically confining stress, CJ 3, acts on the sides of the specimen. The shear stress, Z , and the total normal stress, O N , act on the upper and lower surfaces of the specimen and both are simply calculated from the axial additional force, A P. The stress and strain parameters are defined as follows:
I I I I I
2.724 75.6 o/o
36.3 5% 39.3 70.0 5%
150 96 and one-dimensionally consolidated under a vertical pressure of 49 kPa for seven days. The procedure of undrained shear test is the same as that of conventional triaxial compression test. On the other hand, the procedure of pore water pressure increment test is summarized as follows. 1) The specimen which is isotropically consolidated is undrained-sheared at axial strain rate 0.1 %/min until the maximum shear strain, 7 171i,x of 30 %. 2) The water pressure, which is equal to the excess pore water pressure generated inside the specimen, is imposed on the upper surface of the specimen through the double burette. 3) While the specimen is drained-sheared until 7 max + 50 %, the pore water pressure in the specimen is increasing at a constant rate. The ratio of increase in pore water pressure is 1.96 kPa/min from considering the case histories according to Tokida et al. (1987).
3 RESULTS AND DISCUSSIONS
COS'^ + CJ
(l)
3.1 Undrained shear test
z =( A P /A) sin 0cos 0
(2)
Figure 4 shows the relationship between the vertical displacement, A d,, and the horizontal displacement, Ad,, of the specimen. These results are obtained from undrained shear tests on both normally consolidated and overconsolidated clays. As the specimen is deformed, the vertical displacemen, becomes larger than the horizontal displacement. However, it is regarded that the specimen is approximately deformed in the mode of the simple shear, because the data points are plotted near the Ad, = Ad, line.
CJ
7
=(AP /A)
= ( A d,j/H ‘cos fl ) X 100
(%)
(3)
where ’ * D2) and mdx are a cross sectional area and a maximum shear strain of the specimen, shown in Fig.3.
Then D,
and
are
2.3 Test procedure Both undrained shear test and pore water pressure increment test are performed on clay. Two kinds of test cases are shown in Table 1. The sample is kaolin and its physical property is listed in Table 2. The sample is thoroughly mixed with a water content of
Figure 5 shows the relationships of the maximum shear strain, 7 max, to the shear stress, Z , and the excess pore water pressure, n u , respectively. The triaxial slice shear test can impose a very large shear 723
Fig.6 Effective stress paths of clay during undrained shear
In all cases, the shear stress and the excess pore water pressure become constant values at 7,1,dx 2 30 %, respectively. Therefore, we conclude that the conditions of stress and strain of a specimen reach a residual state at 7 ,ll,x 2 30 %. Figure 6 shows the relationship between the shear stress, z , and the effective normal stress, 0 'N. All effective stress paths move toward a strength line at the residual state. The internal friction angle and the cohesion in terms of the effective stress are d) '\ = 19.3" and c', = 0 kPa, respectively. On the other hand, d~ ' = 18.8" and c'= 0 kPa are obtained from conventional triaxial compression tests. In the cases of normally consolidated clay under 0 7 3 5 98 kPa, the shapes of the stress path are similar to that of overconsolidated clay. It is considered that the clay's behavior is affected by one-dimensional preconsolidation under a vertical stress of 49 kPa.
Fig.4 Relationship between vertical displacement and horizontal displacement during undrained shear
3.2 Pore water pressure increment test
Fig.5 Undrained shear behavior of clay in triaxial slice shear test
strain on the specimen. The shear stress increases with increasing the maximum shear strain. In the only case of normally consolidated clay under 0 7 3 = 294 kPa, the shear stress becomes a maximum value at 7 m,x k 10 %. Then the shear stress seems to reach a constant value at Y,,,, 2 30 %. The excess pore water pressure also seems to reach a constant value at Y,,,,,2 30 5%. It is noted that a similar tendency is obtained from the test on overconsolidated clay under 0 71 = 49 kPa(OCR=8). 724
Figure 7 shows the relationship between the loading value, uL, and the measured value, uM, of the pore water pressure immediately after the pore water pressure is increased. The value of uL is almost equal to that of U,. Thus the pore water pressure in the specimen can be uniformly increased. Figure 8 shows the relationship between the vertical displacement, a d,, and the horizontal displacement, d,, of the specimen during an increase in pore water pressure. As the specimen is deformed, the vertical displacement becomes almost equal to the horizontal displacement. This shows that the specimen is approximately deformed in the mode of the simple shear during the increase in pore water pressure.
Fig.7 Response of pore water pressure in specimen
Fig.9 Shear behavior of normally consolidated and overconsolidated clays during increase in pore water pressure
Fig.8 Relationship between vertical displacement and horizontal displacement during increase in pore water pressure
Figure 9 shows the relationships of the maximum shear strain, ?' m d h , to the shear stress, E , the excess pore water pressure, Au, and the volumetric strain, E \ , respectively. These results are obtained from pore water pressure increment tests on both normally consolidated and overconsolidated clays. As shown in Fig.5, both shear stress and excess pore water pressure reach the residual state where the maximum shear strain becomes about 30 %. In the range over ?' = 30 %, as the excess pore water pressure increases monotonously, the shear stress remarkably decreases and the spccimen simultaneously swells. These shear behaviors seem to be independent oi the magnitude of the effective normal stress and overconsolidation ratio. Figure 10 shows the relationship between the
Fig.10 Effective stress paths of normally consolidated and overconsolidated clays during increase in pore water pressure
shear stress, E , and the effective normal stress, (7 ". The effective stress paths move toward the strength line at the residual state. After the clay specimen reaches the residual state, the shear stress decreases along the strength line accompanying the increase in pore water pressure. These experimental results agree well with the ring shear test results (Suzuki et al. 1999). Therefore, the stress condition of the clay specimen, which has once reached the residual state, moved along the strength line with changing the effective normal stress. Figure 11 shows the relationship between the
,,,dX
725
for supervising this study and Mr. Toshiyuki Kugai, Mr. Chikara Nagase & Mr. Tomoya Yamajo for the experimental assistance.
Normalized effective normal stress dN/ cfNO
REFERENCES Eigenbrod, K.D., Burak, J. -P. & Graham, J. 1987. Drained deformation and failure due to cyclic pore pressure in soft natural clay at low stress, Canadian Geotechnical Journal, Vo1.24, pp.208215. Katagiri, M. & Imai, G. 1996. Deformation characteristics of a saturated cohesive soil subjected to increase in pore pressure, Soils und Foundations, Vo1.36, No.3, pp.1-12. Ogawa, S., Ikeda, T., Kamei, T. & Wdda, T. 1087. Field investigations on seasonal variations of the ground water level and pore water pressure in landslide areas, Soils and Foundutions, Vo1.27, No.1, pp.50-60. Shibata, T. & Hoshino, M. 1968. Triaxial slice shear test on clay, Euchi-to-Kiso, The Japanese Geotechnical Society, Vol.16, No.1, pp.3-9 (in Japanese). Suzuki, M., Umezaki, T. & Yamamoto, T. 1999. Shear behavior of clay subjected to change of normal stress, International Syrnposium on Slope Stability Engineering: Geotechnicul and Geoenvironmental Aspects, IS-SHIKOKU'90, (in press). Tokida, M., Hashimoto, M., Ikeda, T., Ogawa, S. & Kamei, T. 1987. Influence of the increase in pore pressure on the shear characteristics of cohesive soil subjected to the stress history, Proc. of the 22nd Japan National Conference on Geotechnical Engineering, pp.467-468 (in Japanese). lJmezaki, T., Ochiai, H., Hayashi, S. & Uchida, K. 1992. Friction properties between clay and steel sheet-pile, Technology Reports of Kyushu University, Vo1.65, No.6, pp.565-572 (in Japanese).
Fig.11 Relationship between the volumetric strain and the normalized effective normal stress during the increase in pore water pressure in residual state
volumetric strain, E ", and the normalized effective normal stress, CJ 'N / CT 'NO. Here CT 'NO is the effective normal stress just before increasing the pore water pressure in the specimen. The & - 0 " / (J ' N O curves seem not to be linear. The swelling behavior of the clay specimen during the increase in pore water pressure seems to be determined only by the normalized effective normal stress. 4 CONCLUSIONS
The main conclusions are summarized as follows: 1. Triaxial slice shear tests can impose a large shear strain on the specimen, so that the strength line at the residual state can be accurately determined. Using a new developed test apparatus, the pore water pressure in the clay specimen can be uniformly increased. After the clay specimen reaches the residual state, the shear stress decreases along the strength line accompanying the change of the pore water pressure. During the increase in pore water pressure, the volumetric change of the clay specimen seems to be uniquely determined only by the normalized effective normal stress. ACKNOWLEDGMENT The authors express their sincere thanks to Emeritus Professor Hiroshi Kawakami of Shinshu University 726
Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Parameters for curvilineared residual strength envelope S.Gibo Faculty of Agriculture, University of the Ryukyus, Japan
S. Nakamura United Graduate School of Agricultural Sciences, Kagoshima University, Japan
ABSTRACT: The residual friction coefficient was found to decrease with the increasing effective normal stress at the lower half of the effective normal stress, while it was constant at the higher half. Based on this finding, the curvilinered residual strength envelope was divided into two parts: the lower half and the higher half of the effective normal stress. The residual strength parameters were determined at each part, c, was not zero at the lower half of the effective normal stress while it was zero at the higher half, and 4 ,was greater at the lower half than at the higher half. The proposed method might be useful and rational to determine the residual strength parameters from the curvilineared residual strength envelope depending on the magnitude of the effective normal stress. 1 INTRODUCTION Residual strength, the minimum drainage strength acted on the oriented surface of clay mineral particles, is indispensable to the evaluation of stability of the reactivated landslide and the first-time slide occurring in the bedrock with geological discontinuities. The determination of residual strength parameters is very much important, because the suitability of these parameters influences the slope stability analysis results and the optional selection for a suitable countermeasure against landslide. However, the fixed strength parameters can not be given owing to the curvilineared line of the residual strength envelope indicated in the test (Skempton 1964, Bishop et al. 1971, Gibo 1978 1983 1985, Hawkins et al. 1986). The curvature of residual strength envelope does not only depend on the type of landslide soils but also on the magnitude of normal stress (Gibo et al. 1987). Thus, understanding the relationship among the above parameters is very much important for developing a determination method for residual strength parameters.
Figure 1. At the lower effective normal stress decreased with the increasing o level, z 4 o On the other hand, at the higher effective normal stress level, it showed a constant value. And the residual strength envelope curved at the lower effective normal stress level. According to Skempton (1964), the residual cohesion (c,) was nearly to zero in the determination of the residual strength parameters of London Clay. Bishop et al. (1971) considered that the residual angle of shearing resistance ( 4 varies depending on the magnitude of the effective normal stress provided the residual cohesion is zero. On the other hand, Gibo (1987) paid o n’ relationship and the attention to the z ) o ,,’
-
2 CURVATURE OF RESIDUAL STRENGTH
ENVELOPES Hawkins et al. (1986) reported the relationships between residual friction coefficient ( z 4 CT n’ ) and effective normal stress ( CT n ’ ), and residual strength ( z ,) and effective normal stress ( CJ ,,’ ) in
Figure 1 Residual strength envelopes and definitions (Hawkins et a1.,1986)
727
condition of residual shear surface, confirmed the existence of c,. Therefore, it is clear that the strength parameters vary depending on how to estimate the test results and how to draw the strength envelope line.
3 FOUNDEMENTAL THINKING
Gib0 et al. (1987) discussed the residual friction coefficient, z ,/ o ,,’ , and the orientation index of smectite particles on the shear surface as a function of the effective normal stress (Figure 2). SD-3 and T were soil samples from slip surfaces on which well-defined slickenside were observed. SD-1 was soil sample obtained from non-slip surface. The o ,,’ curves relationship between the z ,/ o ,,’ and the orientation index- o ,,’ curves clearly indicated the influence of the orientation of smectite particles on the residual strength. The residual friction coefficient was inversely related to the orientation index of smectite particles. The orientation of smectite particles on the shear surface decreased the residual strength, and this orientation effect was revealed more obviously at effective normal stress below 100kPa. Also, just as shown clearly in Figure I., the curvature of the z ,/ o ,,’ o ,,’ relation is reflected clearly in the z r- o ,,’ relation. At the higher effective normal stress level where the value of z ,/ o ,,’ was constant, z ,/ o ,,’ was equal to tan 41 and c, became zero. On the other hand, at the lower effective normal stress level, z ,/ o ,,’ was not a constant value, then c,/ o ,,’ f 0, i.e. c, f 0. It is obvious that the residual strength varies with the orientation index on shearing surface. At the lower effective normal stress, the effects of residual cohesion on shearing strength can not be neglected.
Figure 2. Residual friction coefficients and orientation indices of smectite on the shear surface as a function of effective nonnal stress (Gibo et al., 1987)
-
-
4 SOIL SAMPLES AND THEIR PHYSICAL AND MINERALOGICAL PROPERTIES The soil samples were collected from the landslide of Taiwan (Gibo et al. 1997) and the Kamenose landslide (Gibo et al. 1987, Hayashi 1992). The liquid and plastic limits of the Taiwan sample were 26.5 and 15.7%, respectively and the clay fraction content was 17.2%. For the Kamenose sample, the liquid and plastic limits were 114.0 and 50.0%, respectively and the clay fraction content was 73.2% (Table 1). The Kamenose sample contained extremely high proportion of smectite having orientation characteristic and a high swelling property (Table 2). Because smectite particles greatly contribute to the formation of the shear surface, a low residual strength could be expected (Egashira & Gibo 1988).
Otherwise, Taiwan sample contained only quartz, mica and chlorite but no smectite, a high residual strength could be expected. 5 DETERMINATION OF RESIDUAL STRENGTH PARAMETERS The residual strength of soil samples was measured using the ring-shear apparatus designed by Gibo (Gibo, 1994). The soil samples passed through a 420- p m sieve, were packed in a shear box with 100 and 60 mm in outer and inner diameters, respectively. The samples were then subjected to shear in an immersed condition until the residual state was attained. To achieve the full dissipation of excess pore-water pressure, the rate of shear displacement in the residual state was set at 0.01 mm/min. The Taiwan sample contained a lot of silt and fine sand, and resulted in a greater residual friction coefficient compared with the Kamenose one. The o ,,’ relationship clearly indicated the z ,/ 0 ,,’ influence of the orientation of clay particles on the residual shear surface. For each sample, z CT ,,’ gradually decreased with the increased o ,, and finally it approached a constant value (Figure 3). In Figure 4a, the residual strength line was drawn provided residual cohesion is zero (Skempton 1964). The residual strength parameters were estimated to be cr=O kPa and4I,=26.O0 . However, at the low normal stress level, the strength was plotted above the line. The residual friction coefficient was found to decrease with the increasing effective normal stress at the lower half of the effective normal stress, while it was constant at the higher half. Based on this finding, the curvilinered residual strength envelope was divided into two parts: the lower half and the higher half of the effective normal stress. The residual strength parameters were determined at each part. Concerning the differentiation of lower and higher levels, the effective normal stresses were divided at certain effective
728
-
4
Table 1. Physical properties of soil sanples(> Qr > Fd, Mi, Kt
Sm:smectite, Mi:miea, Ktkaorinite, Ch:chrolite, Qr:qtiarh, Fd:feldsper
bf Estimated by new riietliod
Figure 4. Residual shear strength eiivclopes and strength parameters for the Taiwan soil sample
b)Kanienose
Figure 3. Relationship betwecii residual friction coefficient and effective nornial stress Figure 5 . ResiduaI strengli envelopes and strength parameters for the Kanicnose soif sample 729
tical method, because the a r varies depending on the magnitude of CT *' . The method of Skempton was convenient to use, but this method did not consider the curvature of residual strength envelope and the existence of the residual cohesion at low effective normal stress. Besides, these residual strength parameters might be underestimated, which could induce a miss-evaluation of slope stability. Therefore, it could not be considered as a good method. From the discussion of different methods it can be concluded that the proposed method is operational but useful and rational to determine the residual strength parameters from the curvilineared strength envelope depending on the magnitude of the effective normal stress.
REFERENCES Bishop, A.W., Green. G.E., Garga. V.K., Andresen. A. & Brown, J.D. 1971. A new ring shear apparatus and its application to the nieasurement of residual strength. Geotechiiique. 21(4): 273-328. Egashira, K. & Gibo, S. 1988. Colloid-chemical and niineralogical differences of smectites taken from argillized layers, both from within and outside the slip surfaces in the Kainenose landslide. Applied Clay Sci.3: 253-262. Gibo, S. 1983. Measurement of residual strength of Shiinajiri Mudstone and evaluation of the results -residual strength characteristics of materials in and close to the slip surface (1)-. Tram. JSIDRE 104: 6 1-68. (in Japanese with English abstract) Gibo, S. 1985. The ring shear bchavior and residual strength, Proc. 4th Int. CoiiJ arid Field IVorksliop oil Laiiclslides, To&~0:283-288. Gibo, S. 1987. Shear strength parameters required for evaluation of stability of slopes. Tszrchi-to-Kiso JGS 35(11): 27-32. (in Japanese) Gibo, S. 1994. Ring shear apparatus for measuring residual strengths and it's ineasurement accuracy. ,Jl. Jpn. Laiddide Soc. 3 l(3): 24-30. (in Japanese with English abstract) Gibo, S., Chen, H. H., Egashira, K., Hayashi, Y. & Zliou, Y. 1997. Residual strength characteristics of soil from the reactivated landslide occurred at the national road across the middle part of Taiwan. Jl. Jpi?. Lai?clslide Soc. 34(2): 50-56. (in Japanese with English abstract) Gibo, S.. Egashira, K. & Ohtsubo, M. 1987. Residual strength of sinectite-dominated soils from the Kamenose landslide in Japan, Caii. Geotecli. Jl. 24(3): 456-462. Hawkins, A. W. & Privett, K.D. 1986. Rcsidual strength. Does BS5930 Help or Hinder?. Geol. Soc. Eiigiiieering Geologp Special Pttblicatioii 2: 279-282. Hayashi, Y., Higaki, D. & Ishizuka, T. 1992. Structure of slip surface formed by rock block slide. Landslides Glissemeiits cle terraiii. DA VID FI. BELL, Proc. 6th Iiit. Syiirp. ; 127-132. Skempton, A. W. 1964. Long-term stability of clay slopes. Geotecliiiique 14(2): 77-101.
730
Slope Stability Engineering, Yagi, Yamagami& Jiang (c) 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Pore water pressure loading tests of a clay S.Ohtsuka, Y. Miyata & H.Toyota Departnierzt of Civil and Environmental Engineering, Nagaoka University of Technolog), Japan
ABSTRACT: The failure mechanism of landslide was investigated with pore water pressure loading test of a clay under the condition of constant deviator stress application. Through the test results, the followings were obtained :(1)Upper and lower yield limits in terms of effective stress were measured for generation of shear deformation. They could give the alternative design parameters. (2)Shear deformation of soil was found to proceed with water migration so that the phenomenon appeared slowly with time. As the effective stress passed the lower limit, shear deformation proceeded with longer time due to dilation behavior caused by plastic deformation. It showed the possible reason that the landslide developed with long time in a different way from the mudslide.
1 INTRODUCTION Landslides in Niigata Prefecture, Japan have been investigated by many researchers. Various types of landslides can be found elsewhere, however, the characteristics of them are summarized in this study as follows:( 1)Landslide occurred due to the increasing pore water pressure in slope by rain and/or melted snow. (2)Slope had the failure potential of shear stress induced by equilibrium in force. (3)Gentle slope failed repeatedly. And landslide took place slowly with time in the different way from mudslide. Pore water pressure loading test was conducted in this study to make clear the failure mechanism of landslide and establish a rational design method based on the obtained failure mechanism of landslide. The deviator stress was applied as constant to simulate the stress state of a soil in slope. Katagiri and Imai(l996) conducted a series of pore water pressure loading tests on a saturated cohesive soil. They investigated a deformation behavior of an overconsolidated soil, especially the yield surface of soil in terms of effective stress. Ogawa(l986) performed the ring shear test by changing the confining stress under the condition of constant shear stress. These researches focussed on the shear resistance of a soil in landslide. Ogawa(l986) found two thresholds in confining stress for the soil to generate shear deformation. They were defined as the upper and lower yield limits, respectively. As the confining stress attained to the lower yield limit, the deformation was observed to increase due to plastic 731
deformation. At the upper yield limit of confining stress, the soil failed unconfinedly. Although the shear strength property of overconsolidated soil has been investigated, it has not been clear that landslides develop slowly with time. Asaoka has pointed out the importance of taking a soil-water coupling behavior into account to understand the shear behavior of clay(Asaoka et al., 1997). This study investigated the soil behavior in pore water pressure loading test from the viewpoint of the soil-water coupling concept to make clear the failure mechanism of landslide. 2 PROCEDURE OF TEST Nagaoka clay passed a filter of 42.5p.m was employed for pore water pressure loading test in this study. The physical properties of the clay are exhibited in Table.l. It was remolded well and consolidated in one dimensional pre-consolidation apparatus under the vertical stress of 47kPa. The soil specimen was set in the triaxial testing apparatus and isotropically consolidated with 200kPa. After five hours since the completion of consolidation, the soil was sheared with a specific deviator stress under the undrained condition. It was defined as the initial state in this study. Applied deviator stress was controlled constant during the successive pore water pressure loading test. In pore water pressure loading, the pore water pressure was loaded at the bottom of soil specimen and was measured at the top of soil specimen. Pore water pressure was enforcedly
increased by IUkPa in a step. After confirming the transmission of pore water pressure from the bottom to the top, pore water pressure was increased step by step. From the undrained shear test, the magnitudes of deviator stress were determined as 50, 75, 100, IZUkPa. Though the employed soil was normally consolidated, the stress state of soil quickly moved to an overconsolidated state due to pore water pressure loading. The expected soil behavior is similar to that of an overconsolidated clay. Table 1. Physical properties of soil Compression index A Swelling index K Critical state parameter M Specific gravity G, Liquid limit w, Plastic limit w,
0.111 0.025 1.53 2.61 49.80% 35.10%
mission time, therefore, becomes the almost same one. On the contrary, it becomes clearly longer with the increase in pore water pressure. This tendency is obvious and the transmission time is found very long near the critical pore water pressure. It is thought to be caused by the generation of plastic deformation. Based on the consolidation theory the coefficient of consolidation, c, is described as c, = k / r n , y , where k and rn, express the coefficients of permeability and volume compressibility, respectively. As rn, increases due to plastic deformation, c, decreases and then, the transmission time gets longer. The overconsolidated soil dilates largely with shear deformation. In progress of shear deformation, the soil requires a supply of water to expand. Because of low coefficient of permeability the shear deformation of soil proceeds progressively with water supply. This mechanism makes the transmission time of pore water pressure longer as the enforced pore water pressure closes to the critical magnitude. 3.2 Overview of failure
3 PORE WATER PRESSURE LOADING TEST
The soil specimen was controlled to be uniform as a soil element, however, it was observed to become non-uniform by the generation of shear band in the specimen. Fig.2 represents the schematic figure of failure mode and the water content distribution after the test. It clearly shows that the uniformity of soil specimen has been broken.
3.1 Transmission of pore water pressure
In pore water pressure loading test, the pore water pressure was measured at the top of soil specimen. Fig.1 shows the measured pore water pressures in time. In the figure two cases for the deviator stresses of 50 and IUUkPa are exhibited as typical cases. Pore water pressure at the top of soil specimen increased obviously with time. The increase tendency in pore water pressure reflects an each loading stage of pore water pressure. It is clear that the transmission time in each loading stage for pore water pressure to transmit from the bottom to the top of soil specimen increases with the increase in pore water pressure. At low pore water pressure, the transmission time seems almost constant. This behavior suggests the soil expands elastically according to the decrease in mean effective stress caused by pore water pressure loading. The trans-
Figure 2. Schematic figure of failure mode and water content after failure (unit:%) The water content is also widely distributed. This distribution of water content is due to water migration caused by non-uniform shear deformation. As Asaoka et al.(l997) pointed out, the localization process in deformation progressed as water migrated. The shear strength of overconsolidated clay deteriorated due to swelling by absorbing water. The softened area naturally deformed further and the localization proceeded successively. This might be an another reason that the transmission time of pore water pressure gets longer as the enforced pore water pressure closes to the critical magnitude.
Figure I . Transmission of pore water pressure 732
3.3 Upper and lower yield limits With the use of measured pore water pressure at the top of the specimen, the mean effective stress of soil specimen can be estimated approximately. Fig.3 shows the measured relationship in pore water pressure loading test between the axial strain and the mean effective stress in the case of the deviator stress, 75kPa. The undrained shear process to set up the initial state is also drawn in the figure. When the mean effective stress is high, the increase in axial strain is small, but it monotonically increases due to swelling. After the mean effective stress of 60kPa the axial strain suddenly increases. The axial strain finally increases unconfinedly after the mean effective stress of 35kPa. As Ogawa( 1986) pointed out two thresholds in confining stress for the soil to generate shear deformation in ring shear test, the similar thresholds in mean effective stress are found in pore water pressure loading test. They are defined here as the upper and lower yield limits in terms of mean effective stress on the generation of shear deformation. The lower yield limit indicates the threshold for the soil to generate the plastic deformation and the upper yield limit exhibits the limit state.
Figure 4.Relationship between volumetric strain and deviatoric strain 3.4 Efective stress path Both effective stress path and void ratio change in pore water pressure loading test are illustrated in Fig.5. In the figure, the stress of soil specimen firstly traces the undrained path up to the prescribed deviator stress. The mean effective stress decreases with the increase in pore water pressure and attains to the upper yield limit defined before. After the mean effective stress reaching the upper yield limit, the deviator stress can not be kept constant and reduces. In the figure, the yield function of the original Cam clay model is illustrated. It is noted that the plastic deformation generates though the effective stress of soil locates inside the yield function, which indicates the Cam clay model can not be applied to simulate the soil behavior of pore water pressure loading test. The localization in deformation gradually proceeds especially after the upper yield limit and the soil specimen gets to be non-uniform. In this stage the soil specimen can not
Figure 3. Relationship between axial strain and mean effective stress Fig.4 exhibits the relationship between the volumetric and deviatoric strains. The upper and lower yield limits are also exhibited in the figure. From the initial state to the lower yield limit, almost linear relationship can be found between the volumetric and deviatoric strains. It means the stiffness of soil keeps constant as a linear elastic body. However, the deviatoric strain gets to generate more than the volumetric strain after the mean effective stress passing the lower yield limit. This tendency is clear after the upper yield limit. It is owe to the plastic deformation. It is noted that the volumetric strain increases largely after the lower yield limit.
Figure 5. Effective stress path and void ratio change in pore water pressure loading test 733
be a soil element and the reliability in stress path is already lost. However, it can be seen that the macroscopic stress gets close to the residual state on the critical state line. The void ratio keeps constant during the undrained shear process. With the increase in pore water pressure the void ratio increases along the elastic swelling line. However, it increases largely apart from the swelling line after the lower yield limit. The difference in void ratio between the measured and the corresponding void ratio on swelling line is the dilation caused by the plastic deformation. It clearly shows that the lower yield limit expresses the elastic limit for pore water pressure loading. After the pore water pressure passing the upper yield limit, the void ratio increases more and forward to the critical state line. 3.5 Strength parameters for design Pore water pressure loading tests were conducted under various deviator stresses of 50, 75, 100, I20kPa. Fig.6 represents the obtained stress paths in a series of tests. Each path is similar to the stress path exhibited in Fig.5. The lower and upper yield limits can be defined for each path as shown in the figure. It is readily seen that the lower yield limits are situated near the critical state line. This fact is noticed because, the critical state line expresses the residual state in the original Cam clay model, on the contrary, the lower yield limit indicates an elastic limit for pore water pressure loading. From the viewpoint of design the critical state line can be employed for the conservative soil parameter in the senses of both elastic limit and residual state. On the contrary, the upper yield limits are obtained in the dry area left side of the critical state line. The plotted upper yield limits seem to be situated on the line, which gives the soil parameters for aggressive design of employing the peak strength.
Figure 6. Upper and lower yield limits under various deviator stresses 734
4 CONCLUSIONS Pore water pressure loading tests of a clay were conducted to make clear the failure mechanism of landslide. The followings were concluded in this study. 1) Shear deformation of soil in pore water pressure loading test appeared taking a long time as the stress state closed to the failure condition of upper yield limit. This behavior could be well understood by the soil-water coupling concept. It is a possible reason why landslides proceed slowly. 2) Upper and lower yield limits in terms of effective stress were observed for the generation of large deformation. The lower yield limit, which denoted the elastic limit, was situated along the critical state line and the upper yield limit for the peak strength was located in the dry area left side of the critical state line.
ACKNOWLEDEMENT This research was supported in part by a grant from Sabo Technical Center. The writers wish to thank Mr. Ikarashi, H. of Kiso-Jiban Co. and Mr. Nakashima, T. of Nagaoka University of Technology for their helps and valuable comments to conduct this research. REFERENCES Asaoka, A., Nakano, M. and Noda, T. (1997). Soil-water coupled behavior of heavily Overconsolidated clay nearfat critical state, Soils and Foundations, Vo1.37, No. 1, pp. 13-28. Katagiri, M. and Imai, G. (1996). Deformation characteristics of a saturated cohesive soil subjected to increase in pore pressure, Soils and Foundations, V01.36, NO.3, pp.1-12. Ogawa, S. (1 986). Ground water behavior and soil strength in Yomogihira and Nigorisawa landslides, 14th Field Investigation Report, Niigata Branch of Japanese Landslide Society, pp.27-38(in Japanese).
Slope Stability Engineering, Yagi, Yamagami & Jiang k) 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Shear behavior of clay subjected to change of normal stress M. Suzuki & T.Yamamoto Departnlent of Civil Engineering, Yanzugcrchi University, Ube, Japan
T.Umezaki Depur~nentof Civil Engineering, Shinshu University,Nagano, Japan
ABSTRACT Ring shear tests, which can simulate the change of pore water pressure in the specimen, are conducted on kaolin clay. The total normal stress is changed during the drained shearing, instead of changing the pore water pressure in the specimen directly. The shear stress of the clay specimen decreases along the residual strength line and disappears accompanying a decrease in effective normal stress. Then the shear stress mobilized on the slip surface is recovered by increasing the effective normal stress.
1 INTRODUCTION According to the present explanation for the occurrence of landslide, the stress condition of the soil element on the slip surface within a slope reaches a certain failure envelope under a constant shear stress as the pore water pressure increases. This phenomenon has been simulated and investigated by imposing an additional back pressure on a cylindrical specimen in triaxial tests (Eigenbrod et al., 1987, Tokida et al., 1987, Katagiri et al., 1996). On the other hand, the soil element on the slip surface after reaching a residual state is subjected to various changes of pore water pressure. Reactivated landslide occurs frequently during rainfalls and snow melting. Thus the influence of pore water pressure on strength parameters is an important factor in considering the stability of reactivated landslide. Ogawa et al. (1981,1987) and Kamei et al. (1 987) conducted ring shear tests which simulate the increase in pore water pressure expediently. The tests are performed by decreasing the total normal stress during drained shearing, instead of increasing the pore water pressure in the specimen directly. Yatabe et al. (1991) expressed doubts for the test results because an excess pore water pressure might be generated in the specimen by decreasing the total normal stress. Therefore, it is very important to 735
grasp the exact value of the true effective normal stress in the specimen. Umezaki et al. (1999) developed a triaxial slice shear test apparatus with the aim of increasing the pore water pressure in the specimen uniformly for a short time. The advantage of this apparatus is that the shear behavior of the clay can be evaluated from the viewpoint of the effective normal stress. However, a ring shear test. which can give infinite shear deformation, is suitable for defining the conception of the residual strength of soil. This paper demonstrates the validity of the ring shear test, which simulates the change of porc watcr pressure in a specimen, in terms of the rate of normal stress. As compared with the results of triaxial slice shear tests, it also describes new findings on the residual shear behavior of clay subjected to the change of pore water pressure.
2 RING SHEAR TEST TO SIMULATE CHANGE OF PORE WATER PRESSURE Figure 1 schematically shows the essential features of the ring shear test apparatus. A ring shear test can give an endless shear displacement to an annular specimen, so that the residual strength can be accurately determined. The specimen is placed in the
Fig.2 Conditions of total normal stress and pore water pressure in the specimen
Fig.1 Essential features of ring shear test apparatus
Table 1 Test cases and initial conditions of specimens CJc*3
Test NO.
1.603
I I
1.667 1.668 4
1.666
5
1.735
~
1.:
"4
U N:
I I
61.5
52.9
Total normal stress
oN*4
"6
(kPa) I
61.3
"1 P ,: Initial wet density "2 w,) : Initial watcr content "3 iJ (.: Consolidation stress
I
(kPa)
A
oN*7
;N*x
Symbols
(kPa/min.)
(rad/min.) 1 .o
0.0025
196
0.98
1.o
0.0025
196
4.9
1.o
0.0025
196
98.0
1 .0
0.0025
98
4.9
0
2.0
0.0025
98
4.9
v
"5 PCR: Overconsolidation ratio(= (I / (I N) "6 0 : rate of shear displacement angle "7 CJ N : change of total normal stress "8 N: rate of total normal stress
4
accurately. A simple method for simulating the change of pore water pressure in the specimen was proposed by Ogawa et al. (1981). The contrivance of the method is summarized as follows. Instead of increasing the pore water pressure in the specimen, the total normal stress which is equal to the increment of pore water pressure is decreased. On the contrary, instead of decreasing the pore water pressure in the specimen, the total normal stress which is equal to the decrement of pore water pressure is increased. If the rate of the total normal stress is too high, an excess pore water pressure might be unexpectedly generated in the specimen. As a result, the effective normal stress on the slip surface can not be evaluated precisely. On the other hand, if the rate of the total normal stress is low enough to dissipate the excess pore water pressure generated in the specimen, the total normal stress is
central part of the apparatus. The inner and outer diameters of the specimen are 60 mm and 100 mm, respectively. The shear stress is applied to the specimen by rotating a turning table. The normal stress, which actually acts on a slip surface, is maintained at a constant value by measuring a frictional force generated between the rigid shear box and the speciemen. The rate of shear displacement angle, 8 , is adopted to ensure the drained condition in the specimen ( Suzuki et al., 1997). Here, a rotating angle, 8 , is used, instead of an intermediate displacement between the inner and outer diameters of the specimen, D. Figure 2 schematically shows the conditions of total normal stress and pore water pressure in the specimen. The ring shear test apparatus is not able to impose the back pressure on the specimen and to measure the pore water pressure in the specimen
736
Fig.3 l’ypical relationship between shear stress and shear displacement angle
equivalent to the effective normal stress (see Fig.2). Therefore, it is necessary to demonstrate the validity of this method by confirming the rate of the total normal stress. A series of tests is performed on kaolin clay. The physical properties of kaolin are as follows; density of soil particles: 0 = 2.724 g/cm3, liquid limit: wL= 75.6 %, plasticity index: I,, = 39.3, clay fraction: F,,,,= 70 %. The test cases and initial conditions of specimens are listed in Table 1. Details of the ring shear test procedure are described elsewhere (Suzuki et al., 1997). ~
3 RESULTS AND DISCUSSIONS
Figs.4 Changes of the normal stress, the shear stress and the vertical displacement with time, respectively.
Figure 3 shows the typical relationship between the shear stress, Z,and the shear displacement angle, 0 , during the ring shear. The specimen is drainedsheared under a constant normal stress. After the shear stress passed through a maximum value, all shear stress gradually decreases and reaches a constant value i.e. a residual strength. It is regarded that all shear stress reaches the residual strength at 6’ 2 5 rad. As the normal stress decreases immediately after the shear displacement angle becomes 0 = 10 rad, the shear stress decreases simultaneously. The quantity of 8 = 10 rad corresponds to D = 400 mm. Subsequently as the normal stress increases again, the shear stress increases.
Figures 4(a)-(c) show the changes of the normal stress, 0 N, the shear stress, Z, and the vertical displacement, v, with time, T, respectively. The data points are the same as those used in Fig.3. As the normal stress decreases monotonously, the shear stress decreases simultaneously. When the normal stress becomes almost zero, the shear stress mobilized on the slip surface almost disappears. It is suggested that this behavior is very similar to ‘the liquefaction of sand’. Following the process, as the normal stress increases monotonously, the shear stress increases again. The latter behavior is different from ‘the liquefaction of sand’. The vertical displacement also simultaneously decreases and 737
Figs.5 Relationships between the shear stress and the normal stress during a decrease in normal stress
Figs.6 Relationships between the shear stress and the normal stress during an increase in normal stress
increases accompanying the decrease and increase in normal stress, respectively. Figures 5(a)-(c) show the relationships between the shear stress, Z , and the normal stress, CJ N, during a decrease in normal stress. These results are obtained from ring shear tests under different rates * of normal stress, CJ N. Both the residual and peak strength lines shown in Figs.5 were determined by conventional ring shear tests on normally consolidated clay under different normal stresses. The angle of shear resistance and the cohesion at the residual strength are 6 = 11.3 and c, = 0, respectively. On the other hand, The angle of shear resistance and the cohesion at the peak strength are 17.6' and c,= 0, respectively. In the cases of
tests under 0, = 0.98 and 4.9 kPa/min, both shear stresses decrease along the residual strength line, accompanying the decrease in normal stress. These results are in good agreement with those of the triaxial slice shear tests which increase the pore water pressure in the specimen by imposing an additional back pressure on the specimen (Umezaki e et al., 1999). In the only case of a test under (i I'! = 98 kPa/min, the shear stress decreases over the residual and peak strength lines accompanying the decrease in normal stress. Considering the shape of the stress path, we hypothesize that the specimen did not have the necessary drained condition. Because the total normal stress is decreased very fast, a negative excess pore water pressure might be
ad=
0
738
Fig.7 Relationships between the shear stress and the shear displacement angle under different overconsolidation ratios
Fig.9 Typical relationship between the displacement and the normal stress
total normal stress is increased very fast, a positive excess pore water pressure might be generated in the specimen. In the range below 0 = 4.9 kPa/min, the rate of normal stress has little influence on the above residual shear behaviors. In order to simulate the change of pore water pressure in the specimen using the ring shear test apparatus, it is very important to change the total normal stress as slowly as possible. Figure 7 shows the relationships between the shear stress, Z , and the shear displacement angle, 8 , under different overconsolidation ratios. Both Z - 8 curves at 8 2 5 rad are independent of the value of the overconsolidation ratio. It should be noted that the residual strength of the clay is not influenced by the overconsolidation ratio. Figure 8 shows the relationships between the shear stress, E , and the normal stress, U N, under different overconsolidation ratios. The data points are the same as those used in Fig.7. In two cases, both shear stresses decrease along the residual strength line, accompanying the decrease in normal e stress under O = 4.9 kPa/min. From these results, it may be concluded that the above residual shear behavior is not influenced by the stress history in the consolidation process. Figure 9 shows the typical relationship between the vertical displacement, v, and the normal stress, 0 N, during the decrease and increase in normal stress. The data points are the same as those used in Fig.3. There exist definite differences in the swelling
.
Fig.8 Relationships between the shear stress and the normal stress under different overconsolidation ratios
generated in the specimen. Figures 6(a)-(c) show the relationships between the shear stress, E , and the normal stress, (7 N, during an increase in normal stress. In the cases of a * test under O N = 0.98 and 4.9 kPa/min, both shear stresses increase along the residual strength line, accompanying the increase in normal stress. This finding suggests that the shear stress mobilized on the slip surface can be recovered by dissipating the positive excess pore water pressure using various drainage methods. In the only case of a test under = 98 kPa/min, the shear stress increases below the residual strength line, accompanying the increase in normal stress. In this case, the drained condition in the specimen is also not satisfied. Because the 739
vertical
behaviors of the specimen when it is subjected to a decrease or an increase in normal stress. The v- 0 curves are the hysteresis.
4 CONCLUSIONS The main conclusions are summarized as follows: In order to simulate a change of pore water pressure in the specimen using a ring shear test apparatus, it is very important to change the total normal stress as slowly as possible. In the above tests, the change of the total normal stress is equivalent to that of the effective normal stress. The shear stress of the specimen, once it has reached the residual state, decreases along the residual strength line, accompanying the decrease in effective normal stress. Finally, the shear stress mobilized on the slip surface almost disappears as the effective normal stress approaches zero. The shear stress of the specimen, which has become almost zero due to the decrease in the effective normal stress, increases along the residual strength line again, accompanying the increase in the effective normal stress. This finding suggests that the shear stress mobilized on the slip surface can be recovered by dissipating the excess pore water pressure. After the specimen reaches the residual state, the residual shear behaviors are independent of the stress history in the consolidation process.
ACKNOWLEDGMENT The authors are grateful to Emeritus Professor Hiroshi Kawakami of Shinshu University for supervising this study and to Mr. Naoki Miyamura & Mr. Hideyuki Ito for the experimental assistance.
REFERENCES Eigenbrod, K.D., Burak, J. -P. & Graham, J. 1987. Drained deformation and failure due to cyclic pore pressure in soft natural clay at low stress, Canadian Geotechnical Journal, Vo1.24, pp.208215. 740
Katagiri, M. & Imai, G. 1996. Deformation characteristics of a saturated cohesive soil subjected to increase in pore pressure, Soils and Foundations, Vo1.36, No.3, pp.1-12. Kamei, T., Ikeda, T., Ogawa, S., Ikeda, T. & Shimazu, K. 1987. Slope stability analysis in landslide and strength parameters, Journal of Japan Landslide Society, Vo1.24, No.3, pp.1-7 (in Japanese). Ogawa, S., Ikeda, T., Cho, S., Kaizu, N. & Noji, A. 1981. Shearing test for determining strength parameters of soil relevant to landslide analysis, Proc. of the 16th Japan National Conference on Geo-technical Engineering, pp.3 65-368 (in Japanese). Ogawa, S., Ikeda, T., Kamei, T. & Wada, T. 1987. Field investigations on seasonal variations of the ground water level and pore water pressure in landslide areas, Soils and Foundations, Vo1.27, No.1, pp.50-60. Suzuki, M., Umezaki, T. & Kawakami, H. 1997. Relation between residual strength and shear displacement of clay in ring shear test, Journal of Geotechnical Engineering, No.5751 -40, pp.141-158 (in Japanese). Tokida, M., Hashimoto, M., Ikeda, T., Ogawa, S. & Kamei, T. 1987. Influence of the increase in pore pressure on the shear characteristics of cohesive soil subjected to the stress history, Proc. of the 22nd Japan National Conference on Geotechnical Engineering, pp.467-468 (in Japanese). Umezaki, T., Suzuki, NI. & Yamamoto, T. 1999. Strength and deformation characteristics of clay subjected to pore water pressure increment, International Symposium on Slope Stability Engineering: Geotechnical and Geoenvironmental Aspects, IS-SHIKOKU’99, (in press). Yatabe, R., Yagi, N. & Enoki, M. 1991. Consideration from effective stress about strength parameters of slip layer clay of landslide, Journal of Japan Landslide Society, Vo1.28, No.2, pp.2026 (in Japanese).
Slope Stability Engineering, Yagi, Yamagami L? Jiang @ 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
A simple model to predict pore water pressures during shearing along undulating surfaces D.J. Petley School of Engineering, University of Warwick, UK
PTaylor WSAtkins North West Limited, Warringtoa, UK
ABSTRACT: This paper is the first part of a study that assesses the influence of undulations on the effective stress across a shear zone during rapid shearing. A simple model is proposed to predict the magnitude of the pore water pressures which may be generated when shearing occurs along non-planar shear surfaces. The model assumes that the non-planar shear surface can be divided into a zone of positive gradient (where compression is occurring) and a zone of negative gradient (where swelling is taking place). In the compression zone, the change in effective stress is obtained from one-dimensional consolidation theory. In the swelling zone, two solutions are considered; firstly, a “free swell model” (where shear zone separation may occur), and secondly, a “forced swell model” (where separation is not permitted). The merits of the two assumptions are considered, and subsequently the free swell model is proposed as the more accurate technique. perpendicular to the undulations will also increase shear resistance and has the potential to generate excess pore water pressures. Undulations perpendicular to the shear direction were first noted during shear box tests in the laboratory by Skempton and Petley (1967) and Morgernstern and Tchalenko (1967). They both noted that shear surface development started with a number of small discontinuities or Riedel shears at the edges of the shear box at angles up to 30” above and below the horizontal. As displacement continues the Riedel shears gradually extend and join together to form an undulating surface. Skempton and Petley (1967) noted that with continued shearing the surface gradually became sub-planar as residual strength is attained. More recently Parathiras (1994) noted that undulations perpendicular to shear direction developed during tests on cohesive soils in the NGI/IC ring shear apparatus. It was also noted that when these undulations developed with amplitudes as low as 0.15mm, a significant loss of residual strength occurred at rates of shearing corresponding with rapid to very rapid, after Varnes (1978). Parathiras (1994) postulated that this fall in residual strength was due to a decrease in the effective normal stress as a result of the “pumping efj’,ct” of the undulations generating excess
1 INTRODUCTION Theoretically shearing across an undulating surface could take place in any direction, however to simplify matters only two directions will be considered. These are illustrated in Figure 1.
Figure 1 : Definition of undulation orientation
Shearing parallel to the undulations will increase the shear resistance of the surface and is unlikely to generate pore water pressures. Shearing 741
positive pore water pressures. Similar falls in residual strength were noted by Petley and Taylor (1997) during NGI/IC ring shear tests, using kaolin against an undulating rigid interface at rapid rates of shearing as low as 60 m d m i n , the amplitude of these waves being 0.5mm. Taylor (1998) also noted significant falls in residual strength in association with undulation development during soil on soil NGI/IC ring shear tests at rapid rates of shearing. The extrapolation of these laboratory findings to actual landslides is very interesting. It is likely that the generation of pore water pressures by uneven or undulating shear zones is the mechanism behind negative rate behavior (a loss of residual strength at faster rates of shearing) and it could be the much debated mechanism behind long run-out landslides. Because Taylor and Petley (1997) and Parathiras (1994) noted that undulations with amplitudes of 0.5mm or less induced negative rate behavior at rapid rates of shearing, it is not unreasonable to conclude that all natural slip surfaces are sufficiently uneven to illustrate such behaviour. If the key to the loss of strength is the attainment of rapid rates of shearing, this could typically be a result of earthquake loading, an occurrence which is known to trigger long run-out landslides. However, more significant undulations can develop in the field both naturally and as a result of large-scale civil engineering. Bromhead (1992) noted that in many tectonically sheared clays, numerous shear zones exist which continuously merge and diverge, resulting in a highly irregular shear zone, containing undulating shear surfaces and lenticular masses of clay. Bridle et a1 (1985) reported that the construction of Emphingham dam (United Kingdom), resulted in perpendicular shearing of undulations in the dam foundation. These were initially formed in the upper Lias clay by valley-ward movements including bulging and cambering. Another example from the United Kingdom is Carsington dam. Skempton and Coats (1985) noted that an undulating shear surface, which was initially formed by the periglacial process of solifluction, was loaded by the dam construction in a direction perpendicular to the shear surface undulations. There is the potential for pore water pressure generation by perpendicular shearing of undulating shear zones, to cause significant falls in strength during earthquake loading in natural slopes and in the vicinity of large manmade structures. This paper presents an attempt to model and understand this pore water pressure generation during rapid shearing. The model results are compared with
observations made using the NGI/IC ring shear apparatus in a second paper (Petley and Taylor 1999).
2 PRINCIPLES AND ASSUMPTIONS OF THE COMPRESSION MODEL If a saturated soil is compressed through a given distance, the change in effective stress, A d , can be calculated from the following relationship: sc= m, A d H (l), where sc is the normal compression distance, m, is the coefficient of volume compressibility and H is the initial depth of the soil. By assuming that the change in effective stress is totally accounted for by pore water pressure generation, it is possible to calculate the magnitude of this pressure (U) by rearranging the above equation to yield: U = sc / m, H ( 2 ) . Furthermore it is then possible to calculate the dissipation of this pressure with time using onedimensional consolidation theory, given the coefficient of consolidation, c,. This has allowed a technique to be developed, which allows the generation of pore water pressures to be calculated during shearing up a positive gradient. The model works by considering one narrow strip of soil starting at the base of the wave-form, which then steadily compressed as shearing takes place and the strip translates to the top of the wave-form. The model divides the wave-form into a number of very small steps, this is illustrated schematically in Figure 2. The soil strip is assumed to spend a short yet finite period of time on each step; the time period will depend on the number of steps the wave-form is divided into and the rate of shearing. When this time has elapsed, the strip is moved instantaneously up onto the next step. The difference in height of each step gives the compression distance, sc, and the vertical distance from the step to the upper fixed boundary gives the soil thickness H. Given that m, is a soil parameter, the pore pressure generated from this movement from one step to the next can be calculated. The next stage is to calculate the dissipation of this pressure using one-dimensional consolidation theory, during the finite time period when the strip is on the step. The remaining pore pressure at the end of this time period is then added to the pressure generated by the next up step movement, before further dissipation takes place. Linking all the steps in this manner provides the pore water pressure distribution
742
Figure 2: Standard Undulation Geometry. along the positive gradient of the wave-form. By selecting a large number of steps and therefore making each step very narrow the model approximates a constant rate of shear. All models need to make a number of assumptions and create boundary conditions so that their limitations are fully understood, this one is no exception. During shearing along a positive gradient, all of the assumptions made in Terzaghi’s theory of one-dimensional consolidation apply. In addition to these, the strip is considered to be a single isolated unit that is unconfined laterally but will not deform in this direction. There are no interstrip forces between this strip and the surrounding soil, and no dilation or increased porosity occurs at the shear interface. Finally the material below the wave-form is assumed to be impervious and therefore drainage is only allowed vertically upwards.
3 PRINCIPLES AND ASSUMPTIONS OF THE EXPANSION MODELS
Modelling expansion of the sample during shearing along the negative gradient is a more complicated problem. In reality this is a three-dimensional problem therefore to try and solve it in what is essentially one-dimension involves some wide ranging assumptions. Two options are considered here. The Free Swell Model, assumes consolidation applies equally to swelling as it does to compression and the equations in the model use values of mvsweiland Cvswell to calculate pressure dissipation. The sample is allowed to swell 743
resulting in pore pressure dissipation but is not forced to stay in contact with the wave-form. This creates two problems, firstly the sample is still assumed to be under an effective normal stress, yet there is no medium to transmit this stress. Secondly pore water pressure dissipation is calculated from vertical drainage but is then used to find the amount of swelling needed to achieve the same dissipation. The second point is accepted as an assumption of the model. The first is overcome by expanding the strip to its full height at the end of the negative gradient. This maintains the geometric integrity of the model without changing the effective stress regime. The Forced Swell Model assumes that the upper body of soil is forced to maintain contact with the undulating surface. This model functions in exactly the same way as the compression along the positive gradient phase, except that the value of sc becomes negative yielding pore pressure dissipation and values of mvsweliand Cvswell are used in the calculations in place of m, and c,. Both the Free Swell and Forced Swell Models were looped together with the model for the compression along the positive gradient. All of the code for these models was written in C and run on a Unix system, both programs produced results in tabular arrays, which could then be plotted using commercial spreadsheets or packages such as Matlab. When the compression and expansion phases were looped together any number of undulations with fixed wavelengths and amplitudes could be modeled.
4 PRELIMINARY MODELLING The initial aim of the model was to produce results that could be compared to NGUIC ring shear test results on remoulded kaolin. Before investigations could commence values of volume compressibility and coefficients of consolidation were determined on remoulded kaolin using standard oedometer tests to BS 1377 Part 5 (1990). Three different tests were conducted and average values were calculated. Both models were then set up using these values and with undulation wavelengths of 100 mm and amplitudes of 0.5 mm; 4 undulations were used in the initial runs. These dimensions match the rigid interface that was used by Petley and Taylor (1997) in the NGI/IC ring shear apparatus. In addition the maximum drainage path length in these tests was 9 mm and therefore this value was initially adopted for H in both models, these dimensions are now
termed “standard undulation geometry’’ and are illustrated on Figure 2. Shear rate was set to 50 m d m i n and the total normal stress, on,was set to 100 kPa. The results from these preliminary runs eliminated the Forced Swell Model immediately. During the compression phases both models generated pore water pressures which peaked at the end of the positive gradient at values of 60 kPa. During the expansion phases the Free Swell Model produced pore pressures which decreased down to a level approaching zero towards the end of the negative gradient. The Forced Swell Model however, generated very large negative pore water pressures of the order -500 kPa. The reason for this negative pressure is the soils inability to swell naturally as much as it has been forced to compress. This is usually well defined on plots of void ratio against the logarithm of normal pressure, during unloading stages in oedometer tests. Therefore when forced to swell past its natural capability, that is where pore pressures have dissipated totally, negative pore water pressures are induced. These were obviously unrealistic and therefore this model was abandoned.
British Standards Institute 1990. BS 1377 Soils for civil engineering purposes. Part 5. Compressibility, permeability and durability tests. Bromhead, E.N. 1992. The stability of slopes. Blackie Academic and Professional, 2nd edition. Morgernstern, N.R. & Tchalenko, J.S. 1967. Microscopic structures in kaolin subjected to direct shear. Geotechnique, 17:309-328. Parathiras, A.N. 1994. Displacement rate effects on the residual strength of soils. PhD thesis, University of London. Petley, D.J. & Taylor, P. 1997. Quick shear with slip of soils against rigid and rough surfaces. In Proc. 2nd Pan-American Symp. on Landslides, 2’ld COBRAE, 1, 435-442, Rio de Janeiro. Petley, D.J. & Taylor, P. 1999. Modelling rapid shearing of cohesive soils along undulating shear surfaces. Proc. IS-Shikoku 99 (in press).
5 CONCLUSIONS Skempton, A.W. & Coats, D.J. 1985. Carsington dam failure. Proc. Syinp. on Failures in Earthworks, 203-220, Institution of Civil Engineers, London
All further studies have been performed on the free swell model and it is believed that this model offers the greatest potential for future development. The second paper in this study Petley and Taylor (1999) goes on to illustrate the use of the model and its implications and correlations with laboratory observations. It is shown that the model does provide encouraging results when compared to actual observations. Future developments of the model could include allowance for the interslice forces between each of the strips analysed, two way drainage paths, the different drainage properties resulting from the dilate shear zone and include the ability to simultaneously analyse a number of different strips at different points in the sliding soil mass. Adding these refinements could make the model more accurate and potentially allow predictions of landslide response to earthquake type loading.
Skempton, A.W. & Petley, D.J. 1967. The strength along structural discontinuities in stiff clays. Proc. of Geot. Con,, Oslo, 2, 29-46, Norwegian Geotechnical Institute, Oslo. Taylor, P. 1998. Fast shearing of cohesive soils using ring shear apparatus. PhD thesis, University of Warwick. Varnes, D.J. 1978. Slope movements and types and processes. In Landslides: analysis and control. Special Report 172, Chapter 2. Washington: Transportation Research Board, National Academy of Sciences.
6 REFERENCES Bridle, R.C, Vaughan, P.R. & Jones, H.N. 1985. Emphingham dam-Design, construction and performance. Proc. Institution of Civil Engineers, 78:247-289.
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Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Modelling rapid shearing of cohesive soils along undulating shear surfaces D. J. Petley School of Engineering, University of Warwick, UK
PTaylor WS.Atkins Northwest Limited, Warrington, UK
ABSTRACT: This paper is the second of a two-part study that assesses the influence of undulating shear surfaces on the effective stress across a shear zone during rapid shearing. This part of the study uses the simple model developed in part one to investigate the influence of rate of shearing, undulation geometry, soil thickness and total normal stress on pore water pressure generation in the shear zone. These preliminary investigations illustrated a correlation between pore water pressure generation and the negative rate behaviour associated with many long run-out landslides. Therefore additional investigations have been conducted and compared with laboratory testing in the NGI/IC ring shear apparatus in which negative rate effects have been recorded and pore water pressures monitored in association with undulating shear surfaces. These comparisons are presented here and it is shown that the model has merits and illustrates the potential for further development. is complete pore pressures start to accumulate from a level above zero pressure. In Figure 1 it is illustrated that at 100 m d m i n this level is around 18 kPa. If the run is repeated over 20 standard undulations the pressure peaks and troughs level out to 82 kPa and 20 kPa respectively. At rates above 200 mm/min Figure 1 shows the soil liquifies (the pore pressure reaches the total normal stress) within four undulation cycles (one revolution of the NGI/IC ring shear apparatus). An additional run has shown that liquifaction occurs within five undulation cycles at a rate of 150 m d m i n . This implies that a critical rate exists between 100 mrnhin and I50 m m h i n above which the residual Mohr-Coulomb envelope will break down and the soil will behave as a liquid with any shear strength being a result of viscosity.
1 INVESTIGATING THE RATE OF SHEARING If undulation induced positive pore water pressures are the mechanism behind negative rate effects, it would be expected that the magnitude of these pressures would increase with rate of shearing. The Free Swell Model was set up for remoulded kaolin on the standard undulation geometry described in part one of this study, Petley and Taylor (1999). The total stress applied was 100 kPa. Eight separate runs were conducted at rates ranging from 1 mm/min to 1000 mmlmin and the results from these are plotted in Figure 1. These reveal the generation of pore water pressures that vary in a cyclic manner reflecting the zones of compression (positive gradient) and expansion (negative gradient). The magnitude of the pore pressures increase with rate of shearing. This is because faster rates of shearing allow less time for drainage on each step and therefore higher pressures accumulate. Figure 1 also illustrates that at rates up to 50 mm/min the shearing is slow enough to allow full pore pressure dissipation in the expansion zone, therefore the wave pattern repeats at a constant level between zero pressure and the peak pressure obtained at a given rate. At rates above 50 mm/min a slightly different type of behaviour is occurring. There is insufficient time to allow full pore pressure dissipation in the expansion zone and therefore once one wave cycle
2 INVESTIGATING UNDULATION GEOMETRY AND SOIL THICKNESS
Studies of the influence of undulation amplitude and wavelength were conducted using 100 kPa total normal stress and at a rate of shearing of 150 mrdmin. Amplitudes were varied between 0.25 mm and 0.75 mm and wavelengths between 25 mm and 100 mm. These studies revealed that pore water pressure levels increased with amplitude and decreased as wavelength increased. Pore pressure
745
Figure 1: Variation of pore pressure generation at different rates of shearing. generation increased with the gradients on the undulations. This may be a function of the Free Swell Model as pressure generation increases with the magnitude of the positive gradient, but is independent of the negative gradient in the expansion zone. If however, undulation induced pore pressure is the mechanism of negative rate behaviour these results are in agreement with Parathiras (1994), who noted that the magnitude of negative rate effects increased with undulation amplitude. An investigation has also been undertaken into the effect of the parameter H on pore water pressure generation. The reason for this is that soil loss (reducing H) from the NGI/IC apparatus has been highlighted as a potential problems in many pieces of work including Lernos (1986), Tika (1989), Parathiras (1994) and Taylor (1998). Two effects are possible as a result of reducing H. Firstly the drainage path is shortened, allowing more drainage to take place by increasing the consolidation time factor T,,, and therefore potentially decreasing the cumulative levels of pore water pressure. Secondly, higher pore water pressures will be developed reducing soil thickness. The model was set up for
remoulded kaolin using the standard undulation geometry, a shear rate of 150 m d m i n and a total normal stress of 100 kPa. The maximum height of the soil strip was varied between 9 mm and 3mm. It is apparent from Fig. 2 that cumulative pore pressure is reduced by decreasing the strip height. This indicates that the effect of increased drainage is greater than the effect of increased pressure generation. During shearing along the positive gradient in the compression zone, the curvature of the pore pressure graphs increases significantly as sample depth is reduced from 9 mm to 5 mm. At 5 mm when the undulation peak is reached the gradient of the plot is virtually zero. Below this depth the degree of drainage is so high the pore pressures actually start to dissipate along the positive gradient, as illustrated on the H=3 mm plot in Figure 2. It is important to allow for this increased drainage in NGI/IC ring shear tests, as many negative rate effects are observed towards the end of the multistage tests, when sample depths are significantly lower than 9 Inm.
746
Figure 2: The variation in pore pressure distribution with strip height 3 INVESTIGATING TOTAL NORMAL STRESS
Previous research by Tika (1989) and Parathiras (1994) has suggested that the magnitude of negative rate effects decrease with increasing normal stress. Then the ratio of the average effective normal stress should increase with to total normal stress, dll;Jon total normal stress. This investigation was conducted using the standard undulation geometry, assuming a remoulded kaolin soil and a rate of shearing of 150 mmhnin. The total normal stress was increased from 50 kPa to 400 kPa and the results are illustrated in Figure 3. A trend of increasing pore pressure generation with normal stress is observed due to the effect of decreasing my values with increasing total normal stress. Lower values of m,, lead to higher pore pressures being generated. Opposing this increased pressure is improved drainage as a result of cv increasing with total normal stress and therefore increasing the time factor Tv;the results in Figure 3 suggest this effect is relatively insignificant. As stated previously the critical ratio with these results is o’lla,I~n which illustrates the magnitude of any pressure induced negative rate behaviour: O’~,~,,/O,, =1 indicates no negative rate effect and o’,,~,/o~, =O indicates soil liquifaction and hence large negative rate effects. The values of G’,,:~,,are taken to be the pore pressures halfway between the pressure peaks and troughs and the results are summarised in Table 1. AS stated previously the critical ratio with these results is O’,~)O~ which illustrates the magnitude of any pressure induced negative rate behaviour: O’,,:~~/G,, =1 indicates no negative rate effect and O’~,~,V/O~~=O indicates soil liquifaction and hence large Ilegative rate effects. The values of dnaV are taken to
Figure 3: The variation in pore pressure distribution and effective normal stress with total normal stress. Table 1: Influence of total normal stress on effective normal stress. Total Normal Stress o,, (kP4
50 100 200 400
Average Effective Normal Stress o’r>:8v (kPa)
0 30 110 280
Ratio
o’”:,hJ“ 0.0 0.3 0.55 0.70
be the pore pressures halfway between the pressure peaks and troughs and the results are summarised in Table 1. This shows that the values of the ratio increase with total normal stress and indicates that the magnitude of undulation induced pore pressure negative rate effects decreases with increasing total normal stress, thus correlating closely to laboratory observations.
4 PREDICTING LABORATORY PORE WATER PRESSURE BEHAVIOUR Taylor (1998) reported a series of NGI/IC ring shear tests that involved the measurement of pore water pressures using miniature transducers installed in a perspex interface with the standard undulation geometry. These are beyond the scope of this study, however the results from Stage F of one test are shown in Figure 4. It is apparent from the fast residual strength graph that significant losses in strength occurred during this test. Of the four transducers installed on the interface only the one in the middle of the positive gradient (dashed line) and the one on the peak of the undulation functioned 747
Figure 4: NGI/IC ring shear test result with pore pressure measurement correctly and these will only be considered here. To model this stage of rapid shearing the properties for remoulded kaolin were used as was the standard undulation geometry. The results file of this test was studied and the soil thickness at the start of shearing corresponded with a value of H = 6.9mm. The rate of shearing was set to 50 m d m i n and the total normal stress to 79.5 kPa, finally a series of three undulations was selected corresponding to 300 mm of shearing in the test. The results of running this model are shown in Figure 5.
from one strip of soil travelling around the annulus. Therefore according to the model results the positive gradient transducer should have recorded a constant pore pressure equal to that predicted at displacements of 0.025 m, 0.125 m and 0.225 m, on Figure 5 this pressure is 43 kPa. The transducer reading is not constant, however its’ average value is of the order 30-40 Wa and therefore the agreement with the model is relatively good. The agreement between the model and the transducer on At the undulation peak is not as close. displacements of 0.05 m, 0.15 m and 0.25 m the model predicts a pore pressure of 45 kPa, whereas the recorded pressures varied between 20 to 40 kPa. In summary, the model appears to be predicting pressures that are of similar magnitudes to those recorded in the laboratory. The fact that the model overestimates the pressure may be attributed to the ability of the soil to drain in three dimensions in the laboratory but only one in the model. Further pore pressure testing and modeling however are required to provide further validation of the model.
5 PREDICTING LABORATORY OBSERVED NEGATIVE RATE BEHAVIOUR Petley and Taylor (1 997) reported two NGI/IC ring shear tests using remoulded kaolin against a planar interface (Test 3) and against an interface with the standard undulation geometry (Test 4). The aim of this section is to see if the magnitude of the pore Pressures predicted by the model can account for the loss of strength observed from Test 3 to Test 4. One difficulty with this task is that the mechanisms behind positive rate effects (an increase in residual
Figure 5: The variation in pore pressure distribution from modeling the parameters of Stage F. When the model results to the test results, it is important to reinember that the two functioning transducers take snap shots of the pore pressures on a positive gradient and an undulation peak of the annulus, The model provides results 740
rable 2: Comparing actual test results to model predictions.
strength with increasing shear rate), viscous effects and particle d~sor~entati~n, Taylor ( I 998), are still likely to be in operation when pore pressures are reducing the effective stress ultimately causing negative rate behaviour. Therefore they will have the effect of reducing the magnitude of the negative rate behaviour. Test 4 was selected because this offset could be estimated from Test 3, during which a positive rate effect was observed. The variation of fast residual strength with in Figure 6 and these increasing rate is il~ustr~ted results are s ~ m m ~ i s in e dthe first three columns of Table 2. The next column in Table 2 illustrates the percentage of fast residual strength lost as a result of changing from a planar interface in Test 3 to an unduiating one in Test 4.
98 kPa for both Tests 3 and 4 was used. This has induced a slight error into Table 2, but it helps to keep the comparison simple, Having caIcuIated the pore pressures required to create the observed loss in strength during Test 4, the model was run using the parameters from Test 4, see Table 3, to see if it would predict similar pore pressures. The results from this modeling are shown in Figure 7, as expected the pore pressures increase with rate of shearing. At rates of 300 m d m i n and 1000 m d m i n the pore pressure reached the total normal stress causing the sampfe to liquefy.
Figure 7: Prediction of pore water pressures during Test 4. Figure 6: Fast residual strength behaviour from Test 3 {planar interface) and Test 4 (standard undulation geometry interface.
When the sample liquefies the average pore pressure is equal to the normal stress, as shown in the final column in Table 2. At rates of 10 mm/min and 50 mm/min the sample did not liquefy and therefore average pore pressures where calculated, by taking the mid-height between the pore pressure peaks and troughs. This enables the completion of Table 2. Comparing the required and predicted pore water pressures in Table 2 reveals that the values are not in close agreement, which is surprising considering
The fifth column provides the pore pressure increase required to cause this loss of strength, this is calculated using the following equation, U = B,, oil(l-%loss/lOO) When using Equation to calculate the values for U, the total normal stress applied to the sample neglecting side friction which equaled 749
Maximum Sample Depth (mm) 7.45 7.10 6.72 5.88
Minimum Sample Depth (mm) 6.45 6.10 5.72 4.88
Undulation Wavelength (mm) 100 100 100 I00
Number of Undulations
Analysis Steps
4 4 4 4
100 I00 100 100
the close correlation observed with the pore pressure transducers. There are two major reasons for this, one relating to the model and the other to Test 4. The average predicted pore pressures in Table 2 are calculated for one soil strip that always starts of with a full undulation height compression, thus generating a maximum pore pressure at the undulation peak. In reality many "strips" of soil start of from the undulation peaks or from the expansion zone, these will all initially generate negative pore pressures, thus reducing the average pore water pressure around the annulus. The model does not account for this and for it to do so would require a different approach. This would be the next logical stage in the development of the model. During the early stages of Test 4, shear displacement was limited by soil loss and it is likely that true fast residual conditions were not properly established, especially when negative rate behaviour was occurring. It is likely that the levels of the ratio of fast residual strength to slow residual strength, provided in Table 2, are too high. Lowering these values for Test 4 would have the effect of increasing the required pore pressures and therefore providing a closer match to the model values.
REFERENCES Lemos, L.J. 1986. The effect of rate on residual strength of soils. PhD thesis, University of London. Parathiras, A.N. 1994. Displacement rate effects on the residual strength of soils. PhD thesis, University of London. Petley, D.J. & Taylor P. 1997. Quick shear with slip of soils against rigid and rough surfaces. Proc. 2nd Pan-American Symp. on Landslides, 2"" COBRAE, 1, Pages 435-442, Rio de Janeiro. Petley, D.J. & Taylor, P. 1999. A simple model to predict pore water pressures during shearing along undulating shear surfaces. Proc. ISShikoku 99. (in press) Taylor, P. 1998. Fast shearing of cohesive soils using ring shear apparatus. PhD thesis, University of Warwick. Tika, T.M. 1989. The effect of fast shearing on the residual strength of soils. PhD thesis, University of London.
The model has been used to investigate the effects of shear rate, undulation geometry, soil depth and total normal stress on pore water pressure generation. The results of this study provide close correlation between current understanding of negative rate behaviour, Parathiras ( I 994) and Taylor (1998), and the pore water pressures predicted by the model, thus:
0
Total Normal Stress (kPa) 95.5 84 82 84
Therefore this paper proposes that undulation induced pore water pressures are a likely cause of potentially catastrophic failures. Given this potential, the model could be used as a starting point for the development of slope stability software that incorporates routines that will calculate such pore pressures. Refinements to the model are required and should include the analysis of more than one strip, the influence of inter-strip forces, the potential for three dimensional drainage and the increased porosity of the shear zone above that of the surrounding soil. A more suitable method for determining the behaviour in the expansion zone is required and could involve the modeling of the soil as a Bingham-Plastic flow using computational fluid dynamics.
6 CONCLUSIONS
@
Rate of Shearing (mdmin) 10 50 300 1000
Negative rate effects increase with shear rate, as do modeled pore water pressures. Negative rate effects increase with undulation height, as do modeled pore water pressures. Negative rate effects decrease under increases in total normal stress, as do modeled pore water pressures.
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Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Apparent cohesion of unsaturated soils as correlated with suction Yongnan Huang Geotechnical Engineering Center, Kiso-jiban Consultrrnts Company Limited, Tokyo,Japan
Kenji Ishihara Department of Civil Engineering, Science University o j Tokyo, Japan
ABSTRACT A series of shearing tests were conducted on loosely compacted specimens using a modified triaxial apparatus which incorporates the measurement of suction. The test results of three different materials revealed that the components of shearing resistance arising from net stress (difference between total stress and air pressure) and suction are independent from each other. Therefore, it was noted that the strength of unsaturated soils can be expressed in the form of Mohr-Coulomb criterion with respect to net stress in which the total cohesion consisting of the effective cohesion and the apparent cohesion attributed to suction. It was observed that the apparent cohesion generally increases with suction at a decreasing rate and reaches the maximum value when suction becomes high enough. By introducing an assumption that the maximum apparent cohesion appears when suction approaches infinity, a hyperbolic function of suction having an initial slope of tan d ' was derived for the apparent cohesion to characterize the non-linearity between apparent cohesion and suction. The validity of hyperbolic function was examined, and values of the maximum apparent cohesion for eight different materials were summarized.
1 INTRODUCTION It is well known that soils in hillsides and embankments are usually unsaturated, where there exists a pressure deficiency in water phase and air phase due to surface tension, referred to as matrix suction or simply suction. Previous studies indicated that suction plays a role to increase the shearing resistance of unsaturated soils (Bishop and Donald, 1961; Bishop and Blight, 1963; Gan, 1988; Toll, 1990). On the other hand, a marked non-linearity of shear strength with respect to suction was pointed out by Escario and Juca (1985) and Frcdlund et al. (1987), and Huang (1994) among others. It may be said that the initial interest on the significance of suction arises from the attempt to extend the effective stress principle to the case of partial saturation (Aitchison and Donald, 1956; Bishop, 1959). The effective stress approach, however, achieved limited success in practice, which in turn led to treating soil suction as another indcpendent stress state variable (Fredlund and Morgenstern, 1977). In this paper, the results of a series of shearing tests conducted on three different kinds of soils are presented and discussed for better understanding the
shearing resistance associated with suction. The test results showed that the strength component attributed to suction, called the apparent cohesion, is independent of the net stress. Based on the experimental observation, a hyperbolic function of suction having an initial slope of tan d ' was proposed to express the non-linearity between the apparent cohesion and suction under the assumption that the maximum apparent cohesion appears when suction approaches infinity. It is noted that the maximum apparent cohesion is an appropriate strength parameter associated with suction. The validity of hyperbolic relation is discussed.
2 EXPERIMENTAL PROGRAM
2.1 Test materials and specinzen preparation Three different materials, clayey sand, volcanic silty sand and artificial clayey silt, which are named as SIMO, UENAE and DL, respectively in this paper, were used in the tests. The physical properties of the soils tested are summarized in Table 1 while the particle size distribution curves are shown in Fig. 1.
75 1
Tablel. Physical properties of soils tested. Index
Mean grain size,
0.12
0.017
0.095
Uniformity coef., Uc
50.0
75.0
Clay content (96)
12.4
3.5 9.5
Silt content (%)
19.8
90.4
28.8
Sand Content (%)
67.8
0.1
58.1
Liquid limit, w (5%)
28.9
NP
NP
Plastic limit. w (%) Max. dry unit weight, 7 dm (gicm') Optimum water
8.7
NP
NP
Djo (mm)
13.1
Figure 1. Particle size distribution of soils tested.
Considering the significance of soil fabric in mechanical properties of unsaturated soils, a close attention was paid to the specimen preparation. All the specimens were prepared by means of compaction at a single molding water to yield the same dry density for each material. Thereafter, the as-compacted specimens, if necessary, were air-dried or sprayer-wetted through the side surface to change the initial values of suction to any specified ones, followed by a certain duration of curing (Huang, 1994). The molding water content was selected as dry of optimum and the dry density was determined to be equal to about 80% of the maximum dry unit weight. It is to be pointed out that small to very small volume contraction or expansion was observed during the process of air-drying and sprayer-wetting, respectively. Hence, the employed method of specimen preparation ensured the identity in soil fabric among the specimens for each material, regardless of the values of suction prior to test.
differential transducer. 2.3 Shearing The conventional triaxial compression test was adopted in the present experimental study. After the isotropic compression in steps under air drained condition, the specimens were compressed axially at constant net confining pressure until the axial strain was greater than 18%. The compression was run at an axial strain rate of 0.05% per minute under air drained but water undrained condition for the specimens of non-zero suction, whereas full drainage was allowed for the specimens of zerosuction, which were obtained by infiltrating free water under imposed all-round pressures.
3 TEST RESULTS AND ANALYSES The typical results of triaxial tests on the specimens subjected to various levels of suction but to the same net confining pressure of 49 kN/m2 are illustrated in Fig.3, where the deviator stress, volumetric strain and suction are plotted against axial strain, respectively. It can be seen that the stress-strain curve corresponding to a higher suction shows a stiffer initial slope, and at the same time, a less volume change corresponds to a higher suction, implying the fact that suction plays a role of stiffening soil structure against external loading. In addition, the specimen of zero-suction shows the lowest shearing resistance as compared with those undergoing the action of suction, and the higher the suction, the higher the shearing resistance. In order to reveal this aspect in more detail, the test
2.2 Appa ra tus The experiments were carried out using a modified version of conventional triaxial apparatus in which water pressure, air pressure, change in overall volume and change in water phase of a specimen can be measured or controlled independently. As illustrated in Fig.2, air pressure is applied to the top of the specimen through a glass fiber filter, and water pressure is measured at the bottom of the specimen through a ceramic disc with an air entry value of 200 kN/m'. On the other hand, the overall volume change in the specimen is determined by monitoring the water level in the inner cell with respect to the reference water table using a pressure 752
results are summarized in the form of deviator stress at failure versus suction as shown in Fig.4. The failure deviator stress was determined as the peak in the stress-strain curve or as the one corresponding to an axial strain of 18 %, because most of the specimens showed work-hardening behavior up to about 18 % of axial strain. It is apparent in Fig.4 that the deviator stress mobilized at failure increases as suction increases for a given net confining pressure, and their relation
Figure 2. Schematic illustration of modified triaxial apparatus.
Figure 3. (a) Deviator stress, (b) volumetric strain and (c) suction versus axial strain.
reveals a marked non-linearity. Of importance is that the curves drawn for the sets of test data of the same net confining pressure are parallel with each other. It is noted that the mentioned parallelism is the case for the three materials tested. This parallelism implies that the increment in the deviator stress due to suction is independent of the level of net confining pressure and vice versa. Therefore, the deviator stress mobilized at failure in the triaxial test can be expressed, from the mathematical point of view, as the simple sum of two functions related to net confining stress and suction, respectively. The feature mentioned above can be interpreted to imply that the frictional component of shearing resistance
with respect to net stress has no connection with suction. It can also be interpreted to imply that the internal friction angle is independent of suction. On the other hand, the contribution of suction to shearing resistance may be classified into cohesion component. Considering the saturated soil as a special case, it is readily understandable that the shear strength of unsaturated soils can be expressed using the MohrCoulomb criterion with respect to net stress in a general way as
753
in which 0 is the total normal stress on the failure plane, U , is the air pressure, c f is the effective cohesion, d ’ is the internal friction angle, and c’ is the strength component attributed to suction, usually referred to as apparent cohesion. Each of the strength components in Eq.1, i.e., the effective cohesion, the friction part and the apparent cohesion, has a clear physical meaning and a particular origin. The apparent cohesion arises from the internal interaction among pore air, pore water and soil particles in unsaturated soil. Because the first two strength components in Eq.1 are identical in both saturated and unsaturated cases, the study of shear strength behavior of unsaturated soils should be concentrated on the point of revealing the characteristics of apparent cohesion. In the case of triaxial tests, the apparent cohesion can be directly separated from the test data using the following equation:
Such obtained results of apparent cohesion are plotted against suction in Fig.5, together with the values of c’ and 6, determined from the data of wetted specimens of zero-suction. As shown in Fig.5, a unique relation between the apparent cohesion and suction is found to exists for each material regardless of the level of net confining pressure, implying little dependency of internal friction angle on suction. Moreover, it is evident that the apparent cohesion shows an increase as suction increases, while their relation is of remarkable non-linearity. It is readily visible that the curves of apparent cohesion VerSUS Suction tend to flatten as Suction increases. Consequently, the apparent cohesion probably reaches its maximum value when suction becomes high enough and then remains almost constant thereafter. In addition, It is noticeable that the magnitude of apparent cohesion is quite different for the three soils tested, aIthough a simiIarity exists in the relation of apparent cohesion and suction. This feature implies that there exists at least one parameter which controls the extent of dependency of apparent cohesion on suction, and it may be treated as the strength parameter associated with suction. The maximum value of apparent cohesion, called the maximum apparent cohesion, c’,,,, seems to be a dominant parameter governing the magnitude of apparent cohesion. It can be said that as soon as the order of the maximum apparent cohesion is known, one may immediately make his mind whether the
Figure 4. Variation of deviator stress with suction.
Figure 5. Relation of apparent cohesion and suction. 754
influence of suction on the shear strength of unsaturated soil has to be considered or not in practical engineering.
4 CHARACTERIZATION OF APPARENT COHESION The similarity in the variation of apparent cohesion with suction, as shown in Fig.5, allows for the possibility of characterizing the relation between apparent cohesion and suction. Let the apparent cohesion be expressed as a function of suction, i.e., cr = f ( u ,
(3)
-Uw)
then, the function f ( U , conditions listed below:
ill,, )
has to satisfy the
Condition (a) means that the apparent cohesion disappears when suction reduces to zero, and condition (b) permits a smooth transition of Eq.l from an unsaturated soil to a saturated soil failure criterion (Escario and Juca, 1985), while condition (c) presents the variation tendency of apparent cohesion with respect to suction. In addition to the conditions mentioned above, it is necessary to introduce at least one parameter which can represent the mechanical property of unsaturated soil associated with suction. Among these, the maximum apparent cohesion, c’,,,,seems to be an appropriate one because it can be considered as the quantitative measurement of available contribution of suction to shear strength of unsaturated soil. Since suction in unsaturated soils may vary within an extremely wide range of value, say from zero to 100 MN/m’ (Croney and Coleman, 1960), it is appropriate to assume that for most cohesive soils, the maximum apparent cohesion appears when suction approaches infinity. This hypothesis is believed to simplify greatly the function for apparent cohesion without losing accuracy. Therefore, an additional condition which the functionf( U , - U , , ) has to satisfy is as follows:
Figure 6. tan d ’/c”versus suction.
Figure 7. Comparison between calculated apparent cohesion and observed one. that the hyperbolic function is a simple one among the others and in agreement with the test data well. Let the hyperbolic function be expressed as
(4) where a , b, c, and d are constants to be determined. Substituting the conditions (a), (b) and (d) into Eq.4 and performing some transformation yields
Nevertheless, there are many functions which can satisfy the aforementioned conditions. It is noted
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Table 2. Summary of the strength parameters. Soil SIMO
C1
dj
kN/m’ 0.0
’
CS
Degree 32.0
kN/m’ 42,s
19.5
DL
1.5
30.9
UENAE
3.3
39.7
6.6
Braehead silt
0.0
33.6
65.6
I Selset clay 1 I Manglashale 1
9.7
11.3
I I
25.1 23.6
1 1
323.2 502.2
(3)
1 1
(4) (4)
Kiunyu gravel
0.0
32.2
32.2
(12)
Glacial till
20.0
25.5
238.5
(10)
I I
It is evident that Eq.5 satisfies the condition (c). Since the internal friction angle of soils falls generally within a narrow range from 20 to 40 degrees, the maximum apparent cohesion becomes the predominant factor governing the magnitude of apparent cohesion. The significant advantage of Eq.5 may be illustrated as the fact that there is only one additional parameter, crm,included in it. c’,,, can be easily estimated by fitting the test data plotted in the form of tan$’(ua -u,,,)/c‘ versus suction with a straight line of unit intercept whose slope is equal to tan$’/cd, , as shown in Fig.6. The values of the maximum apparent cohesion determined in this way for three materials tested are listed in the legend of Fig.6. Detailed re-analysis on the test results available in literature also indicated that the correlation between apparent cohesion and suction can be characterized using Eq.6 with sufficient accuracy (Huang, 1994). The validity of Eq.6 is shown in Fig.7 where the calculated apparent cohesion using Eq.6 plotted against the observed one for eight different materials, three from the present study, and five from literature. As can be seen, all the points are generally concentrated on the 1:1 line in Fig.7 up to a value of 120 kN/m’, implying the efficiency of Eq.6. The values of the maximum apparent cohesion c’,, determined from the test data are summarized in Table 2 together with the effective cohesion and the internal friction angle. It is noted that the maximum apparent cohesion shows a wide range of value, such as from a few kN/m’ up to 500 kN/m’ depending on the type of soils.
5 CONCLUSIONS The results of triaxial tests on three different materials revealed that the components of shearing resistance arising from net strcss and suction are 756
independent from each other. In other words, the internal friction angle is independent of suction. The strength of unsaturated soils, therefore, can be expressed in the form of Mohr-Coulomb criterion with respect to net stress where the total cohesion consisting of the effective cohesion and the apparent cohesion attributed to suction. The test results indicated that the apparent cohesion generally increases with suction at a decreasing rate and reaches its maximum value when suction becomes high enough. It was pointed out that the maximum apparent cohesion involves the quantitative measurement of the contribution of suction to shear strength of unsaturated soil and is an appropriate strength parameter associated with suction. By introducing the assumption that the maximum apparent cohesion appears when suction approaches infinity, a hyperbolic function of suction having an initial slope of tan@’/ci, was derived for the apparent cohesion, to characterize the observed nonlinearity between apparent cohesion and suction. The validity of the hyperbolic function was examined with the present results and the data available in literature, and the values of the maximum apparent cohesion for eight different materials were summarized.
REFERENCE Aitchison, G. D.and Donald, I.B. 1956. E’ffective Stresses in Unsaturated Soils. Proc. 2nd Australia-N. Z. Conf. on SMFE, Christchurch, N.Z., 192-199. Bishop, A. W 1959. The Principle of Effectii,e Stress. Teknisk Ukeblad, 106(39), 859-963. Bishop. A. W. and Donald, I. B. 1961. The E.rperimerita1 Study of Partly Saturated Soil it7 the Triaxial Apparntits. Proc. 5th ICSMFE. 1, 13-21. Bishop, A. W. and Blight. G. E. 1963. Some Aspects of Effective Stress in Saturated and Partly Saturated Soils. Geotechnique, 13(3), 177-197. Croney. D.and Coleman, J. D.1961. Pore Pressure and Suction in Soil.Proc. Conf. on Pore Pressure and Suction in Soils, London, 31-37. Escario, S. and Juca, J . F. T. 1985. Stretigth and Deformutiotz of Partly Satitrated Soils. Proc. 11th ICSMFE, San Francisco, 1, 4346. Fredlund, D. G., Morgenstern, N. R. 1978. Stress state Vnriahle for Umaticrated Soils. J. Geotech. Eng., ASCE, 103(5), 447-466. Fredlund, D.G., Morgenstern, N. R. and Widger, R. S. 1978. The Shear Strength of Unsaturated Soils. Can. Geotech. J.. lj(3). 313321. Fredlund, D. G., Rahardjo, H. and Gan, J. K. M. 1987. Notilineuri~of Strerigtli Envelope for Unsaticrated Soils. Proc. 6th Inter. Conf. on Expansive Soils, New Delhi. 49-54. (10) Gan, J. K. M.. Fredlund, D. G. and Rahardjo, H. 19S8. Detertniriatioti of the Shear Strength Parameters of ati Utisaturarerl Soil using the Direct Sliear Test. Can. Geotech. J., 25(3), 500-510. (1 1) Huang. Y. 1904. Effect of Sitcfion oti Strength atid Deformation Behavior of Utisaturarerl Collapsilile Soils. D. Eng. Thesis, Univ. of Tokyo. (12) Toll, D.G. 1990. A Fratticwrk for Unsaturated Soil Beliuviour. Geotechnique, 40(1), 3 1-44.
Slope Stability Engineering, Yagi, Yamagami & Jiang @ 1999 Balkema, Rotterdam, ISBN 90 5809 0795
Unconfined compression shear strength of an unsaturated silty soil subjected to high total suctions T. Nishimura Ashikaga Institute of Technology,Tochigi, Jujiun
D.G.Fredlund University of Suskutchewan, Suskutoon, Sask., Cunadu
A B S T R A C T T h e r e d o e s not appear t o be data available that s h o w s the relationship b e t w e e n t h e s o i l - w a t e r characteristic curve and experimental shear strength beyond t h e residual s t a t e This paper describes the shear strength o f a compacted unsaturated s i l t y T h i s s t U d y i n vo 1 v e d p er fo 1-1111i n g u n c o n fi n e d s o i I b e y o n d re s i d u a 1 C O n d it i o 11s c o m p r e s s i o n t e s t s on a compacted unsaturated silty soil subjected t o high total suction The soil-water characteristic curves also measured o v e r a wide range o f suctions The highest total suction w a s maximum 9 3 , 6 0 0 k P a c o r r e s p o n d i n g t o a relative humidity of S O 9'0 T h e relationship between shear strength and total suction for the silty soil s h o w s a n essentially horizontal failure surface beyond residual conditions Prior t o the soil r e a c h i n g residual conditions, the failure e n v e l o p e is non-linear 1 INTRODUCTION
U n s at u r at ed n at u I- a 1 s o i 1 s an d art i f i c i a1 1 y c o m p a c t e d unsaturated soils near the g r o u n d s u r f a c e can have high negative p o r e - w a t e r pressure d u e t o evaporation. The ground s u r f a c e is a dynamic boundary, w h i c h is controlled largely by the cl i in at i c cond it io ns . e nv i ronine n t or Geotechnical engineers a r e well a w a r e that e v a p o r a t i v e e v e n t s can greatly exceed for infiltration e v e n t s i n m a n y regions o f the world. Recent studies have evaluated e v a p o r a t i v e rates from soil surfaces. Silvestri, Soulie, Lafleur, Sarkis and Bekltouche ( 1 9 9 0 ) showed that clays were strongly influenced by potential evaporation and result i n settlement I i g htwei gh t structures. on p r o b 1 em s S a t t l e r and Fredlund ( 1 989) demonsti-ated t h a t heave and settlement for expansive c l a y soils a r e influenced by evaporation. Barton (1979) suggested that soil evaporation inay be estimated on the basis of the humidity and water content of the n e a r surface soil. Granger ( I 9 8 9 ) stated t h a t evaporation from unsaturated soil
surfaces is a function o f the actual vapoipressure at the soil s u r f a c e . T h e concept of stress state variables t o d e s c r i b e the behavior of unsaturated soils was introduced by Fredlund and Morgenstren ( I 9 7 7 ) . An e m p i r i c a l , analytical model was developed t o predict the s h e a r strength in terms o f soil suction using a soil-water characteristic c u r v e and sat u I- at e d s h ear strength p a r a in e t e I(Vanapalli, et a l . (1996)). A typical soilwater characteristic curve has one c u r v e foidrying and one c u r v e for the wetting o f a soil. Different saturation stages can be d e fi n e s t h r o U g h t 11e d e sat U I- at i o n p r o c e s s d u e t o increasing soil suction. T h e first f u t u r e is the air entry value. At large increases i n suction, there is a relatively small c h a n g e o f water content at t h e residual zone stage ( i . e . , 1-esidual w a t e r content c o nd it i o 11). Beyond residual soil suction conditions, changes i n the shear strength o f an unsaturated soil have not been well d e f i n e d . T h e c h a n g e i n shear strength beyond residual soil suction conditions ( i . e . , residual zone stage) inay depend on t h e soil type. Laboratory tests are required i n
o r d e r t o e s t i m a t e t h e s h e a r strength and beyond residual water content in unsaturated soil m e c h a n i c s .
2 . P U R P O S E OF T H I S S T U D Y S h e a r strength tests for a soil beyond residual conditions have not been adequately studied T h i s paper describes t h e s h e a r strength b e h a v i o r o f a compacted unsaturated silty soil beyond residual water content c o n d i t i o n s Large total suctions w e r e created i n a c o m p a c t e d silty soil by controlling the relative humidity i n t h e soil T h i s was d o n e i n a relative humidity U nco n fi n e d c o mp re s s i o n tests chain b er w e r e c o n d u c t e d o n unsaturated soil s p e c i m e n s in t h e residual water content range T h e relationship between total suction and s h e a r strength is evident i n the total suction range f r o m 41 kPa t o 93,600 kPa
Fig.1 Relative humidity versus total suction relationship
Table 1 Summary of unconfined compression test results itrain at ire %
17260
3 . TEST PROCEDURE
A silty soil was used i n this test program (i e , a fine-grained cohesionless soil) The statically compacted silty soil s p e c i m e n s had a height o f 100 mm and a d i a m e t e r o f 50 min Initial physical properties o f t h e silty soil specimens had a water content o f 9 6 %, a void ratio of 0 9 4 7 and a d e g r e e o f saturation of 2 7 % All specimens were placed directly into a h u m i d it y r e 1 at i v e t em p e r at u I- e and controlled chamber i n o r d e r t o apply a high total suction T h e chamber could control t h e relative humidity in a range from 20 % t o 9 0 % at a t e m p e r a t u r e o f 3 0 degrees There is a relationship between relative humidity and soils suction (i e , total Fig 1 is suction) a s shown i n F i g 1 plotted using the theoretical model (Fredlund and R a h a r d j o (1993)) The test program selected relative humidifies of Each SS %, SO 9'0, 70 %, 6 0 % and 5 0 % silty soil specimen w a s subjected t o the Total relative humidity for a long time suction values corresponding t o each relative humidity are shown i n Table 1 Soil water leaves t h e soil surface a s result o f evaporation Desaturation o f a soil o c c u r s a s t h e d r i e s When t h e weight o f each soil specimen underwent no further
compresive
humidity %
80
30129
32 4
0 12
70
48158
38 8
60 50
68972 93,590
58 3 59 2
0 19 0 15 0 32
Initial condition
41
a
0 65
28
change, i t was a s s u m e that the soil had come t o equilibrium at the selected relative humidity Each soil specimen was i n a residual condition After soil specimen had reached equilibrium, an unconfined compression test was conducted at residual a rate o f axial strain o f 0 5 mni/min At the end of the unconfined compression test, t h e water content of t h e complete soil specimen was measured i n order to evaluate the soil-water characteristic curve T h e soil-water characteristic curve is a measure o f t h e available soil water at a particular soil suction The soil-water characteristic curve for t h e silty soil w a s evaluated using a pressure plate apparatus (i e , pressure plate method), glass desiccators containing saturated salt solutions (i e , vapor equilibrium technique) and relative humidity t e c h n i q u e over the entire soil suction range The
758
Fig.2 Stress-strain curve for the on unconfined compression test with an Initial matric suction of 41 kPa
Fig.3 Stress-strain curve for the unconfined compression test at a relative humidity of 88% or a total suction of 17,260 kPa
pressure p l a t e method measures t h e soil w a t e r at a variety o f m a t r i c suction values. T h e air pressure in t h e pressure plate a p p a r a t u s w a s increased until a maximum 182 k Pa . The water content of c o r r e s p o n d i n g t o higher values o f total s u c t i o n w a s determining u s i n g b o t h the v a p o r equilibrium t e c h n i q u e and t h e relative humidity t e c h n i q u e . Small soil s a m p l e s w e r e placed into each glass desiccators, and w a t e r contents were measured corresponding t o t h e total suction established in the d e s i c c a t o r s .
Fig.4 Stress-strain curve for the unconfined compression test at a relative humidity of 80% or a total suction of 30,129 kPa
4 . LABORATORY T E S T RESULTS Geotechnical engineers o f t e n required an estimation o f t h e shear strength o f soils at l o w water contents Previous research w o r k on unsaturated soils has not performed shear strength tests at residual w a t e r content conditions This study reports the results of unconfined compression tests at low water contents on a silty soil For c o m p a r i s o n purpose, the initially compacted silty soil with a matric suction o f 41 kPa, w a s tested in an unconfined compression test Stress-strain c u r v e s obtained from the unconfined compression tests a r e shown in F i g s 2 , 3 , 4 , 5 , 6 and 7 T h e stress-strain c u r v e for t h e initial compacted silty soil is s h o w n in Fig 2 Table 1 provides a s u m m a r y o f the unconfined compression t e s t results The compacted silty soil indicates a smooth stress-strain curve a s s h o w n in F i g 2 T h e maximum deviator s t r e s s is reached at an axial strain o f 0 6 5 YO T h e compacted silty soil specimens with a
high total suction s h o w s a distinct peak on the stress-stain c u r v e After reaching t h e in a x i in u m d e v i at o r stress , t 11 e stress - st r a i n curve decreases rapidly Failures occur suddenly i n t h e s p e c i m e n s with a high suction The axial strain at failure for the dried specimens is lower than that o f the initially compacted silty soil T h e value of the strain at failure varies with t h e water content condition T h e shear strength of a compacted silty soil increases slightly at high total suctions 5 DISCUSSION O F RESULTS
The shear strength o f an unsaturated soil is related t o soil-water characteristic c u r v e . The soil-water characteristic curve the d escr i b e s re 1 at i o n s h i p b et w ee n available water i n t h e soil and t h e soil suction, for drying and wetting. T h e shear
759
predict the permeability and s h e a r strength function for an unsaturated soil. T h e soilwater characteristic c u r v e model can be written as an equation as proposed by Fredlund and Xing ( 1 9 9 4 ) ( F i g . 8 ) . Model parameters for the best-fit soil-water characteristic curve for t h e silty soil a r e shown in Fig. 8 . A silty soil has an air entry value o f 3 0 k P a . B e y o n d a suction o f 200 kPa, the soil enters t h e residual state. It is well-known that t h e r e a r e different stages o f desaturation defined b y t h e soilwater characteristic c u r v e . Vanapalli, et al. (1996) suggested four stages as fo 1 1o w i n g : b o u n d ar y effect stage, p r i in a r y transition stage, secondary transition s t a g e and residual stage. T h e soil is essentially saturated in the boundary effect s t a g e . All t h e soil pores a r e filled with water. The soil starts to desaturate i n t h e primary transition stage. T h e w a t e r content in the soil reduces significantly with increasing i n suction. The air-entry v a l u e for the soil lies between t h e boundary effect stage and the primary transition. l n t h e s e c o n d a r y transition stage, the a m o u n t o f water between the soil particle o r a g g r e g a t e contacts reduces a s desaturation c o n t i n u e s . The water meniscus area i n c o n t a c t with the soil particle or aggregates begins t o become discontinuous. The rate o f d e c r e a s e i n water content, t o a change i n suction i n this stage, is less than that i n t h e primary transition s t a g e . There is little water left i n soil pores when the soil reaches the residual s t a t e . T h e water content of t h e u n sat u r a t ed so i 1 re in a i n s re 1a t i v e 1 y c o n s t ant i n the residual s t a g e . Air a l m o s t occupies all t h e soil pores. The w a t e r meniscus i n contact with the soil particles is not continuous and m a y be very small. T h e r e is a little water left in soil pores. Fig. 9 shows the relationship between the shear strength ( i . e . , unconfined compressive strength) and total suction for the residual condition i n t h e unsaturated silty soil. T h e shear s t r e n g t h has a slightly increase i n strength with increasing o f total suction. The ratio o f the increase i n shear strength t o an increase i n total suction translates t o an a n g l e o f 0 . 0 2 degrees. There is a negligible i n c r e a s e i n shear strength because t h e a m o u n t o f w a t e r i n the soil pores is vei-y small T h e effect of total suction on t h e s h e a r s t r e n g t h is
Fig.5 Stress-strain curve for the unconfined compression test at a relative humidity of 70% or a total suction of 48,158 kPa
Fig.6 Stress-strain curve for the unconfined compression test at a relative humidity of 60%
or a total suction of 68,972 kPa
Fig.7 Stress-strain curve for the unconfined compression test at a relative humidity of 50% or a total suction of 93.590 kPa
strength o f an unsaturated soil is related t o t h e a m o u n t o f water i n the void o f t h e soil T h e soii-water characteristic curve for t h e silty soil is shown Fig 8 Several soii-water characteristic curve m o d e l s have been proposed t o empirically
760
1 - Calculated water content
I
1
~
0 Measured water content (Vapor equilibrium technique) i A Measuredwater content (Pressure plate method) j 0 Measured water content (Relative humidity i equlllbrlumue) -__-_____ ~
50 45
r
,
40
g 35
5 c
5
30 25
1-
-----
I
----:.-9 -
Pamameter --Water content at saturation = 31 Oh, Air entry value = 30 kPa Total suction at residual = 200 kPa, Best-fit soil parameters for Fredlund % and Xing (1994) model
60
v,
50
,
Q
I
40 0
I
I
20000
0 0
4
10
100
1000 I0000 Total suctin kPa
100000
40000 60000 80000 100000 Total suction kPa
1000000
Fig.9 Relationship between unconfined compressive strength and soil suction in the residual state
Fig 8 Soil-water characteristic curve for the srlty soil m
4
negligible It is concluded that the shear strength f o r a residual water i n t h e unsaturated silty soil, remain relatively constant The shear strength envelope is postulated in Fig 1 0 for the initially compacted silty soil at a low inatric suction Before the soil suction u p the 4 1 kPa reaches t h e air-entry value, the soil is The essentially in a saturated state failure e n v e l o p e will be tangent t o an angle o f internal friction for the saturated silty soil T h e a n g l e o f internal friction o f silty soil used in this study was 43 degrees Beyond t h e air-entry values, the effect of soil suction translating t o shear strength A non-linear increase in shear decreases strength is shown i n Fig 10 Gan, Fredlund and Rahardjo ( 1 988) observed non-linearly i n t h e failure envelope with respect t o inatric suction for a compacted glacial till w h e n using inultistage direct shear t e s t s T h e tangent of the failure envelope decreases significantly at inat1 ic suctions in t h e range o f SO-100 k P a The a n g l e with respect t o matric suction reaches a fairly constant value when the matric suction reaches SO0 k P a Since t h e shear strength versus total suction relationship was computed a s 0 03 degrees i n F i g 9, the failure surface
100
1 I
I
I
A
1
I
60
/ / i
I
1
/
I
43degrees
40
20 I
/
Air entry value of 30 kPa
0 J
0
20
40
60
80
100
Soil suction kPa Fig.10 Relationship between unconfined cornpresive strength and rnatric suction
indicates a horizontal total suction.
relationship
with
6.C O N C L U S I O N S This paper presents unconfined compression test results and the measurement of the soil-water characteristic curve for a c o m p a c t e d unsaturated silty soil C h a n g e i n shear strength under residual c o n d i t i o n s a r e
76 1
discussed. T h e c o m p a c t e d unsaturated silty soil was b r o u g h t t o equilibrium at relative h u m i d i t i e s o f 88 %, SO %, 70 %, 6 0 % and 50 % . T h e d e v i a t o r stress for t h e soil under residual conditions, reached t o maximum After the v a l u e at a low axial s t r a i n . m a x i m u m d e v i a t o r s t r e s s was reached, the strength suddenly d e c r e a s e d . B e f o r e t h e total suction reached its residual state, the silty soil indicated a non-linear failure envelope. T h e s h e a r strength remained constant under residual conditions.
the prediction o f shear strength with respect t o soil suction. Canadian Geo techni cal Journal, Vol. 3 3 , p p . 3 7 9 392.
REFERENCES B a r t o n , I . J . 1 9 7 9 . A parameterization of the evaporation from non-saturated surface. Journal o f Applied Meteorology, Vol. 1 8 , pp.43-47. F r e d l u n d , D . G and Morgenstern, N . R . 1977. S t r e s s s t a t e variables for unsaturated s o i l s . Journal o f t h e Geotechnical E n g i n e e r i n g D i v i s i o n , ASCE, 103(GT5), pp , 4 4 7 - 4 6 6 , F r e d l u n d , D . G . and Rahardjo, H . 1993. Soil M e c h an i c s for U n s at u rated S o i 1 s , J 0 HN WILEY & SONS, INC. 517pp. Fredlund, D . G . and Xing, A . 1994. Equation for the soil-water characteristic curve. C an ad i an Geotechnical Journal, Vol.; 1, p p . 5 2 1 532. Gan, J - K . M . , Fredlund, D . G . and Rahardjo, H . 1988. Determination of the shear strength parameters o f an unsaturated soil using t h e direct s h e a r t e s t . Canadian Geotechnical Journal, Vo1.25, p p . 5 0 0 510. Granger, R . J . 1989. Evaporation from natural non-saturated surface. Journal o f Hydrology, Vol. 1 1 1, p p . 2 1 - 2 9 . Satter, P and Fredlund, D . G . 1989. Use o f thermal conductivity sensors to measure inatric suction i n t h e laboratory. C a nad i an G e o t e c h n i cal J o u rn a1, VO1.26 , pp.491-498 Silvestri, V., Soulie, M . , Lafleur, J . , Sarkis, G . and B e k k o u c h e , N . 1990. Foundation problems in champlain clays during d r o u g t s . 1 :Rainfall deficits i n Montreal ( 1 9 3 0 - 1 9 3 8 ) . Canadian Geotechnical Journal, Vo1.27, p p . 2 8 5 - 2 9 3 . Vanapalli, S . K . , Fredlund, D . G . , Pufahl, D . E . and Clifton, A.W. 1996. Model for
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Slope Stability Engineering, Yagi, Yamagami & Jiang G 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Shear strength mobilization in shear box test under constant volume I. Kobayashi, H.Ohta & M. Hirata Tokyo Institute of Technology, Japan
A. Iizuka Depurtment ofArchitecture and Civil Engineering, Kobe Uiiiwrsity,Jcipan
Abstract: The specification of strength parameters is an important task in the design work of slope stability. Usually laboratory shear tests are employed to determine the strength parameters. The specimen in the laboratory tests is intended and generally assumed to represent a single point in the in-situ soil medium. And, the strength parameters obtained are interpreted as material properties of the soil element. However, since the uniformity of stressktrain distribution within the specimen is hardly achieved during shearing, the strength parameters thus obtained cannot be essentially regarded as material properties. In this paper, the shear box test under the condition of constant volume is considered and the distribution of stress/strain within the specimen is rigorously examined through numerical simulations based on finite incremental deformation theory. The mobilization of strength is explained associated with the development of shear bands in the specimen during shear.
1. Introduction The specification of strength parameters is a key subject in the slope stability analysis. The soil specimen sampled from the site is subjected to the laboratory shear test to determine the strength parameters. The soil specimen in the laboratory is intended and generally supposed to represent a single point in the soil medium. The uniformity of stress and strain distribution within the soil specimens is assumed. However, in reality, the localized deformation, e.g. slip lines, develops inside the specimen with shear and the uniformity of stress and strain within the specimen is broken. In the strict sense, the strength obtained from such a shear test is not a material property but a solution being obtained under the boundary condition of the laboratory test. Therefore, in general, the strength obtained from the laboratory shear test would be different from the in-situ strength mobilizing a t the site because the geometric and stress conditions in the laboratory test are different fiom those a t the site. The question is how the “strength is different.
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To address this problem, an attempt to reveal the mechanism of strength mobilization in the shear tests has been made throughout the numerical simulation (Iizuka, Kobayashi and Ohta, 1999). Morgenstern and Tchalenko (1967) carried out a series of shear box tests under the constant vertical pressure for remolded kaolin clays and investigated the localized deformation developing inside the specimen with shear. Nishigaki and Mikasa (1979) also discussed the development of localized deformation. They found that the micro fractures of echelon shape developed in the specimen at the earlier stage of shearing and grew to the slip lines. In this paper, such development of localized deformation is rigorously examined throughout the soillwater coupled numerical simulation of shear box test.
2. Constant Pressure Shear Box Test 2,1 Test condition The experimental investigation on formation of slip lines by Morgenstern and Tchalenko (1967) is
SBT under constant vertical pressure shear rate : 0.003mdmin remolded kaolin clay, PI=36 % pre-consolidation pressure: dv0= 430.2kPa vertical pressure:
Fig.2 load and displacement relation of SBT by Morgenstern and Tchalenko (1967)
dui= 215.lkPa
Fig. 1 SBT by Morgenstern and Tchalenko (1 967) introduced in this section. They carried out a series of shear box tests (SBT) under the condition of constant vertical pressure for remolded kaolin clays of which plasticity index is 36. The slurry kaolin clay (water content is 100 %) was preconsolidated with the effective overburden pressure of 430.2 kPa in the oedometer of which diameter is 228.6 mm. Their main purpose of experiment was to investigate the strength anisotropy. Then two types of cuboid specimens were prepared for it: one was trimmed parallel to the bedding plane of preconsolidated clay materials and the other was perpendicular to it. The specimens of 6 0 x 6 0 ~ 2 5mm were sheared in the shear box under the constant vertical pressure of 215.1 kPa a t fairly slow shear rate of 0.003 mm/min against the standard rate of 0.005 mm/min. The test condition is summarized in Fig.1.
2.2 De velopment of localized deforma tion The stress and displacement relations obtained from their SBT are shown in Fig.2. All six specimens (V1 to V6) were sheared under the same condition until each prescribed shear displacement (Fig.2) and the specimens were removed from the shear box to observe the slip lines developing inside. Fig.3 indicates photographs of thus observed slip lines. The symbols of V1 to V6 mean the degree of shearing. Herein, the “pre-cut plane” is a special case that the slip line was artlficially made in advance along the expected shear plane. According to Fig.3, the slip lines appear from both corners (Vl) and gradually develop slantwise toward the inside of specimen (V2 -+V3). The slip lines are connected together and the undisturbed region of diamond shape is formed in the middle part of the specimen (V4). After that, the softening behavior seems to be prominent (V5,
764
Fig.3 shear band development in the specimen observed by Morgenstern and Tchalenko( 1967) V6). It can be said that the mobilization of strength of specimen closely relates to the development of slip lines with shearing. Pradhan e t al. (1995) state that the formation of diamond shape surrounded by slip lines is attributed to the operation in experiment. Namely, they state that the formation of diamond shape is due to the rotation of upper
Fig.4 analysis conditions and input parameters shear box even if the amount of rotation is quite small. However, we do not agree with their statement. The formation of diamond shape is not the sort of thing that is avoidable by some improvement of test equipment and/or experimental operation but essential phenomenon in the direct shear type of tests. Because i) the slip line (theoretically discontinuity of velocity gradient) has nothing to do with the failure line (plane) a t which the stress failure condition is satisfied (see, Yatomi e t al., 1989). The direction of slip line does not necessarily correspond with that of stress failure line. ii) the direction of slip line depends on material properties of clays. The formation of slip lines is the sort of thing that is just obtained by solving the boundary value problem. In this paper, the shear box test (SBT) is considered as a n initial boundary value problem and the simulation of shear band formation is carried out. Herein, the incremental finite deformation theory extended in the field of sowwater coupled two-phase mixture is employed.
model employed is Cam-clay (Yatomi et al., 1989). Therefore, the term of “anisotropy” is out of our scope here. The results of simulation are compared with those of experiment by Morgenstern and Tchalenko (1967), which is introduced in the previous section. The numerical simulation is carried out under the plane strain condition. The model of SBT and input parameters used in the simulation is summarized in Fig.4. The size of the model is 60 mm long and 26 mm high. Since the plasticity index of kaolin clay employed in the experiment is reported to be 36, input parameters shown in Fig.4 are estimated from the plasticity index by following the instructive chart proposed by Iizuka and Ohta (1987). The chart by Iizuka and Ohta is designed to make it possible to determine input parameters of the constitutive model of Camclay type from the plasticity index. The parameters representing the stress history of specimen are set to be the same as in the experiment by Morgenstern and Tchalenko (1967), i.e., the preconsolidation pressure, otv0 is 430.2 kPa and the effective overburden pressure, olv, is 215.1 kPa. Constant volume during shear is assumed in the simulation because it is difficult to apply the constant vertical pressure over the boundary of the upper shear box without any rotation of it. Therefore, the boundary condition in the simulation is, in the strict sense, different from that in the experiment. The hydraulic condition in the simulation is that all
3. F.E. simulation of SBT 3.1 Boundary Value Problem of SBT The sowwater coupled F.E. program has been newly developed based on the incremental finite deformation theory (code: DACSAR-F, see, Iizuka et al., 1998, Kobayashi et al., 1999,). The constitutive
765
Fig.5 distribution of localization (case-1)
Fig.6 distribution of localization (case-2)
boundaries are set to be impermeable but pore water is allowed to flow within the specimen depending on the coefficient of permeability. As to the geometric boundary condition, since the geometric restriction of shear box tests is not obvious, then two cases are considered here. One is that all boundaries (a-b, cd, d-e, e-f, g-h and a-h in Fig.4) are fixed in ydirection (case-1) except the spacing (b-c and f-g), and the other is that both side boundaries (a-b, c-d, e-f and g-h) are released in y-direction (case-2). The shear process is simulated by giving displacement in x-direction to the boundaries of upper shear box (c-d, d-e and e-f) at the constant rate of 0.003 mm/min. The spacing of 2.0 mm between the upper and lower shear boxes is assumed to secure the stability in numerical computation. Furthermore, in order to avoid the difficulty arising from the stress concentration at the corner, middle nodal points are shifted a s shown in Fig.4 (Cook e t al., 1989).
simulation are shown in Figs.5 and 6, which are results of case-1 and case-2, respectively. The distributions of deviatoric strain, volumetric strain and excess pore water pressure, when the shear displacement reaches 8 mm in case-1 and 12 mm in case-2, are compared. Herein, in case-1, the iterative computation did not converge in a time increment of step when the g v e n shear displacement exceeds 8 mm. Much difference in localization pattern of shear deformation (deviatoric strain distribution) is not seen between both cases and the pattern of localized deformation (formation of shear band) observed in the experiment is successfully simulated. It is found from Figs.5 and 6 that the dilation occurs in shear bands a s can be called “dilatancy localization (Iizuka et al., 1998)”, being common to both of case-1 and case-2. However, the distribution. of excess pore water pressure is much different between cases. I n case1, the negative excess pore water pressure develops in the specimen and concentrates a t the middle of specimen. I n case-2, on the contrary, the positive
3.2 Development of shear bands The localization phenomena obtained from F.E.
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Fig.7 effective stress paths in specimen (case-2) pore water pressure develops but its concentration is not seen. Small difference of geometric restriction in the test seems to influence the stress distribution within the specimen. Actual shear box tests are surmised to be between extreme cases of case-1 and case-2. Effective stress paths a t several representative points during shear are shown in Fig.7. They are the results of case-2. I t is found that the strain softening with dilation occurs inside the shear bands. This is consistent with the results by Asaoka et al., 1994 and Kobayashi et al., 1999. The average stresses ( z and 0’)and displacement relations are shown in Fig.8 (case-1) and Fig.9 (case-2) with transition of localized deformation patterns. The average stresses are calculated from the reaction forces working a t F.E. nodes against the given displacement. These “average stress” is only measurable stresses in actual laboratory tests. I t is understood that the growth of shear bands with shear closely relates to mobilization of peak shear stress (strength of specimen) and the softening behavior after the peak.
3.3 interpretation of
Fig.8 apparent stress and dlsplacement relation with transition Of shear band formation
4’
The strength parameters are discussed here associated with the development of localized
Fig.9 apparent stress and displacement relation with transition of shear band formation (case-2)
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Fig. 11 apparent effective stress path (case-2)
Fig. 10 apparent effective stress path (case-1) deformation. The apparent effective stress path of the specimen as a whole (namely, the relationship of average shear stress and average effective normal stress calculated from reaction forces against the displacements given to the specimen) is shown in Fig.10 (case-1) and Fig.11 (case-2). Mohr’s circles are also drawn in the figures. These “average stresses” correspond to the stresses that are measured through the load cell in actual laboratory shear tests. Two stress paths in Figs.10 and 11 do not have much difference in their shapes and show typical ones of lightly over-consolidated clays. The hfference in the geometric restriction considered here does not influence the shape of apparent effective stress path. However, the values of effective internal friction angles specified from Mohr’s circle in terms of the “average” stresses are affected by the geometric restriction as shown in Figs.10 and 11. Moreover, thus specified effective internal friction angles seem to have essentially nothing to do with the 4’ (=26.6 degree) used in the constitutive model as a material property. I t can be said that the 4’ determined from the shear box test should be distinguished from the effective internal friction angle as a material property.
4. Concluding remarks This paper describes the mobilization mechanism of strength in the shear box test. The development of localized deformation within the specimen during shear is examined and the mobilization of strength of the specimen is discussed. The experimental investigation by Morgenstern and Tchalenko (1967) is cited and compared with the numerical simulation. The numerical simulation well explains the formation of shear bands observed in the experiment. It is found that the strength of specimen closely 768
relates to growth of shear bands (slip lines) within the specimen. Furthermore, the effective internal friction angle is discussed. I t is shown that the 4’ determined from the shear box test should be distinguished from the effective internal friction angle as a material property.
References Asaoka,A., Nakano,M. and Noda,T. 1994. Soil-water coupled behaviour of saturated clay nearlat critical state, Soils and Foundations, vo1.34, No.1, pp.91-105. Cook,R.D., Malkus,D.S. and Plesha,M.E. 1989. Concepts and applications of finite element analysis. John WiIey and Sons, pp.247-250. Iizuka,A. and Ohta,H. 1987. A determination procedure of input parameter in elasto-viscoplastic finite element analysis, Soils and Foundations, vo1.27, No.3, pp.71-87. Iizuka,A., KobayashiJ. and Ohta,H. 1998. Dilatancy localization in clay specimen under shearing. Proc of 4tt’ Int. Workshop on Localization and Bifurcation Theory for Soils and Rocks, pp.345-353. Iizuka,A., KobayashiJ. And Ohta,H. 1999. The numerical simulation of strength mobilization in shear box test, Journal of Geotechnical Engineering, JSCE, (under submitting) Kobayashi,I,, Iizuka,A. and Ohta,H. 1999. The transition of localized deformation mode developing in the normally consolidated clay specimen. Journal of Geotechnical Engineering, JSCE, No.6 1 7 l m - 4 6 ,1~ 18, ~ (in Japanese). Morgenstern,N. and Tchalenk0,J.S. 1967. Microscopic structures in kaolin subjected to direct shear. Geotechnique,Vol. 17, pp.309-328. Nishigaki,Y. and Mikasa,M. 1979. Interpretations and applications of soil exploration and test results, Fundamental Engineering Library 4, JSGE, No.4, pp.175-215, (in Japanese). Pradhan,T., Hongo,T. and Mizukami,J. 1995. A discussion theme on direct shear test of soil, Reports of committee II , Proc. of Symposium on Methods and Applications of Direct Shear Tests, pp.12-21, (in Japanese). Yatomi,C., Yashima,A., Iizuka,A. and Sano,I. 1989. General theory of shear bands formation by a non-coaxial Camclay model, Soils and Foundations, Vo1.29, No.3, pp.41-53.
Slope Stability Engineering, Yagi, Yamagami& Jiang 0 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Undrained shear strength of unsaturated compacted clays VSivakumar & I.G. Doran School of Civil Engineering, Queen’s University of Belfast, UK
ABSTRACT: Engineering problems associated with unsaturated soils, whether they are due to natural drying or the compaction process, extend over an enormous range. Typical problems associated with these soils are excessive settlement (or heave) and loss of shear strength during saturation. Reported in this paper is the influence of the compaction procedure on the undrained shear strength of compacted fills. A wide range of samples was prepared using different moisture content, type of compaction and compactive effort. Initial suction caused by the compaction procedure was measured using the pressure plate apparatus. On completion of the suction measurements the undrained shear strength of each sample was determined by shearing to failure. The results indicated that the type and the amount of compaction has a very marginal influence on the initial suction. In contrast the compaction moisture content has a significant influence on the initial suction. The data also indicated a possible relationship between the compaction moisture content and undrained shear strength.
situations are negative pore water pressures (suctions) that are created in the soil. The problems received most involving suction * which have attention are collapse or swelling in clays and loss of strength upon wetting. Traditionally unsaturated soils were considered to result from a drying process caused by lowering the water table in the ground. However the current definition extends to cover unsaturated soils resulting from various sources: gas generation in the offshore environment or in organic subsoils and fills where fine and coarse materials are compacted for civil engineering constructions. Compacted fills are placed at close to the optimum moisture content in order to attain maximum dry density. This inevitably leaves the soil in an unsaturated state and subsequent loading and wetting processes can have a detrimental affect on the mechanical behaviour of the compacted fills. Detailed study and research into the fundamental properties of unsaturated soils leading to a greater understanding of these materials is of paramount importance in the design, construction and use of man-made fills. It has been acknowledged that current codes of practice are insufficiently comprehensive to deal with the problems associated with unsaturated soils.
Z INTRODUCTION Since the study of soil mechanics began in the eighteenth century, through to the twentieth century and the theories developed by Karl Terzaghi in “Erdbaumechanik”, soil has usually been treated as a two phase material, minerals and water. It was on this generalized basis, that Terzaghi formulated the principle of effective stress as given by equation 0‘ = CY - uw. Evidently since much of the developed world enjoys a temperate climate, resulting in generally saturated soil conditions, research has been biased toward problems involving saturated soils. Since the 1950’s research has been extended to unsaturated soils, representing them as three phase materials containing water, air and minerals. With these three phases, the theoretical back ground and associated experimental procedures required for an understanding of unsaturated soil behaviour are intrinsically more complex than those required for saturated soil behaviour. As a result, the ability to synthesise unsaturated soil mechanics has lagged behind its saturated counterpart. The types of problems of interest in unsaturated soil mechanics are similar to those in saturated soil mechanics. Common to all unsaturated soil
769
critical state line on the q:p plane assuming that the soil reaches a critical state at failure. It is has been experimentally established that suction influences the intercept p(s) of the critical state line although the influence of suction on slope M was marginal. The definition of critical state, at least for saturated soils, implies that the initial structure of the soil is of no importance when the soil is taken to failure by subjecting it to sufficient shealdeformation. The validity of this claim in relation tq unsaturated soil was recently examined by Wheeleland Sivakurnar (in press). Figure 1 shows the, relationship between deviator stress and mean ne!. stress at failure obtained from controlled suction, tests on samples of compacted kaolin having widely different initial structures arising from varying degrees of compactive effort, different compaction methods and varying moisture content. It appears that the different initial structures which result from the different compaction procedures have no significant effect on the deviator stress at the critical state.
2 STRESS-STRAIN BEHAVIOUR A major development in the study of unsaturated soils was the introduction of two stress state (Matyas and variables (0-U,) and (U,-U,). Radhakishna, 1968). It is now generally accepted that the volume change and shear strength characteristics of unsaturated soil can be expressed as a function of these two stress state variables. In recent research work the use of the two stress state variables has been extended to more complex mathematical models (Sivakumar 1993), (Wheeler and Sivakumar 1995) and (Alonso, Gens and Josa 1990) thus embracing the elasto-plastic behaviour of unsaturated soils into a single framework. The following variables have been identified as essential parameters in order to develop a rigorous analysis in dealifig with the problems in unsaturated soils.
-03) 9 =(.I s = u , -U, v=l+e v = 1 + e,"
where p, q, s, v and vw are the mean net stress, deviator stress, suction , specific volume and specific water volume respectively. Alonso, Gens and Josa (1990) proposed an elastoplastic constitutive framework for unsaturated soil. A similar framework was reported by Sivakumar (1993) and Wheeler and Sivakumar (1995) in which an attempt was made to extend the modified Cam Clay model to unsaturated soils. This framework assumes the existence of normal compression and critical state surfaces in (p:v:s), (q:p:s) and (p:v,,s) spaces. Fredlund, Morgenstern, Widger (1978) extended the Mohr Coulomb failure criteria in order to establish a relationship for shear strength of unsaturated soils as a function of two stress state variables by following relationship:
z= cl+on tan($') + s tan($")
Figure 1 Deviator stress versus mean net stress at critical state The purpose of this paper is to consider problems associated with the suction created by the compaction procedure and the strength of unsaturated soils when subjected to shearing at constant specific water volume.
(6)
where c' and $' are the cohesion and friction angle in a saturated condition and $b is the angle of internal friction with respect to suction s. Re-interpretation of the above relationship in terms of the stress parameters given by Equations (l), (2) and (3) leads to the following form for the deviator stress at the critical state: (7) 4 = MP + A s ) where M and p(s) are the slope and intercept of the
3 EXPERIMENTAL WORK, RESULTS AND DISCUSSION 3.1 Material Speswhite kaolin in powdered form was used for preparing samples. The liquid limit and plastic limit of the kaolin were found to be approximately 72% and 38% respectively. The specific gravity of the kaolin was 2.65. Previous research on this material (Sivakumar 1993) has established that the value of
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M in Equation 7 = 0.93. The standard Proctor compaction curve is shown in Figure 9. 3.2 Procedure f o r preparing samples. The variables considered in the testing programme were: (a) compaction moisture content (b) type of compaction (which includes both static compaction and dynamic compaction) (c) compactive energy (varying compaction pressure in static compaction and varying hammer weight in the case of dynamic compaction). Prior to the sampling process, a known amount of kaolin powder was mixed with the required amount of water to achieve a pre-selected value of moisture content. Samples were prepared in a specially designed 50 mm diameter and 150 mm height compaction mould. Samples were compacted in 9 layers to ensure uniformity along the sample length. In the first series of tests samples were statically compressed to various compaction pressures (200 Wa, 400 kPa, 800 kPa and 1200 kPa) at a moisture content of approximately 24%. In the second series of tests samples were dynamically compacted by dropping a hammer through a fixed distance at a moisture content of approximately 24 %. The hammer weights used were 175g, 350 g and 525 g. In the third series of tests samples were statically compacted to 400 kPa of stress and the compaction moisture content was varied from 24% to 34%. In the case of static compaction a known amount of moist kaolin was placed in the mould and compressed one dimensionally at a constant rate of axial strain until a pre-selected compression pressure was achieved. 3.3 Suction measurements Suction in the samples after preparation was measured using the pressure plate apparatus illustrated schematically in Figure 2.
The apparatus consists of a high air entry filter with an air entry value of 1500 kPa and pressure
transducers to read the air pressure applied in the chamber and the water pressure in the drainage line. Successful application of the pressure plate apparatus requires proper saturation of the porous stone before each test and this was achieved by adopting the procedure described in Sivakumar (1993). Just before placing the sample on the high air entry filter, the surface of the filter was carefully wiped to remove all the excess water from the surface. When placing the sample on the stone a small amount of pressure was applied to ensure a proper contact between sample and filter. Then the chamber was assembled and pressurized with regulated dry air and the pressure held constant until the end of the test. During this procedure the drainage line was closed and the pore water pressure was monitored.
Figure 3 Typical response from pressure plate apparatus Figure 3 shows the measured pore water pressure plotted against time. The pore water pressure rose to a maximum value approximately equal to the air pressure applied in the camber within a short period of time and then dropped towards an equilibrium value. The steady value is normally achieved within a period of about 5 hours. The magnitude of the suction in the sample after compaction is the difference between the pore air pressure applied in the chamber and the final pore water pressure in the drainage line. Figure 4 shows the variation of the initial suction with the compaction moisture content for a range of samples compacted at 400 kPa. Also included in Figure 4 are the suction measurements obtained on other samples compacted close to the moisture content of 24% but to varying degrees of compactive effort or type of compaction. It is evident that compaction moisture content has a substantial influence on the initial suction but that the influence of compactive effort or type of compaction has a maginal influence on the measured suction values. 771
In Figure 5 the initial suctions are separately plotted in order to examine the influence of compactive effort and type of compaction on the initial suction. The circular data points represent dynamic compaction for which the scale of the compactive effort is marked on the top of Figure 5 and the triangular points represent static compaction, for which the scale is marked on bottom of Figure 5. The amount of compaction (both dynamic and static) seems to have little influence on the initial suction.
subjected to undrained shearing. Each sample was subjected to 200 kPa of confining pressure prior to shearing. During the application of the confining pressure and shearing no drainage of pore fluids was allowed. This implied that the specific water volume within the sample remained constant. No attempt was made to measure pore water pressure or pore air pressure during shearing and the most practical purposes the excess pore air pressure can be assumed to be zero. Figure 6 shows the stress strain behaviour of samples tested in each category in which one of the variables was changed (compactive effort or compaction moisture content or type of compaction). Figure 6a shows the relationship between deviator stress and axial strain for the range of samples prepared by static compaction at a moisture content of 24%. It appears that the sample Compacted to 200 kPa of vertical pressure exhibited plastic behaviour from the start of shearing. In contrast the sample compacted to 1200 kPa exhibited elastic behaviour throughout most of the shearing process. Estimated values of Young modulus E are tabulated in Table 1 and it appears that the elastic modulus is strongly influenced by the compaction pressure.
Figure 4 Influence of moisture content on the initial suction
Table I . Young Modulus obtained from stress-strain curves
1 Static Compaction I Dynamic (J” (kPa) 200 400 800
Figure 5 Influence of compac ion pressure on the initial suction By means of a series of tests t was confirmed that static compaction at 400 kPa produces a sample of similar density to that of dynamic compaction using a 175 g of hammer falling through 300 mm for a total of 81 blows per sample. The differences in the initial suction at higher compaction indicated in Figure 5 may be due to the fact that dynamic compaction causes a much larger amount of shaer deformation than static compaction. 3.4 Undvairzed shearing Subsequent to initial measurement of suction the samples were tested in the triaxial apparatus and
E (MN/m2) N/A 16 28
compaction Mass E (MN/m’) 175 N/A 350 26 525 26
I Change
in moisture content M/c E (MNlm’) 24.6 15 26.7 13 27.9 11
I
Figure 6b shows the stress strain behaviour of samples compacted using various hammer weights. It appears that samples compacted with the 350 g and 525 g hammers behaved elastically at least in the early part of shearing and the sample compacted with the 175 g hammer behaved elasto-plastically from the start of the shearing. In the case of 400 kPa compaction pressure in static compaction and 175 g hammer mass in the dynamic, samples were compacted to the same initial void rstio. However it appears that the stress-strain behaviour of the materials are considerably different. A possible explanation for this difference may be the shear deformation produced by the dynamic compaction process. The Young Modulus of the samples compacted with 350 g and 525 g hammers are si nii 1ar . Figure 6c illustrates the stress-strain behaviour of samples statically compacted to 400 kPa of compression at different values of moisture content, Samples compacted at low moisture content (24%772
in Figure 7. It is generally accepted that the yield locus for naturally occurring saturated soil is approximately aligned along the KO.1A similar effect is likely in the case of unsaturated samples which are compressed in one direction. Figure 7a illustrates the expansion of the yield locus as the compactive effort is increased. The shape of the yield locus for unsaturated soil in the p:s plane has been a subject of great interest and recent research indicates that increase in suction or increase in mean net stress or a combination of both can lead to an expansion of the yield locus as shown in Figure 7b, Sivakumar and Ng (1998). Therefore it is probable (combining both diagrams in a three dimensional space) that reduction or increase in suction can lead to reduction or increase in yield stress even at a given compactive effort.
Figure 7 Yield locus in p:q plane ans p:s plane
Figure 6 Stress-strain curve 28%) initially exhibited elastic behaviour and samples compacted at high moisture content exhibited plastic behaviour throughout. The Young Modulus shear modulus was affected by the compaction moisture content and as expected it was found that Young modulus reduced with increasing compaction moisture content.
A close examination of the stress strain curves shown in Figure 6 indicates that the yielding characteristics of unsaturated soil are also influenced by the variables considered in the programme. Increase in compaction pressure or decrease in compaction moisture content increases the magnitude of the deviator stress at which the sample yields. This can be explained with the sketch shown
Figure 8 illustrates the variation of apparent cohesion p(s) (the intercept of the critical state line on the q axis) with compaction moisture content for the samples tested in the third category where the initial moisture content was considered as the variable and the compaction pressure and type of compaction were unchanged. Looking from Figure 6c it is reasonable to assume that all samples tested in this category have reached the critical state at failure. The magnitude of p(s) was calculated using Equation 7 and assuming M =0.93 (M is the slope of the critical state line). The magnitude of the deviator stress at critical state was estimated from the stress strain curve. Since the confining pressure applied in each test was 200 kPa the magnitude of mean net stress at the critical state is given by q/3+200. Figure 8 shows the magnitude of p(s) plotted against moisture content at critical state (the compaction moisture content). It appears from Figure 8 that the magnitude of p(s) reduces linearly with increasing initial moisture content (at least within the range of moisture content considered) to zero at a moisture content of 30.4% and continue to fall as the moisture content is further increased. Surprisingly the moisture content at which p(s) dropped to zero was approximately the same as the
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value at which the sample exhibited maximum dry density. Figure 9 shows the compaction characteristic of kaolin (standard proctor compaction and static compaction at 400 kPa). At a given moisture content the dry density obtained from static compaction at 400 kPa was considerably less than the standard proctor compaction and the optimum moisture content obtained from static compaction was approximately 30.5%. This moisture content corresponds with the moisture content at which the magnitude of p(s) drops to zero. This indicates that compacting fills to optimum moisture content may mean no loss or gain in shear strength if the fill eventually becomes saturated. If the fill is compacted dry of optimum the shear strength may well be high at the time of placement but will drop when the suction drops to zero during saturation. On the other hand if the fill is placed wet of optimum the shear strength may well be low at the time of placement but when the pore water pressure is dissipated the fill may well gain strength.
Figure 9 Variation of p(s) with moisture content at failure
REFERENCES Alonso, E.E, Gens, A and Josa, A. (1990). Constitutive Model for Partially Saturated Soils, Geotechnique, Vol. 40, No.3,405-430. Fredlund, D.G., Morgenstern, N.R. and Widger, R.A. (1978). The shear strength of unsaturated soils. Canadian Geotech. Journal, 15, No. 3 , 3 13-321. Matyas, E.L. and Radhakrishna, H.S. (1 968). Volume Change Characteristics of Partially Saturated Soils, Geotechnique, Vol. 18, No. 3,432-448. Sivakumar, V.( 1993). Critical State Framework for Unsaturated Soil. PhD thesis, University of Sheffield. Sivakumar, V. Ng, P. (1998). Yielding of unsaturated soils. 2nd International Conference on unsaturated soils, China, Vol. 1, I3 I- 136. Wheeler, S.J and Sivakumar, V. (1995). An ElastoPlastic Critical State Framework for Unsaturated Soil. Geotechnique, Vol. 45, No. 1,35-53. Wheeler, S.J and Sivakumar, V. (in press). Influence of compaction procedure on the mechanical behaviour of an unsaturated compacted clay, Part 2, Shearing and constitutive modelling, submitted to Geotechnique.
Figure 8 Compaction characteristics of kaolin
CONCLUSION A wide range of samples was prepared using different moisture content, type of compaction and compactive effort. It is apparent that the initial moisture content has significant influence on the initial suction and the type of compaction or the compactive effort has marginal influence on the initial suction. It is also evident that the stiffness of the material was influenced by the amount of compaction and compaction moisture content. The apparent p(s) cohesion was influenced by the initial moisture content and it dropped to zero at the value of optimum moisture content.
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Slope Stability Engineering, Yagi, Yamagami& Jiang 0 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Landslide at Malakasa, Greece: Investigation, analysis, remedial works R. J.Chandler Imperial College of Science, Technology and Medicine, London, UK
S.Schina 0.TM. SA Consulting Engineering Company,Athens, Greece
ABSTRACT: An extensive landslide occured in the area of Malakasa, north of Athens, severing the traffic in the Athens - Thessaloniki highway and railway. An intensive programme of site investigation gave information about the nature of the landslide, the geological formations and the ground water level, helping the conception of the phenomenon. The existence of two samples from the slip surface of the landslide gave the opportunity to carry out thin-sample shear tests, while the subsequent slope stability analyses confirmed the results of the laboratory tests. Finally, the design of the remedial works, which are under construction now, is based on the results of the investigation and the analysis of the landslide, leading to the improvement of the stability of the landslide area.
1. INTRODUCTION At 18th February 1995, an extensive landslide, probably one of the most severe in Greece for many years, occured in the area of Malakasa, north of Athens, causing disruption over a wide area, due to the interruption of both rail and road traffic from Central and South Greece to the North. The main damages were the distortion of the welded rail track, the deformation of the highway surface, which reached a heave of 3.0m height and 70m length and the destruction of the earlier remedial works at the toe of the slope, including concrete reinforced piles and a toe wall (Fig.1). Fortunately, no victims were during this disaster, especially because some signs of local movements in the railway became the reason of taking some first remedial measures. The landslide occupied an area of about 30Ox350m from its toe to the fbrthest back-scarp and its maximum thickness was 30m. A combination of factors caused the outbreak of the landslide. The most determinative were an excavation for the widening of the highway at the toe of the slope, the high piezometric level within the landslide mass and the existence of a previous slip surface in the same area.
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Fig. 1 : Plan of the Malakasa landslide
Fig.2 : Site investigation in the Malakasa landslide - Longitudinal section
2. SITE INVESTIGATION In the landslide area, not only inside the landslide limits, but also in the wider area, an intensive programme of site investigation started immediately after the landslide. In the aggregate, the initial and supplementary site investigation included: sampling and non-sampling boreloles, trial pits and shafts, piezometers, inclinometers, observation of fixed points, cracks and pumping tests. The most important conclusions resulted from the site investigation are: e The geology of the landslide mass concerns a mixture of different geological formations, materials of different geological age, structure, origin and mineralogy, which coexist without clear stratigraphy. An indication of the presence in the past of at least one landslide in the same area, which destroyed the original geological stratigraphy. e The main geological formations are: clayey schist, limestone, sandstone and clayey material. The slip surface was mainly found within mass of clayey schist and less in few thin layers of clay within limestone mass (Fig.2). e In the landslide mass, the ground water level is generally high, following the inclination of the soil surface (Fig.3). While in the upper part of the landslide the water horizon seems to be continuous, in the central and eastern part of the landslide, the piezometer data showed an incontinuous horizon. This can be justified by the existence of an impervious layer in a certain depth and a pervious one immediately below this, which is behaving as drainage layer. The main movement of the disturbed area had S-N direction. It started from the south-eastern part 776
above the railway and by loading the downslope (northern) parts it broke out at the toe of the slope, in the highway. On the other hand, the south-western part seems to be swept by the main landslide, showing a movement to north-eastern. The displacement of the main part of the landslide resulted to be approximately 7m from South to North, giving a subsidence of more than 5m in the tension scarp and a heave of 3m of the toe wall. The main part of the disturbed area shows a lack of cracks, fact that indicates the plane movement of the landslide on a pre-existing slip surface. e Finally, the plane character of the main part of the landslide and its three dimensional substance was obvious.
Fig.3 : Plot of pore pressures U (@a) against depth
3. ESTIMATION OF THE RESIDUAL STRENGTH The residual strength (appropriate for the analysis of old landslides) is the minimum constant value attained (at slow rates of shearing) at large displacements. The displacements necessary to cause drop in strength to the residual value are usually far greater than those corresponding to the development of peak strength (of the order of lmm in shear box) and the fblly softened (critical state) strength in overconsolidated clays, where it undrgoes no hrther volume changes in the failure zone, reaching a critical void ratio. Numerous attempts have been made in the past to determine the residual strength, by using different methods. The most important of them are the slip surface tests, the multiple reversal tests, the cut plane tests, the ring shear tests, the thin sample tests and the new cut-thin-sample technique. Chandler & Hardie (1989) carried out numerous tests on thin samples, with a great range of sample thickness and under several vertical loads, giving well agreed results with these of back analyses of landslides in London clay. This technique, with the modification of a new shear box and the addition of the sample cutting before the shearing (new-cut-thin sample technique), was used for the estimation of the residual strength of the Malakasa landslide. The test procedure and the main features of the shear box apparatus described by Schina (1995) are applied in the laboratory of Soil Mechanics of Imperial College of Science, Technology and Medicine. The corresponding shear box was of Casagrande type, as it was modified by Bishop in 1946 (Fig.4).
northern part of the landslide, near to the toe, where the dark grey schist emerged, exposing a heavily slikensided surface. The second one was from the south-westem part of the landslide and it concerns a brown clay. The samples were tested in the direct shear box by using the new-cut-thin samples technique. The samples were consolidated to a normal stress of 0'" = 323.08kPa, giving an overconsolidation ratio OCR=1.50. They stayed under this load for 24 hours, followed by cutting and their relaxation. The rate of shearing, common for both samples, was 0.0262dmin. 3. I Dark grey schist
The wider part of the slip surface is found within the dark grey clayey schist, as it is resulted from the evaluation of the site investigation. For this reason, the results are supported to be closer to the reality and more representative relatively to the other sample. The material was weathered with some coarser particles of less weathered schist, which were removed, as far as it was possible. The initial thickness of the sample was h;=5.14mm and its initial moisture content was w = 25.8%. The final thickness was hf = 3.95mm, while the final moisture content was w = 36.05%. Finally, the initial weight was Wi= 25.5gr. After the consolidation and the unloading, the sample was first sheared twice at a normal stress of ofn= 323.08kPa and then under the following sequence of normal stresses with intervening reversals: 258.83kPa, 162.50kPa, 130.39kPa (twice), 78.47kPa, 53.27kPa. The test results are given to the following shear stress - horizontal displacement plot (Fig.5). 3.2 Brown clay
Fig.4 : Cross-section of the direct shear box
As it was mentioned, the wider part of the slip surface was found within the dark grey schist and less in brown clay. Two samples of different material were scraped from the slip surface. The first one was from the 777
The second sample was of brown clay, characteristic formation of the western part of the landslide, which was moved quite independently to the main landslide. During the preparation of the sample, there were found many coarse particles of big size, which could be estimated as quartz particles, as well as many rotted roots. It was extremely difficult to remove all of them, as their percentage in the sample mass was extremely high. The initial thickness of the sample was hi=4.87mm and its initial moisture content was w = 38.55%. The final thickness was hf = 4.13mm, while the final moisture content was w = 43.34%. Finally, the initial
weight was Wi= 24.5gr. After the consolidation and the unloading, the sample was sheared three times at a normal stress of o’~323.08kPa and then once at 258.83kPa. Unfortunately, as it is shown in the results, it was impossible to attain a residual strength, which would remain approximately constant until the end of the shearing. Despite the repeated shearings, the stressdisplacement plot showed the same behaviour: at
very small displacements a high peak was reached, followed immediately aRer by a very sharp drop to the level of the residual strength. After 2mm of horizontal displacement, the shear stress started to increase, following an inclination almost similar to all shearings, towards the initial peak value. The test results are given to the following shear stress horizontal displacement plot (Fig.6 ) .
Fig.6 : Brown clay - Shear stress-horizontal diasplacement plot
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A rough explanation of this behaviour is referring to the existence of the coarse material within the clay mass. It is probable, these particles to give a high peak at the beginning of the shearing, to collapse immediately after and then to try to “build” their structure again. Whatever the reason, this test was forced to stop, but as a first estimation, the lowest parts of these plots, showing perhaps the residual strength, gave almost the same residual fnction angle with the relevant of the first test. As it was mentioned, only the test results of dark grey schist can be evaluated for the analysis of the
Malakasa landslide. Besides they are the most representative, since the major part of the slip surface was mostly found within this material. The lowest residual friction angle measured was cprr = 11” for o’,= 323.08kPa, showing a residual strength of 62.50kPa, while for normal stress o’,,= 1304kPa the residual fnction angle was 14”. In the following Fig.7, the shear stress as well as the friction angle are plotted against the normal load, giving the range of fnctional resistance for the dark grey schist.
Fig.7 : Dark grey schist - Residual strength envelopes
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4. SLOPE STABILITY ANALYSES In order to carry out stability analyses for the landslide of Malakasa, three different conditions were examined. The first one was the slope conditions before the excavation for the widening of the highway using data from an earlier map of the area. The second one was the slope profile immediately after the last of all the excavations. Finally, the last one was the slope profile after the landslide, gathering in long sections all the available information derived from the site investigation. The analyses performed on longitudinal sections of south-north direction through the slope, in which the best estimation of the slip surface, the ground water level and the tension cracks were represented. For the slope stability analyses, the Janbu’s Simplified Method was used, since the slip surface of the landslide had a non-circular character. For these back analyses the results of the aforementioned laboratory tests were used in order to check the slope stability under the three different conditions. The main conclusions of the slope stability analyses were that before the excavation the slope was stable, while after the excavation at the toe of the slope the factor of safety showed a trend to decrease, reaching a factor of safety of 0.90. Finally, after the landslide, the slope showed limit equilibrium, with a factor of safety equal to 1.OO.
5. REMEDIAL WORKS
As it was derived from the slope stability analyses, two seems to be the most significant factors which affect the stability of this slope: the high ground water level within the landslide mass and the removal of material from the toe of the slope. The design of the remedial works is mainly based on these two factors. The first immediate measure for the improvement of the stability of the landslide area was a local excavation in the upper part of the landslide, between the rear scarp and the railway line. However, the most significant remedial work, which is under construction now, is a net of drainage tunnels, consisted of one main drainage tunnel of N-S direction and six cross tunnels, which are 30-35m apart. The alignment of the tunnels (hypsometrically) will be found very close and always beneath the slip surface. The total length of the drainage tunnels will be 1,375m. The gradient along the tunnels will be smaller or equal to 10%.
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The tunnel network will be connected with the drainage network of the landslide consisted of 300mm diameter drains constructed from the ground surface following a net of lOxlOm and 6” diameter drains, driven from the tunnels to the slip surface. There is provision also for the construction of perimetric drainage holes, around the tunnel lining. Finally, another possible intervention for the increase of the slope stability will be the toeweighting with local rising of the highway and its placing on a well-compacted embankment, providing for an appropriate drainage beneath the embankment.
6 . CONCLUSIONS
The extensive damages and the intricate nature of the Malakasa landslide led to an intensive programme of site investigation, laboratory tests and analyses. As it is concluded, a combination of factors caused the outbreak of the landslide, such as the excavation for the widening of the highway at the toe of the slope, the high piezometric level within the landslide mass and the existence of a previous slip surface in the same area. The design of the remedial measures attempts to reduse the adverse effect of these factors. Finally, it should be mentioned that there was satisfling accordance between the results of laboratory tests and those of stability analysis. For this reason the results of the laboratory tests are evaluated as satisfling and the new cut-thin sample technique as successhl for the fast determination of the residual strength.
REFERENCES Chandler, R.J., Hardie, T.N. (1989),“Thin sample technique of residual strength measurement”. Geotechnique 39, No3, 527-53 1 Chandler, R.J. (199 l), “Slope stabilty engineering: developments and applications”. Institution of Civil Engineers. Thomas Telford, London Schina, S.N. (1995),“Investigation of the landslide at Malakasa, Greece”. MSc Dissertation,University of London Skempas, M.N. (1994),“Dam abutment stability with particular reference to Thisavros Dam”. PhD Thesis, University of London Skempton, A.W. (1985), “Residual strength of clays in landslides, folded strata and the laboratory. Geotechnique 35, Nol, 3-18
Slope Stability Engineering, Yagi, Yamagami & Jiang @) 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Method for determining design strength parameters for slope stability analysis T. Mitachi Graduate School of Engineering, Hokkaido University, Sapporo, Jupun
M.Okawara Faculty ojEngineering, h a t e University, Morioku, Japan
T. Kawaguchi Department of Civil Engineering, Hukodotc Nutionnl College of Technology, Japan
ABSTRACT. A new method of determining design strength parameters for landslide slope stability analysis is proposed. The authors assume that the strength decrease due to the increase of pore water pressure along the potential slip surface can be represented by a function of overconsolidationratio(0CR) defined by the ratio of effective stresses before and after the increase of pore water pressure. Based on the assumptions mentioned above, the authors derive equations expressing the change of effective stress strength parameters(c, $), which varies in between peak and residual state, as a continuous function of OCR. In addition, a practical method of determining strength parameters for stability calculation of landslide slope is proposed. In this method, the strength parameters are given by combining the conventional revem calculation method, which has been frequently used in engineering practice, with the laboratory shear test result5 obtained by cyclic direct box shear apparatus. 1INTRODUCTION
The suitability of strength parameters for stability calculation is the most essential factor in evaluating landslide slope stability. In engineeringpractice in Japan, strength parameters have been almost always determined by an empirical method named as "reverse calculation method". In this method, strength parameters are back calculated as shown in Figure 1 based on the equilibrium condition of sliding earth mass by Fellenius' stability calculation method which is represented by the equation indicating straight line PQ in the figure, where CN, CT and CL are normal and tangential forces acting on the sliding mass and the length of sliding surface, respectively. In calculating strength parmeters by this method, the apparent cohesion q)is assumed in the Grst place as q , k d (kN/m3 (d: thickness of sliding mass (m)) and then the angle of shear resistance $(, is obtained by corresponding point on the straight line PQ in Figure 1 assuming the current safety factor F,, = 1.0. Although theoretical defects have frequently been pointed out on this conventional method, it is still widely used in practice. Studies by Saito (1974) and Yamagami et al. (1984,1992) have been aiming to overcome the defects of the reverse calculationmethod. Gibo et al. (1984,1987) proposed a method to obtain average strength mobilized along the slip surface by taking into account of the type of landslide and by introducing the residual factor R proposed by Skempton (1964) to the peak, fully softened and residual state strength parameters obtained by laboratory shear test. Ogawa (1985) proposed a method to determine design strength parameters for secondary
Figure 1. Determination of strength parameters by conventional "reversecalculation method".
slide by assuming that the clay on the slip surface which reaches once to the residual state shifts to overconsolidated state due to the increase of pore pressure acting along the slip surface. In this paper, the authors derive new equations expressing the change of strength parameters by assuming that the combination of effective stress strength parameters (c, 4) to be used for landslide stability calculation changes in between peak to residual state as a continuous function of overconsolidation ratio, and also propose a simple and practical method of determining design strength parameters by combining the conventional reverse 781
Figure 2.Relationships among the strength parameters for peak, fully softened and residual state of normally and overconsolidatedclay.
calculation method with the strength parameters obtained by laboratory shear test. In contrist to the conventional reverse calculation method, the s i m c a n t feature of the proposed method is that it takcs the material strength characteristics of particular slope into account in the stability calculation. 2 NEW METHOD FOR DE'ERMI"G STRENGTHPARAMETERS
DESIGN
2.1 Strength change due to state change of particular slope The shear strength of soil, which is the controllingfactor of the stability of landslide slope, depends on past stress history and strain level induced on the soil clcment as well as the geological factors. Considering the case of secondary slide, as the increase of pore water pressure results in decrease of effective stress and causes the reduction of shear resistance, the strength parameteIs(c, #) for stability calculation may change in between peak to residual state as a continuous function of OCR. The process of changing shear strength of the soil element along the sliding surface may be modeled by a process of effective stress decrease during direct shear test under constant total normal stress condition. In this paper, it is assumed that the shear strength of clay soils can be represented as follows.
c, = c ,
fC,
=p.o,'
where, tan#,is assumed to be a material constant which is a measure of strength change due only to the change of effective stress 0' under constant void ratio and is independent of stress history of clay soil. Parameter c, defines a strength component which changes with void ratio and is proportional to the equivalent consolidation pressure 0,' defined by Hvorslev (1960) and c, is produced by creep effect and upgradation of clay structure due to ageing and is also assumed to be proportional to the equivalent consolidationpressure since it degrades due to application of the stress exceeding consolidation yield stress. For the sake of simplicity, it is assumed that the sum of two strength components c, and c,, which is defined as c, in this paper, is a linear function of 0,' as shown in Eq.(2), where p is defined as coefficient of cohesion. Otherwise stated, all of the stresses appearing throughoutthe rest of this paper are effective stresses.
Peak strengthparainetersfor overconsolidated state Figures 2 (a)-(c) illustrate the case in which the effective normal stress decreases from the state o,,(point A) to the state oo(point B) and the points C and D denote the drained shear strengths corresponding to each state of o, where (Pp,and #sn are effective angles of shear resistance corresponding to peak and fully softened state of undishirbed and remolded normally consolidated clay,
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Taking into account the change of void ratio due to stress change along the residual state line and the line corresponding to peak strength state, the following equation is obtained.
A ln(ol,/ cr, ) = Y 1n(o, / o,, )
(5)
Combination of Q s . (4) and (5)gives
Denoting N and r as the void ratio for the state o'=l on the normal consolidation line and residual state line, the following equation is derived.
cr,,, /on= exp{(N - r)/A} Figure 3. Effective stress vs. void ratio relationship during constant normal stress direct shear test.
Substituting the relationship obtained by the above equations into Q.(3) and changing the subscript of $ fiom d to p, we obtain, tan 4,
respectively. Straight lines drawn through the points C, and D, and parallel to the h e rc=oltan$, have the cohesion intercepts c,, and c, in Figure 2(d), respectively in accordance with Eqs.(l) and (2). For the case in fully softened state is shown as illustrated in Figure 2(Q. The strength differences between the points C, and C, or the points D, and D,, and those between the points Cpand C, or the points D, and D, in Figure 2(b), may be considered as the strength components c, and c, in Q.(2), respectively. Assuming the effective stress failure envelopes connectingthe points C and D to be straight, tan&, or tan$\, which are the slope of line C P pfor peak strength state or C,D, for fully softened state as shown in Figures 2(d) and (9, can be represented as follows by using the symbols in Figure 2.
(7
= m{ - (OCRx - l)/(OCR - l)}+
tan $r
(8)
where,
Cohesion intercept c,, which is a representative of c, and c, in Figures 2(d) and (0, is denoted as follows.
Combining above equation with Qs.(3) and (9, and changing the subscript of c from d to p, we obtain
c, /oo= m OCRf(0CR' - l)/(OCR - 1)}
(12)
Peak strengthparametersfor normal consolidation state Application of the followingrelationship to the Eqs(8) and (12) where, & , is the representative of $p and 4, , overconsolidation ratio is dehed as OCR=o,,/o[>and are denoted as ocn and o, , equivalent stresses for cr, and o,, respectively. Figure 3 illustrates void ratio versus effective stress relationships at the peak strength state for the u s e of consolidated drained test under constant normal stress o, and the m e of drained test for the specimen experienced consolidation by cr, and subsequent rebound to 0,. Assuming the slope of residual state line (full line) is parallel to that of normal consolidation line (dotted line), the following relationship is derived fiom Figure 3.
and changing the subscript of c and $ from d to s gives following Eqs. (13) and (14) tan$,
= m(1
-A)+ tan$,
(13)
-
c,/o,
=
mA
(14)
Strengthparametersfor resdual state As defined in Eqs.(l) and (2), the angle of shear resistance $, is independent of stress history and change of void ratio, the strength parametexs for residual state are
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Figure 6. Determination of design strength parameters for landslide slope stability calculation.
Figure 4. Change of strength parameters as a function of OCR.
and (12), Eqs.(l3) and (14), and Eqs.(l5) and (16), respectively.
Change of strengthparameters as afinctwn of OCR Figure 4 shows the examples of illustrating the changing trend of the combination of strength parameters by using Eqs.(8) and (12) as a function of OCR for the cases assuming I\ in 4. (10) as 0.1 and 0.2. Based on the examples shown in Figure 4, (cJoJm) versus (tan@Jm) relationship is generalized as shown in Figure 5. 2.2 Method for determining design strength parameters based on the laboratorysheur test results Figure 5 illustrates the change of strength parameters of clay on the landslide slip surface as a function of OCR due to effective stress change. The part of dotted line in the figure illustrates the strength decrease from the fully softened state to residual state as shown in the insertion in Fig5 The strength decrease between the two states is not accompanied by void ratio change and is interpreted due to reorientationof clay particles (Skempton, 1985). If the peak strength parameters are obtained from the monotonic loading direct shear test by newly designed high precision automatic cyclic direct shear apparatus (for example, Okawara et al. 1999) with undisturbed clay specimen sampled from the slip surface of actual landslide site, and strength parameters corresponding to fully softened and residual states are obtained from the cyclic shear test by using the same apparatus with the specimen fully remolded and preconsolidatedfrom the state of slurry, the three sets of the strength parameters should be plotted on the theoretical curved line in Figure 5 as the points A, B and C . Therefore, if we connect the three points A, B and C by folded line as an approximation and draw the line PQ which is the same one as shown in Figure 1 indicating analytically possible combination of (c, $) resulting from Fellenius' stability calculation by assuming current safety
Figure 5. Schematic diagram of possible combination of changing strength parameters as a function of OCR.
As shown above, strength parameters for overconsolidated peak state, for normally consolidated peak state which corresponds to fully softened state, and for residual state are given by the combination of Eqs.(8)
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Hvorslev, M. J. 1960. Physical components of the shear strength of saturatedclays, Proc. ASCE Research Con. on Shear Strength of CohesiveSoils:169-273. Mitachi, T., A San0 and M. Okawara 1996. The relationships between strength parameters obtained from laboratoryshear tests and those for use of stability calculation,Proc. of35th Annual Convention of Japan Landslde Society:345-348 (in Japanese). Okawara, M., T, Mitachi and M. Tanada 1999. Development of an automated cyclic direct shear test apparatus for determining strength parameters for landslide slope stability analysis, Proc. of International Syinposiumon Slope Stability Engineering. Ogawa, S. 1985. On the determination of strength parameters for landslide slope stability calculation,The Foundation Engineering & Equipment, 13(9):18-23 (in Japanese). Saito, M. 1974. A method of determining c and @ parameters for conventional stability calculation, Proc. 9th Japan National Con. on Soil Mechanics and Foundation Engineering:601-6@4(in Japanese). Skempton, A.W. 1964. Long-term stability of slopes, Geotechnique, 14(2):75-102. Skmpton, A .W. 1985. Residual strength of clays in landslides, folded strata and the laboratory, Geotechniyue,35(1):3-18. Yamagami, T. and Y Ueta 1984. A new method for inverse calculation of c and @ona slip surface @art I) - Fundamental concept - , Journal of Japan LandslideSociety, 21(2): 16-21(in Japanese). Yamagami, T. and Y Ueta 1992. Back analysisof strength parameters for landslide control works, Proc. ofthe 6th Int. Syinp. on Landslide:619-624.
factor F, of 1.0 for a particular slope, then we obtain the design strength parameters (cd, &) for secondary slide of this slope by the intersectionpoint E of line ABC and PQ as shown in Figure 6. Even if the measurements of strength parameters have some errors or they are not represents exactly the strengths of corresponding slip surface, the variation of the strength parameters may be plotted around the shaded area in the figure. Therefore, the design strength parameters determined by the point of intersectionE in Figure 6 must be more reliable than those obtained by conventional method based on the assumption of c,=d (kN/m2) (d: thickness of sliding mass (m)) as shown in Figure 1. Case studies for two sites of landslide demonstrating the suitability of the method proposed in this paper are reported in the companion paper (Okawara et al. 1999).
3 CONCLUDING REMARKS As a method of determining design strength parameters for the use of landslide slope stability calculation, a new practical method by combining the strength parameters obtained from laboratory shear test results on a clay specimen sampled from the slip surface of landslide site with the conventional "reverse calculation method" was proposed. The features of this method is as follows.
1. By combining strength parameters corresponding to peak state (c,,, &,), fully softened state (cs, @Jand residual state (c,, &) with the c-tan# relationship which has been used in conventional reverse calculation method, design strength parameters can be determined without assuming the magnitude of c value which is essential in reverse calculationmethod.
2. A simple and practical method for determining design strength parameters proposed in this paper as illustrated in Figure 6 has a theoretical background shown in Figures 4 and 5. 3. Application of the method for determining design strength parameters proposed in this paper makes possible to restrict the range of changing (c, 4) of analytically possible combination along PQ tine in Figures 1 and 6 within the range of possible combination reflecting material strength characteristics. REFERENCES Gibo, S., A. Takei and S. Kohagura 1984. Methods for estimating the parameters of average shear strength along the slip surface, Journal of Japan Landslide Society, 20(4):1-6 (in Japanese). Gibo, S. 1987. Shear strength parameters required for evaluation of stability of slopes, Tsuclzi-to-Kiso, 35(11):27-32 (in Japanese).
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Slope Stability Engineering, Yagi, Yamagami & Jiang (C 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Evaluation of the shear strength for stability analysis of a heavily weathered tertiary rock K.Tsuji, K.Suzuh & H. Hanzawa Toa Corporation Technical Research Institute, Yokohama, Japan
ABSTRACT: A banking embankment with maximum height of 80m was constructed above a heavily weathered tuff stratum overlain by a colluvial deposit. The banking fill was also borrowed from this weathered tuff. In this paper, the shear strength to be used for stability analysis is presented both for the heavily weathered tuff and the banking fill materials, together with factors of safety calculated.
In this paper, the shear strength to be used for stability analysis of this kind of material used for embankment construction and the results of stability analyses are presented. Estimation of shear strength for actual construction has been reported by Hanzawa (1983 and 1993) .
1 INTRODUCTION A large scaled earthwork was carried out at a hill site in Japan. In this project, a banking embankment with maximum height of 80m was constructed on a heavily weathered tuff stratum and the colluvial deposit with N-blows of 10 to 40, while filling material was also borrowed from the weathered tuff. It is very important, therefore, to evaluate appropriately the shear strength of the heavily weathered tuff for the foundation ground and for the heavily weathered tuff to be used as the fill material. For this purpose, a series of direct shear tests was carried out both on the block sampled undisturbed soils and the compacted heavily weathered tuff in partially saturated and submerged conditions.
2 SOIL INVESTIGATION Schmatic diagram of soil profile of the foundation ground and configuration of embankment oare shown in Figurel. Average ground slope is 15 and that of embankment is 29' . The foundation ground is consist of heavily weathered tuff stratum and the colluvial deposit. The colluvial deposit
Figurel. Soil profile Figure2. Soil boring logs and N-values
787
Tablel. Physical properties of soil in the site
I
Soil Wet density ( p I) Natural water content ( W N ) Degree of saturation (Sr) Sand Soil type Silt Clav
Index properties
Liquid limit (WL) Plastic limit (wp)
I Colluvial
I I
deDosit 18.7-19.2 kN/m3 20-22 % 79-81 76 14 % 45 % 41 % 35 % 21 %
I Heavilv weathered tuff 1
I
I
15.8-16.2 kN/m3 30-33 % 69-73 % 2 % 59 % 39 % 49 % 28 %
contains clay, sand and gravel. Soil boring log and distribution of SPT-N-blows obtained from soil investigation at boreholes A and B ( Figurel) are presented in Figure2. From the N-blow distribution profile, the N-blows are scattered between 10 to 40. However higher value of N-blow is caused by large gravel, so it is appropreate to use the N-blow of 10 in the stability analysis. The physical properties of the heavily weathered tuff stratum and the colluvial deposit are shown in Tablel.
It is difficult to predict accurately the excess pore water pressure at failure, therefore, Z r ( U ) is expressed by a function of the 0 'VO, the effective overburden pressure as given in Eq. ( 2) as also indicated in Figure3.
where (I, ap=apparent internal friction angle in undrained direct shear test, and d, d=internal friction angle in drained direct shear test.
3 SHEAR STRENGTH OF COHESIVE SOIL
4 DIRECT SHEAR TEST CARRIED OUT Typical effective stress path of normally consolidated or slightly overconsolidated soft clay and highly overconsolidated stiff clay from direct shear tests are presented in Figure3. In order to simplify the diagram, it is assumed that the cohesion is negligible and the friction angle is the same between the soft and the stiff clay. It is well known that the undrained shear strength ( Z f (U) ) is smaller than the drained shear strength ( Z f ( d ) ) for soft clay, however, this relationship is reversed for stiff clay as given in Eqs. ( l a ) and ( l b ) .
Figure3. Typical effective stress path
7aa
Block sampling were conducted at the most weathered part of the heavily weathered tuff and the part of colluvial deposit without large gravel. After leveling the ground surface of sampling part, block sampling was conducted to acquire the undisturbed samples by pushing the Sampler ( d, =10cm X h=12.5cm, split type) slowly into the ground. Four types of direct shear tests ( DST) were carried out for the undisturbed or compacted samples, as explained here: 1. DST-1: In this test, cylindrical specimen with diameter of 60mm and height of 20mm prepared from undisturbed sample is compressed at the prearranged consolidation load ( CJ 'VC) , until primary consolidation has been achieved, and then sheared under the constant volume condition at a displacement rate of 0.25mm/min. This test was carried out in order to evaluate the shear strength of the foundation ground at the dry season. 2. DST-2: First, the specimen prepared the same as DST-1 is compressed at 1/3 of the prearranged consolidation load and submerged during an hour. Next, the specimen is compressed at the prearranged consolidation load, until primary consolidation has been achieved and then sheared under the constant volume condition at a displacement rate of 0.25mm/min. This test was carried out in order to evaluate the shear strength of
Table2. Direct shear test conditions carried out Test name Soil name Specimen condition Soil moisture condition Degree of saturation Sr (%) Consolidation load (kPa)
DST- 1 H.W.T. C.D. undisturbed partially saturated 79-81 69-73 100,200,300 50,100,200, , 300,400
I
Shear condition Displacement rate Spacing between upper and lower shear box
DST-2 C.D. H.W.T. undisturbed submerged 97-99 90-95 100,200,300 50,100,200, 300,400 constant volume shear 0.25mm/mi n.
I
1 I
DST-4 DST-3 H.W.T. H.W.T. compacted partially saturated submerged 71-80 91-98 50,100,200, 50,100,200, 300,400 300,400
0.50mm
the foundation ground at the rainy season. 3. DST-3: In this test, prepared specimen of compacted ( 3 layer system and each layer is compacted 55 times with rammer weighed 25N) heavily weathered tuff was sheared the same condition as DST-1. This test was carried out in order to evaluate the shear strength of embankment at the dry season. 4. DST-4: In this test, prepared specimen compacted the same manner as DST-3 heavily weathered tuff was sheared as same condition as DST-2. This test was carried out in order to evaluate the shear strength of embankment at the rainy season. The test conditions of four types of direct shear test carried out are presented in Table2. Figure4. Normalized shear stress vs. displacement in DST for undisturbed samples
5 RESULTS OF DIRECT SHEAR TESTS Normalized shear stress, Z / ( 7 ' V C versus displacement,d, from DST-1 and DST-2 are shown in Figure4 and from DST-3 and DST-4 are shown in FigureS. Some acquired knowledge from these diagrams are as follows: 1. It is cleare that T / 0 ' V C for undisturbed samples from DST-1 and DST-3 are greater than Z / (7 ' V C for submerged samples from DST-2 and DST-4, and this tendency is more pronounced in the heavily weathered tuff stratum. 2. Compacted heavily weathered tuff with consolidation load of (7 ' v c = ~ O and lOOkPa in DST-3 present large positive dilatancy, and its relation between T / (7 ' V C and displacement are different from any other specimen compacted as well as undisturbed specimen. 3. Undisturbed sample, both heavily weathered tuff and the colluvial deposit present negative dilatancy when consolidation load, (7 ' V C is equal or greater than 100kPa. Taking the embankment thickness of 10 to 4om into consideration, the shear strength of the foundation ground should be evaluated based on z f (U).
789
FigureS. Normalized shear stress vs. displacement in DST for compacted samples
6 EVALUATION STRENGTH
OF
DESIGN
SHEAR
Typical effective stress path obtained from direct shear tests are shown in Figure6. In this diagram, Z f ( U ) is greater than Z f ( d ) in zone 1 , and
Soil
Zone
Undisturbed colluvial deposit Undisturbed heavily weathered tuff
a
Compacted heavily weathered tuff
a
U 1
Partially saturated condition c'orcap @ ' o r @ a p U'b 33 kPa 29.5' 45 kPa 21.555 kPa 85 kPa
38.0' 23.5'
8o kPa
Submerged condition c'orcap @'or@ap U'b 25 kPa 24.025 kPa 20.025 kPa 45 kPa
,
37.0' 22.5'
,
60 kPa 1
Z f ( d ) is greater than Z f ( U ) in zone fl . When it is assumed that c ' = a and (i, '= 6 d , design mobilized shear strength, Z r ( m o b ) in both zones are expressed as follows:
z f ( m o b ) =C'+ CT
'0
=Cap+ B
- tan (i, ' '0
-tan (1
ap
(zone I 1 (zone fl )
(3a) (3b)
where, 0 'o=effective stress before shear. The effective stress path of undisturbed heavily weathered tuff and the colluvial deposit from DST-1 and DST-2 and compacted heavily weathered tuff from DST-3 and DST-4 are shown in Figure7 and Figure8 with shear strength obtained in the manner shown in Figure6. Evaluated design shear strength parameters, Cap, (i, ap, c' and (i, ' are presented in Table3.
Figure7. Effective stress path from DST-1 and DST-2
Figure6. Typical effective stress path obtained from DST
7 STABILITY ANALYSIS
Two stability analysis for dry and rainy seasons are conducted and results are shown in Figure9. In this analysis, ground water surface is the surface of the colluvial deposit for dry season, on the other hand that is at the middle height of embankment thickness for rainy season. Safety factors obtained from stability analyses are 1.31 and 1.81, respectively, for rainy and dry seasons.
790
Figure8. Effective stress path from DST-3 and DST-4
Figure9. Results of stability analysis
8 CONCLUSIONS
A series of direct shear tests were conducted for undisturbed heavily weathered tuff stratum and the colluvial deposit and compacted heavily weathered tuff with partially saturated and submerged condition and the shear strength parameters applied on the stability analysis are evaluated as shown in Table3. The minimum safety factor calculated is 1.81 for dry season, and 1.31 for rainy season. The banking structure was completed in 1996 and no problem has occured until today. REFERENCES Hanzawa,H. 1983. Three case studies for short term stability of soft clay deposits. Soils and Foundations. 23, 2, 140-154. Hanzawa,H. 1993. Determination of in-situ shear strength for earthworks on soft marine clay, Special Lecture, Nanyang Technological University, Singapore, 1-17.
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Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Effect of degradation on the strength of rock Akira Kobayashi, Kiyohito Yamamoto & Koichi Fujii Iwate Universio, Morioka, Japan
ABSTRACT: Influence of freezing and thawing on rock slope stability is investigated by carrying out the laboratory test. The rocks used for the examination are granite and tuff. The difference between the intact and degraded specimens is investigated for mechanical parameters and distribution of strain by uniaxial compression test. The strain distribution is observed through the image processing. As the results of the test, it is found that the change in mechanical behaviors of granite is quite different from that of tuff. Granite becomes brittle with degradation while tuff becomes hard rock. It is inferred that the volumetric strain distribution may be good measure for failure because the area of expansive strain becomes larger closing to the failure. 1 INTRODUCTION Toyohama tunnel at which collapse of rock slope occurred is located in a cold area (JGS report 1997). The rock slope is exposed to freezing and thawing at such an area in winter. Therefore it is inferred that the degradation of rock mass by freezing and thawing is a big reason of the rock slope failure, moreover the crack growth by icing of ground water. It is the purpose of this study to grasp the degradation of mechanical properties of rock by freezing and thawing. Uniaxial compression test is carried out for measuring some mechanical properties of rock, axial stress-st rain relation, volumetric stress-strain relation and strain distribution. The stress-strain relation is measured by strain gauge. The distribution of strain is measured by image processing with the digital video camera. In this paper, examination is curried out for granite and tuff. The change in mechanical behavior is examined in comparison with the results for intact and degraded specimens.
elastic wave velocity V p . After the compression tests, the moisture content of Kurhashi-granite is measured as 0.14.2%, that of Funyu-tuff is O.8-2.8%. Both intact and degraded specimens are taken from the same rock block. The rock block for the degraded specimen is imposed in freezing and thawing condition. After that, the column specimen is taken from the block. The size of the column specimen is the diameter of 4.5cm and height of 10-llcm. The edges are cut and polished. As shown in Figure 1, eight strain gauges are set on the specimen. Table 1. Basic DroDerties of intact material
Kurihashi-granite
2.7
2.8
F ~ y utuff -
1.7
2.0
1.5
32
0.5
4.59
19
2.24
2 MATERIAL AND LABORATORY EXPERIMENT 2.1 Material The materials used the study are Funyu-tuff and Kurihashi-granite. Table 1 shows the basic mechanical properties, i.e., dry density pd, wet density pt, effective porosity n, void index i and the
Figure 1. Setting of strain gauges and scale of specimen 793
2.2 Degradufion
2.4 Strain distribution with image processing
Instead of measuring the change in temperatures in the rock block during freezing and thawing the temperature in the concrete pillar (40cm x 10cm x 7.5cm) is measured with the thermometer for observation. Figure 2 shows the change in temperature of the concrete. The temperature rises up to +6"C, then falls till -18°C. The thawing process is made by exposing to water. One cycle takes 90 minutes. Degraded specimens experience 240 cycles of freezing and thawing The column specimens are obtained by hollowing out the block. At that time, much damaged parts due to freezing and thawing are excepted.
The strain distributions are obtained from the pictures recorded by the video camera. The specimens have 50-60 marks on the surface as shown in Figures 6 and 7. The behavior of the marks is recorded with the video camera. The pictures are processed with computer to obtain the coordinate of the center of each mark. Each mark contains about 10 pixels. The coordinate of the center of the mark is calculated as the center of the gravity with pixel coordinate. The displacements are obtained by subtracting the initial coordinates from the coordinates after deformation (Figure 4). The tangential displacements are adjusted for the curve surface of column specimen. The triangle element consists of three marks. The strains are calculated for each triangle element with the same procedure as that of FEM. The accuracy of the strain is about 2 0 0 0 ~in t h s case.
Figure 2. Change in temperature during freezing and thawing 2.3 Measurements and analyses
Uniaxial compression test is carried out for each specimen. The number of the specimen is two for each case. The strain of the specimen is measured with strain gauge and the stress-strain relation is examined. About 50-60 marks are set on the surface of the specimen. The behavior of the specimen is recorded with the digital video camera. The displacement of the mark is analyzed after the test and the strain distribution is estimated. Figure 3 shows schematic view of the equipment. Loading is carried out with constant strain rates of 1-4pFsec. Specimens of Funyu-tuff are capped with plaster (Kobayashi 1998), while those of granite are not capped. The axial stress, tangential strain, volumetric strain, secant elastic modulus, secant Poisson's ratio, secant bulk modules-axial strain relations are obtained.
Figure 3. Equipment of uniaxial compression test
Figure 4. Estimation of displacement 3 TEST RESULTS
3.1 Mechanical behmiors Table 2 shows the results, in which compression is positive. In t h s table, is the axial strain at the maximum stress qu, ~ 5 IS0 the scant Poisson's ratio at a half stress of qu, q m a x is the maximum volumetric strain, EyvMan is the axial strain at the Evmap EyvmaxlEyqu shows the ratio of the axial strain at which the volumetric strain changes from compression to expansion to the maximum axial strain. K indicates Kurihashi-granite. F means Funyu-tuff and follwing number of K and F is the number of cycles of freezing and thawing The number of specimens is two for each case. Figure 5 shows various mechanical parametersaxial strain relations. In this figure, the axial strain is normalized with EYqu for the horizontal axis, which is called the axial strain ratio in this paper. The black symbol is the results of the intact specimen and white indicates the ones of the degraded specimen.
794
3.2 Consideration of mechanical behaviors
For Kunhashi-granite, E50 and qu are not much changed after degradation, while the ~vma./Ey4u becomes large after freezing and thawing It is found from Figure 5(c) that the volumetric strain of the degraded specimen is drastically changed from compression to expansion at the stage close to failure. The change in the volumetric strain is expected to be caused by the occurrence of the cracks in the specimen. This phenomenon is observed by AE measurement (Scholz 1968). Thus, the rock after degradation may be failed drastically. The minimum extreme value of the scant elastic coefficient is occurred at the larger axial strain ratio after degradation.T h s is probably because the inner cracks caused by degradation are closed at early stage and the scant elastic coefficientbecomes small. After the extreme value, the elastic stiffness increases till failure. It is concluded from above consideration that the granite becomes more brittle after degradation. For Funyu-tuff, it is found that the secant elastic coefficient and uniaxil compression strength of degraded specimen become larger than those of the intact one. The scant elastic coefficient becomes steady state at mostly the same axial strain ratio for both intact and degraded materials, while the decreasing rate of elastic stiffness from the initial loading is small after degradation (Figure 5(d)). It is inferred that the existing inner cracks are closed by freezing and thawing for Funyu-tuff. &vma./Ey4u after degradation comes to be small and the compressive volumetric strain from the initial loading is also small after degradation. It is concluded that the inner failure may start at the earlier stage than intact specimen. This can be seen in the decreasing of the elastic stiffness of the degraded specimen from the axial strain ratio at whch the peak of the volumetric strain is occurred. Table 2. Test results Rock
Specimen
ESO (GPa)
vso (p)
E,
E-
(p)
JEYP
111.7 54.2 2121 0.28
482
0.60
82.6
34.8 2313 0.22
672
0.63
K240-1
88.4
44.7
1831 0.22
612
0.69
K240-2
101.4
42.5
1848 0.25
662
0.60
granite KO-1
KO-2
tuff
qu @@'a)
FO-1
21.4
5.2 4159 0.28
1077
0.68
FO-2
21.1
5.4 3987 0.26
1183
0.74
F240-1
27.6
7.6 3762 0.30
826
0.52
F240-2
30.6
7.0 4768 0.29
1054
0.61
Figure 5. Mechanical parameter-axial strain relation 795
Figure 6. Strain distribution,and picture that fractures appear 3 -3 Struin distributions
Figures 6 and 7 show the pictures at failure and distributions of shear and volumetric strain at the various stages before failure. The meaning of failure in these figures is that the next frame of the video tape shows the collapse of the specimen. The specimens are collapsed in a moment after t h s frame. The scale of strain is logstrain. The positive sign is expansion and negative one is compression. First picture from the left side, (a), is the strain ~ second ~ ~ one, . distribution at about half of E ~ The (b), is that at the qu.From the third picture to the sixih one, the strain distributions are presented
accordingto time history up to the failure from the maximum stress state. The horizontal axis of each figure indicates the time to the failure. The fractures superimposed on the strain distributions are the ones at the failure. 3.4 Consideration of Strain distributions
For Kurhashi-gitnite, comparing of shear strain distributions of (a) and (b), the big change is not found for the intact specimen, while the direction of shear at the middle part is changed for the degraded one. This may mean the local change in the principal 796
Figure 7. Strain distribution, and picture that fractures appear stress direction. However, it is difficult to find the relation of the fracture pattern to the shear strain distribution for both cases. The direction of cracks at the failure is mainly vertical for both intact and degraded cases (Figure 6 pictures). The area of expansive volumetric strain becomes large at the maximum stress state (b) in comparison with the picture of (a). After the maximum stress state, the area of expansion comes to be large gradually to the failure. The cracks are mainly caused at the expansion area. For Funyu-tuff, the shear strain distribution is not changed so much to the failure state through the maximum stress state fi-om the earlier stage for both
intact and degraded specimens. On the other hand, the cracks at failure are caused at the expansion area and the area of expansion becomes large with time similarly to the cases of granite. The direction of the cracks at the failure is vertical for the intact case, while that for degraded case is a little skewed. 4 CONCLUSIONS
To investigpte the effect of the degradation due to freezing and thawing on the mechanical behavior, uniaxial compression tests are carried out for intact 797
and degraded rocks. The various mechanical properties are compared with the ones after degradation. The historical change in strain distribution of the surface of the specimen up to failure is also compared. As conclusions, the followings are found; 1)The volumetric strain of the degraded specimen of granite is drastically changed from compression to expansion at the stage close to fdure. This means that the granite rock after degradation may be failed drastically. Granite becomes more brittle material after fieezing and thawing. 2)The uniaxial compression strength and elastic stiffness of Funyu-tuff become more hq$ after freezingand thawing However, the inner failure of the degraded case may start at the earlier stage than the intact case. The shear failure may become main cause of the failure for the degraded case, while the tensile failure is main for the intact case. 3)The inner cracks derived from freezing and thawing may be occurred for the granite specimens, while the existing cracks before freezing and thawing may be closed for the tuff specimens. This effect is seen in the change in elastic stiffness. 4)The shear strain distribution on the surface is not related to the failure process, while the volumetric strain distribution is much correlated to the crack pattern. This indicates that the observation of the volumetric strain on the slope surfice is effective for the monitor of slope failure. REFERENCES The Japanese Geotechnical Society 1997 Report on rock slope failure at Furubira. Kobayashi , K. 1998. Diagnosis of degradation of concrete structure. Morikita. Scholz,C.H. 1968. Micro Fracturing and the Inelastic Deformation of Rock in Compression. J.Geophys Res., Vo1.73: 1417-1432.
798
Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Some considerations of Patton model on rock joint shear strength Masanobu Doi & Satoru Ohtsuka Department of Civil und Environmental Engineering, Naguokn Universiry of Technology, Japan
ABSTRACT: It is well known that the shear strength ofjoint included in rock mass considerably affects the stability of rock slope. There are many experimental researches on the shear strength of rock joint. The aim of this study is to evaluate the shear strength of rock joint by using analytical procedure. Following the experiments carried out by Patton( 1966), the analysis is performed under varying the number of asperity and the inclination of asperity. The followings are concluded in this study:(l) the shear strength of rock joint increased with increasing the number of asperity and the inclination of asperity; (2)the change of the shear strength of rock joint against the normal stress was shown to be properly expressed by the change of failure mode; and (3)the proposed analytical procedure was shown to be applicable to jointed rock. 1 INTRODUCTION
The aim of this study is to investigate analytically the shear strength of rock joint by considering the asperity of joint after Patton’s work and make clear of the shearing mechanism of rock joint. The numerical simulation can consider the ideal condition of rock joint different from experiments. Focusing on the effects of asperity along rock joint, the resultant shear strength of joint is investigated for the various asperity conditions. The analytical procedure, which has been developed and applied to rock stability(e.g. Ohtsuka, et al., 1997; Ohtsuka and Doi, 1998), is employed in this study. With the use of finite element discretization, the analytical method is cast into a simple linear programming problem as Maier( 1969) proposed. By introducing the contact condition along the joints into the lower bound method, the redistribution of stress in rock mass and traction along joint is well considered in the stability analysis for the generation of plastic deformation in rock mass and/or both sliding and detachment along rock joint. In this study, the plane strain compression test of a rock specimen including a joint is simulated. The followings are mainly discussed:( 1)The effects of joint asperity such as a number, size and inclination of micro structure are investigated on the joint shear strength; (2)The mechanism of failure mode change depends on both the normal stress and the rock shear strength. Non-linearity in joint shear strength is investigated from the viewpoint of failure modes of sliding along joint, shearing of rock mass (asperity)
Natural rock slope often includes a various kind of joints(cracks), which widely range from micro to huge ones. The existence of rock joints is well known to affect the stability of rock slope. The effects of rock joints on the stability depend on their sizes, geometrical conditions, shear strength properties and so on. These factors should be clarified in the design of rock slope stability, especially the shear strength property of rock joints has been investigated by many researchers. Shear strength of rock joints has been clarified experimentally to depend on the traction along the joint, material properties of rock, especially the contact friction angle and the joint roughness. The experimental studies on joint shear strength can be categorized into the following two types. One is to investigate directly the effects of joint roughness on the shear strength by simplifying the roughness into a regular asperity. The number, size and inclination of micro structures(asperites) constituting the roughness of joint have been widely investigated. The most notable contribution to this research was by Patton(l966). The other is to investigate the shear strength of joints indirectly through the dilation model as proposed by Taylor(1948). The experimental and theoretical studies on joint shear strength were conducted by Ladanyi and Archambault( 1970) and Barton( 1973). They are based on the results by Patton. 799
without sliding along joint, and both the sliding along the joint and rock mass shearing.
determined by solving the boundary value problem.
2 ANALYTICAL, PROCEDURE
The analytical procedure employed in this study is based on the lower bound theorem in plasticity. The rock mass is modeled as an elastic perfectly plastic material. 2.1 Lower Bound Theorem in Shakedown Analysis
with Linear Programming Problem The lower bound theorem assures that a rock mass is stable against the external force F(t) if any time independent residual stress Or, which is statically admissible stress, can be found everywhere in the rock mass. If a rock mass is stable for the applied load, the behavior of it is proven to shakedown to be elastic against any repeated load. When the external force is a monotonically increasing force, the shakedown analysis coincides with the limit analysis. With the use of linear yield function as
Fig. 1 Piecewise linear yield function..
2.2 Stability Analysis Considering Contact
Interaction Along Rock Joint The traction q along rock joint consists of the normal stress q, and the shear stress qs. On the normal stress qI1,the extension in stress is usually not permitted such as qn50 when the extension in stress is defined as positive. If rock joint has a certain cohesion c,, the contact condition on qn is followed by q,Ic,. The contact condition on the shear stress qs can be defined by introducing shear models. Using Coulomb's shear model, the contact condition on q,,. is described as q,, tan #p - cs I q,sI -4, tan I $ ~ + cs where 4 is the angle of frictional sliding resistance along rock joint. These contact conditions on qn and qs are expressed in the following equation:
the shakedown analysis can be formulated as a linear programming problem (Maier, 1969). In Eq.(l), the finite element discretization is introduced into the stress The elastic stress 0' satisfy the equilibrium equation. 0' + 8' = 0" is any safe stress which satisfies a yield function of rock mass. 0, indicates the initial stress which equilibrates with body force. N expresses the matrix constituted from different unit vectors n, and K , the assemblage vector of threshold values k, as n,,o5 k,. The relationship between n, and k, is illustrated in Fig.1. When the yield hnction is non-linear, it is linearized piecewisely, as shown in the figure. The analysis against the external force, F can be formulated with a load factor a for F as follows:
'CS
1 (3)
Eq.(3) indicates the constraint condition on possible stress field in rock mass so that Eq.(l) is replaced by
with the contact condition of Eq.(3) along rock joint. The traction q is a stress vector along rock joint., which is introduced into the equilibrium equation by employing the joint element which was developed by Goodman et. aL(1968). The stability analysis for rock mass including joints is formulated by using the constraint conditions of Eq.(4) and the equilibrium equation as follows:
where s is the ultimate load intensity and B, a matrix correlating the stress vector with the force vector, F. In Eq.(2), the first equation indicates the yield function of rock mass. The second and third equations express the equilibrium equations on elastic and residual stresses, respectively. It should be noted that the redistri-bution of stress is considered with the residual stress which is 800
s =
max
CY
(5)
The contact conditions are considered as the yield function for the joint elements. The detachment and sliding along rock joints are taken as the plastic deformation of joint elements. The redistribution of stress and traction is taken into account with the residual stress (T" and the residual traction 4 along joints. The residual stress 3''and traction ( I r are determined by solving the boundary value problem. The joint element method introduces two springs to rock joint such as the normal stiffness k,, and the shear stiffness k,. However, the physical meaning of introduced stiffness for joint element is not clear. By taking a large value for the joint stiffness, the rational result is obtained from the viewpoint of rigid plastic assumption on joint behavior
plotted data for line A and B are apparently nonlinear. It is readily seen that the shear strength of joint for four asperities is larger than that for two asperities. Fig.4 shows the effect of asperity inclination on the joint shear strength. In this figure, line A, B, and C denote the failure envelopes for joint strength in the case of the asperity inclinations as i = 250 , i = 350 , and i = 450 , respectively. Line D is the same as the line C in Fig.3. The plotted experimental data corresponding to line C and D are also non-linear, but they can be well approximated to bilinear relationships.
3 PATTON'S EXPERIMENTS
In writers' opinion, the experiments conducted by Patton( 1966) affected considerably the later studies on shear strength of rock joint by many researchers. Patton carried out the direct box shear tests on specimens composed of kaolinite and gypsum plaster as shown in Fig.2. Each specimen had 2.95inches (7.49cm) long, 1.75inches (4.45cm) wide, and 2.0inches (5.08cm) height. The results were exhibited as shown in Fig.3and 4.
Fig.2 Some of the different types of specimens (4 asperities : after Patton, 1966). Fig.3 shows the effect of numbers of asperities on the shear strength of joint. Line A indicates the failure envelope for joint strength in the case of four asperities, and fine B, that for two asperities. Line C shows the residual strength for all specimens. The
Patton derived the following conclusions from the results above mentioned. The actual failure envelopes for joint strength are non-linear against 801
the normal stress. The change in the slope of failure envelope expresses the change in failure mode. The inclinations of primary portion in failure envelopes are equal to4 + i as line A, B, and C in Fig.3 or line A, and B in Fig.4. 4 is the angle of frictional sliding resistance for joint surface. The inclinations of secondary portions of failure envelopes are close to 4,. which is defined as the angle of residual strength. The changes in failure mode are found related to the physical properties of asperites along the joint.
rock. The corresponding strength parameters of cohesion, c and angle of shear resistance, 4 under the plane strain condition are exhibited in the table. The stiffness parameters for the joint element are assumed very large as explained before. Table 1, Material Constants. intact rock E 1500.0 MPa, c 5.0 MPa, ~
joint
4 ANALYTICAL RESULTS AND DISCUSSION
In this study, the plane strain compression tests are simulated to estimate the effects of joint roughness such as the number and the inclination of asperities on the shear strength of rock joint. 4.1 Conditions of Calcirlation Fig.5 illustrates schematically the jointed rock model under the confining stress a, and the deviator stress a d. The employed analytical method gives the maximum value of ad at the limit state. Each specimen is nearly equal to 6.0(cm) long and 12.0(cm) height. The mean plane of joint crosses the horizontal plane at angle of 8 . The inclination of asperity is defined as the angle i between the asperity and mean plane of joint. This figure illustrates the case of asperity number, N as 1.
F i g 5 Schematic ofjointed rock model.
The employed material constants for intact rock and rock joint are shown in Table.l. The yield function of Drucker-Prager is employed for intact 802
k,, and k, c,, and c,
4P
.y
0.2
4 30" _
_
~
1oi4H a j m
1.0 kPa variable
8 and $ are fixed at angle of 600 and 300 , respectively, through this study. Therefore, the direction of mean joint plane basically coincides with the direction of failure plane for intact rock since 8 equals to 450 +$/2. The shear strength of joint is estimated on the mean joint plane by considering the normal stress a, and shear strength rj. along the prescribed plane. These stresses are determined by the principal stresses of a ,= a f a d and a = a c, and the direction of mean joint plane, 0 .
4.2 Shear Strength of Joint WIthout Asperites
Before investigating the effects of asperities on the shear strength property of joint, the shear strength of joint in the case of no asperites is evaluated first. Even if there is not any asperity along the joint, the contact resistance still works due to friction property. It might be caused by micro asperity which is categorized into the 2nd and/or higher orders. In the case study, the angle of frictional sliding resistance 4 along the joint is taken into consideration. Fig.6 shows the effects of increasing 4 on shear strength of rock joint. The straight lines indicating the failure envelope in terms of rr and a ,, are well graded with q5p. The inclination of each straight line is obtained to be identical with employed $ p . This fact indicates that the joint slides along the flat joint and the resultant shear strength of joint is described by the simple friction law. However, in the , the failure envelope reaches the case of 1$~=450 Coulomb' failure criteria(dashed line in the figure) of intact rock at high normal stress a,. In this case, the angle of frictional sliding resistance 4 for joint surface already becomes greater than the angle of shearing resistance 4 for intact rock and then, the physical meaning is lost. However, the failure mode naturally changes into the intact rock failure from the sliding failure along the joint.
4.3 EJect of The Number of Asperities on Shear Strength The effect of asperity number on the joint shear strength is investigated. Fig.7 shows the results of computation on the cases of the number of asperities as N=O, 2, and 4.The inclination of asperities and the angle of fiictional sliding resistance of joint surface are kept constant as i =30° and 4 =loo , respectively. The case of N=O corresponds to the case of flat joint. On the whole, it is clear that the increase in N results in large shear strength cf. It is readily seen that the failure envelopes for the cases of N=2, and 4 are non-linear against the normal stress. The inclinations of primary portion of these failure envelopes are exactly identical to the angle of g p + i =40° . The inclinations of secondary portion of these failure envelopes become much smaller. These results indicate the possible change in failure modes with increasing the normal stress 0 , along the joints. That is to say, the sliding failure along joint takes place at low normal stress and the asperities are hlly sheared at high normal stress. The combined failure of the partly sliding along joint and shearing of asperiteis takes place within the range of middle normal stress. The shear strength at the transition point in each failure envelope from the primary to the secondary portion becomes higher with the increase in asperity number. It can be thought that the degree of roughness of rock joint depends on the number of asperities. But the filly failure of intact rock does not occur since each failure envelope does not reach the Coulomb's failure criteria which is shown as dashed line in the figure. The same tendency can be seen for the different i and 4 p . Although there are some differences on the basic conditions between Patton's experiments and these analyses, the results of Fig.3 and 7 are found to be almost same. 4.4 Eflect of Asperity Inclination on Joint Shear Strength The effect of the inclination of asperities on the shear strength of rock joint is investigated here. The results of computation on the cases of the asperity inclinations of i=Oo '15' ,300 and 450 are shown in Fig.8. The number of asperities and the angle of frictional resistance of joint surface are kept constant as N = 4 and 4 =loo in the following analyses. The case of i=Oo is the same with the case of flat joint. The shear strength T~ of rock joint increases with the inclination of asperities i. The failure envelopes for the cases, i=30° and 450 are obtained as nonlinear against the normal stress. However, they seem to be modeled into bilinear models. The inclination of primary portion in each failure envelope is obtained as identical to 4 i. 803
The inclination of secondary portion in failure envelope is smaller than that of primary portion. It seems a little smaller than the angle of shear resistance for intact .rock, but those for i=150 and 3@ are obtained as almost same. However, two lines of secondary portions for i = l Y and 300 are different each other. It is not clear why these two are different. Meanwhile, the results of Patton’s experiments and the conducted numerical analyses seem almost same even though some differences exist in testing methods. The employed analytical procedure is found applied to the analysis of joint shear strength even if the method is based on the framework of continuum mechanics. The obtained results indicate that the shear strength and the resultant failure mode of rock joints are strongly affected by the property of joint asperites and the change in the normal stress o n along the joints. The important factors of joint asperities to affect the shear strength of joint are (1)geometric condition of triangular asperity as an asperity inclination, a number of asperities and others, (2)shear strength of intact rock, and (3)angle of frictional sliding resistance for joint surface. Since the geometric condition of asperites for an actual joint is more complicate, the shear strength of actual joint naturally becomes more difficult to be evaluated by experiments. Furthermore, it is very difficult to investigate the effect of material property of intact rock and angle of frictional sliding resistance for joint surface on the resultant shear strength of joint by experiments. It is possible for the numerical approach to investigate the joint shear strength for various conditions. 5 CONCLUSIONS
In this study, the analytical procedure based on the lower bound theorem in plasticity was employed to investigate the shear strength of rock joint. After Patton( 1966), the effects of the asperity conditions on the resultant shear strength of rock joint were investigated. The followings are concluded in this study. 1. In the case of the flat joint without any asperity, the resultant shear strength of joint obeyed the simple friction law. The change in failure mode was simulated well from the sliding along the joint into the failure of intact rock by considering the confining pressure and the shear resistance along the joint. 2.The shear strength of joint with asperities was investigated widely for the various conditions on geometric condition of triangular asperity as an asperity inclination and a number of asperities. The effects of the shear strength of intact rock and the physical angle of friction for joint surface were also investigated.
804
3. The joint shear strength was obtained non-linear for the normal stress mobilized along the joint. The obtained result was almost same with the experimental results by Patton( 1966) even though there existed some differences in testing methods. This suggested that the applicability of numerical method to estimation of joint shear strength. 4. The shear strength of joint increased with the increase in both the number of asperities and their inclinations. Depending on the change in failure mode, the mobilized shear strength could be modeled into a bilinear relationship for the normal stress. The simulated results were well explained by the Patton formula.
ACKNOWLEDGEMENTS The writers are gratehl to Mr. J. Takeuchi of West Japan Railway Co., Mr. M. Hashiba of Meiken Co. and Mr. Y. Hara of Nagaoka University of Technology for their helps and valuable comments during this. REFERENCES Barton,N.R. 1972. A model study of rock joint deformation, Int. J. Roch Mech. Min. Sci.,Vo1.9, pp.579-602. Goodman,R.E., Taylor,R.L. and Brekke,T. 1968. A model for the mechanics of jointed rock, Proc. of ASCE, 94, SM3, pp.637-659. Ladanyi,B. and Archambault,G. 1970. Simulation of shear behaviour of a jointed rock mass, Proc. of Ilth Symp. RockMech., AIllrlE, pp.105-125. Maier,G. 1969. Shakedown theory in perfect elastoplasticity with associated and nonassociated flow-laws: a finite element linear programming, Meccanica Vo1.4, No.3, pp.1-11. Ohtsuka,S., Yamada,E. and Matsuo,M. 1997. Bearing capacity analysis of rock structures including cracks, Proc. of 9th Int. Con$ of Int. Assoc. for Comp. Mech. Adv. on Geomech.,Vol.1, pp.739-744. Ohtsuka,S. and Doi,M. 1998. Stability analysis of jointed rock slope, Proc. of 3rd Int. Con$ on Mech. of Jointed Rock and Faulted Rock, pp.523-528. Patton,F.D. 1966. Multiple modes of shear failure in rock, Proc. Ist Cong. I S M . , Lisbon, pp.509-5 13. Taylor,D.W. 1948. Fundamentals of Soil Mechanics, Wiley.
Slope Stability Engineering, Yagi, Yamagami & Jiang t) 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
ehavior of jointed model material under biaxial compression Anil Kumar Tyagi Delhi Developnzent Autlzorily, India
KSeshagiri Rao & Anand S.Gupta Department of Civil Engineering, Indian Institute qf Technology, N e ~ pDelhi, India
ABSTRACT. The strength and deforiiiational behaviour of jointed rock mass is the basic input for tlie design of slopes. foundations aiid underground structures. Most of the Civil engineering activities are at sui-Face 01sliallow depth where rock inass is subjected to the negligible or low confining pressure conditions. In-situ testing for the determination of strength and deforniational parameters is mostly expensive and time consuming lience not feasible most of the time. It is always desirable to follow a detailed and experimentally proven approach for tlie preliminary assessment of behavior of rock inass. In the present study a n attempt h a s been made towards this direction to understand the strength reduction and deforniational behaviour of jointed rock under biaxial stress state. 1 INTRODUCTION
joints used were continuous only. Very few studies have been conducted on large size specimens representing a rock mass having discontinuous joints by Hoek and Brown (1980): Ladanyi and Archambault (1972) and Einstein and Hirschfield (1973). It has been observed that very few studies have been made to understand the strength and deformational behaviour of jointed rock inass under biaxial stress condition. In tlie present work, experiments were carried out on a jointed model of rock mass made up of around 288 elements of sand lime brick material. The specimens were subjected to biaxial confining pressure in order to understand the effect of geometry and orientation of discontinuities on strength and failure pattern.
In nature, tlie homogeneity or isotropy in rock inass is very rare. and the influence of joints and other physical defects are important factors which affects not only the strength but also failure pattern of the rock illass. This has been studied by several investigators i n tlie past. The single plane of weakness theory was given by Jaeger and Cook ( 1 969). General shear strength model by combining 1.1-iclion, dilatxicy and interlocking was given by Ladanyi and Archanibault (1972) and RMR classification was proposed by Bieniawski (1974). I-Ioek and Brown (1980) also linked RMR to tlie parameters of their empirical strength criterion for predicting strength of-*jointedrock mass. Since these theories have their own limitations, an improvement has been felt. l i has been proposed by Ramamurthy and coworkers for predicting both the strength and deforniational response of the jointed rocks (Raniamurthy 1993, 1994). The approach computes a weakness coefficient based on properties of' the most critical joint set. The uniaxial compressive strength (aci) and deforinational moclulus (E,) of tlie jointed rock are then linked to those of intact rock through this coefficient. The concept derives it's base from a large number of tests conducted on jointed rocks by Yaji (1984), Aroi-a (1987) and Roy (1993). These investigators used 76 nini high cylindrical specimens and the
2 LABORATORY STUDY In the present investigation for ease of working and reproducing of results a model material (sand-]inie brick) has been selected aiid characterized. A typical configuration of the specimens tested are given in Fig. 1. Tlie specimen is prepared by arranging the individual cut blocks. Tlie test specimens were divided into 3 groups depending on the geometry of' constituting elements, designated as Type A, B and C . The Type-A (Fig. 1) specimens used cubical block as elenients. This has 4 subgroups of' specimens depending on their angle of inclination of'
805
The specimens tested under biaxial compression condition were carefully observed for their failure modes and the strains were monitored in three directions. Strength was measured at varied biaxial stress (oJ values for specimens having elements of varied geometry and inclinations. 3 EFFECTS OF GEOMETRY The relation between the geometry (i.e. h/b i-atio) and ocr (ratio of oC of jointed and oC of intact material) has been shown in Fig. 2. The figure shows that for low h/b ratio, the value of a,, is higher and as the h/b ratio increases, the value of oCrdecreases. It is observed in the present study the effect of geometry is much more pronounced than the parameters in Joint Factor (Ramamurthy, 1993. 1994) concept. It is observed that U,, drops to cl 1n1ost half when h/b ratio is increased from 1:1 to 2 : l . This drop in strength with increase in h/b ratio was due to the fact that number of joints are decreasins with increase in h/h ratio. This anomalous observation may be attributed due to the fact that as h/b ratio increases, center of gravity of the individual element falls out of the base of the element causing reduction in strength.
Figure 1. Configuration of Type-A specimen. critical joint set. Types-B and C used rectangular blocks with different dimensions and a constant inclination angle, as discussed below. 2. 1 Tvpe-A speciineiu The size of the specimen was 15 cni X 15 cin X 15 cni consisting of about 288 elements (2.5 cm X 2.5 ciii x 2.5 cm). The specimen finally formed out of cut blocks consisted of three sets of joints. The joint set I was continuous and inclined at constant angle 8. with the horizontal. Value of 8 adopted was 80". The joints set I1 was orthogonal and perpendicular to set I. The joint set I11 remains vertical for all the specimens and is assumed to have no appreciable effect on variation of mechanical response of the speciiiien. A total of f-our specimens were tested under this category. In subsequent 3 specimens inclination angle, 8 of the joint set I was varied at 60", 40" and 30".
In this ca~egory,the elements have a base width b = 2.5 cni and height (h) = 3.75 cm, thus keeping b/h ratio as 1:1.5. Approximately about 210 elements make the one specimen under this category Four such specimens have been tested at diftei-enr confining conditions. The Type-C specimens have elements with dimensions as b =2.5 cin and h=5.0 cni (b/h = 1:2). A total of 168 rectangular elements would make a single block spccimeii and 4 such specimens have been prepared. For both Types-B and C, the orientation 0 values uas kept 80".
Figure 2. Plot between
806
U',
and h/b at 8=80"
From the Fig. 2 a relationship between h/b and acr,is developed and given as:
Thus by knowing b/h ratio of the element, one can predict value of a', for jointed rockinass having joints at an inclination of 80" with horizontal. In the Fig. 3, plot between J, (Joint Factor) and acrshows that as the number of joints reduces, the strength also gets lowered. This contradicts the expected behaviour noticed in earlier studies.
Figure 4. Plot between (3 and
(gl/03).
5 MODE OF FAILURE The modes of failure in a jointed inass IS a combination of more than one failure niechanisni Out of all combination available for distinct modes were identified (1) splitting, (2) shearing (3) rotalion for sliding along critical joint planes. The derailed modes of failure is indicated in Table 1. For specimen Type-A, opening of vertical joints. staggered joints, peeling of surface of some of the elements in top and bottom layer were observed Crushing of few elements in top and bottom layer is also noticed. The distinct modes of failure are observed in splitting and rotation. For low confining pressure (ai = 7 kPa). initially the mode of failure is splitting upto 50% of the Failure load and thereafter rotation of blocks is observed. As the confining pressure increases, the effect of rotational failure goes on diminishing and it is observed that there is no rotation upto 80% of the failure stress. However the final failure occurs in rotation of blocks. In Type-A specimens for all inclination
Figure 3. Variation of oCrwith J,. . 4 EFFECTS O F ANISOTROPY The effect of anisotropy has been shown in Fig. 4. The figure shows U shaped curve for all the cases of wide base having maximum strength at /3 = 0" and 90" and inininiuin strength at 30" and 60". Where /3 is the angle between critical plane of joint and axis of loading. This is also observed that the effect of anisotropy goes on diminishing as the lateral horizontal increases.
807
angles rotational failure was observed in final stage of loading. Table 1. Detailed Observation on Modes of Failure. 8 Mode of Failure Group b:h
A
1:l
80 "
Primary failure is due to splitting of blocks Sliding on joint set I Rotation at high deformation towards right
A,
1:l
30"
Sliding of blocks at initial stage Final failure at rotation on left side
A2
1:1
40 "
Sliding of blocks; rotation on left side at final failure
A,
1:l
60 "
Sliding of blocks; rotation on right side at final failure
€3
1: 1.5
80
O
Primary failure is due splitting of blocks Sliding on joint set I Rotation at high deformation towards right
C
1:2
80
O
Primary failure is due splitting of blocks Sliding on joint set I Shearing of the intact material Rotation at high deformation towards right
6 CONCLUSIONS The strength of the mass depends on the geometry of the blocks forming the specimens. The reduction in strength is observed when the height of elements increases with respect to width. For specimen C about 50% reduction in strength is observed. It is also interesting to note that as the number of joints reduces, the strength gets lowered. This observation contradicts common results noticed by several researchers. It may be due to the fact that the increase in h/b ratio moves the center of gravity of individual element outside the base of the element causing reduction in strength. Slenderness ratio of the elements reduces the strength of block as it increases. Though the major modes of failure were observed as splitting, shearing and rotation of blocks but specimen failure was primarily governed by
808
splitting and rotation. The effect of inclination of' critical joint set shows that at U = 0" and 90" strength is maximum and 8 = 30" and 60" is least. 7 REFERENCES Arora, V. K. 1987. Strength and deformation of jointed Rocks. Ph.D. Thesis IIT Delhi. Bieniawski, Z. T. 1974. Geomechanics classification of rock masses and its application in tunnelling. Proceeding 3"' Itit. Cong. Rock Mech., Detivet-, Pt. A, pp27-32. De, N. 1997. Strength and deformational behaviour of jointed model materials. M . Tech. Tlwsis IIT Delhi. Einstein, 13. H. and Hirschfield. R. C. 1973. Model studies in mechanics in jointed rocks. JI. SMFE P ~ o c ASCE . Vol. 90 - SM2 ~ ~ 2 2 9 - 2 4 8 . Hoek, E. and Brown, E. T. 1980. Empirical strength criterion for rock masses. JI. Georecli. Engg. Div. ASCE, Vol. 16, pp1013-1035. Jaeger, J. C . and Cook, N. G. W. 1969. Fundamentals of rock mechanics: Chapman and Hall, London, pp5 13. Ladanyi, B. and Archanibault, G. 1972. Evaluation of shear strength of a jointed rock mass. Proceeding 24"' Int. Geological Cotzg., Motirreal. pp249-270. Ramamurthy, T. 1993. Strength and modulus response of anisotropic rocks. Comprehensive Rock Engineering: Pergamon Press, Vol. 1 Clip. 13. pp313-329. Ramamurthy , T. 1994. Classification and characterization of rock mass: Theme Paper. Proc. CBIP Workslzop, Tuntzellitig Itiditi 1994. Roy, N. 1993. Engineering behaviour of rock inasses through study of jointed models. PI?.D. Thesis IIT, Delhi Singh, M. 1997. Engineering behaviour of jointed model material. P1i.D. Thesis IIT IJclIii. Tyagi, A.K. 1997. Strength and deforniational behaviour of jointed model materials. M. Tech. Thesis IIT Delhi. Yaji, R. K. 1984. Shear strength and deformation of' jointed Rocks. Ph.D. Thesis IIT Delhi
7 Slope stability of landfills and waste materials
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Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Stability of slopes of hydraulic-fill dams A.Zh. Zhusupbekov Karaganda Metallurgical Institute, Temirtau, Kazakhstan
A.S.Zhakulin & M. R. Nurguzhin Karagandu National Technical University, Kazakhstan
ABSTRACT: The experimental data of stationary observations of pore pressure in a body of a hydraulic-fill dam have been given. The analysis of the results shows it is necessary to take into account of pore pressure at the account of stability of slopes of hydraulic-fill dams. of a slope on maps of 200x250 in size with a bypass of pulp into additional sumps (a pond zone). They are arranged on the site of the buck waveresistant wedge.
1 INTRODUCTION
One of the stages of a structure erection can appear the most unfavourable in the evaluation of stability for slopes of hydraulic-fill dams. The calculation of stability for a building period requires to take into account the pore pressure in the dain body owing to infiltration from a hydraulic filling beach, and also filtration frorn a settling basin.
FEATURES OF CHANGING OF PHYSICAL PROPERTIES OF EARTH OF A DAM BODY 3
INFORMATION ABOUT OBJECT OF CONSTRUCTION
2
The object of stationary observations of changing pore pressure is the dain of Kopetdagsky reservoir on the Karakuinsky channel (Turkmenistan), intended for irrigation of agricultural areas. The design capacity of the reservoir is 550 in1n.m’ of water, the design height is 23.5 in. The total volume of the earth body of the dain of 16 kin length is 65 rn1n.m’. The erection of the dain was carried out by the perspective and highly productive technology of dispersed of a clay ground in the body of the water development project. The dam has a spread structure. It consists of a waterproof prism (a qualitative hydraulic-fill part) and a buck waveresistant wedge (a bulk part) and it is erected froin local clayey grounds (particles of 0.05 mm up to 40 %). The hydraulic-filling of the earth into body of a dain is made by a unilateral inethod froin the side
The accepted technology of hydraulic-filling and the geological conditions of the territory have caused the difference of the structure of hydraulic-fill earth of a dam body. The site 8.2 kin length located between the stations PIS - 0 to PK - 82 is hydraulically deposited froin 5andy-loainy of grounds of the origin. Unilateral and dispersed hydraulic-filling ensures the deposition of larger sandy fractions in the prism. In the back slope where the pond will be formed a large number of clayey particles are deposited alongside with sandy particles. Accordingly, the given site of the dain is composed of l o a m and clays. The pond zone is characterised the blaky mesa-structure. The physical properties of hydraulic-fill earth have been determined by a radioisotope method. The density and humidity are the main parameters of hydraulic-fill earth quality. Because of weak water-permeability of earth and the increased water receptivity hydraulic-fill l o a m of the pond zones the humidity in them makes 20-22% even after a long sediment. The dain earth occurring lower than depressive curve reaches the even greater humidity. It practicalljs
811
does not vary here and makes 25-30 % and the degree of humidity is close to a unit. The density change in hydraulic-filling height in time in the pond zone shows that the soil ground compaction happens under the influence of a constantly increased load and its own weight. In the pond zone the thickness of dewatering is the lowest and the density is equal 1.221.36 gisin’. The density increased on 0.19-0.21 g/sm’ for different levels from 61n up to loin during a long period of sediment 5-12 months. In the lower zone. where the high degree of humidity is kept during the sediment of maps of soil ground compaction in time is observed but less intensively than in the upper zone. TECHNIQUE OF ORGANISATION OF STATIONARY OBSERVATIONS ON CHANGING PORE. PRESSURE
4
The station PK-2 1+50 was chosen during organising stationary observations on measuring the pore pressure in the hydraulic-filling dam body. Along the dam site seismic mines are representing a metal pipe of I M diameter. The seismic mine chosen for installation of measuring means was in the pond zone. where clay particles sediment when hydraulic-filled, and the process of consolidation goes slower. The lower end of the seismic mine is installed on a ferro-concrete plate of 2x2 in size placed on a contact surface of the foundation was completed. The third layer of the map of hydraulic-filling with an absolute mark of 147.0 in at total height of 7.9 m was completed by the beginning of organising stationary 0b:;ervations on the PK - 21 + 50. To install the measuring means (gauges of pore pressure and stresses) in the pipe at three levels of horizon the windows of 30x10 sin size were made on marks of 138.9; 140.9; 143.6 in respectively. Three gauges of pore pressure and one gauge of voltages were installed on each horizon. Installation of measuring means was made by the method of impressing to the depth of 1.5-2 in froin edges of the seismic mine. The measuring means gauges of pore pressure and stresses are electrical on the stress-measuring basis ha\ ing the whole meteorological certification. During the process of pressing continuous inquiry of the indications of the measuring means was carried
out. After reaching the fixed points the discrete inquiry of the indications every 15 minutes was conducted during 2 or 3 hours. The consequent indications of the measuring means were taken during a month three times a day. The further indications of the gauges were taken prior to the beginning of hydraulic-filling of the map and after the hydraulic-filling of the following layer during three years. The results were processed by the computer according to the laws of mathematical statistics with the evaluation of an error of measuring means and errors of measurement. ANALYSIS OF RESULTS OF STATIONARY OBSERVATIONS ON MEASURING PORE PRESSURE 5
According to the results of stationary observations of changes of the pore pressure during 28 months in the body of the hydraulic-fill dam the following plots were obtained: Distribution of pore pressure and total Distribution of pore pressure on the height of hydraulic filling for different periods of sediment. In Figure 1 the changes of pore pressure and total stress in time in the body of the hydraulic fill dam are represented. In 3 months after the installation of the gauges in the dam body following changes happened: 011 the whole the total stress increased by 0.006 0.008 MPa, what is explicable in the following way. In installing the gauges the condition was disturbed in these areas, for example, on horizons 2 and 3 the ground was actively extruded in thc windows was infringed, from which pressing had been made. Formats additional holes in the ground, which call for activation of the process of relaxation and results in that the stress soinewhat lower than ones natural. In time the stabilisation of the stressed condition of the mass disturbed by the installation of the gauges happens that results in some increase of the total stresses. The values of the total stresses in 3 months after the installation of the gauges practically corresponded to the natural stress condition on the considered horizons.
812
Figure 1. Changes of pore pressure and stresses The pore pressure in 3 months on all the horizons decreased practically on the saine inagnitude equal to 0.006 MPa. This decrease is connected to the bad the inoinent of the installation of the gauges the level of water in the reservoir (LWR) was lowered on the inoinent of inquiry and was on mark 138.1 in, that was aliiiost 1 ineter lower, than on horizon-3, on which our gauges were installed. This lowering resulted in dropping in the pressure at the expense of water filtration. The next inquiry of the gauges was executed in April, i.e. in 6.5 months after the previous one. For this period the next layer was hydraulically up to inark 148.6 in 1.6 in thick. The water surface was also observed the hydraulically filled. Besides on the inoinent of inquiry the level of water in the reservoir was on inark 143.1 ni and, in relation to LWR on inoinent of the previous inquiry, was lifted by 5.0 in. As the result of the changes modifications in the dam body mentioned above the following was marked: the total stresses in all the horizons were increased by one magnitude equal to 0 = 0.03 MPa, that practically completely corresponds to the load of weight of the hydraulically filled layer. The pressure of the first horizon increased by 0.03 MPa at the expense of the hydraulically filled layer 1.6 in, of the second horizon by
0.046 MPa. The pressure in the third horizon changed most essentially, it increased from 0.008 MPa to 0.096 MPa. Such a significant increase of pore pressure took place as the result of hydraulic billing of the layer of 1.6 meters. and owing to the increase of the level oi' water in the reservoir. The difference of the inarks LWR of the first horizon was 4.2 in, of the second horizon, 2.2 in and of the third, 0.5 ni. As the water surface was on the surface of the hydraulically filled map, accordingly the pore pressure increased on the horizons by the magnitude of hydrostatic pressure of water, i.e. on horizon - 1. = 0.04 MPa, on horizon - 2, 6, = 0.07 MPa and on horizon - 3, = 0.096 MPa. The next observation was executed in 2.5 months, the level of water in the reservoir being 140.0 in, i.e. it reduced by 3.1 in in a comparison with the previous level. The pore pressure of the third horizon dropped from 0.096 MPa to 0.065 MPa, i.e. it decreased by 0.031 MPa, that conipletely corresponds to the lowering LWR and says that horison-3 was in the zone of filtration pore of water the dam and in this zone the inagnitude of pore pressure is connected with LWR. In horizon-2 the pore pressure has decreased froin 0.07 MPa to 0.044 MPa. It speaks that the influence LWR has an effect on the inarks of 140.9 rn. The pore pressure decreased least of all in horizon-1 from 0.05 MPa to 0.03 MPa. what
c,,
813
ev
is connected to the process of filtration consolidation of the ground in this stratum froin the effect o i the load applied. The further indications of the gauges were taken in 3.5 months froin tile last time. The level of water in the reservoir was 138.5 in and the inap was in settling for 7.5 months. The total stress in all the horizons increased in coinparison with the previous inquiry that corresponds to the load of the over lying stratum and the density increase. The pore pressure began to drop during the setting of hydraulic filling inap. In the first horizon the intensive drop of pore pressure was observed which was P = 0.01 MPa and in comparison with the previous one on magnitude 0.053 MPa. First of all the reduction of water in the reservoir and the process of filtration consolidation in the Lone of the sandy prism explain it. The pore pressure in horizons 2 and 3 drops less intensively and inakes for the both horizons lipproximately P = 0.01 MPa. It shows that the process of filtration consolidation in these horizons goes less intensively. The next observations were made in 4.5 months, the level of water in the reservoir was 143.2 in, and i.e. 4.7 meters in comparison with the previous level increased it. The next layer was hydraulically filled to the mark of 151.2 meters for that period, the layer thickness being 1.6 in. As the result of the change in the dain bodjr mentioned above the following was marked: the total stress - 0 in all the horizons n a s increased by the magnitude equal to 0.041 MPa. practically that completely corresponds to the load of the weight of the hydraulically filled stratum The pore pressure in horizon-3 was increased to ef, = 0.054 MPa, in the second horizon the pore pressure was = 0.046 MPa and in the first horizon, = 0.035 MPa. The most essential change took place in horizon-3, as it is in the zone of the filtering pore of water through the dam. In horizon-2 the pore pressure was increased froin 0.025 MPa to 0.046 MPa. Hydraulic filling of the next layer and the consolidation of the layer cause the small increase of the pore pressure on horizon- 1 after the hydraulic filling. The last observation of changing the pore pressure and total stresses was executed in Sep-
et
tember, i.e. in 4.5 months from the previous one. For this period the map hydraulic filling was in settling for 8 months, and the level of water in the reservoir was 138.6 meters. The indications of general the total stresses - cr On all the levels were insignificantly increased, what coinpletelp corresponds to the load of the weight of hydraulically filled a stratum. The pore pressure decreased in all the horizons with no exception. In the first and second horizons the drop of the pore pressure caused only by the process of filtration consolidation of the overlying load, was: P = 0.02 MPa and 0.032 MPa, respectively in the first horizon the drop of pore pressure was P = 0.044 MPa, caused first of all by reduction of the level of water in the reservoir by 4.2 meters. In Figure 2 the distribution of the pore ofpressure and the total stresses to the height of hydraulic filling is represented. Froin the plots it is visible, that the process of dispersion of the pore of pressure in the pond zone goes on much slower, than in the zone of the retaining prism of sandy grounds (horizon-3). The dispersion of the pore of pressure in the zone of the retaining prism is caused by that it is in the zone of iiltering pore and depends only on the level of' water in the reservoir. Only the process of filtration consolidation causes the dispersion of the pore pressure in horizons 1 and 2. Also in the pond zone the dispersion of the pore pressure goes on slower, caused by the predominance of clayey particles in the site given. By the results of stationaiy observations of the character of changing the pore pressure in the body of hydraulic fill dain at different depths it is possible to inark the following: -The change of stressed is connected not onlj to the hydraulic filling of darn, but also to the increase of the density during consolidation; -The pore pressure depends both on the hydraulic filled stratum and the level of water in the reservoir; the pore pressure being increased by the magnitude equal to the difference of inarks with LWR in those zones, where goes the filtering pore of water goes through the dam; -The dispersion of the pore pressure in time in the pond zone goes on slower, than in the sandy zone of the retaining prism;
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round (circular) cylinder surface shift satisfLing the equilibrium conditions in the limit condition. Besides the strength characteristic: engagement and the angle of internal friction - 9 are accepted as constant. As the stability criterion the condition is accepted:
-The increase of the marks of the dispersion curve of the filtering pore through the dam body goes on sequentially, with the increase of marks L WR.
where: F-the resultant of the active forces or the moment of these forces in respect to the axis of the shift surface; R- is the generalised calculated value of the forces of the limit resistance to the shift on the considered surface;
Yf,Y,,,Yf,-are Y,-is
reliability indexes on a load;
the reliability indexes on a ground;
y, -is the factor of the working conditions; To search for a dangerous surface of the shift the stability factor is used. The given problem is solved in elastic-plastic statement by the finite element method in the conditions of a flat strain (in non-linear dependence between stress and strains). The results analysis show, that the pore pressure influence, greatly on the evaluation of stability. The period of filling the reservoir and hydraulic filling of the map is the most dangerous in dams. The consolidation of hydraulic fill grounds in the pond zone is considered iii conipleted during the period.
Figure 2. Distribution of pore pressure 6 CALCULATING THE STABILITY OF HYDRAULIC FILL DAMS TAKING INTO ACCOUNT OF THE PORE PRESSURE
Calculating the stability of slopes of hydraulicfill dams was made taking into account of filtration of the pond zone. Settlement case corresponded to the building period, i.e. the designed position in the period of hydraulic filling of the dam and saturation of earth grounds of the slopes with water. The process of consolidation of earth grounds of the dam body is not completed. therefore the account of the pore pressure, was obligation both in the building and operation periods. For the account of the pore pressure the condition was checked up as well:
7 CONCLUSIONS
The conducted stationary observations of changing the pore pressure in the body of hydraulic fill dams showed that in pond zone the process of consolidation in time slowly. It is caused by the dispersion of the pore pressure in the pond zone, clayey particle sediment under the accepted geotechnology. The calculation of the stability of slopes of dam body is influenced by the magnitude of the pore pressure. The period of filling the reservoir with the simultaneous hydraulic filling of the map of setting is the most dangerous for want of to evaluation of stability. The underestimation of pore pressure results in overestimating the sta-
where l) because of its large value of p,and the value of FSI tends to decrease in the lower part and high confining pressure 4 CONCLUSIONS region of the fill due to its small value of 6 N C . The overall safety therefore does not change Concluding remarks drawn from the present study remarkably as compared to that of samples 1 and U , and the circular slip surface passing through are summarized as follows. 1) FEM local factor of safety FSI takes a large deeper portion near the base of the embankment value in the slope surface where OC condition can be critical. It should be noted as in this case develops by roller compaction, and it decreases in that extension of the region of OC stress state does inner deep portion and shows concentration of low not have a direct correlation with increasing overall safety value near the boundary between the OC and safety of embankment. NC conditions. 2) In the analysis of NC-state slope, disregarding pre-compression effects of soil, Fsr entirely becomes lower especially near the slope surface and suggests a higher potential of shallow surface sliding. This situation can arise in cases where embankment becomes wet by impounding of the reservoir or by a rainfall, because pre-compression effects disappear due to saturation. 3) Safety evaluation by FSLdoes not reflect sharply the difference in material properties as compared to that by deformation. The simplified Bishop method of analysis incorporating pre-compression effects, on the other hand, gives an equivalent result to the latter and can be an effective measure for the construction control.
m
m
REFERENCES Lee,K., Ohne,Y., Narita,K. & Okumura,T. 1994. On strength characteristics of cohesive soils having pre-compression effects, AIT technical report, 29B: 69-78. (in Japanese) Mitachi,T. 1976. Influence of stress history on triaxial compression tests of cohesive soils, 20th symposium of JGS: 71-78. (in Japanese)
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Slope Stability Engineering, Yagi, Yamagami& Jiang @! 1999 Balkema, Rotterdam, ISBN 905809 079 5
Earth pressure acting on the side of core block in high embankment K. Nomoto & T. Su,'oimoto Overhectd Transmissionancl Substations Corzstruc'tionOfjce, Tokyo Elestric Pourer Compcrny Incorporated, Jupcin
T. Fujiwara Technology Reseurch Center, TuiseiCorporcitioiz, Yokohanm,Jupan
ABSTRACT: In a large-scale embankment project, 1.3 million m3in volume and 43 m high at its highest, core blocks (about 100,000 m7in volume) were constructed inside the high embankment slope to improve its stability. The core blocks were made of cement-stabilized weathered pyroclastic deposits. In order to monitor the stability of the embankment, made up of volcanic cohesive soil, various types of measuring equipment were installed inside the ground and the core blocks. In addition, large panel-type earth pressure gauges were installed on the sides of the core blocks. First, the rigidity of core blocks relative to that of the embankment is discussed on the basis of measurements made by an inclinometer installed inside the core blocks. Second, measurements by a settlement gauge installed inside the embankment were used to verify the settlement is at a standstill thereby confirming the stability of the entire embankment. The long-term external force acting on the relatively rigid core blocks inside the embankment was then used in the calculation of the coefficients of earth pressure. The values of kv and kh obtained in this manner were kv = 0.4 and kh = 0.4.
1 INTRODUCTION This large-scale embankment project, 1.3 million m3 in volume and 43 m high at its highest, was implemented on the northeastern slope of Mt. Akagi at an elevation of around TP+1,100 m as preparatory work for the construction of a 1,000 kV substation (Nomoto and Tsunoda, 1996). The significant feature of this project was that the "embankment zoning" method was employed,in which the excavated soils at the construction site were classified according to their quality and subsequently used for specific purposes. To improve the stability of the high embankment slope, core blocks (about 100,000 m3 in volume) made of cement-stabilized weathered pyroclastic deposits were constructed inside the slope (Nomoto et al, 1996, Yoshida et al, 1998, Ogasawara et al , 1998). In order to monitor the stability of the cohesive volcanic soil embankment, various measurements of the ground and core blocks were conducted. In addition, special large panel-type earth pressure gauges were installed on the sides of the core blocks. This paper presents the method and results of the large panel-type earth pressure gauge measurement. On the other hand, the measurements by an inclinometer installed inside the core blocks revealed the core blocks had much higher rigidity relative to that of the embankment. In addition, the measurements by a settlementgauge installed inside the embankment confirmed settlement became constant, thereby indicating that the whole embankment
had become stable. On the basis of these observations, the long-termexternal force acting on the relatively more rigid core blocks inside the embankment was used in calculating the coefficientsof earth pressure. The results of calculation are presented below. 2 EMBANKMENT AND MEASUREMENT Figs.1 and 2 show a plan and a typical profile of the embankment which was filled mostly with cohesive volcanic soil. As shown in Fig. 2 the elevation of the top of the embankment was TP+ 1,100 m. The embankment was constructed with a slope of 1:2 (about 27 degrees) and at every 5 m in height there was a 2 m or 5 m wide horizontal step. Core blocks installed at the middle of the slope were made of cement-stabilized weathered pyroclastic deposits by mixing ordinary Portland cement whose weight was equivalent to about 6 % of dry weight of the treated soil (Fujiwara et al, 1998). The ground supporting the core blocks consisted of high quality pyroclastic deposits. The core block shape was decided slope stability analysis; the gradient of the excavated ground was set at I: 1 (45 degrees), the uphill gradient of the ground in contact with the embankment at 1:0.6(about 59 degrees), and the downhill gradient at 1:1. The top of the core blocks was made horizontal and its elevation was TP+1,085 m. Table 1 shows the wet densities and strengths of the volcanic cohesive soil and cementstabilized soil used in the construction of the embankment.
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An insertion inclinometer was placed at the center of the core blocks for better understanding of the effects of uphill embankment on the core blocks by measuring the displacements in both parallel and orthogonal directions to the slope. The earth pressure acting on the core blocks was measured by two panel-type earth pressure gauges installed on the uphill slopeof the core blocks at elevations TP+1,073 m and TP+1,078 m, hereinafter referred to as"EPG1" and "EPG2",respectively (Fig.3). The paneltype earth pressure gauges measured "normal stress" acting in a direction orthogonal to the uphill slope, and also "shear stress"acting in a direction parallel to the slope.
3 EARTH PRESSURE GAUGE INSTALLED ON THE SIDE OF CORE BLOCKS
Fig.4 Structure of a panel-type earth pressure gauge. A settlement gauge, an inclinometer, and earth pressure gauges were installed as shown in Fig.2 to monitor the stability of the embankment. A differential settlement gauge was installed at the top of the embankment,where the embankment was the thickest, in order to investigate the settlement characteristics of the volcanic cohesivesoil.
3.1 Spec@cationsof the panel-type earth pressure gauge Fig. 4 shows a structure of a panel-type earth pressure gauge. The surface of the pressure gauge was made of steel(SS400) and measured 1.O m x 0.5 m. In order to help the transmission of shear strength acting on the uphill
842
3.2 Results of earth pressure measurements Figs. 5 and 6 show time histories of the normal and shear stresses measured by EPGl and EPG2 respectively with the embankment height at corresponding time. Four months after the start of measurement, EPGl was no longer able to measure shear stress, and also normal stress after 15 months. However, the values of normal stress were obtained up to the time of completion of the uphll embankment. The measured normal stress increased in accordance with the progress in filling and reached a maximum value of 227 kPa. Measurements of normal stress and shear stresses by EPG2 lasted for the period of about 2 years. Both normal stress and shear stresses increased with the progress in filling and reached maximum values of 186 kPa and 62 kPa, respectively. To the completion of the embankment they became constant at values of 183 kPa, and 53 kPa in average for the n o d stress and shear stress,respectively. Judging from Figs. 5 and 6 which clearly show that all stress measurementschanged with the progress in filling, it is concluded that the field measurements were carried out in an appropriate manner. The values of all stress have been constant for about 1 year since the completion of the embankment, therefore, it is deduced that the embankment slope is stable. 4 RESULTS OF MEASUREMENTS BY INCLINOMETER AND SETTLEMENT GAUGE.
side, sand particles were glued to the surface of the pressure gauge. Stress acting on the surface of the pressure gauge was measured by straingauge-type load cells, three of which were used to measure normal stress and two for shear stress. The design load was considered 167 kN in the normal direction to the surface of the pressure gauge taking into account the maximum overburden. As shown in Fig. 3 and Photo. 1, eight dummy panels were placed around an earth pressure gauge in order to prevent stress concentration on the pressure gauge.
Figs. 7 and 8 show the distribution of the horizontal displacement at the section of the core blocks and the change with time in the horizontal displacement of the core blocks, respectively. The measurement by an inclinometer placed inside the core blocks showed that at the initial stage of 8 months after the start of filling of uphill side, horizontal displacement of approximately23 mm occurred towards uphill side at the elevation of about TP+l,085m, while the point at TP+1,067m remained still. This relatively large horizontal displacement is considered to be the course of contact between the ground and core blocks. In the uphill side of the core blocks, the increase in displacementat the period of 8 to 14 months, was only 4 mm at the top of the core blocks, and was no more than 2 mm after the completion of the embankment (after 14 to 25 months), thereby indicating the stability of the core blocks. Also, the core blocks constructed using cementstabilized soil was considered to be relatively rigid compared to the uphill side embankment made of volcanic cohesive soil. Fig. 9 shows the time history of embankment settlement made of volcanic cohesive soil. The accumulated settlement from the start of filling was about 26 cm at the foundation level of TP+l,074m, 98 cm at the top of the core blocks at an elevation of TP+1,085 m, and 120 cm at the top of the embankment at an elevation of TP+ 1,100 m. The settlement of the embankment crest immediately
843
after the completion of filling was about 14 mrn a month, nevertheless present settlement is about 1 rnm a month, showing that the settlement almost stopped. This fact suggests that the embankment has become stable and static earth pressure is supposed to act on the side of the core blocks.
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Fig. 11 Relationship between load of filling q and earth pressure.
5 COEFFICIENTS OF EARTH PRESSURE AND THEIR EVALUATION 5.1 Relationshipbetween load ofJillingand earthpressure Fig. 1 1shows the relationship between load by filling q and vertical earth pressure o ",horizontal earth pressure (r h. Based on measurements by the earth pressure gauges, the following relationships between q in the uphill side (r ", (T ,, were obtained. Vertical earth pressure o = 0.45q and horizontal earth pressure (T = 0.40q. Both these relationships show linear relation which indicate that the panel-type earth pressure gauges used in the project were successful in monitoring the changes in the load of filling. In the calculation of load of filling, the value obtained by multiplying the filling height by the wet density of the embankment material ( p [ = 1.55 g/ cm3) was used. The wet density of the embankment material used in the calculation was a mean value obtained from quality control tests on embankment materials carried out during the construction. The equations used to evaluate the vertical and horizontal earth pressure are as follows (Fig.10). Vertical earth pressure: o = m X sin 6 (kPa) Horizontal earth pressure: o = m X cos 0 (kPa) where, m is the measured resultant earth pressure with components p (in the direction perpendicular to the slope) and s (in the direction parallel to the slope):
6 is the angle between resultant force m and the horizontal line.
6=
cy
+ tan-'(s/p)
and cy is the angle of the slope of the embankment measured from the vertical line. cy
values of kv = 0.4 and kh = 0.4 were obtained. The coefficientof vertical earth pressure kv, which had remained at 0.55 during filling, dropped rapidly to 0.4 just before the completion of filling (9 months after the installation of the earth pressure gauges). This could be explained by the release of embankment stress occurred around the side slope of the core blocks as a result of the progressive increase in the load of filling. The constant values of kv = 0.4 and kh = 0.4 were continuously observed from the time when the settlement was relatively large immediately after the completion of filling to the time when the settlement had almost converged 1 year after the completion. This shows that the coefficient of earth pressure was not affected by the redistribution of stress in the ground which occurred with the settlement of the embankment.
6 CONCLUSION
= 3 1 (degrees )
(for a slope gradient = 1:0.6)
5.2 Change in coeficients of earth pressure with time Fig. 12 shows the change in the calculated coefficientsof earth pressure with time. The ratio of vertical earth pressure to the filling load, kv = o v/q (coefficient of vertical earth pressure), and the ratio of horizontal earth pressure to the filling load, kh = o h/q (coefficient of horizontal earth pressure), were 0.55 and 0.45, respectively, though the measured values fluctuated during the initial stage of filling, when the overburden as measured by the earth pressure gauges was small. However, after the completion of filling the constant 845
In this project, large panel-type earth pressure gauges were installed on the sides of the core blocks for the purpose of conductinglong term measurements of earth pressure. This paper is intended to serve as a guide for future in situ evaluations of earth pressure when designing structures constructed in embankment. ACKNOWLEDGMENT The authors would like to thank Dr. Masami Fukuoka, Emeritus Professor at University of Tokyo, for many helpful suggestions through this measurement project.
REFERENCES Yoshida, M., Sugimoto,T., Ichibayashi, Y. andTanizawa, F. 1998. Earth pressure acting on the core block in high earth fill - Measuring results by large scale panel type earth pressure cells -, Proceedings of the 33rd Annual Meeting of the Japan Geotechnical Society: 1677-1678 ( in Japanese). Ogasawara, K., Fujiwara, T., Yoshida, M. and Tanaka, M. 1998. Earth pressure acting on the core block in high earth fill - Evaluation of the measuring result -, Proceedings of the 33rd Annual Meeting of the Japan Geotechnical Society: 1679-1680 ( in Japanese). Nomoto, K., Sugimoto, T., Tanizawa, F. and Ogasawara, K. 1996. Shear Strength of the Boundary Surface of Cement-stabilizedCompacted Soil, Proceedings of the 3 1st Annual Meeting of the Japan Geotechnical Society: 189-190 ( in Japanese). Nomoto, K. and Tsunoda, S., 1996.Design and Execution of Foundation Work of Higashi-Gunma Substation. Electric Power Civil Engineering: 58-62 ( in Japanese). Fujiwara, T., Tanizawa, F., Nomoto, K. and Sugimoto, T. 1998. Construction of cement stabilized core block in high embankment made of volcanic cohesive soil. Proc. of IS-Tohoku98; 199-202.
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Slope Stability Engineering, Yagi, Yamagami & Jiang (c) 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Case study of a liquefiable mine tailing sand deposit W.Wehr Keller Grundbau GmbH, Overseas Division, Offenbnch, Germany
I. Herle, I? Kudella & G.Gudehus Institute for Soil and Rock Mechnics, University of Karlsruhe, Germany
ABSTRACT The slope stability of a loose sandy deposit in a lignite mining area in Germany is investigated. Increasing the slope stability, the slope has been flattened and compacted by blasting. In situ tests have been executed to find out if different compaction methods including blasting have been successful. Selected data from in situ and laboratory tests serve as input for the stability calculation. Liquefaction is not treated as limit equilibrium problem but as a stability one. The stability criterium is the excess kinetic energy after Hill. Input parameters for the hypoplastic constitutivc equation are determined with simple (index) laboratory tests. These soil parameters are verified through the recalculation of triaxial undrained tests. A representative cross section of the slope including the "hidden dam", which is a certain soil volume compacted by blasting, has been chosen to perform the calculations. Varying the input parameters, the influence of different soil and geometric parameters is shown.
I
The result decides whether the areas can be opened for the public.
INTRODUCTION
Soil liquefaction represents onc of the most challenging tasks of modern soil mechanics. Catastrophic events involving soil liquefaction are not only the reason of large material damage but they also take a toll of human lives repeatedly. One can distinguish between the liquefaction triggered by a rapid cyclic deformation (e.g, earthquake) and the spontaneous case. The latter may be more dangerous as it appears that no pre-cursors occur prior to it, and its mechanism and control is still rather unclear. The lignite mining activity in the Lausitz region in Germany has left a large area of loose sand deposits. Prior to mining the groundwater level was lowered. Subsequently mine pits were opened to depths of about 50 m. Pleistocenc sands, which covcr the lignite layers, were continuously deposited in already exploited parts of such pits. However, due to thc excavated lignite the pits could not be totally refilled. After rising the groundwater to the original level some lakes with loose sand embankments arose. These slopes are often subjected to spontaneous liquefaction. Fatal accidents and enormous losses of surface area are reported. In order to prevent this liquefaction the sand is densified by various methods (blasting, falling weight, vibro-compaction) (Raju et al. 1994). The stability of embankments prior and after compaction must bc estimated. This can be done using empirical formulas (Vogt et al. 1991) or a stability analysis.
2 STABILITY CRITERION In case of embankments a limit equilibrium analysis of the slope is often performed. A simple analysis of an infinite drained embankment of inclination I9 with a plane slip surface parallel to the slope surface yields a limit equilibrium for tan 0 = tan 9'without presence of water and tan0 = tany'/2 under the water with seepage parallel to the slope, respectively (9'denotes the effective friction angle). Depending on pressure and density, y' of sand can vary between ca. 30" and 45", i.e. slopes with I9 < 30" (above water) and I9 < 15" (under water), rcspectively, should be stable. However, spontaneous liquefaction of slopes of 0 < 10" has been reported (Foerster et al. 1986). One may object that rapid movements during liquefaction prevents the drainage of sand and consequently "undrained strength parameters" instead of 9'should be used. Figure 1 shows two commonly used procedures for the determination of strength parameters from undrained triaxial tests. The first procedure (Poulos et. al. 1985) adopts only the "undrained cohesion" c, = (01 - 02)/2 from the steady-state line (ss in Figure 1) for a given void ratio. The second one (Sladen et al. 1985) defines a friction angle yu and cohesion c,, from a so-called collapse surface (cs in
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fS /
cs
/
O1'
O2
Figure 1: Schematic representation of
effective ,qre,qspath during an undrained triaja]test with loose sand, steady state (ss) and collapse surface (cs). Fiigure 1). In this way, very low values of 9,and czL, rcspectively can be obtained from triaxial tests (Foerster et al. 1986). Many stability calculations of this kind can be found in the literature (Ishihara 1993, Sladen et al. 1985). However, some authors pointed out that thcse methods cannot capture the liquefaction problem (Gudehus 1993, Lade 1993). The main argument against such calculations is a lack of the physical background. The "undrained" shear resistance implies excess pore water pressures that dcvelop during deformation. But prior to the deformation thcre is no excess pore water pressure in thc sand. Thercforc, the equilibrium theory cannot predict whether the embankment is stable or not. It can at best give a crude estimation of the shearing resistance of the sliding slope when the excess pore pressures are fully developed. The assumption of a localized narrow shear zone or slip surface is also cluestionable. The word liquefaction suggests an analogy with melting of solids, which is rather superficial however. The transition of a saturated grain skeleton to a suspension is also different from plastic flow (or shear melting) so that usual concepts of soil plasticity become meaningless. Shear localization to narrow zones (slip surfaces) cannot be presumed as such skeletons are collapsible (contractant) and not dilatant. A coincidence between the results of the calculation and the in situ observations can be regarded as incidental. A novel approach (Gudehus 1993,1998) makes use of the excess rate of hnetic energy. If after a small perturbation the increase in internal energy exceeds the work of the external forces, the soil remains stable. Otherwise any perturbation will initiate an instability, ~ i.e. a release ofkinetic energy from the system. H stability criterion (Hill 1958) as stability postulate is used for this purpose:
848
The time derivative of the first Piola-Kirchoff stress tensor it, involves the initial configuration as a reference, v3 denotes the velocity field during the perturbation. S2E < 0 implies stable behaviour. Using the integral of the volume the whole soil which moves during failure of the embankment is taken into account. Therefore it is possible to distinguish between liquefied and non-liquefied areas. For example the soil may not be liquefied over the total volume but only over a small area leading nethertheless to an excess kinetic energy. Equation (1) requires a velocity field and the corresponding stress rate. This rate can be obtained from the velocity field using realistic constitutive relations. suitable model Hypoplasticity has turned out as a for this purpose. 3 HYPOPLASTIC CONSTITUTIVE MODEL Hypoplastic constitutive relation is a non-linear tensorial equation that yields a co-rotated stress rate b,, as a function of the granular (effective) Cauchy stress CT:,~, the deformation rate d,, and the void ratio e:
The behaviour of so-called simple granular skeletons without macrovoids is considered in a large range of pressures and densities. The behaviour of sand including macrovoids has been investigated experimentally and theoretically by Herle et al. 1998. A detailed representation of .f in equation (2) can be found elsewhere (Bauer 1946, Gudehus 1996, Wolffersdorff 1996). The deformation rate d,, is defincd as the symmetric part of the velocity gradient. There exists a simple relation between tr,,? and .527 (Hill 1958):
(3) Fable 1: Hypoplasticparameters o f Lausitz sand.
eio
n
a.
13
1.00
0.20
0.25
1.00
constants are needed for the hypoEight plastic Their values for Lausitz sand, which ~ ' equation. ~ is prone to liquefaction, can be found in Table 1. The determination of these constants is straightforward:
the critical friction angle pc can be obtained as the angle of repose; e,o and edo are the critical and minimum (through cyclic shearing) void ratios at zero pressure and they correspond approximately to emaz and enzinfrom standard tests where the pressure is not exactly zero (p,=O is used as a reference pressure); the maximum void ratio in an isotropic state at zero pressure e;o can be estimated as 1.2e,o from idealised grain skeletons consisting of spheres or cubes; the granulate hardness h, and the exponent n can be calculated from the oedometric compression curve with a loose specimen which can be approximated by
ted to the granulometric properties of sand, namely to the mean grain diameter, uniformity coefficient, grain shape and grain mineral. In order to verify the material constants, oedometer and triaxial laboratory tests on reconstituted sand samples and numerical calculations were performed and compared to each other, The agreement is satisfactory (Fig. 2). The constitutive model captures all important features of the behaviour depending on pressure and density.
4 STABILITY CALCULATION (4) (eo is the void ratio at the mean pressure p , = 0); and the exponents cy and /3 can be determined from the peak friction angle and the compression coefficient, respectively, for a dense specimen. It has been shown (Herle 1997) that all these parameters are closely rela-
For a rapid movement it is assumed that the deformation is undrained, i.e. in the liquefied zone the volume is constant. As a first approximation, a mechanism with uniform stretching rate in a triangular zone is proposed (Fig. 3). If the stress is considered only in the centre of the triangle, equation (1) reduces to
The excess kinetic energy becomes positive, indicating instability, if the void ratio exceeds a critical value, which depends on the stress ratio and is in the vicinity of Casagrande’s critical void ratio. The cornputations with this model (Raju 1994) sufficed to find a realistic maximum void ratio of a stable slope.
Figure 3: A velocity field for the stability calculation. Further improvement was achieved assuming various stress states at different points of the ”liquefied triangle” and a partial saturation. The assumption of constant volume is dropped; gas bubbles enclosed in the grain skeleton contribute to the production of kinetic energy when they expand. With intergranular , pressure po and total pressure pressure C ( ~porewater ( T , ~ = a:J - p 0 S z 3 the stability criterion now reads (Kudella 1995):
h2E = -
1,
-
po
)(“‘e ( S -
1) d z k
Computations show that a large amount of gas reduces the danger of liquefaction; however, small gas bubbles can enhance a spontaneous production of kinetic energy.
Figure 2: Measured (top) and calculated (bottom) stress paths in undrained triaxial tests with Lausitz sand at various pressures and densities. 849
The excess kinetic energy has to be calculated by integration of equation (6). Instability has to be presumed, if the integral over the whole deformed soil mass becomes positive. Of course, the assumed triangular deformation zone is only one mechanism of many kinematically possible ones. Therefore, the integral does not yield necessarily a lower bound for stability. Other mechanisms, however, will be analysed in further studies.
usually correlated with the results of field tests through theoretical or empirical relationships. Freeze probing was used as a direct method to determine the in-situ void ratio and degree of saturation of granular materials. In our case a continuous freeze probing profile down to a depth of -18 m allowed a detailed identification of the stratigraphy. A large frozen volume had to be sampled in order to obtain reliable values of the void ratio and degree of saturation (Wehr et al. 1995). Unfortunately, this procedure is expensive, time consuming and complicated. Indirect methods, like measurements of shear wave propagation, are useful if a large area has to be investigated. CPT can be an appropriate alternative if an adequate correlation between the measured soil response and the void ratio is available (Wehr et al. 1995).
5 CASESTUDY We are reporting a case of an artificial lake after flooding which was created in 1962. The soil volume that had been removed to excavate the lignite could not be completely refilled. Open parts of the pit remained, resulting in the formation of slopes and lakes. After the pumps for the groundwater lowering were turned off, the water level rose slowly and first spontaneous liquefactions could be observed in 1964. The rise of the groundwater table will be finished in 2030. In order to increase its stability the slope was flattened in 1977. Compactions by blasting were carried out from 1984 to 1986 and 1992. By that means a compacted body of soil was created in the slope. This "hidden dam" was densified parallel to the slope and the later shoreline in 2030. The function of the dam is to obstruct the undensified deposit behind the dam from flowing into the lake if a liquefaction takes place. In addition the surface of the slope was compacted with vibratory rollers.
5.3 Input parameters f o r calculation After the determination of the soil constants from laboratory tests, and the determination of the in situ state parameters from field measurements, stability calculations have been performed.
5.1 Structure of the deposit
Figure 4: Representative cross section of the slope.
The deposit consists of a 27 m thick refilled open mine. It is composed of very loose sand layers with a fines content due to lignite and silt variing between 0% and 15%. A large amount of macrovoids filled with air is embedded in the grain skeleton resulting in a unusual low degree of saturation. After flooding, the sand below the water table typically has a degree of saturation S between 80 and 90 % due to field experience. Soil compaction in the lower 20 m was achieved by blasting, and in the upper 3 m by vibratory rollers. The original density between -3 m and -7 m depth could not be changed by the compaction methods used.
From the geometric parameters of the slope and the "hidden dam" a representative cross section of the slope for the final water level in 2030 has been chosen (Fig. 4, Tab. 2). Table 2: Input parameters for the stability calculation of the representative cross-section.
5.2 In situ tests
1
Accurate description of the state of soil plays a decisive role in predicting spontaneous liquefaction. The in situ state of granular soils can be defined by the following state variables: relative density, degree of saturation and skeleton stress components. These variables can be measured directly with a great effort only. Therefore, in engineering practice they are 850
cross section of slope height of "dam" height of water level "dam" co-ordinate left "dam" co-ordinate right max.slope angle void ratio in deposit void ratio in "dam" void ratio in front of "dam" degree of saturation
1 35.90 [m] 19.80 [ml 55 13 _Iml_ ['I 6.3
[ml
11 11 w
.c 1 .E,.
I
PnlQZ
[-I
e771CCZ,dLp ~
FQZ',dCCTX
emQc,wQl
~
[-1
j
0.87
0.;: 20.84
Being on the safe side, the maximum measured values of void ratios and slope angles have been taken. For the degree of saturation an average value has been used.
a sufficient soil volume with excess energy has to be involved. The most efficient position of the "hidden dam" is chosen if its position coincides with the maximum of the excess kinetic energy for the case calculated without "hidden dam".
5.4 Program 'Stabil' A computer program Stabil has been developed by Raju 1994. The original computations were based on the assumption of an active Rankine earth pressure field in an infinite slope. As this does not properly represent the flat top of the slope for high water tables, a smooth transition between Rankine stresses in the slope and the active earth pressure in an infinite horizontal plane was modelled using transition functions (Kudella 1995). This procedure can only be regarded as a rough estimate, as the real stress distribution in and behind the slope is unknown. It may in fact be quite different from natural deposits and does, to some extent, still reflect the dumping process by "conservation" of shear stresses in rather arbitrary directions. The unsaturated soil above the groundwater table was assumed to follow the deformation of the underlying soil as a dead load without absorption of energy. Further improvements as SB
SB
modelling of an area of reduced void ratio (the "hidden dam") partial saturation and influence of gas bubbles intcgration by using more than one point graphical output with distribution of the excess kinetic energy (Fig. 5 )
have been added to the computer program by Kudella 1995.
Figure 5: Spatial distribution of the excess kinetic energy (isociirones)in a representative cross section of the slope (integrated excess kinetic energy -1.16
J/(m3s2))
5.5 Results
The following calculation is executed with the representative cross section. Both angles B and I/ (Fig. 3) are varied by the program until a maximum of the excess kinetic energy is found. For the final geometry and water table in 2030 no liquefaction has been calculated due to the excess kinetic energy (see Tab. 3). Table 3: Results of the stability calculation. Cross section excess kinetic energy angle of investigated triangle angle of volume change
1 29
["I
Additional to the above calculations which were on the save side, the slope geometry and the field parameters have been varied to investigate their influence. Each time only one parameter has been changed, keeping all the other ones constant. The slope angle /3 plays a decisive role in the analysis. Before the slope was flattened in 1977 and during blasting in 1985 and 1986, slope angles were between 30 and 35 degrees. The results of the stability calculation show cross sections without a "hidden dam" stable up to /3 = 18", whereas p 2 26" is stable with a "hidden dam". Thus observed liquefaction events before 1977 and in 1985/86 with 30" < @ < 35" could be explained. Varying the height of the water table up to 35 m, which corresponds to the total height of the deposit, no liquefaction due to excess kinetic energy was calculated. Finally the field parameters are varied. The void ratio e=0.87 in the deposit has been varied, taking e=0.76 in the "hidden dam". Evaluating the frozen specimens, void ratios of up to e=0.97 in another cross section were found. Excess kinetic energy was obtained in calculations for e 21.4. However, such high void ratios for sand under the water table have not been measured. The average degree of saturation S was evaluated to be 80% with limits between 72% and 92%. If the calculation is performed with S= loo%, the excess kinetic energy is twice as high as in the latter case. A strong influence of S on liquefaction can be seen, especially if the slope angle is small (Kudella 1995). The variation of S yields no excess kinetic energy.
The spatial distribution of the excess kinetic energy reveals that liquefaction starts inside the soil mass (maximum in Fig. 5). Loosing the overall stability, 851
6 CONCLUSIONS
Herle, I. (1997). Hypoplastitzitat und Granulometrie nichtbindiger Granulate. Publications of the Institute of Soil and Rock Mechanics, Karlsruhe University, 141: Herle, I., Wehr, W., Gudehus, G. (1998). Influence of macrovoids on sand behaviour. 2. Int. Con.. on Unsaturated Soils, Beijing, 60-65 Hill, R. (1958). A general theory of uniqueness and stability in elastic-plastic solids. Journal of the Mechanics and Physics of Solids, 6:236-249 Ishihara, K. (1993). Liquefaction and flow failure during earthquakes. Geotechnique, 43(3): 35 1415 Kudella, P. (1995). Stabilitatsberechnung von setzungsfliefigefahrdeten Kippenrandboschungen. Geotechnik, 18(2): 7-15 Lade, P. V. (1993). Initiation of static instability in the submarine Nerlerk berm. Canadian Geotechnical Journal, 30: 895-904 LiPoulos, S. J., Castro, G., France, J.W. (1985). quefaction evaluation procedures. J. Geotechnical Eng. Div., ASCE, 111(6): 772-792 Raju, V. (1994). Spontane Verflussigung lockerer granularer Korper - Phanomene, Ursachen, Vermeidung. Publications of the Institute of Soil and Rock Mechanics, Karlsruhe University, 134: Raju, V., Gudehus G. (1994). Compaction of 10ose sand deposits using blasting. Proc. XZII ICSMFE, New Dehli, 1145-1150 Sladen, J., Hollander, D., Krahn, J. (1985). Thc Liquefaction of sands, a collapse surface approach. Canadian Geot. Journal, 22: 564-578 Abschatzung der Vogt, A., Forster, W. (1991). Riickgriffweite von Setzungsflieh-utschungen. Neue Bergbautechnik, 21( 10/11): 366-371 Wolffersdorff von, P.-A. (1996). A hypoplastic relation for granular materials with a predefined limit state surface. Mechanics of Cohesive-Frictional Materials, 1: 25 1-271 Wehr, W., Cudmani, R., Stein, U., Bosinger, E. (1995). CPT, shear wave propagation and freeze probing to estimate the void ratio of loose sands. CPT’9.5, Int. Symp. on Cone penetration testing, Linkoping, Sweden, 2: 35 1-356
Many authors are using the ”undrained strength parameters” to calculate the liauefaction riqk nf wRtw saturated slopes. However, such a calculation is not supported by a physical judgement. It implies excess pore water pressures that develop during deformation but there is no excess pore water pressure in the sand prior to the deformation. Therefore a coincidence between calculation results and in situ observations can be regarded as incidental. A novel approach makes use of the excess rate of kinetic energy. It is based on Hill’s stability criterion which requires a velocity field and a corresponding stress rate. The latter can be obtained realistically using hypoplastic constitutive equations with only eight material constants. The determination of the constants is straightforward, because they are closely related to granulometric properties of sand. Liquefaction of a triangular zone has been assumed duc to in situ observations. Various stress states at different points, different void ratios in densified areas and partial saturation with gas bubbles can be taken into consideration. Computations show that a large amount of gas reduces the danger of liquefaction: however, small gas bubbles can enhance a spontaneous production of kinetic energy. A case study of a loose sandy deposit in a lignite mining area is presented. The in situ state of the sand has been determined using extensive site investigation methods including freeze probing. Back calculations of a representative cross section of the slope yield negative excess kinetic energy (stable) for the actual state, but positive excess kinetic energy for the states in 1977 and 1985/86 where the slope angle has been between 30 and 35 degrees. Varying the geometric and field parameters, their limit values for the embankment without liquefaction due to excess kinetic energy are obtained. REFERENCES Bauer, E. (1996). Calibration of a comprehensive hypoplastic model for granular materials. Soils and foundations, 36(1): 13-26 SetForster, W., Walde, M., Dierichs, D. (1986). 8. zungsflieBen im Braunkohlenbergbau. Donau-Europaische Konferenz iiber Bodenmechanik und Grundbau, Niirnberg, 263-269 Gudehus, G. (1993). Spontaneous liquefaction of saturated granular bodies. Modern approaches to plasticity, Kolymbas D. (ed.), Elsevier, 691-714 Gudehus, G. (1996). A comprehensive constitutive equation for granular materials. Soils and Foundations, 36(1), 1-12 Gudehus, G, (1998). On the onset of avalanches in flooded loose sand. Phil. Tmns. R. Soc. London, 356: 2747-2761 852
Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Bilinear model for stability calculation of domestic waste landfills G.Ziehmann Department of Waste Management, Technical University of Braunschweig, Germany
ABSTRACT: This paper presents a new bilinear bearing model for the strength of municipal solid waste (MSW). The bilinear bearing model includes the two different parts of strength of MSW which are the shearand the tensile strength. Part one of this paper explains the new model and shows which experimental equipment is necessary to obtain the different parameters of MSW strength. In part two the results of stability deterinination (shear and tensile strength) are established through the test apparatus presented above. These are compared with shear strength from one- and triaxial tests adapted to MSW. Differences of slope stability calculation according to the different experimental methods are presented within an example.
1 INTRODUCTION
z, = z + t
A variety of different problems occur when calculating MS W landfill stability based on common methods of soil mechanics. These problems are mainly based on two different points due to the heterogenity of MS W: 1. The size of experimental equipment in soil meclianic laboratories is usually in comparison to the size of soil pieces. Normally the lateral length does not exceed 63 mm, where MSW pieces are often more than 1000 inin in length. Because of this, it is necessary to use large-sized experimental equipment. 2. The strength of soil is normally established as shear strength. Due to the size of the pieces included in MSW there are many pieces with one or two longer sides (fibres and foils). These pieces are able to activate a tensile force. To get an exact view of the strength of MSW, it is necessary to determine the tensile strength as well as the shear strength. Due to the fore-mentioned points it was important to develop a model describing the strength of MSW, that determines realistic strength parameters and obtains an exact base for stability calculations.
with: zt =total shear strength z = shear strength t = tensile strength
(1)
The bilinear strength model shown in Figure 1 and 2 was developed, based on knowledge about the bearing behaviour of fibre containing materials like reinforced soil (Jewel1 & Wroth 1987). The tensile strength of fibres and foils can only be activated if the fibres or foils are tightened and anchored on at least two sides. A deformation is than necessary before the tensile strength of MSW is activated. The amount of necessary deformation is a function of MS W compound. This potential interaction between friction and tension forces during deformation is shown in Figure 1 in accordance with constant normal stress. Friction is activated from the beginning of deformation (I). The tension increases due to the number of fibres which are tightened and it's Young's modulus. As demonstrated above it starts at the deformation when the first fibre is tightened (II). After the maximum tensile strength of each fibre is reached or the anchoring strength of the fibres is exceeded the tension is reduced (ILI). Only friction occurs when the deformation increases hrtlier than the point where the tension is reduced to zero (JY). The tensile strength is additional to the deformation also due to the normal stress. If it is considered that the fibres are able to transmit the
2 BILINEAR MODEL OF STRENGTH FOR MS W 2.1 Theoretical foundations
Due to the activation of MSW tensile strength the total shear strength is divided into two different kinds of strength, shear and tensile strength. The general valid relationship is shown in equation (1).
853
anisotropy of MSW sufficiently. Additionally, the separate testing establishes better inforniation about tensile-stress-deformation characteristics. Due to the size of the pieces of MSW a sample box with dimensions of 3 x 1 x 1.5 m and a contend of about 4 m3 was used. The box is filled with layers 20 to 30 cm thick and compacted by a loading plate mounted on an hydraulic excavator. The box has to be moveable, so that it can be transported by a truck, to the place where it is filled with MSW. The normal stress is induced by high pressure pads laying between the load plates and the load gridder. The sample is consolidated by geotechnical recommendations after the normal stress is induced. The consolidation time is normally one to two weeks for each load increment. The test is carried out with at least three different normal loads. The maximum normal load is limited to 500 kN/in2. For the tensile test the sample box is vertically opened in the middle. The front part of the the box is connected to the hydraulic power cylinder, the back part is anchored. During the execution of the tensile test the front part of the box is pulled out, so that in the middle of the sample tensile strength is activated. The pull out load (= tensile strength) is raised path controlled up to the breaking point of the sample. A direct shear test with almost the same size is conducted according to the size of MSW pieces and to the tensile test. The shear surface is 2 m2. The test procedure does not differ from the test in smaller direct shear apparatus. The apparatus used for the tensile test can also be used for the shear test, only the box has to be changed. Before conducting the tests a sample of nearly 300 kg is classified into material groups and by dimensions. Dimensions 1 and 2 (larger than 40 mm and folio or fibre form) are the most important
Figure 1. Model of the interaction between friction and tensile forces
Figure 2 . Bilinear model of shear and tensile strength of MSW
forces out of the deformation zone, the tension has to be added to the friction. Therefore, the total shear strength of MSW according to varying normal stress, increases due to the tensile strength activated by the fibres. The activation of tensile strength is a function of Young’s modulus, maximum tensile strength of each fibre and also the anchoring. The model for varying normal stress is developed because of the anchoring increases with the normal stress. The tensile strength is limited by the maximum tensile strength of each fibre (Figure 2). At the lowest normal stress (o. 120 mm is larger than 10% (by weight) and the height of each layer disposed at the landfill is less than 1 m. 2. It is not possible to estimate the deformation of different parts of the landfill realistically. This is a requirement to get suggestive results when using the linear model. 3. The thickness of layers disposed at the landfill is less than 50 cm and the MSW contains pieces > 40 min. 4. The load on the slope is low.
Figure 9. Results of the slope stability calculation for material 2
For the calculation, a maximum deformation of 20% was used with parameters measured in the oneaxial and triaxial tests. If the deformation is restricted to lower deformation because of landfill specifications (such as embankments, etc.), than the stability number will be smaller, but the trend is always the same. Figure 8 and 9 show the results of the stability calculation for material 1 and 2. By using the linear bearing model (oneaxial and triaxial tests) there is an apparent dependence on the landfill height. The dependence on the unit weight is not so obvious. From the bilinear bearing model the dependence on both, height and unit weight, is obviously much smaller. Even the range of the stability numbers is smaller than by the linear model. This is due to the different assumptions of both models. By using the linear model the tensile strength, expressed as cohesion, is determined independently from the load. Only when using a separate tensile apparatus (bilinear model) the tensile strength depends on the load (= normal stress). This means the tensile strength increases due to the enlargement of load. Therefore, the application of the bilinear model considers the characteristics of the material, while using the linear model leads to higher
6 EXPRESSION OF THANKS
We would like to thank the DFG (German Research Organisation) for financial support. REFERENCES Collins, H.-J., F. Kolsch & G. Zielunann 1997. Veranderung des Tragverhaltens und der mechanischen Eigenschaften von Abfallen durch Alterung und Abbau. AbschluJbericht DFG: Az. CO 76/26- 1 bis -5. DIN 4086. German Industrial Standarad Nr. 18136 (Anonymus). DIN 18136. German Industrial Standard: Nr. 18136 (Anonymus). Gray, D. H. & H. Ohashi 1983. Mechanics of fibre reinforcement in sand. Journal of Geotechnical Engineering: Vol. 109. ASCE. Jewel1 & Wroth 1987. Direct shear test on reinforced sand. Geotechnique 37. Institution of civil engineers. London. 857
Kockel, R. 1995. Scherfestigkeit von Mischabfall im Hinblick auf die Standsicherheit von Deponien. Dissertation an der Ruhr- Universitdt Bochum, Schriftenreihe des Institutes fur Grundbau: Heft 24. Kolsch, F. 1996. Der EinfluR der Faserbestandteile auf die Scherfestigkeit von Siedlungsabfall. Dissertation an der TU Braunschweig, Mitteilungen des Leichtweg-Institutes: Heft 133. Recommendations of the German Geotechnical Society for Landfills (Anonymus) 1997. GDAEmpfehlungen Geotechnik der Deponien und Altlasten: 3. Auflage. Berlin: Ernst und S o h Verlag.
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Slope Stability Engineering, Yagi, Yarnagarni & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
The stabilization of frozen technogenic dumps V. 1.Grebenets & S. N.Titkov Research Institute of Bases and Underground Structures, Moscow, Russia
A.G.-o. Kerimov Research Institute of Bases and Underground Structures, Norilsk, Russia
V. M.Anishin Norilsk City Administration, Russia
ABSTRACT: In Polar and mountain regions, the open mining accompanied by the accumulation of waste masses of loose rock material on the surface of slopes becomes complicated because of cryogenic and glacial pocesses. Complex observations on dangerous displacements of waste rock dumps similar to natural rock glaciers in morphology and creep characteristics have been investigated in the North of Russia as well as in the mountains of Eurasia. Experimental methods for the stabilization of frozen technogenic dumps tested under natural conditions are as follows: 1) artificial change of rock masses movement direction apart from constructions; 2) erection of protecting dams; 3) artificial cooling and freezing of ground by means of the installation of "steam-liquid" thermopiles using the unlimited source of natural ground cold. 1 INTRODUCTION The problem of movement of waste technogenic frozen rock dumps contained toxic components dangerous for the environment is acute in mountains and Polar regions with the intensive economic development and minerals mining operations. The open mining accompanied by the accumulation of loose rock material on the surface of slopes, becomes complicated because of cryogenic and glacial processes. Insufficient ability of natural complexes for the self-rehabilitation requires to work out some special engineering-geological and geotechnical methods in order to promote the stabilization of grounds and to protect engineering constructions from destruction. Processes and phenomena occurring at the slopes as a result of their industrial development are especially dangerous. Numerous excavations and mining sites strengthened the effect of instability of natural slopes. The costs for special protective engineering and geological efforts aimed at stabilization of the situation in Siberia have been increased within the last 30 years almost by 15 times. 2 RESULTS OF OBSERVATIONS In Polar and mountain regions of Russia (Yakutia, Khibin and Trans-Baikal mountains, Norilsk region, etc.) the problem of dumping of waste masses on the surface of mountain slopes under permafrost condi-
tions is extremely important. Dangerous displacements of technogenic rock dumps similar to natural rock glaciers have been observed in Norilsk region on the slope of Mt. Rudnaya (Grebenets et al. 1997). One of the largest in the world technogenic dump is located on the northern slope of Mt. Rudnaya. It was formed as a result of open pit mining. At present, the rock dump volume is about 60 million 1n3 (about 110 million tons). The dumping process of this rock mass has lasted for 25 years and was completed in 1984. The actual height is 105-120 111. The dumping was arranged layer by layer along the slope with the angle 12-15'. The bedrock were covered by Quai-ternaryinorainie sediments and loamy soil from 0,5 up to 6-7 m thick. Lenses of ice were found in many sites (25% of the square of the slope), with the thickness of 0,5-4 m and located closely to the surface. Visible deformations of the dump were observed in summer 1992. On the top platform of the dump vertical subsidences and fractures up to 5-7 m deep and up to 0,5 m wide were observed. Extention of fractures is up to 200-300 m. A "bulging shaft" about 600 m long containing fine grained structure was observed at the bottom part, along the road. Similar morphology is common for natural rock glaciers. Because of bulging of the excavated ground layers and underlying layers shift the "shaft" is up to 6-8 m high, sometimes up to 15 m (Fig. 1). Moving downslope the rock glacier reached the opposite side of the valley thus formed a dam for the small river. The lake appeared in the late
859
1996 and up to the end of Summer, 1998, it was about 300 m long and 50 m wide (Fig. 2). The movement of the technogenic dump makes the road operation (transportation of the ore) very difficult. In 1995 this dump destroyed the shaft and the road. The shift of the dump destroyed the drinking water pipeline. The extent of the frontal part of the dump is 900-1000 m, general displacement at tlie most dangerous site - 50-60 m. Mean velocity of mass displacement is up to 4060 mm/day, and sometimes up to 800-1000 mm/day in separate areas. The average speed of horizontal movement (in all observed areas) within 1993-1998 is about 40 mndday. The increase of displacement speed was promoted by penetration of waters from the communication located in the upper part of the slope. Penetration of water resulted in mobilization of separate parts of the dump. Cryogenic conditions deterioration (decrease of bond strength between ice and ground) inside tlie embankment is caused by the general tendency of permafrost degradation (Grebenets et al. 1994). Frozen ground temperature nearby the dump increased up to minus 2,5-3 OC by 1997 and caused the reduction of stability of frozen debris masses. The technogenic dump at the slope of the Mt. Rudnaya is a body with the complex structure containing ground and ice. It is, in many respects, similar to rock glaciers common for mountain regions all over the world. The main difference between a rock glacier and end moraine is the same as distinction between glaciers and the "dead" ice: glaciers are "alive", because they move (Barsch 1983). Rock glaciers are of considerable variety in respect of shape and size, surface characteristics and internal structure. The shape of rock glaciers can be of tongue, lobe, cover, terrace or front adapting to specific conditions of the relief. The length of rock glaciers varies from a few hundred meters LIP to a few kilometers and width - from several tens meters up to several hundreds meters. The thickness of rock glaciers usually does not exceed the first few tens meters, and the angle of slope of the surface is 10-20'. Movement of rock glaciers is the most characteristic feature which differentiates them from other natural formations similar in structure and composition. Usually, the rate of movement of rock glaciers front is from few centimeters up to few meters per year but deviations are possible. Thus, the velocity of movement of the frontal part of the rock glacier Burkutty (Northern Tien Shan) is 14 m per year (Titkov 1997). Movement of rock glaciers can be not only gradual but catastrophic as well. This is proved by unique technogenic rock glaciers in the Khibin Mountains (North-West of Russia) similar to those in the Norilsk region. Debris material mixed with 860
snow and freezing water During the year-round dumping on the slopes of 30-40' dip formed consequently ice-debris mixture of the ice-content exceeding 30%. This conglomerate became mobile under the pressure of 1,6 kg/sm2 and began moving down a 10-15' slope with the speed of up to 120 m per day. Then the rock glacier reached a rock bar after which the speed increased dramatically and the ice-debris mass broke up into separate blocks which rushed down onto the valley bottom. The volume of the deposed material was about 4x106 m3 (Debris Dumps on Mountain Slopes, 1975) Two main parts of the debris dump of the Mt. Rudnaya can be distinguished: 1) active part (60 YO of the dump), which is the most dangerous; 3) rather stable one at the slope, where the thickness of the sedimentary rock with ice is not big (2-3 m). As a whole, the dump is characterized by essential difference in surface shift speeds, formation and increase of fractures. Fractures are of 2-2,5 m wide and up to 3-4 m deep. Fractures are free of debris that points out the embankment's "living" condition and it's movement In our opinion, two kinds of movement can be observed: besides of technogenic dump's movement as the consolidated ice-rich debris body (like a realrock glacier) there appear crumbling of broken material from the upper parts and shift of separate layers. Shift of layers is the most dangerous process. As it is known, long-term resistance of ice to loads (including shift) is negligible, so the ice has the ability to move under the loading. Obviously, the movement of separate layers of the dump body within the limits of ice-rich layers along with the general movement of this thechnogenic rock glacier is of a special danger because along with the temperature change or increase of deformation the creepage may become into its progressive stage and cause avalanche (collapse) of the whole body or its part. A special problem is caused by the water penetration through the fractures and its further seepage into the body through the channels formed after ice thawing. 3. METHODS OF ENGINEERING Main engineering protective efforts are aimed at the evacuation of the most important objects from the zone of dangerous movement of technogenic rock glacier. Erection of a bulk dam (up to 10-15 m high) capable to support a significant part of the dump mass from collapse is effective for the stabilization of certain dumps,. The opportunity of application of artificial ground freezing techniques at the top of the dump by means of seasonally-cooling devices and natural cold aimed at the improvement of the engineering-geological situation have also been examined. Successful methods in application of similar
Fig. 1 Frontal part of the technogenic rock glacier. Norilsk. August, 1998
Fig.2 Lake formed as a result of damming of the valley by the moving front of the technogenic rock glacier. Norilsk. August, 1998
devices in foundation engineering as well as hydraulic engineering in the Norilsk industrial region have been developed (Grebenets 1990). A complicated engineering problem was solved
while providing the dam reliability of the circulating water supply at one of Norilsk plants: under the conditions of permanent positive fluid temperature in the storage basin ( 5 - 20 "C) thawing zone stabili-
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(ed.), Proceeding of the 6Ih Congress International Association of Engineering Geology: 12851287. Rotterdam: Balkema. Grebenets, V.I., Fedoseev, D.B. & Lolaev, A. B. 1994. Technogenesis influence on the frozen ground. In R. Oliveira & A. Balkema (ed.). Proceedings of the 7'" Congress International Assotiation of Engineering Geology: 2533-2536. Rotterdam: Balkema. Crebenets, V.I., Kerimov A.G.-o. & Baksheev, D.S. Dangerus movements of technogenic rock glaciers, Norilsk, Russia. In A. Marinos & A. Balkema (ed.), Proceedings of International Symposium on Engineering Geology and the Environment: 689-692. Rotterdam: Balkema. Titkov, S.N. 1997. Investigations of rock glaciers of the Tien Shan. Proceedings of I F Int. Conf On Geomorphol.: 274-275. Bologna.
zation (the back of the dam, storage side) is ensured by means of artificial freezing. Dikes were made of artificial loam; a special water-proof core along the longitudinal axis was foreseen. This core was filled with slurry to avoid hollows. Since the significant heat release could provoke thawing of ground and destroy the dike, 170 freon thermal piles of 0.108 m in the diameter, 2-3 m spaced from each other and driven to the depth of 12 m have been installed to provide a water-proof screen. The on-location observations revealed some spots of the temperature as low as minus 5-8 "C along the central lie of the area. Such temperature turns out to be 2-5 "C lower than the initial temperature of the ground prior to the construction of circulating water supply. The geocryological forecast based on natural observations predicted that frozen areas located 25 m away from the fluid coast line in the settler will preserve low ground temperature since thermal piles are under normal operation. Application of seasonally cooling devices may be the effective method of the stabilization of mobile technogenic rock dumps. Many questions, however, are unsolved, such as the methods of the forecast of movement and the effective ways of stabilization of dumps. 4 CONCLUSIONS The problems of technogenic dumps movement in mountain and Polar regions are the most actual ones for ground mechanics investigations. Experimental methods for the stabilization of frozen mobile technogenic rock glaciers are as follows: 1. Artificial change of the movement direction of debris masses apart from protected constructions 2. Erection of protecting dams. 3. Artificial cooling of grounds by means of installation of "steam-liquid" thermopiles using the unlimited source of natural ground cold. Probably, the international scientific cooperation will help to resolve these actual problems. This report has been executed under sponsor support of Russian Fond of Foundational Investigations, grant # 99-05-65352. REFERENCES Barsch, D. 1983. Blockgletscher - Studien, Zusam menfassung und offene Probleme. Abbildung Akademie Wissenschuji Gottingen. K1 (35): 133150. Gottingen. Debris Dumps on Mountain Slopes 1975. Leningrad: Nauka : 175 p.p. (In Russian). Grebenets, V.I. 1990. Antifiltration curtains constructions with natural cold utilization. In D. Price 862
Slope Stability Engineering, Yagi, Yamagami & Jiang (c) 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Stability of MSW mass: Use of an improved limit equilibrium analysis A. Bouazza & I. B. Donald Department of Civil Engineering, Monash University,Melbourne, Vic.,Australia
ABSTRACT: Assessing the slope stability of MSW fills has become a very important aspect of waste containment system analysis and design. As with any stability study the selection of the most probable mode of failure and the proper values for the strengths of the materials are the most critical aspects. The slope stability analysis for MSW landfills is usually performed using conventional method of slices or translational wedge method considering potential failure surfaces at limit equilibrium. Usually a safety factor of 1.3 to 1.5 is considered acceptable, other specific values can also be mandated by regulation. The present paper presents a new method, The Generalised Wedge Method (GWEDGEM) to analyse the stability of a municipal solid waste mass. GWEDGEM is a limit equilibrium method, which satisfies force and moment equilibrium and kinematic conditions. waste slope instabilities counted for 6% to 8% of the total failures encountered i n landfills. The recent waste slope failures reported by Mitchell (1996), Milanov et al. (1997), and Pardo de Santayana & Veiga Pinto ( 1 998) are also a salutary remainder to our profession on the importance of a proper evaluation of the stability of waste repositories. In this respect, assessing the stability of waste fills has become a very important aspect of waste containment system analysis and design. The stability of MSW landfills under static conditions is generally controlled by the following factors: 1) shear strength and compressibility of the foundation soils; 2) unit weight and shear strength of the waste; 3) height of the waste pile and angle of the front and/or side slopes; 4) leachate level and fluctuation within the waste pile; 5) composition of the landfill cap and its resistance to erosion. There are many potential failure mechanisms, which must be assessed in the slope stability analysis of MSW landfills; these include: 1) failure of the side slopes before or during waste placement; 2) sliding failure through the waste pile; 3) sliding along the liner system resulting in lateral translation of the waste material; 4) deep sliding failure through waste, liner and foundation soils.
I INTRODUCTION Landfills, including those designed to contain municipal solid waste (MSW) and hazardous waste, constitute a special class of containment facility that is heavily regulated by federal and/or state laws. Requirements are generally spelled out regarding types and conditions of acceptable waste materials, methods for waste placement and compaction, lining system design and construction, leachate collection system, and monitoring both during and after active operation. Consequently, the attention of the designer engineer has mainly focused on the design of pollution reduction/prevention systems and monitoring to ensure that current legal requirements for non-pollution are met. In this respect, very significant progress has been made in understanding the behaviour and performance of liners, covers, and leachate and gas collection removal systems under different operating conditions. However, stability of waste piles is another aspect i n landfill design, which needs to receive more attention than in the past. A recent survey of reported pollution incidents and other modes of failure affecting U.K. landfills waste disposal sites carried out by Roche (1996) showed that failures due to
863
The U S U soil ~ mechanic methods for slope stability analysis are generally also applicable for the analysis of waste landfill stability. However, most conventional inethods for stability analysis do not allow correctly for internal distortions and hence will not result i n a kinematically admissible failure mechanism. In this paper, a new method which overcomes this problem, the Generalised Wedge Method (GWEDGEM) is used to analyse the stability of a municipal solid waste pile. GWEDGEM is a limit equilibrium method which satisfies force and moment equilibrium and k i n em at i c conditions , 2 EVALUATION OF STABILITY
Stability analysis for MSW landfills are more complex than those for classical earth structures as a result of the difficulties involved in evaluating the physical and mechanical properties of the waste and the interface interactions, as well as the variation of these parameters with depth. In addition, the variation of the waste properties with time inay need to be considered in the analysis. As part of the stability analysis, the shape of the potential failure surface must be evaluated. Failure surfaces passing through the waste are generally circular. On the other hand if the stability along one of the interfaces (waste/liner, liner/foundation soil, etc.) is the most critical, the analysis may need to be performed co11 s i deri n g a non -c i rcul ar fai 1u re surf ace pass i n g along the interface having the lowest strength. The strain compatibility between the waste and the individual elements making up the base liner and cover systems should also be taken into account in the stability analysis of MSW landfills. Pore pressure variations, which can have a significant effect on the slope stability, must be taken into account, especially for foundation soils presenting an undrained behaviour. The slope stability inay also be influenced by the additional shear stresses generated near the base of the landfill as a result of lateral deformation of the waste pile. Given the variability of materials and the variations in strengths for any given material or interface combination, there remains always some concern over the appropriate values to use for analysis and design. The approach taken should depend on the particular problem being studied. If an analysis is being made of the stability of an existing landfill, or back analysis of a failure is required, then strengths that are the most representative of the
actual in-situ values should be used. Worst case scenarios and the probability of their occurrence should also be evaluated. On the other hand, for the design of new facilities and for the development of filling plans, conservative estimates of properties should be used. As suggested by Mitchell & Mitchell (1992), the actual values must be chosen with respect to: 1) the factor of safety, which may either be mandated by regulation or left to the designer; 2) the variability i n properties; 3) the possibility that properties inay change with time; 4) the type of failure being analysed and the consequences if it occurs; 5) whether the condition being analysed will be temporary or permanent. One of the major challenges facing the geotechnical engineer involved in designing landfills is the quantification of relevant geotechnical properties of waste materials. Quantification of these properties is very difficult because: 1) Municipal solid waste is inherently heterogeneous and variable among different geographic locations; 2)There are no generally accepted sampling and testing procedures for waste materials; 3) The properties of the waste materials change with time more drastically than those of soils. Furthermore, the complexity of the mechanical behaviour of domestic wastes makes the problem of landfill stability even more complicated to solve. In any case, in order to perform a simple analysis, basic parameters such as moisture content, unit weight, compressibility, and shear strength are needed. Further details are given in Bouazza & Wojnarowicz ( 1999).
864
3 METHODS OF ANALYSIS Many limit equilibrium stability analyses are currently i n use, although they all exhibit some deficiencies and difficulties in application. Most are based on some form of the inethod of slices but some use multiple wedge analyses. Programs in general use can handle circular or non-circular failure surfaces (either continuously curving or multi-linear for wedge analyses) but the choice of critical failure surface is often left to the operator, though some programs have the ability to search for the critical surface au toinaticall y. Conventional limit equilibrium vertical slice methods such as Morgenstern-Price ( I 965), Spencer (1973), Janbu (1 973) and Fredlund & Krahn (1977) are generally regarded as the best available for stability analyses, but they will not necessarily result
i n a kinematically admissible failure mechanism. Although these methods in principle satisfy force and moment equilibrium the computations often result i n unbalanced forces and moments, depending on the side force function selected and equilibrium conditions are therefore not strictly satisfied. In attempts to remove the convergence problems, which can occur with vertical slice methods. Chen & Morgestern ( I98 1 ) presented example analyses aimed at improving the estimation of side forces on slices, However, the relative simplicity of the vertical slice methods is then lost. A complete method of analysis should satisfy force, moment equilibrium and kinematic admissibility, i.e. the chosen failure mechanism involves neither overlap nor separation of elements. In addition it should be able to handle complex pore pressure distributions, heterogencous profiles, external loading, tension cracks, non-linear and anisotropic strength behaviour and an automatic search for the critical failure mechanism. All the features listed above are included in GWEDGEM, further inforination is provided by Donald & Giatn ( 1989).
Figure 1. Homogeneous slope with phreatic surface
3. 1 Basis of'Method Consider a homogeneous slope with a phreatic surface as shown in Figure 1. Drawn on the same figure is a simple 3-wedge failure mechanism. Free body diagrams of each wedge are depicted in Figure 2; while Figure 3 shows the force polygon of wedge 2. Convention a1 wedge met hods involve drawing force polygons for each wedge in turn. A closing force polygon at the last wedge would mean that force equilibrium is satisfied. This sort of graphical approach is straightforward and certainly instructive. However, it is limited to only a few wedges (2-3 wedges). It is not too bad when the location of the critical failure surface is roughly known. When such information is not available, particularly for slopes, graphical procedures can become tedious since several failure surfaces need to be compared to obtain the minimum factor of safety corresponding to the critical failure surface. The method presented here adopts the simplicity of resolving forces as in the conventional wedge method while simultaneously ensuring moment equilibrium to be satisfied. The idea of a general matrix equation for each wedge allows the extension to n-wedge systems. The failure surface is kinematically admissible since mobilised strengths are used at the
Figure 2. Free body diagrams of the failure mechanism shown in Figure1 .
Figure 3. Force polygon for wedge #2 internal shearing interfaces as well. Consider the free body diagram of wedge #2 and its force polygon, Figures 2 and 3 respectively. Known forces are calculated as follows:
865
s = -('") '
(i =
F
tan $,,,( =
~
1,3,5)
(tan $), F
(i = 1,2,...,5)
(3)
Where i= side numbering, 1 = length of side, Xi= mobilised cohesive component of shear strength at internal interfaces, S, = mobilised cohesive of shear strength at outer failure surface enclosing the slipping soil mass, tanQ,d,,,=mobilised coefficient of friction, Q, and R,=resultant of normal effective stress and mobilised frictional component of shear strength . Resolving forces horizontally and vertically and rearranging in matrix notation.
(4) Where
B=
-
4
C, = W, +X,cos6, -U,sin6, -S,sina, -U,cosa, -X
2 cos&, +U2 sin 6, - R2sin($2(,e,J - 6,)
about a point P, is defined as the product M= I Fld where d is the perpendicular distance between P and the line of action L of F. Equivalently, if r is the vector from P to any point Q on L, the magnitude of moment is given by vector product, M = 1 r x FI. Anticlockwise moments are positive. The latter approach is adopted since the sign of the moment is obtained simultaneously with its magnitude and therefore complicated wedge geometry will pose no difficulty. Therefore for the simple 3-wedge failure surface shown in Figure 1, starting with an initial assuined factor of safety, the developed friction angles, QiCjev and mobilised cohesive components of shear strength can be calculated from equations 1 and 3. Then beginning at the first wedge, from force equilibrium, R:! and Ql can be determined. By estimating the point of application of R2, the point of application of Q I can be evaluated from moment equilibrium. Equal and opposite calculated forces of side #2 is then applied to consider wedge #2. Siinilarly Q?, R4 and the point of application of Q? can be calculated. Finally for wedge #3; SS,Q 5 and the point of application of QS are calculated. But SS is the developed or mobilised shear strength required to just maintain the limiting state of equilibrium. Therefore, F for the last interface can be calculated and if this value is not equal to the original assumed F, a new value of F IS assumed and the above process iterated until convergence is obtained. The converged F is the factor of safety for the assumed slip surface. Several trial slips should be used so that the global minimum safety factor is more likely to be obtained. This leads us to the idea of using automatic search routines to reduce the amount of computation. Giam & Donald (1989) have successfully used multivariable unconstrained methods for the selection of critical failure surfaces. Furthermore, several extensions have been included so that the method presented can be applied to a wide range of problems.
Solving for Qi arid R4 by Cramer's rule, 3.2 Exarnple of application to lan@lls ( C, cos B - C2sin B )
"=
cos(A + B)
A, =
(C,sin A - C2cos A) cos( A + B )
(5)
By estimating the points of applications of R:! and Rj, the points of application of Q f ,Q3 and Qs can be calculated. In general the moment, M, of a force F
The evaluation of the stability of municipal solid waste slopes is done i n very much the same way as the analysis of any other type of geotechnical stability problem. The slope stability analysis is usually performed using conventional method of slices or translational wedge methods. considering potential failure surfaces at limit equilibrium. The safety factor is then determined by comparing the sum of the resisting forces to the driving forces
866
mobilised along the potential failure surfaces. According to current practice, a safety factor of 1.3 to 1.5 is considered acceptable; other specific values can also be mandated by regulation. It is however uncertain whether or not conventional limit equilibrium methods are applicable to MSW because of their ability to undergo large strains without reaching failure. Since these large deformations are not acceptable for the performance of the collection and containment systems, an at-serviceability state approach may be more appropriate. A major difficulty in performing slope stability analyses for MSW lies in the accurate assessment of the necessary physical and mechanical properties of the waste and the hydraulic conditions within the waste and foundation soils. The stability of two landfills is analysed using the GWEDGEM computer program developed by Monash University. The first example considered in this paper is taken from Jessberger & Kockel (199 1). It is related to the rehabilitation work carried out for the domestic waste landfill at Rheiland-Pfalz, Germany. The landfill slope is shown in Figure 4.
Figure 4 Cross section of the Rheiland-Pfalz landfill. The shear strength properties used for the calculation are those proposed by these authors, which were con-elated from CPT results. The stability analysis carried out by Jessberger & Kockel (1991) gave a safety factor of 1.39. TALREN 97 developed by Terrasol (France) is also used for comparative put-poses. TALREN 97 is based on classical slope stability methods considering a failure surface at limit equilibrium. The minimum safety factor obtained with TALREN 97 for circular slip surfaces passing i n the waste pile gave a safety factor of 1.35. This example has been re-analyzed with GWEDGEM using 3, 5 and 9 wedges. Using 5 wedges, the factor of safety corresponding to this critical failure mechanism is 1.29. No further improvement of the safety factor is obtained when using a number of wedges greater than 5. Interestingly, the analysis with 3 wedges gave a safety factor of 1.39 similar to the one obtained by Jessberger 6r Kockel (1991). A summary of the
867
results obtained by various methods is presented in table 1. Table 1: Comparison of safety factors for the Rheiland-Pfalz landfill. Method of Analysis Safety Factor Jessberger & Kockel (1991) 1.39 TALREN 97 1.35 GWEDGEM(3 wedges) I .39 GWEDGEM(5 wedges) 1.29 It is also interesting to report that a slight reduction in the friction angle (from 32" to 28") and the cohesion intercept (from 10 kPa to 5 kPa) gave a minimum safety factor close to 1. This demonstrates the need for a reliable estimation of the MSW properties, which is not often easy to obtain. Mitchell & Mitchell, (1992) stressed the fact that the most critical aspect of the evaluation of stability is the certainty with which the relevant properties are known. They pointed out that while uncertainties about analysis may lead to errors perhaps of the order of 10 to 20 %, uncertainties in strengths may easily result in errors that are twice as great. The second case study involves the internal slope stability analysis of the waste layers within a remediated landfill situated in the city of Porto Alegre, Brazil. The study analyses and compares the results of an original case study by Strauss et al. (1998) with results produced by GWEDGEM. The original study was necessitated by the need to expand vertically the existing landfill cell in order to accommodate extra waste. The main concern for the stability of the landfill is that it is located on a soft clay site. The original cell was originally designed to be 8 in high, after the vertical expansion is completed the cell will reach a height of 26 in. The following results were obtained from GWEDGEM using the original 4 sets of waste strength parameters given in table 2. The analysis of this case study indicated that the weak layer of soft clay sandwiched between the landfill and the foundation soil governed the determination of the final failure surface mechanism. The failure was in the form of deep seated block failure by lateral sliding along the soft layer rather than the conventional circular form reported by Strauss et al. (1 998). The safety factors obtained using GWEDGEM differ significantly from the original values reported by Strauss et al. (1998). This difference may be explained by the different failure surfaces analyzed in the original paper as compared to the current analysis adopting a translational block
type failure. It is also interesting to note that the values produced by the original paper gave very close safety factors for circular and non- circular analysis. Furthermore, the assumption of a uniform bulk unit weight of 7.5 kN/m' for the waste throughout the depth of the landfill is questionable. It is known that the unit weight increases with compression immediately, following application of overburden pressure due to waste placement. The u n i t weight may also increase with the additional compression that occurs over time. A value of 7.5 kN/m' seems to be very low to be representative of the actual unit weight of the waste. Table 2 Waste strength parameters as proposed by Strauss et al. ( 1 998). y(kN/m3)
SetA SetB SetC Set D
UFWL LWL UFWL LWL UFWL LWL UFWL LWL
7.5 7.5 7.5 7.5 7.5 7.5 7.5 7.5
c' (kPa) 0 0 c,, =24 0
4'
(")
35 31
16 16 16 13.5
0
33 28 22
28 22
UFWL=upper fresh waste layer, LWL=lower waste layer.
Tablc 3: Comparison of safety factors for the Porto Alegre landfill. Method of analysis Safety Factor Set A GWEDGEM I .73 Strauss et al. 2. I to 2.3 Set B GWEDGEM I .43 Strauss et al. 2.2 to 2.5 Set c GWEDGEM I .80 2.2 to 2.3 Strauss et al. GWEDGEM I .77 Set D Strauss et al. 2.1 to 2.2
4 CONCLUSIONS A simple and accurate multi-wedge stability analysis has been presented. The method is applicable to landfill stability problems and includes an efficient automatic search for the critical fdilui-e surface. The method is not only applicable to profiles containing weak layers but to all problems which may be handled by circular and non circular slices methods. The stability analysis of municipal solid waste landfills is a challenging task as the mechanical behaviour of waste is of a very cornplex nature. Unfortunately, the present state of knowledge is very limited, resulting i n a renewed interest to have a better quantification of the geotechnical properties of
wastes. Analysis of stability using conventional stability methods for cases where there is a presence of a weak layer can result in misleading safety factors. REFERENCES Bouazza, A. & Wojnarowicz, M. 1999. Geotechnical properties of municipal solid waste and their Implications on slope stability analysis of waste piles. 1 1 t" PANAM Conf. Soil Mechs. & Geotech Engng., Fos de Iguacu, (In press). Donald, I.B. & Giam, S .K. 1989. Iniprovecl comprelzensive lirnit equilibriurii stcihiliQ nrzalysis. Report No 1/89, Monash University, Australia. Fredlund, D.G. & Krahn, J. 1977. Comparison of slope stability methods of analysis. Carz. Geotech. J . 14:429-439. Giam, S.K & Donald, I.B. 1989. Appropriate opt inzim tiorz techniques f o I- fh ilure suflace cleterriiirzatiorz irz geotecliriical stabiliql arzalysis. Report No 3/89, Monash University, Australia. Janbu, N. 1973. Slope stability computations, Embankment dam engineering, J. Wiley & S011:47-86. Jessberger, H.L. & Kockel, R. 1991. Mechanical properties of waste material. XV CGT Ciclo di Colif: di Geotec. di Torirzo. (offprint) Milanov, V., Corade, J.M., Bruyat-Korda, F. & Falkenreck, G. 1997. Waste slope failure analysis at the Rabastens landfill site. Proc. 6"' Zizt. Laridfill Synip., Cagliari, 3:55 1-556. Mitchell, J.K. 1996. Geotechnics of soil waste in ater i a1 i n t eract i ons . P roe. 2I id Zri t. COrzg re SS Eriv. Geotech., Osaka, 3: 1425- 1474. Mitchell R.A., & Mitchell, J.K. 1992. Stability evaluation of waste landfills. Std7iZity cirzd Pei:forrizance of slopes nncl Ernbankinerits, ASCE. Geoteclz. Spec. Pitbl. No 3 1 : 1 188-1520. Pardo dc Santayana, F. & Veiga Pinto, A.A. 1998. The Beirolas landfill eastern expansion landslide. Proc. 3'" Dzt. Congress on Em). Geotecli., Lisbon, 2:905-9 10. Roche, D. 1996. Landfill failure survey: a technical note. Erig. Geology of Wcrste DispcxaL, Geo. Soc., Erzg. Gro. Spec. Publ. 1 1 :379-380 Spencer, E. 1967. A method of analysis of the stability of embankinents assuming parallel interslice forces. Geotc~cluzique,17: 1 1-26. Strauss,M., Bica, A.V.D., Schnaid, F., Brcssani, L.A. 6L Reichert, G.A. 1998. The stability of a remediated landfill on soft clay. Proc. 3'" Iizf. Congress on Erzv. Geotech., Lisbon, 1 :393-398. 868
Slope Stability Engineering, Yagi, Yamagami& Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Stability of bentonite wall by the unified method of molecular dynamics and homogenization analysis Y. Ichikawa, T. Seiki & T. Nattavut Department of Geotechnical and Environmental Engineering, Nagoya University,Japan
K. Kawamura Department of Earth and Planetary Science, Tokyo Institute of Technology, Japan
M. Nakano Department ($System and Information Engineering of Bioproduction, Kobe University,Japan
ABSTRACT: Bentonite is a micro-inhomogeneous material consisting essentially of nanometer scale of clay minerals, mainly montmorillonite. We propose a new numerical simulation procedure, which is a coupling method of the molecular dynamics (MD) and the homogenization analysis (HA). The procedure is called the unzfied MD/HA method (Kawainura et al. 1997; Ichikawa et al. 1998). Here MD is used for defining micro-scale material properties based on the molecular behavior, and HA is introduced to extrapolate the microscopically inhomogeneous continuum characteristics to the bulk-scale continuum behavior. Permeability of bentonite calculated by this unified MD/HA method is well-conformable to the experimental data given by Pusch (1994). In this paper we apply the permeability results to a consolidation problem, and calculate stability of a clay wall designed for waste disposal facility. 1 INTRODUCTION
Bentonite is recently used as an engineered barrier, known as geosynthetic clay liner, for disposal and containment of hazardous waste. When designing the disposal facilities we commonly apply macro-phenomenological models to predict water flow and pollutant transport. However the existing models are not sufficiently effective, because these do not always reflect the atomic-based true physical and chemical behavior, which is essentially important for the transport phenomena in bentonite. In this sense a new scheme is required for analyzing the true behavior of clay. Bentonite is a typical micro-inhomogeneous material. That is, it consists of nanometer size of clay minerals (mainly sodium montmorillonite) , micrometer size of macro-grains such as quartz and feldspar particles, pore water, and air in its microscopic level. A montmorillonite mineral is of lamellar shape with size of approximately 100xl00x ln m , and a group consists of several montmorillonite lamellae and interlamellar water. Two issues are crucial for analyzing the behavior of the rnicro-inhomogeneous material. One is how to determine the characteristics of constituent components of the micro-continuum which are directly affected by their molecular movement: and another is how to relate the microscopic characteristics to the macroscopic behavior. For solving the first problem, we apply the Molecular Dynamics method (MD; Allen 8~ Tildesley 1987: Kawainura 1990); then we employ the Homoge-
nization Analysis (HA; Sanchez-Palencia 1980) for estimating the micro- to macro-behavior. This procedure is called the unified MD/HA method (Kawamura et al. 1997, Ichikawa et al. 1999). The predicted macroscopic properties, for example, hydraulic conductivity is quite compatible with experiment-baxd data. In this paper we discuss the stability and consolidation behavior of bentonite wall, which is designed for waste disposal facility.
2 MOLECULAR BEHAVIOR O F MONTMORILLONTE HYDRATE The structure and physical properties of clay minerals are hardly known by means of experimental methods because of their poor crystallinity. Note that the molecular formula of Xa-montmorillonite hydrate with n-interlamellar water is given by Nalp A12 [Sil1/3A11/3]010 (OH), .nH2O. We call this nH2O system, and if n = 0. it is called the "dry montmorillonite" . We have applied molecular simulation inethods for specifying the true physical and chemical properties of montmorillonite hydrate (Kawamura et a,l. 1997, Ichikawa et al. 1999). The molecular simulation methods belong to a field of computational physics and chemistry. There are two major tools in this field. that is. the metropolis Monte Carlo method (MC) and MD. In MD the motion of every molecule is given by the Newton's equation, and the force is cal869
culated by differentiating an inter-atomic potential function. The key issue is to determine the interatomic or intermolecular interactions quantitatively. We use a new empirical interatomic potential model. That is, the potential function for all atom-atom pairs (i.e., the 2-body term) is composed of the Coulomb, short-range repulsion, van der Waals and Morse terms, and a 3body term is added to the H-0-H interaction because of its sp3 hybrid orbital. Details are found in Kawamura (1992) and Kumagai, Kawamura & Yokokawa (1994). 2.1 Swelling Property of Montrnorillonite Hydmte We can calculate a wide variety of physical properties by using the MD results. In the calculation of swelling property of inontmorillonite hydrate we employ an (NPT)-ensemble MD scheme (N is the number of molecules, P the pressure, and T the temperature) under condition of 300K temperature and O.1MPa pressure. The Verlet algorithm is used for time-integration with a time interval At = 0.4fs, and by the Ewald method the electrostatic energy and force in long range interaction are calculated. We plot the calculated swelling property of Na-montmorillonite in Figure 1 comparing with experimental data (Fu et al. 1990) of Na+Wyoming montmorillonite with the formula Na0.75 [%75 Alo.25l(A13.5Mg0.5)020(OH), .nH20. MD results well coincide with experimental data in spite of slight difference in the molecular formulae.
Figure 1. Swelling property (= basal spacing) for our model compared with experimental data for the Wyoming rnontmorillonite given by Fu et al. (1990). 2.2 Diflusivit y and Vzscosit y The MD model is of one clay-mineral layer with 3,000 water molecules. The size of a basic cell is ca. (3.1, 2.7, 15.9)[nm]. An (NVE)-ensemble MD (V is the volume and E the internal energy) is carried out with a time interval Al = 0.4fs after equilibrating the system with 50,000 steps of (NPT)-ensemble MD calculation. A snap shot for
Figure 2. MD results for the montmorillonite hydrate with external water. the clay-water system is shown in Figure 2(a). We divide the clay-water system into 50 slices with 0.186nm thickness in s-direction, then we can calculate the mean square displacement (1n.s.d.) and the diffusivity (slope of the m.s.d.) for each molecule in the slice. Then by applying the StokesEinstein relationship with its diffusing spare 6 = 0.152[nm], which is obtained by our MD calculation for pure water (without clay mineral), the viscosity of water at each sliced region is determined. Figure 2(b) shows the diffusion coefficient and viscosity in each slice. We find the structurally ordered water layer in contact with the clay-surface, which is called the "ice sheet". Thickness of the sheet is ca. 0.5nm, and is equivalent to two layers of water molecules. In the diffusion layer of 3 to 4nm thickness, the diffusion coefficient is rapidly changed with distance from the clay surface. The viscosity is also changed in this region. We call such a water property the iceberg efjrect. 3 SEEPAGE PROBLEM BY HOMOGENIZATION ANALYSIS (HA) HA is a new type of the perturbation theory developed for a micro-inhomogeneous material with periodic microstructure (Figure 3). We here apply HA to the seepage problem in bentonite with distributed water viscosity in vicinity of montmorillonite. For this probleiri we start with the Kavier870
Stokes equation, and obtain a macroscopic seepage flow equation including the effect of spatial distribution of viscosity. BVf ~
=0
in
Yf,
I
V) =
o
on
r.
Now we introduce a mass averaging operation for Eqn(3)1, and get the Darcy's law:
where is the averaged inass velocity in the unit cell ( jY I : voluine of the unit cell). Averaging of E-2-term of the mass conservation equation derived from Eqn( 1 ) 2 yields the following mucl-o-scale equation [MaSE], called the HA-seepage equation:
Figure 3. Macro- and micro-scale problems in HA.
3.1 HA Fo7niulation of Seepage Problem with Distributed Viscosity
av,o
= 0 in R. (7) Bxa The first order approximations of pressure PE and velocity V,. are given by
We think a flow problem in porous media with a microscopically periodic domain (Figure 3). The local coordinate system y is related to the global coordinates tc by y = X / E . The incompressible viscous flow field is given by
~
v,"(z)= cr2Ko(x,y), P"(x)= P o ( z ) . (8) In geotechnical engineering we usually use the following empirical Darcy's law
BV,"
(9)
- = 0 in RE,,
dxi
v,E = 0
P
on Bs2"f
H=--+< P.9
where v," is the velocity, P' the pressure, Fz the body force vector, q the shearing viscosity, and the water flow region in the global coordinate system (8R,, its boundary). We introduce an asymptotic expansion
where is the average velocity, H the total head the elevation head. Compared this with and Eqns (5)-(8) , we know the correspondence
/ C D (4) where Fs (max.) : Factor of safety considering the allowance in pile stress E R: Resultant of the joint resisting forces along the slip surface 2 D: Resultant of the joint sliding forces along the slip surface Pi' 1.3 : allowance for shear force of pile Note that although Pi' is the allowance for the shear force of the pile, the flexural stresses induced on the landslide prevention piles are much larger than the shear stresses developed. Therefore, rather than the total clearance in shear stress, only a portion proportional to the allowance provided for flexural stresses is used in the analysis. Setting the following: 0 r: induced shear stress, U M: induced flexural stress, 0 allowable shear stress and 0 Ma: allowable flexural stress. The allowance for the flexural stress is given by (T MaU M , and its ratio with respect to the induced flexural stress is ( U M~ - 0 M ) / U M . Next, it is assumed that the allowance for shear stress ( (T '> is proportional to the allowance for flexural stress, i.e., (T s ' = ( T s ( ( T M a - M ) / ( T M =(T ( ( T Ma / ( T M - 1 ) (5) 5
Noting that the allowance for shear force (Pi') is equal to the product of the cross-sectional area of the pile work (A) and 0 s', it follows that A ( (T M-1)=Pi( (T M-1) (6) Pi'= (T I
In the above equation, Pi is the shear force per unit width of the pile obtained in the analysis. Therefore the maximum Factor of Safety, Fs ( max) along the slip surface can be estimated from the allowance of pile shear force by substituting 899
Equation (6) into Equation (4) , Le., / M-1)cos Fs (max) = c R+Pi( 0 M ~ 0
8 }/xD (7)
The resisting force, sliding force and factor of safety along the slip surface considering snow t earthquake load are shown in Table 3. Table 3: Factors of safety considering allowance of pile shear force (snowt earthquake load pile location lower pile center pile upper pile
resisting force (kN) 3372 3619 3678
sliding force (kN) 3184 3009 3134
factor of safety 1.059 1.203 1.173
From the above table, and considering the three -0) mentioned earlier, the order of aspects (0 pile work effectiveness is as follows: center pile ) upper pile ) lower pile. Note that for the case wherein the allowance of pile stresses is considered, it is only in the center pile that the prescribed factor of safety (Fs =1.2) is obtained. In the case of the lower pile and the upper pile, measures such as improvement of pile stiffness are also necessary.
Table 4 shows the resisting force, the sliding force and the factor of safety along the slip surface corresponding to the cases wherein the slip surface passes near the upper portion and towards the lower portion of each pile, respectively. From the table, it is clear that the center pile is the only location wherein the prescribed factor of safety (Fs =1.2) is satisfied for both cases. Note that Fs is less than 1.0 in the case of lower pile and Fs = 1.1 in the case of upper pile. Therefore, the center pile is the only pile position whose effectiveness is the greatest and has an adequate factor of safety considering snow + earthquake loading condition. Table 4: Factors of safety for the case of slip surface passing through the top and towards the lower part of the pile (snow t earthquake load ) pile resisting I sliding factor of location force &NI force O lower pile. is as follows: central pile
>
4 CONCLUSIONS From the results of the analysis, the effectiveness of the landslide prevention piles considering the installation location can be evaluated in terms of the following: (1) the increasing proportion of safety factor of joint elements along the slip surface (2) the magnitude of the over-all safety factor and the sliding resistance of the pile (3) the safety factor considering the allowance in sliding resistance of the pile REFERENCES
Figure 6: Factors of safety with the slip surface passing immediately above the piles
l)Gotoh, M. 1997. Study on the stability of embankment with plate anchors on soft ground by using non-linear finite element analysis, Journal of Japan Society of Civil Engineers, No..567/VI -35, pp.213-223 ( in Japanese) . 2) Fujita H 1990. Slope stability analysis and prevention work planning by FEM, Journal of Japan Landslide Society, Vol. 27-4, 19-26 (in Japanese).
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Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Design and constructional aspects of an anchored slope and gabion revetment system M. H. Kabir & A. M. Hamid Depurtinent of Civil Engineering, Bangladesh University of Engineering and Technology, Dhuku, Bungladesh
ABSTRACT: The design and constructional aspects of a pilot study on an anchored slope and gabion revetmeiit system are presented. A brief outline of the problems and related research in the area is also presented. The structure included, construction of soil slopes with silty soils, anchored by steel wires and masonry blocks. The slopes are protected by wire mesh encapsulated anchored gabion revetmelit system. A 300111 protection system was constructed for a remote village, subjected to high intensity rainfall, submergence, wave action and relatively rapid draw down. Tlie materials and niethods employed were compatible with construction using unsltilled labour. Performance of the structure during last five monsoons, including the worst flooding of the century in 1998, is also reported.
1 INTRODUCTION Tlie paper reports on a pilot project on anchored soil slopes protected by wiremesh encapsulated stone revetnient system. Tlie project site is located in a village nanied Nayapara. in Klialiajuri thana of Netrolcona district, situated i n the northeastern region of Bangladesh (Figure 1). These areas, called Haor areas, are saucer shaped flood basins, inundated annually by rain waters from tlie vast north eastern hilly catchments. Tliese catlinients are situated in, tlie wettest of areas of the world, the hill regions of Assam, Meglialaya and Tripura statcs of India. The flooding period is nornially betwen mid May and mid September. The villages and the settlements in this area are normally constructed on earth filled raised platforms. These tale tlie form of isolated islands during flooding. The slopes of these platform are normally constructed by using very erosion prone local soils. These are subjected to wind iiiduced wave action in 3 to 5 meters water depths. Thc fetch being several kilometers to tens of kilometers, Miith wind speeds often touching I50 km/h mark. Tlie erosion of villages in the IIaor areas is probably one of tlie most severe natural calamities faced by people of this country. Almost every year, a good number of people lose their homestead eithcr totally or partially. The process is progressive which results in increase in homeless aiid landless people on a continuous basis.
A number of types of erosion protection systems are in use. These include, traditional non-engineered systcnis using bamboo and vegetation to engineered structures like niasonry and concrete retaining walls. Recently, slopes armoured with brick or concrete block revctments are also being used, (Figures ?(a), (b), (c) & (d)). Most of tliese structures are inefficient and suffered total or partial failures. After taking up the work in November 1993. studies were carried out to arrive at sound, durable, easy to construct and cost effective solutions to the problems. Special emphasis was placed on tlie remoteness of tlie area and construction by unskilled labour. An integrated reinforced soil slope aiid revetnient system was cnvisagcd. designed and coiistructcd for tlie leading wave faces. This system incorporated, polymer sheathed steel anchor wires, reinforced masonry anchor blocks. geotextile filter and wiremesh encapsulated stone revetment structure. Some aspects of design of the slope of this structure is described in this paper, along with the construction sequence and methodology. Performance of the structure was monitored mainly through visual observation and photographic methods. Tlie structure sustained and performed quite satisfactorily during the last four years’ seasonal monsoon flooding. Tlie flood of 1998, was the worst in this century. The top of the structure was eroded due to inundation by high water levels and wave action.
901
Figtire I The Northeast Region fiemedial ineasiires are being undertaken now 2 I U S E A R C H BElNG UNDERTAKEN
fieserii-cli being undertaken by the authors, generally, include studies on slope and wall type stnictures f'or erosion protection of villages i n Haor areas 'l'hese include, laboratory and analytical studies on behaviour of such structures, under high intensily rainfall, rapid draw down and wave action. The stiuct tires included, anchored gabion faced stepped (tired) walls, anchored walls using tyre, ferrocement and clay tile facing elements and ancliored i-evetinents using concrete and soil cement bloc1,s. Behaviour of geosynthetic horizontal drains ( G H 1)) cuni reinforcement strips and geotextile filters \\ere also investigated. In all the cases sandy bacl; till materials and plate anchor blocks with poly niei- sheathed steel anchor wires were used Cieoteclinical stability as well as revetnient stability was iiivesrisated, in the laboratory, under simulated field conditions. A catalogue of facing types used are presented i n Figure 3. Figure ;(a) shows circular t i c in coiicrete blocks cast in jute fabric form, with steel pin connectors. A structural unit comprising foiii- soil cement blocks, held by a central conical ceiiieiii concrete wedge is shown i n Figure 3(b). Facing elements including, tyres, cubic gabion boxes and continuous gabion mattresses are shown in Figui-es .3(c). ( d ) Rr (e) respectively.
_. 1 he anchored soil slope and gabion revetnient system \?,as constructed for the leading wave faces at Nayapai-a village of Khaliajuri thana (Figure 1). The scliema[ic diagi-ain of this h c n d system is presented i n Figure 4. On the secondary face geojute and vegetati\,c, .s(# erosion protection system was pi o\.ideci 'l'lie construction was completed during the first lialt'of 1994.
Figure 2 Conventional protection structures
Figure -3 Catalogue of facing types used in ~~esearcli 3. 1 7 h N[gx//)lrtzi ~ illup up The Nayapar-a village is located at the western tip oi' the Khaliajiiri thana headquarter, ad-jacent to the bazaar The village is nearly r-ectanylar 111 pian a i i c i elongated rouglily in the east-west directioii This measures approximately 256 i n x 43 m The t o p level of the village was es~ablishedat approximately 3ni above the adjacent Haor- flat level. This \ z a s based on the 19SS flood level, woi-st in a centtir?~. The village was conipletely eroded during the I9SS floods. This was reconstructed in the dry season of 1993-94 to the formation level, by burrowing earth fi-on1 the adjacent areas. 'The hard slopes \\ere consti-iicted on the southern and western faces. T1-w soft slope was constructed on the northern lice, the
902
b) Back analysis of data from survey of tell tule marks on damaged structures in villages around Khaliajuri area. c) Interviewing village people on their experience and observation. For a design wind speed of 100 km/h and water height of 3ni the wave height was estimated as 0.75111.
4 DESIGN OF ANCHORED SOIL AND GABION REVETMENT SYSTEM The details of the anchored soil slope and gabion revetment system are presented in Figures 4(a) & (b). Some aspects of slope structure configurations, stability analyses and the different structural elements, are described in the following.
4.1 Slope Structure ConJigurutioiz The slope is 3m high, divided into two 1.5 meter high sections with a berm of lni in between (Figure 4 (a)). The design of the slope structure incorporates two distinct design philosophies. These are, anchor wire reinforced and coinpacted soil slope and anchored, wireniesh encapsulated stone (gabion) revetnzen/ structure. The first one meter width of the compacted soil slope is held together by secondary anchors to associate the soil mass as a block. This block is held back by two meter long primary anchors. The primary and secondary anchors also held the wiremesh encapsulated stone gabion revetinent structure at grid points. The anchors in this design are intended to serve dual purpose. These are (a) increase of internal stability of the soil slopes, (b) anchorage of the wiremesh encapsulated stone revetment to the soil slope to increase stability against uplift and sliding down failure. A geotextile filter layer was incorporated between the base soil and the revetment structure.
Figure 4. Details of soil reinforced stone revetinent eastern face merged with the high grounds of the bazaar.
3.2 Site Geology and Wave Clinzatology The project site is situated in one of the deepest locations of the I-Iaor areas. These areas were developed by a process of deltaic sedimentation in a slowly subsiding tectonic basin. The surface soil is mainly composed of yellowish gray silts. Occurrences of organic soils in deeper horizons are common. The surface soils are very erosion prone and problematic from filtration point of view. The climate is subtropical with an average annual rainfall of approximately 4,000 min. Over 80% of the rain fall during monsoon season, from June to October. This site is flooded every year to water heights between 2.5 to 3 meters. The water logging is due to slow drainage through Meghna River. In absence of any data on wave action in this remote area, the following approaches were followed to establish design values. a) Calculation of wave height from fetch, water depth and wind data according to Shore Protection Manual (SPM, 1984)
4.2 Stability of Soil Slope Stability of anchor reinforced soil slope requires checking of external and internal stability to ensure adequate factor of safety. a) External stability: This is to ensure adequate Factors of Safety against deep seated base failure of the foundation soil or block sliding failure along the surface of the foundation soil. The foundation conditions are adequate at this site and stability analyses resulted in FOS 2 1.5. b) Internal stability: This for a soil mass, reinforced with anchored tensile elements, are governed by the physical properties o f the soil
903
Figure 5. Stability analysis of the slope
2 [(by cos
2 Tr; (2)
together with tlie strength, size, spacing and length of the tensile elenients and anchor blocks. Latcral stresses and strains with in the soil mass are resisted and counterbalanced by the anchor force and bearing mobilised at the anchor block. Design is therefore a matter of determining the strength aiid spacing of these eleinents for tlie particular type of fill materials to be used and the relevant geometry. Stabilty analyses are performed to calculate the local stability of the soil near a single anchor strip as well as those which consider the stability of wedges of soil with in tlie reinforced zone. Considering the slope geometry, soil and anchor characteristics in tlie reinforced zone, the spacing of tlie aiiclior reinforcement was decided. Since tlie reinforcing eleinents provide internal stability to this zone, it can be assumed to behave nioiiolithically and should have an adequate factor of safety against failure. In this case a safety factor of 1.5 was deemed to be sufficient and series of stability analyses showed that the zone of reinforcement required to be 211-1 inside the soil slope. To determine tlie required horizontal aiid vertical spacing between the tensile anchor elements a method analysis developed by Oltasan Kogyo (1988) was used.
Where FS= factor of safety, u, = I I , ~ , , = pore water pressure, h, =height of water above base of circle, AY, =width of slice, y,,=unit weight of water, AL, = arc length of slice, n = nuniber of slices, i n = number of reinrorcing layers. All other notations are given in Figures 5(b) & (c). Sliding failure analyses were performed in accordance with that suggested by E'ultuolta and Goto (1988). Forces in tlie reinforcing wires were checked to ensure safety against breakage failure. Tliese did not exceed tlie limit of 4.2 1tN witli a factor of safety of 2. Finally the stability of the anchor blocks were checked to ensure a factor of safety of 2 against pull out bearing failure. To provide additional internal strength of tlie embankment and stability of the encapsulated revetnient structures, secondary reinforcing anchor elements were provided at vertical and horizontal spacing of 0.66 ni.
4.3 Stubilitj) Analyses
4.4 Geotextile Filter.
Overall stability analyses of the slopes (Figure 5(a)) were performed using XSTBL program by Sliarnia (1990), which utilizes modified Janbu inethod of analysis. The local stability analyses of tlie upper and lower slopes, incorporating reinforced zones, were performed by Bishop's modified method with tieback reinforcement. In both the cases seepage under rapid draw down condition were critical. The relevant equations for total and effective stress analysis are presented in Equations (1) and (2)
1(W, sin B,)R I=I
904
F,Ly= /=I-
0, - U , h , ) tan $ iCAL/) 8), blast furnace slag, garbage from phosphates production (phosphogypsum pH < 4), mine wastes with pirytes, community wastes etc. Fermented sludge contains in average: Ntotal- 2,2%, P205 - 0,7%, K20 - 0,4% of dry matter. Ca consistency is rather high - 2-4% (Pronczuk 1994). Ten tons of dried sludge have the same fertilising value as 0,65 t of ammonium nitrate, 0,84 t of granulated superphosphate and 0,lO t of potassium salt. Metals consistency are valuable microelements and heavy metals. Waste sludge must fulfil conditions prepared by Ministry of Health and Social Protection (MZiOS 1984): - must be originated from municipal waste treatment stations, - must be well fermented or composted, - can not contain more metals (in mg per 1 kg of sludge dry matter than: 2500 mg Pb, SO mg of Cd, 25 mg Hg, 300 mg Ni and 1500 mg Cr.
As a result of these works have been elaborated three technics of hydroseeding with utilisation of coinmunity waste sludge: 1. Surface technic based on the periodically soil protection against erosion or till seeding time: for seeding must be prepared watered sludge - about 40 ni3 per 1 ha with - 3-6% of dry matter (may be used only sludge or with addition of mulch and seeds), what means 2 t of dry matter per 1 ha. Sludge is the base substance, also colloidal and protective. This technic should be used for seeding humidified and rich soils, exactly as antierosion protection in creating embankments and/or slopes (GDDP 1993). 2. Subsoil technic, when sludge is utilised as fertiliser. Quantity of dewatered sludge should be increased up to 300 m3 per 1 ha with 510% of dry matter content. Such quantity gives up to 30 ton of dry matter per 1 ha. 3. Combined technic, which contains simple agrotechnicai cultivation with fertilising and antierosion protection by covering thin layer of sludge by hydroseeding. As fertiliser is given dewatered, fermented waste sludge with dry matter content 2030%. The second and third technics are preferable for seeding on barren soils, devastated and degraded lands, without soil surface, also antropogenic soil.
All of the technics are consisted in know-how and patents no.: know-how 3/85-4/85 IBDiMPOSTEOR Sopot, Poland; Patent 147319 & Patent 162546. Hydroseeding can be used during total regeneration season, also after first autumn frost, mainly on the southern sandy slopes. 927
Institute and subcontractors gives 3-year guarancy for grass and legumes slopes prepared due to specified technology, of course when are provided necessary nursering. In Poland we use hydroseeders constructed on a base of agricultural sanitation tanks. Hydroseeders are equipped with HSP- 100 pump with capacity 400 to 800 dm3/min and pressure 0,4 to 1,0 MPa. Scheme of carrying out reclamation by hydroseeding method with usage waste water sludge are on the fig. 5 . Technology implemented by Institute for Building, Mechanisation and Electrification of Agriculture IBMER) based on composition of a few main elements: sewage sludge from biological treatment plants, grass and legume seeds and fly-ash (PJBJOR 1996) with woodchips and/or confetti. Properties of sewage sludge were described in technology TG-70. Fly-ash contains several macro and micro minerals iiiiproving soil properties. Typical characteristic of two types of ash in Poland is given in table 1. Quantity of macro and micro elements in ash is secure for plants consumption and quality except high content of Aluminium (Pronczuk 1994 & Duczynski 1990). High pH level (8-11,8 pH) improves soil quality, and fly-ash in mixture can be added to the acid soils in doses up to 200 kg/ha. Increasing level sludge of pH by adding ash creates higienisation effect by hilling bacteries and pathogenic microorganisms (Glazewski 1998). 4 CHOSEN OBJECTS HYDROSEEDED w WASTE WATER SLUDGE
m
e Soding slopes by hydroseeding method on ring road near Siedlce on area 4,9 ha. RDP Siedlce, 1982-84. Scope of works: project, carrying out, supervision.
Table 1. Chemical contents in fly-ash
I I
Component
1. Siiican (Si02) 2. Nitrogen (N) I 3. I Potassium (K2O) 4. Sodium (Na20) 5. Calcium (CaO) 6. Magnesium (MgO) 7. Ferrum (Fe203) 8. Aluminium (A1203) 9. Phospho~us(P205) 10. Sulphur 6)
I
1 1. Cuprum (Cu) 12. Arsenium (As)
1 Plombum (Pb)
I 16. I Chromium (Cr) 17. Nickel (Ni) 18. Boron(B)
I
24,OO
I
1
Fly-ash fkom Fly-ash fiom browncoal pit-coal Y O Y O 62,SO 83,30 0,80 0,16 0,33 1 0,36 I 0,17 0,06 28,30 5,90 7,32 2,lO 3,50 5,O 1 18,33 17,34 0,13 0,36 1.70 1.61
41,OO 34,OO 34,OO
95,50 0,93
I
116,OO 74,OO 4 1,OO
1
Reclamation by hydroseeding method slopes of phosphogypsum dump embankments in WiSlinka on the area 4,2 ha. G d ~ s Factory k of Phosphate Fertiliser Production, 1983-84. Scope of works: project, carrying out, supervision. Soding enbankments of fly-ash and slag dump by hydroseeding on area of 12,4 ha, Warsaw-Zeran,
Fig. 5. Scheme of hydroseeding with tractor and hydroseeder (hydraulic seeder). 928
1
heat Power Station, 1984-85. Scope of works: project, supervision. 0 Reclamation by hydroseeding slopes of ash dump on the area 22,7 ha, Power Plant Bekhatow, 1988-89. Scope of works: project, supervision. 5 NURSE OPERATIONS AFTER HYDROSEEDING We should obtain after hydroseeding proper shaping of the plants, soil should be covered properly, and protected against erosion (Glaiewski & Karpinski 1994). Before we will start with any nursing operations should be checked if soil coverage is properly moved, enriched, roughnesses and tufts are cutted. Moving just after germination protects against growing weeds, and for future causes reinforcement and better growing of plants. Moved grass must be immediately removed from planting area. After moving is very properly enriching soil by ammonia fertiliser. Dose of applied fertiliser depends on quantity of earlier applied waste water sludge. So, the rule of thumb is limitation of applied doses to the small interference amounts. Single dose of ammonia fertiliser should not exceed 30 kg of pure N per 1 ha. From the 15 years practice was stated (GIaiewski, & Ziaja 1995), that, in spring or after the first moving, fertilising of sow by mixture of sludge containing 94% of water in quantity 4 l/m2 with 30 kgha of Aininonia saves in good condition plants growing on slopes. In the second year after seeding, planting should be moved twice in the third year ones. Such practicing will provide to formation of lawn.
6 POSSIBILITY OF HYDROSEEDING TEHNOLOGY FOR PLANTING BUSHES AND TREES In the Road and Bridge Research Institute in 19891990 (MZiOS I990 & Dzieriawski et al. 1987) were realised preliminary research works on implementation hydroseeding for planting bushes and trees on slopes of roadways and highways with utilisation waste water community sludge. Conducted modelling research and results obtained on the experimental plots verified this method as proper for hydroseeding, and good for creating forest on poor, agricultural lands (Dzierzawski & Glaiewski 1995). However, is necessary to prepare elaboration methodology of seeds preparation as stratification and scarification of seeds from chosen trees and bushes in cooperation with dendrologists and foresters. 7 CONCLUSIONS Hydroseeding technology (hydromulching covering) of grass mixtures with application of waste water sludge in comparison to another methods of biologi-
cal reinforcement of earthen structures has the following advantages: - high level of mechanisation and limitation of labour consumption and manpower, - utilisation of waste water sludge beeing garbage substances, difficult for waste water treatment stations and environment instead of mineral and organic fertilisers and partially setting emulsions, - possibility of getting good quality soding without subsoil humification, - limitation of loses caused by water and wind erosion, - shortening of period necessary for recultivation and decreasing work costs. Comparing above with other earth protection methods, hydroseeding enables the high level of works mechanisation, production of good quality soil without use of humus and the utilisation of waste substances arduaus for the environment. Conducting works as well, that slopes soding of earthen structure will be permanent, requires good preparation from botanics, soil and agronomy. Due to the existing practices deposition on the just formed Iayers of the earthen structures waste water sludge with mulch addition will protect slopes, (in the good conditions) against erosion for 3 to 6 months up to reinforcement of soding structure. Comparing above with other earth protection methods, hydroseeding enables the high level of works mechanisation, production of good quality soil without use of humus and the utilisation of waste substances arduaus for the environmental. Technology of hydroseeding is fast and efficient, 3 to 5 times cheaper than traditional humification and sowing, and fulfils present demands on implantation rich green zone.
REFERENCES Duczynski J.P. 1990. Wplyw popiolu z wqgla kamiennego na niektore wlaiciwoici fizyczne gleby piaskowej. Symp. nauk. z okazji jubileuszu prof J. Prohczuh. SGGW, Warszawa: 183-191. Dzierzawski K. 198 1. Zadarniania skarp drogowych budowli ziemnych z zastosowaniem hydroobsiewu. Konj N-7: SN-TJiTO, Warszawa 9’8 1: 68-9 1. Dzieriawski K. & Glaiewski M. 1995. LeSne zagospodarowanie osad6w wtornych z oczyszczalni. Ekoin.@nieria l(2). Lublin: 16-20. Dzierzawski K., Glaiewski M. & Makowski J. 1990. Ingenieurbiologische Bepflanzung der Boschungen - Dynamische hydrosaat mit Anwendung der Abwasserablagerungen. Prace IBDiM 1/90. WKiL, Warszawa: 89-97. Dzieriawski K., Glaiewski M. & Rokicki M. 1984. Badania nad optymalizacja, hydroobsiewu. TG-70. IBDiM, Warszawa (know-how). 929
Dzieriawski K., Glazewski M. & Rokicki M. 1987. Zadrzewianie i zakrzewianie ziemnych budowli komunikacyjnych metodq hydroobsiewu TG-97. IBDiM, Warszawa (typescript). Dzieriawski K.& Kqielewski K. 1981. Hydromechaniczne obsiewanie skarp. TG-61. IBDiM, Warszawa (know-how). EN-4435/M/10/1984. W d i przyrodniczego wykorzystania osad6w Sciekowych z oczyszczalni komunalnych metodq hydroobsiewu. MZiOS, Warszawa. GDDP 1993. Zasady ochrony Srodowiska w projektowaniu, budowie i utrzymaniu dr6g. Dzial 04 - Ochrona Srodowiska w Budowie Drbg., Warszawa. Glazewski M. 1998. Hydroobsiew skuteczny i szybki.Rekultywacja biologiczna elektrownianych odpad6w paleniskowych. EKOPROFITnr 2(18): 14-19. Glaiewski M. 1991. Umacnianie skarp budowli ziemnych TW-3. IBDiM, Warszawa (typescript). Glaiewski M. & Dzierzawski K. 1985. Spos6b rekultywacji nieuzytkow i urzqdzenie do uprawy rekultywacyjnej nieuzytkbw, zwlaszcza na skarpach. Projekt 3/85-4185 IBDiM/POSTEOR. GdahskNarszawa (know-how). Glaiewski M. & Karpinski F. 1994. Ukreplenije sklonow i odkosow gidroposjewom. Awtomobilnyje dorogi Nr 10-1I , Moskwa: 42-44. Glaiewski M. & Makowski J. 1993. Soil and fly-ash dumps reklamation by means of hydroseeding based on sewage sediments. 4-th Inter. Symp. on the Reclamation, Treatment and Utilization of Coal Mining Wastes. Krak6w: 863-872. Glazewski M.& Ziaja W. 1995. Wyniki rekultywacji skladowisk popiol6w przy zastosowaniu hydroobsiewu mieszankami traw i motylkowatych. WMit Nr 4/95: 170-175. KBN nr P06F022508 1997. Temat badawczy. Rekultywacja utworow antropogenicznych metodq hydroobsiewu. Warszawa (know-how). MOSiZN 1990. Sprawozdanie z realizacji I etapu pracy badawczej pt. Zadrzewianie i zakrzewianie nietodq hydroobsiewu. Warszawa (typescript). Pachowski J. 1983. Question I: Earthworks,drainage, subgrade in Poland. XVII World Road Congress Sydney. Australia (discussion). Patent nr 147319 z 1987.11.02. Sposob hydrodynamicznego siewu. Patent nr 162546 z 1990.07.25. Spos6b umacniania skarp o pochyleniu stoku naturalnego i naruszonej strukturze gruntu. PIBJOR 1996. Postanowienie nr 50/96 wyraiajqce opiniq o ochronie radiologicznej odpad6w paleniskowych w postaci zuzli i popiol6w stosowanych do budowli ziemnych oraz rekultywacji. Warszawa. PN-S-02205 1998. Drogi samochodowe. Roboty ziemne. Wymagania i badania. 2.0 1/98. Pronczuk J. 1994. Popioly; melioracje i ochrona. WMiL Nr 2/94. Warszawa: 60. 930
Siuta J. 1988. Przyrodnicze zagospodarowanie osadow Sciekowych. IKS, Warszawa.
Slope Stability Engineering, yagi, Yamagami8,Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Investigation and stabilization of a sliding hillside J. Farkas Geotechnical Department, Technical University of Budupest, Hungary
ABSTRACT: This paper presents the analysis of a case of major surface movement, the underlying causes and the mode of reconstruction work. This landslide occurred at the building site of a storehouse in a cutting made into a hillside in Budapest. During excavation, soon after the natural support at the toe was undercut, a mass of 12 000 m3of soil began to move. An attempt was made to block the movements by the construction of buttress drains but rainy weather still reactivated the movement of the earth mass. Then, to stop the movement, 3 m high stabilizing berm was placed before the toe of the sliding mass. The designer would have liked to add nine new stone ribs to the retaining structure, but stability analysis proved that not even such a solution could have resisted the sliding action. Eventually an anchored slurry trench wall 8 m deep was constructed to support the back face of the excavation. The paper describes the results of ground investigations, the design of the stone ribs, the stability analysis of the hillside and the causes of ground movements. In addition it calls attention to the potential risk of underestimating the inherent danger of constructing on sliding-prone hillsides - on the part of both the designer and the contractor - because unfavourable conditions (awkward underground conditions, or/and a rainy period during construction) may trigger the movement of earth masses with the outcome of significant economic losses. Finally a brief review is given of the effect of moisture content increase on shear strength. 1. INTRODUCTION In Hungary, owing to the topography of the country, construction activity on hillsides is quite common. Human interference with existing conditions often leads to distress and instability of natural slopes. There is a number of known cases where surface movements occurred immediately after completion of excavation or even during construction works. In order to possibly avoid or prevent such movements it is important to have a thorough knowledge of existing site conditions which prevail before any construction activity would take place. It was intended to build a two-storey high storehouse of 50x20 m area on slab foundation in the North-Budapest region, at the foot of the hilly area where the rather steep hill Remetehegy ("Hill of Hermits") emerges from the Alluvial plateau of the Danube on B 103 to 104 m altitude to B 145 to 148 m heights. [B stands for: above Baltic Sea Level]. The plans contemplated to cut off the hill-foot up to the elevation of the two-storey building and to construct an earth-inserted monolithic reinforced concrete building the 50 m long backwall and the
two side walls of which would act as retaining structures. The reinforced concrete ground and floor slabs, as well as the perpendicular partition walls designed in every second module were intended to serve as supports against earth pressure actions. These latter elements were needed to transfer from the stressed floor slabs the loads, which (according to stability calculations) resulted from the earth pressure on the main walls designed as retaining structures. Excavations for the retaining wall sections were designed with slopes. Having known the perils hiding in the underground in the surroundings (water seepage, marks of sliding), it was decided to enhance slope stability by building five stone ribs spaced 8 to 16 m aprat into the back slopes which, at that site, inclined with 56 degrees, to the horizontal. These trapezoidal shape ribs of 3 m width had 13 to 14 m base lengths, 6 to 8 m heights at their backs and faces with inclination of 560 to the horizontal. The arrangement can be seen on Fig. 1. In lack of reliable soil physical data the ribs were designed on the basis of approximative calculations and mainly on engineering guess and experience. Assumed seeping water collected by the stone ribs 93 1
Water seepage occurs mostly on the surface of the basic clay dipping at 100 to 200 inclination towards the valley. Some springs and oozing water can be observed on the area. Nine boreholes were sunk on the site. As an example, the log of B.H. NO 8 is illustrated on Fig. 2. where the actual soil parameters are also given. Shearing stress parameters were derived from (CU) triaxial tests performed on undisturbed samples taken from the borings made after the sliding. Striking the eye is the very low shearing resistance (4 = 70, c = 10 kPa) on the interface between the grey Oligocene clay and the overlying yellowish-grey sandy fat clay. Similar testing results were attained by testing the core samples dug out from the same interface zone in the working pits. Moisture content of the soils in this zone is extremely high (see Farkas and Kovacs, 1996.). From the data gained in the boreholes completed after the sliding, the contour lines of the surface of
was to be discharged through a perpendicular drain at the toe of the ribs. It was planned to use crushed stone backfill material between the earth slopes and, up to the top, the reinforced concrete retaining backwall, after having completed the load supporting members of the building. This deep-drain, parallel to the backwall, would have had 1 m deep clay plug on the top with duly arranged surface drainage. In the course of construction 10 to 12 000 m’ earth had been excavated at and disposed from the foot of the hiU. The working plateau before the excavated slope was at B 106,55 m in May, 1995., when, due to a heavy rainfall, sudden movement of the backslope was experienced. The rainy period lasted for the uncoming weeks whereby the movement of the earth slope accelerated and caused 3 cm slopeward displacement of the stone ribs in the next month. This meant that the critical shearing resistance in the preconsolidated clay became klly mobilised and surpassed a limit value where the shearing resistance reduces while the displacement would continue even at shearing stresses lower than the critical peak value.
Borehole No. 8 W
? 0,o
2. UNDERGROUND CONDITIONS
Oh
Basic rock is the Kiscelli clay fiom the Oligocene (grey, medium and fat clay of Ip = 25 to 36 %) which, 5 to 6 m below its surface, becomes a hard and stiff marl. It is overlain by 5 to 8 m deep hillside debris, consisting mostly of clay, stony clay, interwoven by sandy seams. Water percolates through the seams and so, the area was in movement already during the deposition of the covering strata.
~
-( 14,O1
Figure 2.
Figure 3.
Figure 1. 932
the impervious Oligocene clay were plotted on Fig. 3.
3. STABILITY ANALYSES Several sliding events were observed in the past in the vicinity of the building site. Almost always the displacement of the oxidised clayey zone and the surcharge material over the hard surface of the Oligocene clay could be demonstrated. Experience indicated that the interwoven water-permeable sandy seams served surely as contributing factors to the slidings by having increased the pore-water pressure and decreased the shearing resistance. In our case the movement occurred in the wake of a permanent rainy period. The developed fissure lines and the direction of the movement are represented on Fig. 1. In the line of the movement a gully-llke depression was detected in the surface of
the basic clay (see Fig. 3.) This cross-section was later accepted as determinant for carrying out the calculations with the most probable supposition (based on the stratification in the boring holes and trial pits) that the sliding surface was at the bottomline of the trough (Fig. 4.) at, or quite near to the surface of the Oligocene clay. Principles of the calculation are shown on Fig. 5. Due to the water-absorbing capacity of the stone ribs and the presence of the sandy seams in the overburden clay, uplift forces and seepage forces were not assumed in the calculations. Post factum exploration data revealed that only the back edges of the ribs A, B and C (on Fig. 1.) reached down below the surface of the basic clay, while the bottom level of ribs D and E remained high in the secondarily deposited clay. Insofar the parameters, internal fiiction angle = 70 and cohesion c = 10 H a were used for the calculations on the supposed sliding surface on the interface between the two main deposits (Fig. 2.), the factor of safety against slippage has far not attained the unity, f = 1. Checked was also the safety factor for sliding resistance of the ribs partly restrained by the Oligocene clay. This resistance was supposed to derive fiom the fiiction and adhesion on the embedded bottom and side faces of the rib. The pressure on the sidewalls of the rib was calculated fiom the earth pressure at rest. Finally, the whole mass behind the backwall of the building was taken as a moving solid mass: this way, the driving forces made 47810 kN and resistance forces made 42405 kN, i.e., the stability of the mass behind the
+
933
wall was not adequate, what, otherwise, was proved by the ensued movement of the earth mass in question.
storehouse, our suggestion for stabilizing the situation was to sink a 6 to 8 m deep anchored diaphragm wall from the surface of the berm before the slid earth mass, all along the 50 m long backwall and to the necessary lengths beside the side walls. These were completed in the dry summer period; then the reinforced concrete ground floor slab and the internal foundation blocks were completed. At last, the sidewalls were erected under the protection of falsework atop the slot walls, together with the other two floor slabs. Thin prefabricated drain panels were placed behind the wall sections. This way having the problems with the earth pressure solved the two-storey storehouse has been completed.
4. STABILIZING MEASURES In order interest to stop the slowly creeping movement of the earth mass, a 3 m high earth berm was placed to the toe of the hill, which partly reached up to cover the ribs. This situation is shown in the cross-section of rib D, on Fig. 6. Thereafter the sliding has really stopped, but the presence of the berm obstructed the execution of the foundation work for the planned building. Neither was it possible to wait to see whether or not the sliding would regain forces and start again.
Figure 6. Therefore the structural designer advised to build nine, 1 to 1,5 m wide new retaining stone ribs behind the walls. Control calculations revealed that the new ribs - with bottoms in the Oligocene clay - could increase the safety factors against sliding of the total mass, or against the slip-out of earth masses between the ribs, to f = 1.28, even to f = 1.70, but when the earth berm was removed to give place for the foundation work, these values diminished to f = 1.07 and 1.18. Obstructed in addition was this solution by the fact that the construction of the stone ribs in the remoulded mass would have been rather dangerous and complicated. It has to be mentioned at this place that it would have been opportune to increase the safety by flattening the backslopes, or to build a deep-drain in the background, but these approaches were barred, because that portion of land did not belong to the client. Finally, bearing in mind anticipated construction costs, elapse of time and other difficulties in connection with building of new ribs, which for that matter were still not perfectly certain to resist increasing earth pressure actions on the proposed
5. EFFECT OF MOISTURE CONTENT I N C E A S E ON SHEAR STRENGTH
As shown by the case study, water content strongly influences shear strength on the interface between the grey Oligocenic clay and the overlying yellowishgrey sandy clay. With increasing water content, clay particles adsorb an increasingly thick water fdtq weakening or partly destroying bonds between particles. With the thickening of water film between particles, cohesion decreases, soil at the layer boundary becomes so to say pulpy. Water primarily affects clay minerals and properties of some clay types with noncrystalline components. Clay minerals are "softened" by water (e.g. montmordlonite swells), the texture of clay loosens. When at last shear stress exceeds ''surface active stresses", particles glide on each another. Acquired experience of the author and the results of various sliding types which were investigated and analysed by him show that in 84 percent of 350 cases in Hungary the water seepage on the critical sliding surface took the major part in causing the failure. Such seepage is generally a temporary phenomenon: it presents itself after heavy rains and after thawing in springtime, and originates fiom the infiltration of precipitation on the land overneath the incriminated area (see Farkas, 1983). When cuts are made, the soil under the slope plane gets unloaded, it expands; part of the elastic "energy" accumulated in the earth mass is released, and absorbed by the subsequent displacement. Expansion entrains increase of water content. The rate of expansion depends on the "hidden" deformation energy, due the preloading of the clay. So, the presence and movement of water (in any form) plays a determinant role in a development of close to surface earth movements. On Fig. 7., the exponential correlation between the uniaxial (unconfined) compressive strength and the moisture content of a heavy (fat) clay from a slide is represented. 934
7. Every kind of earthwork which may reduce the stability of a hillside, should be done in the dry season. 8. It is never enough in sliding areas to design earth retaining structures for the pressure at rest, but the equilibrium of the mass above the sliding surface has also to be analysed throughout.
It is an ancient observation - and the case study shows it, too - that correlation exists between the movement of the ground surface and the quantity of precipitation. Among the reasons for such movements, directly, or indirectly - in an overwhelming number of cases - is the role of precipitation. Hungarian records demonstrate that most of the slidings occurred after a long lasting, or intensive rainy period, and during the melting of the snow, respectively, in most cases when the winter was long with plenty snow and thawing was rather slow.
REFERENCES Farkas,J. 1983. Surface motions at clay interfaces. Melykpitkstudomhnyi Szemle. No. 8. pp. 355361. Farkas, J. 1992. Experiences from landslide investigation in Hungary. Proc. of the 6th Int. Symp. on Landslides. Christchurch, New Zealand. Farkas, J. and Kovacs, M., 1996. Investigation of highway cut slope movements. Proc. of the 7th Int. Symp. on Landslides. Trondheim, Norway 1683-1686. Rotterdam. Balkema.
6. CONCLUSIONS 1. In entering the design of structures on slidingprone underground, profound care should be exercised by the engineer, by all means more than in an average case. 2. In performing the design, more detailed and more extensive underground exploration is needed and the governing soil parameters - primarily the shearing strength parameters - have to be tested in sufKcient number. 3. It is good to remember that the shearing resistance in the interface zone is predominantly less than in the under-, or overlying layer. 4. Separately should be examined the possibility for the development of sliding surfaces on the surface of the impervious underlying clay below the oxidised soil zone. Discolouration of the layers dark grey, bluish-grey colour - may call the attention to this situation. (Farkas, 1992.) 5. The role of seeping water is known to have important influence on the development of slidings: it is therefore almost imperative to collect and discharge these perils fi-om the underground. 6. Should it come to the design of retaining drains or stone ribs, care has to be laid on having them founded on, and keyed into, the basic subsoil.
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Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Stability reinforcement of the old embankment sanitary landfills for remediation works E. Koda Department of Geotechnics, WLirsawAgricultural Universiw, Poland
ABSTRACT: The paper presents the stability reinforcement solutions on the two old large embankment type sanitary landfills localised nearby Warsaw, i.e. Radiowo and Lubna. The methods consist of retaining wall, berms, geogrids and tyre mattress. The paper also presents shear strength parameters determined for different kinds of wastes disposed on a.m. landfills as well as stability analysis of the landfill slopes by classical (Swedish and Bishop's) and finite element methods (FEM). Field investigations consisting of morphological analysis, settlement measurements, WST and CPT sounding, back analysis as well as slope failure tests were carried out for determination of parameters. 1. INTRODUCTION Remediation works on old sanitary landfills need the solvency of many problems in the framework of geotechnics [ISSMFEBiTC, 19931. Particular, geotechnical attention on old embankment landfills must be paid to internal and external stability. This is connected with the necessity of determination of waste morphology and physical properties and mechanical parameters of wastes. Determination of waste mechanical parameters needs the introduction of new testing methods or modification of test results interpretation used for soils. Development of new testing methods should be proceeded by longterm investigations and experience. Therefore, methods used for soils, considering specific waste properties are usually adopted for solving current geotechnical tasks in existing landfills, particularly in that old one. Stability analysis of the landfill slopes needs the readjustment of the calculation methods used in geotechnics. 2 SITES CHARACTERISTICS 2. I Radiowo site
Radiowo landfill is located in the north-west part of Warsaw. Since early 60-ties to 1991 the landfill was used as a place where municipal wastes from Warsaw were disposed. Actually, it covers ca. 15 ha area (Figure l), and it is higher than 55 m. Now, only non-composted wastes from Radiowo compostory are stored on the landfill. Radiowo
compostory, the biggest in Poland, with its capacity of approximately 600 ton wastes per a day, gives approximately 300 ton of non-composted wastes. The organic matter content for non-composeted wastes is ca. 4% [Koda, 19971. Central and south parts of the landfill are filled with 10-30 years old municipal wastes, while upper layers in the north part are filled with fresh non-composted products. This significant difference between the two kinds of wastes stored on the landfill influences the diversification of mechanical parameters. Subsoil of Radiowo landfill generally consists of cohesive soils. Locally, non-cohesive soils were founded to the depth of 10m. Groundwater level is at the depth 0-1.0m. On the basis of CPT and DMT tests, following shear strength parameters for stability analysis were proposed: (9'=27", c'=40kPa (cohesive soils) and (9'=33" (non-cohesive soils). 2.2 Lubna site
Lubna landfill is located at the distance of approximately 35km to the south of the centre of Warsaw. The landfill has existed since 1978. Now, it covers area of approximately 20 ha, and it is almost 50m high (Figure 2). Lubna is the only sanitary landfill where all kinds of municipal wastes from Warsaw are stored, i.e. ca. 1500 ton per a day. It is planned to be closed in 2000 year. Subsoil of Lubna landfill consists of non-cohesive soils and muds reaching the thickness of 2-15m, underlayed by boulder and lacustrine clays. Groundwater level is at the depth 0.5-2.0m. 937
Figure 1. The location of test points and the reinforcement construction on Radiowo landfill.
938
Figure 2. The location of test points and the reinforcement construction on Lubna landfill. 939
The in situ tests were performed in 1993-98 for Radiowo and in 1996-98 for Lubna landfill. They were to determine mechanical parameters of wastes for stability analysis, settlement prediction and estimation of bearing capacity for a road foundation. The main purpose of the tests is to utilise the existing landfills entirely, i.e. the determination of shear parameters in order to assure safe slope inclination. The WST sounding was generally performed along the axis and in the vicinity of roads constructed on the landfill. The tests have been repeated when 5m thick wastes had been laid. The sounding results are used for quality control of the road foundation compaction. The average amount of N20 for fresh wastes was approximately 10, but for old wastes - locally of approximately 5. The amount of Nzo increases twice, when disposing wastes were interbeded by sand layers (Figure 3). The CPT soundings in Radiowo were carried out in the northern part of the landfill, to the depth of ca. 25m. The CPT test showed the difference of compaction in disposed wastes. The degree of compaction for fresh non-composted wastes was 1~=0.2-0.5,while for 10 years old municipal wastes it was 1~=0.3-0.7.The CPT tests interpretation procedures, widely used for the determination of shear parameters for soils, were adopted for wastes. The effective angle of friction for wastes was reached within the scope @'=25-45", locally with lower values of 4'=20-25". These values were received after having considered wastes as noncohesive soils. Published test results confirm the existence of wastes cohesion. Therefore, real values of @'will be lower. The CPT test interpretation for wastes, analogically to cohesive soils, gave total shear strength of zfu=80kPafor non-composted and zh=9OkPa for municipal wastes. Figure 4 presents the example of CPT test results for Radiowo site. At the end of observations on test embankment [Koda, 19971, slope failure tests by concrete slabs
Figure 3. The example of the WST test results for different wastes on the landfills.
Figure 4. The example of the CPT test results for Radiowo landfill [Koda, 19981. were performed for verification of shear strength parameters (Figure 5). The ultimate bearing capacity (4) results from failure tests were used for verification of shear strength parameters for noncomposted wastes and for non-composted wastes with sand layers [Koda, 19971. On the basis of back analysis for estimation of bearing capacity of foundation on slopes and according to stability analysis, values of shear strength parameters of noncomposted wastes were established: @'=20" and c'=25kPa (Table 2), while for non-composted wastes with sand layers were: 4'=25" and c'=23kPa. Back-stability analysis by the Bishops', Swedish and FEM (Z-SOIL numerical program) methods for landslides, which took place in 1991 in the north-east part of Radiowo (old wastes) and in 1995 in tubna (fresh wastes), was applied for shear strength parameters verification. For the landslide in the old part (in 199l), failure surface was confirmed by CPT sounding (Figure 6). Slope inclination of the landfill just before the failure was ca. 1: 1.15 and the height of the slope was 46m. The example of the backanalysis results for the three cross-sections (Figure 1) on Radiowo landfill is presented in Table 1.
Figure 5. Scheme of back-analysis of the slope failure tests on Radiowo landfill [Koda, 19971. 940
Table 1. Stability factors from back-analysis of slopes on Radiowo landfill (for +'=26" and c'=20kPa) - cross-section location, see Figure 1. Cross-section Fmin A-A 0.989 0.967 (Eastern 1.03 slope) 1.029 B-B (Eastern 0.984 :;yks
Figure 6. Back-analysis of the landslide on Radiowo landfill in 1991 [Koda, 19971. From the back-analysis for the landslide on Radiowo landfill, the following shear strength parameters were reached: @'=26" and c'=20MPa (Table 2). These parameters were accepted for old municipal wastes on the both landfills in the design stability analysis. From the back-analysis for the landslide in Lubna, calculated minimum stability factor, Fm;,=0.994(cross-section IV-IV - see Figure 2), was reached for the following shear strength parameters: @'=21" and c'=15kPa (Table 2). These parameters were accepted for fresh wastes. In the case of Radiowo landfill, the morphological composition of wastes creates an additional factor influencing mechanical parameters. 4 STABILITY CONDITIONS IMPROVEMENT OF RADIOWO LANDFILL In order to improve stability of the slope (15m high) located close to the street (Figure 1 and 7), there have been done [Koda et al., 19971: the retaining wall, moderate slope inclination from 1: 1 to 1: 1.75, replacement of non-composted waste in the road foundation (of 5 m deep) and the lateral reinforcement with five geogrid layers. In the west part of the landfill, there is a gas pipe
Category
~
(Western slo e
Unit weight Normal stress Shear angle of friction y l k ~ / m ~ ] CT [kPa1 @ 1"l Radiowo 9.0 35 20
:.:: 1 I ~
1.142 1.092
Intercept cohesion c @Pal 25
12.0
50
25
23
14.0
65
26
20
11.0
125
21
15
941
Remarks landslide (in 1991)
slope with cracking
I
~
stable slope
and a railway line (Figure 1). The inclination of the slope is 1:1.25, what causes the danger of the landslide. While the design was preparing, the slope was 20m high, with the final height of almost 55m. Taking into consideration limited area in the close vicinity, the bottom part of the slope was reinforced with the narrow berm. The upper part of the slope was reinforced with one geogrid layer and three layers of tyre mattress (Figure 8). In the bottom of the berm, the drain layer for leachate was made. The surface of the berm was made of cohesive soil and compost. The slope stability analysis was performed with classical methods used in geotechnics and with FEM method (Z-SOIL numerical program). There are no reliable determination procedures of waste mechanical parameters, therefore the use of sophisticated models for stability analysis seems not to be advisable. The lateral reinforcements (geogrid, tyre mattress) were taken into account in stability analysis [Koda, 19971. The stability analysis results of Radiowo landfill, according to Swedish method (without and with reinforcement) are presented in Table 3. All factors of safety for reinforced slopes are higher than 1.3. This fact results from the proposed reinforcement solutions.
Site
Non-composted wastes Non-composted Radiowo wastes with sand Old municipal Radiowo wastes Fresh municipal Lubna wastes
Method Bishops' Swedish FEM Bishops' Swedish FEM Bishops' Swedish FEM
The tests methods slope failure tests, CPT, WST slope failure tests, CPT, WST back-analysis of landslide, CPT, WST back-analysis of landslide, WST
Slope Western Northern Eastern
Crosssection I I1 I11 IV
without reinforcement Reinforcement Swedish FEM 1.04 1.13 berm, tyre mattress, geogrid 1.43 1.49 berm, tyre mattress, geogrid 1.03 1.11 less steep slope, geogrid 1.18 1.23 berm
reinforcement Swedish FEM 1.34 1.36 1.73 1.81 1.68 1.75 1.57 1.62
high) was analysed [Koda, 19981. However, this is difficult and very expensive solution, so that the crib buttress was also replaced by the berm. 6 CONCLUSIONS
Figure 7. The cross-section 111-111and reinforcements of the northern slope [Koda, 19981.
Figure 8. The cross-section 11-11and reinforcements of the western slope [Koda, 19981. 5 STABILITY CONDITIONS IMPROVEMENT OF LUBNA LANDFILL In the case of Eubna landfill, when the slopes are high and of considerable inclination, the berms seem to be the most effective solutions for the slope stability reinforcement. The berm enables to reach additional capacity for waste disposal (Figure 9). On the west slope, in the first step of designing, in order to ensure stability improvement, a crib buttress (15m
Figure 9. The cross-section 11-11of Lubna landfill. 942
It was difficult to estimate shear strength waste parameters only on the basis of CPT soundings. There is no explicit interpretation methods of wastes. Shear strength parameters for non-composted wastes were verified on the basis of slope failure tests, while parameters for old and fresh municipal wastes were verified on the basis of the back-analysis of landslides. The parameters determined in this way are thought to be reliable for the design purpose. Construction of the berms seems to be the most effective method of the stability improvement of old landfills, however it needs the extension of the landfill in the close vicinity. Tyre mattress are cheap and effective method of the slope stability reinforcement in the landfill conditions. There are no reliable determination methods of waste shear parameters, therefore the stability analysis of sanitary landfills should be carried out by the classical methods in geotechnics. The result of the stability analysis should be recommended for the design purpose. The stability factors of slope landfills fiom FEM method are a bit higher than those from Bishop’s and Swedish methods. REFERENCES ISSMFEETC 1993. Geotechnics of Landfr’lls Design and Remedial Works - Technical Recommendations GLR. Ernst & Sohn, Berlin. Koda, E. 1997. In situ tests of MSW geotechnical properties. Contaminated and derelict land, GREEN 2, 247-254, Bolton. Koda, E. 1998. Stability conditions improvement of the old sanitary landfills. Proc. of the 3th Intern. Congr. on Envii: Geot. : Vol.1, 223-228. Lisboa. Koda, E., Fohyn, P., Golqgowski, P. & K.Pejda 1997. Zabiegi wzmacniajqce statecznoik skarp starych wysypisk odpadow komunalnych. Proceedings of the Nat. Con$ on Geotech. in Landfill Constr : 169- 182, Pultusk (in polish).
Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Stabilization and remedial works on some failed slopes along the East-West highway, Malaysia A. Jamaludin Khairi Consult Scln Bhd, Consulting Engineers, Selangor Darul Ehsan, Malaysia
A. N.Hussein Public Works Department of Malaysia, Kuula Lumpur, Malaysia
ABSTRACT: Slope stability problems associated with the design, construction and maintenance of roads in mountainous terrain have always received a great deal of attention from geotechnical engineers as well as government agencies. In Peninsular Malaysia, about half of the total land are hilly to mountainous terrain that occupies, mostly the less-developed forested region in the hinterland. During monsoon seasons, a number of cut slopes and fill slopes fail which result in a great amount of money being expended on slope stabilisation and remedial works. This paper presents some experience on methods of stabilisation and remedial works carried out on some failed fill slopes along the East-West highway located in the northern section of Peninsular Malaysia. In order to achieve the most cost-effective solution a comprehensive assessment comprised of surface and sub-surface investigations, design and method of construction are discussed.
1 INTRODUCTION
alignment cannot be shifted further north due to the Malaysian-Thai border or to the south because of more difficult terrain. The vertical alignment of the highway is limited by the reservoir water levels of the Temenggor dam and the two bridges at Banding Island. It rises from approximate elevation of 1OOni (a.s.1) to reach approximate elevation of 1050m(a.s.l) at the highest point on the main range at about midway of the route. 90km of the highway or 75% of the whole length traverses through mountainous terrain amidst densely grown rain forest. Such unique features of geographical location with forested mountain barrier at high altitudes create an ideal situation for heavy and frequent precipitation, resulting in annual average rainfall record of 3600mm. During the 1994 monsoon, 5 failed fill slopes were identified which require immediate remedial works to avoid further deterioration to the highway. The location of the five affected fill slopes are shown in Figure 2 although only two locations at km 33.7 (site 1) and km 42.8 (site 2) are discussed in this paper.
Malaysia is geographically situated in the heart of Southeast Asia monsoon belt in which high incidences of heavy, intensed and prolonged rainfall periods are fairly common. In keeping with the rapid development schemes more roads will inevitably have to be built over mountainous, rugged and rolling terrains. Combination effects of these two factors of terrain and weather gradually create slope stability problems, which require solutions entailing innovative approach in the design and construction as well as cost-effectiveness.
2
RACKGROUND OF PROJECT SITE
The 112 km East-West highway linking Jeli town in the eastern region and Gerik town in the western region, Figure 1, represents the only road connection in northern Peninsular Malaysia. Since its opening in July 1982, the highway has greatly facilitated the previous long, tiresome journey between the east and west coasts. Economically, the highway has served apart from being a trunk road, provide an infrastructure that opens up and helps accelerate development in the resources-rich but inaccessible hinterland regions of Kelantan and Terengganu states. Due to several site constraints, the present
3
GEOLOGICAL HIGHWAY
SETTING
OF
THE
General geological setting shows that the western portion of the highway alignment comprised of 943
Figure 1 Location of the East-West highway
Figure 2 Geological sequence and longitudinal profile of East-West highway
944
interbedded sequences of fine sandstone, siltstones and shales with local occurrences of tuffaceous material. These rocks which are lightly metamorphosed belong to the Baling Group of Lower Paleozoic age. Beyond Banding approximately about km 38 meta- morphic rocks are more dominant with occurrence of phyllites, quartzite and schist. As shown in Figure 2 the East-West highway runs through a variable terrain underlain by bedrock materials that have had a diverse geological history and have been subjected to tropical weathering process. (Cook, 1996) Figure 3 Subsurface profile at location 1 (site 1 )
4 SITE INVESTIGATION Detailed site investigation and engineering analysis were performed to derive for the most cost-effective remedial solution. The site investigation comprised of surface and sub-surface investigation to identify the failure mechanisms of the fill slopes. The scope of work for surface investigation involved identifying the extent of the catchment area contributing to surface and subsurface flow, assessment of the surface geology from nearby cuttings and locating points or area of seepage. For the subsurface investigation four boreholes were drilled at each location of the failed embankments. Interpretation of the subsurface profile of the two locations is as shown in Figures 3 and 4. These boreholes were carefully positioned so as to get the subsurface profile at a typical crosssection of the failed scar.
Figure 4 Subsurface profile at location 4 (site 2)
6 CAUSES OF FAILURES 5
GROUND CONDITIONS
The failed fill slope at site 1 is a partial cut and filled structure in which the crest is located immediately across a cut slope of quartzite sandstone and rocks. There was no drainage structure seen on the fill slope except for the roadside drain at both sides of the highway. Part of the roadside drain was found to be broken which allows infiltration of surface water into the fill slope. This rapid infiltration of surface water at the crest of the fill slope leads to the weakening of the underlying soils. Perched groundwater was also observed from the fill slope face located approximately half way down the fill slope. The source of this subsurface water is suspected to have originated from the slope hollow located in between the cut slopes. At site 2 the failed fill slope is situated on a small ravine indicated by a slope hollow at the upslope section. The ravine is crossing at a skewed angle approximately in the south to west direction. It is suspected that just after the construction of the
Based from the site investigation, the fill material forming the fill slope at site 1 consists of gravelly sand. The thickness of the fill layer varies from 2 to Gin with SPT (N) values ranging froin 5 to 10. Bedrock was observed at depth within 12 to 15m below the existing ground level. Groundwater level was high varying fi-om 0.8m at the crest to 7.0m towards the downslope section. For site 2, the embankment comprised of about 6 to 7m thick of fill material tapered to the original ground at the downslope section of the fill slope. Results from the borehole logs showed that the soil consists of loose to medium dense silty sandy gravel with SPT (N) values ranging from 4 to 12. The fill material used is assumed to have been originated from the cut slope material. Below the fill material the ground is stiff with average SPT value of 15. Hard layer is encountered at depths between 15m to 20m. 945
highway surface water has infiltrated and weakened the fill slope. These small failures were left unattended and as time advances creates oversteepening at the toe. The already existing oversteepened gradient of the fill slope coupled with the exceptionally intensed rainfall has triggered the formation of tension cracks on the pavement surface. The slip surface is moderately shallow and limited to within the fill.
7 CONCEPTUAL DESIGN Various design concepts for the remedial works were proposed based on results of the site investigations and interpretation on the causes of failure. The remedial works proposed were aimed to limit further the failure from affecting the highway, prevent further instability in the existing slope and improvement to the surface and subsurface drainage. For site 1, bored pile retaining wall was proposed with improvement to the existing slope by regrading to a stable gradient. The proposed gradient of the regraded slope is 1:1.5 after various options were assessed to meet the stability requirement. Initial proposal for total reconstruction of the fill slope was not feasible due to the site constraint, which made it not cost-effective. Realignment of the highway was also considered, however, this option was found to be not suitable due to the presence of existing bridge structures at both ends of the failed fill slope. Closed turfing was recommended to prevent surfacial erosion. A typical cross-section of the remedial works for site 1 is shown in Figure 5 . A series of horizontal drains were also incorporated to arrest any elevation of groundwater level. Roadside subsoil drain was also installed at the upslope section of the fill slope. For site 2 reconstruction of the embankment to a stable gradient for the top two berms followed by reinforced soil using geogrid for the bottom two berms. The height of each embankment is 6.0 metres with installation of berm drains at each intersection. Figure 6 shows a typical cross-section of the remedial works carried out at site 2, Reinforced earth was considered for the bottom slopes to enhance stability and to avoid excessive amount of earthwork construction. Subsurface drainage was incorporated by installing sand drainage blanket layer. Benching with sand between the excavated surface and fill material was also incorporated.
Figure 5 Remedial works at location 1 (site 1)
Figure 6 Remedial works at location 4 (site 2)
8 GEOTECHNICAL ANALYSIS Stability analysis was performed using a computer software incorporating modified Bishop’s method of circular failure analysis. This method was used to complete a rapid search of many surfaces for near critical failure surfaces. These analyses were performed at several intervals along the repaired fill slope. The geotechnical parameters adopted in the analysis for the remedial design are shown in Table 1. Table 1 Geotechnical parameters used for analysis Soil layer Unit Cohesion Friction weight value angle y (kN/m3) c (kN/m2) 4 (degree) Fill layer 18 5 30
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Original Ground
19
10
35
Hard laver
19
15
38
The parameters used for the fill material were based on the back analysis of the existing failed and unfailed sections of the fill slope and experience gained from other remedial works along the EastWest highway. A minimum long-term factor of safety of 1.3 was adopted. This was considered satisfactory as the soil shear strength parameters used were conservative and the failure surface was reasonably well defined from the site investigation. The fill slope profile was modelled into 3 layers as fill material, original ground and hard layer based from results of the borehole logs. The results of the geotechnical analysis are shown in Figures 7 and 8.
enable a long-term remedial solution of the fill slope. 2. Improvement of the drainage system to the fill slope was also emphasised to improve the stability of the fill slope by reducing the surface infiltration and erosion caused by rainfall. 3. Subsurface drainage in the form of sand blanket, which serves to drain seepage water through the fill material, causes a lowering of the groundwater table. 4. Remedial works using cantilevered bored piles and slope reconstruction with geogrid reinforcement was chosen for the failed fill slopes. REFERENCES
9
CONCLUSIONS
Cook, J.R. 1996. Engineering Geology of the EastWest Highway. Seminar on East-West Highway Long Term Study. Public Works Department & Public Works Institute (IKRAM).
1. Careful assessment and sufficient site investigation works in deriving the causes of failures
Hengchaovanich, D. 1984. Practical Design and Construction of Roads in mountainous terrain. Seminar on Design and Construction of Roads in Mountainous Terrain in Malaysia Geotechiiical Control Office. 1979. Geotechnical Manual for slopes. Geotechnical Control Office, Engineering Development Department, Hong Kong.
Figure 7 Results of stability analysis for site 1
Figure 8 Results of stability analysis for site 2 947
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Slope Stability Engineering, Yagi, Yamagami 23 Jiang @) 1999Balkema, Rotterdam, ISBN 90 5809 0795
Landslide controlling measures at construction sites nearby King’s palace at Narendra Nagar D. Mukherjee, K. Kishor & 0.PYadav GTE Division, Centrul Road Research Institute, New Delhi, India
ABSTRACT The Narendra Nagar area is quite famous from the historical point of view. Very old constructed King’s palace is situated here on the hill top on a vast flat grassy land. Slightly away from the palace on the same hill range many construction works such as Health Resort Complex, Kitchen area and Service Block etc.. are being carried out at different locations by a private consultant group. After starting the construction work the area has become more vulnerable and prone to fail. At seven different locations incidents of slope failure and sinking of the road have created problems not only to the moving vehicles but also endanger the construction sites from the point of view of stability. Detail landslide investigation work has therefore been conducted and different suitable remedial measures have been implemented accordingly to restore stability of affected hill slope areas. 1 INTRODUCTION
2. LANDSLIDE AT SERVICE BLOCK SITE
Present landslide affected areas under study are existing nearby the King’s Palace at Naraendra Nagar in the Garhwal Himalayan region. The Narendra Nagar area as a whole because of its geographical location experiences impact of high intensity of rain fall specially during the monsoon. The King’s Place is situated here on the hill top on a wide spread flat area. Slightly away from the palace on the same hill range many construction works such as Health Resort Complex, Kitchen area and Service station etc. are going on at different locations. After starting the construction work the area has become more vulnerable and prone to failure. At several locations incidents of slope failure and sinking of the road are frequently visible which have created not only problem to the moving vehicle but also endangered the construction sites from the point of view of stability. Moreover as the area experiences appreciable amount of rainfall , the probability of landslides will always be there and consequently will be aggravated in due course of time unless some suitable remedial measures are adopted in time. Out of seven landslide affected areas only two of them are discussed in this present paper.
The investigation of the affected area reveals that the overall nature of the hill slope material is highly susceptible to slide if proper remedial measures are not being adopted. At various places folding and fractured nature of the rock are visible. Major portion of the hill slope contains poor quality of rock formation. Because of highly jointed and weathered nature of rock, probability of ingress of rain water deeper into the hill slope is quite high at several locations. The rock formations of the overall area does not maintain its uniform homogeneous nature because of which it exists erratically as dissected fractured rock blocks entrapped within the finer matrix and are endangering the stability of the hill slope in this area. The average slope angle also varies from less than 35 degree to even up to 55 degree or more at different places. The failure mode of the hill slope depicts the direct relationship with the discharge of the surficial runoff in this area. Tilting of the trees, denuded hill slopes, sinking of the roads and the cracks developed in the collection tank indicate active nature of mass movement in this area. The present landslide affected area has covered about 1500sq.m area. Whole of the affected area is
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covered with loose debris composed of small fragments of slates and soils. Small gullies have already developed on the debris covered area at the down hill side near the toe part in a dendritic fashion,‘G’ as seen in Fig. 1. From the toe portion slightly upward approaching towards the road comes sharp vertical excavated rock face and a flat area for the future service block. The exposed rock formation here seems to be highly disturbed ,folded, crumpled and weathered. Moreover drastic change in the structural geological properties of the rock and lithology indicate that the area was under the influence of high tectonic disturbances in the geological past and a probable fault plane might be existing in the nearby vicinity. The vertical cut slope face is standing at present without any supporting wall. At this place the nala has become narrow and the filled up debris material is comparatively steeper than the debris of the uphill area above the road level. On this slope the water is flowing down through the hume pipe culvert existing just below the road. The road shows sign of sinking. A big wall is existing on the road level to retain the flow of debris materials. A catch pit has been made adjacent to the wall for collecting water flowing through the road side drain. This water is allowed to pass quickly through the pipe culvert down below the road level which is affecting the proposed site for service block and also the down hill slope area. Just above the road level within the debris slope area a rubble cement masonry check wall “R” has been made to restore the loose debris. Further upward on the debris covered hill slope another wall “W’ meets wide spread flat grassy land where King’s Palace is situated. Huge amount of rain water gets accumulated within this flat grassy land and gradually enters the hill slope. A few temporary drains have been made here to remove the accumulated rain from the vast flat grassy area. These drains are not following proper gradient and a big ditch is there within the path of the drain through which enormous amount of rain water passes deeper into the hill slope. The outlet of the drains are ending at the peripheral part of the flat area and causing series of small scale of failure nearby the road.
3. LANDSLIDE NEAR BY HEALTH RESORT
The hill slope area under study includes the site of health resort at the top of the hill and two roads on the same slope.. This area is under the grip of mass movement activity which is evidenced by the occurrence of minor slump of hill slope materials and the sinlung of both upper ‘U’ and lower ‘L’ roads 950
1. perforated tin sheet & log retaining structure; 2. Drum retaining structure; 3. Vegetation turfing; 4. Pipe culvert; 5. Existing concrete path ; 6. Surficial cemented drains; 7. Crate walls; 8. Existing retainig wall; 9. Retaining wall; 10.box drain; 11. Grassy flat land Figurel. Landslide affected area at service block respectively. Slope failures have occurred on both sides of the existing pipe culvert. The natural gully connects both roads. The general trend of slope profile seems to be of moderate nature and the slope angle varies from 30 degrees to even upto 50degrees. The tilting of the trees here indicates the movement of the hill slope materials. Huge quantity of water along with the unwanted debris materials from the construction site flows down from the uphill and has affected both the roads and slope in this area. The slumped portion of the slope nearby the pipe culvert may extend hrther and may create problem in near hture if protection is not taken at this initial stage. The failure on the left side of the culvert on upper road “U” is about 7.5m long (along the road) and 10.7m high. Road side drain is lacking in this area due to which ingress of the rain water causing considerable sinking of the road at this location. Almost identical nature of failure is also observed at the lower road ‘L’ which is situated on ridge house road exactly below the upper road ‘U’. Geomorphologically the uphill slope area here on the left side of the culvert
are more steeper than the right side. Such convex to straight type of slopes comprising of weak constituent material may tend to fail easily if gets saturated with rain water. On the uphill slope surface various tension cracks are developed which are covered with local vegetation as seen in Fig.2. Such cracks unless treated properly may create landslide in near future. On the down hill slope just below the collection tank two retaining walls RW 1 &RW2 have already been constructed as shown in Fig.2. No weep holes have been provided in these walls. The over flowing water from the collection tank will therefore saturate the filled up soil mass between the walls. Due to lack of weep holes the entrapped water will exert pressure to the retaining structures and may damage them in future. More over if the velocity of water flowing out from the pipe culvert is quite then it may cross the collection tank and directly hit the back fill materials of retaining walls from a great height and thereby cause erosion along with damage of existing structure. Although the collection tank has been made here with a purpose to reduce the impact of erosion by the falling water from the pipe culvert yet the overflowing water fiom the collection tank may also create erosion or undercutting activity at the base of the tank. The collection tank as well as the retaining structures therefore may become unstable in this area. Hence suitable measures will have to be provided here so that the draining out of water may be maintained smoothly without creating any damage to the collection tank and the retaining structure existing in this area. Here downhlll slope beyond the lower most wall contains huge amount of loose debris materials resting with steep inclination. To restore the debris intact additional remedial measures are required to be adopted here at fbrther lower level towards the downhill direction reaching upto the toe part. The road stretch is existing without any surfacing work. Sinking of the road has also seen nearby the culvert area. The road side drain is also not been provided.
4 RECOMMENDATION MEASURES
OF
Legend: A,B,C,- Sliimped areas; UdZ-upper & lower road; RWI &RW2- Retaining walls; DRI-Box drain; DR2- Angle drain; TC-Terzsion crcaks; TD- trench cum surface drain; TI &T2- toe walls; GB- Gabion wall; SC- chute & CT- collection tank. Figure2. Landslide affected areas nearby health resort
(i) The vast flat grassy land containing depression areas sporadically permit huge amount of rain water deeper into the hill slope and thereby saturates the constituent materials of the hill. It is therefore suggested that the wide spread flat area at the hill top should be reshaped either into a domal type of structure with cemented surface drain all along its periphery or it should be made inclined to a particular direction with a continuos surface drain at the lower most region of the inclined surface. This surface drain should be connected to other cross drainage so as to maintain quick disposal of rain water towards the down hill slope region through pipe culverts, as seen in Fig.l.The ditch present in the path of the drain ‘D’ and the catch pit nearby the road side ‘C’ should be plugged with cement concrete. (ii) The road side drains existing in poor condition with uneven gradient and without any cement work should be reconstructed with cement work. Both angle and rectangular drains should be provided as shown in Fig. 1 &2.
FEMEDIAL
For improving stability of the affected hill slope areas several types of suitable remedial measures have been suggested for implementation before the monsoon. 4. I Provisioii of Siri-jkial Drains
Proper drainage network is required in both the areas for quick disposal of rain water so as to reduce the increased pore water pressure developed within the hill slope materials.
4.2 Trench cuni sirrfnce drains
It is a combination of surficial and sub surficial drain by which maximum water from the hill slope can be drained out quickly. Both surficial run off as well as sub surficial water can be drained out by this type of drain. Such drains are known as surficial cum trench drains. The cross sectional diagram of the drain is shown in Fig. 3 . 95 1
There is no hard and first rule for dimensions as it may require changes according to the existing field condition. 4.3 Provision of Check walls
At various locations specially at the down hill slope regions where loose debris materials are covering a vast areas should be provided with sausage/wire crated check walls as shown in Fig.1 &2. In the small slumped areas near by the road on the cut slope face wire crated walls must be put at different locations. After construction of the crate walls the gullies should be refilled or plugged with local boulders or soil. Depending on the field conditions cheaper remedial measures such as bully check dam structure, bully crib wall, perforated tin sheet and log retaining structures and drum retaining structures can also be used. For steeper slope condition perforated drum retaining structure should be used. Whereas, other light retaining structure are suitable for gentle slope. The details of such remedial measures are described in Fig.4. Towards downhill side of the debris covered area at the toe region gabion/ sausage walls should be provided. Such wall has the advantages of being able to withstand large deformations without cracking.
Figure3, Trench cum surface drain
4.4 Vegetative tirifilig zisiqg biodegradable geogrids A major part of the slide area is devoid of vegetative cover except for a few tall trees present on the slope. As such, the slopes are experiencing erosion due to flowing water fiom rain. The common technique for preventing surface erosion is the promotion of vegetative growth on the denuded slopes. Considering the present condition of the slope, natural growth of vegetation may not take place on the slope easily. It is therefore recommended that the technique of promoting growth of vegetation with the help of biodegradable natural geogrids made up of natural fiber viz. Jute or Coir may be used. Jute or coir mesh is a netting made up of the jute or coir with square shaped opening of 2.5 cm size usually available in the roles of about 1.20m width and 50 or lOOm length.
Figure4. Perforated tin sheet and log retaining structure scale surficial slides light retaining structures such as log perforated tin sheet structures, perforated drum retaining structures and log crib walls etc. are more effective and economical. Wire crate retaining walls are highly effective, specially in loose debris covered hill slope areas. Choice of remedial measures for controlling landslides varies from place to place even in the same rock formation and it actually depends upon the existing field conditions.
CONCLUSION
ACKNOWLEDGEMENT
Use of biodegradable jute geogrid for promoting growth of vegetation on the denuded surface of the landslide affected hill slope has been found to be very effective for restoring ecological balance and improving hill slope stability. For controlling small
Authors are thankhl to Dr. P.K.Sikdar, Director, Central Road Research Institute, New Delhi for his kind permission to publish the paper.
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REFERENCE A CRRI report, 1998 Field investigation and correction techniques for improving landslide affected areas at Kinwani, Narendra nagar, Garhwal, U.P.
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Slope Stability Engineering, Yagi, Yarnagami & Jiang 0 1999 Balkerna, Rotterdam, ISBN 90 5809 079 5
Reduction of land cutting effects by the application of lightweight embankments J. Nakano, H. Miki, H. Kohashi & A. Fujii Soil Mechanics Division, Materials and Construction Department, Public Works Research Institute, Ministry of Construction, Tsukuha, Japan
ABSTRACT: This paper presents a study on feasibility and seismic stability of the embankment methods using the lightweight materials for the purpose of reducing the effect of land cutting in road construction on the steep slopes, The application of lightweight embankment methods on mountainous roads possesses the potential for meeting the various social needs, such as the conservation of natural environment in road construction, the simplification of the disaster management of roads, the progress in recycling of waste soil, the reduction of construction costs, and so on. By the approach of trial designs of cross section that is assumed to consist mainly of fill, it is proved that lightweight embankment methods have the advantage over conventional embankment methods in the cases of the steep rocky slope and the slope that has thick colluvium deposit.
1 INTRODUCTION When using general methods to construct a road on steep slopes in mountainous regions, it is still necessary to execute extensive cutting and filling works along with large slope protection works in order to obtain the designed road width. The large-scale alteration of lands caused by this extensive cutting and filling works not only threatens the natural
Figure 1.
environment, but also may create dangerous conditions that require extremely labor-intensive disaster management works following the completion of each road, especially in Japan where many disasters are caused by intensive rainfall, earthquakes etc.. Constructing a road in such a location by building embankments using lightweight materials (EPS blocks, Air-foam mixed stabilized soil, Expanded-beads mixed lightweight soil, etc.) can reduce the alteration
Concept of the reduction of the effects of land cutting by using the lightweight embankment
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(slope gradient: 1 :2, surface soil thickness: 1 m) Case 2: gentle slope made of thick colluvium deposit (slope gradient: 20°, deposit thickness: 10 m) Case 3 : steep slope made of soft rock (slope gradient: 1:1, surface soil thickness: 1 m)
of lands. This method also has the potential to lower construction costs because it cuts down on the quantity of slope stabilization works, slope protection works, and reforestation works that must be required. From these points of view, this study demonstrates the feasibility of lightweight embankment methods on mountainous roads, the type of slopes in which the application of lightweight embankment methods is advantageous, and also verifies the seismic stability of high embankment using the lightweight materials on steep slopes.
[Alternative construction methods] Type 1: Cantilever retaining wall method Type 2: Multiple anchor reinforced earth wall method (Type 1-2 are the methods using normal fill materials) Type 3: Air-foam mixed stabilized soil method Type 4: Expanded-beadsmixed lightweight soil method Type 5: EPS block method (Type 3-5 are the methods using lightweight materials)
2 SUTUDY ON FEASIBILITY OF THE LIGHTWEIGHT EMBANKMENT METHODS 2.1 Method of trial designs The trial cross section is designed to meet the assumed grade (Type 3- class 2 in “Road Structure Ordnance”, design speed: 60 km/h, width: I 1 m), by fixing the center line of the road so that it consists mainly of fill on the premise that the quantity of slope cutting will be reduced. Trial design cases are shown below. Three cases as natural ground conditions and five cases as alternative construction methods are assumed, and verified the slope stability (rotational slip) and the stability of embankment (sliding, tilting, bearing capacity). Table 1 shows the assumed mechanical properties of natural ground and fill materials.
Table 1. ProDerties of natural ground and fill material
2.2 Results of the.fiasibility study In every case, the aforementioned methods are compared from the viewpoints of stability, workability, maintenance after construction, environmental impact and construction costs, in order to demonstrate the
matural ground conditions] Case 1: gentle slope made of soft rock
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impact. Of the three lightweight embankment methods, the mixed soil methods (the air-foam mixed stabilized soil method, the expanded-beads mixed lightweight soil method) are preferable to the EPS block method because the cost of fill materials are lower and the excavated soil can be recycled for filling work.
3 STUDY ON SEISIMIC STABILITY OF THE LIGHTWEIGHT EMBANKMENT METHODS 3.1 Method of seismic trial designs
Figure 2. Examples of trial cross section design (Case 3 )
feasibility of the application of lightweight embankment methods. Table 2 presents an outline of the results. In natural ground condition Case 1 (gentle slope made of soft rock with thin surface soil), methods using normal fill materials, such as the cantilever retaining wall method, the multiple anchor reinforced earth wall method, are advantageous because of their low cost; they can be completed by executing only a small quantity of retaining wall work, cutting etc. to ensure their stability. But in ground condition Case 2 (gentle slope made of thick colluvium deposit) and in Case 3 (steep slope made of soft rock with thin surface soil), if the methods with normal fill materials are used, the retaining wall, anchors, or other work tends to be large-scale, and as the embankment load increases, large-scale excavation work or other special countermeasure work are required to ensure slope stability. In these two cases, the use of lightweight embankment methods is advantageous because it lowers the construction costs, reduces earthwork and excavation work, and can minimize the environmental
Horizontal seismic coefficients (kH= 0.15, 0.20, 0.25) are tried to act on embankments made of lightweight material (Air-foam mixed stabilized soil, Expandedbeads mixed lightweight soil, EPS blocks) with heights of 8 m or more (3 cases: H = 8 m, 11 m, 15 m), constructed on a natural ground conforming to Case 3 (steep slope made of soft rock with thin surface soil) to verify seismic stability of slope (rotational slip) and of the embankment body (sliding, tilting and bearing capacity) provided by the trial design. 3.2 Results of seismic stability study Table 3 presents the results of the seismic stability evaluation. The shaded parts of the table represent the cases and factors whose seismic stability cannot be ensured, and in these cases some countermeasure works must be implemented to guarantee seismic stability. Because embankments constructed with EPS blocks do not provide adequate stability to prevent tilting in any of the cases, it is necessary to anchor the base course to the natural ground. It is proved that in all cases using EPS blocks, the stability will be ensured
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by installing anchors with a stipulated strength at a pitch of 4 m in the longitudinal direction of the road. In the cases of air-foam mixed stabilized soil embankment higher than about 10 m, countermeasures to prevent sliding of the embankment body and to increase the friction between embankment and natural ground are required. And in the cases of expanded-beads mixed lightweight soil embankment of 15 m high, it is necessary to implement stabilization measures to prevent rotational slip of the internal parts of the embankment. Furthermore, when the mix proportion ratio of expanded-beads is particularly high, it is better to verify whether or not it is possible to count on an adequate shear strength under high earth pressure. Therefore, concerning the application of lightweight embankment on steep slopes, in the cases of low embankments, it is usually advantageous to use lightweight mixed soil methods such as the air-foam mixed stabilized soil method, the expanded-beads mixed lightweight soil method, from the viewpoint of material and work cost. However, in the cases that the embankments are higher than 10 m and that an adequate seismic resistance is required, it is possible that the use of EPS blocks would be advantageous because the use of lightweight mixed material must be accompanied by special countermeasure work to stabilize the embankment body. 4 CONCLUSIONS By the approach of trial designs of cross section that is assumed to consist mainly of fill and seismic trial designs of high embankments using lightweight materials, this study draws the following conclusions concerning the application of lightweight embankments on slopes in mountainous regions. 1. On steep slopes of soft rock and on slopes with a thick layer of colluvium deposits, it is difficult to ensure stability of vertical walls backfilled with normal fill material, and lightweight embankment methods are advantageous in such cases from an overall evaluation accounting for workability, maintenance after construction, environmental impact and construction costs. 2. There are cases where the seismic countermeasures against sliding and tilting are necessary for the lightweight mixed soil embankments higher than 1Om. 3. Concerning a comparison of lightweight embankment methods, the lightweight mixed soil methods are usually advantageous by reason of its lower construction costs, but in the cases that the embankments are higher than 10 m and that an adequate seismic resistance is required, it is possible that the use of EPS blocks would be advantageous.
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5 RECOMMENDATIONS This study on feasibility of the embankment methods using the lightweight materials was confined to a study based on trial designs of cross section, but it is still necessary to verify its feasibility based on case studies of actual road planning which includes designs of longitudinal section. In this study, the usual horizontal pseudo-static method was used for the verification of seismic resistance of lightweight embankments, but in the future it will be necessary to perform an analysis of seismic response properties of embankments using the lightweight mixed soil. Furthermore, it will be necessary to conduct a study on effective seismic countermeasure works for various lightweight fill materials and a study on the effects of the action of soil pressure on the front wall structure during an earthquake. REFERENCES EPS Development Organization 1993.2. EPS niethd Ricotosho Japan Road Association 1999.3 Manual for slope pmtectiod Highway earthwork series Japan Highway Public Corporation 1996.8. Design and execution guide for lightweight embankment metlwd using air-forni inixed lightweight soil Miki, H. 1994. Qpes mid@aims of lightwe@ embankment tnethod, fisoko v01.22 No. I0 Miki, H. NLW trend of earth structure in highway earthwork series etc.,
[email protected] No.2 Okamoto, T. & T. Inoue 1996.6 A s t u 4 on the execution of lightweight enibanhwnt using air-niilk, Kphugiho Vol.I9 Soil mechanics division PWRI, Public Works Research Center & other 14 companies 1997.3. Technical nianualfor the airfoani mixed stabilized soil method, Report of cooperative research No. 170 Soil mechanics division PWRI, Public Works Research Center & other 16 companies 1997.3. Technical n~anualfor the expanded-beads mixed lightweight soil nietliod, Report of cooperativeresearch No. 17 1 Soil mechanics division & Soil dynamics division PWRI 1992.3. Design and execution nianual for lightweigh enibanhnent method using exparded-plystyrcne, PWRI Documents No.3089
Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Relaxation effect in retaining wall on passive mode Erizal United Graduate School of Agricultural Sciences, Ehime Universib, Matsuyamu, Japan
Toshinori Sakai & Sad& Miyauch Faculty of Agriculture, Ehime Universizy, Matsuyama, Japan
ABSTRACT: Retaining wall, sheet pile, anchor and footing foundation are some of important structures and close relationship with the stability problems. The stability analyses usually use to find safety factor of those structures. In the retaining wall problems, numerous investigators have evaluated the earth pressure only on active or passive mode by separately. While in under construction, before retaining wall is operated on passive condition, the relaxation process (active condition) is acted for a certain time. In this paper attempt to explain the effect of relaxation process on the retaining wall problem. The experiments were conducted on the air-dried Toyoura sand in plane strain condition and height of sand mass was 10 cm. To make relaxation process was by pulling the wall until a certain relaxation displacement (RJ before the wall was moved toward the sand mass. The peak passive thrust and the zone of localization until & = 0.2 mm was similar. There was no influence of relaxation until Rd = 0.2 mm. under construction, the retaining wall is acted the relaxation process (active condition) before is operated on passive process for a certain time. Unfortunately, at present, the analytical or experimental studies have not been enough to explain the influence of the relaxation process on the passive earth pressure acting on the wall. In this paper, we attempt to explain the influence of relaxation process on passive earth pressure by comparing the experimental results with finite element analysis.
1. INTRODUCTION The determination of forces acting on structures, which are connected to or in direct contact with sand mass, is of paramount importance in applied geotechnical engineering. Safe and economical design of engineering structures such as retaining wall requires a sound knowledge of the active or passive stresses exerted against them. Retaining walls are frequently use to hold back the earth and maintain a difference in the elevation of the ground surface. Traditionally, civil engineers calculate the active and passive earth pressure against the wall following either Coulomb or Rankine's theory. Another popular method to estimate the earth pressure is the logarithmic-spiral method proposed by Terzaghi (194 1). There have been a number of researchers working on associated with earth pressure as Terzaghi (1932), Rowe and Peaker (1 965), Arthur and Roscoe (1 965), James and Bransby (1970), Richards and Elms (1992) and Fang et al. (1994). Nakai (1985) and Tanaka and Mori (I 997) evaluated the retaining wall problem by using finite element analysis. Until now, the researches in retaining wall problem have been evaluated separately only on active and passive mode. Actually, it is necessary to consider the effect of relaxation process during the construction of the structure related to the retaining wall. Because,
2.
TESTING APPARATUS AND ANALYTICAL METHOD
The testing apparatus consisted of soil bin, model retaining wall and driving tool. The soil bin was fabricated of steel members with inside dimension of 300 X 500 X 1000 mm (in Fig. 1). Both sidewalls of soil bin were made of 10 mm thick glass plates. The selection of the width of the soil bin was governed by the friction effect along sidewalls. Terzam (1932) suggested that the retaining wall should be twice as wide as it was high. Arthur and Roscoe (1965) reported that the side friction was not a large factor influencing the behavior of the retaining wall when wall was as wide as it was high in passive earth pressure tests.
959
located below the base of the wall serves to hold the bottom is 100 mm of steel to accommodate the entire log-spiral failure surface. The driving tool consisted of small jack and speed control system. The small jack could horizontally pull and push the model retaining wall, and speed could be controlled by automatic speed control system. In the test, the speed was 0.005 m d s e c . The tests were performed using Toyoura sand (Gs = 2.64; emm= 0.98; emln=0.61; D,,= 0.16 mm; U, = 1.46 and fines content = 0 %), and the sand mass was prepared by pouring the air-dried sand. The dry density was 1.64 - 1.65 g/cm3. The height of the sand mass above wall base (h) was 10 cm. The relaxation process was conducted by pulling the wall until a relaxation displacement (RJ of 0.0 mm (no relaxation process), 0.2 mm, 0.4 mm, 0.6 mm, 0.8 mm, 1.0 mm and 2.0 mm. The tests were performed by pushing the wall toward the sand mass (passive mode) after the relaxation was conducted (active mode). A finite element analysis, which was proposed by Tanaka (1997) was considered the shear band thickness (w) as characteristic length into a constitutive equation. The constitutive model for non-associated strain hardening-softening elastoplastic material was introduced. This model was based on the yield function of Mohr-Coulomb type and the plastic potential function of Drucker-Prager type. The element employed for the analysis was 4-noded Lagrange type element with reduced integration. Dynamic relaxation method with return mapping algorithm was applied to the integration algorithm of elasto-plastic constitutive relation including shear band effect. The finite element mesh used for this analysis is shown in Fig. 2. The input data for the analysis was based on the data obtained from the test by using air-pluviated dense Toyoura sand (Tatsuoka, 1986). The dry density (yd), residual friction angle (+4r), Poisson's ratio ( U ) and initial shear modulus (Go) were assumed to be yd = 1.64 g/cm3, @r = 34', LF 0.3, Go= 80000 kN/m2. The input data for shear band thickness (w) was 3 mm. This value was based on the data obtained from the test reported by Sakai (1997) and Erizal (1997).
Fig. 1. The testing apparatus In this experiment, tests were conducted on the ratio of height of sand mass and width of wall as 10/30. The movable retaining wall was made of aluminum with 300 mm wide, 225 mm high and 60 mm thick. Two earth pressure cells were attached on the model retaining wall to measure the distribution of earth pressure on the wall, as shown in Fig. l(c). According to the general wedge theory (Terzaghi, 1941), the passive failure surface developed in the backfill would extend below the base of the wall. As shown in Fig. l(b) the fixed bed
Fig. 2. Finite element mesh 960
3. EXPERIMENTAL RESULTS
AND
ANALYTICAL
3.1. Experimental results Fig. 3 shows the relationship between earth pressure and displacement curves at three difference R,. It is shown that the peak value of earth pressure and the
earth pressure of cells No. 1 and No. 2 acting on the wall. The values of peak passive thrust are similar until & = 0.2 mm. The values of peak passive thrust decrease with increase of R, within the range from 0.2 mm to 1.0 mm. In the range over & = 1.0 mm, the values of peak passive thrust are similar. These phenomenons can be explained by observing the zone of relaxation and shear band development
Fig. 3. Relationship between earth pressure and displacement curves displacement appeared peak earth pressure are similar at R, = 0.0 mm and 0.2 mm. The peak value of earth pressure at & = 1.0 mm are smaller than the results at R, = 0.0 mm and 0.2 mm. The earth pressure of cell No. 2 at & = 1.0 mm reaches the ultimate value without a prior peak. Fig. 4 shows the relationship between peak passive thrust (P,) and relaxation displacement (&). The passive thrust is calculated by summing the
inside sand mass as shown in Fig. 5 and Fig. 6. Fig. 5 shows the photographic representation of the relaxation zone inside sand mass at R, = 0.2 mm, 1.0 mm and 2.0 mm, respectively. It shows that the relaxation almost give no effect inside sand mass until R, = 0.2 mm. But the relaxation gives effect inside sand mass over R, = 1.O. The relaxation zone inside sand mass are similar at & = 1.0 mm and 2.0 mm. Fig. 6 shows the comparison of shear band development at & = 0.0 mm and 1.0 mm. It can be seen that the distance between the wall and the tip of shear band development reached on the sand surface at & = 1.0 mm is smaller than at & = 0.0 mm. It shows there is evidence of relaxation effect inside sand mass at R, = 1.O mm. 3.2. Analytical results
Fig. 4. Relationship between peak passive thrust and relaxation displacement
The verification of the results of triaxial compression test by the finite element method using one element (2 cm X 4 cm) are carried out employing the material properties with and without shear band. The calculated stress-strain-volume change relationship under o3= 98 kPa is shown in Fig. 7. Fig. 8 shows the relationship between total passive thrust and displacement curves for the
96 1
Fig. 5. Photographic representation of the effect relaxation inside sand mass
Fig. 6 . Photographic representation of shear band development for wall displacement = 20 cm
Fig. 7.
Simulated stress-strain volume of triaxial test
results obtained by experiments and analysis. It is shown that the results obtained by analysis are good agreement with the experiments. The relationship between peak passive thrust and relaxation displacement obtained by experiments and analysis is shown in Fig. 9. The analytical results also shows that the peak passive thrust decreases with increases of Rdwithin the range from 0.2 mm to 1.0 mm. Fig. 10 presents the contour of maximum shear strain inside sand mass obtained by analysis. These results are similar with the photographic representation of the relaxation zone inside sand mass as shown in Fig. 5. It is also shown that there is no relaxation effect inside sand mass until = 0.2 mm. For Rd = 1.0 mm and 2.0 mm, there is evident of relaxation effect inside sand mass. 962
Fig. 8. Relationship between total passive thrust and displacement curves
Fig. 9. Relationship between peak passive thrust and relaxation displacement obtained by experiments and analysis The contours of maximum shear strain in passive condition obtained by analysis can be seen in Fig. 11. It is shown that the distance between the wall and the concentrated zone reached on the surface at & = 1.0 mm is smaller than at R, = 0.0 mm. This phenomenon is similar with the observation of shear band development in the experiment. 4.
(c) Active condition, R,
= 2.0
mm
Fig. 10. Contours of shear strain obtained by analysis
DISCUSSIONS
From the experimental results, it is shown that there is no relaxation effect inside sand mass up to 0.2 mm of%. The Peak Passive thrust at R, = 0.2 mm is similar with at = 0.0 mm.
Within the range of & from 0.2 mm to 1.0 mm, the relaxation effect inside sand mass increases with increases of Rd and the peak passive thrust 963
International Conference on Soil Mechanics, Vol. 11, Montreal, Canada: 363-367. Erizal. 1997. A study of a progressive failure in retaining wall on passive mode. Master Thesis. Ehime University. Fang, Y.S., Chen, T.J. and Wu, B.F. 1994. Passive earth pressure with various wall movements. J. Geotech. Eng., ASCE. Vol. 120, No.8: 13071323. James, R.G. and Bransby, P.L. 1970. Experimental and theoretical investigations of a passive pressure problem. Geotechnique 20( 1): 17-37. Nakai, T. 1985. Finite element computations for active and passive earth pressure problems of retaining wall. Soils and Foundations 25(3): 98-1 12. Richards, R.Jr. and Elms, D.G. 1992. Seismic passive resistance of tied-back walls. J. of Geotech. Eng. Vol. 1 18(7): 996- 1011. Rowe, P.W. and Peaker, K. 1965. Passive earth pressure measurement. Geotechnique 15 (1): 57-78. Sakai, T. 1997. A study of a particle size effect of trap door problem with glass beads. Int. Symposium on deformation and progressive failure in geomechanics: 145-150. Sakai, T. and Tanaka, T. 1998. Scale effect of a shallow circular anchor in dense sand. Soils and Foundations. Vol. 38: 93-99. Tanaka, T. and Mori, H. 1997. Three-dimensional elasto-plastic finite element analysis of short pile and retaining wall. Proc. Of the I" Kazakhstan National Geotech. Conf Akmola, Vol. 1 : 32-37. Tatsuoka, F., Sakamoto, M., Kawamura, T. and Fukushima, S. 1986. Strength and deformation characteristics of sand in plane strain compression at extremely low pressures. Soils and Foundations, Vol. 26(4): 79-97. Terzaghi, K. 1941. General wedge theory of earth pressure, ASCE Tram.: 68-80. Vardoulakis, I., Graf, B. and Gudehus, G. 1981. Trap-door problem with dry sand: a statical approach based upon model test kinematics. Int. Jour. Numer. and Anal, Methods in Geomech., Vol. 5 : 57-78.
Fig. 11. Contours of maximum shear strain obtained by analysis decreases with increases of R,. Over R, = 1.0 mm, the relaxation effect inside sand mass are similar and the peak passive thrust are also observed same. 5.
CONCLUSIONS
This study evaluates the influence of the relaxation effect inside sand mass on retaining wall by comparing the experimental results with finite element analysis. The tests are performed by pushing the wall toward the sand mass after the relaxation is conducted. The conclusions from the results can be summarized as follows; 1. The calculated results by finite element analysis show good agreement with the experimental results. 2. From both analytical and experimental results, it is shown that the values of peak passive thrust are similar and no effect of relaxation inside sand mass until R, = 0.2 mm. 3. Over R, = 0.2 mm, the values of peak passive thrust decrease with increase of R, and the relaxation influences the shear band propagation inside sand mass. 4. The relaxation effect can be explained by observing the zone of relaxation and shear band development inside sand mass. REFERENCES Arthur, J. R. F. and Roscoe, K. H. 1965. An examination of the edge effect in plane-strain model earth pressure tests, Proceeding 61h 964
Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Stabilization and geoenvironmental restoration of the main central channel in the Fucino plain, Italy - A case history G.Totani, F! Monaco, M. Leopardi, A. Farroni & A. Russo Spena Fuculty of Engineering, University of L'Aquilu, Ituly
ABSTRACT: The Fucino Lake, once one of the largest lakes i n Italy (= 170 kin'), was completely drained i n 1854-76. The main central channel is still at present the most important catchment drain of the reclaimed Fuciiio plain. The banks of the channels, excavated i n soft silty-clayey soils, have been subjected i n the past to a series of sliding/erosion phenomena, which caused partial filling and reduction of the hydraulic capacity of the channel. This paper illustrates the engineering process followed for design of the remedial works, taking into account the nature of the soils, the environmental peciiliariry of the site and the high seismicity of the region. An i n situ soil improvement technique (jet-grouting) was chosen as the most suitable for the stabilization works, which enabled to restore the full hydraulic capacity of the channel and improve the stability of the banks without damaging the natural environment. 1 THE FUCINO LAND RECLAMATION: HISTORICAL BACKGROUND The Fuciiio Lake was once one of the?largest lakes i n Italy. with a surface of about 170 km-, filling a wide valley of tectonic origin. The lake was encircled by mountains 2000 to 2500 in high. It was a "closed" lake, characterized by the absence of important natural effluents. The water inflow from rivers and streams, tributary to a drainage basin of about 890 kin', was counterbalanced, i n the low water periods, by evaporation and by some percolation through the underlying fissured limestone. Being a closed lake, the level of the Fucino Lake was highly variable (about I2 m difference between maximum and minimum level). This involved large hazards for the people living along the coast, whose strong protests had already induced Julius Caesai- to investigate the possibility of reclaiming the lake by discharging the waters into the nearby Lii-i basin. This idea was given concrete form in 52 A.D., when Emperor Claudius, using 30,000 slaves and over 1 1 years of work. constructed the "Roman Tunnel" or "Effluent Claudius", 5653 in long and about 10 in' in section. This tunnel had been working until the V century A.D., when it was obstructed due to negligence and the Fucino Lake tui-ned again into a closed lake. After various and vane attempts to restore the effluent carried out over the centuries (since 1200 through 1SOO), i n 1854 Prince Alessandro Torlonia ordered the project for the construction of a new drainaae tunnel. The new tunnel, called "Torlonia Tunne?', 6283 ITI long, followed the route of the "Roman Tunnel", but the outlet elevation was considerably lower and the cross section more than
double. The tunnel was completed in 1876. Its maximum flow capacity was 50 m'/s. The reclamation works also included: - the main central channel, a catchment drain crossing the Fucino i n east-west direction for 8 kin and connecting the outlet with an expansion basin of about 24 kin' called "Bacinetto"; - two perimeter drains intercepting the high waters north and south; - the Bacinetto channel, a catchment drain 3.6 kin long continuing the main central channel and separated from this by a gate-bridge; - a network of secondary and tertiary drains flowing into the main central channel, about 260 kin total length. All the drained waters were collected i n thc Bacinetto and from here, through the gate-bridge, into the main central channel and by gravity down to the outlet. Any excess waters were retained in the Bacinetto until outflow through the effluent was possible. Beginning from I876 several drawbacks, i n contrast to the project assumptions, were highlighted. In fact, the maxiinum water flow which the effluent tunnel could discharge was only 40 in-'/s. and the Bacinetto was very frequently flooded. Furthermore, due to soil subsidence following the reclamation works (about 1.26 in in 1876) and settlements (= 30 cm) induced bp the 1915 earthquake (a terrific event which completely razed all the nearby towns and villages, causing more than 15,000 dead), the level of the free water surface above the outlet elevation was higher in large areas even for weak floods, and one effluent was no longer sufficient for the whole basin. Moreover, beginning
965
Figure 1 . Fucino land reclamation
-
Present hydraulic layout
from 19 18. the Bacinetto was completely cultivated and consequently lost its role of expansion reservoir. For this reason, in 1942 a second Fffluent was constyucted. 6240 in long, having 1 1 in- section and 20 m’/s flow capacity. In this way, the overall flow capacity increased to 60 in3/s. In 1951 the hydraulic system was further improved. The Fucino basin was subdivided into three zones (Figure 1): - Low Waters Area (27 km’ surface, 648.50 in a.s.1. minimum elevation) including the Bacinetto, surrounded by channels and small areas provided with pumping stations for mechanical discharging. - Medium Waters Area (75 km’ surface, 651 m a.s.1. minimum elevation), surrounded by a series of channels collecting the waters into the main central channel in ordinary flow conditions; in case of flood, the hydraulic level in the channels is highei- than the average ground surt‘ace elevation; by closing two gates, medium waters are allowed to flow into expansion tanks and from here to be pumped up into the main central channel. - High Waters Area (38 km’ surface, 653 in a.s.1. minimum elevation), with permanent gravity drain age. 2 GEOLOGYANDHYDROLOGY The Fucino plain, as it is today, results from the massive reclamation works carried out over the centuries, beginning from the Roman age. These works led to complete reclamation of the ancient lacustrine basin, established in the Quaternary period 966
in a large and deep morphological depression of tectonic origin formed during the Apennines orogenesis, surrounded by high mountains of carbonate Mesozoic-Caenozoic rocks. During the Middle and Upper Pleistocene, fine grained materials of variable lithological composition, originated from erosion of the nearby mountains and transported into the lake by various tributary streams, sedimented inside the basin. The upper portion of these sediments, forming the present Fucino plain, results from the last, very recent deposition phase (Figure 2). The recent lacustrine sediments are formed by predominantly fine grained soils, composed by irregular alternations of silts, clayey and/or sandy silts, silty sands and sands, in layers and lenses of variable thickness, with nearly horizontal bedding planes. The thickness of the lacustrine deposit is more than 300 in, locally even more than 1000 m. The upper =: 40 in were deposited during the late Pleistocene, the top = 5 + 6 m during the Holocene. The main central channel, which reaches a maximum depth of about 13 m (bottom elevation), was completely excavated i n this deposit. Groundwater table is present in the lacustrine deposit. The groundwater level, measured by piezometers installed in boreholes, is about 5 + 6 in higher than the elevation of the channel bottom at = 20 + 25 m distance from the channel, and close to the ground surface at -- 100 m distance. This reflects the drainage action exerted by the channel, helped by the presence of more permeable sand layers in the upper portion of the deposit. Being i n direct hydraulic connection with the channel, the groundwater level tends to vary as the water level inside the channel varies.
Figure 2. Fucino plain - Schematic geological map
3 GEOTECHNICAL CHARACTERIZATION Several site investigations, including boreholes ( 16 to 30 m depth), cone penetration tests, CPT (20 + 23 m depth) and flat dilatometer tests, DMT (20 + 23 in depth), were performed along the banks of the main central channel. Laboratory tests were run on undisturbed samples taken froin the boreholes. The typical soil profile and basic physical properties, resulting froin laboratory tests performed on samples taken at different depths and locations, are shown in Figure 3. The soil deposit is constituted predominantly by sand/silty sand layers in the upper 6 -+ 8 m and by soft clayey silt of medium plasticity with frequent, irregularly distributed sand lenses below this depth. Typical CPT profiles are also shown in Figure 3. A series of "base" minimum values of the cone resistance qc = 0.5 + 1.5 MPa has been observed in all the CPT soundings. The values of qc, tend to increase slightly and gradually with depth, varying from to 0.5 -+ 1 MPa to 1.5 i 2 MPa (undrained shear strength cl, = 25 -+ 100 kPa). Higher, remarkably variable qc. values correspond to loose sand/silty sand layers, more frequently found in the upper 6 i 8 in (i.e. above the elevation of the channel bottom, "drained" to some extent by the channel itself). Typical DMT profiles are shown in Figure 4. The DMT profiles obtained at different locations clearly reflect the marked heterogeneity of the deposit. The profiles of the horizontal stress index from DMT, K,, show that the clayey silt layer is slightly
overconsolidated, since KD values ( K D = 3 + 4) are systematically higher than 2 , indicating normal consolidation. The overconsolidation ratio inferred froin DMT according to the correlation proposed by Marchetti ( 1980) for uncemented cohesive soils is OCRDM7. = 2.5 (for K D = 3 . 3 , i.e. = 50 9% higher than the average reference value determined from oedoineter tests (OCR&,,, = 1.6). This deviation is probably related to the particular soil microfabric, due to the presence of a high carbonate CaCO; content (25 + 50 %), which gives the Fucino clay significant interparticle cementation. (Note that, due to the sedimentation conditions, the lacustrine Fucino deposit has never been subjected to significant mechanical overconsolidation). The c,, profiles determined from DMT show that c,, slightly increases with depth from = 50 to 100 kPa. These values are in agreement with the c,, values determined froin CPT and from laboratory UU triaxial tests tests (cl,k,l7 = 50 + 80 kPa). The drained shear strength parameters determined in the laboratory by CIU triaxial tests and direct shear tests are the following: angle of shearing resistance CD' = 28" i 32", cohesion (in terms of effective stress) c' = 0 to 5 +- 8 kPa. The permeability of the clay is very low (coefficient of permeability k = 3 +- 4 x 10-*cm/s). Higher values were obtained in the upper predomi iiantl y sandy layers. All the above values are in good agreement with data reported by other researchers (a detailed characterization of the Fucino clay can be found in A.G.I., 1991).
967
Figure 3. Soil profile, typical CPT profiles and physical properties
Figure 4. Typical DMT profiles
4 ENGINEERING PROBLEMS AND STABILITY ANALYSES
where the slopes are higher. As a consequence, the channel was partially filled and the maximum flow capacity was reduced to 40 m’ls. Significant excavation works had to be undertaken in order to reshape the cross section and restore the full hydraulic capacity of the channel. It was therefore necessary to evaluate the stability
After land reclamation and construction of the main central channel, a series of slidinglerosion phenomena involved the channel banks over the years, particularly in the final portion of the channel,
968
conditions of the channel banks following the excavation, and to verify if any stabilization works were required. The slopes which have been subjected in the past to sliding can be considered, at present, nearly in a limit equilibrium state (i.e. factor of safety Fs = 1). This assumption has enabled to perform a backanalysis of the slidings, based on the exact knowledge of the geometry of the slopes and on a reliable assessment of the soil profile and the groundwater table position. The back-analysis was carried out assuming a constant value of the angle of shearing resistance CD' = 30°, since both laboratory testing and available literature data (A.G.I., 199 1 ) indicate that this value is slightly variable (and, however, in a range of minor influence on the results of the stability analyses). The range of c' values determined by back-analysis (for F s = 1) is -- 5 + 7 kPa, in good agreement with the laboratory data. The results of the back-analysis, combined with the laboratory testing data, helped select the geotechnical parameters to be used in design: natural unit weight y = 18 kN/in3; angle of shearing resistance CD' = 30"; cohesion (in terms of effective stress) c' = 6 kPa. Stability analyses were subsequently carried out in order to evaluate the effects of the excavations required to restore the original profile and hydraulic capacity of the channel, in absence of any stabilization work. The factors of safety, calculated for several different sections, were generally inadequate, and even close to 1 for the highest slopes. The analyses also showed that the potential slip surfaces were relatively deep, reaching on average = 3 t 6 in depth below the toe of the slope, coinciding with the channel bottom. All the above results refer to static conditions (i.e. no seismic actions taken into account). Since the analyses indicated that the channel banks after the excavation would become unstable even in static conditions, it was realized that a severe earthquake (to be necessarily taken into account, in view of the high seisrnicity of the region) would produce for sure a series of widespread slidings along the slopes. It was therefore concluded that stabilization works were absolutely necessary i n order to prevent the channel banks from collapsing as a consequence of the excavation.
5 DESIGN OF STABILLZATION WORKS The designers were asked to plan remedial works which could fulfil1 the following requirements: - restore the full hydraulic capacity of the main central channel; - ensure the stability of the channel banks for sliding/erosion, even in case of earthquake; - preserve the existing environment, without affecting significantly the natural habitat and vegetation established over the years on the channel banks, and possibly help renaturalization of the site.
The last requirement precluded the use of two large categories of stabilization works currently in use: - works involving large excavations (retaining walls, gabions, reinforced earth, etc.); - works requiring the use of heavy and bulky equipments (diaphragm walls, sheet piles, etc.). The selection of the design solution was finally oriented towards an in situ soil improvement technique. In particular, the jet-grouting technique was identified as the most suitable in this case for the following reasons: - limited extension of the influence zone of the treatment (no damage to the existing vegetation); - practical absence of pollution; - high mechanical strength of the treated soil; - treatment feasible even at shallow depths; - light equipments; - possible inclination of the grouted columns; - possible insertion of steel reinforcement i n the grouted columns. The jet-grouting technique was used at the same time both as a retaining structure for protecting the excavation and as a stabilization treatment against general sliding of the slope. The design layout and details about the geometry and dimensions of the jetgrouting treatment (columns diameter, spacing, inclination, etc.) are shown in Figures 5 and 6. It should be noted that: - the jet-grouting treatment below the bank road level was aimed at preventing sliding caused by excavation and improving the general Factor of safety (minimum value allowed by the Italian regulations Fs = 1.3); - the jet-grouting treatment above the bank road level (along the cut) had the true i-ole of a "retaining wall". The jet-grouting technique and the layout of the stabilization works also enabled to: - reduce settlements of the bank road; - permit free seepage of groundwater through the grouted columns (sub-horizontal drains prevent/reduce pore pressures on the retaining wall also i n case of higher groundwater level); - turf and bush plant on the banks, helped by use of liydrosowing (large volumes of natural moist soil are left between one column and another). In order to verify the effectiveness of the selected design solution and to define the final layout of the jet-grou ti ng treatment, stability analyses were carried out for several sections of the channel, taking into account three different hydraulic conditions: empty channel; ordinary flow conditions; maximum flood conditions. For ordinary flow conditions, stability analyses were also carried out taking into account very severe seismic actions. In the stability analyses, the "blocks" of treated soil, formed by the jet-grouted columns and the natural soil in between, were characterized by "equivalent" average strength parameters. In all the examined cases, the factors of safety obtained were acceptable in both static (Fs = 1.4 t 1.5) and seismic conditions (Fs = 1.1 t 1.2).
969
Figure 6. Stabilization works by jet-grouting - Cross section of the channel and construction sequence
6 CONCLUDING REMARKS
main central Fucino channel is perfectly operating and the channel banks, which have been given a new stable and permanent profile, are covered with vegetation. A nature reserve has also been established in a nearby site. This case history may be considered as a prototype which could be possibly used/improved i n further applications, whenever taking into account in design the environmental features of the site is as much important as the pure engineering practice.
The case history presented in this paper is an example of environmental engineering design, involving the contribution of different specific expertises (geology, hydraulic and geotechnical engineering, historical geography, ecology, landscape architecture). In this case, the selection of an in situ soil in improvement technique (jet-grouting), combination with proper design of the layout of the stabilization works, careful planning of the construction stages and optimization of the execution techniques, has enabled to restore the hydraulic capacity of the channel and improve the stability of the banks, and at the same time to preserve the existing environment and help renaturalization of the site. At the present time, the
REFERENCES A.G.I. (Burghignoli, A. et al.) 1991. Geotechnical Characterization of Fucino clay. Proc. X ECSMFE,
Florence (Italy). Marchetti, S. 1980. In Situ Tests by Flat Dilatorneter. ASCE Jnl GED, 106-3: 299-321.
970
Slope Stability Engineering, Yagi, Yamagami & Jiang GI 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Slope stabilization in residual soils of Peru A.Carrillo-Gil Universityof Engineering, Lima, Peru
A. Carrillo-Acevedo A. Currillo Gil SA. Consulting Engineers, Lima, Peru
ABSTRACT: The main objective of this paper is to present a real case occurred in residual soils fiom the Peruvian amazon plane, in order to show the positive effects of the stabilizationwith drainage and the pore pressures dissipation that previously had originated large landslides in the season of the annual water level decrease of the Amazon River. This happens in a very short time and decreases something more than 12 meters in a fast way. This effect decreases the shearing strength of the saprolitic soil underlying, producing instability in its banks and important damages in the works of civil engineering over the surface. The results of the practiced instrumentation allowed a better planing and distribution of the drains in the affected area as well as an interpretation of the registered movement with biaxial inclinometers and the water pressures with pneumatic piezometers. All of them were associated with the extensive rains of the area the movement of the riverbed and the rapid drawdown of the water, minimizing the risk and creating’ better possibilities for fbture investments. 1-INTRODUCTION
The stability of the riverbanks in the Peruvian Amazon jungle presents a great number of technical problems not existing in other places, since in very few regions of the world are present the atmospherical, environmental or hydrological conditions that prevail in this region, adding to these factors the lack of conventional construction materials. The erosion and sedimentation phenomena that alternatively occur in both margins of the amazon rivers, and the continuous course changes between the subsequent years, present additional problems and large challenges to the application of the knowledge of the geotechnical engineering. To offer some explanation to the movement of the meanders of the Amazon river, it is present below the factors that can originate them: Soil with very low gradient and smootly sloped toward to the East, in the order of 1: 20,000, that offers greater or smaller resistance to the water flow. The changes of water level between flood and ebb times, that reach fluctuations fiom 10 to 12 meters. The tectonic movements in the Amazon zone are small, however the surface of the land bark suffers level changes, originating possible displacement in the bed of the rivers. According to what is shown previously, the Amazon river has impacted strongly on the riverbank
between the years 1948 and 1972 and on the others riverbank, between the years 1993 and 1994 , being produced phenomena of instability. Phenomena go advancing downstream initially as erosion to end afterwards as sedimentation and therefore stabilization of the slide critical area. During more than 40 years they have been producing landslides that have considerably damaged different types of engineering works placed in the banks of the Amazon river in the region of Peru, when the river impacts directly on the critical border, and increasing gradually according to the river is going far (Figure 1). 2-GEOLOGYCAL SETTING
AND
GEOTECHNICAL
The general geology considers that a large part of the Amazon region has stayed covered during the interglacial periods of the quaternary by an interior sea of shallow water when the level of the oceans had 100 meters above of the existing now (330,000 years ago) it also began to fluctuate during several glacial and interglacial periods forming terraces throughout the water courses, dropping to 100 meters below of the original level during the last Glacial Era (17,000 years ago) and remaining in 971
Figure 1.-
View of slope tipical failure in tropicals soil of Peru, Iquitos, 1994
these deep channels the large rivers, between them the Amazon river, raising afterwards to the current level (6,000 years ago). The accomplished studies establish that in the high jungle and in the limits of the low jungle are found so much igneous rocks as sedimentary, while in the low jungle prevail saprolitic soils originated by the sedimentary rocks of the terciary and quaternary and they are formed mainly by sandstones, shales and clays. The general description of the geomorphology of the Amazon region indicates that the low jungle is substantially flat and as said remain, its height varies between 80 to 400 meters above mean sea level. Due to this small difference of elevation the rivers flow slowly, getting in the dry station the apperance of lakes. This region of the Amazone plain, can be indicated as advanced erosion type. The Amazon plain is characterize by its great humidity and soil covered by a dense tropical vegetation. 3- SLOPE FAILURE MECHANTSM
The statistical analysis of the movement of the Amazon River performed between the years 1991 and 1996 clearly established that the landslides have occurred during the stage of water level decrease in the river. This is different in instead of what occur in other parts of the world where the rains that are presented during the decrease of the level of the water of the those which river originate the
landslides. We consider that as a phenomenon of rapid drawdown that affects the bank, because of the water level decreases to an average of 12 meters in a very short time. This rapid drawdown is interpreted as a process that increases the undrained deformation of the saturated zone in the affected banks. In other words, the reaction of the stability of the banks to the rapid movement when the water level decreases is similar to the response occurred in an open cut in which is produced a forced alleviation, due to material that previously was offered as lateral support and that was suddenly removed. In this case, as a consequence of the imbalance produced by the rapid drawdown of the river, there is water that remains within the porous structure of the soil, since its level does not decrease to the same speed that the water level. This phenomenon causes an increase in the weight of the bank body, as in the pore pressure with the soil. This effect reduce the shearing strength of the soil, which, together with the effects of the river, causes the ladslides ( if it has not been possible to evacuate the water tricked within the soil of the bank).
4-LANIDSLDES CONTROL MEASURES The system of installed deep drainage is efficient and it has generated an adequate drainage during the critical stage of drawdown of the Amazon river in 1996, 1997 and 1998. In the better behavior area we 972
Figure 2.-
Measured Horizontal Displacements
put 31 horizontal drains of 30 meters of length, spaced each 3 meters with a slope of 3" and diameter of 4". In the adjacent section we installed wells with radial drains that arrived to lengths understood between 15 to 25 meters The measures analyzed indicate a small displacement in direction to the Amazon river in the stage of water level decrease, and backward displacement when the water level rose The comparison of the results obtained demonstrate that he movements registered before have reduced considerably, probably due to the effective operation of the deep drainage system, and the additional effects produced by the sedimentation that originated due to movement of the riverbed of the Amazon river. 973
The results of the final piezometrics measures indicate that, as a rule, the dissipation of the pore pressures in almost all cases has been effected in correspondencewith the decrease and increase of the water level . So, we found a good behavior in the drainage system installed in the critical zones. The piezometers that were installed in the zone of the last landslide (1994) from the beginning of their readings showed irregularities with respect to the dissipation of the accrued pore pressures &er of the decrease of the river. It must be noted that in the location zone of these instruments was not practiced nor any deep drainage system or treatment for maintenance .
Figure 3 .-
Cross Section With Piezometrics
5- CONCLUSIONS The deep drainage system, insta lled in the studied area (by means of wells with radial drains as well as by horizontal drains) has contributed effectively in the stabilization of these banks, and the analysis of all the measures taken during the several months of work with the instruments, prove that there is a substantial improvement in the stability conditions of the platforms included in the study, conditions that can improve in the hture due to more sedimentation 974
that presumably could be produced in the place by effect of the change of the Amazons riverbed. The results shown in this paper provide a global vision of the stability problems of soils in the Peruvian wet tropic, generated by the changing morphology of the rivers that originate important risk situations in some cases, and increasingly growing stability in others that permits to establish the development of new behavior standards for the riverbanks of the Peruvian Amazon that in the hture can be predictable with certain aproximation
considering their evolution in the geological time of hundreds of years, since now in certain areas it has already passed the danger, and maybe within 100 or more years, the problem return to be present and the safety factors of the banks decrease gradually until to become unstable and to produce large landslides as they occurred in sites and dates of study, considering finally that the Peruvian Amzon is located in a region of a very singular world in light of their geotechnical occurrences and of climate that create very difficult wet tropical soils to predict and handle in the construction of the earth works. ACKNOWLEDGMENT The permission of The Maritime Authority of Peru (ENAPU-PERU S.A.) to publish this paper is gratehlly acknowledged REFERENCES A. CARRILLO-GIL, S.A.,Consulting Enginers, 1998, Stability Control of the Riverbanks in Iquitos, Peru, Technical Report to ENAPU S.A.,Lima, Peru. CARRILLO-GIL,, A., 1978, Characteristic of Tropical Soils in Peru, Latin American Magazine of Geotecnique, Vol. IV, NO4 pp. 207-216, Caracas Venezuela (in Spanish). CARRZLLO-GIL, A., 1983, Stability Problems in Iquitos, Peru, Proc. WI Pan-American Conference on Soil Mechanics and Foundation Engineering, Vancouver -Canada. CARRELO-GIL,A.,CARRILLO,E.,CARDENAS, J.,ROBALINO M.,( 1994), Characterization of Tropical Soils of Peru, X National Congress of Civil Engineering, Lima, Peru (in Spanish). CARRILLO-GIL, A.,CARFULLO, E.,CARDENAS, J.,1995, Properties of the Peruvian tropical soils, Proc. X Pan-American Conference on Soil Mech. and Foundation Eng., Guadalajara, Mexico( in Spanish). CARRILLO-GIL,A.,DOMINGUEZ,E., 1996, Failures in Amazon riverbanks, Iquitos, Peru. Seventh International Symposium on Landslides, Trondheim, Norway. CARRILLO-GIL, A., 1997, Peculiarities of tropical saprolitic soils of Peru, XIF' International Conference on Soil Mechanics and Foundation Engineering, Hamburg, Germany. CARRILLO-GIL,A., 1998, Analysis and Design in the Tropical Soils of Peru, WII GEO'LIMA '98,Lima,Peru.
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Slope Stability Engineering, Yagi, Yamagami& Jiang 0 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Case study of a cut slope failure in diatom earth A.Yashima & H. Shigematsu Gifu Universi@,Japan
S.Okuzono Kyushu Industrial Universi@,Fukuoka, Japan
M. Nishio Japan Highway Public Corporation, Japan
ABSTRACT: A cut slope failure happened in a diatom earth during the road construction at Takasu in Gifu Prefecture, Japan. An undisturbed sample was taken with the block sampling technique to investigate the cause of the failure. From the experimental findings, it is found that the soil structure can be easily destroyed by a small disturbance. Once the soil structure is destroyed, the sediments behave in a liquid manner due to a high water content. The trial field drainage tests were carried out to find out the optimum drainage pipe length for the countermeasure against slope instability. The excavation of the diatom earth slope has been successfully conducted with the proposed pattern of drainage pipes, stabilizing piles and surface replacement by the improved soil.
1 INTRODUCTION
In north mountain area of Gifu prefecture, Japan, diatom earth is widely distributed. It is a 1a.custrine sediment deposit during the la,te Pliocene to Pleistocene epoch. A high factor of safety of the cut slope was originally assessed due to high unconfined compressive strength of the raw material. During the road construction work, however, the slope failure occurred as shown in Fig.1. To investigate the cause of the failure, site investigations using Swedish weight sounding and borings were carried out firstly. The investigations revealed that there is an existing slide surface in the slope. Thin seam of soft and wet material was found out along the existing slide surface. Ohmori et al. (1998) investigated the mechanical properties of clay seam along the sliding surface to understand failure mechanism of soft rock slopes. They found an implication related to the choice of strength of clay seams and ground water in the slope stability analysis. Then, an undisturbed block sample of diatom earth was taken from shallow depth at the site close to the slide surface. Laboratory tests on the sample were carried out to understand the mechanical properties on raw and disturbed diatom earth. The laboratory test program consists of liquid limit test, unconfined compression test, triaxial compression test, isotropic consolidation test, soaking test a.nd microscopic observation through the scanning electron microscope. 977
From the experimental results, it is concluded that the soil structure can be easily destroyed by a mechanical disturbance. Once the soil structure is destroyed, the sediments behave like a liquid b e cause of the high water content. Three landslide countermeasures, preventing piling, replacement and drainage pipe were studied. The trial field drainage tests were carried out to obtain the optimum drainage pipe length for the countermeasure against slope instability. The excavation of the diatom earth slope has been successfully conducted with the proposed pattern of drainage pipes, stabilizing piles and surface replacement by the improved soil. 2 PROPERTIES OF DIATOM EARTH 2.1 Liquid limit
An undisturbed block sample of diatom earth was taken in the vicinity of the slide surface. The sampling site is shown in Fig.2. The liquid limit was firstly obtained for the material passed through 0.42 mm sieve by a putty knife. To understand the influence of disturbance of the diatom earth on the physical property, the liquid limit of the specimen passed once through the sieve is compared with that of the specimens passed three times and five times as well as the specimen ground into powder. The liquid limit chart for diatom earth with different disturbance history is summarized in Fig.3.
Figure 1. Slip line in the cut slope a t road construction site.
Figure 3. Liquid limit. chart of diatom earth passed through 0.42mm sieve by a putty knife.
Table 1. Mechanical and physical properties of Takasu diatom earth. unconfined compressive strength(kPa) q, pre-consolidation pressure (kPa) pc compression index C, swelling index C, natural water content (%) w, liquid limit (%) w,. *:passed once through 0.42mm sieve
Figure 2. Sampling site of Takasu diatom earth
The liquid limit on the diatom earth is found to be much lower than the natural water content, as shown in Table 1. The liquid limit of the fully disturbed diatom earth is surprisingly low, implying that once the soil structure of the diatom earth is destroyed) the material behaves like a liquid. Therefore, the mecha.nica1property of the existing seam in the slope is considered to be very sensitive to the change of water content. The microstructures of diatom earth were photographed through the scanning electron micrcscope (SEM). SEM micrographs of raw sample,
384 720 2.94 0.15 205 153-
passed once through 0.42 mm sieve and ground into powder are shown in Photo.l. Micro diatoms with an extremely large void are observed in the raw sample. The existence of the large void is one of the main reasons why the diatom earth has a very high natural water content. On the other hand, once the diatom earth is ground into powder, only small particles with relatively small void can be observed, as shown in Photo.l(c). This reduction in void space due to the mechanical disturbance corresponds to the significant reduction in liquid limit of the diatom earth by grinding into powder. 2.2 Slaking property
There were many eroded gullies observed on the slope surface after the exmvation. The dry-wet cycle was considered to deteriorate the micrG structure of the diatom earth. To investigate the slaking property of the diatom earth, soaking tests on raw and air-dried samples were carried out. In Photo.2, a remarkable destruction of the structure on the air-dried sample caa be seen. Once the diatom earth is air-dried, the micrestructure can be easily destroyed by wetting(Maekawa and 978
Photo 2. Slaking properties of Takasu &atom earth (a)raw sample a n d ($)air dried sample.
Mechanical properties of diatom earth stabilized by lime and cement were reported by Tateishi et al. (1992). The stabilized diatom earth was found to have a strong micrestructure and behave in a brittle manner. 2.3 Mechanical property
Photo 1. SEM micrographs of diatom earth (a)raw sample, @)passed once through 0.42nini sieve and (c)ground.
Miyakita, 1991). On the other hand, a raw sample did not change its structure even after one month in the water. From this finding, it is found tha-t t o prevent the progressive surface erosion of the slope, a quick surface treatment is necessary after the excavation. For this purpose, a surface replacement by an improved soil was carried out at the construction site.
Unconfined compression tests on the sample were carried out and the typical experimental results are shown in Fig.4. The strain where a peak strength occurs is about 3 % for the diatom earth, being much larger than that of Ja.panese sensitive alluvial and Pleistocene clays investigated by Yashima et a1.(1998). A sharp reduction in the strength asfter the peak strength is observed. The failure state of the specimen is shown in Photo.3. Vertical cracks can be seen in the specimen. From the stress-strain relation and failure state, it is found tha.t the diatom earth is a brittle material with high unconfined compressive strength. If the unconfined compressive strength of 400 kPa is used for the slope stability analysis, the calculated factor of safety is more than 10. Consolidated-undrained compression tests on normally consolidated and overconsolidated samples were carried out. The stress-strain and pore water pressure-strain relations and effective stress paths are shown in Fig.5. The larger the confining pressure is, the less the reduction in the strength after the peak strength will be. The undisturbed samples have a rather high compressive strength.
979
Figure 4. Experimental results of unconfined compression test.
Photo 3. Failure state (a)side a n d (b)top. Figure 5 . Experimental results of undisturbed diatom earth (a)stress - strain relations, @)pore water pressure - strain relations a n d (c)effective stress paths.
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Figure 6. R m e history of &charge froin drainage pipes with different length.
Figure 7. Slope profile with three countermeasures; drainage pipes, stabilizing piles and surface replacement by the improved soil.
From unconfined and triaxial compression tests on the raw samples, it is found that if the diatom earth is kept undisturbed, the strength is rather high and high factor of safety of the cut slope can be guaranteed (Nagaraj et al., 1998). On the other hand, if the diatom ea,rth is mechanically disturbed or air-dried and wetted, the microstructure can be easily destroyed and the material behaves like a liquid.
3 COUNTERMEASURES AGAINST SLOPE INSTA BI LI T Y The driving forces that cause the failure of a cut
981
slope arise from the own weight of the diatom earth as well as from the water pressure actiiig in the existing sliding seam. Based on the laboratory experiments and field observations, the philosophy and procedure for the countermeasures against the slope instability can be summarized as fol1ows:drainage pipes, stabilizing piles and surface replacement by the improved soil. To withstand the driving force due to own weight of the cut slope, the stabilizing piles were first installed near the toe of the slope. Then a surface replace ment was conducted with a cement-lime mixed diatom earth to prevent the progressive erosion of the slope surface. In order to lower the groundwater table and r e
Koba, T. 1992. Diatom earth stabilization by lime and cement. Proc. 27th JSMFE annual meeting, pp.2377-2380.
duce the pore water pressure in the existing seam, the installation of the drainage pipes were planned. The trial field drainage tests with different pipe lengths were carried out at two neighboriiig cut slopes. Time histories of the discharge from the drainage pipes with different length were monitored at two cut slopes and the test results are summarized in Fig.6. From the figure, it is found that the amount of drained water through pipes with length of 10 m and 15 m was not sufficient, while the discharge through pipes with length of 20 m, 25 m and 35 m are found to be almost same and sufficient. Based on the test, the optimum length of 20 m for the drainage pipe was determined. Further excavation of the diatom earth slope has been successfully conducted using the countermeasures such as pattern of drainage pipes with the length of 20 m, stabilizing piles near the toe of the slope and surface replacement by the improved soil. The slope profile with three countermeasures is shown in Fig.7. 4 CONCLUDING REMARKS Slope failure along the existing seam is considered to be triggered by the excavation, surface erosion and the rise of ground water table at the road construction site in Gifu Prefecture, Japan. To design the countermeasure against the slope instability for the further excava,tion, laboratory tests and trial field drainage tests were conducted. From the experimental evidences, three measures were determined such as pattern of drainage pipes, stabilizing piles a.nd surface replacement by the improved soil. The further excavation of the diatom earth slope has been successfully conducted using the proposed countermeasures. REFERENCES Nagaraj,T.S., Onitsuka,K., Tateishi,Y. and Hong,Z. 1998. Is diatom earth a collapsible ma,terial? Proc. Int. Sympo. o n Problematic Soils, Sendai, pp .257-2 60. Ohmori,K., Ohta,H., Hirose,T., YasutaniJ. and Tazaki,K. 1998. Strength and mineral composition of clay seams along the sliding surface. Proc. Int. Sympo. o n Problematic Soils, Sendai, pp.633-636. Yashima,A., Sh ig e ma tq H . and Oka,F. 1998. Effect of internal structure as related to geotechnical properties of Osaka Pleistocene clay. Proc. Int. Sympo. o n Problematic Soils, Sendai, pp.571-574. Maekawa, H. and Miyakita, K. 1991. Effect of repetition of drying and wetting on mechanical characteristics of a diatomaceous mud-stone. Soils and Foundations, Vo1.31, No.2, pp. 117-133. Tateishi, Y., Onitsuka, K., Yoshitake, S. and 982
9 Stability of reinforced slopes
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Slope Stability Engineering, Yagi, Yamagami L? Jiang 0 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Centrifuge model testing of reinforced soil slopes in the perspective of Kanto Loam G.Pokharel, A. Fujii & H. Miki Soil Mechanics Division, Public Works Research Institute, Tsukuba, Japan
ABSTRACT: A series of unreinforced and reinforced soil slope model tests were conducted on a representative problematic soil (Kanto Loam). In this paper, the centrifugal loading test results observed at the ultimate failure-state are presented. The primary objective of these tests were to identify the effectiveness and applicability of anchor plate attached soil nailing method in the stabilization of natural and cut slopes especially made of the problematic soils which losses friction as the water content increases. The centrifuge model tests illustrated that the anchor plates are not only effective in increasing the safety factor, but also in reducing the settlement. Therefore, the proposed method exhibits wide applicability. Further studies are recommended to investigate the applicability and limitations of the method in varying drainage conditions. 1 GENERAL The problematic soil is usually described as special soil that behaves completely different upon change in water content and its behavior does not fit in the conventional theories applicable to most of the widely available soil types. Residual soils, Volcanic soils, Collapsible, Loess soils, Kanto Loam, etc. are described as special soils that need to be treated differently and sometime most of these soils are referred as problematic soils (Agha et al. 1991). The usual standard design parameters are not enough in designing the reinforced soil structures on aforesaid problematic soils and other design parameters should be identified, e.g. drainage characteristics, swelling behavior with the water content, etc. In the present research, the effectiveness of soil nailing method for the stabilization of natural soil slopes made of Kanto Loam is discussed through the centrifuge model testing. A n alternative method of reducing the cost, length and number of soil nails has been proposed by attaching anchor plate at the embedded end of soil nails. The effectiveness of the anchor-plates was illustrated by comparing the failure mechanisms of the unreinforced and reinforced soil slopes with and without anchor plates. The model test exhibits promising results. Meanwhile, the results also illustrated the effectiveness of the centrifuge machine in studying the failure mechanism of natural and cut slopes under gravity loading. Overall, this paper presents these perspectives of nailed soil slopes in detail and recommendations for further studies have been also made at the end.
2 REINFORCED SLOPE MODEL TESTING 2.1 Model testing scheme In order to propose a new soil nailing technique appropriate to specific problematic soils, typical representative problematic soil types should be identified first as mentioned in the previous chapter. In this paper, Kanto Loam was selected as a representative problematic soil among the most commonly available soil types in Japan, which are considered to be problematic soils. The model test results concerning Kanto Loam will be described in detail. In this series of model testing scheme, the four types of models (Table 1) were identified in order to assess the effectiveness of major components, e.g. soil nails, number of nails, facing and anchor plates. Figure 1 illustrates a typical schematic longitudinal sectional view of the model slopes, its X-sectional
985
(a) Longitudinal Sectional Details (b) X-sectional View (c) Panel-Nail-Anchor-Plate Connections Fig. 1Schematic view of model slopes reinforced with anchor plates attached soil nails
slopes was fixed to yd=0.65 tf/m' which is quite closer to the values mentioned in the JGS Soil Testing Manual. The unreinforced model slope having water content, w,,=lOO% failed interestingly at 20g acceleration and others failed at F=40g (Case 1~ ~ 9 0 % and) F=14g(Case 3 w-110%) loading. The remaining tests were decided to conduct at the water content of w=lOO%. Conventional laboratory tests on soil sample to determine basic engineering properties and soil nail pullout tests on the proposed soil type were respectively carried out. The cohesion and angle of internal friction for the soil (c-@) were determined through triaxial tests (UU), and found as c=l.ltf/m* and @=3.7" (degrees), respectively. The schematic view of the pullout test mould is illustrated in Figure 2a, and the pullout test models for linear bars and planar reinforcements are qualitatively compared. The pullout test results are presented in Table 2. Pull out tests were conducted with varying confining pressures, but, the confining pressure did not show
view and the panel-nail-anchor plate connections. The various sizes of the model slopes were indirectly controlled by the specifications of the centrifuge machine that was used in this modeltesting program. The centrifuge machine has standard sample box of size 500x400x130mm. The selection of model slope size and shapes has to take into account of the size of the sample box. The inclination of the model slope face was assumed to be 1V: 0.2H, and was decided first based on the assumption that the majority of soil nailing work is carried out on natural and cut slopes with steep face. The other reason is due to the size of sample box and possibly large failure surface for the highly plastic Kanto Loam. This will not only free the unreinforced slope from the effect of rear rigid boundary but also the reinforced slopes because of the enough distance between embedded end of the soil nails and the rear rigid boundary. The model-testing scheme under present research program consists of four series of work division, as follows: a. Determining the basic engineering properties of the soil type. The Kanto Loam is highly plastic clay and its plasticity varies greatly with the change in moisture content. Therefore, the first work in this stage was to decide moisture content so that rest of the tests could be conducted at a single moisture content level. The moisture content at which the model slope fails (equivalently 4 . 8 m high slope at Fs=l.O in l g gravity loading) at 20g was assumed to be the model testing moisture content for the remaining model slopes. A set of trial unreinforced model slopes were made with three different water content -9096, =loo% and -110% around the average natural moisture content of Kanto Loam (w,-100%) and loaded in centrifuge until failure. The dry density of the soil mass in these model
Figure 2 Schematic view of pullout test models. 986
any significant changes, and it should be attributed to the very small angle of internal friction, 4. The pull out test data shown in Table 2 is for the 100mm embedded length of the 5mm-diameter soil nails (at a vertical pressure of 5tf/m2). For anchor plate attached nails, the nail length did not show significant effect on ultimate pullout load when the tests were carried out on 10cm, 15cm, and 20 cm long bars. The details of the bar size and idealizations are discussed in the next paragraphs. Table 2 Ultimate pullout load for lOOmm long nails. Type of Soil Nail
Ultimate Pullout load
Sand coated soil nails Anchor plate attached soil nails
0.168 kg/cm length
17.5 kg/nail
b. Design of soil nail configuration: bar size, length and spacing, in order to maintain safety factor of 1.2 (equivalently 6m high slope) under static loading condition. This stage utilizes the data from the step 1. The size and spacing of the nails were configured based on the pullout tests conducted under the same soil conditions (i.e. the same dry density and water content). The design was initially expected to be based on the two-wedge method, and the stability was to be confirmed by Modified Bishop's Method. But, the very low internal angle of friction of the Kanto Loam (at w=lOO%) and the self-weight loading alone made the search of critical two-wedge failure mechanism impossible or tending to the slip circle failure mechanism almost similar to the failure mechanism predicted by the Modified Bishop's Method. The failure surface predicted using the UU test results was found very large surface compared to the model size and use of long nails might get influence of the rear side boundary (Fig.1). Thus, the anchor plate attached nails were designed to satisfy the safety factor of 1.2 and the model reinforced with nails without any anchor plate attachment was expected to serve a comparison. c. Centrifuge model testing of the reinforced slope. Slope models prepared based on the configuration proposed on stage 2, were tested in the centrifuge machine. Thus, the successive chapter examines two aspects: first the conventional design methods and its applicability, and then, further examines the applicability and effectiveness of the soil nails. Based on these assessments, a new approach is proposed to suit these special soil types and the proposed enhancements will be expected to apply in the stabilization of natural and cut slopes.
2.2 Model slope preparation and testing The basic components in conventional soil nailing methods (Fig. 1) usually consist of three basic components: (a) soil 'nail (b) grouting around soil nails and (c) facing material. Sand coated metallic reinforcing bars of 5mm diameter were used and 8mm thick acrylic plates were used as facing panels. Five panels were placed vertically and each panel has surface area of 48 mm x 130 mm. The 130mm side was on z-axis and equal to the thickness of the models in z-direction. The facing panel and reinforcing bar connections were nut bolt type and rigidity was increased by applying strong glue on both sides of panel. Similarly, the anchor plate attached nails also had nut bold joint behind the anchor plate and the front of the plate has flushed surface. All these connections are rigid type which means the joints on both side anchor plate or facing panel connected ends were not supposed to rotate. The both surfaces of the sample box along z-axis were coated with fine grease in order to satisfy the plain strain condition along z-axis. Grid was also made on the front surface in order to take photograph at different acceleration levels and make it easy to trace the deformation of soil mass with respect to the grid. Red colored fine sand lining inside the soil allowed the observation of failure surfaces and trace it to paper as shown in Figure 3. The soil in the model apparatus was compacted in such a way that the dry density of yd=0.65 tf/m3 is achieved. To ensure the uniformity of the soil density, the soil mass was divided into 10 layers i.e. 2 layers over each panel height and soil weight was computed to the respective layer (Fig. 1). Thus, the height of each fill was 23.5 mm and dry density was assumed to be accurate enough and practically acceptable. The first trial run with the same soil density and water content in foundation soil showed an effect of the foundation stiffness and the failure surface passed through the bottom of the toe and bulging was also observed. Then the foundation soil was made of strong material and the failure surface was forced to pass through the toe of the slope. This is a usual approach to force the critical slip surface to pass through the toe of the slope in the design of nailed soil structures. The bottom surface of the panel was wedge shaped and a vertical cut was made in order to avoid the resistance against sliding of the slope face. It is because the contribution of panel resistance is not directly accounted in the conventional design and analysis methods. Dry sand was loosely filled in this vertical cut to increase the workability while preparing the slope models. As early as the model preparation was completed,
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Figure 3. Slip surface and deformations observed at the end of model testing. the models were immediately installed in the centrifuge machine and centrifugal loading was carried out. This was due to the high water content in the soil mass and delay might cause the drying of the soil surface and the result may not represent actual assumptions made in the idealizations.
3. RESULTS AND DISCUSSIONS The four model slopes (ref. Table 1) were tested in this series of model testing. The set consists of (1) unreinforced soil slope (Fig.3a). (2) Reinforced with sand-coated two bars per layer (Fig.3b). (3) Reinforced with sand-coated bar: single bar per unit facing panel (Fig.3~)and (4) Reinforced with 20mm square anchor plate attached soil nails with panel facing (Fig. 3.d). As the primary objective of current research was to investigate the effectiveness of reinforcing bars, the deformations only at the ultimate failure mode are presented in Figure 3. The unreinforced slope (Fig. 3a) failed at F=20g. The second model slope (Case 2) failed at F=24g. The reinforced slope with single bar per facing panel with (Case 3) and without (Case 4) anchor-plate failed at F=22g and F=30g, respectively. The unreinforced and reinforced slopes without panel facing (Case 1 & 2) failed due to the slip failure 988
close to the slope face. The increase in the number of nails did not produce a significant improvement, and it should be attributed to the lower pullout capacity of bar due to the negligible angle of internal friction. The slope with anchor plate on the embedded end (Case 4) first showed a crack behind the rear end of the top most soil nail and the successive slip surfaces were seen towards the slope face. The top surface showed a very high settlement in the case of reinforced slope without anchor plate compared to the anchor plate attached case. This verifies the effectiveness of anchor plate not only in increasing the failure load, but also in reducing the settlement. 4 CONCLUSIONS Promising features of the proposed anchor plate attached soil nailing technique were illustrated through a series of centrifuge model tests. The anchor plates are effective not only in increasing the safety factor, but also in reducing the settlement and therefore, exhibits wide applicability of the method in the stabilization of natural and cut slopes especially made of the problematic soils which losses friction as the water content increases. Further research is essential to investigate the effect of
drainage condition. Meanwhile, the centrifuge model test is a cost-effective method in studying the failure mechanism of soil slopes under self-weight loading. REFERENCES Agha, A., RK Katti & N. Phien-wej 1991. Problematic soils and their engineering behavior, Proc. gth ARC on SMFE, Bangkok: 223-253, Rotterdam: Balkema
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Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Dynamic behavior of vertical geogrid-reinforced soil during earthquake A.Takahashi, J.Takemura & J. Izawa Tokyo Institute of Technology,Jupun
ABSTRACT: Most of seismic design codes of geogrid-reinforced stri.ictures are based on pseudo-static limit equilibriiim approaches. However, in the 1995 Hyogoken-nambu Earthqiiake, geogricl-rc3infort:ed stxuctures were not damaged serioiisly. It implies that the importance of the permanent clisplacement as a design criteria in the evaluation of the seismic performance of geogrid-reinforcod structures. In this study, centrifuge model tests were conducted to study the dynamic behavior of geogrid-reinforced soil during earthquake. An attempt was made to discuss the effects of length and spacing of geogrids 011 the perinaiient deforinatioii of the reinforced soil.
1 INTR,ODUCTIO& Iii the 1995 Hyogolteri-ria.iribuEa.rtliqua.ke, a iiuiiiber of geogrid-reinforced retaining ~vallsperformed well, compa.red with a.ny other t,ypes of ret,airiing wa.lls (Ta.t,suoltaet al. 1996). Although soiiie tlisIhceriieiits were observed in such wa.lls?iio cat,a.strophic failure took pla.ce even for the seisiriicity greater t1ia.n the design value. This inea.nt the iiiiportmice of periiia.nent displaceineiit iii the seisiiiic clesigri of geogrid-reiiiforee(1 soil st,ruct,ure.
In order t o gain insight irit,o trhe belia.vior of geogrid-reinforced soil struchres, a series of ceiitrifuge model tests were performed by tlie authors (Taka.liaslii et a.1. 1998). R,esult,s of the tests for st,eep embarikriieiits with the slope angle of GO degrees showed that, t lie perniaiieiit deforiiia.tion riiode of t,lie reinforced ernbaiikmeiit varied with tlie leiigth of geogricl. By cornparing two emba.iikiiieiits with same tot,al pla.c:erneiit,leiigt,li of geogricl but, differeiit leiigtli a.iid spa.cing, it nras fouiid that, cleforriia.tioiiiii tile eiiibariltriieiit with longer leiigth aiid larger spaciiig becairie sriialler tliaii that with shorter length a.nd sinaller spacing. 111 this st,ucly, ceiitrifuge model tests were carried out oil \:ertica.l geogrid-reinforced soil. Effects of leiigtli a.iicl spaciiig of geogricls oil the seismic perforiiiaiice of tlie reiiiforced soil were discussed. especially for t,lie periiiaiient cleforiuatioii. 2 2.1
TEST
PROCEDURE .4SD COSDITIONS
Test 1jroc:"dure
Geogecliriical ceiitrifuge usecl iii the tests was T. I. T . hIa.rl< I1 Ceiitrifiige (Takeiiiura et al. 1989). AIodel setup used is slion.11 iii Figure 1. Iiiagi saiicl with drj' (delisit!. of I.L~:I~//TL'' aiid water coiiteiit of 27%, I \ ~ ~ IIISCYI S for iiiakiiig the irioclel gro~iiid. Bi\sic. propert,ies of Iiiiigi saiicl are g i \ m iii Tal.)le 1. Fr ic tic )ii i i i i gle . (1 iws o1.) t,ai 1i e(1 fr oi 11 t,I ia xi ii 1 ( Y ) i 1ipressioii tests iiiider driiiiiecl coiiditic.)ii. Coliesioii. I ' as bi~ck-cillclilate11f r ~ i i ithe failwe lieiglit OIJservecl iii a ceiitrifuge test, oii iic:)ii-reiiif(~rced1.ert.ical slope. hlodel geogritls used ill t>lietests was a 991
Table 1. Material properties of Inagi sand 2.66 SDecific eravitv Mean grain size D50 0.2mm Uniformity coefficient U, 3.2 4.2kPa Cohesion c* Internal friction angle q Y 33deg. * pd = 14kN/m", w = 27% "
I
I
Table 2. Material properties of model geogrid
5 8.0(4.0 x 1 0 2 ) k N / m Tensile strength T f Elongation - ..&UYO at, break E f Thickness 0.2(10)mm in pa,rentheses, prototype scale
Figure 2. Schema.tic view of inodel facing plates
Figure 3. Process of riiodel preparatioii
glass fiber made fly-guaxd, of' whicli properties a.re listed in Table 2. In order to avoid t,he local fa.ilure at the vert.ica1face in the wa.11, a.lurninuni made fa.ciiig phtes were adopted ils shonm in Figures 1 arid 2. One piece of geogrid was a.t,tacl-iedt.o one pla.te, a.iid these plates were connected iii hinged conclitioii each other as s k io n ~iri t~liefigure. An aluminurn iriodel coiit,aiiier wit,li iritier sizes of 450rrini in aidth; 150rniri in breadtli a i d 250riiiri iii height, was wed. R.iibber slieet,s were placed a t 1iot.l-i side of the contaiiier for alxorbiiig of stress ~vavesfroiri t,he side boundaries. This coiita.iner has a det~aclla.blebasc plate aiid a rid pla.te. so that riiodel ground can be prepared iii t,lie tipsitle-don;ii posit~ioii.Teiriplates were pla.ced in the turiietl up side donii coiita.iiier as shon~niii Figure 3. Inagi s a i d with water coiiteiit of 27%) m s statically conipacted t,o the bulk density 7, = 17.8hi\?/iri" layer by layer usiiig a. hellofrani qliiicler. Tlie iriotlel Seogrid wis placecl on each hyer i>.lid optical nia.Ilters for displikceiiient i r i ~ ~ ~ ~ ~ ~\ WeRi ~alsii i e i lt >t l i ~ c at ~d the f'roiit surface of' tlie ground. This coiiipaction wa.s corttiiiued u p to tlie top level of the 1)ox. .Uter coinpletioii of soil coriipa.ctioii) tlie base pla.te was attaclieti t,o t,lie container arid t.he container was turned t,D the right pnsitiori. Tlie rid arid t,he template was then t,akeii off a,s showii iii Figure 3(c).
Table 3. Amlied siiiusoidal wa.ves iii the test
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,
4th
,
l'(d.2)
100 (2)
20
20 (0.4)
100 (2)
40
Ha.viiig prepa.red the riiodel, t.he container n~as set. i>ii the shakiiig table iriouiit.ed oil t,he ceiit,rifiige. Sliaking t.ests were conducted uiitier SOG ceiitxifugal acceleratioii by siiiusoidal m v e s with ii. frequeiicy of' 100Hz: wliicli is eqiii\aleiit. t.1, 2Hz iri t.he protot.ype scille, to t.lie slltiliiug t.able. Fo~our IviIVes n i t h diff'ereiit. coirdit~ic.)iis as sliowii i i i Table 3 were input t,o each model. Typical time Iiist.ories of the iiiput siiiusoidal n-ayes ill'e. S ~ O W I iii J Figure 4. During slialiing, i~x:eIer~tti~)~i aiid tlisplaceriierit. of t.ho rciinforccd soil xvprc' inciasnred at the loc;tt,ioiis sliown in Figure, 1. Phot,ographs m r e tdtcn beforo and aft,c.r shaking t o ohsc:rvc~thc c1isplac:onieiit~of t,arget,s on thc front, stirface of t,hc rciiiforccid soil.
Figure 5. Observed acceleration a t All aiid A 2 1 ( C x c 4 aid 5 in S t q 2)
Figure 4. Iiiput waves acceleration time histories (Cascl 2) Table 4. Test coriditioiis L (min) s (inin) Test case Case 1 150 (7500) 30 (1500) Cast. 2 120 (6000) 30 (1500) Case 3 90 (4500) 30 (1500) cast>4 120 (6000) 15 (750) caw 3 90 (4500) 15 (750) iii parentheses. prototype scale
2.2
3 TEST RESITLTS AND DISCUSSION A11 data prcwntod in this swtion arc in t h c prototyp" SC*;tlP.
3. I
Test coPdZtZo7L.s
Table 4 gives tlie test c:oiiclitioiis adoptetl iii this st,ucl\:. Height, of tlie reiiiforced soil ivall was 150iiiiii3 7.5iii iii the prot,otype scale. Effect, of length of geogricls (L) ori the periiiaiieiit, (leforiiiai g a t d i i l Cases 1. 2 arid 3. To gaiii iiii iiisiglit iiito the effect of SpiLciiig l.wt8n.eengpogricls (s) 011 tlie d(Jfirl the ~ ; j ~ l stage > . of sli a k i ii g . It should be iiotecl t h a t tlie relati1.P locatioii o f i~CceIerollietei~ -411 to the reiiiforced zoiie iii Case .i is clifiereiit, froiii that iii Case 1. . A l l located oiitside of the I~iiif(.)rcedzoiie i i i Case 5,wliile iii Chse 4 the location of' A l l was the boundary between reinforced and non-reinforced zone. To obtain further information on the interaction between the reinforced zone and the soil behind it, relative accelerations between A21 and A l l are showri in Figure 6. The relative acceleration is taken as negative
993
Figure 6. Observed relative acccleration at, A21 t,o A l l (Case 4 and 5 in Step 2)
Figure 7. Tiirie histories of settlement at L1 (Case 3)
when the fortsiarcl acceleratioii at A21 was larger than that A l l . In Case 5, large negative relative acceleration was observed between the reinforced zone and the soil beliind it, coiripai-ed with Case 4. This inay iiriplv that large relative displaceirieiit occurred at reinforced and non-reinforced zone arid largei impact load act to reinforced zoiie from the soil beliiiicl in Case 5 tlian Case 4. ,4cceleratioii responses in Cases 2 arid 3 were esseiitially the sarne as observed iii Cases 4 arid 3. 3.2 Perm anPnf deform at ion, of iiertzccd remforced sod s l o p Time liistoiies of settlement at tlie shoulder of the
soil slope. L1 in Case 3 are slio~vnin Figure 7. The settkiiieiit, gra.tlually accuinulated with time witlioiit aiiy clrairiatic iiicrease. Observed rleforiiia.tioii of iiioclel grouiid due to sliakiiig B I P sliowii for Cases '3. 3: -4 a.iitl 5. respecti~.ely iii Figitre S. Altlioiigh t,lic, iriagiiitude of displaceineiit differs for differeiit coiidit ioiis, deforiiiatioii modes were two-part. wedge tj-pe iii all the cases. T1ia.t is a triangle active failure beliiiitl the reinforced soil acc:oiiipanirig with the horizont,al displac:eiiierit, of tlie reiiiforcecl soil. Iii all tests: lijrge liorizoiital displaceirierits of 1~x11s\VPR 01)serirecl. especia.11,~iii Cases 3 m t l 5. Perlilitileiit, set tleiiieiits clist,riliution at t l i e shallow dept 11 iii tlie soil aiicl 1iorizoIital pei~iiaiiwttlisI)l;tceiiieiit.sclistri1)iitioii i1.t the s l ~ p ef t i w a f t , P r t l i ~ 4th sliakiiig are slion~iiii Figures !) ~ i i d10. re, li\rge xet,tleii\riits J T ~ Y spec:t,ively. 111 ill1 t lie rced zoiie wliere t.lir a(:t.ive faillire wedge foriiiecl. III Cases 3 a i i d 5 wit,li shorter goegrids, tlie settlement became larger at, both reinforced zone a,rici the active failure wedge than the models with longer geogrids (Cases 2 and 4). The effect of the spacing between geogrids could not Iieeii clearly seen in tlie permanent sett,lement,s. In Cmes 3 aiid 5, very large liorizoiit.al displaceirieiits were also observed. Iri these cases. the effect of the spacing betweell geogrids n ~ found ~ s in tlie liorizoiita,l displaceriierits to soirie esteiit., i.e. the longer spa.cirig causes the larger displaceirieiit. On t,he other liand, iii Cases 1: 2 a r i d 4 with loiig geogrids, small liorizoiital peririaiieiit, displacemeiits were observed, aid 110 obvious effect of the spacing could be seen. From these figures, it, cari he sa.id tlia.t, the spaciiig between geogricls do Iiot iiiucli affect, tlie settleirients of reinforced soil l.)i.tt.t,liP liorizoiit,al displaceriieiit,s of the slope w h m t,lie le~igtliis short. Coiisideriiig tlie defoririatioii inode, it, caii he said tdia t t,lie I m e slicliiig aiid defoririatioii of t.he reirifijrcecl zoiie cause t,lie liorizoiit,al displaceinerit, o f t.he slope surface. 111 Figure 11. observed liorizoiital displaceirierits at section A and 13 of soil n d l s (see Figiire 8). a e shi-)n~iifm Cases 2. 3 , 4 aiid 5. Horizoiit.al displaceiiient,s at, t,lie ltot,t,oin of sectioiis A aiid B corresponcl t,o the base slidiiig of reiiiforcrP(1 zoiie. Large base slidiiigs t.ook place iii tlie (rases n-it,Iisliwter geogricls. Honw.er: thpw is rge (lifterpiice iii tlie lime slicliiig I)etn-eeii the lvitli the s mi e leiigtli a i i t l tliffewiit. spaciiig (CilSrs 2 arid -I 01 Cilses 3 aiid 5). I t iinplies that the s1)xiiig of geogrids does tiot. iriucl-i affect, o i i t.lie I m e sl icl i ii g . At, the ~ipperportioii of the reiiiforced zoiies, alIiiost the sil.ille Iiorizolital displacenieiits a101ig t,he elevatioii n w e oljser\:ecl. 0 1 1 the other hillid, at the lower portion, the liorizoiital tlisplacenieiits iiicreasecl witli the eleva.tion. This iiidicates that the shear clefamation of the reinforced zoiie iziainl!; 994
~
Figiire 8. Observed deformation of model ground
(a) With long geogrids (Cases 2 and 4)
Figurc 9. Pcrmancnt, sctt,lemcnt,s distribiit h i aftcr a11 shaking
(h) With short, geogrids (Cases 3 and 5)
Figure 11. Observed displacement of soil wall
Figurci 10. Horizont a1 1minaiimt displace111cnts of slopc sllrfacc~ aftc.1. all sh aliing
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t,oolt place at tlie lower portion. To gaiii iiisiglit, into the effect of t,lie reiiiforceiiieiit 011 t,he lateral espa.iision of reiiiforcecl zoiie, relilt,iI-e periiiarient displacPriieiit,s of reinforced zc)iies lx%n-erri sectioiis -4 a.~iclB are sliowii iii Figure 12. The rela.t,ive clisplaceIiielitl is tillten as positi1:e ~rlieiitlie sect,iori -4 1iio17es ~iioret.liaii sectioii B. i.e. reiiiforced zoiie iiicre tlie resiilts of Case 5 irliere t,he relaierit at, the upper portion is iiegatii-e, which nieaiis liorizoiital coInpressioii, it c a ~ be i said tShat the lat,era.l erpaiisioii of reiriforcecl zoiie became sriialler as tlie leiigtli of geogricls increased aiid tlie spacing betnreeii geogricls elecreasect.
l~ecariiesriialler a s tlie leiigth of geogritls increased a.nd tlie spacing betweii geogricls clecreased.
(5) Lateral ( y w i s i o n of rc4nforccd mno l)o(miio siiiallcr as t.hv l(~ngt1iof googri(1s incmwod an(l t,ho spacing I , ( l t \ t r ( > t L l i googrids dtl(:r(tiis(’d. hlost of siiuplified iiietliotls to c>stiiiiate a periiiaiieiit tfispla.ceiiieiit of i1 reiiiforceti soil 1ia1.e o d y consiclered a base slidilig. This s t u c l ~ .s l i o n ~tlie importance of an estimation of a permanent cleformation of a reinforced soil itself, which is highly affected by the spacing between geogrids.
ACKN0TVLEDGEMEi.T The preseiit studv was supported by JSPS under the Japan-US Cooperative Science Prograiri with SSF. This support is gratefully acknowledgecl. REFERENCES
(I) Although niagnitirde of displaccimcnt diffmd for cfiffmlnt rciiiforc.c.nicnt condition+ twopart, n.cdgc~type dcformatioii niodo nv~:, ob-
(2) Largcl sclttJleiiic~ntsn-cw ohsclrv(3.d btihiiid thc winforccd zonc w l i ~ trho ~ 2t.t i w failiiw n w l g t ~ ~vabfornitd. In tlio modol:, ivith shortor googrids, tho sc\ttloiiicmt lm anit1 lxrgclr at both roinfou c d zone a n c l thv xcsti\-c failiiro ~wdgct than tho niodcls with l o n g t ~gcwgricls ( 3 ) L q e base slidiiigs tool< place iii thr case5 with sliortei geogiids. Howevei. tlir spcicing of geogi1ds llah 110 I l l U C l l rff.ct 011 tll? l m e slidiiig (4) Shear tlefoiriiatioii of tlie ieinfoiced zoiie inclinly took place at tlie lower poition. It
Figurc 12. Observed relat,iw displacements of reinforced zones bet,ivcen sections A and B 996
Takakiaslii, A.. Taltemura. J.. Tsutsunii. F. & SaeTia. W. 1998. Slialtiiig table tests on geogridreinforced eiribanlmient duririg earthquake with ceiitrifuge. PIOC10tli Earthquake Eiigiiieeriiig Sympo:,ium. 1’01. 2,1551-1556 (in Japancw) T;ikcmirra, .J.. Kiniiirrz, T.. & Siicmasa, S . 1989 D(~~~(~lopI11(~iit of Eart~h~j11;ili(~ siiiiiilators at Tokyo Institiito of Twhn port. No. 40. Dq,t. Cil-il tiitcl of T(v.hnolog>r.41-60
F.. Tateyxiia. XI. 5_. Koseki. .J. 1990. P ~ i f oiii i a i i w of soil retaiiiiiig n.~llsfoi 1 iii P 1r i ha ii k i 1i e 11t s Soi1s ii 1ic1 Fo 111id at ic n 5 . SI ) I S ~ I I 0P1 1 Grot~chiiiCid Aspects of the , J u ~ 17 199I, H (1 1 1.00 mm, we have compared the apperture size with the adhesion values (Table 3). It has been found that as the apperture size increases the adhesion value also increases. We do not recommend to introduce some of the adhesion values acquired in these tests into stability calculations. Based on evaluations of friction tests with geosynthetics and soil (Bluniel & Stoewahse 1998) recommend that adhesion values derived froni friction test results should be introduced into stability calculations only in special cases, e.g. for interfaces between cohesive soils and geoniembranes. 1073
Except for three reinforcements, both free shear tests and partially fixed shear tests have been carried out. In these experiments the least value of friction angle between soil and geotextile is 30.5" with sample no. 4 and the highest value is 33.8" for geotextile sample no. 6, with their bond strengths turning out from 0.77 to 0.88 respectively. Martin et al. (in Koerner 1998) have found soil (concrete sand) - to - woven geotextile hction angle to be in the range 0.77 - 0.84. This same result has been found in this work for the reinforcements tested on the free shear box with the sand. These values are recommended to be used for routine first hand design purposes of woven geotextiles.
Figure 3. Shear stress vs. shear displacement for sand geotextile interface (free shear test).
Figure 4. Shear stress vs. shear displacement for sand geotextile interface (free shear test).
As most of the reinforcing woven fabrics are made of polyester yarns or polyester tape fabrics, we have carried out 5 different tests with this high modulus polymer material. Table 2, shows a 6 % of difference in efficiency E4 (see eq 2 ) for polyester geotextiles, the bond coefficients to be in the range from 0.77 - 0.83. From Figure 3 we can also see as the horizontal displacement increases, the value of the shear stress can differ for the same polymer material (polyester) as much as 20 % for the same shear displacement value and same confining pressure. This can be explained by the difference in material roughness (Table l), apperture size (Table 3) and variations of weaving technology, where by it can affect directly the physical, mechanical and the hydraulic properties of the fabric (Koerner 1998). The polypropylene geotextile sample no. 6 on the other hand has turned out to have the best interface properties (See Table 2 and Figure 5). The values of the coefficient of interaction for this reinforcement are 0.84 for a free shear box test and 0.88 for a partially fixed shear test. Figure 4 demonstrates that the tensile strength of a material does not necessarily be directly proportional to the interface shear strength of the same material. Sample no. 3 and sample no. 4 are presented to the market under the same coniinercial name. Their difference except other things is in their tensile strength. It seems that shear parameters do not depend on the tensile strength of a material. Their bond coefficients (COI) are 0.88 and 0.84 respectively. As a matter of fact sample no. 6 with the least tensile strength (35 kNm) from all the materials tested was found to have the best shear parameter.
Figure 5. Shear stress vs. shear displacement for sand geotextile interface (free shear test).
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As it can be seen from the graph of horizontal displacement versus shear stress (Figure 5) for a free shear test, 5 types of reinforcements are presented where by, all of the polyester specimens were compared to the polypropylene geotextile (sample no. 6). The results show that sample no. 6 throughout the whole shearing, being all boundary conditions kept constant has the predominance in the shear stress values from 80 % - 100 % (Figure 5). An absolute prevailing of the shearing stresses can be read for the same vertical stresses values from 0 5 mm of the horizontal displacement value. Note that shear stresses developed during the shearing along the soil /geosynthetics interface, for lower displacement values are usually adopted for design purposes (European Standard - Draft 1997). Figure 6 shows the comparison of shear stress versus displacement for a polyester specimen and the polypropylene sample tested on the partially fixed shear mode. Test results obtained depict that the polypropylene geotextile specimen tested with the same boundary conditions as the polyester material is almost 100 % better as far as the shear parameters are concerned, even when shearing progresses. The roughness of the polypropylene geotextile tested has the biggest influence on the shear parameter (Table 1). This is because of a large amount of soil mass is mobilised, when a rough reinforcement is used and intum an increase in interface friction angle (Baslik 1985). The relatively higher opening size ratio of this polypropylene reinforcement than the polyester specimens is also the other factor, where is due a higher value of interface friction angle is recorded. In such cases the
Figure 6. Shear stress vs. shear displacement for sand geotextile interface (partially fixed shear test).
soil is pushed into the appertures, where by the soil mass in the interface is not very much distorted, thus friction is increased. While comparing the nature of the graphs (Figure 5 and Figure 6), tests conducted under the free shear mode do not actually reach their maximum shear stress values within the given displacement. They have an increasing tendency, when shear progresses (Figure 5). In what concerns the tests on the partially free shear mode the maximum shear stresses values are reached for displacements within the range 0 to 10 % of the length of the shear surface (Figure 6).
6 CONCLUSION The following conclusion can be drawn for the woven geotextiles tested with a sand on a direct shear apparatus : a. Peak soil - to - geotextile friction angles for the tested woven geotextiles have turned out to be 30.5' - 32.6' (Table 2), which corresponds to a CO1 value of 0.77 - 0.84 (free shear mode) and it can have a higher value up to 0.88 (33.8'), when tested in a partialy fixed shear (the authors recommend experiments to be conducted on more sample population) mode. b. The bond coefficients of the polyester woven geotextiles are found to be in the range of 0.77 0.83 (Table 2), which corresponds to a 6 % of difference in efficiency (E#). This difference might be influenced by the type of the weaving technology, the apperture size and the roughness of the fabrics tested. As much as a 20 % difference in shear stress has been found when the nature of the graphs were studied for the same shear displacement values and same confining pressure (Figure 3). c. The polypropylene woven geotextile has been found to have the best shear strength parameters. A CO1 value of 0.84 and 0.88 (Table 2, Figure 5 & Figure 6) has been calculated for a free shear test and the partially free shear mode respectively. From 0 - 5 mm of the shear displacement value, the polypropylene geotextile has shown to have the biggest shear stress value for all confinment stresses applied. Note that shear stresses developed during the shearing along soiVgeosynthetics interface, for lower displacement values are usually adopted for design purposes (European Standard - Draft 1997). d. The shear stress versus shear displacement curves for the partially fixed shear tests reach
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their maximum shear stresses values in the range of 0 - 10 % (Figure 6) of the length of the shear surface. In what concerns the graphs from the free shear box test (Figure 5) there is a progressive increase of the shear stresses values as the displacemnet increases.
ACKNOWLEDGMENT This paper is prepared for the grant of the Ministry of education of the Slovak republic entitled : Friction of soils and geosynthetics. The authors would like to thank Mr. Oswald for the presented microphotographs.
REFERENCES 1. Baslik, R. , Janotka, I., Ayele, T., 1997. Geosynthetic and Mineral Materials Used for Landfill Liners and Covers. Sixth International Landfill Symposium, 13-1 7 October, Sardinia, ITALY vol. IIIp. 293 - 301. 2. Baslik, R. 1985. Geotextiles for reinforcement. VUIS, Bratislava. 3. Blumel, W., Stoewahse, C., 1998. Geosynthetic interface fiction testing in Germany - effect of test setups -. Sixth International Conference on Geosynthetics, Atlanta, USA, p. 447 - 452. 4. European Standard Draft prEN IS0 12957 -1. 1997. Geotextiles and geotextile - related products - Determination of the friction characteristics - Part 1 Direct shear test (ISO/!DIS 1295 7 -1 : 199 7). 5 . Koerner, R.M. 1998. Designing with geosynthetics. 4" ed. Prentice Hall, New Jersey 6. Lopes, M.L., Ayele, T., 1998 : Influence of Reinforcement Damage on the Pull-out Resistance of Geogrids. Sixth International Conference on Geosynthetics, Atlanta, USA, p . 1183 - 1188. 7. Matys, M., Hric, S., Polko, I. 1997. Hydro power plant iilina - Shear tests of soils and geosynthetics and the design of sealing of the approach channel. Geotechnics 97 - The of Modern Technology of Rudiment Construction, Podbanskk, SLO VAKIA,p . 43-45. 8. STN 72 1030 (Slovak Technical Standard). 1987. Laboratory direct shear box drained test of soils. 9. STN 73 6025 (Slovak Technical Standard). by 1997. Soil structures reinforced geosynthetics. Technical requirements.
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10 Probabilistic slope stability
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Slope Stability Engineering, Yagi, Yamagami& Jiang 0 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Localized probabilistic site characterization in geotechnical engineering S.Pumjan & D. S.Young Department of Mining Engineering, Michigan Technological University,Houghton, Mich., USA
ABSTRACT: Probabilistic modeling of geotechnical strength parameters for large scale in-situ characterization of geological formations is presented in this paper. The actual measurements of strength parameters from a limited sample test (hard data) are combined with other sources of information (subjective soft data) to characterize the localized probabilistic model of in-situ strength parameters. The sequential Gaussian cosimulation technique was introduced to achieve the localized probabilistic characterization of in-situ strength parameters. The bivariate probability distribution was applied for the probabilistic joint model of strength parameters (c-4). In other words, the full statistical distribution of strength parameters are obtainable for every local point within the geological domain. These localized probabilistic models provide the necessary input for probabilistic structural analysis, such structures as slopes, dams, foundations or underground openings, where c and 4 are applied for the structural design.
1 INTRODUCTION
Probabilistic modeling of geotechnical strength parameters for large scale in-situ characterization of geological formations requires a formidable task to accomplish based on sparsely located samples available. This is due to the fact that the physical reality of strength parameters are varied spatially throughout the formation and the probabilistic description of their spatial variability is almost impossible with limited samples located sparsely. Also, the interdependence between the strength parameters should be taken into account in the probabilistic modeling technique, which makes it more complicated statistically. To overcome these difficulties, the actual measurements of strength parameters from a limited sample test (hard data) must be complemented by other sources of information (subjective soft data) such as quantitative I i thological descriptions, geophysical data, and other physical properties determined from simple laboratory and field tests. The soft data are generally abundant, and easy and cheap to obtain, but do carry valuable information that should be used in the estimation model for probabilistic site characterizations. Consequently the soft data was incorporated with the actual hard data
to overcome these difficulties through the geostatistical techniques and to improve the quality of the probabilistic modeling of strength parameters. The sequential Gaussian co-simulation technique (SGCOSIM), developed by Almeida ( 1 993), was introduced to achieve the localized probabilistic characterization of in-situ strength parameters, which can combine hard data of strength parameters with other sources of information to simulate the spatial variation of geotechnical strength parameters in probabilistic terms. Therefore, the probabilistic distribution of strength parameters are available for every local element (or point) within the entire geological formation. The strength parameters, cohesion (c) and internal friction angle (@) in this case, were jointly simulated based on the sequential principles, which duplicates the probabilistic structures of strength parameters as well as their spatial variations within the geological formation. Also, the Markov-Bayes assumption was applied to simplify the co-simulation process as compared with the traditional co-kriging method that requires the cross-covariances. Cross-covariances are not easily obtainable, if not impossible, for geotechnical projects due to the small number of sample data. Technical details of SGCOSIM are given in the next section.
1079
Therefore, when applying this conditional simulation algorithm, it is necessary to check that the studied variables are at least bivariate normally distributed. The check is implemented by the way that the experimental indicator variograms corresponding to specific cutoffs are compared to the corresponding Gaussian theoretical ones. The comparison allows the judgment of accepting or rejecting the multi-Gaussian model (Almeida 1993). Conditional simulation of strength parameters using SGCOSIM. Compute the statistical parameters of bivariate normal distribution (BVND) from the simulated data for ;very local block: BVND (uc,u+,o;,a&p,+) where U , and o, are the mean and standard deviation of c, U+ and o,+are mean and standard deviation of 4, and pc,+is the correlation coefficient between c and @ (Tatsuoka 1971).
2 SEQUENTIAL GAUSSIAN CO-SIMULATION SGCOSIM is developed for joint conditional simulation of several interdependent random variables. The sequential simulation (Deutch and Journel 1992) implies that the univariate conditional cumulative distribution function (ccdf) is estimated sequential1y.
where 2,is a random variable, z, is a particular realization of Z,, rz is the original data, and N is previously simulated data. By assuming the multiGaussian model, all ccdf's are normal distributions which are fully characterized by their means and variances. These statistical parameters are estimated locally by the simple kriging system. The joint spatial variability of several variables is reproduced by conditioning to a prior model of covariances and cross-covariances. The derivation of conditional distribution is now conditioned to data of the same type, data of different variables, and previously simulated data found within a neighborhood of the location being simulated. The spatial cross-correlation between variables is injected through the cross-covariance structure, and the estimation of the conditional mean and variance of ccdf calls for the use of the simple co-kriging system. The distinctive advantage of this simulation algorithm is that several spatially interdependent variables are jointly simulated. Also, it allows the integration of different sources of information (hard and soft data) in the mapping of several primary variables (hard data) without the complexity of the traditional co-kriging system. The cross-covariances of primary (hard) variables and secondary (soft) variables are inferred from the covariances of hard variables based on the Markov-Bayes assumption. A step-by-step procedure of the probabilistic characterization of the in-situ strength parameters by using SGCOSIM is outlined below: Preliminary statistics of hard variables (c-4) and soft variables (soil density in this case). Structural analysis: the covariances of strength parameters and the cross-covariances of strength parameters and soil density. Checking for bivariate normality: SGCOSIM is built on the multi-Gaussian assumption.
CASE STUDY The localized probabilistic characterization of the insitu strength parameters using SGCOSIM was carried out on the soil slope at the Mae Moh mine, Lampang province, northern Thailand. A total of twenty-seven soil strength data, in terms of cohesion ( c ) and internal friction angle (@),were tested within the area of 600m x 400m: Northing (1900m to 2500m), Easting (-3800m to -4200m). These soil strength data were obtained from direct shear test, by which at least three soil specimens were tested at various normal loads of specific magnitude. A Mohr Coulomb's failure criterion was then applied to determine the strength parameters: cohesion and friction angle. The strength parameters are modeled as primary (hard) variables. Associated with the strength data was the soil density data obtained from the laboratory soil density test. A total of twentyseven soil density data was collected at the same location of strength data. In this study, soil density is modeled as a secondary (soft) variable. As it has been outlined, the localized probabilistic characterization of the in-situ strength parameters using SGCOSIM starts with preliminary statistics of variables under the study. 3.1 Preliminary statistics Figure 1 shows the normal probability plots of cohesion, friction angle, and density variables. As seen in Figure I , the cohesion and friction angle show 1080
performing SGCOSIM the normal score transform is used to transform the experimental histogram of random variables to the standard normal distribution with zero mean and unit variance. Therefore, the variables under the study can be distributed in any shape. The experimental histograms of cohesion and friction angle are transformed to the standard Gaussian distribution by the normal score transform method. The normal score transformation function used in this study transforms one variable at a time and ensures the normality of the transformed variables. However, the bivariate normality assumption can be acceptable by considering the normality of each variable and their correlation coefficient, as well as physical conditions: two shear strength parameters were tested on the same sample taken randomly (Young 1977, 1993). The general statistics of primary and secondary variables are given in Table 1. Table 1. The general statistics of studied variables. Variables
Number of data
Mean
Variance
cov"
cohesion (kN/m2)
27
22.10
70.40
0.379
internal friction angle (degree)
27
18.29
41.69
0.353
density (ton/m3)
27
2.14
0.0036
0.027
%ov is the coefficient of variation. It was found that there is an essential correlation between primary and secondary variables as seen in Table 2 which shows the correlation coefficients of the input variables. Table 2. Correlation coefficients between variables. Figure 1. Normal probability plots of cohesion, friction angle and density variables. almost a straight line and confirm Gaussian distributions. Thus, up to this point, it is safe to assume that the cohesion and friction angle are normally distributed. On the other hand, the density shows a skewed distribution, and its normal probability plot cannot be fitted easily by a straight line, which indicates that the density variable is not normally distributed. However, this should not pose any problem in the simulation model. Because in
Variables
Correlation coefficients
cohesion and friction angle
-0.343
cohesion and density
0.408
friction angle and density
-0.335
From Table 2, the cohesion and friction angle are negatively correlated, the same with the correlation coefficient of friction angle and density having a moderate negative correlation. On the other hand, the cohesion and density are positively correlated. The moderate correlation coefficient between cohesion 1081
words, the stationarity of variables under the study remains unchanged.
and friction angle, pc+= -0.343, implies that these two variables are interdependent. Thus, the joint conditional simulation of these two interdependent variables is required. It has been mentioned that the advantage in incorporating the secondary (soft) variables into the mapping of the spatial distribution of primary (hard) variables depends on the degree of interdependence between primary and secondary variables. Therefore, the moderate correlation coefficients, as shown in Table 2, Pc.density = 0.408 and P+-density = -0.334, justify the use of soft information in the simulation of the primary variables, and improve the quality of their models. Finally, the joint conditional simulation of the interdependent strength variables by integrating the knowledge of density variables is achieved through the use of SGCOSIM.
3.3
SGCOSIM is built on the multi-Gaussian random function model. The multi-Gaussian model requires that the variables have to be multivariate normally distributed. Therefore, before applying this simulation technique, it is necessary to check if the variables are at least bivariate normal. In practice, this is checked by comparing the experimental indicator variogram calculated at different cutoffs to the theoretical indicator variogram calculated at the same cutoff value (Almeida 1993). In this study, the experimental indicator variograms of cohesion and friction angle at three different quartiles [the first quartile (25%), the second quartile (50%), and the third quartile (75%)] were calculated and then compared with the theoretical indicator variograms at the same quartiles. Figure 3 shows the experimental indicator variogram and theoretical indicator variogram corresponding to the first quartile for the cohesion. As it can be seen, the comparison between these two variogram models was reasonably positive in cohesion. The same conclusion was drawn for friction angle although it was not shown in Figure 3. However, by judging from the global picture, the two variogram models show a good agreement, and hence the multi-Gaussian random function model is deemed appropriate for this study.
3.2 Spatial variability analysis To determine the overall structure of spatial variations for cohesion, friction angle, and density, the experimental variogram calculations were performed in two major directions: horizontal and vertical. However, the very erratic point variograms observed from both directions did not carry any significant spatial structure, and thus a complete description of spatial structure cannot be drawn. The erratic point variogram calculations are mainly attributed to the limitation of the experimental sample data: only 27 samples are available in the total slope area of 600m x 400m. Also, a directional variogram search reduces further the possibility of finding sample pairs. Hence the omni-direction variogram was searched. In doing onmi-direction variogram calculations, the variogram was searched in every direction. The experimental variograms and their models of variables under the study, and the transformed variables, are given in Figure 2. The variogram models and their parameters are given in Table 3. It is important to note that when the omnidirection variogram is applied, the spatial variability is assumed to be the same for all directions. In general, when the information is sufficient to carry out the directional variogram calculation, the anisotropic variability can easily be added into the simulation model by comparing the ranges of variograms in different directions. As shown in Figure 2, the experimental variograms were all modeled reasonably by the spherical function. The ranges of variables before and after transformation are almost the same. This indicates that the transformation process does not affect the spatial structure of variables. In other
Checking for bivariute normality
3.4 Joint conditional sirnulatiorz with SGCOSIM In this study, the localized probabilistic characterization of strength parameters was applied on the cut slope with the dimension of 130m x lOOm x 60m in length, width and height. The cut slope was discretized by a grid with 26 x 20 x 30 nodes for the conditional simulation corresponding to block size of 5m x 5m x 2m. Then, twenty simulated values of cohesion and internal friction variables are stored at the center of the cell blocks. Finally, the statistical inference of BVND parameters for each cell block, were then computed based on these (uL,uS,o;,o&pLS) 7 7 simulated data. Figure 4 shows the distribution of the mean values of cohesion and friction angle, U , and ucp,respectively at the specific cross-section -3900 East. As can be seen in these pictures, the simulated mean values, computed out of 20 simulated values, varied locally for both cohesion and friction angle variables. The transition from the higher mean values to the lower mean values was well recognized. The low mean
1082
1.20
Everfmental variogram calculation & Modelling
7
1.00-
0.80-
0.60-
1'
d
0.40-
0.20-
,cohesion(hard) variable I
0.0
'
"
'
I
'
'
"
20.0
10.0
I
'
1
"
30.0
''
o
f 40.0
transformedcohesion (hard) variable .
0.0
w 10.0
Dlstame
m
20.0
i
30.0
40.0
Distance
Figure 2. Raw variogram and normal-transformed variogram of cohesion variable.
Table 3. Variolrram models of variables.
I Normal score transformed data
I Original spatial data Model
Nugget
Range (m>
Sill
Model
Nugget
Range (m>
Sill
cohesion
sph"
5
15
65
sph*
0.1
16
0.9
phi
sph*
2
15
38
sph"
0.05
15
0.95
density
sph"
0.0005
14
0.003
sph*
0.1
15
0.8
0.60
o.80]
0.40
0.20
0.00 0:o
10.0
20.0
0.0
30.0
10.0
20.0
30.0
Distance
Distance
Figure 3. Experimental and Gaussian model-derived indicator variograms of the first quartile for cohesion variable. values of strength parameters indicate the low soil strength area or weak zone. At the same time, the high mean values of strength parameters indicate the high soil strength area or strong zone. The output from the localized probabilistic characterization of the in-situ strength parameters using SGCOSIM is the statistical parameters of BVND (uc,u+,a:,oi,pc+) for each cell block. Table 4 illustrates the statistical parameters of BVND for cell blocks.
4 CONCLUSIONS
The most important conclusion, in general, which could be drawn from this work is that the probabilistic modeling of geotechnical strength parameters can be accomplished based on sparsely located samples available. The geostatistical simulation technique, sequential Gaussian cosimulation (SGCOSIM), was used to achieve the 1083
provide the necessary input for probabilistic structural analysis at a desirable confidence level, such structures as slopes, dams, foundations or underground openings where c and Q, are applied for the structural design. Moreover, risk assessments can be achieved for both the potential structural failure and the uncertainty of the in-situ characterizations. Finally, the probabilistic characterization of strength parameters for geotechnical sites was carried out by using cheap soft data, which will improve the engineering structural analysis in realistic terms as well as its risk assessments.
REFERENCES
Figure 4. The mean values of the cohesion and friction angle variables at -3900 East. Table 4. Example of output data file from SGCOSIM. U,
U6
0,
06
Pc6
21.6689
21.6854
6.7780
5.4724
-0.5508
22.7488
18.1594
6.7740
3.9782
-0.2880
24.6360
17.8106
7.3438
6.3982
-0.5670
24.0786
19.3124
7.0520
6.4766
-0.1671
21.9617
18.3561
5.5744
4.8500
-0.0022
Almeida, A. S. 1993. Joint simulation of multiple variables with a Markov-type coregionalization model, Ph.D. Dissertation, Stanford University, Stanford, CA, 199 pp. Deutch, C. V. & A. G. Journel. 1992. GSLIB: Geostutistical Software Library, 340 pp, New York: Oxford University Press. Journel, A. G. & Ch. J. Huijbregt. 1978. Mining geostutistics, 595 pp, London: Academic Press. Pumjan, Sunthorn. 1998. A localized probabilistic approach for slope stability analysis, Ph.D. Dissertation, Michigan Technological University, Houghton, MI, 180 pp. Tatsuoka, M. M. 197I . Multivariate analysis, 3 10 pp, New York: John Wiley & Sons. Young, D. S. 1977. Probability analysis of rock slopes and its application to a pit slope design, pp. 5C5-1 - 5C5-6, Proc. 18"' US Syrnposiuin on Rock Mechanics. Young, D. S. 1985. A generalized probabilistic approach for slope analysis, Internutionul Journal of Mining Engineering, 3:2 15-218. Young, D. S. 1993. Probabilistic slope analysis for structural failure, Int. J. Rock Mech. Min. Sci., & Geonzech. Abstr., 30(7):1623-1629.
localized probabilistic characterization of the in-situ strength parameters. SGCOSIM takes into account the physical reality of strength parameters which are their spatial variability and their interdependency. Also, SGCOSIM combines hard data of strength parameters with soft data of soil density to simulate the spatial variation of geotechnical strength parameters in probabilistic terms. In other words, the full statistical distribution of strength parameters are obtainable by using cheap soft data for every local point within the geological domain, where geotechnical engineering is applied. And these localized probabilistic models of strength parameters 1084
Slope Stability Engineering, Yagi, Yamagami& Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
A localized probabilistic approach for slope stability analysis D. S.Young & S. h m j a n Deparmzent of Mining Engineering, Michigan Technological University,Houghton, Mich., USA
ABSTRACT: A localized probabilistic approach for slope stability analysis is presented in this paper, which is also applicable for the risk assessment of slope failure at a specified confidence level. The probabilistic analysis as well as its localization was achieved through the probabilistic characterization of the in-situ strength values, that provide the full statistical distribution of strength variables at every point within the geological formation. An actual case study was made for a surface coal mine to demonstrate its superiority and to show its technical details.
1 INTRODUCTION
The localized probabilistic analysis of slope failure was formulated into two stages of analysis. In the first stage of analysis, the probabilistic characterization of the in-situ strength parameters was carried out by the geostatistical conditional simulation technique called sequential Gaussian cosimulation method (Almeida 1993). The second stage of analysis involves the probability calculation of slope failures by combining the probability regions of strength values and the deterministic slope analysis methods.
I . 1 Probabilistic site characterization The sequential Gaussian co-simulation method generates the local joint probability distribution functions (joint pdf's) of strength parameters, which are conditioned to available experimental samples in the slope area. Therefore, it offers various advantages over other conditional simulation methods and is ideal for the geotechnical site characterization, where a probabilistic structural analysis and its risk assessments are desirable. Many primary variables can be simulated sequentially with the secondary variables and their covariances are reproduced in the simulation. It includes not only the correlation between the primary strength parameters but also the correlations between the primary hard data and secondary soft data in the estimation of the primary variables, which will
improve the quality of primary variable models based on the secondary variable information. The spatial variability of both primary and secondary variables are also quantified and incorporated into the simulation process through covariances and variogram models. In the slope analysis, the primary strength parameters are cohesion values ( c ) and internal friction angles (4) of the slope materials. The secondary information could be other soil parameters including pocket penetrometer tests, liquid limits, plastic limits, or geophysical data, and geological interpretations. Finally the sequential Gaussian co-simulation estimates the full statistical distribution of primary variables at every point in the slope area. The technical details of probabilistic site characterizations are available in other publications (Pumjan 1998, Pumjan and Young 1999). As a simple example, the localized probabilistic model of soil strength parameters is given in Figure I , where the average strength values (c-values only) are given for every element within the coal mine slope area. It provides the input needed for the local slope analysis in probabilistic terms.
2 PROBABILISTIC SLOPE ANALYSIS
The localized probabilistic analysis of a slope was solved by combining the local joint probability model of the in-situ strength parameters and the threedimensional deterministic method of slope analysis. 1085
Figure 1. Mean values of c-variable. Then, the safety of a slope is measured in terms of the probability of failure at the specific confidence level desired. The technical procedures of probabilistic slope analysis are given in the following sections.
Figure 2. Isodensity contour ellipses at various confidence levels and the critical boundary line.
2.1 Bivariate normal probability The strength parameters, c and @ values, are not normally distributed in general. Therefore, these primary variables have to be transferred into the Gaussian variables before the bivariate normal probability is applied for the probabilistic slope analysis. Then, the whole process will be generalized for any type of variable distribution. The Gaussian transformation of random variables can be achieved through the orthogonal polynomial function of the Hermite model whose weighting function is the standard Gaussian density function (Young 1985). A computer software for this transformation is available in GSLIB (Deutch & Journel 1992) which was used in this paper. Once the strength parameters are accepted originally as Gaussian variables or transferred into normal scores, a simple bivariate normal probability analysis can be made for the slope stability by drawing the critical boundary line on the probability region projected on the c-@plane. 2.2 Probability regions For quantitative probability analysis, it is customary to represent a binormal distribution slirface by plotting isodensity contour ellipses with various confidence levels, as shown in Figure 2 (Young 1985, Tatsuoka 1971). The contour ellipse defines the probability region with a specific confidence level of its own region. Therefore, the probability region defines the actual sample statistics, and a realistic slope analysis can be achieved on this region. Also, the specified confidence level can be converted into the risk level that could be tolerable in a slope. Consequently the risk assessment on a slope failure can be obtained easily from the probability region on the c-4 plane.
2.3 Critical boundary line The probability of slope failure (p,) can be determined by locating the critical boundary line on the c-@ plane, which will divide the probability region into two zones; stable and instable zones. Then, the probability of slope failure is simply the probability volume above the instable zone bounded by the critical boundary line and the probability region of a specified confidence level (see Fig. 2). The critical boundary line can be obtained from the back calculation of a deterministic method of slope stability analysis. The three-dimensional slope stability analysis method, called Hovland’s method, is adopted in this case (Hovland 1977). It is an extension of the two-dimensional ordinary slicing method (or Swedish method) and a limited equilibrium method based on columns in the place of slices. The back calculation made on the critical failure surface with a given safety factor of design criteria will generate a series of necessary strength values for the slope (Young 1985). These are the strength values required to maintain the given safety factor within a slope and draw a critical boundary line as plotted on the c-4 plane, where the probability region was projected before (see Fig. 2). 2.4 Probability of jailure The final probability of failure is obtained simply by integrating the volume of the bivariate density surface above the instable zone in Figure 2. The numerical integration of pf can be done by using a subroutine BNRDF in IMSL STATLIBRARY (199 1). Since the full binormal distribution of strength values is available for every point, including the base of each column sliced by the critical failure surface, from the input model of localized probabilistic site
1086
characterization, the probability calculation can be made on each column and pf can be localized in this approach. By knowing pf at every point or every column, a realistic safety analysis and interpretation can be made on the slope such as the spatial distribution of pf in a slope area, better ways of locating and monitoring slope failures, and remediation plans or searching potential failure areas that are not completed yet in this paper. Also, the global probability of slope failure was reached by making a weighted average of local pf’s in terms of base areas of columns along the critical failure surface. The technical details of complete slope analysis are referred to Pumjan (1998).
3 A CASE STUDY ON A COAL MINE
A case study was carried out on the soil slope of a surface coal mine to illustrate the technical superiority of local and global pf’s over the conventional methods. The study area was arbitrarily selected and a total of 27 soil strength data (c and 4 values) were tested from an area of 600m x 400m x 60m. The direct shear tests were made to obtain the soil strength values under Mohr-Coulomb’s failure criterion. The test results are given in Table 1. The strength parameters were modeled as primary (hard) variables and the soil density data obtained at the same locations of strength data were simulated as a secondary (soft) variable in this case. The results of localized probabilistic characterization are given in Figure 1, as an example where the average values of c-variables were plotted in the elemental model of slope. The general statistics and spatial variability analysis as well as localized site characterization for this case are available in other publicatio~~s (Pumjan & Young 1999). This is the input model of soil properties in the slope area and provides the local statistical distribution of soil data needed for the probability analysis. Table 1. Soil sample tests. Parameters Mean
Variance
c (kN/m’)
22.11
0.379
0 (degree)
18.29
0.353
2.14
0.028
densitv (ton/m3)
3.1 Probability of slope failures The probability of slope failure was analyzed using the design safety factor of 1.0 and the confidence level of 99.99% in this case. The critical failure surface was searched using the automatic search subroutine based on the limited equilibrium method. When the slope angle of 30” and height of 30m were applied, global p;s and deterministic safety factor of Hovland’s method were summarized in Table 2. In this case the deterministic safety factor of 1.15 by Hovland’s method is acceptable for mines, but the global pf of 39.5% may not be tolerable easily. It demonstrates clearly the superiority of pf over the deterministic analysis as a measure of slope safety. Also, pf gives the risk involved with the slope safety, since pf was calculated with a specific confidence level on the probability region. In other words, the probability of overall slope failure is 39.5% with a 99.99% confidence level. So, pf is an ideal measure of slope safety with its risk involved, and it is a complete analysis of slope safety.
Hovland’s
Global pi
Height
Slope Angle
S.F.
at 99.99%
30m
30 O
1.15
39.5%
Also, local pf’s were calculated and projected on the base plane of the elemental block model of the slope and contoured in Figure 3, which also shows the columns used in the Hovland’s method and local pf analysis. The local pf’s in this area ranged from 10% to 90%. As it may be expected, higher p,’s are located in the central portion of the sliding mass and this portion will have a significant influence on the overall failure. It demonstrates that the local pf draws more detailed structural safety over the global pf, which is applicable for better design and analysis of slopes as well as monitoring for slope failures and reinforcements of potential failure areas. 4 CONCLUSIONS
A localized probabilistic approach for slope stability analysis was developed by combining the 3-D deterministic method of slope safety calculations and the sequential Gaussian co-simulation technique for the localized probabilistic site characterizations. Both local and global pf’s give better methods of designing and analyzing slope stability problems over the deterministic methods. Both pf’s can be applied immediately for the risk 1087
Figure 3. The critical failure surface and contours of the local probability of failure. assessment of slope failures and the uncertainty analysis of the geotechnical site characterizations. The pf calculation is simple enough, as shown in the case study, to be a routine engineering practice for slope analysis. The localized probabilistic approach can be easily extended to other geotechnical structures such as dams, foundations, retaining walls and underground openings including mines, tunnels, and caverns. The local pf provides a realistic picture of slope stability conditions and locates the vulnerable areas where the local failure is likely to initiate. It is applicable for slope reinforcements and remediations. The pf is the conditional probability of failure that was conditioned to the available information, the state of knowledge of a slope and the design parameters. Two important design criteria are included in pf; the design safety factor and the desirable confidence level in the slope analysis. The single deterministic safety factor of a slope is not enough to analyze the slope stability, as seen in the case study where a stable safety factor of 1.15 was compared with pf = 39.5%.
Hovland, H. J. 1977. Three-dimensional slope stability analysis method. ASCE J. Geotech. Eng. Div. GT9:97 1-986. IMSL STATLBRARY. 1991. Version 2, International Mathematical and Statistical Libraries, Inc., Houston, TX. Pumjan, Sunthorn. 1998. A localized prohabilistic approach for slope stability analysis, Ph.D. Dissertation, Michigan Technological University, Houghton, MI, 180 pp. Pumjan, S. & D. S. Young. 1999. The localized probabilistic site characterization in geotechnical engineering, Proc. o j 37'' U.S. Rock Mechanics Symp., Rotterdam:Balkema (in press). Tatsuoka, M. M. 1971. Multivariate analysis, 3 10 pp, New York: John Wiley & Sons. Young, D. S. 1985. A generalized probabilistic approach for slope analysis, International Journal of Mining Engineering, 3:215-218.
REFERENCES Almeida, A. S. 1993. Joint simulation of multiple variables with a Markov-type coregionalization model, Ph.D. Dissertation, Stanford University, Stanford, CA, 199 pp. Deutch, C. V. & A. G. Journel. 1992. GSLIB: Geostatistical Software Library, 340 pp, New York: Oxford University Press. 1088
Slope Stability Engineering, Yagi, Yamagamid Jiang 0 1999 Balkam, Rotterdam, ISBN 90 5809 079 5
Probabilistic analysis of structured rock/ soil slopes - Several methods compared Dawei Xu Department of Mine Engineering, BHP Iron Ore, Newman, WA., Australia
Robin Chowdhury Department of Civil and Mining Engineering, Wollongong Universiv, N.S.W , Australia
ABSTRACT: This paper presents the applications of the probabilistic analysis method to structured or inhomogeneous rocWsoil slopes via a window’s version computer package, Proslope@Version 1.0,developed by the first author. Through a realistic case study of a slope in an open cut mine, three approximate approaches for the probabilistic analysis, ie Monte Carlo Simulation method (MCSM), conventional First Order and Second Moment method (FOSM) and Hasofer & Lind’s First Order and Second Moment method (FOSMHL), are applied to examine the similarities and differences among them. Five most widely used probability distributions are employed to simulate the basic input random variables for MCSM and to examine the influence of these distributions on the failure probability of a slope. Meanwhile, the influences of the correlation between cohesion and internal friction angle in the same geological domain and uncertainty of the groundwater level on the failure probability of a slope is also closely examined.
1 INSTRUCTION
Monto Carlo Simulation Method (MCSM), Conventional First Order and Second Moment Method (FOSM), e Hasofer & Lind’s First Order and Second Moment Method (FOSMHL), and Point Estimate Method (PEM). Through a realistic case study of a slope in an open cut mine, this paper presents some interesting and significant findings for applying the first three approximate methods on the basis of a computer package, Proslope@ Version 1.0,developed by the first author. Proslope@ Version 1.0can be used to estimate the reliability index and corresponding failure probability of structured or inhomogeneous rocMsoil slopes. The mechanical properties of rock and /or soil materials such as cohesion, friction angle and unit weight in different geological domains can be considered as the basic input random variables for this package. The impact of the groundwater level uncertainty on the failure probability of a slope can also be taken into account. Five widely used probability distributions, Normal, LogNorm, Beta, Triangle and Uniform, are integrated in Proslope@ Version 1.0 for the user to carry out Monte Carlo Simulation. The factor of safety of a slope is calculated by a ‘rigorous’ limit equilibrium solution, 0
In general, it is inevitable for geotechnical engineers to deal with uncertainties, which may be associated with geological structures, rocWsoi1 mechanical properties and groundwater, when they conduct the stability assessment of rocWsoil slopes. However, the conventional deterministic approaches for the rocWsoi1 slope stability analysis cannot quantitatively consider these uncertainties. As a result, the failure risk assessment for rocWsoil slopes cannot be carried out systematically by these methods. For open pit mining, these uncertainties not only impact on the stability assessment of slopes but also influence mineral recovery and financial returns. To avoid the limitation of the deterministic analysis approaches, a number of geotechnical engineers and research workers, such as Wu and Kraft (1970), Moss and Steffen (1978), Priest and Brown (1983), Miller (1984), Chowdhury and Xu (1 992), Xu (1 997), Xu and Chowdhury (1998), have developed probabilistic methods for the stability and reliability assessment of rocWsoi1 slopes over the last three decades. To date, the most widely used approximate estimation methods for the probabilistic analysis of rock/soil slopes are,
1089
Morgenstern and Price (M&P) method, in Proslope@ Version I . U.
2 MORGENSTERN AND PRICE METHOD Within a deterministic framework associated with the limit equilibrium concept, it is customary to define a factor of safety, FOS, as an index for stability, safety or reliability of a rocldsoil slope. Morgenstern and Price (M&P) method (1 965) is one of the most popular among relatively ‘rigorous’ generalised procedure of slice (GPS) methods associated with the limit equilibrium concept. In this method, two values of the factors of safety which are based on both force and overall moment equilibrium equations need to be calculated as follows
In slope probability analysis, a performance function. G(X) = (FOS-I). is always used to describe that the slope is in a ’safe state’ (G(X) > 0) or in a ‘failure state’ (G(X) < 0). Therefore, the failure probability of a slope can be defined as follows:
P,
=
Pr [G(X) < 01 or P,
=
Pr [FOS < 11
(4)
The performance function, G(X), is primarily related to the slope layout, rocMsoil material profiles, the slip surface location, the mechanical properties of rocMsoil materials (unit weight, cohesion, friction angle) and the groundwater or piezometric level. The reliability of a slope may be designated by a
Based on the assumption that the FOS follows a standard normal distribution, the relationship between Eqs. (4) and ( 5 ) is:
~ ( , A X +(w, , + AT, - ~ , ) t a n ~ , ) y , , , ~ s e c c l ~ / ~ , , ~ F ,= I;’_--_ (2)
in which, the relationship between interslice normal and share forces E, and T, is assumed as, T, = I f (x,)E,
3.1. Dq%itions ofperformctnce firnction, fuilzrre prdmbil ity cind re 1ici hi1ity index
P,
=
co(-p,
3.2 Basic input random variables for slope probability analysis
(3)
where, I is an unknown coefficient and f (x,) is a prescribed function with respect to coordinate x,. By adjusting the value of the coefficient 2, a relatively ‘rigorous’ solution, i.e. F f = Fm, can be obtained. This means that the factor of safety has to be calculated by an iterative procedure.
The basic input random variables for Proslope@ Version 1.0 can be the mechanical properties of rocMsoil materials and upper bound level of groundwater or piezometric surface. The uncertainty in identifying the failure mechanism of rocMsoi1 slopes is not taken into account by this package. However, if the true percentage of the failure model is designated as Plnode1 then the failure probability for a slope may be expressed as follows:
3 PROBABILITY FRAMEWORK OF SLOPES As the M&P method is an inexplicit mathematical formula with respect to the basic input variables, it is impossible to determine the accurate mean value, p, standard deviation, G, of FOS and failure probability P,. In general, some approximate approaches mentioned in the previous section must be used to conduct the probabilistic analysis of slope stability.
It is obvious that when the errors in identifying the = 1, the failure failure model can be ignored, ie Pmode, probability of the slope will then be equal to P,.
3.3 Monte Carlo simulation method (MCSM) For the slope probabilistic analysis, the probability distribution of the basic random variables and performance function or factor of safety can be
1090
simulated by the Monte Carlo simulation technique with a large number of calculations. Within Proslope@ Version 1.0, the skewed Triangle and Beta distributions can be simulated by assigning the distribution parameters. Moreover, for the Normal and LogNorm distributions, the distribution tail can be truncated by setting up the upper and lower boundaries of the random variables. The empirical failure probability of a slope can be defined by the ratio between the number of values of the FOS < 1 and the total number of simulated FOS values. The empirical distribution of FOS can also be examined with a histogram. The first and second statistical moments of FOS can be estimated by the following equations:
where, N is the number of simulations and Fi is a simulated FOS. The results obtained from Eq. (8) can then be used to calculated the reliability index and corresponding normal failure probability of a slope via Eqs. ( 5 ) and (6).
minimum distance from the surface G(X) = 0 to the origin of the uncorrelated reduced random variates. The formulation for calculating reliability index 0 1 4 ~ can be written as follows:
in which,
(
*
is the gradient vector at the most
x:)
on probable failure point x*= (x;,xf,. . . , failure surface (G(X*) = 0) and pLs is the vector of the mean value of the basic input random variables, respectively. As the performance function based on M&P‘s method is inexplicit the true failure point on the failure surface is initially unknown. As a result, the reliability index PHLneeds to be obtained by an iterative procedure which has been thoroughly discussed elsewhere (Chowdhury and Xu, 1992). The failure probability of a slope can then be calculated by Eq.(6) once p1.1~is obtained. It is necessary to note that PllL will be equal to when G(X)is a linear function.
3.4 First order second monienl (FOSM) In this method, the first and second moments of a performance function G(X) can be approximately estimated by the following equations:
in which, [C] is the matrix of co-variance of the basic input random variables, and VG is the vector of the partial derivatives of the performance function about the basic input random variables at their mean values. Once the first and second moments of the performance function of a slope are obtained by Eq.(9), the reliability index and corresponding failure probability of this slope can then be calculated by Eqs. (5) and (6) respectively.
4 REALISTIC CASE STUDIES One realistic cross section from an open cut mine is selected to conduct the probabilistic analyses on the basis of the different assumed cases. As shown in Figure 1, the most likely failure surface in this cross section may consist of two faults at the upper and middle parts of the slope and then cross and follow shale bedding planes at the lower part of the slope.
Phreatic Surface
,
3.5 Hasofer and Lind’s FOSM Hasofer and Lind (1 974) proposed an alternative definition for the reliability index which is designated here as pk,~ and is defined as the
t
’
L
Cross Bedding Planes
Figure 1. A cross section for slope stability and probability analysis.
1091
Table 3 Probabilistic analysis results in case 1.
Only one rock unit, the McRae shale, is above this potential slip surface. 10,000 simulations were used for MCSM in the following case studies.
4.1 C'use 1: basic input data rmdoni vuriubles
(IS independent
1
norniul
FOSM
I-1,
1.1285
I
I
FOSMHL
1
MCSM 1.1314 0.1000
0,
0.1014
Reliability Index
1.2668
1.2997
1.3 I44
Normal P,
10.26%
9.72%
9.44%
In this case, the mechanical properties of rock materials and the upper bound level of groundwater are assumed to follow the normal distribution without significant tail truncations. The mean values, standard deviation, upper and lower boundaries for these basic input random variables are listed in Table 1 and 2. Based on the deterministic analysis, the factor of safety for this slope is 1,1285. Table 1 Shear strengths for probabilistic analysis. Central
1
Cross
I
Along
1
Figure 2. Probability distribution of FOS for case 1.
4.2Case 2: Infliience
To investigate the influence of correlation between c and 4 on the failure probability, the magnitudes of pc4, the coefficient of correlation between c and 4 , were assumed to be equal to -0.5 and 0.5 respectively. The mean values and standard deviations for all the basic input random data are same as that in Tables 1 and 2. The calculated results are shown in Tables 4 and 5 respectively.
*The unit of cohesion is kPa. Table 2 Parameters of unit weight and phreatic surface for probabilistic analvsis. CI
Unit Weight Phre. Surface
0
25 (kN) 1.5 (kN) 512 6
111
2 in
IJB
LB
29 (kN)
21 (kN)
520 m
505 m
of correlation coqficient (pccJ
4 on the value of Pi.-
between c and
Table 4 Probabilistic analysis results for pci= -0.5. The results obtained from the probabilistic analysis are presented in Table 3. The histogram of the simulated FOS is shown in Figure 2. For the given conditions, Table 3 indicates that the failure probabilities obtained from three methods are not significantly different. However, the empirical Pi based on MCSM is lowest in comparison with that of FOSM, FOSMHL and normal P, of MCSM. In contrast, P, from FOSM is higher than that of other two methods whereas P, obtained from FOSMHL is quite close to that of the normal P, from MCSM. The difference of P, between FOSM and FOSMHL means that the performance function based on M&P's method is not but close to linear with respect to these basic input random variables. Figure 2 shows that the histogram distribution of FOS generated by MCSM is very close to the normal distribution. As a result, the normal P, is very close to the empirical P, for MCSM.
I
FOSM
I
FOSMHL
I
MCSM
0,
0.0812
Reliability Index
1.58 18
1.6338
1.6517
Norinal P,
5.68%
5.1 1%
4.93%
Empirical P,:
I
0.0805
4.56%
Table 5 Probabilistic analysis results for pci= 0.5.
1.1377
The results in Tables 4 and 5 indicate that there are also no significant differences amongst the three approximate methods for the given conditions. By comparing Tables 3, 4 and 5, it is of interest to note
1092
that the failure probability of the slope will be significantly influenced by the magnitude of pcd.The failure probability will be decreased while pcl is being changed from the positive magnitude to negative magnitude. The normal and empirical P, values for MCSM shown in Tables 4 and 5 are also very close to each other whatever the magnitude of pcc is. Figure 3. Histogram distribution of one cohesion truncated at left tail.
4.3 Case 3: Influence of upper and lower bound of c and $ on the value of PI; The impact of the tail truncation for the normal distribution is examined in this section via setting up that the upper and lower boundaries of c and 4 are equal to 1-1 rf: 2 0 respectively. Other conditions are same as that in Sections 4.1 and 4.2. As FOSM and FOSMHL cannot consider the influence of the upper and lower bounds on the value of P,, MCSM is alone used in this case. To compare the results from MCSM, the results of FOSM and FOSMHL, shown in Table 3, are copied into Tables 6 and 7 respectively. Table 6 Probabilistic analysis results for a given lower boundaries of c and b (U - 20). FOSM
FOSMHL
MCSM
PF
1.1285
1.1447
0,
0.1014
0.0967
Reliability Index
1.2668
1.2997
1.4962
Normal P,
10.26%
9.72%
6.73%
FOSM
PF
I
0 1
FOSMHL
I
0.1014
1.1213
I
-
Reliability Index
1.2668
1.2997
Normal P,
10.36%
9.72%
Empirical P,
MCSM
1.1285
I
0.0955
1.2694 10.22% 10.16%
Figure 4. Probability distribution of FOS for c and 4 truncated at left tail.
4.4 Case 4: Influence of dflerent distributions of basic input random vai*irrbleson PI;-
5.89%
Empirical P,
The histogram distribution of FOS accordance with the basic input random data which are truncated at the left tail is presented in Figure 4. Form Figure 4, the probability distribution is still very close to the normal distribution under the given conditions.
I
In this example, unit weight, cohesion, friction angle and upper level of phreatic surface are assumed to follow the triangle, Lognorm, beta and normal distributions. The skew coefficient for the triangle distribution is 0.4. The mean and standard deviation values for other random variables are same as that in the Case 1. However, the upper and lower bounds need to be changed to some degree to meet the requirement of the particular distribution. The histograms for one set of simulated cohesions. internal friction angles and unit weight are respectively shown in Figure 5 , 6 and 7. MCSM is alone used to estimate the failure probability of the slope. The mean and standard deviation of FOS are 1.1302 and 0.100 1 respectively. The reliability index, normal P, and empirical P, are 1.3009, 9.66% and 8.94%. Figure 8 shows that the distribution of FOS is quite close to the normal distribution in this case. However, it does not mean that the distribution type of the basic random variables do not impact the distribution shape of FOS. Some
1093
geotechnical engineers to carry out the probabilistic analysis of structured and inhomogeneous rocMsoil slopes. Through a realistic case study on the basis of this package, it is of interest to note that the failure probability of slopes will be impacted by the shape of the probability distributions associated with the basic input random variables. It can also be found that the results calculated from FOSM and FOSMHL are quite close to that from MCSM for some cases. Consequently, it is possible to use a searching technique in conjunction with FOSM or FOSMHL to find the most critical slip surface associated with the maximum failure probability.
Figure 5. Distribution for one set of simulated cohesions.
6 REFERENCES
Figure 6. Distribution for one set of simulated internal friction angles.
Figure 7 Distribution of simulated unit weight.
Figure 8 Probability distribution of FOS for different distributions of basic input random variables. cases have indicated that the distribution of FOS is no longer close to the normal distribution.
Chowdhury, R.N. & Xu, D. 1992. Reliability index jor slope stability assessment - two methods conipared, J. Reliability Engineering and System Safety, Vol. 37, pp. 99-108. Hasofer, A.M. & Lind, N.C. 1974. Exact and Invariant Second Moment Code Forinat, Jnl. Engg. Mech. Div., ASCE, Vol. 100, pp I 1 1 - 121. Probabilistic rock slope Miller, S.M. 1984. engineering, Research Report, Dept. of The ARMY, U.S. Army Corp. of Engineering. Morgenstern, N.R. and Price, V.E. 1965. The analysis of the stability of general glip surface, Geotechnique, Vol. 15, pp.70-93. Moss, A. S. E. and Steffen, 0. K. H. 1978. Geotechnology crizd probability in open-pit mine planning, in Proceedings 1 l t hComm. Min. & Metal. Congress, Hong Kong, pp. 543-550. Priest, S.D. & Brown, E.T. 1983. Probubilistic stability analysis of vciricrble rock slopes. Trans. Instn. Min. Metall. 92, pp. A1-A12. Wu, T.H. & Kraft, L.M. 1970. Safety Ancrlysis of Slopes. Proc, Am. Soc. Civil Eng., J. Soil Mech. Foundation Div., 96: pp. 609 - 630. Xu, D. 1997. Probability ancrlysis of rock rind soil slopes in [I conzplex geology environnient, Mining Geology Conf., Launceston, Australia, lo- 14 Nov. Xu, D. & Chowdhury, R.N. 1998. The probctbilistic cinulysis of three dimensional wedge stability several inetods compared, Proceedings of ICEM2, Wollongong Australia, 10-13 Feb, pp. 1019- 1025.
5 CONCLUSIONS A comprehensive computer package has been developed to provide a powerful tool for 1094
Slope Stability Engineering, Yagi, Yamagami 8 Jiang (c) 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Reliability analysis and risk evaluation of the slopes of open pit mine Qing Yang, Jiankui Jiao & Maotian Lum Department of Civil Engineering, Dalian University of Technology, People’s Republic of China
Dazhai S h Baotou University of Iron and Steel Technology, People’s Republic of China
ABSTRACT: In this paper, the progressive failure probability of the slope of the Baiyunebo Open Pit Mine has been evaluated using Monte-Carlo method, based on the 2-D reliability model of progressive slope failure. Then according to the calculation results, the risk of investment and benefit is assessed through applying economic decision theory to slope engineering. 1 INTRODUCTION With the development of the techniques of the surface mining, the height of the slope increases considerably. At the same time, the contradiction between the stability and the economics gets more serious. When the slope angle is increased 1 degree, the great economic benefit can be obtained, however this means that we must undertake much more risk. The development of the reliability analysis theory makes it possible to establish the relationship between the reliability and economic analysis. In the processes of the reliability and economic analysis, the results must be undertaken some extent risk since the most of data are indeterminate. In this paper, the progressive failure probability of the Baiyunebo Open Pit Mine is evaluated using Monte-Carlo method, based on the 2-D reliability model of progressive slope failure. On the basis of the results of reliability analysis, the risk of the unreliability costs of the slope of this mine is assessed using the Monte-Carlo method.
2 RELIABILITY ANALYSIS OF THE PROGRESSIVE FAILURE 2.1 Basic concepts The fact that movements of the slope can occur for many months or years before the slope finally collapses suggests that the failure process is progressive rather than instantaneous as is assumed in most forms of stability analyses. The fact that the parameters along the sliding surface are not uniform, such as, stress, strength of rock masses and the force
of the water seepage, play an important role in this progressive failure process. In this process, the failure starts from the local area which has the biggest failure probability and the failure may spread and propagate to the neighbor area and then the whole slope collapses finally, or the failure stops to propagate and the whole slope is also stable. In the process of the analyses, when the local area is in failure, the strength is reduced to residual strength and the shear stresses surplus are undertaken by the neighbor area units. The unfailing area still has the peak strength. The failure probability is used to evaluate the slope stability and the conditional probability is used to express the possibility of the failure propagation. The threshold value of the progressive failure probability is determined by considering many factors. At the present time, the failure probability is determined as 3 X 10-2-- 1 X 1O-* for the whole slope and 10 X 10-2for the bench slope. The level of risk undertaken can not be taken higher. Because of the lower technique conditions and the lower the ability of compensation to the risk, and considering the factor that slope of Baiyunebo Open Pit Mine is higher and the most rock masses have not been exposed, the engineering geology is of indetermination. But in the 2-D condition, the failure probability is conservative, therefore the failure probability may take higher in some extent. In this paper, the failure probability is taken as 5 X 1O‘2 for the whole slope and 8 X 10-*for the slice failure probability. Because the importance of the slice is less than that of the whole slope and higher than that of the bench slope.
1095
2.2 Mathematical Model Taking the surplus-thrust method of slope stability analysis as an example, the calculation method of progressive slope failure is illustrated as below. Suppose that the slope is divided into a number of vertical slices, which are numbered 1 to n from beginning at the toe. Figure 1 shows the force condition of the i-th slice. In the analysis, the following hypotheses have been made: @the slice’s weight W,, the normal force NI and the seepage force U, are all through the slice’s barycenter; @the equivalent horizontal force SUIPQ, caused by blasting acts on the slice’s barycenter; @the shear strength of sliding plane is generated by the MohrThen, , + ~the . Coulomb criterion; @y,=a,, ~ , + ~ = a force equilibrium equations of the i-th slice are written as follows.
S,= W, sina,+Q,,, ~os(a,+~-a,)-Q, SUIPQ, cosa,
(1)
N‘, = W, cosa,+Q,+,sin(a,+,-a,)-U, -SUIPQ, sina,
(2)
+
In which a, and a,,, are represented the dip angle of the sliding faces of the i-th slice and the (i+l)-th slice respectively; Q, and Q,+I stand for the interslice thrusts of these two neibouring slices respectively, and given by
zero, that is Qi+l= 0. The margin of safety SM is defined as the difference of the shear resistance and the shear force. Based on Mohr-Coulomb criterion, the margin of safety of the i-th slice can be written as:
+
SM, = c,L, N’, tg$,-S,
(5)
where L, is the length of the sliding plane of the i-th slice; c, and $, respectively represent the cohesion and internal friction angle in the length of L, . Under the condition that the i-th slice is not failure, these two strength parameters are the means of the peak strength parameters c,, &, otherwise, they will be the means of the residual strength parameters c,, $r The margin of safety indicates the stability of slice or slope. If it is less than zero, the slice or slope is in an unreliable state. 2.2.1 Determination of The Initial Failure Slice At the beginning, each slice undertake the peak shear strength. The initial margin of safety of the i-th slice is SM,= c,,L,+N’,tg$ ,-S,, and its initial failure probability is P,=P[SM,d)O]. The initial failure slice is defined as the slice which failure probability is the largest and more than the given failure threshold in all of the slices. It is thought that the failure of slope starts from this slice and expands one by one. And the failure of slope will not take place if the largest failure probability is less than the given failure threshold.
1-1
Q, = ~ [ c , L ,-SUIPQ, COS^^ +(W, COS^, -UJ 2.2.2 Calculation of Progressive Failure Probability
J=l
- SUIPQ sinaJ)tg4, - W, sina,]
(3)
I1
Q,+]= C [ W , sina, -SUIPQ, cosa, -c,L, ,=I+]
,
- (w, cosa, - U , - SUIPQ sina, ) tg4, J
Suppose that the slices from j to k are failure, then their strengths are reduced to their residual strengths. In order to undertake the surplus shear, the failure probabilities of their neighbor slices have to increase.
(4) k
It is necessary to point out the problem about the value of Q,+,.The case that QI+lpec;ificatims of' safety requirements ibr structurLs" (1981), the DIN 1054 100 (1995) edited by the German standards institute (DIN) and the Eurocode {CEN 1994) In this context the application of partial fFctors of safety to the characteristic values of the basic random variables gives the design values Based on the works carried out at the Department of Geotechnical Engineering, University of Weimar (Nottrodt 1990, Weiss 1991), the statistical data of the basic random variables and partial kctors of safety are g,;venin Table 1
and probabilistical (reliability indexes) iiivestigations.
'I able 1 Statistical data of the basic 1. ai iables and p;utial factoi s of safet) __
Ccefficimt I)islributioll l'ai tin1 factor of r'ariatioii of safL.ty Fi-iction angle cp [ 1 Cohesion c [LN/ni'] Suichaige q [kN/ni'] Sod weight Y KN/ni'l
7 5 %, 25 % 40 96
5 ?6
Log-iioimal 1 25 Log-normal 1 6 Extiem 13
Noinial
10
2 INCLINATION OF THE INTE,RNAL SLIP SURFACE The failure of the lower anchor of a double-propped retainins wall will be investigated here as an example (Fig. 2).
Figure 2 Failure mechanism of the lo\~eianchor of a doublcpropped i etaiiiiiig w:dl
The failure mechanism, according to the kinematical theory, as shown in Figure 2 is described mechanically with the following limit state equation : E p ~ ~ ~ ( 6 p - +0C,COSC~ 2 + ~ )+ C,,COS(~,,+~,-cp) + +A,cos(B,-c~ + e ) + C , W , C O S- ~ - Ii(G,+P,)~in(B,-cp) + [(G,+P2)~h(Ol1-cp)+ (1) krguie 1 Failure in the deep slip sui-face (a) Failuie model alieiK-anz (1 053), @) Simplified model for iiilection anchor (c) Kiiicmatical model foi- inlection anchoi
The system safety or reliability is expressed in terms of the reliability index l3 The computation and evaluation of the reliability index I3 is based on the fii-st-order reliability method The limit state equations are 1he basis of all the deterininistical (anchor lengths)
+AlcOs(O~l-cp+e)]",I
=o
in which sin(0 12 +e2-2 c p ) m2=
,,
S i n ( 0 12 +e -2 Cp
(2)
More details to the formulation of the limit state equatioll are given in Belabed { 996a)
1116
The critical inclination of the internal slip surface is computed iteratively with the following fimction :
an inclined internal slip surface is placed at the intersection point of the deep slip surface with the injection anchor. This is the exclusive dityerence between both models.
(3)
3.1.1 Case I In this case, no anchor force acts on the rear active slip body number (2) (see Fig. 4). Comparative analyses (Belabed 1996a) have given that this rear active slip body (the resultant Qlzin Fig. 4) may be replaced with sufficient accuracy by the active earth pressure force E, (see Eq. 4). By the simplilied model, the active earth pressure force E, acts on the fictitious anchor wall.
where A 8 is a term that will be varied. It takes positive and negative values. I<esults (Fig. 3) show that the critical inclination (OI2) is, with sufficient accuracy, equal to the ixtclination of the active slip surface according to Coulomb (BI2=n/4+(p/2)From Figure 3 it is apparent that the more the inclination ( 8 , J differs fi-om the inclination of the active slip surface according to Coulomb, the greater is the syC..:ciiireliability index (0).
Figwe 3. 1)cteimination of thc critical inclination of the internal s l y :;urf21cc.
Figure 4. Kmematical model by the failure of the lower anchor of B double-propped retaining wall (Case I ) .
The same investigations were carried out for triplepropped retaining walls. The most critical internal slip siirhce is approximately 5" steeper than the active slip surface according to Coulomb, when the internal slip surface cuts two anchor rows. This difference could become larger with the increase of the intersected anchor rows (by more than three anchor rows). Thus, firrther investigations are necessary.
The limit state equation corresponding to the kinematical model (Fig. 4):
3 APPROACH WITH THE FICTITIOUS ANCHOR
WALL In this study a comparison between the kinematical model nearst to reality which is based on the kinematical theory of rigid bodies and the simplified model is carried out. By the simplified model (Fig. 1b), a fictitious vertical anchor wall is placed at the intersection point of the deep slip surface with the injection anchor. By the kinematical model (Fig. Ic),
Epcos( 8 p- 8, + 19 ) + C,CO s 'p + C,2cos (-5+ 0 4 -
i(G, +P,)sin(8,-'p) + E p s (-+--@,> x 319
4
2
, T) + 2 -
}=0
For more details to the formulation of the limite state equation see Belabed ( 1996a). The results are illustrated in Figure 5 They show a satisfactory agreement between the kinematical and the simplified models both deterministic (anchor lengths) and probabilistic (D-level). In this case, the simplified model is a satisfactory approach for the kinematical model. The use of a fictitious vertical anchor wall instead of the inclined internal slip surface is a good approximation.
1117
showed that the resulting (Q2) which acts on the internal slip surface is much greater than the active earth pressure force (Ea). That's why, contrary to case (1) the rear active slip body (2) may be not replaced by the active earth pressure force. Otherwise, it would lead to undermeasurements. The approach with the fictitious vertical anchor wall (simplified model) is misleading because the effect of intersected anchor forces is neglected (see Fig. 7).
II
E8
L-.l
8
. C
c
Simplified model 4010
3515
30110 25/20 cp [o]/c[kN/m2]
20135
Figure 7. Siiiiplificd model by the failure of the lower anchor of a retaining wall with three rows ofanchors. (case 2). 11010
3515
30110
25/20
20135
The limit state equation of the kinematical model (Fig. 6):
"1/c [kNlm 2]
(p [
Figure 5 Anchor lengths and reliability indexes (cas : 1)
3 1.2 rase 2 The anchor forces (A, and A,) act on the rear active slip body number (2) because the corresponding anchors are intersected once from the internal slip surface (see Fig. 6). Thus, the intersected anchors prevent partly a soil loosening.
Figure 6 Kmematical model by the failure of the lower anchor of a retaining wall with three IOW of anchors (Case 2)
(:omparison ofca]culations (Belabed 1996a) have
E p c o s ( y j 3 + c p +clcos(p ) +c,,cos(e,,-cp +e3)+ + C2m3cosq+ (A, +A,)cos (03- (p + e ) - I (G1 +P,)Sin(e,- cp ) + [ (Gz+Y,)sin - cp) + ( 5 ) +(A1+A2)cos(8,,-(g + e ) ] m 3 i =0 ~
in which
For more details to the formulation of the limit state equation see Belabed (1996a) The results are presented in Figure 8 which shows in detail the courses of anchor lengths and reliability level (l3-level) for that concerned failure model By the failure of the medial anchor, the simplified model yields longer anchor lengths than the kinematical mods1 The maximal difference is about 7% The difference is clearly larger (appl-oximately 20%) by the failure of the lower anchor (Fig 8). Nevertheless, both models yield equal reliability indexes l3 (Fig 8, In this case it to advise, the simplified model not instead of the kinematical model
1118
to investigate since the simplified model underestimates here the system safety (reliability). A more exact modelling of these failure mechanisms is exclusively guaranted with the kinematical m d e l because therewith the system safety is not underestimated and correct used. Furthermore, an economical design of anchor lengths (up to about 2.00 m shorter) is to expect by the investigation of failure mechanisms with the kinematical theory. Finally, it is also established that the influence of anchor forces on the system safety is significant by anchored retaining walls with several anchor rows.
which acts direct behind the front active slip body (Fig. 9) and corresponds loads as a result of e.g. adjacent buildings. Their value and width (b) will be varied. The statistical data of the surcharge q' are assumed to be identic with those of the unlimited surcharge q (Table 1). The results of the investigations are summarized in Tables 2 and 3.
Figure 9. Influence of a bolh sides limited surcharge
Kinematical model Simplified model 4
+ - t - - - - - - r ~ i
4010
3515
30110
25/20
20135
cp [o]/c[kN/mZ] 5
I
Y
a 4 x a ,v 1-
x3
.-*
.I
w
.a
.*
T 2 p?l
Kinematical model Simplified model 1
4010
3515
I
I
30110
25/20
20135
It can be seen from the tabulated results that for q'=O, both models, the kinematical and the simplified models, provide equal anchor lengths and reliability indexes (0). For q' = 100 kN/m' and b = 3 m, the simplified model underestimates the system safety. Then it yields longer anchors and smaller reliability indexes than the kinematical model. This is the result of the assumption of a fictitious vertical anchor wall at the point of the intersection of the deep slip surface with the anchor. Computations have showed that the additional earth pressure, in consequence of q', on the fictitious anchor wall is greater than that on the inclined slip surface. This becomes clearly with very high loads (9' = 100 kN/m2). Finally, it is established that by high loads the kinematical model is exacter than the simplified model. The consistency of the kinematical rigid bodies theory could be so confirmed.
'p [o]/c[kN/m2J
Figure 8. Anchor lengths and reliability indexes (case 2). Table 2. Influence of a both sides limited surcharge on anchor lengths La [m].
After the French recommendations (Habib 1989) the simplified model may be here as well investigated but with consideration of the influence of the intersected anchor forces on the earth pressure force behind the fictitious anchor wall. 3.2 Influence of a both sides limited surcharge
The analysis was performed for a both sides limited and uniformly distributed vertical load, surcharge q',
cp/ol/clkN/mz]
30/10 25/20 20/35 q' = 0 Kinematical model 5.53 6.38 7.47 8.23 8.28 Simplifiedmodel 5.53 6.38 7.47 8.28 8.52 q'=30kN/m2 ; b = 1.50m Kinematical model 5.63 6.49 7.60 8.39 8.49 Simplifiedmodel 5.74 6.60 7.72 8.57 8.86 q' = 100 kN/m2 ; b = 3.00 m Kinematical model 6.07 7.03 8.25 9.20 9.52 Simdifiedmodel 6.76 7.78 9.07 10.12 10.65
1119
4010
35/5
'f;ible 3. lnlluciice of a both sides limited surcharge on reliability inclex I3 [-I. cp["l!clkN/m2]
40/0
35/5
30/10 25/20 20/35 (1' = 0
Kiiicmatic:d model 2 72 Snn~11ific.dmodcl 2 72
3 23 3 40 3 23 3 39 (1' = 30 hN/mZ , Kinematical model 2 76 3 25 3 44 Simplifiedmodel 2 8 2 3 29 3 47 q' = 100 hN/mz , Kincmatical model 2 7 1 3 16 3 38 Simplrfiedmodel 2 48 2 78 2 97 _________- _ _ _ _ _ _ _
3 25 3 27
2 80 2 86
b = 1 50 3 30 2 85 3 35 2 94 b = 3 00 ni 3 36 3 0 1 3 06 2 92
-
4 CONCLUSIONS
I h s d on thc iesults obtained iii this work, the f'dlowing recommendations are proposed as an irniversal valid solution to t1v stability assessment of the overall system (wall-ground-anchor) of anchored I ctaining walls 1 Failure mechanisms should be investigated with the kinematical rigid bodies theory which yields the most realistical results 2! The internal slip surface may be, for simplicity, approximately replaced by a fictitious vertical anchor wd1, when the corresponding rear active slip body is subjected to no high loads (surcharges) and/or to no anchor forces 3 When at most one anchor intersects the internal slip siirface, the unfavourable inclination of the internal slip surface to the horizontal can be assumed with a sufficient accuracy equal to that of the active slip surface according to Coulomb 4 When more than one anchor intersects the internal slip surface, it is recommendable to assume tliat ihe internal slip surface is 5" to 10" steeper than the active slip surface according to Coulomb The studies have showed that the use of the statistic-probabilistical safety concept in structural design and stability assessment of structures is of a great importance Then it permits, compared with the conventional (global) safety concept, an objective comparison between several structure failures
Methcden. D.Sc. the.~is,I3auliaus-Uriiversi~~t Weimar,(ieiiiiany. I,. 1 996b. Genauere Modellierung delBelabed, Biucluneclinnismen bci mehrfach vci-ailkci-kii Stiltzwiinden. Uairteclimik 7 3 :776-780. CEN 1994. Geotechiiical design, general iules. Ei{i*opeaii C'oiiiiiiittee.kw Slc7iickli.iliznlioll ((-Ehg, 151so~:odc.I\Jedci.l:itiils Normalisatie-iiistituut ("I), Ilelft. I'restaiidai-cl (IINV). DIN 1054,100 1995. Siclierlicitsn:ichu.eisc in1 Ixrd- und Grundbau. I)eutsclies Iiisliliit fur Noiwii{tig. Criifjler, G. 1982.Anwendung des statistischen Sicheiheitskonzepts auf verankerte Wande und veiiiagelte W a d e . Z'oitr-upe deer. Buugr.rrr~dtagz~irg, Bruitiischtt~eig:49-82,Ileutsche Gesellschafl fur Geotechnik. Cidler, G. 1987. I4vnagelle Gelaiitlespiiiiige - li.ngveihdteii i i i i d S'lancisichei.~~/leil.VeriifTentlicliungeii des lnsliluts fur Bodeilmechmk und Felsmechanik, Universitiit Karlsruhe, I-Ielt 108. Galer, G. & (3, Gudehus 1989. Anchored Walls - Model tests and statistical design. Piweet/iiigs OJ [lie /I"' /iiteinnlioiin/ Coilfererice 011 Soilhfecliniiics uiitl1;oiriidalioii Ei~giiieeiiiig, Rco tie jaiieir~o:829-8~2. (hldschcider, M. 8r. 11. Kolymbas 1980. i3ci-t.cliiiung dcr Standsicherheit mehi-fachverankerter Stutz\v;ande. Geotrcliiiik 4: 156-164. GIilJSIUAU 198 1. (iiundlageii zur P'estleguiig \.on Sicherheitsanfordeiungen fur bauliche Anlagen. Published by Deiitsclies liistiti(t Ncmiriiiig (DIN) e. V.. Bcuth Verlag. Berlin-Kdn. I-Iabib, 1'. (Editor) 1989. I~ecoiiinieiic~~rtioiis for. /lie desigii, calculatioii, coiistiwctiori aiiti iiioiii[oi.iiig o j groiiiitl aiichoiages (translation from the French). A A : Balkeniaff~ottei-~~nl/I3ro~l~eld. Kranz, E. 1953. Ilbeer. die I >i.nrikei.iiiip voii ~ ~ J i l i i ~ ~ ~ ~ i ~ i l Mitteilungen Wassabau und 13aulimchung, I l d t 1 1, 2. Auilage, Berlin. Verlag 137ist & Sohi. Nottrodt, €3.-P.1990.Reitrag zur Einfuhi-ung seii~ipro\,ahilistisclier Methoden in der Geoteclmik. 11. Sc. diesis, 13auliausUniversitat Weiinar, Gennany. Weiss, W. 1991. Zur Siclicrheitsl~eive~tuiigini Giundbnu. I,'ei.~ffeiitlichiiiigeii dcs (;i.iiiidbmrriii.stitir/s del. I'ecliiiisclieii Uiriveixitat Bei.liii, tiefl 20.
REE;EREiNCES Ander.wn, W.F., 'I'.N Hanna . & M.N. Abdel-malek 1983. Overall stability of anchored retaining walls. Jozrlsinl cfgeolechnical eiigiiieering 1 1 : 1 4 16-1433. Delabed, I,. 1 995. Standsicherheitsuntersuchung zweifach der kinematischeii \~ei.ankerter Stutzwande mit Stai-rkiirpeimctliode. Geotechnik 19:17 1 - I 74. E:l:labed, I,. 19962. Zuverlassigkeitsuiitei-suchung des ?'ragsystems "i.nehrfach verankerte Stiltzwande" rnit probabilistischen
1120
11 Landslide investigations
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Slope Stability Engineering, Yagi, Yamagami& Jiang 0 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Methodological study of judgement on landslide occurrence Miau-Bin Su Civil Engineering Department, National Chung-Hsing University, Taichung, Taiwan
Lian-Chang Chan & Gee-ShorngLee Taiwan 2nd Engineering Station of Soil and Water Conservation, Taiwan
ABSTRACT: Landslides occurred at Li-Shan in central mountain area of Taiwan are divided into many large blocks action. In order to predict the stability of slopes in the future, time series analysis was applied to study the relationship between rainfall intensity and groundwater level fluctuation. Hourly data on rainfall in four of six monitoring stations were applied to form its ARMA(p,d,g) model. Then, transfer function for each pair of rainfall intensity and groundwater level change was searched. Discussion on the simulated result told that, for different hydrological situation, the transfer function defined are different. For each slope studied, its representing transfer function can be applied to predict groundwater level change inside slope, so as its factor of safety in regard to slope stability.
1 LNTRODUCTION Li-shan landslide is located in Li-shan village which is in the intersection of two main routes in central mountain area. As can be seen from figure 1, there
are Tai-8 which is the central cross-island parkway starting from west coast to Hwa-Lian county in east coast, and Tai-8branch route which starts from LiShan village to I-Lan county in the north-east part of Taiwan island. The landslide studied is a major
Figure 1. Geographical map for Li-Shan landslide location. 1123
disaster happened in April 1990 caused road service interrupted and some building in this area cracked and subsided. In order to resume this area for service and utilization, a remidiation project was planned and executed starting from 1995. The landslide was judged still active currently. In order to avoid sudden interrupt of road service, a predefined procedure for the judgement of landslide occurrence has to be set in advance. That is the goal of this research. What reported in this paper is part of the methodology in searching for a reliable method in predicting the landslide occurrence in the future. Geological study in this area showed the characteristic of old colluvium stratum and highly weathered slate. Figure 2 as the map viewing from the surface showed the landslide can be divided into many blocks, each has the shape of round or horseshoes. Tension cracks can be found in many locations. Because of the highly fractured surface, seepage action is very strong and some spring can be found seeping out from downslope in part of this area. Infiltration of rain water via percolation from surface and wild crack are concluded as the main factor influencing the stability of slopeland. Six monitoring stations were set to record the rainfall,
ground movement, and groundwater level fluctuation, etc.. In order to predict the change in stability of slopes based on groundwater pressure change, a methodology were set to use time series analysis for recorded rainfall and groundwater level change data to see its relationship so as to its influence to the stability of slopes. Figure 3 is a geological profile showing the material underneath. In Li-Shan area Geological material was formed by metamorphic rock mostly slate. Near surface, there are strata having different degree of weathering. Highly weathered stratum labelled ‘‘w1” is a soil like and highly impervious material as can be seen from boring sample. Groundwater fluctuation was concluded as the major cases of landslide in this area. The remedial action was concentrate on water control. Surface drainage system and groundwater control using horizontal drain, collecting well together with drainage gallary were planned.
Figure 2. Distribution of sliding blocks. 1124
Figure 3. Geological profile for B1 and B5 sliding block
2 TIME SERIES ANALYSIS Time series analysis was used to find transfer function between rainfall data and G.W.L. change using monitored data. After optimized transfer function is defined for each sliding block, it can be applied to perform forcasting work.
is left over to cover the rest influence of other hydrological factors. In here, data from B-9 station on 97/8/28 for rainfall and G.W.L. change are used to demonstrate the procedure of Time Series Analysis. In this storm, total accumulated rainfall was 247.3 mm and maximum G.W.L. change is 2.88m. The complete transfer function calculated is as following:
2.1 Transfer Function
0.0 1 63& 0.00926B4
Dynamic model was used to explain the relation between manipulated variable X and controlled variable Y. It can be defined as the transfer function between X and Y. A complete transfer function model can be written as (1):
yt'
1 - 0.9286 1 B 1
Xt4+
.................( 2 )
Sat 1 -0.33779 B + 0.49 I8 18B its transfer function is 0.0 1 6 3 4 - 0.00926B4
(1)
with U
=m(B)
: part of Y, used to describe
6(B)xt-b
6(B)
= -a : Disturbance term, nothing to
N
do
6(B) t-b
xt-l .....................
+ 0.01634
X,-l
(3)
+
set Y ,=Y ,' Ytbl X,., Xt-s : the real monitored data Yt': fitted value for G.W.L. The transfer function defined was then put into the original rainfall record together with observed G.W.L. change as in eq.(4) to perform the fitted value estimation for G.W.L. change. The result are 7
with X,. In order to define the relation between rainfall and G.W.L change within sliding blocks, only f
1 - 0.92861B can be expanded to Y , = 0.92861 Y t m l
(4)
4tB)
( y = o(B)x
yt'
) is explained, the disturbance term
1125
7
put into figure 4 to show the verification. As can be seen from the plot, data matches well. It shows the applicability of using transfer function to describe the relation between rainfall and G.W.L. change for each sliding block.
showed a rising section and a decending section. But, transfer function derived from B9-4 shows the greatest destructive potential so is chosen as the optimized transfer function to represent B9 block and in the future, can be applied to the forcasting work.
Figure 4. Verification of transfer function using fitted values for B9-4.
Figure 5. Unit hydrograph of G.W.L. change derived from four different rainfall record of station B9.
Same procedure as described in above is applied to other record of storm for the same monitoring station, as named B9-1, B9-2, and B9-3. Data used are tabulated in table 1 for each storm record. As can be seen from the table, delayed time gets smaller when rainfall intensity gets larger.
The chosed optimized transfer function was then applied to other recorded storms to see the simulation accuracy. Figure 6 and 7 are the simulated graph for B9-1 and B9-3 individually. The result are mostly acceptable.
3 RESULT 2.2 Optimization of transfer function Four transfer function derived from different storms were compared to find the most representative one for data from B-9 station. The procedure for optimization is as followings. A unit rainfall lasting for an hour is applied to each transfer function in above to calculate its response as an unit hydrograph which are shown in figure 5. The response of G.W.L. change along time is shown. Every hydrograph Table 1. Data for analysis for B9 station Observation Accumulated Initial Date rainfall (mm) G.W.L.
The above mentioned procedure were applied to data from four monitoring stations and its result are concluded as following. Table 2 showed data used for analysis for each sliding blocks. And, table 3 showed the optimized transfer function for each monitoring station. Unit hydrographs were then drawn in figure 8 to show the response characteristics for each sliding block.
Maximum ARIMA G.W.L. Model Fluctuation
Transfer Function
0.011 1
B9-1
96.5.20
110.5
-27.64
1.06
(2,1,0)
B9-2
97.6.10
179.0
-26.99
1.90
(2,1,0)
4 = 1-0.98383B xt-5 Y--
B9-3
97.6.14
156.4
-26.35
1.13
(0,1,5)
B9-4
97.8.28
247.3
-27.38
2.88
(6,1,0)
1126
yt’
0.01213 1 -0.9701 1 B
*‘E’
0.0 1408 1-0.96668B xt-3
6 3 4 0.oO926B x= 0.011-0.9286 Xt-1 1B
I
1 4 ,
1 1 h
E
2
$
lI.x
1
I
Date:l996.5.20 Accumulated Rainfall: 1 1 0 . 5 m m Initial G . W . E . : 2 7 . 6 4 m
1.1
Initial G.W.E.: 26.
1 h
E 0.8
v
2
11.6
0.6
0; 0.4
11.4
0.2
11.2
0
0
-t-co~~--~am-t-m Q - D m i D W P
Time(hr)
Time(hr)
Figure 6. Simulated B9-1 G.W.L. change using optimized transfer function.
Figure 7. Simulated B9-3 G.W.L. change using optimized transfer function 0.14 I
4 CONCLUSION
-
B 1 Sliding Block
0.12
Time series analysis on monitored data in Li-Shan landslide area were performed to find transfer function between rainfall and groundwater level change. The result showed it is applicable to predict groundwater level change for each slope based on accumulated rainfall intensity. Then, it can be applied to predict dynamic change of slope stability and forms the basis of slopeland management in landslide area.
-B4 Sliding Block
0.1 0 h
E0.08
4 3 0.06 d
0.~~1fi.:, 0.02
I - -
-
-. _- _ , , . ,c - -, - ,- _, _,_ ,_ _
-.-.-.__ ---.---.____
0.00
0 1 2 3 4 5 6 7 8 9 10111213141516171819202122232425
Time (hr)
Figure 8. Unit hydrograph of G.W.L. change for B 1, B4, and B9 sliding blocks.
Table 2. Data used for analysis for each sliding block Maximum Accumulated Initial Station Date Rainfall G.W.E. Fluctuation G'W!'* (mm> (m> (m)
Maximum G.W.E. in Analyzed Period (m>
B1
1996.7.30 1997.6.10 1997.8.17 1997.8.28
356.7 180.3 132.9 247.3
1891.03 1886.76 1881.70 1880.59
14.71 8.83 5.76 16.82
1905.74 1895.59 1887.46 1897.41
B4
1997.3.21 1997.6.10 1997.8.28 1998.2.23
165.8 179.3 247.3 171.3
1892.51 1895.62 1895.88 1898.02
2.30 2.17 2.13 1.05
1894.81 1897.79 1898.01 1899.07
B5
1997.3.21 1997.6.10 1997.6.14 1997.8.28
173.3 180.3 155.9 248.3
1959.39 1959.83 1960.21 1959.58
0.28 0.58 0.22 0.66
1956.67 1960.41 1960.43 1960.24
B9
1996.5.20 1997.6.10 1997.6.14 1997.8.28
110.5 179.8 156.4 247.3
1897.05 1897.70 1898.35 1897.31
1.06 1.90 1.13 2.88
1898.11 1899.60 1899.48 1900.19
1127
Table 3. Optimized transfer functions for analyzed data for each sliding block. Elevation of Maximum min Station’s G.W.E. during Sliding Optimized Transfer Function analized Block Ground Surface(m) Period(m)
0.1231+0.W5B1O -0.W731B12 4-3 1-0.91WB
1923.92
1905.74 1878.66
$=
1937’70
1901.21 1890.69
q=
B5
1967.05
1960.43 1945.92
& = 1-0.98054B 4 - 1 1
B9
1924.70
1903.74 1895.50
B1
B4
0.03755-0.02934-l? -0.0075d
1-0.99533B
0.000532
Q01634+Qrn$ 1-QM861B Xt-1
REFERENCES Box, and Jenkins 1994.Time Series Analysis: Forcasting and Control, prentice-hall Inc., Gupta, Yash P. and Somers, Toni M. 1989. Availability of CNC Machines: MultipleInput Transfer-Function Modeling, Transaction on Reliability, 38(3):285296. Maidment, David R. and Miaou, Shaw-Pin, Apr. 1985. Transfer Function Models of Daily Urban Water Use, Water Resource Research, 2 1(4):425-432. Trier, A. and Firinguetti, L. 1994. A Time Series Investigation of Visibility in an Urban Atmosphere-I, Atmospheric Environment, 28(5):991-996.
1128
4-1
Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
The retrogressive slide at Nipigon River, Ontario, Canada K.Tim Law Department of Civil Engineering, University of Hong Kong, People’s Republic of China (Presently: Carleton University, Ottawa, Ont., Canada)
C. E Lee Department of Civil Engineering, University of Hong Kong (Formerly: Onturio Hydro, Canada)
ABSTRACT The Nipigon River landslide was a retrogressive slide that occurred in Northern Ontario, Canada. This landslide is similar in form to those widely reported in Eastern Canada and the Scandinavia countries. However, while those landslides occurred in highly sensitive clays of marine origin, the Nipigon River landslide involved lacustrine silty soils of moderate sensitivity. The landslide was triggered by an initial slide primarily caused by a rise in the groundwater regime in a weak soil. The mechanisms of the subsequent retrogressive landslide have been investigated using existing theories. The investigation shows that application of theories largely derived from experience with marine soils is inadequate. They tend to underestimate the potential of landslide occurrence and the landslide regression distance involving soils of lacustrine origin. Static soil liquefaction is suggested to explain the underestimation.
1. INTRODUCTION A large retrogressive landslide occurred on April 23, 1990, on the bank of Nipigon River at 8 km north of the town Nipigon in northern Ontario. The failure had a maximum width of 285 m and extended about 350 m inland. It involved silty soils of lacustrine origin. The landslide debris created a temporary blockage, which raised the water level by 2 m at 8 km upstream from the failure site. Because of the remoteness of the failure, there was neither casualty nor loss of private property. Yet the slide did create some substantial environmental and economic impact. The slide debris generated clumps of trees that resulted in the relocation of water intake facility for the town. The drastic increase in the turbidity of the water threatened the habitat for the fish living in the river. Near the scarp area of the landslide, a TransCanada Pipeline was displaced laterally by 8 m, though without rupture. A large capacity fibre optic cable running adjacent to the pipeline was severed. Repair of both of these facilities incurred a substantial cost. Retrogressive landslides of this magnitude are not uncommon for slopes in sensitive marine clays in Eastern Canada and in the Scandinavian countries. A lot of experience therefore exists in dealing with this type of failure. The literature, however, contains
very few records of large retrogressive slides in lacustrine deposits. This failure therefore provides an opportunity to check whether the experience with the sensitive marine soils is applicable to lacustrine deposits. This paper describes the slide, the ground conditions and the analysis of the slide based on existing concepts on the mechanisms leading to retrogressive landslides. Conclusions are drawn on the applicability of existing theories on retrogressive landslides in silty soils of lacustrine origin. 2. DESCRIPTION OF THE SLIDE The site plan for the failure is shown in Figure 1 and a cross sectional view of the landslide is given in Figure 2. The failure site is at an outside bend on the Nipigon River that drains into Lake Nipigon to the south. The Nipigon River Valley was probably formed on a fault zone deepened by the continental icesheet. In the past, postglacial lakes inundated the area, depositing a thick layer of lacustrine soils of varved silt, fine sand and some clay. Crustal rebound brought the general area to the present elevation. The Nipigon River is in its early stage of flood plain development. The lacustrine deposit in this part of the river has been subjected to erosion and slope 1129
failure extends to about 350 m inland and has a maximum width of 285 m. Elongated ridges of soil normal to the direction of landslide movement are observed. These ridges suggest that both translation and rotation have been involved in the motion of the landslide mass. The ground surface has subsided by about 6.5 m after the landslide.
failures. The present failure is located in this lacustrine deposit and is largely a part of the river evolution process.
The TransCanada gas pipeline was exposed and displaced laterally by 8 m at the scarp of the landslide. The pipeline, suffering no rupture, was 0.92 m in diameter and was originally buried in a 3.7-m deep trench backfilled with granular materials. A large clear-cut area was found in the vicinity to the north of the landslide. Based on a witness report and the recorded time of the severing of the fibre optic cable, the total time between the start and the end of the whole slide was about three hours.
3. GROUND CONDITIONS
Figure 1 Site plan of failed area 106
I
I
I
Trans-Canada
-
/
86
I
I
Pipe i i n e
---- --- ---
---
S o f t to very soft Interbedded silt and clayey silt
1
!
!
!
I
I
0
100
200
500
400
500
Horizontal
Distance
(rn)
Figure 2 Cross section of failed area The flowbowl of the failure shows a typical bottlenecked shape that is commonly found in retrogressive slides in the Scandinavia countries. It is of note that the majority of the retrogressive landslides in sensitive marine clays of Eastern Canada do not have a bottleneck (Carson 1977). The
A site investigation was conducted by Ontario Hydro shortly after the landslide. It consisted of strength measurements using the field vane shear device and the piezocone penetrometer tests, piezometric readings, standard penetration tests (SPT) and soil sampling. The locations of some of the borings are shown in Figure I . The soil profiles and properties at the three borehole locations are shown in Figures 3 through 5. The borehole information generally indicates 4 main units in the subsoil. The top unit (Unit 1) is a 2 to 3m thick loose sandy silt. It has a STP resistance of 2 to 8. This is followed by Unit 2, which is a 3 to 4 m thick soft to firm clayey silt. It has an undrained strength of 20 to 40 kPa. It is not considered sensitive as its sensitivity ranges from 2 to 4. The effective strength parameters from consolidated undrained triaxial tests on piston samples are c’ of 2.8 kPa and 4’ of 30’. Unit 3 is a 2 to 5 m thick compact to dense sandy silt with a STP resistance ranging fiom 18 to 35. The lowermost unit (Unit 4) is a thick deposit of laminated soft clayey silt with an undrained strength of 20 to 50 E a . It has a sensitivity of about 10 to 15 and tends to liquefy upon disturbance. While there is little difference in soil strength between locations at Borehole 1 and Borehole 2, there is an appreciable strength increase towards Borehole 3, which is about 400-m inland from the riverbank. One borehole through the flowbowl near the scarp shows the absence of the sandy silt layer and suggest that the failure surface lies in the clayey silt
1130
Figure 6 Relation between stability number and retrogression distance of slides in Eastern Canada (from Mitchell and Markell (1 974)) Figure 4 Soil Profile at Borehole 2 layers (Unit 1 and Unit 2). Piezometric readings were made around the failure mass at about a month after the failure. These readings were affected by the landslide scarp and the weather condition and hence they were probably lower than those at the time of the failure. Yet the watertable was still quite high at the time of measurement. It varied from 2.5 to 5 m below the ground surface. There was a tendency for groundwater flowing from the clayey silt layers (Unit 2 and Unit 4) into the sandy silt (Unit 3).
4. MECHANISMS LANDSLIDES
FOR
RETROGRESSIVE
There are two main schools of thought on the mechanisms of retrogressive landslides involving soft sensitive marine clays in Eastern Canada and Scandinavia. The first postulates that a retrogressive landslide is formed by a series of successive rotational slips while the second assumes lateral spreading is the main mechanism.
4.1 The successive-slip mechanism Bjerrum (1955) first proposes the concept of a series of successive slips that constitute a retrogressive landslide. He considers an initial slide
1131
has to occur first. The strength along the slip surface of the initial slide is reduced to the remolded value. The remolded strength of soft sensitive clays is generally so low that the sliding mass will move significantly away from the landslide scarp area, leaving a fairly steep slope behind. If the strength in the remaining slope is low, a next rotational slip will follow. The process will be repeated until the material in the remaining slope is sufficiently strong to stop the retrogression process. While the initial slide may fail in a short or long term condition, the subsequent slips occur in an undrained manner. Other researchers such as Eden (1956), Meyerhof (1957) and Mitchell and Markell (1974) support this view. Mitchell and Markell (1974) present some interesting field observations mainly involving retrogressive landslides in soft to firm sensitive clays in Eastern Canada. Based on the observations, they propose that for a retrogressive landslide to occur, the ratio of yH/c,, has to be greater than 6, where y and cz, are the unit weight and undrained shear strength of the soil composing the slope, respectively, and H is the height of the initial slide. In addition they show that the distance of retrogression can be estimated from the empirical relationship as shown in Figure 6. The requirement of yWc,, > 6 is consistent with theoretical soil mechanics in that yH/c,, 2 5.5 is needed for a slope to fail in an undrained manner. The slightly higher required yH/c,, suggests that when a slice of slope slides down during the retrogression process, it does not necessarily completely move out of the way for the next slice to slide down. It will therefore provide sonie counterbalance load. To overcome this counter-balance load, a higher yH/c,, is needed for the retrogression process to continue. This concept is reasonable for application to clayey soils. For liquefiable soils (even under the static condition), the liquefaction process may swiftly and drastically reduce the undrained strength. For this type of soil, it is therefore not surprising that a lower value of yH/c,, is sufficient to sustain the retrogression process. Concomitantly, for the same yWc,, value, the retrogression distance will increase with an increase in the liquefaction potential of the soil involved in the slide.
4.2 The lateral spread mechanism
wedges are formed in the sliding process, with their tips pointing alternately upwards and downwards. Softening occurs at the tip of the downward pointing wedge as a result of the high stress concentration there. This leads to subsidence and lateral spreading of the wedges. The resulting movement of the soil mass is characterized mainly by translation. According to the proponents of this theory, the translational movement is more consistent with their observed morphology of the landslide debris. Based on this theory, the subsidence of the landslide mass, dh, can be estimated from:
@I/
'
where h , is the thickness of the sliding mass. In most cases this is not normally measured as part of a landslide investigation and it is often substituted by H,the height of the initial slide. Extending his own treatment of the theory, Carson (1979) proposes the following equation for calculating the distance of retrogression, R:
;[ R=
(2M - M ' )
rl
-2
+
2l
N 1
hl
where N = yfI/c,, , S = sensitivity of the soil = undisturbed strengthhemolded strength, M = Ah I h, and a = inclination of the sliding surface for the retrogression. The above approach makes use of a number of assumptions to the extent that one should consider Eq. 1 and Eq. 2 semi-analytical. In fact, different assumptions have been used in deriving the equations. Eq. 1 assumes zero strength on the sliding surface while Eq. 2 assume the remolded strength is mobilized at the sliding surface. 5 . ANALYSIS OF THE SLIDE There are three aspects in analyzing a retrogressive landslide: ( I ) the initial slide, (2) the subsequent retrogressive slide and (3) the termination of the retrogression.
5.1 The initial slide
The concept of lateral spreading as a mechanism for retrogressive slides originates from Odenstad (1946). The concept has been elaborated by Carson (1977 and 1979). This theory assumes a series of
The initial slide is essential in triggering a retrogressive slide. The initial slide for this failure was described in details by Radhakrishna et al.
1132
(1992). Briefly, the initial slide occurred under the drained condition with a high groundwater regime. A warm spell began at the site 5 days before the failure. This provided a thawing condition particularly at the clear-cut area. The snow melt water recharged the slope through the granular backfill in the trench for the TransCanda gas pipeline at the crest of the slope. This increased significantly the pore water pressure in the slope. With erosion of the toe reaching the crtical condition, the initial slide took place. 5.2 The subsequent retrogressive slide The retrogressive slide is examined based on the two schools of thought on the mechanisms for a retrogressive slide: the successive-slip mechanism and the lateral spreading mechanism. Both require an intial slide for triggering the subsequent retrogressive failure. This has been satisfied in the present case. Both concepts assume the process to take place under the undrained condition. This slide involves two layers of clayey silt that are likely to fail in an undrained manner because of the relatively short time for the whole failure to complete. 5.2.1 The successive-slip inechanism Based on the successive slip mechanism, Mitchell and Markell (1 974) suggest a retrogressive landslide can only occur when yH/c,, > 6. The combined average undrained strength of the clayey silt involved in the Nipigon River slide is about 30 kPa. 9x8/30=5. This is in The corresponding y17’/cc,,=1 contrast to what has been observed by Mitchell and Markell (1974). According to them, therefore, this retrogressive failure should not have happened. Further the distance of retrogression for this slide reaches 350 in, much larger than what has been recorded for cases in sensitive marine soils with a similar yH,/cL,value as shown in Figure 6. Therefore, what Mitchell and Markell (1974) have found applicable to the sensitive marine soils in Eastern Canada are not quite applicable to this lacustrine deposit in northen Ontario. If applied to this case, their findings will underestimate both the potential of occurrence and the retrogression distance of this slide.
retrogression distance, R , the following quantities have been used for the application: S = sensitivitiy = 4 to 15 for the soils involved; A4 = Ah/hl; where hl = the thickness of the slide mass. The value of hl was not measured and two possible values have been assumed. The first is by assuming the slope height of the initial slide as h,, ie., hl = 8 m. The second is to assume that the sliding surface reaches the lower clayey silt layer as shown in Figure 2, ie., hl = 10 m. The inclination, a, of the sliding surface is not known either. Two possible values are again taken. The first value corresponds to a horizontal sliding surface as is assumed by other researchers for similar studies. In this case, a = 0. A second value of a is taken to corresnpond to the inclination of the surface of the lower clayey silt layer (Unit 4), ie., a = 0.7’. With all these values the retrogression distance for the various cases are calculated and shown in Table 1. The results in Table 1 show that within possible ranges of the various parameters, the calculated regression distance varies from 24 to 131 m. In general, the regression distnce increases with an increase in sensitivity or an increases in the inclination of the sliding surface. Within the range of pertinent values, the calculated regression distance is signficantly lower than the measured vlaue of 350 in. This theory therefore again underestimates the subsidence and the regression distance. Table 1 Calculated retrogression distance based on lateral spreading concept for various pertinent values Thickness Of sliding Mass, hl
Sensitivity
Surface, a
(In)
8
1 1 Inclination
4
0
0.7 15
0
0.7 10
4
0
0.7 15
Retrogression Distance R (m) 26 27 98 117 24 26
0
89
0.7
131
5.2.2 The lateral spreading mechanism
5.3 The termination of retrogression
Application of the lateral spreading mechanism can be carried out by means of Eq. 1 and Eq. 2. Based on N = yI?/c,, = 5, the subsidence Ah estimated from Eq. 1 is 4.42 ni. This is significantly lower than the measured value of 6.5 m. To estimate the
The retrogression of the slide terminated at a distance aout 350 m inland. There, the strength of the clayey silt (Unit 2 and Unit 3) has an average of about 40 kPa as indicated by Borehole 3. This is appreciably higher than those at Borehoel 1 and 1133
Borehole 2. The corresponding yH/c,, value is about 3.8. This value is significantly lower than the one (6.0) suggested by Mitchell and Markell (1 974).
morphorlogy of the flowbowl does show thai both translational and rotational motions have been involved in the mass movement of the landslide.
6. DISCUSSION
As this slide is to our knowledge the first retrogressive slide in lacustrine soils ever recorded and reported in details, it is not surprising to find some descrepencies between the measurements and the estimated values based on experience with soft sensitive marine clays. Some possible reasons for the descrpencies are listed as follows. (a) The successive slip theory underestimates the potential for the occurrence of this slide. Based on this theory, the yH/cc,,value estimated using the undisturbed strength suggests that the retrogressive slide should not have occurred. The clayey silt invovled in this landslide is noticeably liquefiable. This behaviour will tend to reduce the strength of the soil such that the real yH/cc,, value is higher than the estimated value and hence the noted descrepency. This reasoning also applies to the measured low yH/cNvalue of 3.8 at which the retrogression process stopped. (b) Both theories underestimate the retrogression distance of the slide. The reduction in strength due to possible liquefaction can again provide an explanation in both cases. When static liquefaction occurs in the slope, the liquefied strength can drop below the remolded strength because of the high pore pressure generated during the undrained failure of the slope. A lower strength leads to a higher retrogression distance in the successive-slip theory as shwon in Figure 6. For the lateral spreading theory, if the liquefied strength is indeed lower than the remolded strength, the actual sensitivity during retrogression will be larger than that estimated using tlie remolded strength. Since the sensitivity used in applying Eq. 2 is based on the reinolded strength, which is lower than the actual value, an underestimate of the retrogression distance is a natural consequence. (c) The measured subsidence is larger than the estimated value based on the lateral spreading theory. This is most likely caused by the possibility that at least part of the retrogressive movement is indeed the result of successive rotational slips. This will give rise to a subsidence larger than that is given by translational motion and softening of the tip of downward pointing wedges assumed in the lateral spreading theory. Observations on the
7 CONCLUSIONS A retrogressive landslide involving lacustrine soils along the Nipigon River is studied in this paper. particular attention is paid to mechanics of the retrogressive nature of the slide and the following conclusions can be drawn: (a) The analysis and the field observations suggest that both translational and rotational motions have occurred in tlie retrogressive landslide. (b) There are problems in using past experince gained in aiialyzing retrogressive landslides in sensitive marine clays for the analysis of this retrogressive slide in lacustrine silty soils. (c) By using past experience, one underestimates the potential of occurrence and the retrogression distance of this slide. (d) The main reason for the problems is due to the high liquefaction potential of the silty soils.
8
WFERENCES
Bjerrum, L. 1955. Stability of natural slopes in quick clay, Geotechnique, 5: 101- 109. Carson, M.A. 1977. On the retrogression of landslides in sensitive muddy sediments. Can. Geot. J., 14: 582-602 Carson, M.A. 1979. Reply to discussion, On the retrogression of landslides in sensitive muddy sediments. Can. Geot. J. 16: 43 1-444. Eden, W.J. 1956.The Hawsburry landslide. Prod. Of the 10'" Canandian soil Mechanics Coizf Oftawa, 14-22. Meyerhof, G.G. 1957. The mechanism of flow slides in cohesive soils. Geotechnique, 7: 41 -49. Mitchell, R.J. and Markell, A.R. 1974. Flowsliding in sensitive soils. Can. Geot. J., 1 1: 1 1-31. Odenstad, D. 2946. The landslide at Skottorp on tlie Lidan river. Proc. R. Swed. Geot. Inst. ,4: 1-38. Radhakrishna, H.S., Bechai, M., Lau, K.D., Hale, I., Law, K.T., 1992. The Nipigoii River Landslide. Proceedings Canadian Dam Safety Conference, Whistler, B. C.
1134
Slope Stability Engineering, Yagi, Yamagami& Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Simplified model for estimating a scale of sliding debris Mitsuharu F u hd a & Seiji Suwa Geo-ResearchInstitute, Osaku, Japan
ABSTRACT : Some llgurcs of landslides have been categorized on the basis of accumulated data. These have bcen utilized as a primary step to classify types of landslides and to determine a scale of counterparts. However, these classifications of landslides aren’t taken into account as a whole by soil mechanics. Therefore , the purpose of this paper is to propose a new simplified method to predict landslides behavior. The fundamental idea of this method depends on an arch action effect. Furthermore, the application of a conclusion derived from this idea is shown to be useful for designing groups of piles as a countermeasure.
1.INTRODUCTION The countermeasures generally adopted for prcvcnting landslide activity used to be chosen on the basis of investigation results carried out in a typical section of the debris of a landslides. However, as the existing classification results of landslide points out, a whole portion of landslides debris doesn’t move homogeneously at the same speed and in the same direction. It has a tendency t o slide as if some appropriate discontinuous large blocks arc generated in a sliding debris and move interacting with each other. Although these segmentation affects on a selection of counterparts and spacing, there were few ideas presented to explain these behaviors. However, it is clear that these methodological patterns in the three dimensional directions are governed mainly by a boundary condition which acts at both ends of a debris. Therefore, in this paper, a new simplified method is shown to classify landslide patterns by applying thc concept of arch action. This concept was proved to take an important role i n a prediction of earth pressure at rest by G.P. Tshcbotarioff (19.51). The mechanism derived i n this research will be applied as one example t o ;i design of grouped piles for prevent ing landslide.
2. U T I L1 ZAT 1 0 N 0 I; ARCH ACT10N
when the debris attains an ultimate condition by fully sliding. The following three governing equations can be derived from the concept shown above.
f
I,
%I”
f RI,”
Fig.1 Postulated curve o f arch action
Where P,,, is ;i intensity o f force which acts on 1; debris in the horimntal direction, L is ;I distance, 1 is a height of convictcd line, and R,,,, R,,, ;ire corresponding reaction forces at both ends. The relationship between L and f is listed in Table 1 and Fig 2.
~1
Tshcbotarioff introduced the effect of arch action to analyx a distribution of earth pressure which acts on a retaining wall. This useful concept of structural mcchanism is extended in this research to define types of landslides. A fundamental assumption in this idea is that ;I CUI-vc oi arching axics is postulated ;is ;i quadratic line. and arch action is dcvclopcd
Table 1 Height of convicted curve
1135
100
200
+L
According to this relationship, some conditions of slope stability can be classificd as fc~llows.
F
The result of equation (3) is the approximate angle between a horizontal line and the line connecting the fulcrum to the crown of this curve and the angles are assumed to be 16.7 here. Although, these slant angles are actually considered to vary following the strength condition of the debris, on the viewpoints of a primary stage of research, equation (3) is fixed to a function. Furthermore, gravity affection and slant forces applied on the debris emerge when applying this concept to three dimensional slope bchavior. However, as secondary assumptions, three following definitions are postulated.
PI,, < P,,, : There is no occurrence of slope moving and this relation corresponds to keeping stable. However, actually, there is small move which is constrained by elastic theory. P,, = P,, : Slope starts to move as a debris. However, it is an important factor in this idea to determine P,,,. Generally, there are some types of earth pressures encountered at the same location. They are active earth pressure at the surface, passivc earth pressure at the bottom portion and earth pressure at rest in the middle depth, although their boundary conditions appear vague.
1)Arch action acts parallely to a sliding plane. 2)Intcnsity o f external force PI,, acts parallcly to a sliding plane. And uniform distribution of earth pressure can be defined approximately in depth d i rcct io n . 3)Crack and separation lines are generated along the both ends of a debris. Additionally, frictional force is fully mobilized from a surface to the bottom of a debris, and a direction o f acting is parallel to a sliding plane.
Furthermore, a redistribution of earth pressure can be found after a slope has slid. However, in this investigation, the earth pressure at rest is assumed approximately from a surface to the bottom o f a debris from the practical view point. P,,,,= K * p , * z , K = 1 - s i n d ' (7) Equation (7) includes the assumption by Jakey. Since earth pressure tends t o increase towards a lower side, the two blocks defined in Fig.3 is mod e 1c d.
Fig.2 Skelton of arch action
Equilibrium of forces which act at each end arc defined as follows. R,, = c' + (R,,,, - U ) tan 6 ' (4) Since a shear force develops to be maximum at both fulcrums under the condition designated by Fig.1, a collapse zone is devolved at both ends, if slope starts to move governed by shearing force. In other words, equation (4) is regarded as ;i main i'actor to restricting the bchavior of a slope. Thcrclorc, some cquations shown previously c;in be arranged ;is basic cquations to investigate an interaction relationship occurring in a n inner portion o f ;i slope. Combining cquations (1) and (2), equation ( 5 ) is taken.
Substituting cquations ( 1 ) and ( 5 ) for equation (4). equation (6) is obt;iincd. Equation (6) means an cquilibrium condition 01' inner pressures when ;I collapsc /one develops along the surrounding of ;i debris. Here PIlLdefined by equation (A) is earth ~ ~ r c s s u generrallp. re 2(c'-u . tan$')
PilL=
L( 1 - 1 .h(ihtan$')
(6)
Stable zone Sliding debris Fig.3
Kinematic modeling of sliding debris
In other words, the important problem of whether a surrounding boundary o f a slope will develop into a collapse or not is probably rclatcd to critical points. However, when encountering two mixed zones, a surface portion of debris resists sliding, although the lower portion will slide in a fractured Tone. If the lower portion slides then the length L of arch action won't keep constant i n depth. Howcvcr, generally, the appearance o f the pattern on the surface o f a slope is ;I fundamental way of defining ;i slope. As a result, a whole portion 01. debris is ncccssary to reach an ultimate condition. Thcrclorc the Icngt h 01 debris L has to he postulated ;IS ;I primiiry assumption. From this concept, cquation ( 8 ) is arranged.
L=
2(c'-U . tan$') K . I), . Z( 1 - 1.66htan@')
(8)
Examples o f calculation ;ire shown as follo\vs. For conl'irming the utility 01 equation (8). ;iparamctric calculation \\/:is perli)rmcd. These rcsults ;ire shown in Tiiblc I! and 3.
1136
I
...
r
25
25.9
12.9
8.6
6.5
5.2
3.4
2.6
1.7
1.3
30
175
87.4
58.3
43.7
35
23.3
17.5
11.7
8.7
(unit : m)
0 10 20 30 40 SO
1
2
3
4
5
7.5
10
15
20
175 141 108
87 71 54 37
58 47 36 25 13 2
44 35 27 18 10 2
35 28 22 15 8 1
23 19 14
18 14 11 7 4 1
12 6 7 5 3 0
9 7 5 4 2 0
74 40 6
20 3
10 5 1
1 -
Fig5 Schematic shape of debris affected by pore water pressure
(unit : m) Calculation of case 1 was carried out under the condition of zero pore pressure, wet density of 17.6KN/m3, and cohesive strength in effective stress c' of 30 kPa. From Table 2, the length of a debris is remarkably affected by the inner friction angle d 7 and the depth Z. And the length of a debris is evaluated 5 to 25m under an inner friction angle . On the contrary, it becomes lower than 25 remarkably larger if an inner friction angle is greater than 25 . On the other hand, the length(L) tends to decrease abruptly reversible to depth Z. In short, the shape of a debris resembles the figure shown in Fig.4.
segregation under an inner friction angle ranging from20" t o 3 0 " . Clay soil sliding : A conglomerate of small blocks appears under the condition of an inner friction to 25 . angle ranging from 5 Comparing flgures of landslides and the results shown in Table 2 and 3, both scales of landslides and mechanisms are similar with each other. Fig.6 shows a representative shape of the cross section of debris. Length of cross section L gradually decreases in depth if without ground water. On the other hand, under a submerged condition, the shape of the bottom becomes flat and tends to be thin.
Fig4 Shape of cross section of a debris Examples shown in Table 3 were performed under the variance of pore water pressure and the assumption that the inncr friction angle d ' is 30 and the wet density is 17.6KN/m'. Table 3 shows that pore water pressure makes an clfcct of shortening a length o f debris L. Fig.5 shows a variance 0 1 schematic shape o f a debris affected by pore water pressure. Based upon these cxamplcs. the bchavior mode 0 1 sliding ohtaincd hp the parametric study strictly corresponds with thc existing knowlcdgc accumulated by cxpcricncc (Y.YOSHIOKA , 1980). Rock sliding : A sliding block moves i n a body under the condition that an inncr friction angle ranges from 35 O t o 45 O . Wcalhcrcd rock sliding : Sonic blocks move jointly cacti undcr a n inncr friction angle ranging more than 30 O . Fr ;ic t u r c d d cbr is s 1i d i 11g : R c in ;i I- k ah 1y. ;id v;i nc c s 1o
I Width :1 v
Fig.6 Width variance of a sliding debris subjected to ground water level
Fig.7 Relation bctwccn width o f clay layer B and cross sectional length L
3. I< ELA'I'I0N I3 E TW E EN C KOS S S EC'I' 10N A L LENGTH AND LANDSLIDE CLAY Fig.6 shows the relationship between the thin clay layer at the bottom of a debris and cross sectional length L. There ;ire three types ol relationship classified between both items. Classilication ( 1 ) corresponds with the case that the cross sectional 1137
length is longer than the average width of clay layer at the contact area. Generally, slope analysis is carried out at the centcr line of a debris and the following equation is utilized.
F,=
c, . t S+ (W 'cos 0-us * t,)tan$'+P, W .sin P
(9)
Where c, and 6'are strength parameters along a sliding plane, t S is a longitudinal length of sliding line, U, is pore water pressure, W is weight of slice and ,8 is angle of bottom of sliding plane to a horizontal axies. In this equation, P, means an external force which acts at the both ends of a debris. Since side effect affects an inner mechanism of sliding through an arch action, P,can't be neglected. If P, isn't involved, strength of sliding plane should be larger. On the other hand, case (111) means that slope can keep stable, even though a clay layer exists underneath from a potential debris. Therefore, if landslides occur under case ( 111 ), side effect doesn't intluence the equilibrium of force surrounding a slice on a slip plane. In this case, P, can be neglected in a calculation based on the equation (9). Case (11) has a position between case ( 1 ) and case (111). Considering a procedure of countermeasure to resist against landslide on the view point of classification shown above, there is a case that improvement both sides of a debris does result. However in case 111, the procedure at the surrounding of a debris is possibly designed in vain. Furthermore, the measure of relative positioning of boundary conditions can be applied to the designing of groups of pilcs to prevent landslides. Generally, spacing of piles installed is designed subject to 2 or 3 times the distance of the diameter of a pile. In other words, spacing of 0.6-2.0m is required under the diameter of a pile of 0.30-0.40m. On the basis of Fig.6, spacing o f piles installed is instructed by the cross sectional length 0 1 debris at the bottom mne :tttachcd ;I slip plane i n case ( I ). For example, a clay landslide possessing an inner Iriction angle o f 10 and thickness o f 5m requires ;i spacing 01
ci1sc (
I)
l . l m according to Table 2. On the contrary, debris with a inner friction angle of 30 and thickness of 5m requires a spacing of more than 10m. In this case, the designing of piles should be instructed according to the allowable strength of the pile. However, arch action can't effectively develop on a sliding plane under case (111) so countermeasures would need to be chosen using different methods for these piles. Table 4 shows the results by comparative study.
Slope aiiiilysis Sitlc effect ;ilIkcis ;I cqu i 1ihri I I 111 of forces ac~ing 0 1 i c slicc.
CONCLUSIONS
c'ou ~ i i e r m iisiiri. c Spiiciig of pilcss iiisiitl led shou 1 d be designed 011 the hiise of'
,
I'lii~ri~lim, ~iiidysis cross scctio~i;illciigili slioultl hc p c r f i m i c d ilic slitliiig pI;iiic. cx~~isidcriiig siriicliirc of ii slicc.
iii
1138
There are many factors for estimating the equilibrium of sliding debris. Of these, simplified arch action effect, which was instructed by Tshebotarioff, was applied to classify the behavior mode of landslides. The conclusions are shown as follows. 1)The scale of sliding blocks are effectively evaluated on the base of a new simplified model. These results are proved to be similar to the actual phenomena of landslides. 2)The cross sectional shape of a landslide in depth is proved to vary depending on the strength of the sliding debris and the existence of ground water. If the level of the ground water increases in a sliding debris, the bottom shape of debris may change to be flat and thin. 3)The spacing of piles installed and it's effectiveness is instructed by the correlation between the width of sliding block and the clay layer along the sliding plane. And case study results proved to resemble past experiential data on landslides be havior.
REFERENCES Tshebotarioff, G.P. 1951. Soil Mechanics, Foundations, and Earth Structures, pp.273-276. Yoshioka, Y. 1980. Newly Systematized Civil Engineering 77 Erosion Control Works * Landslide Steep Slope Disaster, GIHODO, pp. 133-37. In Japanese.
Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Landslide prediction using nonlinear dynamics model based on state variable friction law K.T.Chau Department of Civil and Structural Engineering. Hong Kong Polytechnic University, Kocrdoon, People’s Republic of China
ABSTRACT: This paper suminarizes tlie work by Chau (1995, 1999) on inodeling landslide as a consequence of bifurcation of a steadily creeping slope when it is subjected to perturbations. In particular, the translational slip of infinite slope is modeled by a nonlinear dynamics system after incorporating nonlinear state variable fiiction laws into the slip surface (one and two state variable laws by Cliau, 1995 and Chau, 1999 respectively). According to the state variable friction laws, the shear strength (T) along the slip surface depends on the creeping velocity (V) and state variables, which evolve with the ongoing slip. Linear stability analysis is applied to study tlie stability in the neighborhood of the equilibrium solution (i.e. stable creeping slope) of the system. If the one state variable friction law is used in landslide modeling, velocity strengthening in the laboratory always implies the stability of a creeping slope containing tlie same slip surface under gravitational pull. For two state variable friction law, however. velocity strengthening observed in the laboratory does not necessarily imply stability of slope.
1. INTRODUCTION Landslides are very wide spread geologically and Crequent in occurrence. They pose serious threats to highway. railway. residential areas and other public f’acilities on mountainous terrain in inany countries. The traditional way to assess whether a slope is safe relies mainly on the use of factor of safety by applying a limit equilibrium of the soil or rock inass (Sltempton, 1964); however, such approach has failed repeatedly in predicting tlie changing stability of slopes with time. To overcome this shortcoming, Chau (1 995, 1999) proposed that landslide can be understood as a consequence of the unstable slip (or bifurcation) of a creeping slope when it is sub-jected to small external perturbations, such as tlie effect of rainfall. Although the idea of inodeling landslides as unstable creeping of slopes is not new (Davis et al., 1990, 1993; Savage & Clileborad, 1982), none of the previous studies employed the approach of nonlinear dynamics and stability analysis. In the analyses by Chau (1995, 1999) a new approach, motivated by the experimental observations by Ruina( 1983) for dry rock surfaces and by Sltenipton (1985) for fully saturated clay layers, is proposed using state variable friction laws to model
creeping along a plane of weak surface. such as a persistent rock joint, a rock joint filled with wet gouge or soil, or a soil interface. l’he nonlinear parameters involved in the formulation by Chau (1 995, 1999) may evolve due to enviroimiental impact (e.g. rainfall) such that a previously stable slope inay become unstable when a sinall perturbation is imposed (i.e. bifurcation occurs). Therefore. the bifurcation analysis of Chau (1995, 1999) provides a plausible explanation to why slope failure may occur during a particular rainfall or earthquale, which is not the largest in tlie history of the slope. The main limitation of Cliau’s (1995, 1999) analyses is that progressive process of failure (e.g. Bjei-ruin, 1967; Bishop, 1971 ; Palincr & Rice, 1973; Rice. 1973) is not incorporated. To model the failure of slopes subjected to frequent rainfall, progressive failure in fluid-infiltrated solids needs to be considered. The pore water pressure fluctuation during the shear zone propagation or shear sliding inay have significant effects on the failure process and this remains an area for active research (Rudniclti, 1987, 1991; Rudniclti & Clien, 1988; Rudniclti & Hsu, 1988; Iverson & LaHusen, 1989; Davis & Bolton. 1997). These analyses are, however, out of the scope of the present presentation.
1139
I n this paper, both tlie one state variable approach by Chau (1995) and tlie two state variable approach by Chau ( 1999) are presented. The main motivation for using the two state variable approach is from tlie experimental observations on quartzite (Ruina, 1983), dolomite (Weelts & Tullis, 1985) and granite (Tullis & Weeks, 1986). They found that two state variables are often needed for a inore coniplete description of tlie shear stress evolution with deformation subjected to sudden changes in the creeping velocity in experiments. Tlie two state variable frictional law, wliicli is to be discussed in the next section, is first proposed by Ruina ( 1983) and has primarily been applied to eaitliqualte niodeling (Gu et al., 1984: Blaiipied & Tullis. 1986; Rice, 1983; GU. & Woiig, 1994). It should be emphasized here that altliough tlie stability of a spring-slider system with a two state variable friction law has been well studied (Gu et al., 1984; Blanpied & Tullis, 1986; Gu & Wong, 1994), tlie present stability analysis is different from these works since the present nonlinear system is formulated by considering gravitational pull instead of by spriiig-slider system. That is, a nonlinear system different from the spring-slider system is considered hcre, and thus direct comparison of tlie present stability analysis and theirs cannot be made despite the same friction law is used. As remarked by Chau (1 995), many landslides around tlie world were triggered by rainfall (Pun & Li, 1995: GEO. 1995, 1996; Keefcr et al., 1987; Iverson & Major, 1987; Icasliiwaya et al., 1989; Icim et al., 1991: Pierson et al., 1991; Polloni et al., 1991; Premcliitt et al., 1994), there were also many landslides triggered by eartliqualtes (Ladd, 1935; Pearce & Watsoii, 1986; Updilte et al., 1988; Madole et al., 1996). In the present context of analysis, the effect of rainfall and earthquakes enters the present analysis only through tlie imposition of a sudden drop or jump in eitlier the friction stress or the creeping d o c i t y , as discussed by Chau (1994) and Cliau & Chan (1 994). 2. STATE VARIABLE FRICTION LAWS Presuming that tlie memory dependence of any slip surface can be represented suitably by some set of parameters, which themselves evolve with ongoing slip and depend on tlie current mechanical state of the slip surface. one may cast the following forin of fi-iction law (e.g. Rice. 1983):
r = F(Y y,Y, , , y,, ), d q j / d t = G ( J < Y . Y J > . ,qJ,), I
(1)
where i = 1,...n and V is the sliding velocity on the slip surface within tlie slope. The set of parameters Y,, Yz,..., Y,, are called state vmiubles. 3. ONE STATE VARIABLE ANALYSIS If we assuming that the mechanical state on the slip surface can be characterized by one state variable, we have (Ruina, 1983; Cliau, 1995):
where V is tlie slip velocity, Vo is the reference velocity, A and 13 are empirical constants, q1 is the threshold stress level, t is tlie time variable. and I, is a characteristic decay length scale. Tlie state variable 0 is introduced to characterize the current mechanical state of the slip surface. The parameters, A, 13, 1,. q), and 0, distinguish one slip surface from the otliers. even though they may compose of tlie same material. In general, 0 remains constant for steady state, but evolves for unsteady slip. This velocity dependent friction law is motivated by experiments by Skempton ( 1 98S), which is sltetclied in Fig. 1 together with tlie prediction by (2) and (3). Assuming translational slide of slope, Chau (1995) obtained a set of two coupled first ordcr differential equations. Then, linear stability analysis is applied to obtain the regime classification in Fig. 2 for tlie equilibrium solution
(4)
where y = pgh sin ilA and so = ~ ~ l . 4 . The important conclusion is that if tlie one state variable friction law is employed, instability occurs wlieii p > 1, which coiiicides with the condition of velocity-weakening observed in laboratories. We will show later than this conclusion is not true if two statc variable friction law is used. In Fig. 2, tlie parameters are K = pI.;,'/A, = BA, and h = WL.
4. TWO STATE VARIABLE ANALYSIS Actually inore detailed study on the experimental results reveals that two state variables friction law is always needed if a better fit of experimental data is required (e.g. Ruina, 1983). Therefore, a slope
1140
stability analysis using tlie followiiig two state variable friction law is considered by Chau (1 999): = r(,+ e,+ e! + A 111(v/ J:, (5)
__b_
p = - (Ae'+
where V is tlie slip velocity, V,, is tlie reference velocity (which is somewhat arbitrary), A . B, and B2 are eiiipirical constants, q1is the threshold stress level, I is tlie time variable, and L , and L, are tlie characteristic decay length scales for tlie state variables 8, and 8, respectively. Both of these current state variables (0, and 0,) evolve with oiigoiiig slip and are introduced to characterize tlic current iiiechanical state of tlie slip surface. 'The parameters, '4. B,. and B,, indicate tlie amount of 'immediate inercase in shear resistance after a velocity increase along the slip surface, and the subscquent drops in shear resistance due to tlie first and sccoiid state variables respectively Figure 3 shows tlic prediction of frictional stress variation with the changing in sliding velocity given by (5-7). Again, using force ecluilibl-ium for infinite slopes. Ciiau (1 999) obtained a system of three cozpled first order d i ffereiiti a1 equati o11s. The equi 1ib r i uiii so1u t i o i i for tlie system in tlie s-v-0 phase space is (Cliau, 1999)
I?'/K)
Figure 2. Regime classification for thee possible types of equilibrium point of slope equilibrium in the p-q space for tlie case of oiie state variable friction law (after Cliau, 1995)
I--- L , 4
Figure 3. Frictional stress variation with the changing in sliding velocity predicted by tlie two state variable friction law given by (5-7).
1 - D,?+ P i
Figure 1 . Frictional stress variation with the chaiigiiig in sliding velocity observed experimentally by Skemptoii (1 985) and by equations (2-3).
where s = d A , so = z,,lA, y = y,,,h sin i cos i/A,p, = BJA, 0 = 0JA. v = lii(?WJ7 and p2 = B,/A ( Y , ~and ~ h 1141
new dimension of research for slope failures. The potential use of such theories to model snoi+ avalanche seems to be a possible direction of research in the future (This possibility is proposed to me b) Prof. Talteshi Ito of the Akita National College of Technology, Japan). When two state variables fiiction law is uscd, Chau (1999) obtained tlie classification of slope instabilities shown in Fig. 5 .
\/
UNSTABLE
/\
STABLE
E>O
E40 m 15 m-40 m 0.2(within 5years) -~ ~
0.2 - >0.02
(within 5 to 50years)
.
The event is probable and may occur within inediurn to long term. The event is not expected, but could occur under extreme adverse conditions within extended long term period.
0.02 - >0.002
(within 50 to 500years) 0.002 - >0.0002
(within 500 to 5000years)
__-
The event is possible but may occur only in exceptional circumstances and adverse condition.
Field inspection sheets for Hazard Assessments for Natural Slopes have been fully developed at this stage but are not included in this paper due to space limitations. A weighting technique is used which recognises the research findings of several previous workers such as Stevenson (1977), Vecchia (1 978), Sinclair (1 992), Anbalagan (1 992), Chang (1 992), RTA (1995) and Kumar et a1 (1 996). The seven influencing factors for slope performance considered are (1) site history, (2) landslide indicators, (3) bedrock geology type and landslide material type (e.g. rock or soil) and the appropriate geotechnical properties, (4) geologic structures (e.g. adverse bedding plane, faults, and joints), (5) morphological factors such as slope angle, seepagdground moisture condition, erosion and vegetation, (6) preventative or remedial works installed and their performance, and (7) adverse human impact. Point scores are give to the above mentioned factors and the probability/hazard rating is determined according to the total score using Table 4.
4 TRIAL HAZARD ASSESSMENT ON SELECTED SITES, WOLLONGONG AREA A preliminary trial hazard assessment was carried out on the 6 selected sites in the Wollongong area to test the applicability of the new field inspection sheets for hazard assessment of natural slopes. Hazard and risk assessments were also made using the other four methods (e.g. AS/NZS, GTR, RSA & RTA). Two professionals with very different levels
>0.0002 (greater than 5000years)
of experience in the slope instability field (i.e. inexperience and extensive experience) carried out the trials. The results are tabulated in Tables 5 and 6 for the experienced professional and the inexperienced professional respectively.
5 DISCUSSION All the methods compared in Tables 6 and 7 are based on the probability (hazard) - consequence matrix approach. Assessment by the AGS approach is not included in these tables because that approach considers risk directly and not in terms of its two main components, the hazard and the consequence. The results indicate that the assessments for probability (and, therefore, hazard) can be significantly different depending on the individual assessment method used. For example, in Table 6 for site 2, the probability is assessed as moderate or likely or high or very high and this is indeed a very significant variation that justifies the need for a more comprehensive method. Turning now to the UOW comprehensive method, there are few significant differences between the assessments made by the experienced and inexperienced professionals for each of the 6 sites (compare column 6 in each of Tables 6 and 7). Similar is the case with the assessments by the RTA method for each of the six sites. On the other hand, the differences between their respective assessments are greater using the other less comprehensive methods Another interesting point is that the AGS method
1313
Table 5 . Hazard assessments results derived from 5 qualitative methods on 6 selected sites by an
~-
Fable 6. Hazard assessments results derived from 6 qualitative methods on 6 selected sites by an
results in the risk category ‘Very High’ for all the six sites. This is because these are all landslide sites, each of which is included in a Land Instability Database of the Northern Illawarra (Flentje, 1998). While each site has a history of movement, some are currently active whilst others are currently inactive. While the assessments by the RTA and UOW methods are consistent with these sites being landslides, only the UOW method captures the differences in the future hazard of landsliding between the different sites. Thus the initial use of the UOW method at these 6 sites clearly justifies the development and utilisation of comprehensive field data sheets and it is expected that the method will also be vindicated for consequence assessments that have so far not been completed
6 CONCLUSIONS Several qualitative hazard and risk assessment methods and approaches have been studied and their main features outlined in this paper. Each of these approaches has been used successfully in the context for which they were developed and within the agencies that developed them for particular
applications. Yet it should be recognised that each method has both its merits and limitations. The AGS approach, for instance, has the merit of simplicity but it requires an experienced geotechnical engineer to make an assessment and the use of considerable subjective.judgement. Hazard and consequence are not separated. Moreover, it is not possible to distinguish the level of risk between different sites, each with a previous history of instability. Methods which consider a probability (hazard) consequence matrix approach allow systematic assessment of risk and are, therefore, more valuable. Such methods may vary from those that are mainly qualitative to those that are increasingly quantitative. Consistency of assessments requires that careful thought be given to influencing factors for slope stability, the weight to be given to each factor, and the way in which data on each is recorded during site inspections for hazard and risk assessment. Effort must also be made to have clear and unambiguous definitions for the terms used. In this paper, the UOW approach has been introduced briefly and the need for the development of data sheets for hazard, consequence and risk assessments has been highlighted. The method has been used on six sites as a trial for hazard assessment and other methods have also been used for Comparison.
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Based on this limited field application, reliable, consistent and repeatable assessments are obtained. There is remarkable correspondence between the assessments of an experienced and an inexperienced professional, which is not achieved when other methods are used by the same pair. 7 REFERENCES Anbalagan, R., 1992. Terrain Evaluation and Landslide Hazard Zonation for Environmental Regeneration and Land Use Planning in Mountainous Terrain. Proceedings of the Sixth International Symposium, David H. Bell (ed.), Christchurch, New Zealand, A. A. Balkema, Rotterdam, Vol. 2., pp. 861-868. AS/NZS, 1995. Standards Association of Australia/New Zealand. Australian Standardmew Zealand Standard 4360: 1995. Risk Management. Chang, S.C., 1992. The Simprecise Mapping and Evaluation System for Engineering Geological and Landslide Hazard Zonation. Proceedings of the Sixth International Symposium, David H. Bell (ed.), Christchurch, New Zealand, A. A. Balkema, Rotterdam, Vol. 2., pp. 905-91 0. Chowdhury, R. N. & Flentje, P.N., 1998. Effective TJrban Landslide Hazard Assessment. Eighth Congress of the International Association of Engineering Geology and the Environment, A Global View from the Pacific Rim. September 21-25, Vancouver, British Columbia, Vol. 2., pp. 871-878. Cruden D. M. & Varnes D. J., 1996. Landslide Types and Process. Landslide Investigation and Mitigation, Special Report 247, Transportation Research Board, National Research Council, Turner A.K. & Schuster R.L.(eds.), National Academy Press, Washington D.C., pp. 36 - 75. Fell .R., 1994. Landslide Risk Assessment and Acceptable Risk. Canadian Geotechnical Journal, Vol. 3 1, pp. 261 - 272. Fell R. & Hartford D., 1997. Landslide risk management. Landslide Risk Assessment, Cruden & Fell (eds.), Balkema, Rotterdam, pp. 5 1 - 109. Flentje, P.N. 1998. Computer Based Landslide Hazard and Risk Assessment (Northern Illawarra Region of New South Wales, Australia). Doctor of Philosophy Thesis, University of Wollongong, New South Wales, Australia. Unpublished, 525p. Flentje, P.N. & Chowdhury, R. N., 1999. Quantitative Landslide Hazard Assessment in an Urban Area. Proceedings of the Eighth Australian New Zealand Conference on
Geomechanics. Editor: Dr. Nihal Vitharana. February 15-17, Hobart, Tasmania. Institution of Engineers, Australia, Vol. l., pp. 115-120. GTR, 1998. A report concerning Hazard and Risk associated with sites identified after the August 1998 rainstorm in Wollongong, NS W, Australia. Report commissioned by Wollongong City Council Emergency Services Department. Confidential report prepared by a Geotechnical Team including one of the authors. Unpublished report. Hutchinson, J. N., 1995. Landslide Hazard Assessment. Proceeding of the Sixth International Symposium on Landslides, Christchurch, New Zealand, Volume 3, pp. 1805 - 1841. Kumar, K., Tolia, D.S. & Kumar Satish, 1996. Landslide Hazard Evaluation in a part of Himalaya. Proceedings of the Seventh International Symposium on Landslides, Kaare Senneset (ed.), Trodheim, Norway, A. A. Balkema, Rotterdam, Vol. 1, pp. 239-244. RTA, 1995. Roads and Traffic Authority of New South Wales, Geotechnical Engineering Unit (Scientific Services Branch, Technical Services Directorate), September. Guide to a Slope Risk Rating System, unpublished report. RSA, 1997. Rail Services Australia, Railway Geotechnical Services, August. Geotechnical Risk to Rail Operations - NS W, unpublished report. Sinclair, T.J.E., 1992. SCARR: A Slope Condition and Risk Rating. Proceedings of the Sixth International Symposium, David H. Bell (ed.), Christchurch, New Zealand, A. A. Balkema, Rotterdam, Vol. 2., pp. 1057-1064. Stevenson, P.C., 1977. An Empirical Method for the Evaluation of Relative Landslide Risk. International Association Engineering Geology Bulletin, Vol. 16, pp. 69-72. Turner A.K. & McGuffy V. C., 1996. Landslide Types and Process. Landslide Investigation and Mitigation, Special Report 247, ‘Iransportation Research Board, Turner A.K. & Schuster R.L.(eds.), National Research Council, National Academy Press, Washington D.C., pp 12 1 - 128. Vecchia, O., 1978. A Simple Terrain Index for the Stability of Hillsides Scarps. In: J.D. Geddes (ed.). Large ground Movements and Structures. New YorMToronto, John Wiley, pp. 449 - 461. Walker, B.F., Dale, M., Fell, R., Jeffrey, A., Leventhal, A., McMahon, M., Mostyn, G., and Phillips, A., 1985. Geotechnical Risk Associated with Hillside Development. In Australian Geomechanics News, No. 10, December 1985, pp. 29-35.
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Slope Stability Engineering, Yagi, Yamagami L? Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Data-bases and the management of landslides R. M. Faure CETU, Centre d’ Etude des Tunnels, Lyon, Frurice
ABSTRACT :This paper deals with the description of landslides and the management of them in data bases. It presents a new computer tool for storing information on landslides and exchanging data of landslides on the net. The works of TC 1 1 are related. Cet article traite de la description des glissements de terrain et de leur classifications. L’aspect gestion des glissements de terrain et echange de donnees sur Internet est presente et les travaux du TC 1 1 sont evoques. KEY-WORDS : landslides, data-bases, classification, data exchange
1 - INTRODUCTION For more than a century land planers are looking for a better knowledge of natural risk for the settlement of new buildings, roads and other constructions. One of these natural risks are landslides. During the development of soil mechanic we found the great diversity of this phenomena and we understood why, in the middle age, in Europe. landslides are sometimes called earthquakes. This great diversity conducts to lot of classifications as each author is interested by some aspects of the phenomena. In the fist part of this paper, we summarise the works done by T C l l for the description of a landslide. It is a very necessary job, if we want to speak seriously of landslide. The international community of searchers needs a common language to exchange data on landslides. So we point out the role of a multilingual glossary for the description of a landslide and main classifications are presented to illustrate the different point of view of authors. In the second part of this paper we try to find the differxt kinds of data-bases on landslides used in the world with the analysis of data-bases described in papers presented in conferences, symposiums and revues. Some are quite rich, others quite poor in information and we try to guess the purpose of their building as the use of them may be different. The last part of this paper presents a computer tool
system for the management of landslides .and the exchange of data on landslides to improve research. As a conclusion, we propose with TC 11, an organisation for an efficient international exchange system, in respect with the personal willing of each researcher.
2 - DESCRIPTION OF LANDSLIDES The description of a landslide is a very difficult purpose as landslides may be observed as different actors.
2 - 1 - Landslides are world crust modellers. With the new possibilities of earth detection past landslides are now recognised, and it is obvious that their role in modelling landscape is important. At the Christchurch conference we can heard of the falling of the top of Mount Cook, and New Zealand become not so high. We know that in the Alp:, a slide of more than 20 000 000 m3 occurs each 25 years, that is demonstrated by the analysis of R. Schuster. (see table 1)
2 - 2 - Landslides can injluence humun relations. 2 - 2 - 1 - The lost of lives The lost of lives is one of the dramatic effect of a 1317
Table I : Major landslide disasters in Alps. (j'kom Eisbacher et al., 1984 in TRB report 247) Year 12 19 1248 1348 14 19 1486 1499 1515 1569 1569 1584 1618 1669 1806 1814 1881 1592 1963
Name Plaine d'Oisans Mont Granier Dobratsh Massif GanderbergPasseier Wildsee Zarera Kienholz Biasca Hofgastein Schwaz CorbeyrierYvorne Piuro 1 Salzburg IGoldau Antelao Massif Elm St.Gervais Vaiont Reservoir
Country France France Autriche Italie
Number of deaths thousands 1500 to 5000 heavy losses -400
Type of slope failure Failure of a landslide dam Rock avalanche Earthquake triggered rock falls Failure of a landslide dam
Suisse Suisse Suisse Autriche Autriche Suisse
300 -400 -600 147 140 328
Rock avalanche Debris flow Failure of a landslide dam Debris flow Debris flow Debris flow
Italie Autriche Suisse Italie Suisse France Italie
-1200 250 457 300 115 177 1900
Rock avalanche Rock topple and rock falls Rock avalanche Rock avalanche Rock avalanche Debris flow Rock slide in a reservoir
-
Table 2 :Socioeconomic losses in majors ,Jupun lundslides disaslers (I 938 to I98I) .from Ministry of construction in TRB report 247.
Year Juillet 1938 Juillet 1945 September 1947 July 195 1 June 1953 July 1953 August 1953 September 1958 August 1959 June 1961 September 1966 July 1967 July 1967 July 1972 August 1972 July 1974 August 1975 August 1975
Name Hyogo Hiroshima Gumma Kvoto Kumamoto Wakayama Kyoto Shizuoka Yamanashi Nagano Yamanashi Hyogo Hiroshima Kumamoto Niigata Kagawa Aomori Kochi
Number of residents dead or Number of houses destroyed miss ing or badly damaged
505 1154 27 1 114 102 460 336 1094 43 130 32 92 88 115 31 29 22 68
1318
130192 1984 1538 15141 4772 5122 19754 277 3018 81 746 289 750 1102 1139 28 536
landslide. Frank landslide in 1902, in Canada killed 70 peoples. In Norway, in sensitive clays the landslide of Verdal killed 116 peoples, Vai'ont landslide more than 1500 and this last autumn mud slides killed more than 300 peoples in South of Italy. (Del Prete et al., 1998). About politics, we know that catastrophic events enhance the grants about them, but for what duration? and the example of the fall of Mount Granier in Savoie (France) in 1248, shows that the local Prince use the phenomena for domination purpose. After 1248, the county of Savoy enlarged quickly as the Count of Savoy said that this big slide which killed more than 3000 peoples was a God judgement against men who wanted to belong in the neighbour county.
2 - 2 - 2 - Works.
2 - 3 - 1 - Vocabulary
The vocabulary use to describe landslide is in all the books about them. A special attention must be done when using this words. In the multilanguage glossary made by TC11, the definition of the words is done. (WP/WLI,1993) It is not easy to have in different countries the same meaning, because slides depend of local geology and their approach by a specialist depend of his scientific background. So, it is a good reason to read and use this kind of glossary.
2 - 3 - 2 - Shape. The shape of a landslide is sometime simple but usually it is complex and the measurements of it have to be clearly defined. TC 11 recommendations give sketches for a better knowledge. One of the most useful measurement is the volume, but to attempt its value it needs lot of data. And the range of volumes is very large as it is shown in the table 3.
The influence on human works is also important. To avoid landslide zones huge works have to be done. On the motorway leading to the tunnel of Frejus, between France and Italy, a big bridge was built to overpass a landslide zone. On the same road, a bridge built on a slow landslide is draw up each year as the landslide goes down. Footings and piles of the bridge has to be separated and a sophisticated system allows the displacement of the pile on the footing. Height piles are equipped. (Gamier et al., 1987)
The activity of a landslide attempts to include the time in the description. A landslide may be active, suspended, reactivated, inactive. An inactive landslide may be dormant, abandoned, stabilised or relict. (WP/WLI, 1993)
2 - 3 - Lundslides is cilso a local phenomena.
2 - 3 - 4 - Distribution of activity
As local phenomena a landslide can be described with its own words.
The distribution of activity indicates how the landslide evolves. It can be advancing, retrogressive. widening, enlarging, confined. diminishing, moving. (WP/WLI, 1993)
2 - 3 - 3 - Activity.
Table 3 :Some volumes of landslides 1940 1984 1970 1980 1987 1982 1800 1963 1248 191 1 1980 - 15 000 BC - 30 000 BC
Prats de Mollo la Preste Le Thoronet corny Le Friolin Val Pola Ancona Valezan Mont Toc Mont Granier usoy Mont Saint Helen Flims Alika 1319
France France France France Italy Italy France Italie France Pamir USA Switzerland Hawai (USA)
1 000 000 m3 2 000 000 m3 4 500 000 m3 10 000 000 m3 30 000 000 m3 100 000 000 m3 150 000 000 m3 280 000 000 m3 500 000 000 m3 2 200 000 000 m3 2 300 000 000 m3 12 000 000 000 m3 300 000 000 000 m3
Table 4 : The speed scale Speed Class 7 6 5 I 4 3 2
Description Extremelv raPid Very rapid Ratid I Moderate Slow Very slow
usual measure > 5 rdsecond > 3 m/minute > 1.8 m/hour I > 1.3 m/month > 1.6 rdyear > 16 mm/year
speed in mm/s > 5000. > 50. > 0.5 I > 0.005 > 0.00005 > 0.0000005
I
I Extremely slow
I
I
1
< 16 m d y e a r
I
I 1500
- Dimensions of landslides - Typology - classification depending of
data-bases are I: ‘esented in t / ?proceedings of Trondheim ISL Brasil, Rio de 593 - Yearly distribution Janeiro area - Seasonal distribution - Spatial distribution - Landslide types and material types - Volume - Damages Poland >200 - Deep rocky landslides (from a map) India, Hirnalaya 14 - Types of landslide - Causes - types of material Sri Lanka 114 - Shapes, areas, length, width
Purposes
- Assessment of landslide hazard - inventory along main roads - location of hazardous areas.
1996(Tome 3, ~ 1 8 4 6 3 1978)
I - Improvement of knowledge of geologic factors. - Communication with urban planners - distribution of landslide features and impact parameters
- date of landslides - geomorphological impact - correlation with faults. - Mitigation with local control measures.
- Mobility of landslides and vulnerability of structures
India, N WHimalaya
- Studies of slopes
- Risk zones
- Assessment of vulnerability of slopes - Control measures
- Historic landslides - Time repartition
- Land planning - Hazard evaluation
50
- Use of landslide report - WASSSiSSlDB project
- A complete description of
Spain, Los Guajares Mountains. Colombia
134
- Correlation between 18
- Definition of a methodology
Bulgaria
I60
Cruden
Alberta, Canada 156
Faure, Pairault Fernandez, I rigaray, Chacon ForeroDuefias, CaroPeiia Frangov, Tvanov, Dobrev
France
landslide factors
- Zonation of mass movement and erosion.
- A guide for land-use p’,.nners
- correlation with triggering
- Time analysis
factors and geology
1322
Iriguaray, Fernandez, Chacon Matkovic, Miklin Koukis, Tsiambaos, Sabatakis Naithan i, Prasad Polloni, Casavecchia Rybar Schoeneich, Bouzou Shunmin, H u i mi ng Chowdury F le n tj e Agostoni Laffi Mazzoccola Sciesa Presbitero Ghayoumian Shoaei Shariat Jafari Aliegra Barisone Bottino Rodrigues Pejon
Spain, Granada bassin
542
- Use of a GIS - 1 1 factors
- Determination of main
- 8 models of slopes - from geology to computer - Historic landslides
mitigation
factors
I - 6 Types of landslide
Greece
1116
- Landslide control - Remedial works
- Frequencies of lithology,
sliding material, causes. consequences, measures, etC - Thematic maps - Geological aspects - Outstanding rainfall
- Remedial measures
Czech Niger
- Landslides and geology - Huge landslides and geology
- Basic geological structure - Influence on water storing
China, Tibet
- Complexity of landslides and
capacities - Use of fractal dimension
India. Himalaya 10 areas >100
Northern Italy
I Australia,
- Rain triggering condition
I fractal dimension of soils. I328
Wollongong area Italy, Lombardia
Iran
- Use of GIS
I - Hazard mitigation - Urban planning
- Historical record - Monitoring
I700
1300
- Landslides and floods
- Prevention of risk
- Historical records - Use of GIS
- Hazard mitigation
- Different aspects of
- To avoid the loss of agricultural grounds.
- Urban planning
landslides
- Hazard map - Risk map
- Adapted landslide form.
Italy, Susa valley
I
Brazil
67
- Adapted sheet for mass movement and erosion
4 - 2 - Details about two datu-bases. We present here two data-bases, this one built by Cruden in Alberta (Cruden et al. 1990) and this one built in France at LCPC.
4 - 2 - 1 - Landslide report. WP/WLI, 1990 gave on one sheet a landslide report on which we find the following data : Date of the report and the date of the landslide. The locality and the co-ordinates. with elevation of the crown, the toe and the tip. The references of the reporter. (name, affiliation, address, phone) The geometry, length, width, depth of surface of rupture and displaced mass. The volume The damage that is, values. injuries, deaths. With this kind of report Cruden (Cruden. 1996) gives an interesting occurrence analysis of landslides
I - Inventory map
1
in Alberta. But the information is minimum, no technical features are displayed, and any scientific conclusion is impossible
4 - 2 - 2 - A French data-base. In 1989 the LCPC, in France started a data-base on landslides. (Lacube et al.. 1989). Recently. the BRGM joints it to improve this data-base that we briefly present here. Seven screens are used for the description of a landslide. The use of lists facilitate the entry of data and all is stored in an relational data-base under MS Access. a) Identification The identification and localisation are based on the use of French administration references. b) Description The description contains three parts : geometry, geology and geotecnics. A scheme with measurements is used for the geometry. stratigraphy and facies and materials are the geology, the measures of shear characteristics are used for the geotecnical aspect. 1323
c) Genesis and evolution This part is for the causes, natural or man made, and the induced phenomena. d) Damages Dead and injured peoples, cost of destroyed buildings and indirect cost are noticed. e) Studies, survey and works The kind of studies, survey and works is defined. f) Costs and decisions Legal features are written on this sheet with an estimation of the cost. g) Information The origin of the information is clearly registered. 4 - 2 - 3 - Comments.
In the second data-base an effort is done to define more completely the landslide but the aspect of classification is missing. These data-base are oriented towards administrative purposes and it would be difficult to use them as research data, although some geological descriptors are used.
4 - 3 - The available tools in 1999. The list of the previous data-base never shows exchange intention of data. But for the international community an exchange challenge of data exists for allowing a more powerful research and facilitate common works. We have now networks, like Internet, that give the possibility of quick exchange. but in a specific format. This format HTML (Hyper Text Mark up Language) is the most used language on the Net, and its versatile possibilities must be explored to assume the exchange challenge. ‘The GIS systems (Geographical Information System) are also new tools that allow the use of maps. but the use of them is very heavy. For landslides it is not necessary to have so powerful tool as we need only a position on a map. The appurtenance to a district can be efficiently managed with literal input. 4 - 4 - Two experiments. 4 - 4 - 1 - The WASSS system
The WASSS (World Area Slope Stability Server) was decided when we find the necessity of a database to get new rules or to confirm the rules used in the expert system XPENT. (Faure et al; 1992) With the time the WASSS system evolves and grows. We can now identify five releases. (Table 6) WASSS tried to simulate a world wide organisation, without a real leader, gathering different complex
events for exchanging data for a research use. WASSS4 installed two years ago partly failed, about the data-base itself but the comparisons between calculus methods is very appreciated. Some letters of readers are friendly encouragement to continue. Nevertheless it gives interesting observations that we use for building WASSS5. Thanks to my students going through the world and Canadian and Japanese co 1leagues.
4 - 4 - 2 - The Sechilienne observatory. The very huge landslide of Sechilienne, near Grenoble which can dam the valley of Romanche is an important risk that the authorities want to mitigate. (Faure and al, 1998) The actors involved in the escape and survey process are very numerous and the specialities of each one are very different. What is the common language between an historian, a town planner, a geologist or a law man? For this pre rupture case, it is all the data about a small territory that we have to store. As to manage with efficiency the coming crisis, all the data of any kind, the site with its urban zones is studied and surveyed since 1985, were stored on a CD-rom. The definition and the realisation of it, was a three years long job, with lot of meetings and give us some ideas about storage and exchange of complex data. At the opposite of WASSS4, we have there a local big problem. with a well known manager and a team of near 30 specialists and the main difficulty is in the presentation of data aqd a common language for sharing them. The aim of the observatory is to present to the population living near the menace. the best solution of mitigation. For this a CD-Rom is now pressed every 3 months to be up-dated with the last measurements, but the main part of it is the history of the site and the compilation of all the studies done. Some similar recent cases are also presented in it for convincing people that it really may happen. All the past studies are also presented as to avoid the desire of some manager to get time though new studies that are already done. The hanged 25 millions cubic meters of rock, are since ten years, advancing at the rate of 2 to 4 cm each month. It is urgent to do something. 5 - WASSS 5, FOR EXCHANGING LANDSLIDE DATA. 5 - 1 - The philosophy of the project. From the two previous experiments, we can say that the exchange of data is one of the most powerful mean to improve knowledge and make easier
1324
Tiible 6 : The five releuses ojkVASSS
I Svstem release WASSS 1 (1994) WASSS 2 (1 995) WASSS 3 (1 996) WASSS 4 (1997) WASSS 5 (1999)
I Characteristics Fist trials, we used Visual Basic and a PC Extension on Internet, the description of cases is very simple Utilisation of the description elaborated by TC 1 1, the number of fields grows up and the multilanguage aspect is truly evident Local bases and central base co-exist, but with great difficulties to maintain the commtibilitv between them. Combined use of HTML features and small bases, introduction of maps as front-end.
management. But for complex localised events, such are landslides, floods, tunnels, etc.. ., the free exchange of data is difficult for the following reasons : The owner believe that his data is not worthy enough to be displayed. As each owner is a specialist of his domain, he is never sure of the interest of his data for other specialist. In fact, that is not important for one may be very useful for another man. * The owner is frightened by the incompletness of his data. This point is very important because it is always difficult to masters a wide range of subjects. For management reasons some data are confidential and their use is specific. * The centralisation of data is heavy to manage.
Q
Q
Most of these reasons belong to human behaviour, and in the team for the Sechilienne project, two social observers were very active. They shown that the property of data is a very sensible point.
So we try to bypass these difficulties by a new approach in WASSS 5 . The philosophy or approach of this problem, can be defined by the following points :
* The data are shared in two parts for management and exchange. Firstly : Meta-data (some elementary important data and data on the data) are stored in a small local data base (with MS Access for example) for management and owner purposes. The meta data are identification, localisation, classifications and main measures on the event. Secondly : Detailed and complete datd are stored on HTML pages, linked with the
previous data base. In these HTML pages, it is possible to define with great accuracy any kind of event with sketches, maps, spread sheets, text and photographs. The first HTML page represents the meta-data and is automatically generated when feeding the small local data base. The data created by one is stored and remains on his own computer until he decides himself to give them directly to his colleagues. Each owner manages his own data base, as he wants, and can also store some partial data. The conditions of exchanging the data in WASSS 5 are summarised by: The existence of a well known site where a very small data base with only the net addresses of the different owners of local data bases on landslides is displayed for all. So every one can contact and can be contacted by anyone. A glossary of terms about the landslides can be down load from this site, as to use a common language for describing the facts in the more accurate manner. This glossary is made in several languages. Some models and instructions for feeding the data-base are also provided, and all documents for a common culture of exchange are available. As every one knows all the owners of landslide data-base, one can ask directly (a forum is projected), for a special kind of event. The answer is send by e-mail. With this organisation the exchange of data is only the exchange of HTML pages. This very simple exchange is decided by the owner of the data when he is asked by a colleague. When the addressee receive the desired pages, he add them in his own data base, the meta data for feeding its own data-
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base being on the first HTML page. He can so improve, with its personal representation, his data for a better research or management. By this manner we solve: 0 The integrity and the full customisation of all data bases. No data-bases are similar as they give an answer to different needs or specifications because organisation and laws are different in each country. There is no leader for this project. links are only between researchers. e The property of data on each HTML page the name of the builder and references are clearly written, and the exchange is triggered by the owner who attaches at his e-mail, the only pages he decides to send. e The simplicity of the exchange as only HTML pages go on the net or intranet. o and we enhance the contacts between searchers at the most effective level.
5 - 2 - The ~MrlG-RibfFsystem for WASSS 5 . The code MAG-RMF is small and can be used on any kind of usual computer. As it manages pictures of maps a wide screen give some comfort. Oriented object technology is used to produce it. MAG-RMF is the visual front-end for managing the links between four units. A set of image of maps (bitmaps) for land representation. 0 A set of contextual object representations (icons) for visual representation of any kind of view (aspect of the landslide) enable on the system. For example, with a mouse click, one can choose a location view, a risk view, a damage quantification view or an accesses view. 0 A data-base for storing and managing the meta data. To day, with MS-Access but any kind of relational data base is possible. e A set of HTML pages for an accurate description of landslides. We present hereafter the data-base through its front end. This front end (one screen) for feeding the data base is shared in four equal parts. (figure 1) The first one is called ‘references’. In it, we find the name of the landslide, its rank in the data base, the date of the landslide, the date of its first
entrance in the data-base and the update date, the link to the most representative image of the event. the link to the first HTML page, the WGS coordinates and the parameters of objects icons. The second part is called ‘classification’ and is a set of ten pre-defined lists, each list being the item of a classification corresponding to one aspect of the event. When one decides an application, it is the more difficult part of the analysis. Because the use of classifications is not easy, a lot of considerations must be taken in account depending of the potential users of the software. Happily the code is easily versatile and an incremental process to the final definition can be done. We propose a set of classifications for landslides hereafter. The third part is called ‘knowledge’ where the user can store, six measures about the landslide and six binary numbers (yes or no) which indicate if some information is developed in the HTML pages. The fourth part is a working place for searching the references and the links to image, HTML pages and other data base when one enters in the data-base a new event. When we use MAGRMF for managing the data-base, this fourth part is used for the display of the image attached to the event. The description above shows the meta data and how they are defined. As they are stored in a relational data-base (MS Access) queries can be made by a SQL language which allows any combination of parameters. The varied data about the landslide (maps, sketches, profiles, pictures and text like report) are stored in HTML format and these pages are linked with the meta-data. It is possible to reach any information in these pages with a browser (like Altavista) dedicated to the computer in which are the data.
5 - 3 - The geriernl use of MAGRMF. 5-3
- 1 - Maps front-end.
When one enters MAG-RMF application, he can feed the data-base or manage it. As all stored landslides are localised, a map front end is use for a quick view of the positions of them. Displayed maps are only bitmaps, so the user can works with its own usual (home made) maps. Zooming is going from a father map to a son map, all maps being stored in a tree with a maximum of nine maps at each level.(see figure 3). All the maps are referenced in WGS co-ordinates.
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Figwe I : The display of the meta-data 5 - 3 - 2 - Icons representation. Each landslide is marked by three icons. The meaning of each of icons must be decided before building the data-base. These three icons are chosen among different icons, that are small bitmaps. For my own data-base the three types of icon I use are: for the first one a small draw, representing either a fall, a slide, a spread, a topple, a flow. for the second I use five orange circles, with growing radius corresponding to the importance (volume) of the slide. The circle is red if there is loss of lives. for the third icon I use coloured squares corresponding to the risk. Green when the landslide is stabilised or without risk (inactive), orange when the landslide is evolving (active) and red when there is a heavy risk caused by the landslide. For managenent, one can chose the display on the maps through the symbol of object icons, a click of mouse is the user choice of the rank of icons. And, if for example, icons of rank two represent the damage made by the events, a damage map is automatically built. We obtain so, different views of the data-base. (figure 3)
5 - 3 - 3 - Links between different types of in fom a t i o n. A click on the icon leads to a choice of display for meta data, image, or HTML pages. After this choice all the corresponding information is displayed. As the meta data are stored in a relational data base, queries in SQL manner can be done. On the maps only the selected icons are displayed. And one can also use a browser for searching a chain of characters in the HTML pages.
5 - 4 - The ten classifications used for landslides in WASSS 5 . The MAG-RMF code includes ten classifications. For WASSS 5 we have chosen the following ones and a list of item for each classification allows a quick input. 1 ) Geomorphological. Block fall, rock fall, rock topple, soil topple, rock slide, soil slide. debris avalanche, rock spread, soil spread, flow, subsidence.
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Figure 2 :Displaying the son maps of a map (Zoom)
Figure 3 : Three views ojthe same group ojlundslides
2) Geotechnical, Peak rupture = first slide, residual rupture = reactivated slide, creep, liquefaction. unknown. 3) Risk, Pre-failure with low risk. Pre-failure with high risk, Post failure, Post failure with possible reactivation. unknown. 4) Activity of the landslide. Active, suspended, reactivated, inactive dormant, inactive. 5 ) Distribution of the activity Confined, advancing. retrogressive, widening, enlarging, diminishing, moving. 6) Main vulnerability. People, goods, main road, road, river, railways, other network. agricultural lands, small value land, other. multiple menace, unknown. 7) Main cause. Soil weathering, rainfall, erosion at toe, earthquake or volcanism, deforestation, man work, unknown. 8) Water causes. Water table raise, irrigation, artesian, surface infiltration, none, unknown. 9) Shape of the failure surface. Subsurface planar, deep planar, subsurface circular,
deep circular, non circular, multiple, composite. 10) Manager. Government. state or region, city, experimental
5 - 5 - Main measurements as meta-data. Some data are declared as meta-data for their importance and meaning. For example the length of a landslide is a useful information and the existence of a survey give quickly the evidence of the dangerousness of the landslide. In this application of MAG-RMF, the six retained measures are the length, the width, the depth, the volume, the average speed and the maximum speed. The information on existence of detailed data are about survey, remedial works, geological or hydrological investigation, geotechnical investigation, cost elements and back analysis. If. for these last item, the answer is yes the user can find an HTML page giving details of the item. 6 - PERSPECTIVES.
In a next f h r e . we want to enhance MAG-RMF on four ways.
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0 The set of icons must be enlarged for giving a great choice of representation, although each user can add its own icons. The number of icons for the display of one event will be increase to five. So, the display of information will be more powerful. But other kinds of icons shall be added such are, linear shape, surface shape. proportional icon to a field of the data-base, icon showing a direction. Linear shape and surface shape are drawn, reading a set of points stored with a link to the landslide, so it is possible to draw in scale, on all the maps. the limits that the user wants to display, With this, MAG-RMF will allow an accurate representation of all the landslide on the map. When input. these contours can be drawn with the mouse, at the greatest scale to be more accurate, and with different drawings. 0 A link between two data-bases will be set. So it will be possible to change of scale representation, as maps can be different. Information will be also of an other type. For example, about landslide, we can have a data-base only for one landslide when it is and experiment field. The main data base is for all landslides, and others one are only for some very well documented landslides, including escape scenario, monitoring features and so on. e Each element may have several parts corresponding to different classification. For them we shall enlarge the data-base adding only a counter on the screen for entering each classification. The first use will be for a data-base about tunnels. which cross different geological soils. 0 An other release will allow the total customisation of MAG-RMF giving to the user the possibility of changing all the labels and the writings.
Other subjects can be treated with MAG-RMF. As said before, we are working for a French data-base about tunnels and cut and cover, that are numerous in urban areas. With an appropriate use of icons, we are also carrying MAC-RMF as a tool for the management of versatile risk on an administrative area. Certainly lot of other domain exists. But, for the user, MAG-RMF is a small code. friendly using, that can give a strong help in managing localised events. All data being stored in HTML pages, they are easily used in other applications. It is possible, with the use of icons. and the choice of the maps, to customise the application. as to give to the decider the best representation of its domain. The basic release of MAG-RMF for developing a
local data-base is available on the net at : http://wasss.entpe.fr
7 - CONCLUSION As the MAC-RMF system, in its basic release, is free on the net, we hope that lot of local data-bases will grow. It is an help for all landslides managers. The main interest of MAG-RMF used with the WASSSS concepts, is certainly the soft management of the exchanges of data and the possibilities of customisation. The exchange triggered by the owner when he is ready to do it, is a good feature for the system and can win some shyness in the presentation of data. The list of all data-base will be maintained on the WASSS server if searchers indicate the existence of their data-base. With this tool the concepts of the WASSS 5 project can be easily set up. We hope so a rapid increase in the exchange of data about landslides. BIBLIOGRAPHY
Brnbb E. E., 1984, Innovative approaches to landslides hazard and risk m a ~ p i n g . 4 ISL, ' ~ Toronto, V O ~1, ~ ~ 3 0 7 - 3 2 4 Champetier de Ribes G.,1987, La cartographie des mouvements de terrain des ZERMOS aux PER. Bul. LPC,n" 150-151, pp 9-19 Crziden D.il%,de Lugt J., 1990, The world inventory of historic landslides.. 6Ih IAEG congress, Amsterdam, pp 1573-1 578 Cruden D.M., 1996, An inventory of landslides in Alberta, Canada.,7Ih ISL Trondheim, Vol 3, pp 1897-1 882. Del Prete Ad, Guadagno F.M, Hawkins A.B., Preliminary report on the landslides of 5 May 1998, Campania, southern Italy., Bull Eng Geol Env, 57. ppl13-129, 1998. Eisbacher G H., CIugue J.J., 1984, Destructive mass movements in high mountains. hazard and management. Geological Survey of Canada paper 84-16, Ottawa, 230p Frrure Re&-Michel., 1999, Some ideas and a tool for exchanging complex ground data. 2 1 Urban Data Management Symposium. Venice Faure R.h%,Tailhan J., Cligniez V., Gandon B., 1998, Presentation de l'observatoire de Sechilienne. Revue Internationale de geomatique., vol 8 n"3, pp 47-57, Hermes ed. Faure R.M., Pairault T., Hama M., Turcott-Rios E., 1998, The 4Ih release of WASSS. 7'h Int. Symp of Geology Engineering, Vancouver Gurnier P., Plaut E., Joziveaux D.,Le Dellioti P.,
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Tonello J., Corte J.F., 1987, Un pont a geometrie variable - reamenagement des fondations du viaduc du Charmaix (Savoie), Travaux n"619, pp3 1-37 Hutchinson J.N., 1988, Morphological and geotechnical parameters of landslides in relation to geology and hydrogeology., Proc 5'" ISL, Lausanne, vol 1, pp3-35. lAEG commission on Landslides, Suggested Nomenclature for Landslides. Bull IAEG. no 41 pp 13-16 Lacube J., Durville J.L., 1989, Un essai de fichier informatique sur les mouvements de terrain., Bull. L. LPC, no 161, pp 86-9 1. Leroueil S, Locat J., 1998, Slopes movementsGeotechnical characterisation, risk assessment and mitigation. 8'h IAEG congress, vol 2, pp933-944 Leroueil S., Rankine lecture 1999, Geotechnique ikfascarelli Didier, 1994, Ingenierie des pentes instables: approche orientee modeiisation de la connaissance. These de I'INSA de Lyon, France. Sassa K., 1985, The geotechnical classification of landslides., Proc. of 4'h Int. Conf and Field Workshop on landslides. Tokyo, Vol 1 pp 3 1-40. Vnrnes D.J., 1978. Slope movements, types and processes. TRB special report no 176, Landslides analysis and control, pp 1 1-33 Vnunat J., Leroueil S., Faure R.M., 1994, Slopes movement : a geotechnical perspective. Proc 7'h IAEG congress, Lisboa, ppl637-1646 TRB report 247, Landslides. investigation and mitigation. K. Turner and R. Schuster Editors ISBN 0-309-06 15 1-2 WP/WLI, 1990, A suggested method for reporting a landslide. Bull IAEG, n041, pp5- 12 WP/WLI, 1991, A suggested method for a landslide summary. Bull IAEG, n"43, pp 10 1- 1 10 WP/WLI, 1993, A suggested method for describing the activity of a landslide., Bull IAEG n"47, pp5357. WP/WLI, 1993, Multilingual landslide glossary, Bitech publishers, Richmond BC, Canada, 59pp.
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Slope Stability Engineering, Yagi, Yamagami & Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Seismicity in the development of the geological process in the Republic of Tajikistan Svetlana Vinnichenko Tajik State Geological Department ‘Tajikglavgeology’, NGO ‘Man and Nature’, Dushanbe, Tajikistan
ABSTRACT: It has been determined by the author that more than 1500 national economy facilities are located in dangerous areas, which are subject to active and dangerous landslides, mudslides, floods, gully erosion, earthquakes and other geological processes. Determination of the reasons lying behind these processes and establishment of regularities in their development is a necessary requirement for engineering - geological assessment of territories in question and follow - up designing of engineering protection schemes. The present work presents a new methodical approach to zoning of regions of Tajikistan according based on the conditions of development of all kinds of processes and their applicative associations. Under this approach the leading role is attributed to seismisity.
1
GENERAL CHARACTERISTIC OF SEISMOLOGICAL CONDITIONS OF TAJIKISTAN
Among the factors that influence the development of the present geological processes in Taj ikistan the most important are seismisity, precipitation and manmade disasters of which seismisity are considered to be the most interesting and significant. Tajikistan is situated in one of the most active seismic zones of the Earth - Pamir - Gindukush. Modern history of its development is characterized b y ,great tcctonic grows with the periods of seismic activity. According to the geotectonic data the highest seismisity was registered in the upper l’leistocene period (0 - 111) where the traces of strong earthquakes were registered in the zones of all great tectonic borders. The traces of the earthquakes and the scale of the seismic activity are identified in the zones of diffcrent types of seismodislocations of the disjunctive and gravitational character and are clearly seen in the modern relief. Nigh seismisity is also characteristic of the modern history of Tajikistan. There are data on a number of the disastrous earthquakes for the last century: Karatag (1 903), Sarez (1 91 I), Faizabad ( 1933, 1943, and 1947), Garm (1 941), Khait ( 1 949), and Pamir (1 976, 1988).
The epicenters of these earthquakes are along the brinks of tectonic joints (Hissar - Kokshal, Zaamin - Karavshi, Ilyak - Vahsh) and almost all great regional intrastructural breaks. Like in any mountain territory, seismisity in Tajikistan is of great interest not only as a geological process, but as a factor that influences the conditions, character, activity, variety and peculiarities of other geological processes and phenomena.
2
IMPORTANCE OF SEISMISITY IN D 1 3 E L 0 I’M I3N‘I’ 0 1: G E 0 LOG ICAL PROCESSES IN TAJIKISTAN.
Depending on the history of the geological development, the structure of the territory and the activity of the present tectonic life, seismisity can serve either as a background, as a cause or as a reason of practically all types of the present engineering geological processes. This thesis is confirmed by the following gravitational processes: landslides, falls, stonefalls and their paragenetic types. It is this group of processes that often determines the present character of the mountain regions of the Republic of Tajikistan, the present condition of slopes and stability of the territories.
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Currently about 50.000 landslides and falls are registered in Tajikistan and almost 9.000 of which are influenced by the seismic factor. That is why in a number of her works the author suggests the following classification of landslides:
2.1 Seismogenic In this case the seismic factor is the main reason that causes a landslide or a fall. Seismogenic landslides occur in all major seisniogenic tectonic zones, seams intrastructural regional breaks and the newest ruptures. The registered seismic force is 6-8 on the Richter scale. If the earthquakes of the high seismic force are the main reason of the gravitational landslide phenomena, then the latter (landslides) in their turn are at present considered to be the main estimates of' high seismisity. In connections with this American seismologist I. Allen (1 978) says that geological data especially those concerning the Present Quaternary period (0 - IY ) are considered to be the most valuable means to estimate the seismisity breaks and gravitational displacements registering large but not deep earthquakes are more widely spread than it was thought before. But they were not studied. Based on these assertions the signs of the seismisity of the landslides are the following: - Connection with the luiown earthquakes; - Simultaneity of' their formations on large areas. The epicentral areas of the earthquakes are the most affected; - Connection with the seismodisjunctive and seismotectonic dislocations; - Unusually complicated meclianism of the displacement; - The size, the area. the depth of the affected masses of rocks, the range of the displacement; - Specificity of the development; - The nature of the landslides that agrees with the structure of the slope; - Incompleteness of the displacement; All these sigh of the seismisity were found in Tajikistan through 20 known earthquakes of the 14 - 16 energetic groups. One of the strongest earthquakes in Khait (1 949) can confirm these seismic signs. The epicenter of the earthquakes was in the plane of Hissar - Kokshal tectonic joint that separates the structures of South-Tyanshan, SouthTajikistan and Painir geostructural regions in the spurs of Karategin mountain range. Depth of the earthquake center is 15 - 20 km; Magnitude is 7.6 - 8'2. Seismic force is 10 - 11.
Energetic group is 17. Covered area is 7000 sq. lun. The total number of the displacements that occurred during main tremor and the foreshocks is 1,100. The stone - detritus fall over 400 million cubic metres moved as far as 16 km. in a few minutes (photo 1). The landslides that caused the fall occurred in the rocky granite gneiss. The analysis of the territories in the zones of large breaks showed that:
Photo 1. The Khait stone-detritus fall (I 949).
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During the period of modern history of the development of the geological processes in Tajikistan the most widely spread processes are landslides; Seismogenic landslides can be considered to be the main relief forming factor ( photo 2); Seismogenic landslides accompanied with the disjunctive dislocations form one paragenetic complex of seismognic phenomena that decrease the firmness of the slopes (fig.1); It is the last circumstance that engineers' geologists are interested in, because old landslide circuses and masses of seismogenic landslides being dangerous are at the same time the most suitable areas for the development (slight slopes with the ground water outlets). (photo 2)
Unlike the seismogenic landslides they are formed during the earthquakes of the medium force
Photo 2. The present landslide activity in ancient landslide circus caused by irrigation 2.2 Re Iai iveIy se ismogenic (RSG) Here Seismisity is the cause the landslides on the structurally prepared and affected slopes. They are characteristic of the pleistoseistic regions of all known present and old registered earthquakes. The structures of the slopes, hydrogeological conditions, maninade congestions are the factors of preparedness. The registered seismic force is 4-6 on the Richter scale. (Faizabad - 1947, Aini - 1963, Sharora 1989). For this group of landslides seisniisity is only the reason even when the landslides happen during the earthquake. The essence of this classification lies in the fact that before the landsliding a slope or a mass of rocks undergoes a period of “preparedness”. It can be the structural predisposition of the slope: the fall of rocks “along the slope”, the existence of the feeble rocks, inaninade activities and erosion that undermine the foot of the slope, the forination of the horizons of the overmoistured rocks inside the mass of rocks, feeble contact of the massifs etc. The thesis of the relatively seismogenic (RSG) landslides can be formulated as follows: - Oil the one hand landslides prepared by a whole complex of various factors can occur on the slope without any earthquakes; - On the other hand the landslides cannot be caused even by the strong earthquake if the slope is “not prepared”; Relatively seismogenic (RSG) landslides differ from the usual noiiseismogenic landslides for which the mechanism of displacement, the size and the scale of its manifestation are not typical.
of the energetic group on the structurally predisposed slopes. It is interesting to stress that “incomplete” typically seismogenic landslides and seismic parceling are in most cases the reason of the relatively seismogenic displacement. The best example of the RSG landslides is the Missar earthquake (Sharora, 1989). The epicenter of the earthquake was in the Hissar valley 15 krn away fiom Dushanbe in the zone of the Ilyak structural disjunction. Earthquake of the medium force occur in the area regularly (the average frequency is once in 15 10 years). But up to 1989 the development of the landslides hadn’t been registered. In 1989 in the epicentral zone of the earthquake there were 3 large landslides (1,2 - 2,5 million cubic metres) and 2 landslides of the “incomplete displacement” (200 thousand cubic metres j. The landslides were formed on a large erosive terrace and on gently sloping erosive gully of the structurally denudative plateau covered with a layer of dry loess sediments of the polygenetic composition. The main reason of the landslides was the forming of the horizons of the technogenic overinoistured rocks inside the loess layer at a depth of 5 - 30 metres. The formation of a small horizon is connected with the areas of the irrigated agriculture where the watering norms during 25 years were 5 - 6 times higher. Seismic vibration caused such great texotrophy moves of the overmoistured horizon that it broke through 5 - metre layer of “dry” loesses and splashed out in mud-floows. Along the erosive gully the bottom of which has the incline of 15 degrees, the landslide Okkuli covered 2,8 kni in two and halfhours. The Oltkuli landslide came down n large erosive terrace during 4 minutes and buried a part of the settlement. The total number of people buried is 187. 2.3 Non-seismogenic The development of landslides occurs with a high seismic background (1-2 on the Richter scale). The main factors are climatic, hydrogeological and maninade dynamic. (Landslides in the Hyssar valley, the activity of ancient landslides in the zones of land-reclamation and house building). (Photo 3 .)
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3.1 Seismisity CISa reuson - no protection from the landslide is available. The main uciivities: 1. Resettlement of the population from the endangered zones, 2. Ban on the development of the seismically affected slopes and territories, 3. Preparedness for the liquidation of the possible damage. They are recommended for all mountain valleys.
3.2 Seismisity us the cause
Photo 3. Thc typical IandsIide$oiv in loess rocks 017 flit hills .sh~rllo~~s. The consideration of the peculiarities of the interdependence of the landslides and the earthquakes is of great practical value for Tajikistan because the landslides of various types determine special kinds of territories where people are occupied with the economic activities. The general characteristics of the landslide areas of Tajikistan are given fig. 5.
3
BASING OF THE SCHEMES OF ENGINEERING PROTECTION OF MOUNTAIN AREAS.
A great number of landslides niake it difficult and economically ineffective to protect from them. Moreover, very often it is just impossible to do it because of the rapidity, scales and the complecacy of their activities. Proceed from this, in order to ground the diagrams of engineering protection of the territories experiencing ‘’ the attack of the dangerous and disastrous landslides” it is necessary to have an individual approach based on the evaluation of the seismisity as the leading factor that causes landslides. The peculiarities of the territory protection in the mountain regions of Tajikistan where the seismisity is very high are as follows:
The nmin uciiviiies: I . To evaluate the vulnerability and the risk of the territories with the feeble slopes, 2. To observe the rules of land - utilization on the territories with unsoiled rocks, 3. To point out slopes and zones with the complicated structure, 4. To work out measures that could help to fix slopes, 5. To make a wide use of preventive measures These recommendations are useful for the high foothill plains and all mountain hollows of the Central and South - Eastern parts of Tajikistan. 3.3 Seisnzisity ( I S the htrckgrozind
The rmin activities: 1 . To make detailed engineering - geological survey of slopes and territories affected by the landslides, 2. To organize monitoring, 3. To single out the main factors that impact the development of landslides, 4. To work out and to implement measures preventing landslides, 5 . To use the territory rationally. These recornmendations are obligatory for all developed territories of the South - Western part o f Tajikistan characterized by the development of large loess layers. The measures helping to protect the territories with landslides should include: 1. Careful analysis of the analogues under similar conditions 2. Psychological preparation of the population 3. Training the population the rules and standards of nature-utilization 4. Preparation of the population for mitigation of natural disasters
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3. Special eiiEineerinS-geological map the conditions of 1and s1ides ’ d eve 1op in en t in .lilandi. l..igiire
L‘Y p 1ai 11i ng
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4
MAIN CONCLUSIONS
I'he present work illustratively shows a way in practical usage of scientific - geological research as a basis for engineering protection schemes. Pcc~iliaritiesin the development of gravitational geological processes, which have been determined for various differently structured territories of 'Tajikistan, are typical for all regions of Central Asia. After a more intensive and detailed study, suggested zoning could be applied practically to all mountainous and seismically active regions of the world.
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Slope Stability Engineering, Yagi, Yamagami& Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
Evaluating rockfall hazard from carbonate slopes in the Sele Valley, Southern Italy Mario Parise National Research Council, CERIST, Bari, Italy
ABSTRACT: Prediction of failure in rock slopes above or in the vicinity of towns and communication routes is a preliminary but fundamental step in the mitigation of damage related to rockfall processes. This paper deals with the analysis of the rockfall hazard at the eastern border of the valley of the Sele River, in the Southern Apennines of Italy: following the geologic and geomorphologic description of the area, the historic rockfall activity along the Mt. Valva - Mt. Marzano ridge is described, giving particular emphasis to the effect of the November 23'd, 1980, Irpinia earthquake (M = 6.8). Structural analysis of the carbonate walls, and the overall rockfall hazard are then discussed, focusing in detail the attention on the rock slopes above the towns of Valva and Colliano.
1 INTRODUCTION Carbonate mountains are very common in the Southern Apennines of Italy, forming most of the highest ridges and peaks in Campania and Basilicata regions. Even though less affected by slope instability when compared to areas characterized by outcroppings of mostly argillaceous sequences, carbonate massifs are not completely immune from such hazard. In particular, the steep slopes bordering the calcareous massifs are often subject to detachment of rocks by fall or toppling processes.
The location of important communication routes, as well as the presence of inhabited areas in the proximity of the borders of carbonate mountains make these processes highly dangerous to the anthropogenic environment, in terms of both vulnerability and risk. This paper deals with evaluation of the hazard related to rockfalls along the eastern border of the Sele River valley, at the boundary between Campania and Basilicata. It is the first contribution within the framework of a research project aimed at assessing the rockfall susceptibility in carbonate slopes, in the attempt to reduce this landslide hazard.
2 GEOLOGICAL AND GEOMORPHOLOGICAL SETTING
Figure 1 - Index map. Triangles mark the rain gauges in the area. The rectangle indicate the approximate location of Figure 2.
The borders of the valley of the Sele River consist of carbonate rocks forming the Picentini Mountains to the west and the Mt. Valva - Mt. Marzano Mt. Ogna ridge to the east (Fig. 1): limestones, dolomitic limestones and dolomites of TriasCretaceous are the main lithofacies. The intervening valley between the carbonate terrains is marked by the presence of shales, marls, chert limestones, sandstones, varicoloured clays, and clayey-marly-arenaceous flysch (Late Cretaceous Paleocene) ; all these successions are characterized 1337
by the presence of mainly clayey lithofacies. Alluvial deposits (Pleistocene - Holocene) are present along the course of the Sele River and its main tributaries. The towns of Valva and Colliano (and the hamlet Collianello) are located at the south-western foot slope of the Mt. Valva - Mt. Marzano ridge (Fig. 2). The ridge, NW-SE oriented, has relief energy on the order of several hundreds of meters. Steeply inclined to subvertical walls border the massif toward the valley of the Sele River; talus accumulations connect the carbonate outcrops to the flysch successions in the valley. As a result of the Apennine tectonics, which had already begun in the Middle-Late Miocene and continued throughout the Plio-Pleistocene, the rock units present in the area suffered large scale dislocations and associated intensive deformations (D'Argenio et al. 1973). However, present morphostructural setting of the upper valley of the Sele River has been shaped mainly during the last Pliocene-Quaternary events (Aprile et al. 1979).
The hydrographic network in the carbonate landscape has a rectangular pattern that testifies strong structural control. Water courses develop following the network of faults and fractures, and they are often entrenched as a consequence of the rapid Plio-Quaternary orogenic uplift, which, in the Picentini area, is on the order of several hundred metres (Capaldi et al. 1988). At least three orders of erosional surfaces hanging above the present base level have been identified on the eastern border of the valley. At present, they are not continuous; however, their remnants can be followed throughout the whole ridge extending from Mount Valva to the south-east. These morphological surfaces, and their distribution at elevations from about 800 m up to 1200 m a.s.l., again testify to the strong uplift that characterized the neotectonic evolution of this sector of the Southern Apennines of Italy (Amato et al. 1992). The areas where carbonate materials crop out are only marginally influenced by mass movements; these include mostly rockfalls and topples limited to the steep slopes bordering the calcareous massifs. These slopes are usually bare or covered with bushes and shrubs; only locally they are vegetated with trees. The massif foot slopes are marked by scree accumulations, made of small to medium pieces of rock which have become detached from the rock mass and have fallen as individual pieces. Locally, high-gradient fans built by alluvial as well as channelized and open-slope debris-flow processes (sensu Cruden & Varnes 1996) are present; the fan deposits have dip slope greater than 2 5 " , and present an indurated crust made of well cemented carbonate breccia. 3 CLIMATE AND SEISMICITY
Figure 2 - Map of the study area (location in Fig. 1). Explanation: 1 ) rockfalls (numbers refer to Table 11); 2 ) measurement station; 3) water course; 4) inhabited areas; 5 ) main roads. Contour interval is 100 m.
Climatic conditions vary widely within the study area because of significant differences in elevation, physiography, and other factors. The first part of this session attempts to provide a general indication of such matters along the western side of Mt. Valva - Mt. Marzano ridge, insofar as these can be extrapolated from available stations. Average rainfall in the surrounding of the study area ranges between 875 and more than 1600 mm per year (respectively, at Balvano and Senerchia rain gauges; see Fig. 1 for location). A general trend toward an increase in average rainfall when approaching the carbonate massifs is observed.
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Table I - List of great earthquakes felt in the study area (afier Boschi et al. 1997). I0 is epicentral intensity, l L local intensity (localities between brackets), M magnitude. dute 08/09/1694 0910411853 1611211857 071061191 0 2311 111980
epicetitml CIIWI Irpinia-Basilicata Irpinia Basilicata Irpinia-Basilicata lroinia-Basilicata
I,,
I, VlII (Valva and Colliano) 1X VI (Colliano), IV-V (Valva) XI VI (Valva) Vlll VI (Colliano) X VlII (Colliano and Valva)
. X-XI
Temperature data are not available at the western border of the Mt. Valva - Mt. Marzano ridge; however, analysis of data from the closest thermometric station (Materdomini, 570 m a. s. 1. elevation, distance from the study area about 8 km) indicate that in the fall-winter seasons, temperatures around 0" C are common. Temperature values at higher elevations, and in the proximity of the calcareous massifs, are expected to be much lower, which points to the crucial role played by the water in the discontinuities: the frequent daily variations in temperature below and above 0" C, as a matter of fact, favour the enlargement of joints and fractures in the rock mass, due to the cyclic expansion and contraction associated with the freezing and thawing of water. As regards seismicity, the upper valley of the Sele River is a very active seismogenic zone: main earthquakes felt in the study area are listed in Table I, which also provides the local intensity at Valva and Colliano for each event. All listed earthquakes are well above the magnitude threshold for the trigger of rockfalls (M = 4.0 according to Keefer 1984, and to Mc Calpin & Nelson 1996). As a matter of fact, several rockfalls were observed during the last great earthquake, on November 23rd, 1980 (Fig. 3; see session 4.1).
A4 6.8 5.9 6.9 5.8 68
On the other hand, no historic documents have been found about the occurrence of similar phenomena during the other earthquakes listed in Table I. However, since reported magnitudes are in at least two cases equal or greater to that of the 1980 earthquake, the possibility of rockfall occurrence during those earthquakes cannot be excluded.
4 ROCKFALL HAZARD EVALUATION 4.1 Historic data A list of 19 rockfalls (Table 11) that occurred on the eastern border of the Sele River valley was obtained through review of geological literature, historical documents, and newspaper clips (Parise 1995); an additional case, not included in Table I1 due to limited size of detached rock, was observed during the 1999 field surveys. Sixteen out of the nineteen cases listed were seismically triggered, and fifteen of these refer to the latest great seismic event, that is the 1980 earthquake. The limited number of pre-1980 cases makes clear the difficulty in obtaining any notice regarding rockfalls in the past. However, interview with local inhabitants attest that frequent but small-size rockfalls periodically occur every winter-spring season. The importance of the 1980 earthquake in triggering a number of rockfalls in the valley of the Sele River was also confirmed by comparing air photo data sets which were shot before and after the seismic event: source areas of rockfalls are clearly shown in the post-earthquake photos by the white colour of the fresh exposed rock that produced a contrast with the light grey tone which usually characterizes the carbonate rocks. The largest rockfall source areas are still easily recognizable after about 20 years from the event, both in the field and on air photos (Fig. 4).
Figure 3 - Isoseismal map of the November 23rd, 1980, Irpinia earthquake (after Postpischl 1985, simplified). L = Laviano, and C = Contursi for reference.
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Table I I - List of rockfalls in the study area. IL is the local intensity, Ed the epicentral distance, I. the epicentral intensity. Rockfall numbers refer to figure 2. 170
1 2 3 4 5 6 7 8 9 I0 11
12 13 14 15 16 17 18 19 (I) ("
ilcitc 08/09/1694 19/1011901 2611 111901 2311 111980 2311 111980 2311 111980 23/11/1980 23/11/19SO 2311 111980 2311 I11980 23/1 111980 2311 111980 2311 111980 23/1 111980 2 3 1 313930 2311 111980 2311 111980 2311 111980 04/06/1982
locutlol? Colliano Cliffabove Colliano Cliffabove Colliano Collianello Collianello Collianello Mount Marzano (SW slope) Moiiiit Marzano (S slope) Mount Marzano (S slope) I Piaiii - Mt. Valva I Piani - Mt. Valva Mt. Valva Mt. Valva Mt. Valva Mt. Valva Mt. Valva Mt. Valva Madonna degli Angeli Costa la Spina, Val di Raio (Colliano) ('I
seisiizzc trizger -Yes") - IL= VIII: Ed = 23 kni NO
No Yes - I, = VIII; Ed = 6.3 kin Yes - IL = VIII; E, = 6.3 kin Yes - 1,. = Vlll; E,, = 6.3 kin Yes - IL = VI11; Ed = 5.2 km Yes - I, = VIII; Ed = 5.5 kin Yes - I, = VIII; Ed = 5.7 km Yes - 1,. = VIII: Ed = 4.4 km Yes - I , = VIII; Ed = 4.4 k m = 4 1 km = 3.9 kni Yes - I , = VIII; Ed = 3.8 kin
Yes - I,
=
VIII; Ed = 4.8 No
~
refcrence Serva 1981; Esposito et al. 1998 Parise 1995 Parise 1995 Agnesi et al. 1983 Agiiesi et al. 1983 Agiiesi et al. 1983 Agiiesi et al. 1983 Budetta 1983 Budetta 1983 Agiiesi et al. 1983 Budetta 1983 Agnesi et al. 1983 Agnesi et al. 1983 Agnesi et al. 1983
Parise 1995
Irpinia and ~ a s i ~ i c earttiquakz: a~a I,, = I X . collapse sinkhole of' likely karst origin. Area 100 rn', diameter 120 111. depth 70-80 in
4.2 Sti-uctuml survey Macroscopic and mesoscopic structural analyses have been carried out in the area shown in Figure 2. Interpretation of aerial photographs allowed the identification of the main lineaments (Fig. 3, the tectonic nature of which has been successively proved in field surveys; moreover, this phase of work greatly helped in establishing both spacing and location of the measurement stations. Mesoscopic structural studies were then conducted by means of survey of main joint sets and analysis of prevailing directions of kinematic indicators, such as faults with slickensides and shear zones. Eleven measurement stations, whose location is shown in Figure 2, have been selected as statistically representative samples. Due to the massive aspect of the rock mass, bedding was only in a few cases identifiable with certainty; measured beddings are not consistent in the study area, showing the effect of faulting that produced local changes in the strata attitude. However, a general trend dipping toward the northern sectors seems to be present. State of fracturing in the rock mass is high to very high: at least two closely-spaced discontinuity systems have been detected at each station, plus additional minor systems. Main joints are usually highly dipping or subvertical; the presence of discontinuities
dipping less than 50" is greatly subordinate. As regards orientation of the main discontinuity systems with relation to local slope direction, a first remark has to be made: in each and every measurement station, a joint set consisting of subvertical fractures having about the same direction of the local slope is present. This system is clearly related to a decompression effect in the outer portion of the
Figure 4 - View from the southwest of Mount Marzano: in the foreground, the inhabited areas of Colliano and, poorly visible above the rocky spur in the center of the photo, the hamlet of Collianello. The arrow points to the scar left by one of the rockfalls triggered by the 1980 earthquake (# 7 in Fig. 2).
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Figure 6 - Measurement station # 1 location): two main joint sets predispose rockfall and toppling processes. In the Picentini Mountains, delimiting the valley to the west.
(see Fig. 2 for the rock mass to background, the of the Sele River
5 DISCUSSION
Figure 5 - Rose diagram of lineaments as obtained from photointerpretation, expressed as percentage of cumulative length.
rock mass. Most of the joints belonging to this system are usually tight, a few open as much as some millimetres. In addition to that above described, one or two other joint sets are generally present, about perpendicular to the local slope. Dipping is very high to vertical, and joints aperture from loose to open. Among identified joint systems, those oriented N 140-170 and N 90-110 are the most common; a subordinate joint set, with strike about N 50-70, is also frequent. Analysis of joint set characteristics with relation to slope orientation helped in assessing the most frequent type of failure in the rock mass along the western border of the Mt. Valva - Mt. Marzano ridge: rockfall by detachment along one of the more pervasive and diffuse discontinuity systems seems by far to be the main type of failure. At those sites where the joint set having about the same direction of the slope prevails, toppling failure could also be possible; this is in particular true when joints are loose to open. The possibility of rock detachment by wedge failure is by far the less common, due to high dipping or subverticality which characterize the great majority of the joints.
Location of utility lines, roads and houses in the proximity of rockfall-prone mountains pose the need in identifying the more likely sources of detachment, and assessing the related hazard, in order to mitigate the probable losses (Fenti et al. 1979; Keefer 1993; Franklin & Senior 1997). The town of Valva and Colliano, and the hamlet of Collianello, are placed well within the area potentially hit by arrival of fallen rocks. Following the free fall, detached rocks can easily travel downslope by bouncing, rolling, or by a combination of the two processes as well, and reach the inhabited areas and the main communication routes visible in Fig. 2. This is confirmed by the presence of scattered, individual, medium- to large-sized blocks at the foot slope of the carbonate massifs, close to the uphill outskirts of the towns. Geomorphological and structural surveys performed so far at the eastern border of the Sele River valley have pointed out to a moderate to high susceptibility of the carbonate slopes to rockfalls. Several factors, including fracturing in the rock mass, high relief energy, slope steepness, concur to predispose the slopes to such phenomena, whose trigger can be either climatic or seismic. Structural surveys are still ongoing; they consist, in addition to identification and measurement of the main joint sets, in the collection of the most important parameters 1341
needed to the complete characterization of the discontinuities (e.g. persistence, spacing, roughness, presence and type of filling, etc.). Such matter, already available for several measurement stations, was not treated in this paper for the sake of brevity. The ongoing research will proceed through implementation of some of the methods proposed in the scientific literature for evaluation of the rockfall susceptibility, and by critically comparing the obtained results; particular attention will be focused on the eventual differences in rockfall susceptibility along the two borders of the valley of the Sele River. REFERENCES Agnesi, V., A. Carrara, T. Macaluso, S. Monteleone, G. Pipitone & M. Sorriso Valvo 1983. Elementi tipologici e morfologici dei fenomeni di instabiliti dei versanti indotti dal sisma del 1980 (Alta Valle del Sele). Geol. Appl. e Idrogeol. 18 (1): 309-341. Amato, A., A. Cinque, N. Santangelo & A. Santo 1992. I1 bordo meridionale del Monte Marzano e la valle del F. Bianco: geologia e geomorfologia. Studi Geol. Camerti 1992/1: 191-200. Aprile, F., L. Brancaccio, A. Cinque, S. Di Nocera, M. Guida, G. Iaccarino, F. Ortolani, T. Pescatore, I. Sgrosso & M. Torre 1979. Dati preliminari sulla neotettonica dei fogli 174 (Ariano Irpino), 186 (S. Angelo dei Lombardi) e 188 (Eboli). CNR - Prog. Fin. Geodinamica, publ. no. 251: 149-178. Boschi, E., E. Guidoboni, G. Ferrari, G. Valensise & P. Gasperini 1997. Catalogo dei forti terremoti in Italia dal 461 a.C. a1 1990. Istituto Nazionale di Geojisica-SGA: 644 pp. with CD-ROM. Budetta, P. 1983. Geologia e frane dell’alta valle del F. Sele (Appennino Meridionale). Mem. e Note Ist. Geol. Appl., Napoli, 16: 53 pp. Capaldi, G., A. Cinque & P. Romano 1988. Ricostruzione di sequenze morfoevolutive nei Picentini Meridionali (Campania, Appennino Meridionale). Suppl. Geogr. Fis. Din. Quat. 1: 207-222. Cruden, D.M. & D.J. Varnes 1996. Landslide types and processes. In A.K. Turner & R.L. Schuster (eds), Landslides. Investigation and mitigation. Transp. Res. Board, Nut. Res. Council: 36-75, Washington, D.C.
D’Argenio, B., T. Pescatore & P. Scandone 1973. Schema geologico dell’Appennino Meridionale (Campania e Lucania). Conv. Moderne Vedute sulla geologia dell ’Appennino, Atti Acc. Naz. Lincei, Roma, 9. 187: 49-72. Esposito, E., A. Gargiulo, G. Iaccarino & S. Porfido 1998. Distribuzione dei fenomeni franosi riattivati da terremoti dell’Appennino Meridionale. Censimento delle frane del terremoto del 1980. Proc. Int. Con$ “Prevention of hydrogeological hazards: the role of scientific research ”, Alba, Italy, 1: 409-429. Fenti, V., S. Silvano & V. Spagna 1979. Methodological proposal for an engineering geomorphological map. Forecasting rockfalls in the Alps. Bull. Int. Ass. Eng. Geol. 19: 134138. Franklin, J.A. & S.A. Senior 1997. Rockfall hazards - Strategies for detection, assessment, and remediation. Proc. Int. Congress “Engineering Geology and the Environment ” , Athens, 1: 657-663. Keefer, D.K. 1984. Landslides caused by earthquakes. Geol. Soc. Am. Bull. 95 : 406-42 1. Keefer, D.K. 1993. The susceptibility of rock slopes to earthquake-induced failure. Bull. Ass. Eng. Geologists 30 (3): 353-361. Mc Calpin, J.P. & A.R. Nelson 1996. Introduction to paleoseismology . In J.P. Mc Calpin (ed .), Paleoseismology , Academic Press, San Diego: 1-32. Parise, M. 1995. Raccolta di notizie storiche relative a fenomeni franosi innescati da eventi climatici e/o sismici nelle aree di interesse del Progetto CEE. Rapp. Int. n. 34, CNR CERIST, Bari: 75 pp. Postpischl, D. (ed.) 1985. Atlas of isoseismal maps of italian earthquakes. CNR, Prog. Fin. Geodinamica, 114 (2A). Serva. L. 1981. I1 terremoto del 1694 in Irpinia e Basilicata. In CNE-ENEL, Contributo alla curatterizzazione della sismicita del territor-io italiano.
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Slope Stability Engineering, Yagi, Yamagami 8,Jiang 8 1999Balkema, Rotterdam, ISBN 90 5809 079 5
Effect of soil slope gradient on motion of rockfall S. Kawahara & T. Muro Department of Civil and Environmental Engineering, Ehime Universiv, Matsqama, Japan
ABSTRACT: The objective of the present paper is to investigate the motion constants and the energy consumption of rockfalls in consideration of the sinkage. A series of laboratory experiments was executed in the combination of cylindrical rockfall masses and slope gradients on a flat homogeneous slope made of Toyoura standard sand. As a result, the qualitative influences of the rockfall masses and the slope gradients on the motion constants including the acceleration, the ratio of the residual velocity to free fall and the slip ratio, and the energy consumption including the soil compaction energy and the kinetic energy were clarified. 1 INTRODUCTION
Roads constructed in mountainous districts like Japan are exposed to the danger of rockfalls owing to local downpour and weathering. The estimates of the impact force caused by a rockfall and the falling point are very important for the design of rockfall prevention work. Analyzing the results of several rockfall experiments carried out in different fields, the linear relationship between the motion constants of the rockfall and the slope gradients were clarified (Ushiro et al. 1997). However, the experimental conditions including the unevenness of the slopes in the respective fields could not be arranged in the same perfectly, and also the accuracy of the measurements was not always precise. The objective of the present paper is to investigate the motion constants and the energy consumption of rockfalls in consideration of the sinkage. A series of laboratory experiments was executed in the combination of cylindrical rockfall masses and slope gradients on a flat homogeneous slope made of Toyoura standard sand. As a result, the qualitative influences of the rockfall masses and the slope gradients on the motion constants and the energy consumption were clarified.
2 EXPERIMENTAL METHODS The soil used was Toyoura standard sand. The soil properties are shown in Table 1.
The apparatus shown in Figure 1 consisted of three main parts: a rockfall, a soil bin and a crane. The rockfalls shown in Figure 2 were columns made of aluminum and steel. The dimensions are shown in Table 2. The density of aluminum of 2.69 g/cm3 is almost the same as granite. The wires were attached to the sides of the column at intervals of a semicircle. The soil bin filled with the soil had an effective length of 230 cm, a width of 30 cm and an effective depth of 25 cm. Each of the soil divided into six layers was compacted using a weight having a mass of 7.5 kg and a drop height of 30 cm. The relative density of the soil D,was 73 %. The only top layer was remold after each run. The rockfalls were subjected to motion on the soil slope of the gradients 8 ranging from 15 through 35 degrees using the crane. The moving distance was measured using a displacement transducer. The moving distance per roTable 1. Soil properties. Properties Density of soil particles Coefficient of uniformity Coefficient of curvature Mean grain size Maximum dry density Minimum dry density Optimum water content Water content Initial dry density
1343
Figure 1. Apparatus for motion of rockfall.
Figure 3 Relationship between motion distance s an( time t Figure 2. Rockfall
Table 2. Dimensions of rockfalls. Material Diameter d (cm) Aluminum 7.5 10.0 Aluminum Aluminum 12.5 Steel 10.0 Steel 12.5 Width 9.0 cm
The correlation coefficients are very high ranging from 0.978 to 0.998. Therefore, it can be concluded that the motion of the rockfall is unifoiin acceleration.
Mass m (kg) 1.022 1.888 2.995 5.284 8.401
3.2 Acceleration The above acceleration includes the influence of the wire tension of the displacement transducer. When the wire tension T (= 1.667 N) acts on the rockfall, the equation of translation motion along the slope is given as follows:
tation was measured using the trace of the wires attached to the rockfalls. These experiments were repeated at least twice.
mu’ = mg-sin 0 - pmg-cos 0 - T
3 RESULTS AND DISCUSSIONS
where a’ is acceleration before revision, g is gravity acceleration (= 9.8 m/s2) and p is the coefficient of equivalent friction. Therefore, the true acceleration n can be calculated as follows.
3.1 Molion distance
(2)
Figure 3 shows the relationship between the motion distance s and time t. In general, when the motion of a material is uniform acceleration a, s - t curve is expressed as a parabola. Figure 4 shows the influence of the slope gradient tan 0 on the acceleration U. The acceleration U increases linearly with tan 0. The inclination of the
L
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Figure 4 Influence of slope gradient tan 0 on acceleration a.
Figure 5 Influence of slope gradient tan 0 on coefficient of equivalent friction p.
regression line little changes irrespective of the rockfall mass i n . The acceleration a decreases with the rockfail mass m because the motion resistance increases with the sinkage of the rockfall. 3.3 C’oefficient of eyuivalent,friction
The coefficient of equivalent friction p includes not only surface friction but also motion resistance owing to the sinkage of a rockfall. The coefficient of equivalent friction p is obtained from the equation of translation motion as follows. Figure 6 Motion of a rockfall. p = tan 0 -
g.cose
(4)
Figure 5 shows the influence of the slope gradient tan 0 on the coefficient of equivalent friction p. The coefficient of equivalent friction p also increases linearly wit11 tan 0. The coefficient of equivalent friction p increases with the rockfall mass m.
a
V ___
J2gh
(5)
where v is translation velocity and h is falling height shown in Figure 6. Furthermore, a is calculated using p as follows.
3.4 Ratio of residual velocity to.free-jal1 The ratio of residual velocity to free-fall a is defined as follows:
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4
I
a=
P -tan 0
Figure 7 Influence of slope gradient tan 0 on ratio of residual velocity to free-fall a. Figure 7 shows the influence of the slope gradient tan 0 on the ratio of the residual velocity to free-fall a. The ratio of the residual velocity to free-fall a also increases linearly with tan 8. The ratio of the residual velocity to free-fall a decreases with the rockfall mass m.
Figure 8 Influence of slope gradient tan 0 on slip ratio i.
1 E , = -mv2 2
= mas
(a:
v = &)
Secondly, the kinetic energy based on rotation is obtained using i as follows:
3.5 Slip ratio A slip ratio i in braking state is defined and calculated as follows:
(7) where Y is the radius of the rockfall column, o is the angular velocity, N is the number of the rotations and 1 is the moving distance at N rotations. Figure 8 shows the influence of the slope gradient tan 0 on the slip ratio i. The slip ratio i decreases with tan 0 and the rockfall mass m.
3.6 Kinetic energy ratio of rotation to translation
In this section, the kinetic energy ratio of rotation to translation was calculated. Firstly, the kinetic energy based on translation is obtained as follows.
= -mv ’ ( ~ + i ) * =-ma(l+i)’s 1 4 2
(9)
where I is the moment of inertia of a column. Therefore, the kinetic energy ratio of rotation to translation E i E , is calculated using only i as follows.
Figure 9 shows the influence of the slope gradient tan 8 on the kinetic energy ratio of the rotation to the translation EIE,. The kinetic energy ratio of the rotation to the translation E / E y as well as i decreases with tan 8 and the rockfall mass m.
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Figure 9 Influence of slope gradient tan 0 on kinetic energy ratio of rotation to trailslation EjE,.
Figure 10 Influence of slope gradient tan 8 on energy consumption ratio EIE,,.
3.7 Energy consumption
4 CONCLUSIONS
In this section, energy consumption was calculated. In general, potential energy is transformed into kinetic energy. The rest is consumed in soil compaction and surface friction. The surface-friction energy is negligibly smaller than the soilcompaction energy. Therefore, the soil-compaction energy E, is obtained as follows:
A series of laboratory experiments was executed in the combination of rockfall masses and slope gradients on a flat homogeneous slope made of Toyoura standard sand. The conclusions are summarized as follows. 1. The motion of the rockfall is uniform acceleration. 2. Both the acceleration of the rockfall and the ratio of the residual velocity to free fall increase linearly with the slope gradient. 3. Both the acceleration of the rockfall and the ratio of the residual velocity to free fall decrease with the rockfall mass on the soil slope because the motion resistance increases with the sinkage. 4. Both the slip ratio in braking state and the kinetic energy ratio of the rotation to the translation decrease with the slope gradient and the rockfall mass. 5 . The ratios of the soil compaction energy to the potential increases considerably with the rockfall mass because the ratios of the kinetic energies to the potential decrease considerably.
E, = El,- E,, - E, where
E,,is potential energy.
E,, = mgh = mgrsin 0 Figure 9 shows the influence of the slope gradient tan 8 on the energy consumption ratio EIE,,. The ratios of the soil-compaction energy to the potential E,/&,decreases with tan 0, especially in the steel. The reason for this is because E,IEl, little increases with tan 0, whereas EJEl, increases owing to the increase in a.The ratio of the kinetic energies to the potential EJE,, and E/E,, decrease considerably with the rockfall mass m, thus causing the ratios of the soil-compaction energy to the potential EJE,, to increase considerably.
1347
REFERENCE Ushiro, T., Yoshida, H., Yano, M., Takahashi, K. & Yagi, N.1997. A study of parameters for motion of falling rocks on slopes and jumping height. Journal of construction management and engineering. No.58 1/VI-37: 49-58 (in Japanese).
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slope stability Engineering, Yagi, Yamagami 6: Jiang 0 1999Baikema, Rotterdam, ISBN 90 5809 079 5
Study of accidents caused by rockfall in Kochi Prefecture TUshro, Y. Matsumoto & N.Akesaka Daiichi Consultants Company Limited, Kochi, Japan
N.Yagi Ehime University, Mutsuyumu, Japan
ABSTRACT: This report aims at evaluation and study the critical behavior of rockfall situated along the roads in the mountainous region. There are approximately ten traffic accidents caused by rockfall in Kochi prefecture every year. Three of them are fatal accidents resulting in death of the drivers and passengers. A field analysis is carried out by studying the geology and tracing the path of rockfall. The motion of rockfall is analyzed based on the data gathered in the field as well as from experiments. The obtained results from the field studies and experiments are compared and the effects of trees on the rockfall motion are evaluated and discussed.
1 INTRODUCTION The total area of the Kochi Prefecture, shown in Fig. 1, is 7,107km2 and its 84% is covered by mountains. The total length of roads in Kochi prefecture is 12,615km consisting of 1,074km national road, 1,063km regional road, 964km prefectural road and 9,515km local road. These roads mainly pass through the steep mountainous region where only
39% of them are improved. The Kochi prefecture, with annual precipitation of 2,60Omm, is one of the most heavily rained regions in Japan. Its average monthly precipitation during the rainy and typhoon season exceeds 300mm. Because of the climate and geography of the region, many rockfall accidents and incidents have been reported every year. In what follows some important of them will be analyzed and discussed.
Fig.1 Locations of Accident by Rockfalls in Kochi Prefecture
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source of ~~~.~ll
Horizontal distancex (m)
0
60
0
Y 2o
(m> 40
*
20
40
I
I
80 I
100
120
I
I
140 1
Velocity V(m/s)
on tree h=0.8m
H
Scars on treeh=0.8m Scars on treeh=0.7m
60
Remains of Roa
80
rs on tree h=0.6m
100
Road 120
Rest Point
Fig.2 Shima Kitagawa Rockfall
2 SHIMA ROCJSFALL (SHIMA, KITAGAWA) 2.1 Overview At 7:50 am of Mar. 2, 1996, a rockfall of approximately 10 tons hit a car running on the ToyoYasuda prefectural road (Route 439). The driver was injured and died after being rushed to the hospital. The topography of the site is a valley, where the steepness of the slope is 41G up to remains of the road at the height of 25m from the road surface and 3 6 O for the higher points, see Fig. 2. The geology of the site, which falls into the Shimantogawa group, consists of sandstone and shale. The bedrock is exposed and formed scarp face of 80m above the road. There are many boulders of 0.5-0.8m on the talus slope between the roads and the remains of the roads. Vegetation of the slope is a broadleaf tree of 45 years old. There are also 25-29 years old cedars above the scarp face. 2.2 Source of rockfall The sandstone block caused the accident had size of 2.6mX 1.3mX 1.8m (with an estimated weight of 10 tons). It was rested against a cedar tree 18m away from the road. The source of rockfall was scarp face of 82m above the road and there was evidence that the rock had loosened from it, see Fig. 3. The bedrock was
1350
Fig.3 Source of Rockfall
Fig.4 Deformation Caused by Rockfall
fractured and had open cracks on its surface. The deformation caused by the rockfall is shown in Fig. 4.
props of the fence were pushed into the ground by 13cm by the impact of rockfall. The beams of fence were deformed to W shape, which matches to the bottom shape of the boulder. This indicates that the boulder jumped over the fence. There were no slip marks left on the road. Therefore, the boulder hit the running car before the driver could push the break. Impact of the rock on the right side of car at the head-light rotated the car clock wise and left the tail of car hit to the guard fence, see Fig. 4. Figures 2 and 5 show movement of the rockfall estimated from traces and tracks. 3 FUTAMATA KITAGAWA)
Fig.5 Rockfall Accident Diagrams (Rebound 6 ) 2.3 Mechanics of rockj5all motion Many scars, splits, and shear failure were found on trees along the path. These scars were located less than 1.0 m above the root of the trees. Also 6 dents caused by rockfall were found. A guard fence had been installed along the side of the road and the
ROCKFALL
(FUTAMATA,
3.1 Overview At 8:20 am of August 11, 1988, a boulder with an estimated weight of 1.0 ton hit a pickup truck running on the Toyo-Yasuda prefectural road (Route 439). The boulder crushed and killed the driver's wife sitting next to the driver. The driver was seriously injured and hospitalized for 2 months. It was raining at the time of accident. That is, the rock was loosened due to the rain, which was considered the cause of accident. The geology of the site is Southern Shimanto belt, which consists of sandstone and shale. The slope was covered with 1-2m thick residuals. There were many 0.3m-0.5m boulders on the slope. However, boulders of equal sizes located only over 47m above the road caused the accident. The source of rockfall was 47.6m above the road. There was a scarp face supplying boulders above 20m of the source. The average steepness of the slope up to 18.lm above the road, where remains of the road existed, was 49' . The average steepness between the remain of road and the source was 41' and it was steeper for higher points. The average steepness between road and the source was 44. Vegetation of the slope was broadleaf trees for between road and remains of the road. The slope above remains of the road was a forested area consisting of cedars trunk having size of 30cm. ~
Fig.6 Flight Path of Rockfall (Futamata,Kitagawa)
3.2 Mechanics of rocvall motion The boulder that hit the truck was angular sandstone boulder having size of 1.Om X 0.7m X 0.55m and estimated weight of 1.0 ton. This boulder left scars on trees along the path. Most scars on the trees were left within l . l m above ground. 1351
Fig.7 Cross section of Futamata Rockfall Slope
Fig.8 Path of Rockfall From traces of the rock, it is estimated that the boulder initially rolled or slid on the slope, then jumped at point 11, bounded on the retaining wall, and finally hit the roof of car. There was no dent and therefore no jumping till point 11. The estimated rockfall path is shown in Fig. 8. The site slope has no significant change and its contour lines are nearly parallel. Therefore, rockfall moved straight but was deflected by trees.
4 SAITSUNO ROCKFALL (SAITSUNO, OTSUKI) 4.1 Overview At 2:20 p.m. on August 12, 1988, a boulder caused by landslide hit a car running on Route 321. The boulder broke through the windshield, hitting steering wheel, passing through side of the driver’s
head, hitting a passenger’s head on the left rear seat, and escaped from rear window. The passenger was killed instantly and the driver was seriously injured. There was a heavy rain earlier in the morning, and it loosened the slope and caused landslide and rockfall. The geology of the slope, with Southern Shimanto formation, consists of sandstone and shale. Its steepness is approximately 53’ . Although the slope was densely vegetated by cedars, the boulder passed through sparsely vegetated previous debris. There was a rockfall prevention work in the site, but there was no prevention work at the location of accident. The reason was probably due to existence of a 5m space between the road (with 3.7m width) and mountain which has given false sense of safety.
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4.2 Mechanics of rockfall motion There were evidences of exfoliation of boulder in the previous debris at 48m above the road, thus this place was considered as source of the rockfall. There were eight sandstone boulders of size 0.5m X 0.3mX0.2m and many boulders with diameters of 0.05m-0.lm in the site. It was concluded that exfoliated boulder disintegrated along the way down on the slope or on the road upon impact. There were broken and scared trees at the passage of rockfall and remains of impact of rockfall were left on the slope. We concluded that rockfall rolled from source, rebounded and rolled on the remain of road, hit retaining wall and jumped, finally hit the car as shown in Fig. 9. It is estimated that the final rebound velocity to be 10.0 m/s and impact velocity to the car to be 12.1 m/s.
5 CHARACTERISTICS OF ROCKFALL ON SLOPES
5.1 Velocity conservation coefficients of rockfall Figure 2 and Figure 5 shows the estimated velocity of rockfall. We estimated the flight path of rockfall and calculated velocity from 3 points of flight path while rockfall was in the air. We also estimated coefficient of friction on the slope to be p = O S and resistance of the slope to be C, , which was constant throughout the slope. We calculated C, under the conditions of zero initial velocity and no change in velocity when the rockfalls motion changed from a linear (slide or roll) state to a nonlinear (rebound) state. The estimated velocity of the rockfall can be given by the equation 1 as follows: 7 tan0 z - p
2
pcos0
7 tan0 s -p
)i +
v,, -
2
(g = acceleration of gravity, 8 = slope angle, t = time) The velocity conservation coefficients
a,,=v/J2gH for calculated velocity was a, = 0.5 for Shima and Futamata, and a,, = 0.75 for Saitsuno.
Fig.9 Flight Path of Rockfall (Saitsuno,Otsuki)
Figure 10 shows relationship between velocity conservation coefficient a , and steepness of slope. Data for this figure includes actual rockfall (Kitanada, Ooto) and field rockfall experimental data. All the field experiments were performed on slopes without vegetation cover for easier observation. When the slope was talus, velocity conservation coefficients of actual slopes were 25% smaller than experimental slopes. This could be the effect of trees. There was no difference in velocity conservation coefficient between actual and experimental slopes for rock slopes. This is because there was less trees on rock slopes thus energy loss by trees was smaller. 5.2 Rebound height of rockfall Figure 1 3 shows relationship between estimated rebound height from scares on trees and steepness of slope. Data for this figure includes actual rockfall (Kitanada, Ooto) and field rockfall experimental data. We used maximum rebound height in actual slopes, reduced 95% of average for confidant rebound height in experimental slopes. The rebound heights of actual slopes were 1.0m smaller on talus and 1.5m smaller on rock surface than experimental slopes. This can be explained by energy loss by trees.
6 CONCLUSIONS From analysis of actual and experimental data, the following conclusions can be drawn from this study:
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Fig. 10 Relationship between Velocity Conservation Coefficient and Steepness f f ' Max. of Rockfall Slope (Maximum)
Talus Slope Takamatsu Sonohara-A
- - Talus Slope without - - Slope with trees
trees
Saitsuno 0 0.6
0.7
0.8
0.9
1.0
1.1
1.2
1.3
1.4
Steepness tan0
Fig.11 Relationship between Jumping Height and Steepness
1. Actual rockfall slopes had trace of rockfall on trees and surface based on which the rockfall motion can be estimated. 2. Rockfall velocity on actual slope was 25% smaller than the experimental data. This is resulted by energy loss caused by trees. 3. Rebound heights of the actual slopes were 1.0m smaller on the talus slope and 1.5m smaller on the rock slope than those of the experimental slopes. This can be explained by energy loss caused by trees. 4. If there is a flat area such as remain of road on slope, it acts as a jump stand and increases rebound height. That is, it increases the danger. 7 REFERENCES Investigation Committee of Rockfall Accident on Toyo-Yasuda Prefectural Road. 2996. Report of rocwall accident investigation.
Japan road society.1983. Rockfall measure handbook. (In Japanese) Management office of Oodo dam. Ministry of Construction. 1996. Report of roc,?$all accident investigation. (In Japanese) Public Works Research Institute .1980. Report of field rockfall experiment. (In Japanese) Sasaki Y, Taniguchi E, Funami K, Tanimoto M Horiguchi M . 1981. Experiment of rockfall jumping height.24th Journal of Japan Road Society Meeting. Ushiro T, Murakami,T. 1983. Estimation rockfall jumping hight. Journal of the symposiuin about rockfall impact force and design rockshade. Vol.I .pp48-54. (In Japanese) Yoshida H., Ushiro T., Masuya H., Fujii T., 1991. Evaluation the impact force on rock shade considering slope characteristic. J. of structural mechanics. Vo1.37A. pp1603-1616 (In Japanese)
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Slope Stability Engineering, Yagi, Yamagami& Jiang 0 1999 Balkema, Rotterdam, ISBN 90 5809 079 5
The coefficient of restitution for boulders falling onto soil slopes with various values of dry density and water content K.T.Chau, J. J.Wu & R.H.C.Wong Department of Civil Structural Engineering, Hong Kong Polytechnic University,Kowloon, People’s Republic of China
C. E Lee Department of Civil and Structural Engineering, University of Hong Kong, People’s Republic of China
ABSTRACT: Rockfall has long been a serious problem to inountainous regions of the world. The design of rockfall mitigations requires estimations. on the horizontal and vertical travel distances of the probable rockfall events. The most popular approach for estimations is the use of computer progranis in rocltfall simulations, and it is important to note that the main parameter controlling the prediction is the coefficient of restitution. In view of this, we have recently done experiments simulating boulders failing onto soil slopes. We find that normal and tanegntial components of the coefficient of restitution ( R,, and R, ) increases with the dry density of the soil when the moisture content is less than the optimum water content. which leads to optimum soil compaction. When the optimum water content is exceeded, both R,, and R, remain roughly constant regardless of the values of the dry density of the soil.
1. INTRODUCTION It was estimated that 36 percent of the land area on the Earth is composed of mountains and 10 percent of tlie world’s population live on mountainous regions (Gerrard, 1990). Rockfall is one of the natural processes of the denudation of mountains, and it is also a phenomenon that causes tremendous hazard to human society. Ditches, cable nets, rocltfall shelters, and rock fences are sonie conimonly-used rockfall mitigation techniques to alleviate the hazard of rockfall (Chau, 1997). The design of these rockfall mitigations require estimations of the size of the rocltfall, the horizontal and vertical travel distances of the probable rocltfall events, the bouncing height at various positions along the slope, and the impact energy of rocltfalls. The most popular approach for these estimations relies on tlie use of computer programs in rockfall simulations (e.g. Sprang, 1995; Pfeiffer and Bowen, 1989; Huiigr and Evans, 1998a-b). No matter what computer program we adopted for the rockfall simulations and for the acquirement of the required rockfall parameters, the most important input parameter controlling the final output of the rocltfall statistics is the coefficient of restitution. It is wellknown that the coefficient of restitution is not a
material parameter (Wu, 1985; Sprang, 1995), but yet there is no coinprehensive effort being made to quantify the coefficient of restitution as a function of the impact conditions. Since the main parameter that affects significantly the computer predictions of impact energy, bouncing height and travel distance of rocltfall is the coefficient of restitution. Therefore, a reliable estimation of the coefficient of restitution is of profound importance in rockfall prediction and for designing countermeasures against rockfall. In view of this, we have recently implemented a comprehensive experimental research program to investigate the effects of impact conditions on the coefficient of restitution. It is hoped that more reliable rockfall statistics can be achieved with the more refined input to rockfall simulation computer programs. In particular, two high speed cameras of capability of capturing 332 frame per second were purchased, artificial rock fragments of various shapes and sizes (50, 60 and 70inm) were cased using plaster. Artificial slope surfaces were made of either plaster or compacted soils. A rockfall apparatus. including a rockfall platform, a slope platform, and a measuring board is manufactured. Many experiments are still ongoing (Chau et al., 1998, 1999). The present paper summarizes some of our recent
1355
before and after impacts) if the rotational energy is neglected:
experiments on the effect of the density of soil slopes on the coefficient of restitution during impacts. Coefficient of restitution can be found by field tests (e.g. Wu, 1985; Evans and Hungr, 1993), by back analysis (e.g. Kobayashi et al., 1990; Fornaro et al., 1990; Pfeiffer and Bowen, 1989; Paronuzzi, 1989; Descoeudres and Zimmermann, 1987; Azzoni et al., 1991) or by theoretical estimation (e.g. Bozzolo and PamiIli, 1986; Kobayashi et al., 1990); 11owever, to date there is no attempt to correlate the coefficient of restitution with the dry density as well as water content of soil slopes. Therefore, the main purpose of this paper is to present some of our recent laboratory observations on the cocfficient of restitution as a function of the dry density of soil slopes.
1 -
R,
mv,’
=--2
1 2
-
V, VI2
When the rotatiollal energy is included, coefficient ofrestitution can be &filled as:
R,,
1 2 1 L?7VI2+ IW12 2 2
= 3
(3
the
mv,’+ I@,’
(4)
Finally, Descoeudres and Ziniinermann ( 1987) followed a definition which is different from everyone else and is defined using the ratio between the rebounding impulse and the incoming impulse:
2. COEFFICIENT OF RESTITUTION 2.1 A brief review on coeflcient of’iwtitution Various definitions for the coefficient of restitution have been proposed, but there seems no consensus on which dehition is more appropriate for rocltfall prediction. One of the most commonly used definitions is the tangential and normal components of coefficient of restitution defined in terms of velocity:
If the mass of the boulder is the same before and after the impact (i.e. no clear fragmentation or cratering mechanism is observed). the definition is the same as those given in (2).
(1)
where Y,,, and V,,, are the magnitudes of the normal component of the rebounding and incoming velocities respectively. Similar definitions apply to the tangential coniponents denoted by subscript “t”. This definition appears to be the most popular one, which has been used by many authors (see Table 1 for details). Another popular definition for the coefficient of restitution is those adopted from impact dynamics of spheres, which is simply defined as the ratio between the magnitudes of the rebounding and incoming velocities:
Table 1. The definitions of the coefficient of restitution adopted by various authors Eqn. 1
2
3 4 5
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Authors/ References
Budetta and Santo (1994), Fomaro et al. (1990), Azzoni et al. (1991), Pfeiffer and Bowen (1989), Wu ( 1 985), H u n g and Evans ( 1 988, I993), Giani (1992), Hoek (1990), Kobayashi et al. (1990). Richards (1988) Spang and Raiitenstraucli (1988), Paronuzzi (1989, JRA (1983), Spang and Sonser (1995), llabib (1977), Richards (1988), Azzoni et al. (1995b) Cliau et al. (1999) Azzoni et al. (1995), Bozzolo and Painini (1986). Descoeudres and Zimmermann ( 1 987)
the study by Wu (1 985), the coefficient of restitution has been given irrespective to the iinpact angle. In particular, Wu (1 985) found, by dropping rocks and on either rock slopes or on wood platforms, that RI, decreases and RI increases linearly with the impact angle which is angle between the slope surface and the direction of the incoming boulder. Other parameters of rocltfall that may influence tlie value of the coefficient of restitution has not been incorporated into many of these studies. In particular, effects of the impact energy, tlie impact strength of both slope and boulders, the density of soil (in case of soil slopes), tlie shape of the boulders, the surface roughness of both boulders and slopes, and the rotational energy on the coefficient of restitution have not been studied thoroughly. Table 2 shows some typical values for the most popular definition of the coefficient of restitution ( R,, and RI ). Since the total energy loss for each impact can be estimated by (Hungr & Evans, 1988; Evaiis & I-Iungr. 1993):
where a is the angle of impact (see the definition by Wu, 1985). Thus. using this equation, R , defined in (2) can be related to R,, , RI and a as:
(7) hi addition, a number of field tests carried out in Japan by the Japan Railway Association showed that tlie ratio between the linear kinetic energy and the rotational kinetic energy is approximately fixed. In particular, we have (JRA, 1983):
E = E,,+ E,.= E,,(l+ p)
(8)
where the lower bound for p is about 0.1. Therefore, we have another relation between the definitions for the coefficient of restitution given in (3-4): (9) The Japan Road Association (JRA) defined tlie coefficient of restitution for average decrease of tlie velocity along a slope (not for a particular impact) versus the velocity of pure gravitational fall of a
boulder. This view point is quite different froin all other studies cited above. I n particular. the coefficient of restitution a is defined as (JRA. 1983): V =aJ2gH
If the boulder does not hit the slope surface, a will be 1. But when a number of impacts occurred during a boulder falling through a height of H,the velocity V should decrease with height and yield a < 1. The coefficient of restitution is approximated by (JRA. 1983)
where 8 is the slope angle and p is the average coefficient of frictional loss. Table 2. Some typical values of the coefficients of restitution (defined in (1)) given by Hoek ( 1 990)
R,,
RI
Descriptions
0.53 0.40 0.35
0.99 0.90 0.85
0.32 0.32 0.30
0.82 0.80 0.80
clean hard rock asphalt roadway bedrock outcrops with hard surface and large boulders talus cover talus cover with vegetation soft soil, some vegetation
Foriiaro et al. (1990) also compiled a very interesting plot for R,, versus R, , in which tlie pairs of R,, and R, seem to fall into an elliptical region for rock slopes, fall into a triangular strip for dctritical slopes, and fall into the lower half domain for fine to medium debris slope (see Fig. 1). Theoretical prediction of the coefficient of restitution is no easy matter, as the actual value depends on the actual contact mechanism between the boulder and the slope. Inelastic deformation (either inelastic crushing or plastic yielding) at the contact zone will affect tlie observed value for the coefficient of restitution. An ongoing effort is being made to quantify analytically the value of the coefficient of restitution; we quote two simplified theoretical predictions for the coefficient of restitution here. The first one is the prediction b j Bozzolo and Paniini (1986): 1357
Figure 1. The domain for the coefficient of restitution for various slopes in the R,, - R, space (iiiodiiied from Foriiaro et al., 1990)
where Rc,,] is the experimental value for the coefficient of restitution, and the second term in the bracket is the theoretical prediction assuming a point contact without sliding. By assuming that the result force at tlie contact is acting parallel and opposite to the resultant incoming velocity but the range ofthis reaction is restricted by the angle of maximum friction
v (3-3b)
/
n
Bj = VolejnFN,dCdq,
Therefore, the momentum and mass conservation is closed within columns (elements) in terms of nodal velocity and depth. The basal elevation fbnction B(xy) is represented in an uneven spacing coordinate grid. To achieve a higher-order basal inclination, the gradients B, and By are computed a priori by the finite difference of B ( x j ) before bilinearly interpolated for the elemental or nodal ones.
(3-3c)
where M is the consistent mass matrix, B is the nodal force vector, and the continuity constraint is released. In practice, the following two-step predictorcorrector procedure is adopted in order to avoid an implicit solution of the algebraic system: (3-4a) U* = (MO" + A t @ M - d , on+] = 20*- ~ o * ~ - d , (3-4b) in which 0' is the intermediate velocity, and M-d the inverse of lumped mass matrix. Worth special mention is that the correction step is quite crucial for unsteady problems, otherwise spurious dissipation will be introduced. Moreover, the volume of an element is unchanged during motion and hence accumulative error can be avoided. When element vertices proceed to new positions = 2"+ AL\t(B"" + U " )/ 2 , (3-5) the mean height at an element center is readily updated by (3-2). The nodal height is then redistributed from the least squares approximation QhnC'= R , (3-6a) in which Q, = J,"+'JN,N,d
