Fundamentals of Durable Reinforced Concrete (Modern Concrete Technology Series (E. & F.N. Spon).)

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Fundamentals of Durable Reinforced Concrete

Modern Concrete Technology Series editors Arnon Bentur

Sydney Mindess

National Building Research Institute Technion – Israel Institute of Technology Technion City Haifa 32 000 Israel

Office of the President University of British Colombia 6328 Memorial Road Vancouver, B.C. Canada V6T 1Z2

There have been significant advances made in concrete technology in recent years. However, the results of this recent research are still either widely scattered in the journal literature, or mentioned only briefly in the standard textbooks. The aim of this series is to examine, to some depth, each topic of interest to provide a uniform, high quality coverage which will, over a period of years, build up to form a library of reference books covering all the major topics in modern concrete technology. Primarily the books are of single or dual authorship although, for certain books, edited volumes consisting of a collection of chapters has been considered more appropriate. The material is presented at a level suitable for senior and postgraduate engineers working in concrete technology. Also available in the series 1 Fibre Reinforced Cementitious Composites A. Bentur and S. Mindess 2 Concrete in the Marine Environment P. K. Mehta 3 Concrete in Hot Environments I. Soroka 4 Durability of Concrete in Cold Climates M. Pigeon and M. Pleau 5 High Performance Concrete P-C. Aitcin 6 Steel Corrosion in Concrete A. Bentur, S. Diamond and N. Berke 7 Optimization Methods for Material Design of Cement-based Composites Edited by A. M. Brandt 8 Special Inorganic Cements I. Odler 9 Concrete Mixture Proportioning F. de Larrard 10 Sulfate Attack on Concrete J. Skalny, J. Marchand and I. Odler 11 Determination of Pore Structure Parameters K. Aligizaki 12 Fundamentals of Durable Reinforced Concrete M. G. Richardson

Fundamentals of Durable Reinforced Concrete

Mark G. Richardson

London and New York

First published 2002 by Spon Press 11 New Fetter Lane, London EC4P 4EE Simultaneously published in the USA and Canada 29 West 35th Street, New York, NY 10001

This edition published in the Taylor & Francis e-Library, 2004. Spon Press is an imprint of the Taylor & Francis Group © 2002 Mark G. Richardson All rights reserved. No part of this book may be reprinted or reproduced or utilised in any form or by any electronic, mechanical, or other means, now known or hereafter invented, including photocopying and recording, or in any information storage or retrieval system, without permission in writing from the publishers. The publisher makes no representation, express or implied, with regard to the accuracy of the information contained in this book and cannot accept any legal responsibility or liability for any errors or omissions that may be made. British Library Cataloguing in Publication Data A catalogue record for this book is available from the British Library Library of Congress Cataloging in Publication Data A catalog record for this book has been requested

ISBN 0-203-22319-5 Master e-book ISBN

ISBN 0-203-27744-9 (Adobe eReader Format) ISBN 0-419-23780-1 (Print Edition)

Dedicated to the memory of Tom McCormack (1949–2001)

Disclaimer

The publisher makes no representation, expressed or implied, with regard to the accuracy of the information contained in this book and cannot accept any legal responsibility or liability for any errors or omissions that may be made. The reader should verify the applicability of the information to particular situations and is urged to consult with appropriate professionals prior to taking any action or making any interpretation that is within the realm of a professional practice.

Contents

Preface Acknowledgements

xi xii 1

1

Framework for durability by specification Context 1 Key issues 2 Historical review 3 Specifying durable concrete: the options 11 Durability and the next generation of standards 18 Summary 18

2

Probabilistic approach to durability design Design life 20 Structural design analogy 22 Approach to design for durability 26 Future research needs 33 Application in practice 35 Summary 37

20

3

Permeability and transport processes Pore structure and the hydration process 40 Transport processes and rates 43 Measurement of permeation properties 46 Factors influencing the permeability of site concrete 49

38

4

Corrosion of reinforcement in concrete Nature of corrosion damage 51 Electrochemical process 54 Polarisation curves, the ‘Evans Diagram’ 60 Passivity 61

51

viii

Contents Corrosion mechanism in carbonated concrete 62 Corrosion mechanism in chloride-rich concrete 63 Influences on corrosion activity 65 Influence of cracking 67 Modelling the rate of corrosion 69 Monitoring corrosion activity 71 Summary 76 77

5

Carbonation Carbonation and corrosion 77 Chemistry of carbonation 78 Detection of the carbonation front 79 Primary factors influencing carbonation rate 81 Mathematical modelling of the rate of carbonation 85 Application of models to service life prediction 94 Carbonation: exposure categories in EN 206–1 97 Specification by performance 99 Summary 100

6

Chloride ingress Chloride ingress and corrosion 101 Detection and expression of chloride levels 105 Critical chloride level for corrosion 107 Primary factors influencing chloride ingress 111 Mathematical modelling of chloride ingress 114 Application of models to service life prediction 121 Chlorides: limitations and exposure categories in EN 206–1 124 Specification by performance 129 Summary 132

101

7

Alkali–silica reaction Background 133 Manifestation of the problem 135 Mechanism of expansion and reaction 139 Primary factors influencing the reaction 141 Other factors influencing ASR occurrence 145 Modelling and service life prediction 148 Specifications to minimise the risk of ASR 149 Summary 159

133

Contents

ix

8

Freeze/thaw effects Background 160 Primary factors of influence 162 Air entrainment 165 Developments in specification and design practice 166 Freeze/thaw attack: exposure categories in EN 206–1 170 Developments in testing 173 Specification by performance 176 Summary 177

160

9

Chemical attack: sulfates Introduction 179 Physico-chemical aspects 180 Factors influencing sulfate attack 184 Approaches in specification and design practice 188 Sulfate attack: exposure categories in EN 206–1 190 Developments in testing and specification by performance 192 Summary 193

179

10 Chemical attack: acid and seawater attack Introduction 194 Physico-chemical aspects 196 Factors influencing attack 199 Mathematical modelling of acid attack 201 Approaches in specification and design practice 203 Acid and seawater attack: exposure categories in EN 206–1 204 Specification by performance 206 Summary 206

194

11 Cracking in reinforced concrete structures Introduction 208 Mechanism of cracking 209 Chronological aspects of cracking 212 Cracking and the design phase 213 Cracking during the construction phase 220 Cracking during the service phase 225 Cracking and corrosion of reinforcement 229 Summary 230

208

x

Contents

12 Abrasion, erosion and cavitation Surface deterioration 232 Abrasion 233 Erosion 235 Cavitation 235

232

13 Weathering and efflorescence Weathering 237 Efflorescence 238

237

References Index

241 255

Preface

This book has been written at a time of change in European concrete practice, particularly in respect of specifying for durability. The introduction to European practice of non-harmonised standard EN 206–1 Concrete – Part 1: Specification, performance, production and conformity represents a significant step in raising awareness of the need to consider each potential deterioration mechanism when specifying concrete from a durability perspective. Despite this significant step, those who contributed to the drafting of the standard would not see it as reaching the journey’s end in respect of the methodology for specifying durable concrete. It will in time be seen as a first generation European standard. The new standard will do much to enhance the durability of future European concrete infrastructure although it is but a first step in a new direction. The exciting demands of flagship infrastructure projects, which are literally spanning the divide between regions and countries, require further development of our detailed understanding of deterioration mechanisms in concrete and reinforced concrete. Equally, the requirement that our future structures should represent significant examples of sustainable development, demands that we get the balance right between optimum use of materials and the costly risk of failure during a defined service life. Thus mathematical models of degradation are required which can be used in the probabilistic analysis of durability and life cycle costing. The next generation of European concrete standards will hopefully embrace to a greater extent the art and science of a well-reasoned engineering solution for the design of durable concrete structures. Current and future directions in the specification of durable European concrete form the context of the book. The emphasis is therefore on design and specification issues to the general exclusion of site practice. It is not intended to convey an impression that the production of a durable concrete artefact can be achieved without due regard for the skills of those who toil in all weathers and in difficult working environments to construct the structures we can be proud of. The subject matter of the book is but one link in the quality chain. This book is therefore intended as a basic text for three groups of readers who will play a major role in designing the quality structures of tomorrow.

xii

Preface

First, it provides an overview in respect of deterioration mechanisms for specifiers who find that, in applying the principles of new European Standard EN 206–1, they want to know a little bit more about each phenomenon that can lead to durability failure. Second, it presents a state-of-the-art review for postgraduate researchers about to embark on the task of advancing our understanding and modelling of a particular degradation mechanism. Third, it is a source of reference for undergraduates of engineering, architecture and building technology, and students of advanced concrete technology who find a need to read up on a particular deterioration process as part of their coursework.

Acknowledgements I wish to express my thanks to Tony Moore and Richard Whitby, Spon Press, for their encouragement during the various stages of preparing the book. I acknowledge with thanks permission from Seoirse Mac Craith to use his photographs in the production of Figures 1.1, 4.2, 7.3, 10.1 and 13.1. I am grateful to Mary Dunphy and Myles Christian for their expert advice on matters pertaining to the illustrations. The advice of Dr Peter O’Connor on chemical aspects of the subject matter is greatly appreciated. I express my heartfelt thanks to Rosemary Flynn, Anne Duffy and Joseph Duffy for their invaluable assistance during the preparation of the manuscript. Finally, I thank colleagues in various sectors of the concrete technology world for their friendship and supportive interest in the publication.

1

Framework for durability by specification

Context Economic forces make concrete the most suitable material for the majority of the world’s infrastructure. Concrete forms an indispensable element of motorways, airfields, harbours, canals, wastewater treatment facilities, and water supply schemes. High-rise buildings, once the preserve of steel, are now being executed in both concrete and steel/concrete composite construction. Concrete railway sleepers are displacing timber in many parts of the world. Thus concrete continues to play its crucial, if understated, role in the planet’s development. Practices in the concrete industry therefore have significant global effects. The most important construction issue emerging to the fore at the turn of the millennium is sustainability. Can our new structures be provided in a way that does not have a negative impact on the balance sheet of our planet’s finite resources? How do we determine an adequate specification for concrete, project by project, in a manner that will achieve satisfactory performance over the required life of the structure without squandering the earth’s resources? Over-specification is both wasteful of resources and unjust to the client. Under-specification, on the other hand, leads to premature and costly repair work, often with considerable disruption to third parties, such as road users, and ultimately a higher cost in financial and environmental terms. Towards the end of the twentieth century over a quarter of a million concrete bridges in the United States of America were classified as deficient and the number was increasing by 3500 per annum. The scale of the problem in Europe was somewhat less but was not insignificant. Figure 1.1 illustrates a typical example of corrosion damage to the soffit of a bridge. Clearly the state of the developed world’s highway infrastructure at the end of the twentieth century did not represent sustainable development. It well illustrates the challenge for concrete practitioners in the twenty-first century: a fundamental aspect of sustainability is the specification and achievement of durable concrete.

2

Fundamentals of durable reinforced concrete

Figure 1.1 Evidence of corrosion and spalling in a concrete bridge soffit

Key issues Durable concrete is quality concrete. A quality product is one that meets predetermined expectations. These expectations may be set out in a specification and can vary from one project to another depending on serviceability requirements. Durability is not an end in itself. In the context of European Commission directives, durability is not specifically listed as one of the essential requirements of a finished structure. However the achievement of essential requirements, such as mechanical resistance and stability, over the required life of the structure involves consideration of the implications of durability failure. It is also worth stating at the outset that durable concrete need not be maintenance-free concrete. These views may present a different starting point for the achievement of durable concrete than those that may once have existed, when the properties of concrete were equated with those of bedrock. Specification of durable concrete involves a mutual understanding by both the specifier and the end-user of what is meant by durability. A review of traditional codes of practice shows that we are not yet at the stage where the specifier and end-user engage in a structured framework that includes full consideration of the key issues. These key issues are not as yet definitive but should include the following:

Framework for durability by specification • • • • •

3

the design life of the structure or its individual elements; the serviceability requirement; a quantifiable description of the criteria that define serviceability failure; the acceptable level of risk; the permissible extent, if any, of maintenance.

Full application of these issues to future practice will require three developments. First, development and acceptance of mathematical models of deterioration that can be easily applied in practice. Second, development and acceptance of universally applicable tests for properties of concrete from which the likely future satisfactory performance of the concrete may be verified. Third, the adoption of a probabilistic approach to durability design with agreed values of the acceptable probability of failure or reliability index. The concepts involved are neither new nor complex. However they have yet to gain a significant foothold in practice due to the semi-traditional craft nature of guidance on concrete durability in codes of practice and product standards. A short review of the key issues may be helpful in introducing the concepts. Fundamentally, the design life is a notional value determined by the designer as a function of the required service life, defined by the user, and an appropriate factor of safety. The serviceability requirement may be descriptive, for example the onset of a stage in the corrosion–spalling cycle. Serviceability failure could be quantified in a number of ways, for example a maximum limit on the percentage of surface area in a deteriorated condition. Probability analysis is already implicit in the routine design for structural resistance of load-bearing elements and could equally be integrated in design for durability. The permissible extent of maintenance would depend on access, aesthetic, or economic considerations. For example, foundations may be inaccessible over the full service life; patch repairs might be unacceptable in visual concrete; functional obsolescence might render it uneconomic to invest in maintaining structures beyond a certain age. These key concepts underpin developments in the production of satisfactory specifications for durable concrete that can provide a better level of reliability than that experienced heretofore.

Historical review Strides have traditionally been made in the exploitation of concrete technology in respect of the structural performance of the material. Even today high strength concrete, part of the new generation of high performance concretes, is gaining market share. Strengths in excess of 100 MPa are not uncommon. Strength, however, should not be the main issue of interest to the concrete practitioner. Structural failures are rare but durability failure is all too common. Many studies, such as that by Mac Craith (1985), have highlighted the potential deficiencies in concrete over its service life, particularly in respect of corrosion of embedded reinforcement. Nevertheless, numerous

4


Figure 1.2 Chronological relationship between mix parameters and durability

examples exist of durable concrete structures in even the most hostile exposure conditions. Deficiencies exposed by serviceability failure must therefore point to shortcomings in execution of the concrete or inadequate specification. The historical decline in durability levels lay, partly, in the development of


5

Figure 1.3 Durability grade concept

higher cement strengths. The sequence is illustrated in Figure 1.2. The achievement of durable structural concrete has traditionally been linked to cement content. Prior to 1960 the cement content required to achieve specified strength was generally considered adequate for durability also. In selected instances of severe exposure the engineer might exercise judgement in specifying a higher cement content than that required for strength alone. Developments in the 1960s included increases in the strength of cement, refined tolerances in mix design and batching procedures, and increased efficiency through greater employment of statistical methods in quality control. The net result was that concrete strength grades and workability requirements were maintained with lower cement contents. Water/cement ratios were appreciably higher than a decade before and this led to higher permeability concretes. Concern for the serious implications regarding durability was voiced by some and the problem was addressed. An example is that of the inclusion in the United Kingdom Code of Practice CP 110:1972 of specific durability recommendations for a range of exposure classes. The recommendations took the form of limits on minimum cement content and maximum water/cement ratio. National durability grades The heightened awareness among designers of the need to specify for durability as well as strength led to anomalies. Specifications sometimes included redundant parameters with the result that the producer was forced to endure inefficient use of materials. Additionally the range included mixes that satisfied the designer’s intention regarding strength but not the durability requirement. This problem was overcome by the introduction of the ‘national durability grade’ concept, as proposed by Deacon and Dewar (1982). The basis of the concept is illustrated in Figure 1.3.

6


The ‘national durability grade’ introduced a link between specification for durability and a measure of its potential attainment. This was achieved through developing the accepted link between durability and impermeability. Achievement of impermeability was introduced in Code of Practice CP110 in the United Kingdom in the 1970s by control of maximum water/cement ratio and minimum cement content. However this in itself was an incomplete loop because the most commonly specified and tested parameter was twenty-eight day cube strength. Deacon and Dewar looked at the problem from another angle. Compressive strength is related to water/cement ratio and cement content: if one specified a particular concrete grade could one then be assured of meeting particular targets in respect of maximum water/cement ratio and minimum cement content? If this could be established for a large sample size, for example on a national basis, one could specify the two key durability parameters through concrete grade alone and test accordingly. Examples of the relationships determined by a survey of national practice are illustrated in Figure 1.4. This shows the results of a survey conducted by the Irish Concrete Society in 1998, analysed by West and Keating (1999). It may be seen that for any given grade of concrete within the national population one can establish an absolute maximum water/cement ratio and a minimum cement content or deduce statistical values, such as those applicable to a 95 per cent confidence interval. The earlier free-fall in durability level was arrested by the introduction of the durability grade concept. Nevertheless durability grade was seen as being a contributor to durable concrete but not the final solution. Two significant shortcomings existed. The first involved the variation of national material properties with time and the difficulty of tracking these changes in codes of practice. The second involved the continued use of exposure conditions categorised on the general basis of environment rather than on specific deterioration mechanisms. National durability grades can only be based on a survey of practice at a particular time. The resulting relationships, for example Figure 1.4, are critically dependent on the characteristics of local materials and these characteristics may change with time. A key factor, for example, would be the average cement strength. Figure 1.5 illustrates differences between grades established in different countries and at different times in the same country. A comparison is presented of United Kingdom and Irish practice in the 1990s and a comparison between Irish practice in the mid-1980s and late 1990s. It will be noted that higher cement strengths in the UK led to the requirement for higher concrete grades for a given cement content than those in the Republic of Ireland. Equally the gradual increase in strength over time of Irish cement is reflected in the higher concrete grades for a given cement content recorded in the later survey. The second shortcoming was that national durability grades were introduced into an existing system of exposure classes based, in the main, on a qualitative description of exposure condition. These qualitative descriptions,


7

Figure 1.4 Example of relationship between mix parameters and characteristic cube strengths based on a survey of national practice

categorised for example as ‘mild’, ‘moderate’, ‘severe’ etc., made reference to environments rather than specific deterioration mechanisms. This may have had the effect in some instances of over-simplifying the process for some specifiers who then overlooked relevant deterioration mechanisms. This shortcoming is being addressed in the next phase of development prompted by the introduction through the Comité Européen de Normalisation (CEN)

8


Figure 1.5 Variation of cement content and concrete grade relationship with time and between regions

of European Standard EN 206–1 Concrete – Part 1: Specification, performance, production and conformity, and its associated national documents, for example BS 8500 in the UK and DIN 1045–2 in Germany. European standard EN 206 Development of a European standard for concrete was a protracted process, indeed as it was for cement, extending over twenty years. The process began in the early 1980s with two proposed standards that appeared as prEN 206 and prEN 199 (ready-mixed concrete). These proposed standards failed to get the required support for publication as EN documents. The drafts were later merged into a single document and achieved European prestandard status as ENV 206 in 1989. Further development over the following ten years brought the document to EN status with the positive vote being recorded in the spring of 2000. Development of standard EN 206, and the relevant parts of design code Eurocode 2 such as cover to reinforcement, provided an opportunity to take a more rational approach to specification and design for durability. Durability-threatening mechanisms are considered in turn: risk of reinforcement corrosion; and the effects of carbonation, chloride ingress, freeze/thaw, and chemical attack. This has been framed in an exposure classification system which has eighteen subclasses designated by alphanumeric codes (Figure 1.6). Durability is specified either through the traditional practice of limiting values of concrete composition or by performance-related methods.


9

Figure 1.6 Exposure classification system in European standard EN 206–1

The former is likely to be more widely used. The standard requires that the intended working life of the structure shall be taken into account. Allowance is included for anticipated maintenance. In relation to the approach of limiting concrete composition, the common parameters are permitted types and classes of constituent materials; minimum cement content; maximum water/cement ratio; and (optionally) minimum strength class. In some cases additional requirements may need to be imposed, for example air entrainment or use of sulfate-resisting cement. Despite extensive deliberations it did not prove possible to frame a single set of values in EN 206–1 for use across Europe. Standard EN 206–1 could not cover all aspects of European concrete practice in a unified manner. The standard therefore requires or permits national standards bodies to publish provisions valid in the place of use and the relevant limiting values are presented in national complementary documents. Such an approach is not uncommon in European standards practice. Guidance on local conditions may be found in national annexes to European standards and normative references may be found in complementary standards. In the case of EN 206–1, for example, the British Standards Institution has published relevant values and framework in complementary standard BS 8500. The performance-related method is quite different. It allows the durability requirements to be determined in a quantitative way. Consideration is given to matters such as the intended working life and the criteria that would define

10 Fundamentals of durable reinforced concrete durability failure. The required parameters of the concrete may then be determined in one of three ways. The first utilises a comparison between the durability requirements of the project and those assumed in the traditional approach of limiting values of concrete composition. Parameters may then be determined through refinement of the values published in national complementary documents to EN 206–1. The second method involves use of performance criteria based on approved tests and involving conditions which are representative of those to be encountered. The third approach involves the use of predictive models. International research, particularly over the past two decades, has significantly brought forward an understanding of the phenomena that influence deterioration. Massive investment in a number of high profile European transport infrastructure projects has encouraged the application of this research to practice but it has been on a limited basis. The approach adopted differs from traditional practice by its consideration of the required service life, the relevant deterioration mechanisms and the use of predictive durability models. The core aspects of the first generation of mathematical models of deterioration are well developed but these prototypes require further development before being refined for use as routine design tools. Basically it is desired that one could design for durability in a probabilistic sense in a similar way to the current methods of design for structural resistance. Of equal importance, however, is that it is compatible with the drive to increase the use of performance-related specifications in European practice. It did not prove possible for the CEN Technical Committee responsible for EN 206–1 to issue the first version of the standard to be used in practice with durability parameters determined on a more scientific basis. It must be acknowledged, however, that the introduction of exposure classes based on deterioration mechanisms, rather than environments, is a major step forward. However the output is still based on national experience of local materials in local environments over the last couple of decades. It would have been preferable if the change in concrete practice could have advanced more directly to the extensive use of performance tests or design methods. Progress in this direction is being encouraged and is being pursued, for example, by those engaged in DuraNet, a network for supporting the development and application of performance-based durability design and assessment of concrete structures. It is intended that later versions of EN 206 will embrace developments in these fields. The development of durability-related performance tests and criteria is progressing steadily. The primary topics being studied are carbonation, freeze/thaw performance, and sulfate resistance. Other characteristics may prove worthy of investigation. For example, criteria for the acceptance of concrete cover based on a statistical evaluation of the achieved cover rather than a single minimum value. Tests for chloride ingress and abrasion resistance may soon prove applicable to more widespread practice. There have been inevitable setbacks – for example a test for water penetration was evaluated


11

as a measure of concrete quality but it is not being pursued further. Nevertheless, the core aspects of many performance tests for durability are well established although a lot of work remains to be done in achieving an acceptable level of precision.

Specifying durable concrete: the options Three main approaches may be distinguished for the specification of durable concrete: • • •

all-encompassing prescriptive approach; deterioration-specific prescriptive approach; durability design method and performance testing.

The method used through recent decades is represented by the first approach. The second approach is a key feature of the introduction to practice of European Standard EN 206–1. It will be the most commonly used methodology for some time to come despite its recognised shortcomings. The third approach, introduced through an informative annex in the first version EN 206–1, provides an alternative method which is potentially more reliable but is likely to be used in a minority of cases until research progresses further in the areas of deterioration modelling and performance-based specifications. All-encompassing prescriptive approach The all-encompassing prescriptive approach involves consideration of deemed-to-satisfy limits. It is exemplified by the durability clauses in codes of practice such as British Standard BS 8110 and National Standards Authority of Ireland IS 326. The tables relating concrete quality and cover to reinforcement, in the form of that presented in Table 1.1, is based on the concept of the national durability grade. Control of the minimum cement content and maximum water/cement ratio may be achieved through specification of an appropriate minimum grade of concrete. The most noteworthy issue is that deterioration mechanisms are not explicitly considered. The specifier uses instead an all-encompassing environmental classification system based on a qualitative description of the conditions of exposure (for example, ‘moderate’ or ‘severe’). Consideration of specific threats, such as alkali–silica reaction, is covered by reference to clauses or other documents containing prescriptive advice. This all-encompassing methodology relies to a considerable extent on the exercise of engineering judgement. Guidance is given on interpreting the descriptions but it cannot be exhaustive. Examples arise where the specifier will feel that their final choice remains somewhat subjective. Doubts about the adequacy of this prescriptive approach for specification in chloride-laden

12 Fundamentals of durable reinforced concrete Table 1.1 Format of traditional prescriptive approach to durability Exposure classification

Nominal cover

Mild Moderate Severe Very severe Most severe Abrasive

NC2 – – – – –

NC1 NC4 – – – –

NC1 NC3 NC5 NC6 – –

NC1 NC2 NC3 NC5 – –

NC1 NC1 NC2 NC3 NC6 *

W5 C1 G1

W4 C2 G2

W3 C3 G3

W2 C4 G4

W1 C5 G5

Concrete properties Max. free water/cement ratio Min. cement content Lowest grade of concrete *Additional requirements may be noted

environments have been expressed by, for example, Browne (1986) and Bamforth (1994). The dilemma faced by the specifier may be considered by an analogy with structural design. A designer could not size a reinforced concrete beam with an acceptable level of reliability if the information on span and loading was not quantified but was merely classified as ‘moderate’ and ‘severe’ respectively (Figure 1.7). It is readily apparent that such a system would lead to an unacceptably low factor of safety in certain circumstances and the wasteful use of excessive material in others. The all-encompassing prescriptive approach is not objective. It cannot be used in economic optimisation because it does not take account of the required service life nor the required reliability. The method is cumbersome in adapting to the benefits of emerging technologies in the form of new materials and construction techniques. A code based solely on the allencompassing approach cannot, for example, easily harness the potential benefits of new technologies. The specifier, for example, would find it difficult to do a cost–benefit analysis on the use of controlled permeability formwork or corrosion inhibitors because their use would not specifically change the prescriptive requirements of the code. The shortcomings of the all-encompassing prescriptive approach may be seen by an examination of four key aspects: • • • •

validity of a prescriptive approach; exposure class selection; the relationship between strength and mix parameters; site practice.

A fundamental issue is the validity of a prescriptive approach. The specifier is relying solely on the perceived relationship between a descriptive exposure


13

Figure 1.7 The dilemma posed by a qualitative description of the design constraints

class and prescribed mix parameters (minimum cement content and maximum water/cement ratio) to ensure adequate protection. Reliance on simplified empirical relationships or previous satisfactory experience is total. The second issue is the identification by the specifier of the appropriate exposure classification based on descriptive clauses. The description of what is covered by the exposure classes in national codes has usually been quite comprehensive and so accurate judgement is likely. Nevertheless it is not the most efficient method of ensuring that all relevant deterioration mechanisms have been identified and considered. The likelihood of over- or underspecification is significant. The third issue is the assumed relationship in a national code between concrete grade and current industry practice regarding minimum cement content and maximum water/cement ratio. This relates to the fact that the concrete durability grades quoted in codes are based on industry norms in a country as surveyed at a particular time. Changes to practice due to changes in material characteristics, such as cement strength, may alter the relationships. It is impractical to expect codes of practice to rapidly adapt to subtle market changes. Thus the weakest link in the chain influences the durability of the concrete but the chains themselves are not amenable to regular scrutiny. The fourth point relates to the implicit assumption that achievement of key parameters such as specified cube or cylinder strength will be accompanied by

14 Fundamentals of durable reinforced concrete proper site practice regarding compaction and curing so as to achieve the full durability potential of a given mix. Poorly cured concrete may have an unacceptably high permeability in the cover zone despite being made with an appropriately specified concrete of high quality. The durability properties of the cover zone in the member as-built, other than the depth of cover, does not specifically form part of the acceptance criteria. Prescribing the right mix does not necessarily guarantee the achievement of the required level of impermeability. Recognition that the all-encompassing prescriptive approach has failed in certain cases in the past has resulted in more demanding values being placed on water/cement ratio and other criteria. There is a limit however on how far one can push the numbers without causing other problems with concrete such as ease of placing and compaction. Deterioration-specific prescriptive approach The potential shortcomings of the all-encompassing approach will be addressed, in part, by the deterioration-specific approach of European Standard EN 206–1. Specification remains prescriptive in that minimum binder content, maximum water/binder ratio, and (optionally) minimum concrete grade are output from consideration of exposure subclasses. However the specifier is forced to consider the most onerous condition from a combination of deterioration mechanisms and environmental conditions in the determination of limiting values of concrete composition. Some national complementary standards will also allow a trade-off between concrete quality and cover. The essential advance over the all-encompassing approach is that of making the exposure/environmental classifications more comprehensive. The downside is that it makes the process more complicated through the introduction of the greatly increased number of exposure classifications. Practitioners probably feel that there are too many classes while some researchers feel that there are still too few! The numerous subclasses are required to take account of environmental characteristics that may influence the rate of deterioration, for example wetting and drying cycles in the case of carbonation. The details of the eighteen subclasses are presented in Table 1.2. The basic format of the limiting values table is presented in Table 1.3. This illustrates an overview of the structure. The limiting values of concrete composition and properties are set nationally and published in accompanying documents to the European standard. Minimum strength class is presented using a dual designation system based on cube and cylinder strengths. The standard was formulated on the basis of cylinder strengths and the cube strengths are merely an approximation to the cylinder strength – hence the somewhat awkward dual reference system, exemplified by a ‘Grade C30/37’ concrete.


15

Table 1.2 Summary of exposure classes and environments in EN 206–1 Degradation phenomenon

Subclass

Environment

No risk of corrosion or attack

X0

Corrosion induced by carbonation

XC1 XC2 XC3 XC4 XD1 XD2 XD3 XS1 XS2 XS3 XF1 XF2 XF3 XF4 XA1 XA2 XA3

Unreinforced concrete: all exposures except freeze-thaw, abrasion, chemical attack Reinforced concrete: Very dry Dry or permanently wet Wet, rarely dry Moderate humidity Cyclical wet and dry Moderate humidity Wet, rarely dry Cyclical wet and dry Exposure to airborne salt Permanently submerged Tidal, splash and spray zones Moderate water saturation, no de-icing agent Moderate water saturation, de-icing agent High water saturation, no de-icing agent High water saturation, de-icing agent or sea water Slightly* aggressive environment Moderately* aggressive environment Highly* aggressive environment

Corrosion induced by chlorides other than from seawater Corrosion induced by chlorides from seawater Freeze/thaw attack

Chemical attack

*Quantified in respect of the chemical characteristics of groundwater (SO42, pH, CO2, NH4, Mg2) or soil (SO42, acidity)

The approach addresses one of the four shortcomings raised in relation to the all-encompassing prescriptive approach – that of exposure class selection. By presenting the specifier with an expanded suite of exposure classes, which more closely reflect specific deterioration mechanisms, it is more likely that specification for durability will be more reliable. Nevertheless the other three concerns remain: validity of the prescriptive approach, changing relationships over time between concrete grades and mix composition and achievement of potential durability through proper site practice. Durability design and performance testing The third approach to the specification of durable concrete is radically different to the foregoing. The durability design method involves consideration of each relevant deterioration mechanism and the expected service life of the structure in a quantitative way. Appropriate material parameters may then be determined based on an acceptable probability of failure. The all-encompassing prescriptive approach was introduced to practice at a time when the deterioration mechanisms were less well understood. Research, particularly over the last two decades, has identified the dominant

Maximum water/cement ratio appropriate to each exposure subclass Minimum strength class appropriate to each exposure subclass Minimum cement content appropriate to each exposure subclass Minimum air content appropriate to XF exposure subclasses Other requirements: for example freeze/thaw resisting aggregates, sulfate-resisting cement, etc.

Carbonationinduced corrosion XC1, XC2, XC3, XC4

No risk

X0

XD1, XD2, XD3

Chloride other than sea water XF1, XF2, XF3, XF4

Freeze/thaw attack

XA1, XA2 XA3

Aggressive chemical environment

Values published nationally by standards authorities

XS1, XS2 XS3

Sea water

Chloride-induced corrosion

Table 1.3 Framework for limiting values of composition and properties of concrete appropriate to exposure class


17

mechanisms and the key parameters controlling the rates of deterioration. Further work on the proposal and refinement of mathematical models of deterioration is continuing. The models for carbonation, chloride ingress and corrosion propagation are the most advanced to date but still require refinement. Excellent progress has been made in the last few years on adopting the probabilistic approach, commonly used in structural design, for determination of durability parameters. The durability design approach allows consideration of the type of structure, material properties, microclimates, required life, quality of site practice and the probability of failure. Thus it addresses the shortcomings of the prescriptive approach. The specifier may envisage a prescriptive approach as being part of European standards for some time to come but the employment of durability design and the development of performance-based specifications is to be encouraged, especially on critical projects. Standards for certain phenomena continue to develop along the prescriptive route, for example alkali–silica reaction. Nevertheless significant advances are being made on topics related to corrosion of reinforcement. Three avenues are available for determining the appropriate parameters: • • •

satisfactory experience; performance test methods; predictive models.

The avenue of specification based on satisfactory experience allows account to be taken of long-term satisfactory experience with concrete specified for similar works in a particular environment. The proposed works would involve application of similar materials and practices in an environment where the satisfactory history was established. Essentially it is a refinement of the prescriptive approach but the link between specification and performance is clearly established. The second alternative route involves performance test methods. A specified performance would be defined for a relevant deterioration mechanism. The concrete producer would then prove the adequacy of a proposed mix by reference to a relevant approval test. The test would demonstrate the potential of the proposed mix to meet the defined performance level. The test might be conducted specifically for the contract. Alternatively the concrete mix might be accepted where adequate performance has been established by previous tests on concretes of similar materials and weaker mixes (for example, higher water/cement ratios). Control testing could then be used to monitor an agreed key element of the mix. This would highlight any departures from the agreed mix. The third possibility involves the use of predictive models. Considerable progress has been made on the mathematical modelling of deterioration mechanisms, especially those related to corrosion initiation. The route

18 Fundamentals of durable reinforced concrete provides the key to better design of new works and management of the durability of existing stock. The predictive models will relate the rate of deterioration, for a given mechanism, to key measurable parameters of the concrete or mix constituents. Acceptable models will be calibrated against data from the in-service behaviour of structures. The development of the models involves consideration of the often complex chemical and physical interactions present in a deterioration mechanism. Translation of the models into workable design aids may involve incorporation of simplifying assumptions. This would not necessarily detract from the integrity and sophistication of the approach since these steps would contribute in a quantifiable way towards the factor of safety which is an integral element of prudent design and management of risk. The practitioner will find more advanced treatment of these topics and software in publications under development, such as that by The Concrete Society and Taywood Engineering Ltd.

Durability and the next generation of standards The launch into practice of the first generation of European concrete standards, especially EN 206–1 and EN 1992, in the early years of the second millennium will not fully reflect the state of knowledge regarding durability design which existed at the end of the first millennium. The nature of standards and codes preparation is such that they must inevitably lag behind technical developments. Considerable progress on durability design has been made, notably through the DuraCrete project of the European Commission’s Fourth Framework Programme and in practice in the Netherlands. It is likely that the next generation of standards will still embrace a form of deemed-to-satisfy approach for routine building projects. However the inclusion of more sophisticated quantitative approaches in a normative way for critical projects will probably be evident. The research which will underpin the changes to the second generation of codes and standards will be focused on validating deterioration models from laboratory and field trials and on the broader issue of applying risk analysis to durability specification. The risk analysis approach will involve a significant change to specification for durability by forcing specifiers and, significantly, clients to consider service life in a probabilistic way. The client will get what they pay for. Indeed the extension of the risk analysis approach to embrace life cycle costing will be an obvious step.

Summary At the turn of the millennium the risk of concrete failing to perform satisfactorily over its service life should be low, given the current understanding of deterioration, yet it remains unacceptably high. The cause of the problem lies in the gulf that exists between simplified descriptions of exposure conditions


19

and the complex deterioration mechanisms that act alone or in concert. The need to bridge this gulf is a pressing one. Durability of concrete needs to be ensured in future works from both an economic and an environmental viewpoint. The economic argument is a straightforward one – premature expenditure on repair or demolition and the cost of disruption to production reduces a client’s competitiveness and reflects badly on the concrete industry. Equally important in modern society are the green issues. Unforeseen maintenance involves wasteful use of the earth’s scarce resources of both materials and energy. Demolition due to unserviceability involves production of construction waste for which the planet has a dwindling capacity to absorb. Future generations will not look kindly on today’s specifiers if load-bearing frames of reinforced-concrete buildings are not durable enough to allow building rejuvenation through façade replacement and interior re-fit. Substantial research on concrete durability is in progress throughout the world. In Europe it had been hoped that the fruits of this research would have been fully incorporated into the first generation of European concrete standards EN 206–1 and EN 1992. Regrettably the required package of research knowledge and in-service experience is not sufficiently complete to allow this. Nevertheless progress is apparent in that specifiers will be required to consider the various deterioration mechanisms in a structured framework rather than relying on a qualitative exposure classification system. The prospects for ensuring the durability of concrete structures over the period of their intended service lives is now promising. The introduction of new materials and construction techniques offer higher quality concretes with adequate impermeability. Equally exciting, however, is an enhanced understanding of the parameters that influence the rate of reinforced concrete deterioration. This knowledge is introducing a new way of approaching durability through design methods based on scientific and engineering principles. Prescriptive approaches still have a role to play but it behoves the specifier to make the best use of the considerable advances made in recent decades in understanding the properties of durable reinforced concrete. The following chapters present the specifier with background to the factors which influence the selection of limiting values in the EN 206–1 framework and provide an overview of research developments to those about to engage in the challenging quest of developing predictive models, performance tests, and probability-based durability design methods.

2

Probabilistic approach to durability design

An integral concept of specification for durability in structures is that the material properties should meet the performance requirements over a defined life. This concept has been in the background in standards and codes of practice for over 50 years. For example in the United Kingdom, Code of Practice CP3, Chapter IX on durability (Council for Codes of Practice for Buildings 1950) included definitions of both the ‘designed life’ and the ‘satisfactory life’. More recently British Standard BS 7543 (1992) has been published to cover the durability of buildings and their elements. Significantly, in the quest to provide a rational basis for achieving design life, a joint RILEM/CIB group developed a systematic methodology for service life prediction (Masters and Brandt 1989). This has been adopted for standardisation by the International Standards Organisation (1998) as ISO 15686 ‘Service Life Planning of Buildings’. The RILEM/CIB methodology built on earlier work from Sweden (Sentler 1983) and in the Netherlands (Siemes et al. 1985). This was further studied and a clearer focus on the way forward was provided by the publication of RILEM Report 14 (Sarja and Vesikari 1996) and CEB Bulletin 238 (Schiessl et al. 1997). Meanwhile a significant impetus to the development of the tools necessary to apply the service life principles was provided through a European project ‘DuraCrete’ (Probability Performance Based Design of Concrete Structures). Although the stage has yet to be reached whereby codes and standards will fully embrace the concepts, degradation models of a deterministic nature have been used on major infrastructure projects in Denmark and a stochastic model was used to document the required service life in a contract in the Netherlands.

Design life An overview of the concepts of design life and service life is illustrated in Figure 2.1 in a development of work by Tuuti (1982) and Sarja and Vesikari (1996). This shows that the structure reaches its design life when the maximum tolerable level of damage is reached. An acceptable value of the service life may then be evaluated, for example by statistical considerations. This introduces the concept but in practice the degradation mechanism can itself


21

Figure 2.1 Concept of service life prediction based on durability considerations

be modelled stochastically and the risk of reaching the limit of damage is considered probabilistically. This is explained in later sections. The notion of design life has been described (Rostam 1984) as the combination of possible technical life with economic considerations. Possible definitions of design life have been advanced. For example: ‘the minimum period for which the structure can be expected to perform its designated function without significant loss of utility, and not requiring too much maintenance’ (Somerville 1986); ‘the period of use intended by the designer’ (Concrete Society 1996); and ‘the period of time after installation during which a building or its parts exceed the performance requirements’ (ISO 1998). A definition of the ‘working life’ is presented in EN 206–1 (Comité Européen de Normalisation 2000a) in terms similar to that of Somerville’s design life. The design life of a building would be determined based on the service life expectation of the client. It is unlikely that clients will be very specific but it should be possible for them to indicate the category of structure in descriptive terms such as ‘temporary’, ‘normal’, or ‘major infrastructural’. The designer can then relate this to the classification of structures in standards and codes that contain information on service life expectation. Other than the case of temporary structures, the choice for non-renewable elements will typically be between a fifty-year life or a 100-year life. Even this simple choice allows the specifier to get the client thinking about the balance between first cost and life-cycle cost. The days are gone when clients could expect their reinforced concrete assets to be forever maintenance-free.

22 Fundamentals of durable reinforced concrete Design life may thus be seen in the context of supporting decisions relating to specification of material properties and could therefore be extended to life-cycle cost analysis. However it must be seen as a notional concept – the structure will not necessarily reach the failure criterion at the appointed year. Those assessing the durability of existing concrete structures often find variable corrosion activity, which demonstrates the variable nature of the phenomena involved. It does not mean however that a rational approach to specification for durability is impossible. Specification for durability is amenable to mathematical analysis and design but the methods should take account of variability. This implies the use of a statistical approach to the problem.

Structural design analogy The use of statistical methods is highly appropriate in durability design because the fundamental requirement is to minimise the risk of failure. This introduces the concept of risk analysis. Durability design can be achieved in the context of a defined probability of failure. Equally one may refer to a defined level of reliability. The approach is directly comparable with that used in design for structural resistance. In the past specifiers and clients or owners may have had a vague idea of what their expectations were in respect of service life but it was not specifically considered in practice, except in rare cases such as temporary structures. Harnessing the design life concept allows the selection of appropriate materials on a rational basis. This involves the need to model the degradation in performance with time-dependent functions in the context of an overall approach to durability design. A number of approaches are possible and several have been advanced (Siemes et al. 1985, Sarja and Vesikari 1996, Siemes and Rostam 1996). Consideration of the various approaches to durability design involves an appreciation of the meaning of safety factors, mean values, design values, distributions, performance functions, and capacity functions. At first the concepts may seem new to the specifier who has traditionally worked only with ‘deem-to-satisfy’ rules for durability but a comparison with traditional structural design methods demonstrates that the principles are the same. The fundamental principles that underlie the application of probability analysis to durability design may best be introduced, therefore, by first reviewing a familiar example from structural design codes. The example chosen for demonstrating the principles is that of providing adequate resistance to bending in the design of a reinforced concrete beam. Consider the case of a singly-reinforced concrete beam, simply supported at each end, and carrying a uniformly distributed load (Figure 2.2). The beam may fail in a number of ways but for the purpose of this example only the phenomenon of bending will be considered. The design of the beam is based on the assumption that it will fail if


23

Figure 2.2 Conditions assumed in structural design analogy

the maximum demand bending moment, occurring at mid-span, exceeds the ultimate moment of resistance of the section. Thus we may impose a constraint: RS0 where R ‘resistance’ (ultimate moment of resistance of section) and S ‘load’ (maximum bending moment). To convert this relationship into a form that yields design values for the section we formulate relationships describing the resistance of the section based on concrete failing first, the resistance of the section based on the reinforcement yielding first, and the maximum demand bending moment. Thus: R f ( fcomp, b, d) and R f ( fsteel, As , z) while S f (W, L) where fcomp is the compressive strength of the concrete, b is the width, d is the effective depth, fsteel is the tensile strength of the steel, As is the area of steel reinforcement, z is the lever arm, W is the total load, and L is the span. It may be further noted that z is a function of d and that d is a function of the total

24 Fundamentals of durable reinforced concrete

Figure 2.3 Theoretical profile of actual bending moment in service compared to maximum and minimum moments of resistance.

depth h. The design problem becomes one of selecting values and either validating that the constraint of the formula is met or solving the formula for an unknown value. The issue of uncertainty now enters. The intended values of width and total depth are generally assumed to remain constant during the service life and are readily achievable, within specified tolerances, during construction. What of the assumed concrete strength and load, however? The actual concrete strength that will be achieved in the structure cannot be known with certainty at the design stage. It will depend on the materials selected by the producer, the degree of compaction achieved by the operatives, and the curing conditions, which will be influenced in part by the weather at the time of construction. The load is variable also. The imposed load component in particular will fluctuate on a daily basis to a degree that is primarily dependent on the function of the building but what of the unexpected loads that may arise over the lifetime of a building? Design based on the worst case – weakest possible concrete and potentially heaviest load together with an allowance for design and construction blunders – would unnecessarily result in ungainly structural elements with a high moment of resistance (Rmax). These would be uneconomic, would severely limit the technical advancement of span, and would fail to meet the requirements of sustainable development. On the other hand, design based on the most optimistic case – concrete achieving its full potential strength, no unexpected load combinations, and no allowance for blunders – would result in elements with low moments of resistance (Rmin) which would have a high probability of failure. This is


25

Figure 2.4 Concept of design strength and design load based on statistical considerations

illustrated in Figure 2.3, which charts a theoretical profile of the bending moment values (Sactual ) resulting from load combinations that vary with time, as may be expected in reality. It is a question of balancing economy and safety in an acceptable way. This is achieved by the use of characteristic values of strength and load based on probabilistic considerations and by the application of safety factors. The characteristic values of strength and load are determined by consideration of the mean values encountered in practice, their variability and the application of statistical parameters. The characteristic strength is determined by reducing the mean strength by an amount based on a chosen multiple of the standard deviation, while the characteristic load is based on values above an anticipated mean. These values are modified further to produce design values through the application of partial safety factors (Figure 2.4). Thus the design problem may be solved in a deterministic way despite the fact that the values of the moment of resistance (R) and design bending moment (S) are calculated in an approach that incorporates allowances for the variability encountered in practice. The underlying probabilistic nature of the problem is further reinforced by the fact that the magnitude of both the factors of safety and the allowance for variability in the design codes has been selected to yield an acceptable probability of failure. The concept is illustrated in Figure 2.5. Thus while it appears that the problem being solved is of the form:


Figure 2.5 Incorporation of probabilistic concepts into an essentially deterministic model

RS0 it is set in the wider context of ensuring that the probability of failing to satisfy this condition is less than a maximum allowable failure probability (Pf max). This may be stated as follows: P {R S < 0} Pf max The application of these principles to durability design was considered by RILEM Commission TC 130–CSL (Sarja and Vesikari 1996). Further development in application of the principles to practice will hopefully influence the durability clauses in the second generation of European standards for concrete.

Approach to design for durability The mathematical solutions to the problem of design for durability are not quite as straightforward as those presented in the structural design analogy. Allowance for the multitude of variables and their distributions in durability design problems can sometimes lead to complex mathematical solutions.


27

Thus a number of approaches have been proposed, each of which may be more readily applicable in any given case depending on the deterioration mechanisms involved. The application in each case, however, involves aspects that are familiar from the principles of the approach to structural design. Although the terminology in the literature varies for similar approaches, broadly the approaches may be described as the ‘Lifetime Safety Factor Method’, the ‘Intended Service Period Design’, and the ‘Lifetime Design’. The first of these is essentially deterministic but the others are stochastic design methods. Lifetime Safety Factor method The concept of the Lifetime Safety Factor method first involves consideration of the following: •

•

• •

a function (R or R(t)) which describes the resistance of the structure; it may or may not be time-dependent; it may be based on mean values of input parameters; a function (S or S(t)), which describes the ‘load’ (for example, the chloride level) on the structure; it may or may not be time-dependent; it may be based on mean values of input parameters; the safety margin (R(t) S(t)); the mean service life (tmean).

The resistance of the structure may reduce with time until failure is reached when it equals a pre-determined value of the ‘load’. In such cases R(t) is described as a ‘performance model’. Equally, the resistance of the structure may be constant but the load may increase with time until failure is reached when it equals a pre-determined value of the resistance. In such cases S(t) is termed a ‘degradation model’. Note that the requirement may be phrased in a limiting manner: for example, the ability of a concrete element to keep a given parameter below a certain level. Degradation models are common in durability design problems. For example, in the case of the carbonation phenomenon the depth of carbonation increases with time and may eventually reach the reinforcement; in the case of chloride ingress, the level of chlorides at the reinforcement builds up over time and may reach a critical corrosion threshold level. Consider the case of a degradation problem. The conditions which prevail during the mean service life (Figure 2.6) are such that: R S(t) 0

t tmean

and R S(tmean) 0


Figure 2.6 Profile of a degradation model for durability design

It may be seen from Figure 2.6 that, due to the distribution of values of S(t), the possibility of the condition {R S(t) 0} being met occurs in advance of tmean and that the probability of this occurring increases with time. Thus the maximum allowable failure probability must be considered (Pf max) leading to a design constraint that: P {R S(t) 0} T Pf max

T tmean

and this leads to a target service life (tg) which is the time at which Pf max occurs. The challenge for the specifier is to meet the target service life (tg) that represents the client’s expectation of service life. To achieve this duration of satisfactory performance it is necessary to specify based on the anticipated mean service life (tmean), which becomes the design service life (td). This introduces the concept of the ‘Lifetime Safety Factor’, which is described by the relationship: td t tg where td design service life t ‘Lifetime Safety Factor’ tg target service life Introduction of probability to durability design in a similar way to that of


29

structural design for the traditional form of ‘loading’ requires a more simplified route than that of calculating probabilities. It may be shown (Sarja and Vesikari 1996) that, once values for ‘Lifetime Safety Factors’ have been calibrated, design would involve the following steps: • • •

• •

•

agree the target service life of the structure; determine the design service life based on the Lifetime Safety Factor and the target service life; apply the relevant degradation model or performance model using the design service life and select appropriate material properties, sections sizes, and/or protective measures; check that the reduction in the safety margin, for example R S(t), from time t 0 to time t tg is less than an allowable value; if the reduction in safety margin is too great, redesign using higher performance materials, or larger sections, or introduce additional protective measures; if the redesign fails to produce a satisfactory value of the minimum safety margin consider agreement with the client of a shorter target service life.

The Lifetime Safety Factor method provides a good introduction to the concept of extending structural design principles to durability design. The next development is to extend consideration of the probability of failure as a significant criterion. Two other methods, which are essentially both sides of the same coin, have been described (Siemes and Rostam 1996) as the ‘Intended Service Period Design’, and the ‘Lifetime Design’. They approach the same problem from different perspectives but yield the same outcome. The choice of method depends on the information available at the design stage. Intended Service Period Design method The principles underlying the Intended Service Period Design and the Lifetime Design methods are similar to the Lifetime Safety Factor method. However the methods extend the use of probabilistic analytical tools. In relation to the Intended Service Period Design, consider the following: • • • • •

the function (R or R(t)) which describes the resistance of the structure; the function (S or S(t)), which describes the load on the structure; the condition which represents a technical failure (R(t) S(t) 0); the maximum allowable failure probability (Pf max); the target service period (tg) during which it is expected that the probability of technical failure will not exceed the maximum allowable level.

The probability of the difference between resistance and capacity becoming negative at least once during the target service period may be calculated at any time t:


Figure 2.7 Probability distribution curves showing the effect of different load parameters

P {R(t) S(t) 0}T

where T tg

This value may then be compared with the maximum allowable value of the failure probability to calculate the probability of the former being less than the latter. This may be stated as follows: Pf,T P {R(t) S(t) 0}T Pf max

where T tg

Thus it is possible to chart a series of values of Pf,T for different values of t and check whether or not the required level of reliability is achievable during the target service life. The effect of changing parameters in the resistance or load functions may also be examined by plotting the probability distribution curves. Typical curves are illustrated in Figure 2.7. The shape of the curves will be familiar from the traditional quality control procedure of assessing by statistical means the acceptability of a batch of goods while only testing a sample. Thus design would involve the following steps: • • •

agree the target service life of the structure by consultation with the client; apply the relevant degradation and/or performance model using the mean values of the input parameters together with their distributions; check the probability of the resistance falling below the required level during the service life, or equally check the probability of the load exceeding the available level of resistance;

Probabilistic approach to durability design • •

•

31

chart the failure probability through consideration of the acceptable maximum level of failure; if the failure probability is unsatisfactory, redesign the mean values of the input parameters by specifying higher performance materials or larger sections and/or reduce the distributions by specifying stricter quality control measures if this is feasible; if the redesign fails to produce a satisfactory level of failure consider if the client could accept a shorter target service life or a higher risk of failure.

Lifetime Design method The principles of the Intended Service Period Design may also be used in the Lifetime Design but the problem is approached in a different way. For example in the former one might consider the probability of the level of chloride at the reinforcement reaching 0.4 per cent by weight of binder during the target service life. In the latter, one considers the probability of the service life being less than the target service life for the condition whereby the chloride level is 0.4 per cent by weight of binder at the reinforcement. The first method is used where the distributions of the performance and load are known. The Lifetime Design method is used where the distribution of the service life is known or may be assumed to follow a certain profile. The Lifetime Design method uses the functions R(t) and S(t) to formulate a relationship in terms of the life of the structure (L): L f (R(t), S(t)) Knowing (or assuming) the appropriate model of the service life distribution and the maximum permissible value of the probability of failure it is possible to evaluate the target service life. The probability of the calculated life of the structure being less than the target service life may be determined: P {L tg 0} The concept is illustrated in Figure 2.8 for a deterioration process whose time-dependency promotes a higher risk of failure in earlier years. A further calculation may be made of the probability of failure. This is the probability of the difference between the expected life and the target service life being less than an acceptable value: Pf P {L tg 0} Pacceptable As before, the effect of changing parameters in the resistance or load functions may be examined by plotting the probability distribution curves as illustrated in Figure 2.9.


Figure 2.8 Lifetime design: service life distribution

Thus design would involve the following steps: • •

• • •

•

agree the target service life of the structure by consultation with the client; apply the relevant degradation and/or performance model, using the mean values of the input parameters together with their distributions, to determine the calculated life of the structure; check the probability of the calculated life falling below the target service life; chart the failure probability through consideration of the acceptable maximum level of failure; if the failure probability is unsatisfactory, redesign the mean values of the input parameters by specifying higher performance materials or larger sections and/or reduce the distributions by specifying stricter quality control measures if this is feasible; if the redesign fails to produce a satisfactory level of failure consider if the client could accept a shorter target service life or a higher risk of failure.

Use of the Lifetime Design method involves knowing the service life distribution. The distribution will depend on the deterioration process under consideration. The distributions will not necessarily be normal. Some phenomena lead to a situation whereby the values are not evenly distributed about the mean with damage occurring more frequently in earlier years than later. This would be typical in the case of deterioration models involving the


33

Figure 2.9 Probability distribution curves showing the effect of different resistance parameters

square root of time. These models include the significant cases of corrosion initiation by carbonation and chloride ingress. The service life distribution for such phenomena may be modelled by, for example, the log-normal distribution.

Future research needs The translation of theory to practice represents a great challenge for researchers and specifiers in the field of durability design. The reaction of many specifiers to the introduction of the extended prescriptive approach of European standard EN 206 is likely to be that it seems overly complex. Imagine then the challenge of introducing the probabilistic approach to durability design into practice. Nevertheless specifiers recognise that an improvement in the method of specification for durability is required in the case of structures which are required to have a long service life. The traditional prescriptive approach has a role to play but it cannot fulfil the need in all cases. The inclusion of probabilistic methods of durability design in codes and standards involves a number of stages of development. Significant progress has been made in the 1990s but much work remains to be done. The path ahead may be charted as follows: • •

development of accepted models of deterioration and the conversion of these models into readily-applicable design tools; calibration of the models with experience of real structures;

34 Fundamentals of durable reinforced concrete • • •

selection of favoured methods of probabilistic durability design; definition of applicable limit states; determination of acceptable levels of failure probability.

Many models already exist for various deterioration mechanisms and a review of the most widely referenced ones are presented in later chapters. It is clear that in some cases there is great diversity in the models available. This is caused by the complex interaction of physical and chemical phenomena involved in each deterioration mechanism. Much of the research done in the production of the models is very valuable but a need exists to focus attention on a smaller number of models for development into internationallyacceptable design tools. Models used as design tools need to be relatively straightforward so that they can be used on an everyday basis. Complex formulations based on parameters that cannot be readily specified or tested are unlikely to find favour. The refinement, or to put it more correctly, the coarsening, of the models will involve replacement of terms which are difficult to test routinely. Practitioners will want to be able to specify on the basis of formulae that include properties that can be checked for compliance. Replacement of terms through conservative assumptions may involve loss of exactitude but this will be compensated for by an increase in reliability in service. Coarsening of the models need not imply that oversimplification will result. The projects employing probabilistic durability design methods will most likely be high profile, multi-million Euro projects where adequate design time will be affordable. Furthermore the use of software will ease the demands of statistical analysis. Complex models are unlikely to be required on routine low to medium budget projects because the prescriptive approach of EN 206 would suffice. Input parameters for the models will need to be established by calibration with practice. The type of distribution appropriate to each input parameter will need to be studied and agreed. The typical magnitude of standard deviations will also need to be assessed. A wealth of data exists in both published research studies and in unpublished commercial test reports on deteriorated structures carried out as part of investigations. The challenge is to harness this data in a focused manner. Serviceability Limit States need to be debated and agreed. A wide range of possibilities exists even in the single case of crack development in cover concrete from corroding reinforcement. For example in some structures the serviceability limit state would be reached at the appearance of the first crack whereas in others a certain level of spalling would be tolerable. Circumstances will vary – the risk of concrete spalling is less acceptable in the case of a bridge over a motorway than for a bridge over a minor river. Research is also required on defining the acceptable level of probability failure or, if preferred, the reliability index. The codes tend to specify the reliability index rather than the probability of failure. The relationship


35

Figure 2.10 Relationship between acceptable level of failure probability and reliability index

between the acceptable level of failure probability and the reliability index is illustrated in Figure 2.10. The terms are related by the following formula: Pf y ( ) where Pf probability of failure y normal distribution function reliability index For example, if is two the area outside that multiple of standard deviations from the mean, on one tail of the curve, is about 3 per cent, giving a failure probability of the order of 102. A starting point may be made by studying the values in existing codes of practice and draft Eurocodes. To indicate the order of magnitude of probability of failure appropriate to durability studies Bamforth (1999) referred to typical code values. To take the extreme case of collapse, the reliability index for the ultimate limit state was 3.8 for a fifty-year life indicating a probability of failure of the order 104. The reliability indices for serviceability limit states drop down to about 1.5, indicating a probability of failure of the order 102. This order of magnitude may be appropriate for durability design but the various durability limit states of serviceability need to be explored in greater detail, particularly in the context of sustainable development.

Application in practice Clearly much work remains to be done before probabilistic durability design methods are fully enshrined in codes of practice. However it is heartening

36 Fundamentals of durable reinforced concrete to review what is thought to be the first case of the use of the DuraCrete methodology in practice. The example, reported by Breitenbucher et al. (1999), pertains to the design of the concrete lining for the Western Scheldt Tunnel in the Netherlands. The methodology was used to fulfil a requirement that the contractor document the achievement of an anticipated life in excess of 100 years. The tunnel was to be constructed in chloride-contaminated soil. The tunnel segments were to be jointed together with interlocking nibs and recesses. It was recognised that in time the joints may leak, thus exposing the joint surfaces to chloride ingress. The cover to reinforcement at the joints was identified as being critical – a compromise was required between minimising the cover for structural reasons and providing adequate cover for durability reasons over a service life in excess of 100 years. The serviceability state chosen for analysis was that of the onset of corrosion. The acceptable probability of failure was derived by a study of the reliability indices in a Dutch code of practice and in a draft Eurocode. In the absence of better data a requirement was set that the minimum reliability index would be in the range of 1.5 to 1.8, equating to an acceptable failure probability of about 102. Chloride ingress was modelled by a formula based on a solution to Fick’s second law of diffusion. Further detail on the background to this formula is presented in Chapter 6. The form of the equations used was as follows: x(t) 2C(Crit)

√

kt DRCM,0 ke kc

(tt ) 0

n

t

kt DRCM,0 D0

(

C(Crit) erf1 1

CCrit CSN

)

where x concrete cover C(Crit) critical chloride content CSN surface chloride level DRCM,0 chloride migration coefficient measured at time t0 D0 effective chloride diffusion coefficient at time t0 kt, ke, kc constants to take account of method of test, environment and curing on the value of D0 n age exponent erf 1 inverse of error function t0 reference period t exposure period The condition tested was for a cover equal to or in excess of 35 mm. Quality control procedures for achieving the specified cover were anticipated to be good and so a mean value of 37 mm was adopted with a standard


37

deviation of 2 mm, exponentially distributed. The chloride migration coefficient was determined by a rapid chloride migration method to be 4.75 1012 m2/s (mean value) with a standard deviation of 0.71, normally distributed. The critical chloride content was adapted from a literature review, taking account of the anticipated humidity, and was taken to be 0.70 per cent by weight of binder with a standard deviation of 0.10 per cent, normally distributed. The surface chloride level was taken as 4.00 per cent by weight of binder with a standard deviation of 0.50 per cent, normally distributed. Mean values and standard deviations were assigned for the coefficients and age exponent. The value of the reference period, 0.0767 years (28 days) was, of course, deterministic. An analysis of the use of a 37 mm mean cover yielded a reliability index of 1.5 at 100 years and was therefore acceptable.

Summary The first generation of European standards enter practice at a time when key concepts of probabilistic durability design are well understood. However the backup data required to allow inclusion of the design methodologies in codes and standards is not yet available. The need exists for refinement of the deterioration mechanism models and correlation of laboratory and field data. The precise direction that the second generation of standards will take depends on the timing of their publication and on research developments in the interim. These developments will be in the context of a framework embracing the key concepts of service life design and probabilistic analysis. Meanwhile the successful use of service life prediction models, both essentially deterministic and stochastic, in major civil engineering infrastructural projects has been demonstrated.

3

Permeability and transport processes

The durability of concrete is essentially influenced by processes that involve the passage, into or through the material, of ions or molecules in the form of liquids and gases. The service life will be dependent on the rate at which these species may move through the concrete. The passage of these potentially aggressive agencies is primarily influenced by the permeability of the concrete. Permeability may be defined as the ease with which an ion, molecule or fluid may move through the concrete. This definition is somewhat imperfect because the processes involved in fluid and ion migration include the distinct mechanisms of capillary attraction, flow under a pressure gradient and flow under a concentration gradient. These mechanisms are characterised by the material properties of sorptivity, permeability and diffusivity respectively. The term ‘permeability’ has often been popularly used, however, in an allembracing manner to refer to properties that influence ingress. The permeability of concrete to a given agent, for example carbon dioxide, is a function of the pore structure, the degree of interconnection of the pore structure and the moisture content of the permeable pore structure. The diameter of most ions and gas molecules are smaller than the pores in concrete so even the highest quality concrete will be permeable to some extent. The permeability of a concrete will be predominantly influenced by the permeability of the cement paste, especially the quality of paste in the cover concrete and at the interface with aggregate particles. The capillary pore structure is particularly significant. Permeability is a function of the degree of interconnection between the pores, the pore size distribution and its tortuosity. The moisture state is also important and can be beneficial or deleterious. Pores that are water-filled reduce the permeability to gases but may allow ionic diffusion. The significant relationship between permeability and the rate at which durability-threatening mechanisms proceed in concrete is apparent from a brief consideration of the phenomena. The most common problem is corrosion of reinforcement. The rate of corrosion is related to ease of ingress of moisture and oxygen. Transfer of ions through the concrete is also a rate-controlling feature. Corrosion is preceded by depassivation of the reinforcement. This may be caused by carbonation or chloride ingress. Carbonation rates are a


39

function of both physical and chemical phenomena but clearly the ease of ingress of carbon dioxide is a key feature. Depassivation due to chloride ingress is caused by the build up of chlorides to a critical level, which is related to the ease of ingress of chloride ions from external sources. Sulfate attack is caused by the ingress of sulfate ions, typically from groundwater. Deterioration by freeze/thaw behaviour is a function of the number of freeze/thaw cycles and is related to flow of water and its distribution within the pore structure of the concrete. Alkali–silica reaction can occur in many concretes but it only becomes damaging when sufficient moisture can be imbibed from the permeable structure to cause the production of gel in amounts which cause expansion. The transport processes involved in the passage of potentially harmful agencies through concrete are: • • • • •

gaseous diffusion (oxygen, carbon dioxide); vapour diffusion (moisture movement); ionic diffusion (chlorides, sulfates); absorption and capillary rise (chlorides dissolved in water); pressure-induced flow (aggressive groundwater, freeze/thaw).

It should also be noted that in certain mechanisms, for example carbonation and sulfate attack, the rate of ingress is also influenced by an associated chemical reaction. The phenomenon of adsorption should also be noted. It is a phenomenon in which molecules from one material may adhere to the surface of another. Adsorption is relevant to concrete technology, for example in the case of gas molecules adhering to the solid surface of the pore walls in concrete by Van der Waals forces or chemical bonds. This is due to the surface energy: the molecules at the pore walls have excess energy because they are not surrounded. The surface area of a given pore volume increases as the pore radius decreases leading to a consequent increase in water adsorption. Thus the relative humidity of concrete can be higher than the surrounding environment. The relative humidity of the pore structure can be a significant factor in the rate of transport of other species in concrete. In most concretes gaseous and ionic diffusion will be the critical mechanisms. Gaseous diffusion is rapidly accommodated in unsaturated concrete and ionic diffusion occurs in saturated conditions. Thus it is often concretes subject to cyclical environments that deteriorate most rapidly. Examples include the tidal zones of coastal structures and structures subject to wetting and drying. The former allows continual ingress of chlorides by both ionic diffusion and by absorption leading to rapid depassivation of reinforcement. Concurrently there is access for moisture and oxygen, which promotes corrosion following depassivation. Structures subject to wetting and drying will carbonate as the relative humidity falls and, following depassivation, will experience reinforcement corrosion as the relative humidity rises.


Pore structure and the hydration process Permeability is obviously related to the pore structure but it is important to differentiate between porosity and permeability. Porosity is a measure of the proportion of a material represented by voids. Permeability is a measure of the ability of one material to move through another. Although permeable materials always require a pore network there is not necessarily a direct link. Lightweight concrete and air-entrained concrete, for example, may have relatively high porosity but can have low permeability. Poorly cured, high water/ cement ratio concretes will have both high porosity and high permeability. The extent to which a concrete is permeable derives from a number of factors. The most important of these, assuming well-proportioned materials and good compaction, are the water/cement ratio and the degree of hydration. It is the reaction between the unhydrated cement grains and the water that produces gel, which is particulate in nature with a pore structure that predominantly influences the permeability. The aggregate is usually impermeable or is surrounded by a layer of gel that is sufficiently impermeable to isolate the aggregate from the permeable network. Hardened concrete may be considered as a three-phase material consisting of solid matter, water and air. The relative proportions of each phase depend on the age of the concrete and the nature of the environment to which the element is exposed. The greatest change occurs during hydration with some of the water becoming chemically bound and some evaporating. The water-filled spaces found in fresh cement paste become filled, partially or totally, by hydration products. The production of sufficiently impermeable concrete is made possible by the fact that the gel formed during hydration has a greater bulk volume than the parent cement grains. An excellent insight into the process of hydration may be gained from the work of researchers such as Powers (1958). The unhydrated cement grains may be assumed, in a slight simplification, to be formed of silicates, aluminates, and aluminoferrites. The dominant compounds are tricalcium silicates (3CaO.SiO2) and dicalcium silicates (2CaO.SiO2). These silicates are predominantly granular in nature. The silicates, on hydration, produce calcium silicate hydrates (for example 3CaO.2SiO2.3H2O) and calcium hydroxide. The calcium silicate hydrates have varying physical properties but it is now thought that most of them are fibrous in nature with straight edges and lengths up to ten times their width. The calcium hydroxide is more clearly crystalline. The calcium silicate hydrate crystals interlock and form both physical and chemical bonds. It is thought that the physical bonds are more significant in giving concrete its structural properties. The calcium silicate hydrate crystals are so small that the product is regarded as a gel. Crosslinking of fibres leads to a particulate network with interstitial spaces. The clusters of gel particles will have spaces within them known as gel pores. Gel pores, sometimes characterised as ‘micropores’, exist as interlayer spaces between the calcium silicate hydrate sheets. The gel pores occupy about one


41

Figure 3.1 Representation of the pore structure in concrete

third of the gel volume. Larger spaces are formed by the boundaries of the clusters and these are known as the capillary pores (Figure 3.1). Capillary pores may be described as the space originally occupied by the mix water. Hydration of cement grains can continue in the water-filled capillary pores. The resulting gel will be porous but it forms at the expense of the capillary pore volume and therefore the overall effect is one of reduction in pore volume. This is due to the fact that hydrated grains occupy about twice the space of unhydrated grains. Clearly the state of the capillary pore structure at the end of the hydration stage is critically dependent on the water/cement ratio and the quality of curing. Figure 3.2 indicates schematically the influence of water/cement ratio and curing. The graph shows the growth in gel at the expense of water during hydration. It also shows that at higher water/cement ratios the remaining free water creates a large capillary space while poor curing has the same effect by reducing the volume of gel produced. The production of durable concrete therefore requires the water/cement ratio to be: • • •

low enough to limit the initial volume of the capillary pore network produced by the mix water; high enough to provide a water-filled capillary pore network with an initial volume at least twice that of the unhydrated cement; combined with a curing regime which enables the capillary network to remain water-filled long enough to ensure that the hydration process is not stopped through lack of water.

Consideration of the water/cement ratio requirements leads to a requirement in the range 0.4 to 0.5. Regarding a lower limit, Neville (1995) demonstrates that below a water/cement ratio of 0.38 the capillary pore


Figure 3.2 Influence of water/cement ratio and quality of curing on the pore structure of concrete

volume would be insufficient to allow complete hydration of the cement grains. As the water/cement ratio increases above 0.38 the space available becomes progressively greater than that required. However the production of impermeable concrete does not demand that the capillary network should become completely filled by gel during hydration – it is sufficient if the gel partially fills the network in a manner which makes the capillary pore network discontinuous or tortuous. At the upper end of the range, the use of a water/cement ratio in excess of 0.7 is unlikely to produce a pore network of acceptable impermeability for structural purposes. In addition to gel pores and capillary pores, voids caused by entrapped or entrained air will be found in hardened cement paste. The relative size of these pores is represented graphically in Figure 3.3 based on studies by researchers such as Setzer (CEB 1989) and Mehta (1993). Gel pore diameters are of the order of nonometres and range from 0.001 m to 0.008 m. The permeability of such pore diameters to potentially aggressive species is not significant. In theory, all capillary voids could be filled in a concrete with a very low water/cement ratio of 0.38. In practice full hydration is rarely achieved and in any case workability requirements demand higher water/cement ratios. Thus it is inevitable that a capillary network will exist. The challenge is to produce a concrete which has the least capillary network for a given workability. The achievement of capillary pore diameters in the range 0.01 m to 5 m is possible in well-hydrated, low water/cement ratio concretes but values can reach 50 m in lesser quality concretes. Air voids are formed by entrapped or entrained air and pore diameters


43

Figure 3.3 Relative size of the pore types found in concrete

range from 100 m (entrained) to 2 mm (entrapped). The size and distribution of entrained air depends on the use of air-entraining admixtures. Entrapped air should be released during compaction but clearly some entrapped air remains, particularly on inclined shuttered-surfaces. Voids may also be found in aggregates. Aggregate voids contribute to the porosity of hardened concrete. Normal weight aggregates have porosities in the typical range of zero to 20 per cent. Lightweight aggregate voids may total 50 per cent. The aggregate voids are, however, isolated from the permeability network if a satisfactorily cured and compacted cement paste completely surrounds each aggregate particle. Moving from pore structure to permeability, it should be noted that the capillary pore space will have a fluctuating moisture level in service, dependent on the environment, and this will influence permeability and the rate of ingress of aggressive species. The relative amounts of water and air remaining in the pore structure of hardened concrete depend on the temperature and relative humidity of the surrounding atmosphere. If the atmospheric conditions are cyclical so too will be the permeability of the concrete.

Transport processes and rates The principal transport processes in concrete are gaseous diffusion, ionic diffusion, absorption and liquid flow under pressure. Gaseous diffusion Gaseous diffusion is a transport process by which a flow of matter is observed to occur under the influence of a concentration difference. The actual transport mechanism is by random molecular motions with diffusing molecules moving independently of one another. These molecules constantly undergo collisions with the molecules of the material through which they

44 Fundamentals of durable reinforced concrete are diffusing and, although the motion is random, an average fraction of diffusing molecules move past a given section at any one time. A higher number of molecules move from an area of high concentration to an area of low concentration than vice versa. The deterioration models for concrete where gaseous diffusion is involved generally include Fick’s first law of diffusion. This law may be used to describe the rate of diffusion of a gas into a uniformly permeable material: J D

dc dx

where J mass transport rate (g/m2s) D diffusion coefficient (m2/s) dc/dx concentration gradient (g/m4) x distance (m) The moisture content of concrete greatly influences the rate of diffusion. The diffusion rate of gases such as oxygen and carbon dioxide is 104 to 105 times greater in the gas phase than in solution. Ionic diffusion Ionic diffusion requires a concentration difference and saturated conditions. Ions, such as chlorides and sulfates, exploit free water as a passage for ionic diffusion. Ionic diffusion may be represented through Fick’s second law of diffusion and, in certain conditions, by the error function solution of Crank (1975): C 2C D 2 t x Cx x 1 erf C0 2(D t)0.5 where Cx concentration of species at depth x (x in metres) C0 concentration of species at the surface D diffusion coefficient (m2/s) t time of exposure (s) erf error function A common use of the formula is in chloride profile analysis in which instance the concentrations are often expressed as percentage chloride by mass of cement. The error function is available in references such as Crank (1975) or Lawrence (1981). It may also be found in commercial spreadsheet packages. One problem in using spreadsheet packages is that some find difficulty in handling problems with parameters of very diverse magnitudes.


45

The effective diffusion coefficient may typically be of the order 1012 m2/s whereas time, for example, could be 109 seconds (approximately 30 years). This problem may be overcome by first evaluating the coefficient using the value in years and then converting to the required units. Absorption Absorption is a process in which fluid is taken into the spaces within a material. It involves the ingress of a fluid, such as water, by capillary action. It is related to the pore structure of the sample but not to its permeability. Absorption may be significant in concrete subject to freeze/thaw cycles or in situations where potentially harmful agents, such as chlorides, are dissolved in the water which has cyclical access to the member. The transport of liquid by capillary rise is caused by the pressure differential across the meniscus. Capillary rise is characterised by the equation of Washburn (1921) but in concrete technology it is often adequate to determine the ‘sorptivity’ of a sample. Sorptivity may be defined by the following relationship: V S t 0.5 A where V volume of material absorbed in time t (mm3) A cross sectional area of sample in contact with water (mm2) S ‘sorptivity’ (mm/min0.5) t time (min) Liquid flow under pressure Pressure induced flow is often used in test methods to characterise the permeability of a material. Its relevance to concrete in service is limited to such cases as that where the element is under a hydrostatic head, such as groundwater flow. Liquid ingress under a pressure head may be modelled empirically by D’Arcy’s Law and the D’Arcy coefficient may be converted to the intrinsic permeability by a conversion factor. The following relationship results: Q k g h A L where Q flow rate (m3/s) A cross-sectional area (m2) k intrinsic permeability (m2)

density of fluid (kg/m3) g acceleration due to gravity (m/s2) viscosity of fluid (Ns/m2) h head loss across the sample (m) L thickness of sample (m)


Figure 3.4 Initial Surface Absorption Test

The intrinsic permeability is a property of the material independent of the fluid concerned. In the case of concrete the coefficient of permeability based on water flow (at room temperature) is often quoted as a measure of its permeability: Q h K A L K coefficient of water permeability according to D’Arcy (m/s)

Measurement of permeation properties Many test methods are available some of which are a direct measure of permeability, others detect surface absorption or porosity but may be used to rank the concrete in relative terms. A comprehensive report on site testing for permeability was published by the Concrete Society (1987). Four test methods in particular are commonly used. These may be categorised as follows: • • • •

Surface Absorption Test, Water Absorption Test, Sorptivity Test, Pressure Permeability Tests.


47

Figure 3.5 Water Absorption Test

Surface Absorption Test Surface absorption is commonly measured by the Initial Surface Absorption Test (ISAT) but can also be determined by use of the ‘Autoclam’ device, which is primarily an apparatus for pressure permeability tests. The ISAT has the advantage of being covered by a British Standard (BS 1881: Part 208). It does not measure permeability directly but it may be used for comparative purposes. The test is limited to providing information on the characteristics of the surface of the concrete only. The method involves sealing a cap onto the concrete surface and monitoring the rate of seepage of water into the surface and near-surface pores (Figure 3.4). Clearly the moisture content of the surface has an influence on the results. On-site testing must therefore take account of current and recent weather patterns that would significantly influence the surface condition. Water Absorption Test The Water Absorption Test is also covered by a British Standard (BS 1881: Part 122). It does not measure permeability directly but detects the absorption characteristics of a core recovered from a concrete member. It therefore gives more than just surface absorption values. The method involves immersing oven-dried cores in water to detect the uptake of water in a specified time (Figure 3.5). Sorptivity Test A simple sorptivity test may be used to study absorption by capillary action. The test involves placing one surface of a concrete sample just in contact with water and monitoring the amount of water absorbed over time (Figure 3.6). It has been found that the amount of water absorbed is proportional to the square root of time and the sorptivity may therefore be computed.


Figure 3.6 Sorption Test

Figure 3.7 ‘Figg-type’ Permeability Test

Pressure permeability tests Commercially available kits have been developed to monitor the air and water permeability of concrete through flow under a pressure difference. Examples include the ‘Figg-type’ tests (Figure 3.7) and the ‘Autoclam’ test (Figure 3.8).


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Figure 3.8 ‘Autoclam’ Permeability Test

Both methods detect the quality of concrete in the cover zone. The ‘Figg’ method involves drilling a hole in the cover concrete and sealing the top of it. The device achieves a pressure differential between the hole and the surroundings, forcing a flow (of air or water), the rate of which can be monitored by the pressure change over time. The ‘Autoclam’ Permeability Test was developed by the Queen’s University of Belfast for use both in the laboratory and on site to measure air and water permeability. A base ring is bonded to the concrete surface to isolate the test area. The main body of the apparatus is then screwed to the ring. A cylinder and piston are used to pressurise the test area. The change in pressure over time is monitored. The apparatus can also be used without pressure to measure surface absorption.

Factors influencing the permeability of site concrete Durability is intrinsically linked to impermeability. The capillary pores, initially filled with water, accommodate the hydration products of the chemical reaction. These products occupy double the space of the original solid constituents being hydrated. Clearly therefore water/cement ratio is a critical parameter in the production of low permeability concrete. In theory, all capillary voids could be filled with a concrete with a very low water/cement ratio. In practice full hydration is rarely achieved and in any case workability requirements demand realistic water/cement ratios. Thus it is inevitable that a capillary network will exist. The challenge is to produce a concrete which has the least capillary network for a given workability. This cannot be achieved by specification alone. Site practice must also be considered. The readily-controllable aspects which can have the greatest influence on permeability are mix design and specification, compaction, and curing.


Figure 3.9 Influence of water/cement ratio on permeability

Concrete mix design should aim to produce the lowest possible water demand consistent with durability and workability requirements. Dewar (1985) has shown that consideration of mean particle size, voids ratio diagrams, and particle interference effects can be used in mix design to determine the minimum water demand required. Software is now available to assist in this process. The water/cement ratio should be as low as possible consistent with workability requirements. This will usually entail the use of a water-reducing admixture. The significance of the water/cement ratio is illustrated by the trend shown in Figure 3.9. It may be noted that exceeding a ratio of 0.6 can lead to an exponential rise in permeability. Proper compaction is essential in the production of impermeable concrete. Entrapped air bubbles may have relatively large diameters and would therefore provide linkage between otherwise discrete capillary pores. Poor curing is probably the greatest single cause of permeable concrete. The quality of the cover zone is as important as its depth in terms of the defence mechanism against durability threats. The quality of the cover zone is critically influenced by the curing regime. Inadequate curing prohibits the required blocking of capillary pores by the products of hydration. Thus the production of durable concrete involves the selection of suitable materials; a mix design with low water/cement ratio which reduces the interparticulate spaces to a minimum yet produces a workable concrete; and the employment of proper site practice in respect of compaction and curing to ensure that entrapped air is expelled and that the highest possible degree of hydration is achieved.

4

Corrosion of reinforcement in concrete

Nature of corrosion damage Corrosion of reinforcement is the greatest cause of durability failure in the world today. This situation came about because designers and specifiers of works constructed in earlier decades were reliant on codes that did not fully take account of the phenomenon. The drafting of codes involves reliance on experience and there was insufficient evidence and understanding of the corrosion phenomenon in concrete to better inform the drafters. The emerging generation of standards will improve matters by forcing us to explicitly consider the mechanism of failure. It is important therefore to appreciate some fundamental aspects of the corrosion phenomenon. Corrosion is an electrochemical process whereby a metal undergoes a reaction with chemical species in the environment to form a compound. The chemical species are principally oxygen and water. Steel reinforcement has a natural tendency to corrode if access to oxygen is possible in a moist environment. This is because it is formed of metals found naturally occurring as ores to which they wish to revert. The durability of reinforced concrete requires conditions in which the dissolution of metal atoms is not supported and that the reinforcement be inaccessible to oxygen and moisture. Two selfdefence mechanisms are employed to achieve this. The first involves a naturally occurring protective film on the reinforcement, which requires certain conditions for its survival. The second involves cover concrete of sufficient depth and impermeability. Regarding the first issue, the high pH level of fresh concrete leads to the formation of a passive skin on the surface of reinforcement. This skin prevents corrosion occurrence by preventing contact with oxygen and moisture. The passive film may be broken down in time through carbonation or chloride ingress reaching the steel. Regarding cover, the rate of corrosion depends on the rate at which oxygen and moisture may penetrate the cover. This has a two-fold influence. First, oxygen and moisture are required to feed the process. Second, the concrete must be moist enough to have an electrical resistance that is low enough to allow the creation of an electrochemical cell. The damage to the structure is manifested in a number of ways: the

52 Fundamentals of durable reinforced concrete cross-sectional area of reinforcement available for load-carrying is locally reduced; the surface of the structure may exhibit cracking and spalling; rust staining on the surface may become apparent even in the absence of cracking. In the case of ferrous reinforcement the compounds formed through corrosion are hydrated iron oxides, for example ferrous hydroxide Fe(OH)2, ferric hydroxide Fe(OH)3, and a secondary reaction forms rust. The initial reaction is generally the formation of ferrous hydroxide and the secondary reactions produce a form of rust dependent on the environmental conditions, examples include FeOOH, Fe2O3.H2O, Fe2O3, or Fe2O3. Schematic descriptions are as follows: 2Fe(OH)2 O2 → FeOOH and 2FeOOH → Fe2O3 H2O Also 4Fe(OH)2 O2 2H2O → 4Fe(OH)3 and 4Fe(OH)3 → 2Fe2O3.H2O 4H2O The resulting oxide has a lower density than the parent metal and so a volume increase occurs when a mass of metal is replaced by the new compound. This leads to particular problems in reinforced concrete structures. Corroding reinforcement in concrete not only loses valuable cross-sectional area but the concrete may also have to accommodate the increased volume of the iron oxides produced. The resulting rust may occupy a volume between two and five times that of the parent steel. If this cannot be accommodated expansive pressures result. The low tensile strength of concrete may be exceeded by the pressure caused through build-up from even a microscopically thin layer of corroding reinforcement. Cracking may result and further development to the stage of delamination and spalling is not uncommon where depth of cover is low. The various stages of this phenomenon are illustrated in Figure 4.1 and an example of corroded reinforcement is illustrated in Figure 4.2. The precise form of damage to the concrete depends on the bar diameter, bar spacing, and the depth of cover. The first sign of distress could be popouts or long thin cracks along the line of the reinforcement. A form of attack described as ‘pitting corrosion’ may also occur, as illustrated in Figure 4.3. A pit develops on the reinforcement surface. This form is more insidious than spalling because it can lead to loss of cross-


53

Figure 4.1 Stages in corrosion-induced damage

section without evidence of distress being apparent on the surface. The loss of effective cross-section is very significant compared to that occurring where there is general corrosion on the complete perimeter. This form of deterioration can occur where small anodic sites are in combination with a large cathodic area. Such conditions can occur where moisture conditions vary across a structure, due to drainage conditions or leaking joints, leading to localised anodic sites. This mechanism of attack is usually related to the presence of chlorides and is dependent on the relative amounts of chloride and hydroxide ions. If the chloride ions predominate the loss of Fe2 ions is accelerated and a pit develops. If the OH ions predominate one gets precipitation of FeOH which repairs the passive oxide film on the reinforcement. The loss of cross-section in pitting corrosion can occur rapidly and critically reduce the load-bearing capacity of the reinforced concrete member. Unlike carbonation-induced corrosion, where cracking and spalling can serve as an early indicator of a problem, chloride-induced pitting corrosion may be quite advanced before evidence is apparent. Deterioration in these cases


Figure 4.2 Example of corroded reinforcement

may be so advanced at the time of discovery that the structure may be beyond economic repair. An interesting aspect of the corrosion cell in reinforced concrete is the fact that damaged areas, for example zones of spalling, may only represent the anodic sites. Patch repair of concrete around a spalled area may not be effective if the cathodic sites remain permeable – in time the patches will spall since the cell is allowed to continue functioning. It may be necessary to enhance the concrete’s impermeability, for example by coating, to ensure an effective repair.

Electrochemical process Measures to control corrosion of reinforcement require an understanding of the processes involved. Corrosion is an electrochemical process and therefore


55

Figure 4.3 Pitting corrosion

a basic understanding of electrochemistry and its application to the particular case of reinforced concrete is required. Oxidation and reduction Corrosion is related to the flow of electrons. It is necessary therefore to recognise that elements may gain or lose electrons. An element or compound that loses electrons is said to be ‘oxidised’ and one that gains electrons is said to be ‘reduced’. The elements and compounds of interest in the case of reinforced concrete are iron, water, and oxygen since the metal undergoes a reaction with chemical species in the environment, principally oxygen and water, to form a compound. When metal oxides are formed the metal atoms lose outer electrons. The metal atoms are oxidised and the oxygen is reduced. The following examples are presented in a form whereby the number of electrons is equal in each case as this is a significant controlling factor in the consideration of the rate of corrosion. Oxidation example: Fe → Fe2 2e cation

Reduction example: H2O 1⁄2O2 2e → 2OH anion

The electrochemical cell Oxidation and reduction processes will be detectable if conductors of electricity are immersed in a solution capable of conducting electric current. A classic example of a cell in which a flow of electrons is readily detectable is


Figure 4.4 Basic form of an electrochemical cell

that shown in Figure 4.4. This illustrates two electrodes connected together and immersed in an electrolytic solution. The cell may be easily constructed in a laboratory and is familiar as a form of battery. The positive electrode is the anode while the other is the cathode. An electrochemical cell may also occur with a so-called ‘mixed electrode’. This describes a condition in which the anode and cathode are on the same material. The occurrence of a mixed electrode may occur in reinforced concrete if a number of conditions apply: • • •

the passive film on the reinforcement must be destroyed locally; the concrete must be moist enough to act as an electrolyte; the cover must be permeable to oxygen.

The consequent scenario is illustrated in Figure 4.5. This shows the formation of anodic and cathodic sites on a reinforcing bar. Metal oxidation occurs at the anode. Electrons are freed and these can flow through the reinforcement to the cathode. The electrons are consumed in reduction processes leading to the formation of hydroxyl ions. The permeability of the cover concrete allows moisture to ingress from the environment and a sufficiently saturated state develops at which the moist concrete may function as an electrolyte. Furthermore the permeability of the cover, perhaps allied to its insufficient depth, can allow the penetration of carbon dioxide or chlorides to an extent that local depassivation of the reinforcement will occur. Oxygen could then penetrate the cover to feed the cathodic reaction. The concrete is


57

Figure 4.5 Mixed electrode in reinforced concrete

moist enough to serve as an electrolyte and a build up of rust products becomes apparent on the reinforcement. The example illustrated in Figure 4.5 shows microcell corrosion with anode and cathode in proximity on the same bar. The cathodic area may equally be widespread along the reinforcement layer. Macrocell corrosion is also possible in reinforced concrete members. For example one layer of reinforcement may become the anode and another the cathode. Equally one part of the structure may be depassivated and become the anode while a much larger area may remain passive and become the cathode. Such high cathode to anode area ratios can lead to very significant increases in corrosion current. Corrosion rate Individually the anodic and cathodic processes would lead to an accumulation of positive and negative charges respectively on the reinforcement. This is not sustained because the hydroxyl ions diffuse towards the anode where they meet the counter-diffusion of ferrous ions. The resulting combination will cause electrical neutralisation if the anodic and cathodic processes are coupled together in the form of a corrosion cell with no excess electrons. If there is no external source of electrons, the electrons produced by oxidation will be fully consumed by reduction. Thus the oxidation rate and reduction rate must be equal. Hence the anodic reaction rate must equal the cathodic reaction rate. This equality controls the rate of corrosion. Thus the rate of electron flow reflects the rate of corrosion. The rate of flow may be referred to as the ‘corrosion current density’. Essentially it is the number of electrons flowing per unit area and is represented graphically in Figure 4.6. A value at the upper end of the range would be 100 mA/m2. Of greater interest is the rate at which material is being lost from the reinforcement. A relationship may be established through Faraday’s Laws


Figure 4.6 Corrosion current density

between the measured current flow and the annual material loss since the mass of substance liberated is proportional to the quantity of electric charge that liberates it. Faraday’s First Law of Electrolysis states that the mass of substance liberated is proportional to the quantity of electric charge that liberates it. Faraday’s Second Law of Electrolysis concerns the masses of different substances liberated by the same quantity of electrical charge: the masses are proportional to the ratio of the atomic mass and the valence. Thus the loss may be calculated from: m

MC zF

where m mass loss M atomic mass C electric charge z valence F Faraday constant (96487 coulomb) This ratio of atomic mass to valence is called the chemical equivalent. In the case of iron (Fe) and the ferrous ion (Fe2) the atomic mass is 55.85 amu and the valence is two. Therefore the chemical equivalent is 27.95 and oxidation of 27.95 g of Fe occurs per Faraday of electrical charge. Given that a coulomb is the electric charge conveyed in one second by a current of one ampere, it is interesting to review the effect of a corrosion current density of one milliamp per square metre. This equates to 0.001 coulombs/m2/s therefore the rate of loss would be: (55.85) (0.001) (1) 2.89 107 g/m2/s (2) (96478) This would equate to a loss of about 10 g/m2/year. A corrosion current density at the upper end of the range, such as 100 mA/m2, could theoretically lead to an annual loss of approximately one kilogramme per metre squared of metal. Weight losses higher than those predicted by Faraday’s Laws have been detected in practice. Rodriguez et al. (1996) reported that the loss in bar cross-section can be


59

estimated through Faraday’s laws from the measurement of corrosion rate using the linear polarization technique: 0 x x 0.115 Icorr t where residual bar diameter (mm) 0 nominal bar diameter (mm) coefficient depending on type of corrosion x attack penetration (mm) Icorr corrosion rate (A/cm2) t time from start of corrosion (years) The coefficient has a value of two where the corrosion is homogenous. The value rises to a figure in the range four to eight if pitting corrosion occurs. Electrode potential The corrosion phenomenon depends on the ability of electrons to transfer across the interface between the metal and the electrolyte and vice versa. This parameter is described as the ‘electrode potential’ of a particular metal in a particular electrolyte. Thus in reinforced concrete the process is driven by the potential difference between the reinforcement and the cover concrete. The potential difference develops at the interface of the reinforcement and the concrete because of the tendency of the metal ions to dissolve and the difference in environment (metal and concrete) on either side of the surface. An excess positive charge builds up near the interface. The potential cannot be measured directly and so it is measured with respect to a reference electrode. Electrical potentials may be surveyed in existing structures by devices such as the ‘half cell’, described later. Reference electrodes include, for example, saturated hydrogen electrode; saturated calomel electrode (SCE) and silver/silver chloride electrode (SSC). Where potentials are quantified it is necessary to indicate the reference electrode type. Corrosion of steel in concrete typically occurs in the potential range 450 mV SCE to 600 mV SCE. Electrical resistance Flow between anodic and cathodic sites during corrosion depends on the electrical resistance. The resistance is dependent on the moisture state of the concrete. Only oven-dry concrete has a high enough resistance to prevent conduction. Saturated concrete has a resistivity in the range of 10 to 60 m and moist concrete can be up to 100 m (Neville 1995). The resistivity increases one hundred fold as the relative humidity falls from saturation to


Figure 4.7 Polarisation curve showing the reversible potential (Erev)

50 per cent (Lawrence 1990). Resistivity low enough to cause problems is of the order of 50 to 100 m, which is typical of concrete in moist environments. The current is conducted by ions in the pore water or, if necessary, through the gel water. Concrete absorbs water rapidly but can be slow to dry out. Thus moist conditions can persist even when rainfall is sporadic. The presence of chloride ions in the pore water would also decrease the resistivity.

Polarisation curves, the ‘Evans Diagram’ An insight into the durability of reinforced concrete through the control of corrosion rate may be gained by consideration of ‘polarisation curves’ and ‘Evans Diagrams’ as demonstrated, for example, by Bentur et al. (1997). These curves graph the relationship between electrical potential and current. A polarisation curve may be generated by experimenting with two electrodes. A potential difference is applied between the two electrodes and the current flow is monitored. If the potential difference is that between the anode and the cathode, the cathode can be electrically positive with respect to the anode. The favouring of either anodic or cathodic reactions will be detectable at different potentials but there exists a potential at which the reactions are balanced – this is described as the ‘reversible potential’. The more polarised the potential, that is the further the potential is from the reversible potential, the greater will be the current (Figure 4.7). In the case of corroding reinforcement it is of great interest to consider the potentials at the anode and cathode. The anode will initially have a certain potential difference with the electrolyte and so too will the cathode. A common electrode potential is generated between the steel and concrete which


61

Figure 4.8 Evans Diagram showing the potential (Ecorr) adopted by a corroding metal

is intermediate between the individual anodic and cathodic systems. The corrosion process occurs at the potential where the rates of anodic and cathodic reaction are equal. It is very informative therefore to plot the relationship of potential and current in a manner that shows where equilibrium of current occurs. This is achieved by plotting curves on a common axis, that is ignoring the sign for the current. Such a plot is known as an Evans Diagram. The potential that the metal adopts as it corrodes is determined by the point of intersection of the curves (Figure 4.8). It is the value of electrode potential detectable when the metal is corroding freely. The corresponding current may also be assessed. This potential (Ecorr) and the corresponding corrosion current (icorr) may be determined by considering the kinetics of the anodic and cathodic reactions in a given situation. Design of durable structures involves a rudimentary appreciation of the way a corrosion cell behaves. A number of consequences of the electrochemical nature of reinforcement corrosion are more readily understood by reference to an appropriate Evans Diagram.

Passivity The pore liquid phase in concrete is mainly a solution of sodium hydroxide (NaOH) and potassium hydroxide (KOH) due to the dominant alkalis in the

62 Fundamentals of durable reinforced concrete cement. These are sodium oxide (Na2O) and potassium oxide (K2O). This yields a solution of high alkalinity. The high pH level of fresh concrete protects a passive skin formed on the surface of the reinforcement. This skin, of nanometre order of thickness, prevents corrosion occurrence by preventing further oxidation of the metal atoms. Reinforcement in fresh concrete initially begins to corrode and the corrosion product forms the passive skin on the bar. The skin prevents the dispersal of ferrous ions and is a conductor of electricity. It is therefore unable to support an electrical potential difference and the force driving the metal ions into solution is removed (Lawrence 1990). The film is formed through a reaction between iron and water. This reaction yields a twin-skinned film consisting of oxide (Fe2O3 or Fe3O4) with an outer hydrous oxide layer. The skin is stable at pH levels above 11.5. When anodic sites form on reinforcement the reaction is that of oxidation of the metal ions to yield ferrous ions and the anodic current rises. However a point is reached where passivation occurs and the current reduces significantly. The Evans Diagram for a reinforcing bar in a passive environment is illustrated in Figure 4.9. It may be noted that the intersection of the curves usually occurs at a point where the electrode potential is very high and the corrosion current very low. The corrosion current would be low enough to ensure an adequate service life and the steel is considered to be in a passive state. Curiously depassivation and therefore corrosion can occur where a very low rate of oxygen diffusion is encountered, such as in submerged structures. In such a case the cathodic curve moves back to intersect the anodic curve at a low potential. However the rate of corrosion is not significant in terms of the overall service life of structures.

Corrosion mechanism in carbonated concrete The phenomenon of carbonation is fully described in Chapter 5. The carbonation front, which delineates areas of differing pH, advances from the surface towards the passive reinforcement at a rate dependent on environmental factors and material properties of the concrete. If peaks on the carbonation front reach the reinforcement the passive film is locally destroyed, as it is unstable at low pH values. If the passive film breaks down iron oxidises to form ferrous ions (Fe2) with the following half-cell reaction: 2Fe → 2Fe2 4e The ferrous ions pass into solution. The electrons flow through the reinforcement to the cathode where they are adsorbed by the electrolyte. The cathodic reaction produces hydroxyl ions (OH), in the presence of oxygen and moisture, as follows: 2H2O O2 4e → 4OH


63

Figure 4.9 Evans Diagram for a reinforcing bar in a passive environment

The hydroxyl ions move from the cathodic site through the moist concrete, the electrolyte, towards the anodic site where ferrous hydroxide (Fe(OH)2) is formed. The anodic reactions may be represented as follows: 2Fe2 4OH → 2Fe(OH)2 The ferrous hydroxide is unstable in the presence of oxygen and so the reaction may proceed to form rust. The reactions can be represented by the following: 2Fe(OH)2 O2 → FeOOH 2Fe(OH)2 H2O 1⁄2O2 → 2Fe(OH)3 Further advances of the carbonation front will increase the depassivated area. Widespread corrosion may then follow with the development of cracking along the lines of the reinforcement.

Corrosion mechanism in chloride-rich concrete The issue of chloride ingress is dealt with more fully in Chapter 6. The anodic reaction is of particular interest in the case of concrete subject to high

64 Fundamentals of durable reinforced concrete chloride levels. As the chloride concentration increases the anodic curve, vital to passivity, alters as illustrated in Figure 4.10. It may be seen that the effect is to make the potential more negative with a consequent effect on the corrosion current. The process first involves the oxidation of iron to ferrous ions (Fe2): 2Fe → 2Fe2 4e The cation combines with chloride ions to form chloride or oxychloride compounds (FeCl2 and FeOCl), for example: 2Fe2 4Cl → 2FeCl2 The process then becomes self-propagating, due to the acidic conditions created and the recycling of chloride ions. This occurs through the hydrolysing of the chloride compounds, for example: 2FeCl2 4H2O → 2Fe(OH)2 4HCl Alternatively 2FeCl2 4H2O → 2Fe(OH)2 4H 4Cl and 2FeOCl 2H2O → 2Fe(OH)2 2Cl There follows a consequent recycling of the liberated chloride ions. Although corrosion product is being produced so too is hydrogen chloride and hydrogen (H) or hydronium ions (H3O). The increased acidity of the anodic area helps to prevent precipitation of corrosion product. The chloride and oxychloride compounds are therefore more stable and are free to migrate further. The increased acidity also encourages further oxidation of the iron. The rate of corrosion is influenced by the ability of the cement matrix to bind the liberated chloride, the resistivity of the electrolyte and the availability of oxygen and moisture at the cathode. The binding ability is dependent on the characteristics of the cement. Sulfate-resisting cement, for example, has a lower binding capacity than, say, ordinary Portland cement. The issue of oxygen and moisture at the cathode is not straightforward because even low rates of oxygen supply may lead to severe pitting corrosion. This effect occurs because the anodic sites may be localised but the corresponding cathodic sites may be spread out over a wide area. The cumulative effect of even low rates of oxygen supply to large cathodes may be significant. The fact that the corrosion product is discouraged from precipitation due


65

Figure 4.10 Evans Diagram showing the influence of chloride concentration in concrete

to the acidic conditions, and the existence of highly active and localised anodic sites in combination with large area cathodic sites, accounts for the phenomenon of severe pitting without early warning through visible signs of deterioration at the concrete surface. When cracks do develop the corrosion product will then be deposited along the crack where the effect of the acidic zone will become less marked with distance from the reinforcement. By the time rust staining becomes apparent at the surface the extent of reinforcement deterioration may be structurally significant.

Influences on corrosion activity Corrosion activity is influenced by local environmental factors including oxygen supply, relative humidity and temperature. Influence of oxygen supply The passive film may be broken down through carbonation or chloride ingress. The rate of corrosion will then depend on the rate at which oxygen may penetrate the cover. This is illustrated in Figure 4.11. The oxygen supply to the reinforcement is a function of the permeability of the cover. Good


Figure 4.11 Evans Diagram showing the influence of oxygen supply in concrete

quality concrete with low permeability will restrict oxygen supply. Equally, a condition in which the concrete is saturated may not lead to corrosion even in permeable concretes. Submerged parts of reinforced concrete structures in the ocean are not subject to the same risk of corrosion damage as those parts above water despite the ingress of significant amounts of chloride. This is due to the restriction of oxygen ingress because gaseous diffusion is very slow through the saturated pores and there may be little dissolved oxygen in the water. Nevertheless the problems associated with large cathodic areas cannot be ignored. It may be noted from Figure 4.11 that the cathodic reaction curve becomes steeper as the oxygen supply diminishes leading to a reduced corrosion current as expected. Relative humidity A corrosion cell cannot occur if the concrete is too dry to serve as an electrolyte or too wet to allow ingress of oxygen. Sufficiently dry conditions are typical inside buildings and corrosion does not occur despite the high probability of reinforcement depassivation through carbonation. Corrosion activity is most vigorous at relative humidity values above 80 per cent as illustrated in Figure 4.12 using data from Andrade et al. (1986) and Parrott (1994). These levels are rarely found indoors. Air conditioned buildings are designed


67

Figure 4.12 Influence of relative humidity on corrosion rate

for an atmosphere of typically 50 per cent relative humidity which, although ideal for promoting conditions conducive to carbonation, should not support the development of corrosion activity. Equally corrosion activity reduces as the relative humidity approaches saturation since the effect of permeability reduction would become dominant. Temperature The rate of corrosion increases with increasing temperature. The effect is not significant at typical European humidity levels but it could be in hot humid climates. Browne (1988) reported that an increase in temperature from 20°C to 40°C could increase the rate of corrosion by a factor of five.

Influence of cracking The influence of cracking on corrosion rate is discussed more fully in Chapter 11. It has been a subject of debate in the context of acceptable limits for crack widths. The jury is still out on the issue. The Building Research Establishment (1993) and Concrete Society (1995) have indicated however that live cracks, parallel to the reinforcement, represent a much higher corrosion risk than dormant perpendicular cracks.


Figure 4.13 Corrosion activity in cracked concrete: significance of uncracked zone over the cathode

Various issues must be weighed up. On the one hand there is the seemingly self-evident point that cracks which extend to the reinforcement provide easier access for carbon dioxide and chlorides leading to depassivation, and to oxygen and moisture thus encouraging the corrosion process. On the other hand, the observed rate of deterioration of cracked structures is not universally higher than non-cracked structures. Thus the various influencing factors need to be considered in isolation. The first issue relates to whether or not the cracking occurs in the initiation or propagation phase of corrosion. If cracking results from the build up of corrosion product it is clearly limited to the propagation phase. Service life is the sum of time to initiation of corrosion plus the time required for corrosion propagation to cause an unacceptable level of deterioration. The initiation phase will usually be relied upon to form the greater part of the service life. If the cracks are caused by factors other than corroding reinforcement they could influence the initiation phase to some extent. On the other hand cracks caused by the expansive forces generated by corrosion product obviously occur after depassivation and therefore will have minimal effect on the overall service life if the initiation phase is of adequate duration. Despite this there has been understandable confusion in the past about the link between cracking and corrosion rate. The second issue is that of orientation of the cracks. If the cracks are parallel to the reinforcement and over it they will pose a more significant problem than if they are perpendicular, due to the area of reinforcement potentially exposed to harmful agencies. Fortunately cracks parallel to, and over, reinforcement are more commonly associated with the effects of corrosion propagation rather than its initiation and should therefore be associated with the shorter phase of the service life. One notable exception is the case of plastic settlement cracks.


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A third issue is the dynamic state of the cracks. Cracks are capable of selfhealing in certain circumstances – autogenous healing. This effect can be caused by the precipitation of calcium carbonate in the crack leading to clogging of the crack and thereby preventing further ingress of harmful species. Live cracks may also become clogged by debris but are less likely to autogenously heal. A fourth issue involves the density of cracking. Individual cracks that intersect the reinforcement locally may encourage anodic activity but access to the cathodic areas will not be different to that pertaining in crack-free surfaces (Figure 4.13). Therefore the rate of corrosion should not be different, as it will continue to be influenced by the accessibility of the cathodic sites. A final issue is that of crack width. Other than noting that autogenous healing is more likely the narrower the crack, it does not appear that crack width itself is a significant factor. A relationship between crack width and rate of corrosion during the propagation phase is not established.

Modelling the rate of corrosion Early proponents of the design life approach to durability management assumed that only the initiation phase should represent the design service life. This conservative assumption was based, in part, on the difficulty of accurately predicting that part of the propagation phase during which the structure remained serviceable. This difficulty may be overcome in time by further research of corrosion damage mechanisms in which case it will be possible to design for corrosion propagation with the same degree of uncertainty as initiation mechanisms. It may seem prudent to continue with the approach of considering service life as being represented by the initiation period only, given the uncertainties inherent in durability modelling. However the influence of relative humidity on carbonation and corrosion rates, for example, may demand that more attention be devoted to design of the propagation phase. At certain relative humidity levels the propagation phase can be longer than the initiation phase, as illustrated in Figure 4.14. A model for corrosion rate based on electrochemical activity has been developed by Lawrence (1990) but as a starting point for propagation period modelling researchers have simply concentrated on the cause and effect of damage to the structure. The detectable effect, which may be used to define the limit of acceptable damage, could be cracking. For example, it could be the first cracking detectable through magnification; the first visible cracks to the naked eye; the first evidence of spalling; or deflection over and above that anticipated when the reinforcement is in good condition. The cause would be reinforcement corrosion defined as the depth of reinforcement corrosion that leads to the limiting effect. Thus: tp f (CD, CR, )


Figure 4.14 Distribution of initiation and propagation phases at different values of relative humidity

where tp propagation period CD corrosion depth which causes the limit of acceptable damage CR corrosion rate reinforcement diameter Parrott (1994) has illustrated such an approach in the case of carbonationinduced corrosion. He used the following relationship to demonstrate a prototype model, noting that some parameters had to be left out in the absence of more information: tp

CD CR

In the example he took the propagation period to be the time to appearance of the first visible crack. He determined illustrative examples of the input parameters derived from many published results in the literature. He found that the corrosion depth which causes visible damage is about 100 m. The corrosion rate varies from about 0.3 m/year at 50 per cent relative humidity to a maximum of about 50 m/year at 98 per cent relative humidity. He noted that the model could, in time, be refined to take account of influences such as the diameter and geometry of the reinforcement cage.


71

The modelling of corrosion rate has yet to be fully integrated into the first generation of design life models. Durability design involving corrosion initiation models only are emerging as the more common approach. Their use as optimisation tools may, however, benefit from the combination of initiation and propagation period models. The disadvantage of including the corrosion propagation period in determining the service life is that the uncertainty of the models is very different from that associated with initiation period models. Further research is warranted to reduce the uncertainty. An alternative approach in the interim would be to assume a propagation period of zero and therefore use it as an additional safety margin. This could be a good approach while experience is built up of the reliability of the first generation of deterioration models in service.

Monitoring corrosion activity Monitoring corrosion activity in concrete, especially in the field, can be achieved by detecting two key aspects of electrical activity – the potential and the resistance. The location of corrosion cells may be determined by electrical potential contour plots taken on concrete surfaces. A connection to the reinforcement must be made and the technique relies on electrical continuity through the reinforcement. Measurement of electrical potential alone will indicate the location of corrosion but does not indicate the severity. Resistivity measurement in conjunction with potential mapping may better assist the investigator. High rates of corrosion are associated with large potential gradients and low concrete resistivities. In theory, knowledge of potential and resistivity would allow calculation of the current flow and therefore the rate of corrosion. In practice it is difficult to get sufficiently accurate values but estimates are informative. Potential (half-cell) mapping The current flow between anodes and cathodes on the reinforcement leads to variation in potential that is detectable at the concrete surface. A map of equipotential lines on the surface can be used to detect active corrosion cells. The principle is illustrated in Figure 4.15. The principle is based on bringing two half cells together. One half cell is the reinforcement in the concrete, the other half comprises part of the test equipment. These are connected together to form a full cell. The cell potential is measured. The part of the potential contributed by the test equipment is constant and so variations detected across the circuit are attributable to the reinforcement/concrete half cell. This permits identification of areas where corrosion activity is highest. The test apparatus consists of a circuit formed by an electrical connection to the reinforcement, a high impedance voltmeter, a reference electrode, an ionic bridge and a sufficiently moist concrete surface (Figure 4.16). The ionic bridge is required between the reference electrode and


Figure 4.15 Principle underlying the use of potential mapping on concrete surfaces to detect corrosion activity

the concrete surface. It could consist of a sponge soaked with an electrolytic solution such as water or a weak sodium hydroxide solution. Commercial devices are available in which the bridge is incorporated. If a circuit cannot be formed, due to a lack of continuity in the reinforcement, it may be possible to conduct a survey by using two electrodes on the surface – one fixed and one moving – connected through a voltmeter. The approach to investigation generally involves a cover survey to mark out the reinforcement location. A survey grid is then established at about onemetre intervals, the reference electrode is placed over the grid points, and the potential is noted. The grid is progressively reduced to a spacing of about 200 to 500 millimetre intervals where negative potentials of interest are detected. A commercially available device is available which can be wheeled across surfaces to produce a continuous plot of potentials. Results are stored in a data logger and a dot matrix printer can generate variable density contour plots. The potential can only be measured with respect to a reference electrode such as copper/copper sulfate (CSE) and silver/silver chloride (SSC). It is necessary to indicate the reference electrode type where potentials are quantified. For example a potential of 130 mV (SSC) corresponds to zero on the CSE scale. Value ranges have been reported which have been calibrated with field observations to give an indication of the relationship between electrical


73

Figure 4.16 Test set-up for ‘half-cell’ potential mapping

potential and corrosion risk. The findings of Van Daveer (1975) are widely used but other value ranges have been reported. A general summary is that electrical potentials more negative than 350 mV (CSE) indicate a very high risk of corrosion and that electrical potentials more positive than 200 mV (CSE) indicate a very low risk of corrosion. Judgement must be made on the extent of risk if intermediate values are encountered. It is important to note that the equipment can only identify active corrosion sites. The absolute values of electrical potential can be affected by carbonation, oxygen starvation, chloride content, and temperature. Therefore its primary strength is as a device for comparative studies of corrosion activity on similar members of a particular structure. Electrical resistivity measurement The measurement of resistivity basically involves passing a current through the concrete at a known voltage. A device commonly used is the Wenner four point apparatus and the basis of the technique is illustrated in Figure 4.17. An alternating current is passed between two outer probes and the voltage between two inner probes is measured. The resistivity (R) can be determined from the following relationship:


Figure 4.17 Test set-up for electrical resistivity measurement

R2a

V I

where a probe spacing V voltage between inner probes I current between outer probes The probe spacing must be such that the resistivity of the cover concrete is being determined. If the spacing is too wide the electrical field will be interfered with by the reinforcement; if it is too narrow it will be overly influenced by surface effects. The resistivity readings are significantly influenced by the moisture state of the concrete. It is therefore important to survey the concrete in the state of greatest interest, taking account of the local climate. Low values, for example in the range of 50 to 100 m, indicate the risk of a high corrosion rate; intermediate values in the range of 100 to 200 m indicate the likelihood of a low to moderate corrosion rate. Values in excess of 200 m indicate that corrosion rates are likely to be low. Other devices are available for measuring resistivity on site, such as one commercially available as part of a linear polarisation resistance measurement system. Linear polarisation resistance measurement The half-cell mapping technique can be used to identify corrosion activity but does not give an indication of the rate of corrosion. Equally resistivity measurements can only indicate the likely corrosion rate if it occurs. Taken


75

Figure 4.18 Test set-up for linear polarisation resistance measurement

together the two techniques can yield very useful information but if a measure of the corrosion rate is required this can best be achieved through the polarisation resistance technique. The technique involves making an electrical connection to the reinforcement followed by the application of a small current and measurement of the shift in potential. The corrosion rate may then be deduced. If a second electrode is introduced to a reinforced concrete system and an external current is applied the anodic and cathodic potential is shifted, or polarised, from the corroding potential (Ecorr). An external corrosion current then exists between the two electrodes and can be measured. At very low external polarisation voltages a linear relationship exists between external current and the polarisation. The polarisation resistance (Rp) is defined as follows: Rp

E I

where E change in potential I applied current The polarisation resistance is inversely proportional to the notional corrosion current at the mixed potential (Ecorr) and therefore the corrosion rate may be detected. Experience with a commercially available device that has been developed for on-site polarisation resistance measurements is described by Rodriguez et al. (1995). The device confines the applied current by means of a circular sensor with concentric counter electrodes (Feliu et al. 1990). The sensor houses a central reference electrode and a pair of electrodes between the concentric rings (Figure 4.18).

76 Fundamentals of durable reinforced concrete The polarisation resistance is determined and the corrosion rate (Icorr) is deduced from the following relationship: Icorr

K Rp

where the constant K is taken as 26 mV for corroding reinforcement and double that figure for passive reinforcement. Rodriguez et al. (1995) report that rates less than 0.2 A/cm2 indicate reinforcement in the passive condition; values up to 0.5 A/cm2 indicate low to moderate rates; values in the range of 0.5 A/cm2 to 1.0 A/cm2 indicate moderate to high rates. Values greater than these are indicative of high corrosion rates.

Summary An understanding of the electrochemical process of corrosion is required to fully appreciate the characteristics of durable concrete. Prevention of corrosion is best approached by creating conditions where the reinforcement remains passive and where the ingress of moisture and oxygen is severely limited. The threat to passivity comes from carbonation and chloride ingress. Corrosion activity may take the form of homogenous corrosion or pitting corrosion. The latter is commonly associated with chloride ingress and represents a significant threat to durability. Ingress of oxygen and moisture may be limited by the specification and achievement of impermeable concrete through a low water/cement ratio, good compaction and good curing and in most cases the influence of cracking will not be significant. Where corrosion activity occurs it can be detected through techniques involving the electrical properties of concrete and a corroding cell.

5

Carbonation

Carbonation and corrosion Carbonation is the term used to describe the effect of carbon dioxide on a material. The phenomenon rarely leads to structural problems but carbonation-induced corrosion can lead to unsightly spalling on structures. The repair costs, for example to multi-storey office development façades, can be considerable. Carbonation of cementitious materials results in a lowering of the pH – making the material less alkaline – and hence the term ‘neutralisation’ is also sometimes used in the literature. Reinforcement in concrete is embedded in an oxygenated alkaline solution. The reinforcement will not corrode if the protection afforded by the passive film – a thin layer of oxide deposited on the steel – remains substantially intact. This insoluble oxide film prevents oxygen reaching the steel and inhibits corrosion. The reinforcement is said to be ‘passive’ when it is in this state. Corrosion of reinforcement can commence however if the passive oxide film protecting the reinforcement is destroyed, the cover concrete is sufficiently permeable to oxygen and moisture, and the concrete is moist enough to serve as an electrolyte. The lowered pH in zones of carbonated concrete may threaten the continuity of the passive film. It is important therefore to specify cover concrete that is capable of resisting the penetration of the carbonation front as far as the reinforcement during the service life of the structure. The pH of the pore solution of fresh concrete is approximately 12.6. The alkaline nature of concrete is principally due to the presence of calcium hydroxide, Ca(OH)2, formed during the hydration of cement. Dissolution of the calcium hydroxide leads to the presence of hydroxyl ions in the pore water and this gives concrete its high pH. The calcium hydroxide is susceptible to reaction with carbon dioxide from the atmosphere. The reaction proceeds in the presence of moisture as water provides a medium for the reaction. Concretes in service will almost always contain sufficient moisture for the reaction to proceed. The reaction involves the production of calcium carbonate (CaCO3). Conversion to calcium carbonate influences the surrounding pore fluid pH, which falls to about 8.3. This fall in pH results in

78 Fundamentals of durable reinforced concrete a carbonation front which divides carbonated and non-carbonated zones. While carbonation does not adversely affect the concrete itself the durability of reinforcement may be compromised. This is because the passive ferrous oxide layer on the reinforcement breaks down when the surrounding pH falls below 9.

Chemistry of carbonation Initially carbon dioxide diffuses through the surface of the concrete due to the concentration difference between the atmosphere and the concrete pore structure. A thin skin of carbonated concrete develops which may be less than a millimetre in thickness. Further penetration is primarily a function of the concrete permeability and the amount of calcium hydroxide available for reaction. Carbon dioxide passes unhindered through the carbonated layer and is available for reaction with the next layer of calcium hydroxide. It may progressively penetrate further into the concrete over time and ultimately part of the carbonation front may reach the reinforcement and cause depassivation (Figure 5.1). The carbon dioxide can react with constituents of both unhydrated and hydrated cement. In the context of durability the latter is of greatest interest. The various reactions have been summarised by, for example, Aschan (1963), Berger and Klemm (1972) and Papadakis et al. (1989). The most significant reaction is that involving calcium hydroxide and therefore carbonation of concrete is generally summarised as follows: H2O

Ca(OH)2 CO2 → CaCO3 H2O The steps of the reaction are first the dissolution of calcium hydroxide followed by reaction with dissolved carbon dioxide: Ca(OH)2 → Ca2 2OH Ca2 2OH CO2 → CaCO3 H2O Additional reactions involving unhydrated constituents and calcium silicate hydrate are possible whereby the CaO in these compounds combines with carbon dioxide to form calcium carbonate and hydrated silica: 3CaO.SiO2 3CO2 yH2O → SiO2.yH2O 3CaCO3 2CaO.SiO2 2CO2 yH2O → SiO2.yH2O 2CaCO3 The mineralogical forms of calcium carbonate are calcite, aragonite and vaterite. Calcite, the most stable, is the ultimate form found in carbonated concrete (Gaze and Robertson 1956). Sauman (1971) found that the carbon-

Carbonation

79

Figure 5.1 Ingress of the carbonated zone to the reinforcement

ation process produced some vaterite first but this was gradually converted to the more stable calcite. This may be influenced by the degree of hydration. It has been reported by Kondo et al. (1969) that the carbonation of hydrated C3S produced calcite whereas unhydrated C3S produced vaterite. The vaterite level decreases if moisture is available to maintain the process of hydration. Cole and Kroone (1960) reported that vaterite is formed first and that this is transformed into aragonite and finally into poorly crystallised calcite, following Ostwold’s law of successive reactions. The process of carbonation in concrete is not uncontrollable. It is possible to produce durable concrete in which the rate of carbonation may be so slow that, in engineering terms, the carbonation front may be regarded as having reached a limit. This limit may be of the order of a millimetre or less. This situation can occur through a further lowering of the diffusivity in highly impermeable concrete by deposition of carbonation products. The calcium carbonate occupies a greater volume than the reacting constituents.

Detection of the carbonation front There are many different methods for locating the carbonation front. These include acid/base indicators, mineralogical analysis, differential thermal analysis, thermogravimetry, x-ray diffraction, neutron radiography, infra-red spectroscopy, and chemical analysis. The simplest method to use is the acid/base indicator phenolphthalein. This indicator solution method is the basis of almost all carbonation studies, especially due to its usefulness in field tests. The indicator changes colour at a pH of approximately 9. Below this figure it remains colourless but above pH 9 it turns purple. It requires little skill in use and gives reproducible results. The test is carried out on freshly broken surfaces brushed free of dust and sprayed with indicator


Figure 5.2 Forms of carbonation profile encountered in practice

Figure 5.3 Influence on carbonation profile of biaxial penetration of carbon dioxide

solution. Readings on structures in service are best carried out on cores. Alternatively on-site testing may be carried out by drilling a 20 mm diameter hole and exposing the edges of the hole with a hammer and chisel. Phenolphthalein may then be sprayed onto the freshly broken surface. The smooth drilled surface is not amenable to testing. Concrete that is difficult to

Carbonation

81

expose may be examined by drilling closely spaced holes and breaking out the concrete in between. The indicator solution is usually prepared from a gramme of phenolphthalein powder per 70 ml of ethanol and 30 ml distilled/ion exchanged water. The depth of carbonation may therefore be regarded as the average distance from the surface of the concrete element to the zone where phenolphthalein indicator solution changes colour to purple, indicating that carbon dioxide has not reduced the alkalinity of the hydrated cement in that zone. The depassivation threat from carbonation is adequately assessed by locating the front with phenolphthalein even if the indicator does not mark the border between calcium hydroxide-dominated and calcium carbonate-dominated zones with absolute precision. The carbonation front does not always advance at a constant rate due to inhomogeneities in the concrete and dense aggregates are not coloured by phenolphthalein. Thus it may be necessary to record both the average and the maximum depth of carbonation (Figure 5.2). The average depth gives an indication of the quality of the concrete and the influence of the local environment. The maximum depth is important from the point of view of durability being threatened if a sufficient number of peaks on the front reach the reinforcement. The phenolphthalein test may also be carried out on site by breaking off pieces of concrete and spraying the exposed concrete. This may be misleading if readings are concentrated on corners of columns and beams. These areas are the easiest to break off but they may have high carbonation depths that may not be representative of the general structure. The corners will experience biaxial penetration of carbon dioxide, as illustrated in Figure 5.3, and the permeability of concrete placed in the corners of shutters may not be representative of the member due to compaction difficulties.

Primary factors influencing carbonation rate Carbonation rate is significantly influenced by several factors that interact and the combined effect of these may exacerbate or ameliorate the process. The primary factors, determined from both field observations and theoretical considerations, are as follows: • • • •

diffusivity/permeability; reserve alkalinity; the environmental carbon dioxide concentration; the exposure condition.

Diffusivity and permeability The carbonation phenomenon is governed by the diffusion process whereby material migrates from an area of high concentration to one of lower

82 Fundamentals of durable reinforced concrete concentration. Permeability, on the other hand, is generally characterised by flow under the influence of a pressure difference. Nevertheless the permeability properties of a concrete are often used to indicate its quality with respect to carbonation resistance. The lower the permeability the greater the resistance to the inward diffusion of carbon dioxide into the concrete, for a given moisture content. The requirements for achieving low permeability concrete are well known and include: low water/cement ratio, proper compaction, adequate and timely curing. Strength is related to these three factors and so it may be postulated that the carbonation rate should be inversely proportional to strength. While strength is not a significant factor in itself, lower rates of carbonation have generally been observed in stronger concretes. The rate of carbonation in high quality concretes may be so low that the carbonation depth tends to a limit of a millimetre or less. Low depths of carbonation may also be detected on occasion in weak, highly permeable concretes exposed to rain. In such cases the diffusion of carbon dioxide will have been fortuitously hindered by water-filled pores. Carbon dioxide diffusion is 106 times slower in water than in air. Nevertheless carbonation rates in tropical climates, such as in Singapore, have been found to be higher on average than in comparable concretes in temperate climates despite higher relative humidity levels. A possible explanation for this has been proposed by Roy et al. (1996) who note that the higher temperatures in tropical climates may influence the diffusion coefficient. It may also be noted that a 10°C rise in temperature is generally associated with a doubling of reaction rate in chemical reactions. Reserve alkalinity The ability of a concrete to resist the progress of the carbonation front is related to the volume of calcium hydroxide present. This reserve of alkalinity may be considered as a buffer with the calcium hydroxide acting as a sacrificial defence system – the more calcium hydroxide present the more effective the filter against carbon dioxide. The calcium hydroxide volume is proportional to the calcium oxide content of the cementitious binders. The reserve alkalinity of concrete made from blended cements, such as Portland fly ash cement is lower than that made with Portland cement. In relation to carbonation however this disadvantage is counterbalanced by the lower permeability of well-cured blended cement concretes. Indeed the influence of the water/cement ratio on the carbonation rate is greater than the cement content. The benefits of materials such as pulverised fuel ash in the carbonation context are related to the reduction in water demand for a given workability. Further support for the proposition that strength plays a significant role in diminishing the carbonation rate comes from consideration of the role of cement content. Under favourable hydration conditions strength is

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83

Figure 5.4 Influence of environment on the rate of carbonation

proportional to cement content. The higher the cement content the greater the amount of calcium hydroxide produced and, therefore, the higher the reserve alkalinity. Environmental carbon dioxide concentration The rate of carbonation increases with increasing carbon dioxide content. Highest rates, other than in particular industrial situations, are found inside buildings. Atmospheric carbon dioxide levels, currently quoted as 360 parts per million (ppm), are rising. Hayward (1997) reported that the concentration in 1800 averaged 280 ppm but that this had risen to 340 ppm by 1990. The forecast, should the present trend continue, is for an exponential rise to 700 ppm by the year 2100. The level of carbon dioxide in the atmosphere varies slightly from a minimum in coastal environments to a maximum in urban environments. Highest levels typically occur in the interior of buildings. This has been found to influence the rate of carbonation as illustrated in Figure 5.4 using data from a study in Ireland (Richardson 1988). Rates in the coastal environment are low perhaps due to high humidity but of course chloride-induced corrosion would pose a greater threat to structures in coastal and marine environments. Rates in rural and suburban areas are typically lower than urban areas but in practical terms the differentiation is not meaningful as microclimates with high carbon dioxide levels can occur in any area. The rates inside buildings are high due to low humidity levels, and consequent semi-dry permeable pore structure, as much as being due to the higher


Figure 5.5 Influence of relative humidity on the rates of carbonation and corrosion

carbon dioxide concentration. However the low humidity levels discourage corrosion initiation and so the internal environment is not always a cause for concern. It should be borne in mind that carbonation and corrosion are not inexorably linked. In certain concretes the carbonation front may advance beyond the cover concrete but the relative humidity of the atmosphere may not be high enough to sustain corrosion. Rates of carbonation are highest in the range 50 to 75 per cent relative humidity. Below about 45 per cent relative humidity the water in the pores is not in a state that encourages dissolution of calcium hydroxide or carbon dioxide and this reduces the reaction rate. Above 75 per cent relative humidity the influence of water filling the pores becomes significant. The critical relative humidity level with respect to corrosion is about 80 per cent, below which there is insufficient moisture to sustain the process. The trends are illustrated in Figure 5.5. It should be noted that very low rates of carbonation have also been observed in outdoor concretes of all environmental categories. This further highlights the significance of exposure to moisture. Exposure condition Low rates of carbonation have been observed in outdoor concretes of all strengths and in all environmental categories. This highlights the significance of the degree of exposure and hence, the degree of moisture in the permeable pore structure. It would appear that moisture plays a fundamental role in the phenomenon and greatly influences the carbonation rate. This is most probably due to the significant effect on the diffusion coefficient of the

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85

Figure 5.6 Influence of exposure condition on the rate of carbonation

moisture content of the pore structure. The permeability of wet concrete is a fraction of that for dry or semi-dry concrete. This is well illustrated by analysing data from the study of concrete in Ireland categorised as follows: • • •

external, exposed to rain; external, sheltered from rain; internal.

The results are illustrated in Figure 5.6. These trends agree with studies in other countries of temperate climates using the same categories. The external environment in Ireland includes moderate to high relative humidity. Rainfall in excess of one millimetre per day occurs on average half of the time. It may be anticipated therefore that concrete in the ‘external, exposed to rain’ category would be wet and rarely dry. The permeability of these concretes would therefore be low, irrespective of the quality of the concrete. Concrete in the ‘external, sheltered from rain’ category would experience moderate humidity. The durability of these concretes would be critically dependent on the concrete quality. Concrete in the internal environment would be dry, potentially leading to high permeability and would also be exposed to high carbon dioxide concentrations. The likelihood of corrosion of reinforcement in such an environment is low.

Mathematical modelling of the rate of carbonation The phenomenon of carbonation provides one of the earliest examples of the application of mathematical modelling to deterioration processes in concrete.

86 Fundamentals of durable reinforced concrete A formula was published in 1928 by Uchida and Hamada showing the depth of carbonation to be proportional to water/cement ratio and the square root of time. Since then many models have emerged which indicate that a square root relationship provides a reasonable basis on which to predict future behaviour, being somewhat conservative for most exposure conditions. Observations in the field have shown that carbonation rates lower than that predicted by the square root model are experienced where the concrete is subject to rain. Nevertheless a significant aspect of the (approximately) square – root relationship x k √t to specification and site practice is that a reduction of cover by a quarter could halve the time to depassivation. Empirical formulae This square root relationship may be stated as follows: – x k √t where x depth of carbonation k ‘k-factor’ dependent on diffusivity, reserve alkalinity, carbon dioxide concentration, and exposure condition t time This approximate relationship has been verified by observations of concrete in the laboratory and in service. For example, Alexandre (1976) published the following model: –— x √ 2k1

–

√t

where k1 material coefficient Equally Alekseev and Rozental (1976) published the formula: n– x A √t

where A coefficient n 1.92 for water/cement ratio 0.6 n 2.54 for water/cement ratio 0.7 Inclusion of the water/cement ratio is logical since it would have a significant influence on the permeability of the concrete and hence on the ease with which carbon dioxide could diffuse into the concrete. The significance of water/cement ratio was long recognised. The formula of Uchida and Hamada (1928), referred to earlier, may be stated in the following form: x

– w0 0.3 √t √0.3(1 3w0)

where w0 water/cement ratio

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87

Similarly, formulae by Kishitani (1960) can be presented in the following forms for water/cement ratios (w) below 0.6 and equal to or above 0.6 respectively: x

4.6w 1.76 – √t 7.2

x

– w 0.25 √t √0.3(1.15 3w)

It may be noted that the second of Kishitani’s formulae is basically that of Uchida and Hamada but with w0 substituted by (w 0.05). The difference may be accounted for by the moisture state of the sand used in their experiments. Further development of these formulae were published by Kishitani (1964) and Hamada (1969) with the introduction of a factor ‘R’ which took account of cement type, aggregate type, and agents such as airentrainment admixtures. Tsukayama et al. (1980) published a formula along similar lines to that of Kishitani but with a slightly lower value of the material factor. It may be stated in the following form: – x (1.187w 0.493) √t The foregoing set of formulae concentrated on the influence of water/cement ratio, recognising its significant influence on permeability. The effect of cement content, which influences the permeability and reserve alkalinity was implied but not specifically included. This was addressed by Smolczyk (1969) who introduced the influence of cement content by inclusion of compressive strength in a formula for carbonation: xa

–

( √wN––– b) √t T

where a and b are coefficients and NT represents the compressive strength at T days. Another variation is that of Weirig published by the Commission Carbonatation (1972) as follows: x

– –––– 1.12w 1.47) √t (√582w so sf

where s is the loss on ignition and osf is the specific surface. Other empirically derived formulae have been published by, for example, Richardson (1988). Theoretical verification of the square root relationship The empirical models are limited in their usefulness because they cannot be extrapolated to predict performance in all circumstances. Therefore a more


Figure 5.7 Parameters in mathematical model of carbonation rate

fundamental approach to modelling the carbonation phenomenon has been undertaken by several researchers. A theoretical basis for the square root relationship has been proposed since the 1960s by researchers such as Alekseev and Rozental (1976), Meyer et al. (1967) and Engelfried (1977). The theory involves equating the two processes involved: diffusion of carbon dioxide and the chemical reaction between carbon dioxide and calcium hydroxide. The relevant parameters are listed below and illustrated in Figure 5.7. n quantity of carbon dioxide diffusing through element (kg) D diffusion coefficient (m/s2) A area (m2) r concentration of CO2 at surface (kg/m3) rx concentration of CO2 at depth x (kg/m3) c amount of alkaline material available in unit volume of concrete to buffer carbon dioxide (kg/m3) t time (s) The diffusion process is governed by Fickian diffusion. Thus, using standard sign convention we may state: dn D A or

r rx dt x

Carbonation dn D A

89

| r rx | dt x

The chemical processes may be stated as: dn c A dx Hence c A dx D A

| r rx | dt x

and x dx D

| r rx | dt c

By integration we may solve for x. In addition it may be assumed that the carbon dioxide concentration at the depth of carbonation will be zero. Thus x

√

2Drt c

The quantity of carbon dioxide in the atmosphere (parameter ‘r’) is about 360 ppm, which equates to 7.2 104 kg/m3. The amount of alkaline material available in a unit volume of concrete to buffer the carbon dioxide (parameter ‘c’) may be calculated as the product of the cement content, the fraction of calcium oxide in the cement (for example, 0.65 for Portland cement), and the ratio of molecular weights of carbon dioxide (44) to calcium oxide (56), a value of 0.786. A slight restatement of the above formula is that of Kondo et al. (1969): x

√

2 D c – √t c0

where the denominator is the product of the amount of reactant per unit weight and the density. Formula based on reaction engineering A fundamental model of carbonation of concrete was developed by Papadakis et al. (1989) based on reaction engineering principles. The model considers the mass conservation of carbon dioxide, calcium hydroxide and calcium silicate hydrate. The formula was derived by considering the kinetics of the hydration and carbonation reactions, the decrease in porosity

90 Fundamentals of durable reinforced concrete consequent on the fact that the products of carbonation have higher molar volumes than the calcium hydroxide and calcium silicate hydrate, and the influence of humidity. The model has been found to give good agreement with experimental results at typical relative humidity levels (Papadakis et al. 1990). For fully hydrated concrete it is of the form: x

– 2 [CO ] D √t ( √[Ca(OH) ) ] 3[CSH] 2

e,CO2

2

where [CO2] molar concentration of carbon dioxide (mol/m3) [Ca(OH)2] molar concentration of calcium hydroxide (mol/m3) [CSH] molar concentration of calcium silicate hydrate (mol/m3) De,CO2 effective diffusivity of carbon dioxide (m2/s) A further nineteen formulae are employed in the solution of the model involving parameters such as the cement composition, water/cement ratio, aggregate/cement ratio, carbon dioxide concentration, and the relative humidity. Certain simplifications can be applied to the model while retaining its applicability (Papadakis et al. 1992). These are as follows: [CO2] 42 yCO2 [Ca(OH)2] 3 [CSH]

33 000 1 ( c / w)(w/c) ( c / a)(a/c)

[

De,CO2 (1.64 106) p1.8 1

RH 100

]

2.2

with p

(

) 1 w/c( / 0.3)(w/c) c

w

c

w

where yCO2 ambient carbon dioxide content by volume

c mass density of cement (kg/m3)

w mass density of water (kg/m3)

a mass density of aggregate (kg/m3) w/c water/cement ratio a/c aggregate/cement ratio p porosity of hardened cement paste RH relative humidity A simplified version of the expression for effective diffusivity, which was found to fit well with experimental data, is:

[

De,CO2 p2 1

RH 100

where 1.2 0.1

]

2

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91

Thus for ordinary Portland cement concrete the depth of carbonation may be modelled by the following formula: xc 350

(

) 1 w/c( / 0.3)(w/c) ( 1 RH 100 ) c

w

c

w

{[1 (

) (w/c) (

) (a/c) ] y t } c

w

c

a

0.5

CO2

Deviation from the square root relationship Consideration of a range of material properties led Daimon et al. (1971) to publish a formula that took account of the deviation from the square root relationship through the concept of an induction period. The formula was of the form: –— –––— x √ ke √t ti where ke

kt f

kt

2 D0 P T 0.75 2731.75 R c0

and

where ti ‘induction time’ t exposure period ke rate constant obtained experimentally kt theoretical rate constant f tortuosity D0 diffusion coefficient at 0°C porosity of specimen P partial pressure difference across the carbonated layer T temperature R gas constant c0 amount of reactant in unit weight of concrete specimen

density of specimen Tendency to a limit In practice, particularly in permanently or cyclically wet environments, the carbonation depth tends to a limit. The limiting value of carbonation depth is thought to be due to two factors. The first involves the back diffusion of material from the pore water in the non-carbonated zone to the carbonated

92 Fundamentals of durable reinforced concrete zone due to the concentration difference. The second involves the reduction in diffusion coefficient caused by the deposition of calcium carbonate in the concrete pores. The reaction products of carbonation occupy a greater volume than the constituents that react. Thus a limit may be reached when the amount of carbon dioxide reaching the carbonation front is low enough to just about react with the counter diffusing hydration products. An alternative possibility is that the pores may become waterlogged for periods thus slowing the diffusion of carbon dioxide to an extent that change in carbonation depth is not detectable. This has been addressed by several researchers including Martin et al. (1975), Schiessl (1976), Frey (1993), Bakker (1994), and Sickert (1997). Schiessl (1976) demonstrated that the Fickian diffusion model could still be used taking these factors into account. First, the diffusion coefficient was not constrained to be constant and was allowed to decrease with time due to the reduction in permeability caused by deposition of calcium carbonate in the pore structure of concrete. The diffusion coefficient would also vary with time due to fluctuations in moisture level. Thus the diffusion coefficient (D) was replaced by a mean value (Dm): Dm D (1 f x) where f environment factor Second, the counter diffusion of calcium hydroxide was taken into account by assuming that an additional amount of carbon dioxide (b kg/m2) was required to enter the concrete to react with the counter diffusing calcium hydroxide. The equations for the physical and chemical processes therefore became: dn D (1 f x) A

| r rx | dt x

and dn c A dx b A dt Thus: c A dx b A dt D (1 f x) A

| r rx | dt x

and dx

[D (1 c f x) | r x r

x

|

]

b dt c

The solution to the equation as time approaches infinity is the value: x

D | r rx | d

Carbonation

93

where d f D | r rx | b Thus the solution of these equations for the depth of carbonation approaches a limit as time approaches infinity, thus validating the link between theoretical and observed behaviour. The departure from the strict square root relationship is also accounted for. The considerations taken into account by Schiessl were also studied by Frey (1993) and further examined by Sickert (1997). Frey proposed the following formulation: 2

dk (e Aw 1) (1 ew t ) 0.5

where dk depth of carbonation (mm) t time (years) A parameter (for example, 0.11 year1) w carbonation characteristic number (for example, 0.25 year0.5) The carbonation characteristic number is dependent on the concrete composition. The formula reaches a limiting value as time approaches infinity of: 2

dke (e Aw 1) Bakker (1994) further investigated the tendency towards a limit in circumstances where the concrete was subject to cyclical wetting and drying. The effect of wetting and drying cycles on carbonation is illustrated in Figure 5.8. Bakker showed that if one considers the wetting time tw and related drying time td to be as follows: tw t n t n1 td

x n1 B

then: –––— xn A √teff n where teff n td1 td2

(xB ) 1

2

...


Figure 5.8 Influence of wetting and drying cycles on the rate of carbonation

. . . tdn and A B

√ √

(x B ) n1

2

2 Dc (C1 C2 ) a 2 Dv (C3 C4) b

if xn carbonation depth after n cycles (m) n number of cycles a amount of alkaline substance in concrete (kg/m3) Dc effective diffusion coefficient for carbon dioxide at a given moisture distribution in the pores (m2/s) C1 C2 carbon dioxide concentration difference between air and the carbonation front (kg/m3) tdn length of nth drying period (s) Dv effective diffusion coefficient for water vapour at a given moisture distribution in the pores (m2/s) C3 C4 moisture difference between air and the evaporation front (kg/m3) b amount of water to evaporate from the concrete (kg/m3) td length of drying period (s) tw length of wetting period (s)

Application of models to service life prediction The use of mathematical models for service life prediction involves the choice of a model that balances accuracy with simplicity. The empirical relationships

Carbonation

95

are limited in applicability but, on the other hand, more complex models are not readily applicable due to the difficulty in determining some of the parameters. The application of the models to practice has been explored by Parrott (1994), who considered in particular the diffusion aspect, and by the CEB who are further examining the significance of the chemical buffering effect. Parrott (1994) refined the basic square root formula to take account of the deviation from the square root relationship as relative humidity increases. The diffusion coefficient was modelled by the air permeability of the cover concrete. The amount of alkaline material was represented by a factor related to the calcium oxide. The relationship may be stated as follows, using the notation of Parrott: d

a k 0.4 t ni c0.5

where d depth of carbonation at the end of the initiation phase (mm) a coefficient k air permeability of the cover concrete (units of 1016 m2) ti duration of initiation period to start of corrosion c calcium oxide content in hydrated cement matrix that can react with carbon dioxide (kg/m3 of the cement matrix) n power exponent. Value close to 0.5 for indoor exposure and decreases with increasing relative humidity The carbonation rate in the internal environment may thus be modelled – by Parrott’s equation by the x k √t relationship and this accords with observations under controlled conditions. The rate of carbonation is lower in – exposed environments and departs from the √t relationship. Hence Parrott suggests the introduction of a power exponent n, which is close to 0.5 for indoor exposure but decreases with increasing relative humidity above 70 per cent. Additionally Parrott takes account of the variation in air permeability value (k) through a factor related to relative humidity. The use of a parameter related to calcium oxide content demonstrates that the relationship could be used in the design of durable concrete using parameters that are more readily determined than the more theoretical ones that form part of the classic carbonation formulae. The application of Parrott’s formula in practice requires evaluation of specific parameters, some of which may require calibration with extensive field experience. The depth of carbonation (d) at the end of the initiation phase would in practice be taken as the minimum depth of cover. The coefficient (a) is used to calibrate the equation with observations from structures in service. Parrott has assigned a value of 64 for the coefficient, based on an extensive literature review. The air permeability is dependent on the relative humidity but may be estimated from tests on a specimen dried at 60 per cent relative humidity by the following relationship:


Figure 5.9 Decline in coefficient ‘m’ in Parrott’s (1994) carbonation formula as relative humidity increases above test value

k m k60 where m coefficient k60 permeability of specimen dried at 60 per cent relative humidity The value of m varies from 1.00 at 40 per cent relative humidity to one hundredth of this value at saturation level as illustrated in Figure 5.9. The calcium oxide content depends on cement composition, exposure condition, and proportion of cement reacted. It represents the alkaline buffer. For example Parrott quotes values for a CEM I cement from 460 kg/m3 at relative humidity levels in the range 40 to 70 per cent rising to 610 kg/m3 at saturation level. The model being explored by the CEB is based on the following form of the carbonation relationship: xc

√

2 k1 k2 Deff Cs – t0 √t a t

( )

n

where xc depth of carbonation (m) at time t (s) k1 constant related to execution k2 constant related to exposure condition Deff effective diffusion coefficient (m2/s) Cs concentration of carbon dioxide a chemical buffering capacity t0 age at which Deff is determined n constant related to the environment

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97

Estimates of the relative buffering capacity of Portland cement and slag concretes have been made based on published data. These examples show how accessible the development of a model for durability design could be, given the extensive research on durability issues in recent decades. Refinement of the input parameters and calibration with observations in practice are required to allow further simplification of the models for the safe introduction of design for durability in practice.

Carbonation: exposure categories in EN 206–1 Reinforced concrete may deteriorate through corrosion of reinforcement induced by carbonation. Unreinforced concrete is not at risk because changes caused to the material by carbonation are not detrimental. Carbonation rate is significantly influenced by the moisture state of the permeable pore structure. The wetter the concrete the slower the rate of carbonation. Equally, in conditions of low humidity corrosion is unlikely even if the carbonation rate is high. These facts are reflected in the exposure classes identified for consideration in EN 206 in relation to the risk of corrosion induced by carbonation. Four exposure classes have been distinguished in respect of corrosion induced by carbonation and these are designated as XC1, XC2, XC3, and XC4. An informative annex in EN 206–1 presents indicative limits for maximum water/cement ratio, minimum cement content, and minimum strength class for each exposure condition, based on an intended working life of fifty years. These are informative values only, based on the mean of values currently representative of European practice for CEM I concrete. They are not therefore necessarily appropriate for use in all countries, nor for intended working lives that differ significantly from fifty years. They do however serve as useful comparative benchmarks. Specifiers must consult the corresponding advice on limiting values in national annexes or complementary standards valid in the place of use. Examples include BS 8500 published for practice in the United Kingdom and DIN 1045–2 for use in Germany. Exposure class XC1 Exposure class XC1 covers reinforced and prestressed concrete that is either dry or permanently wet. The risk of corrosion in either case is low but for differing reasons. In the case of dry concrete it is likely that carbonation rates will be high but that corrosion will not occur due to the low humidity. In the case of permanently wet concrete, carbon dioxide will not have ready access through the water-filled pores and so carbonation rates will be low. Examples of dry concrete in this context include the surfaces of members inside buildings or structures where the relative humidity will remain low at all times during the service life. Processes or activities that lead to periods of high humidity within the building would require consideration of an alternative

98 Fundamentals of durable reinforced concrete class. Examples of permanently wet concrete obviously cover the case of the surface of members that will be submerged at all times during the service life. Any likely change of use of the building or structure during its life would also need to be considered if this could alter the dry or permanently wet condition. Given the low risk to durability posed by carbonation in these conditions, the limiting values on concrete composition need not be onerous. In the case of XC1, the informative annex to EN 206–1 shows the minimum cement content as 260 kg/m3 and the maximum water/cement ratio as 0.65. It will be recognised that such values do not represent onerous restrictions on the production of structural concrete. Hobbs et al. (1998b) demonstrated that, due to the low risk of corrosion, the requirement for a 100-year intended working life in the United Kingdom should not demand a change to the requirements for a fifty-year intended working life and this is reflected in the BS 8500 recommendations. Exposure class XC2 Exposure class XC2 covers reinforced and prestressed concrete that is wet, rarely dry. The risk of corrosion in such concrete is low but it is higher than for class XC1. Carbonation may occur during the drying periods but since these will be rare the overall rate of carbonation will be very low. Examples of wet, rarely dry concrete could include reservoirs or water towers if their service conditions kept them full most of the time, and members below ground level, including foundations. The informative values for class XC2 in EN 206–1 indicates a typical requirement for an increase in cement content of 20 kg/m3 and a decrease in the maximum water/cement ratio by 0.05 in comparison with class XC1. This is reflected in national recommendations in the United Kingdom and in Ireland but not in Germany, thus emphasising the need for specifiers to be mindful of relevant national provisions. Hobbs et al. (1998b) demonstrated that, in the United Kingdom, the requirement for a 100-year intended working life should not demand a change to the requirements for a fifty-year intended working life. Exposure class XC3 Exposure class XC3 covers reinforced and prestressed concrete that is exposed to moderate humidity. The risk of corrosion in such an environment is significant if the depth of carbonation can reach or exceed the depth of cover during the service life. Examples of such situations are concrete surfaces in the external environment that are sheltered from rain and surfaces in the internal environment where humidity is higher than normal. The latter condition can arise, for example, in buildings housing processes related to the brewing industry or in commercial laundries.

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99

The informative values for class XC3 in EN 206–1 indicates a typical requirement for an increase in cement content of 20 kg/m3 and a decrease in the maximum water/cement ratio by 0.10 in comparison with class XC1. The study by Hobbs et al. (1998b) demonstrated that, in the United Kingdom, a significant reduction in maximum water/cement ratio and increase in minimum strength class would be required to extend the intended working life from fifty to 100 years. The values were of the order of 0.15 and approximately 15 N/mm2 (based on cube strength) respectively. This is reflected in the recommendations of BS 8500, which also allows a trade-off between concrete composition and cover. Exposure class XC4 Exposure class XC4 covers reinforced and prestressed concrete that is cyclically wet and dry. The severity of this category in comparison with XC3 varies across Europe depending on the local climate. In certain regions the drying cycles would allow carbonation to proceed at a reasonable rate while the wetting cycles would promote rapid corrosion once depassivation has occurred. In other regions the climate would generally be moist enough to reduce the drying effect and thereby limit the window of opportunity for carbonation. In the case of moist climates this exposure class can be treated as being equivalent to exposure class XC3, while in other regions it represents a more severe condition. Examples of concrete covered by class XC4 include surfaces in the external environment that are exposed to rain, external or internal surfaces exposed to contact with water or high humidity on a cyclical basis through natural or industrial processes. The informative values for class XC4 in EN 206–1 show a 40 kg/m3 increase in the minimum cement content and a decrease in maximum water/cement ratio of 0.15 in comparison with class XC1. However Hobbs et al. (1998b) indicate no change in requirements from class XC3 due to the effect of the United Kingdom climate on the rate of the combined carbonation/corrosion process and this is reflected in Irish and British standard recommendations.

Specification by performance There is a desire to move away from the specification of durable concrete through limitations on mix composition. Durability would instead be specified by performance. The benefits of this would be significant in that clients would get the performance they required while the producers would be free to use their expertise in producing the optimum mix. Performance testing for carbonation is by necessity a lengthy process and therefore carbonation testing cannot be used as a routine control test. However the Concrete Society (1996) describe how such tests could be used as a type approval test. This would involve tests on concrete samples to establish that the performance of a particular mix is satisfactory. Additionally the

100


test could establish the relative performance of a mix with one of known performance. In the case of carbonation Harrison (1996) notes that such a type approval test should establish the relationship between concrete strength and depth of carbonation so that strength could be used as the control measure for a given production unit. The test being developed by CEN is based on the method of measurement set out by RILEM Committee CPC–18 (1985) and employs the phenolphthalein method. Two concrete prisms of 100 100 mm cross-section are subjected to controlled conditions of temperature, relative humidity and carbon dioxide concentration. Five readings are taken on each face and the average calculated for each prism and thence for both. The readings are taken at intervals of up to a year. The test is being developed to overcome earlier problems whereby repeatability was good but reproducibility was not.

Summary Carbonation should not be a significant problem in future works in temperate climates, such as in the United Kingdom and in Ireland. The quality of concrete required from the viewpoint of strength will generally assure resistance to carbonation. The carbonation depth often reaches a limit at very low values or may proceed at a rate that is slow enough to prevent the depassivation of reinforcement within the design life of the structure. Serviceability problems that do arise will generally be due to inadequate construction technique, for example poor curing, or inadequate cover due to errors in specification or construction. Consideration of the simplified square root relationship between depth of carbonation and time emphasises the importance of achieving the specified cover. A reduction of about one quarter in the cover could lead to a halving of the time to depassivation of the reinforcement. Rates of carbonation inside buildings will be significantly higher than outdoors and the carbonation front may proceed through the full depth of cover within a decade or two. However corrosion may not ensue if the relative humidity indoors remains low. Caution is however required where the environment promotes significant wetting and drying cycles which promote cycles of high carbonation rate followed by conditions conducive to corrosion propagation. The maximum rate of carbonation may be up to double the average rate of carbonation due to local inhomogeneities. Therefore isolated spalling may provide early warning of the potential for full depassivation of the reinforcement within the service life of the structure. Mathematical modelling of carbonation is well advanced and the prospect of designing for durability through the application of straightforward calibrated models is good. Difficulties still exist with the reproducibility of carbonation performance tests but this problem is being addressed.

6

Chloride ingress

Chloride ingress and corrosion There is universal recognition that corrosion of reinforcement due to chloride penetration is the most significant threat to the existing reinforced concrete infrastructure of developed countries. Structures such as road bridges and harbour facilities are exposed to chloride-rich environments through application of de-icing salts or from the natural environment. Evidence of corrosion initiated by de-icing salts in post-tensioned grouted duct bridges caused major concern in the 1980s and 1990s leading to a four-year moratorium on such techniques by the Highways Agency in the United Kingdom. In the marine environment the tidal and splash zones are recognised as high corrosion risk areas as illustrated in Figure 6.1. Inland structures near coastal regions can also deteriorate through corrosion induced by windblown salts. Concrete permanently submerged in seawater may allow significant chloride penetration but significant corrosion may not occur due to the low level of oxygen supply. Structures such as tunnels for rail and road traffic may be exposed to saline groundwaters of similar characteristics to seawater. In highway structures drainage paths or leaking joints may channel runoff in a manner that exposes localised areas to high concentrations of chloride. Chloride-induced damage has also affected building structures. Car park ramps and decks have deteriorated due to water, laden with de-icing salts, dripping from cars. Building façades have prematurely deteriorated due to corrosion induced by calcium chloride used as an accelerator in precast work. Kropp (1995) notes that chlorides may also be introduced in fire-damaged structures through exposure to thermally decomposed polyvinyl chloride fittings and furnishings. Many buildings contain PVC and it liberates hydrochloric gas at temperatures in excess of 80°C. Structures subject to wetting and drying cycles created by weather or tidal patterns are particularly vulnerable to high chloride uptake. There has been a slow realisation that specifications based on prescriptive advice in codes of practice may have underestimated the requirements for durable concrete in chloride environments. Particular attention has been drawn (Bamforth 1994, Grantham 1999) to the potential inadequacy of

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Figure 6.1 Corrosion-induced damage in a tidal zone

Portland cement concrete to provide the required level of durability over a typical service life and this has generated an interesting debate. The passivity of reinforcement is dependent on the stability of the passive film formed on it when the steel is immersed in the alkaline environment of fresh concrete. The passive film is rendered ineffective in circumstances where the chloride level in the surrounding concrete exceeds a critical level. A low level of chloride can be present and tolerated in durable concrete. Chlorides can inadvertently be introduced as trace elements in, for example, brackish water, aggregates and admixtures. Excessive levels, however, may occur over the service life if significant external sources of chloride are available. The internal sources of chloride are currently limited to tolerable levels by specification. For example, European Standard EN 206–1 limits the chloride content in a range from 0.10 to 0.40 per cent in the case of concrete containing reinforcement. The strictest limitations apply to prestressed concrete. Aggregate standards also limit the chloride content of aggregate for use in concrete. The use of seawater, chloride-bearing aggregates or admixtures such as calcium chloride accelerator is thus strictly controlled. Chloride-induced corrosion is generally focused on a small area, which forms a pit surrounded by uncorroded reinforcement. The process is illustrated in Figure 6.2. This can lead to rapid loss of cross-section and

Chloride ingress 103

Figure 6.2 Process of pitting corrosion in a chloride-rich environment

critically reduce the load-bearing capacity of the reinforced concrete member. Deterioration could be significantly advanced before damage is apparent at the surface. The mechanism of attack is described more extensively in Chapter 4. During the corrosion cycle ferrous ions become available to combine with the chloride ions to form compounds such as ferrous chloride (FeCl2). Hydrolysis of these products over time releases chloride ions with a consequent reduction in the anode pH. The corrosion rate increases because the oxidation of the iron is encouraged in such acidic conditions. The recycled chloride ions exacerbate the problem. An influencing parameter is the chloride ion:hydroxyl ion ratio at the reinforcement. It has been found (Tritthart 1989) that the higher the hydroxyl ion concentration the greater the fraction of total chlorides represented by free chlorides. If the chloride ions predominate the loss of ferrous (Fe2) ions is accelerated. If the OH ions predominate precipitation of FeOH ions help to repair the reinforcement’s passive oxide film. The reduction in pH discourages precipitation of the corrosion product. The pit at the anode may develop rapidly due to the combination of a localised anode with a comparatively large cathode. An insight into the influence of chloride level on the electrochemistry of the situation may be gained from the departure from a passive condition as illustrated in the Evans Diagram presented in Figure 6.3. The possible mechanism of chloride ion interaction with the reinforcement and passive layer has not been fully resolved. The studies of Foley, published in 1970 and 1975, are often quoted, as for example by the American Concrete Institute Committee 222 (1985) and by Sagoe-Crentsil and Glasser (1989).

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Figure 6.3 Evans Diagram showing departure from a passive condition due to the presence of chloride

Three theories have been advanced – adsorption theory, oxide film theory, and transitory complex theory. The adsorption theory postulates that chloride ions are adsorbed onto the surface of the reinforcement in preference to dissolved oxygen and hydroxyl ions. The reaction rate of iron with the chlorides is higher and soluble complexes are formed. The resultant dissolution promotes formation of a pit. The oxide film theory suggests that the passive oxide layer is more open to penetration by chloride ions than other anions. Defects and pores in the film are thought to allow chloride ions access to the reinforcement leading to pitting corrosion. The transitory complex theory postulates that a soluble complex of iron chlorides forms from the chloride ions and ferrous ions in competition with the ferrous hydroxide reaction. The diffusion of the iron chloride complex away from the anode is thought to destroy the passive film. The complex later breaks down with precipitation of iron hydroxide liberating the chloride ion. The chloride ion is then free to recommence the cycle. Chlorides may be present in three states: free chloride ions in the pore solution, chlorides strongly bound, and chlorides loosely bound. The free chlorides are the most significant contributors to the corrosion risk. They may be introduced from an external source or carbonation may release bound chlorides from internal sources. The aluminates are able to combine with

Chloride ingress 105 internally or externally introduced chlorides up to a certain concentration of chloride, and prevent them remaining as dangerous free chlorides. However the lowering of pH by carbonation releases some of these bound chlorides ions into the pore solution. The strongly bound chlorides are chemically combined, mainly as calcium chloroaluminates (3CaO.Al2O3.CaCl2.10H2O) and calcium chloroferrites (3CaO.Fe2O3.CaCl2.10H2O) through the calcium aluminate hydrates (for example 3CaO.Al2O3.6H2O). There is a two-stage reaction involving the production of calcium chloride from calcium hydroxide and sodium chloride. The loosely bound chlorides are adsorbed by the pore walls made up from calcium silicate hydrate such as 3CaO.2SiO2.3H2O. The threshold level of chloride for corrosion is not fully established because the critical level at which passivity is lost and corrosion commences is difficult to define. The reasons for this include the influence of environment but measurement also causes difficulty. The nature of the chlorides is significant but while free chlorides are most relevant it is the total (free and bound) chloride level that is detected in tests. In addition the chloride ion to hydroxyl ion ratio is an influencing factor governing corrosion activity but it is not readily determinable. The method of measurement (acid or water soluble) and expression (mass by weight of cement or concrete) can also make comparisons difficult. Thresholds quoted are therefore generally single point values of total chloride, not merely the harmful free chlorides. This gives compatibility with the most common analytical techniques – generally based on acid-soluble chloride. Total acid-soluble chloride level, expressed as a percentage by mass of cement, is traditionally used in specifications.

Detection and expression of chloride levels Tests for chloride level are generally based on acid-soluble chloride techniques that yield a value for total chlorides. Tests for water-soluble chlorides are referred to less often, despite the fact that water-soluble chloride levels better describe the presence of free chlorides. Results are expressed as single point values of percentage chloride by weight of cement. Effective chloride diffusion coefficients may be determined if sufficient single point values are determined for various depths. Dust samples obtained by drilling concrete surfaces are commonly analysed for total chloride content from which chloride diffusivity may be determined. The method is suitable for tests on in-situ concrete as well as laboratory-based research. Methods are also available for determining chloride diffusivity directly by laboratory tests involving specially prepared concrete discs. Determination of free chloride content could be achieved by chemical analysis of pore fluid but the sampling technique requires extremely high pressure and the required apparatus is not commonly available. It should be noted that results can be expressed as percentage chloride by weight of concrete in certain cases. This may arise in the appraisal of existing

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structures in instances where the cement content may not be known with certainty. An estimate of the percentage chloride by weight of cement may be made by multiplying by an appropriate factor. A factor between 6.5 and 7.0 is typically used for concretes with cement contents of the order of 300 kg/m3. The factor drops to about 5.0 in the case of cement contents of the order of 400 kg/m3. Sampling Dust samples may be obtained by drilling the member face or by drilling the side face of cores recovered from the member face. Tests on broken off pieces are also possible but not recommended. The most common practice is to recover dust samples by drilling perpendicular to a member face. The sample is representative of the depth band covered. The specific bands selected may be related to the depth of cover but intervals commonly used are 0–10 mm, 10–25 mm, 25–50 mm, 50–100 mm and 150–200 mm. The dust from the first 5 mm depth interval, measured from an exposed surface, should be discarded due to loss of chloride through rainwater washout. An even greater depth may have to be discounted if carbonation is deep. This is because carbonation releases bound chlorides, and a wave of free chlorides which migrate inwards in advance of the carbonation front may produce unusually high readings in a particular depth band. Deep depth bands can be used to establish a background chloride level, if they are beyond the zone of inwardly diffusing chlorides. In the case of investigation of compliance with limits on chloride level between ten and twenty samples are required if the chloride levels are close to the limits allowable. Concrete Society Technical Report No. 32 (1989) recommends use of a drill bit of 10 mm diameter minimum and 25 mm maximum. Several holes should be drilled per location to produce a representative sample of at least 30 g. Many test agencies use a 20 mm diameter drill bit and a minimum of three holes per location to produce a combined representative sample of at least 25 g per depth interval. Smaller sample sizes could be misleading if they were predominantly made up of dust derived from the drilling of aggregate. Cores may be recovered from members and dust drillings taken at specific depths via the side face of the core. It should be noted however that some loss of water-soluble salts and redistribution of chlorides can be caused by the use of lubricating water on the drill bit of the core driller. The effect is more significant if small diameter cores are used. It is possible, although not desirable, to conduct tests on pieces broken off an element. Concrete recovered by this method should be broken into lumps and crushed to produce a 25 g sample from at least a 50 g piece. It is clearly not a preferred method as it is difficult to determine the depth bands with accuracy. Tests on spalled pieces are not recommended due to the likelihood of carbonation, thus presenting a chloride profile that may not be representative of the structure as a whole.

Chloride ingress 107 Measurement Tests for chloride level are generally based on acid-soluble chloride techniques. This yields a value for total chlorides that overestimates the durability threat from free chlorides. Nevertheless it should be noted that bound chlorides can be released in certain circumstances and that specification clauses that impose limits on chloride content are usually based on total chlorides for reasons of practicality. Furthermore Tuuti (1982) found that the free chloride content can be proportional to the total chloride content. Chloride content can be determined in a number of ways. The Volhardt titration method using silver nitrate is often used, as described for example in British Standard BS 1881: Part 124 (Section 10.2), but alternative techniques are available. A method for determination of chloride ions by potentiometry could be used. The sophisticated technique of X-ray fluorescence spectrometry is advantageous in that the time required for analysis is short. It also allows cement content to be determined from the same sample. Furthermore, repeat tests are possible. However the method is comparative and so standard samples must be available with known chloride contents. Rapid detection methods An alternative to drilling dust samples on site for later analysis in a laboratory is the use of tests for the rapid determination of chloride level. Such tests trade off accuracy for speed. The techniques are useful for a quick check on specification compliance but the level of accuracy would probably rule out reliance on the test in borderline situations. A number of commercially produced kits for rapid determination of chloride level are on the market. In the case of one kit a 5 g sample is treated with nitric acid and anhydrous sodium carbonate at room temperature. Chloride concentration is determined using a proprietary test strip. The method is thus quick and inexpensive. Another commercially available kit involves treating a 5 g sample with nitric acid and sodium bicarbonate. Silver nitrate is added drop-by-drop until a colour change is noted. Chloride concentration is determined by recording the number of drops required. The accuracy of results with these kits is not claimed to be as good as the titration methods. The benefits are speed and low cost – commendable features of screening tests.

Critical chloride level for corrosion The depassivation of reinforcement in concrete by chloride ions is in competition with the repair of the passive oxide film by the hydroxyl ions. This has led to the concept of a critical chloride threshold level. The critical level at which passivity is lost and corrosion commences is not clearly defined and significant variations exist in the literature on the chloride level which was found to initiate corrosion. Therefore the critical amount of chlorides

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necessary to initiate corrosion has been a matter of much debate and consideration has been given to the factors that may influence it in any given situation. The onset of corrosion and therefore the critical chloride threshold level is dependent on several factors, including the following: • • • • • • •

the chemistry of the binder, especially the C3A content the proportion of the total chlorides that are free chlorides chloride ion to hydroxyl ion ratio the water/binder ratio the hydroxyl ion concentration temperature and relative humidity electrical potential of the reinforcement.

The interaction of these factors is such that the critical chloride content has not been uniquely determined due to its dependence on the properties of a particular concrete and its service environment. For example there is some conflict on the influence of pH. However it has been noted that reducing the water/binder ratio increases both the corrosion threshold and the pH whereas the addition of silica fume has the opposite effect in each case. Nevertheless there is currently a large amount of published data on suggested values for the chloride ion content below which there is an acceptably low risk of corrosion of the reinforcement. A review of the literature reveals a trend of agreement that 0.4 per cent chloride by mass of cement represents an acceptable benchmark for design purposes. Everett and Treadaway (1980) suggested the following descriptions of corrosion risk and these were adopted in Building Research Establishment (1982) Digest 264 and Concrete Society (1984) Technical Report No. 26: • • •

less than 0.4 per cent chloride by mass of cement: low risk; 0.4 to 1.0 per cent chloride by mass of cement: medium risk; greater than 1.0 per cent chloride by mass of cement: high risk. A similar set of criteria for corrosion risk was used by Browne (1980):

• • • •

less than 0.4 per cent chloride by mass of cement: negligible risk; 0.4 to 1.0 per cent chloride by mass of cement: possible risk; 1.0 to 2.0 per cent chloride by mass of cement: probable risk; greater than 2.0 per cent chloride by mass of cement: certain risk.

Codes have traditionally indicated total chloride levels which should not be exceeded in fresh concrete. These levels have been set conservatively based particularly on failures to building façades and prestressed structures by chloride-induced corrosion. A review of some national standards and European Standard EN 206–1, for example, reveals values in a range from 0.1

Chloride ingress 109 to 0.4 per cent chloride by mass of binder including the following examples: 0.4 per cent in the case of reinforced concrete, 0.2 per cent where sulfate resisting Portland cement is used, 0.1 per cent for prestressed and heat cured concrete. The lower value in the case of sulfate resisting Portland cement reflects the lower level of tricalcium aluminate available to bind the chlorides and the fact that if sulfate attack occurs the free chloride ion levels increase due to the breakdown of calcium chloroaluminates. It should be emphasised that these are limits on the total chloride content of new works, not the free chloride content, and therefore do not represent the corrosion threshold as such, which would be a higher value. Buenfeld (1986) stated that even 0.2 per cent chloride by mass of cement can lead to corrosion if the chlorides are being introduced from an external source and a large proportion remain free in the pore solution. This assertion, together with the typical code limits above is in agreement with the Concrete Society (1989) report, which recommends investigation for reinforcement corrosion if the measured chloride content is greater than 0.6 per cent by mass of cement. Extensive durability failures in highway structures subjected to de-icing salt application led American Concrete Institute Committee 222 (1985) to take a very conservative approach in framing recommendations. They suggested levels of 0.20 per cent for reinforced concrete and 0.08 per cent for prestressed concrete. The ACI committee decided that the large amount of conflicting data on the corrosion threshold values and the difficulty of defining the service environment throughout the life of a structure necessitated such conservative values. The literature includes several comprehensive reviews of published critical chloride threshold levels. A wide range of values exists from a minimum of 0.06 per cent to a maximum of 2.5 per cent by weight of cement. Pettersson (1992) presented a review based on six published studies. The range was from 0.06 per cent to 2.5 per cent. It was noted that different properties of the concrete and its service environment influenced the range. Variables included pH, cement type, curing regime, water/binder ratio and the use of admixtures. Bamforth (1996) published a review derived from over twenty sources including Funahashi’s review of seven studies in 1990. The values for critical chloride threshold levels varied from 0.06 per cent to 2.2 per cent by weight of cement. The wide range was partially accounted for by different methods of test. Glass and Buenfeld (1995) noted a range of published threshold values of 0.17 to 0.7 per cent for field exposure tests and 0.4 to 2.5 per cent for laboratory-based trials. They suggest that determination of a unique chloride threshold level applicable to a wide range of structures is not achievable. The range of critical chloride threshold levels published at present covers a wide range and the sensitivity of the influencing parameters is not clear. Glass and Buenfeld (1995) suggest that it may be beneficial to consider the critical chloride threshold solely as an indicator of corrosion risk. The results of a study of bridges in the United Kingdom showed that a chloride content between 0.35 and 0.5 per cent by weight of cement gives a corrosion risk of

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Figure 6.4 Typical critical chloride content for good quality uncarbonated concrete

below 25 per cent. Chloride contents in excess of 1.0 per cent give a corrosion risk of over 70 per cent. They concluded that without further work, no improvement could be made to the suggested chloride threshold levels of 0.4 per cent for buildings exposed to a temperate European climate and 0.2 per cent for structures exposed to a more aggressive environment. Equally the Concrete Society (1996) stated that while the threshold level of chloride can lie between 0.17 and 2.5 per cent the best that can be done at present is to use criteria in the 0.2 to 0.4 per cent range. A unique threshold level of chloride for corrosion is thus not established, however a possible relationship between critical chloride content, relative humidity and concrete quality has been published by the Comite EuroInternational du Beton (1989). This relationship was used as a guide in the design of the Western Scheldt Tunnel in the Netherlands (Breitenbucher et al. 1999), in what may have been the first use of mathematical models in a durability design-based specification. The trend of the relationship is indicated in Figure 6.4. The precise elevation of the curve depends on the quality of the concrete and the extent of carbonation. The better the concrete quality the higher the critical chloride level. Carbonation of the concrete lowers the critical chloride level. The range of values covered by the graph in the CEB (1989) guide shows that a critical chloride level of 0.4 per cent chloride by mass of cement for reinforced concrete is a suitable design level for good quality uncarbonated concrete.


Primary factors influencing chloride ingress The rate of chloride ingress through the cover to reinforcement depends primarily on the following material and environmental factors: • • • • • • • • •

chloride diffusivity of the concrete; sorptivity of the concrete; ability of concrete to bind chlorides; water/cement ratio; chloride diffusivity of the aggregate; degree of exposure to chloride source; temperature; carbonation; hydrostatic head (if applicable).

Chloride diffusivity The chloride diffusivity of concrete is clearly a key issue in its ability to endure in chloride-rich environments. The uptake of chlorides at the surface is primarily a function of sorptivity but once inside the concrete the transport of the chloride ions will be by diffusion. Diffusion occurs in pores that are totally or partially water filled. Potential durability is therefore often characterised by a diffusion coefficient or effective diffusion coefficient. Bamforth (1994) drew particular attention to the shortcomings of Portland cement concretes in the chloride environment through estimation of typical chloride diffusivity using Fick’s second law. He applied the error function solution by Crank (1975) to chloride concentration profiles generated from data derived through surveys of structures in service. Blended cement concretes performed well but the effective diffusion coefficients for Portland cement concretes were found to be at a level that would not afford protection when used with practical levels of cover. This assertion was challenged by Spooner (1995) on a number of grounds. While the ability of normal Portland cement concretes to consistently provide durability in chloride environments is a source of debate, the potentially enhanced performance of secondary cementitious materials, reported earlier by, for example, Higgins (1986), is noteworthy. The decrease in chloride ion diffusion with age is more significant in the case of pfa and slag concretes than in the case of Portland cement concretes. Fookes (1995) noted that in the severest microclimates blended cements were necessary otherwise a cover of over 200 mm would be required! In the case of very high chloride-level exposure conditions specifiers may also consider the use of supplementary materials such as microsilica or metakaolin. Sorptivity The transport of chloride ions into concrete has a capillary suction component in addition to diffusion. Capillary suction of water containing

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chlorides occurs into air-filled pores near the surface zone. The effect of absorption is significant. Concrete subject to wetting and drying can experience a greater uptake of chlorides than concrete subject to diffusion alone. The salt-laden water can be absorbed by the initially dry pores. The pores near the surface may become saturated. A drying cycle will evaporate the water leaving the chlorides behind which can then diffuse inwards under the influence of the concentration difference. The surface pores are then available for further chloride absorption. A reduction in surface absorption can significantly enhance service life by lowering the uptake of chlorides. A technique which may achieve this is the use of controlled permeability formwork (CPF). Such formwork systems allow air and water to escape from the concrete under controlled conditions without loss of cement or aggregate particles. This improves the quality of the surface and the performance of the outer layer of concrete by significantly reducing the water/cement ratio. Bamforth and Price (1993) report that chloride ingress reduction through the use of CPF could at least double the service life. Ability of concrete to bind chlorides The threat to durability is posed by free chlorides. The greater the binding capacity of a concrete in respect of chlorides the better it should perform. The ability of concretes to bind chlorides depends on the alkalinity of the binder, its fineness and on the tricalcium aluminate (C3A) and tetracalcium aluminoferrite (C4AF) content. The aluminates are very effective in binding chloride ions introduced at an early stage. They are not as readily available at later ages to bind externally introduced chlorides because they will by then have reacted with other ions. Typically the C3A content would be of the order of 7 to 12 per cent in Portland cement. Sulfate-resisting cements contain less than 3.5 per cent tricalcium aluminate and should therefore be less effective although this has been questioned by Arya et al. (1990). The alumina content of pulverised fuel ash (pfa) and slag (ggbs) is considerable and this conveys benefits in respect of chloride binding. However for all binders the influence of C3A content becomes less significant as chloride concentration increases. The better performance of pfa concrete in comparison with Portland cement concrete may also be due to lower chloride diffusivity. It is thought that pfa concretes may have a more extensive capillary network, which allows greater absorption of chlorides, but lower diffusivity ultimately leads to less chloride at the reinforcement. Harrisson (1995) ranked the resistance to chloride penetration of concretes made with different cements as decreasing in the following sequence: Portland cement/ggbs; Portland cement/pfa; Portland cement; sulfate-resisting Portland cement. The beneficial effect of microsilica was also noted. Pettersson (1992) also demonstrated the beneficial effect of microsilica – although the critical chloride concentration decreased in microsilica concretes, the corrosion rate was lower leading to a net beneficial effect.

Chloride ingress 113 Resistance to carbonation is also important in this context because carbonation can lead to the release of bound chlorides. Water/cement ratio Given the traditional link between water/cement ratio, permeability and hence durability it is unsurprising that chloride diffusion through high water/cement ratio concretes is more rapid than for dense low water/cement ratio concretes. Hobbs and Matthews (1998a) studied the relationship between effective diffusion coefficient and water/cement ratio using published data from ten studies of CEM I concretes. The coefficients were standardised through the Arrhenius equation to allow for temperature variation. The best-fit equation was determined as follows: Dce 0.04(1166w/c) 1012 Thus the very significant influence of water/cement ratio is apparent. Chloride diffusivity of the aggregate The rate of chloride ion ingress into concrete is influenced by both the diffusivity of the cement paste fraction and by the diffusivity of the aggregate. Hobbs (1999a) drew attention to the chloride diffusivity of aggregate and noted that consideration would need to be given to the aggregate volume in certain circumstances. Aggregates can have chloride diffusivities in excess of or lower than the cement paste fraction and so the aggregate volume may need to be limited or increased to achieve the desired limit on chloride diffusivity. Controls may need to be considered to ensure that the aggregate selected for a structure in a chloride environment does not have a detrimental impact on the required chloride resistance of the concrete. Hobbs (1999a) states that highly permeable aggregate could increase the chloride ion diffusivity of concrete by a factor of ten. Degree of exposure to chloride source Detrimental amounts of chloride may be expected to penetrate into high quality concrete if the conditions of exposure are such that chloride-laden water is frequently in contact with the structure. Damage in bridges tends to be concentrated in locations where water containing de-icing salt drains or leaks. Pritchard (1986) suggests a range of details designed to minimise the problem. These include deck continuity to avoid joints, accessible deck joints, deck drainage and drips. Member geometry may also influence durability. Columns are less durable than walls and beams are less durable than slabs. Presumably this is due to the differences between biaxial penetration and, effectively, uniaxial penetration.

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Temperature The resistance of concrete to chloride ingress is temperature related. Dhir et al. (1993) report that the resistance of Portland cement concrete is dramatically less at 45°C than at 5°C. The diffusion coefficient is influenced by temperature and can double when average temperatures climb from about 10°C to 20°C. In addition such seasonal temperature variations may lead to bound chlorides taken up in the winter being released as free chlorides in the summer. Carbonation Carbonation releases bound chlorides thus making free chloride ions available to ingress into the uncarbonated zones. This can increase the harmful free chlorides ion concentration in the cover zone and may bring the level at the reinforcement to a critical level. Hydrostatic head Parts of structures subject to a hydrostatic head in a chloride source will suffer a further force driving chlorides into the cover concrete. Such conditions most commonly occur in sections of structures submerged in seawater.

Mathematical modelling of chloride ingress Chloride ingress from the external environment occurs by diffusion and by capillary suction. In the early stages of exposure chlorides are transported into concrete by absorption. The absorption effect may reduce with time unless the concrete is subject to wetting and drying. Mathematical models of chloride ingress currently being developed are primarily based on chloride diffusion although attempts have been made to take absorption into account. The following review illustrates the variety of approaches to modelling chloride ingress that could be used as starting points in the development of service life prediction tools and performance-based specifications. Fickian model with apparent diffusion coefficient Models based on consideration of diffusion alone are constructed around Fick’s second law of diffusion and the error function solution by Crank. Fick’s second law of diffusion concerns the rate of change of concentration with respect to time. It may be stated as follows for diffusion in a semi-infinite, homogenous medium, where the diffusion coefficient (D) is independent of the dependent and independent variables: C 2C D 2 t x

Chloride ingress 115 with boundary conditions of: Cx 0 at t 0 and 0 x Cx Cs at x 0 and 0 t where Cx chloride concentration at depth x at time t Cs surface chloride concentration Crank’s solution of Fick’s second law of diffusion can be stated as follows, using an apparent diffusion coefficient: Cx x 1 erf Cs 2(Dca t)0.5 where Dca apparent diffusion coefficient t time of exposure erf error function Many models are based on Fick’s second law of diffusion and on the simplified assumption that the diffusion coefficient is constant. The determination of a diffusion coefficient in an existing structure involves a number of stages. First the concrete is sampled at a series of depths to determine the chloride levels. The chloride levels are determined by laboratory analysis. It is best if the final depth of sampling is such that the background chloride level is detectable. This background level is deducted from the other values and a concentration gradient is plotted of the chloride that has diffused into the concrete. Figure 6.5 illustrates such a plot. It may be noted that the surface chloride level is depressed due to the effect of salt being washed out in rainwater. In concrete structures such effects can lead to a variation in the surface chloride concentration (Cs) and therefore a notional surface chloride concentration is used (Csn ). This can be achieved by extrapolation. Knowing the age of the structure, a best-fit curve may be fitted and the value of the apparent diffusion coefficient may be determined. This determination of apparent diffusion coefficient may conveniently be carried out on spreadsheet packages that include the error function. It may be found that problems can arise from the combination of very small numbers (for example, apparent diffusion coefficients of the order of 1012 m2/s) and very large numbers (for example, exposure periods of the order of 108 seconds). The difficulty can be overcome by specifying the period in years and converting the apparent diffusion coefficient to the required units at the end of the calculation. Some studies have questioned the validity of predictions based on the simple diffusion model in the case of chloride ingress. Chatterji (1995) argues that a model solely based on Fick’s second law is questionable due to the ionic nature of chlorides. It is also argued that the diffusion model does not take account of the transport of chlorides by absorption and that this effect

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Figure 6.5 Best fit curve based on Crank’s error function solution to Fick’s second law of diffusion

reduces with time. Furthermore, reliance on the measurement of total chloride levels within a concrete as an indicator of future corrosion risk may be questioned for the following reasons: • • • • •

chloride diffusion values for a given concrete may not be constant over time, decreasing with continued hydration; diffusion rates may vary with depth from the surface; surface chloride concentrations increase with time if the concrete surface is subject to wetting and drying cycles; the ability of different cementitious binders to bind chlorides has not been fully researched; profiles based on results from accelerated laboratory tests do not correlate well with behaviour of concrete structures in practice.

Chloride ingress 117 Nevertheless it is generally accepted that, while not perfect, the Fickian model provides the best starting point for the first generation of predictive models. The initial and boundary conditions are capable of being fulfilled in laboratory trials but deviations in real structures are significant. Research is continuing to enhance the model for application in practice. For rapid comparative studies it may be useful to note that formulae based on Fick’s second law of diffusion lead to a square root of time relationship with depth of penetration. Element modelling During the 1980s Schiessl examined the use of finite difference models to take account of the criticisms of using a fixed value of diffusion coefficient alone without allowing for other effects. Later Bentz et al. (1996) developed a finite difference model to examine the influence of pfa in reducing chloride ingress in the case of a marine tidal exposure zone. The model uses a direct finite difference implementation of Fick’s second law of diffusion and corrects for the effects of initial sorption, changing surface concentration with time and chloride binding by the hydrates. It was designed to model a specific case of pfa concrete in a marine tidal exposure condition. It was found to be an improvement over the use of Crank’s solution, particularly in the modelling of low diffusivity, high pfa content concretes. One and two-dimensional penetration was studied. The following approximation of Fick’s second law was achieved, in the case of the one-dimensional model, assuming central difference form for spatial derivatives: C c 2ci ci1 D i1 t x2 Temporal derivatives and spatial derivatives were determined and a linear set of equations was set up. These were solved for each time step and a concentration profile was determined. The initial condition was set by evaluating experimentally the chloride concentration at 28 days due to sorption. The background level was taken to be 0.1 per cent chloride by mass of cement and this value was set elsewhere in the model. The boundary condition was a chloride concentration that increased linearly with time. The first value was based on the 28-day result that accounted for the initial sorption effect. Later values were determined by measurement. The point of reference was midway through the second depth interval sampled. This avoided the problems associated with the variable chloride concentrations that can occur close to the surface. The time stepping was started using a number of different diffusion coefficients and the output was compared with profiles that had been established with data from a marine exposure site at one, two, and four years’

118


exposure. The diffusion coefficient with the least squares of differences was selected as the best fit. Onyejekwe and Reddy (2000) proposed a Green element method analysis. Fick’s second law is used as the governing partial differential equation in the following form: D

2C c 0 x2 t

x0 x xL

together with the free space Green’s function of: 2G (x xi ) x2 where (x xi ) is the Dirac delta function to yield the simplified equation:

(

c(xi , t) c

G c G x x

)

xxL xx0

xx0L G

( c t ) dx 0

The variation of concentration within each element of the model was described by interpolation functions, which for the eth element are: e1 () 1 and

e2 ()

where is a local co-ordinate with its origin at node 1 of the element. The dependent variable c and c/ x are specified through summation of the relevant interpolation functions. These equations are used to develop an equation that is discretised at each node. The authors demonstrated good correlation with published experimental results including the consideration of time dependent effects on the value of the diffusion coefficient. Model based on Nernst-Planck equation Nagesh and Bishwajit (1998) addressed the issue of modelling the non-linear chloride diffusion process by developing a model using Nernst-Planck equations. The model addressed the significance of adsorption in unsaturated concrete and the influence of movement on dissociated ions by the electrostatic fields created by other ions. For unsaturated conditions the chloride ingress involves both capillary suction and diffusion. The authors demonstrated that: J Jc Jd and

Chloride ingress 119 Jc Cs D(Cs , s) Jd Dd (Cv)

ds dx

dCv dx

where J total flux Jc flux of mass of chloride by capillary suction Jd equivalent ionic diffusion flux of chloride Cs , Cv free chloride concentrations s solution saturation of concrete Dd chloride diffusion coefficient in concrete assuming it to be concentration dependent only Replacing Cs and Cv (kg/m3) by Cm (kilogrammes of free chloride per kilogramme of concrete) and replacing s (volumetric solution content) by sm (kilogrammes of solution per kilogramme of concrete) yields a total flux of: J

[{D(C

m

sm) c

} {

}]

Cm dsm dC c Dd (Cm) m sm dx dx

where c density With dJ d(CT) dx dt and CT c Cm where CT is the total concentration of bound and free chlorides and is a factor dependent on the porosity of the concrete and the solution saturation, then:

[

dCm d Cm d dCm D(Cm , sm) sm Dd (Cm) dt dx sm dx dx

]

Boltzmann’s Transformation was applied to reduce the partial differential equation to an ordinary differential equation in terms of a new variable and yielded: d Cm d dC dCm D(Cm , sm) sm Dd (Cm) m 0 d sm d d 2 d

[

where x t 0.5

] [

]

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The diffusion coefficient could then be determined using the boundary conditions of equal to infinity and Cm equal to the initial free chloride concentration in concrete (C1). An approximate finite difference form of the above equation was formulated from which values of the chloride diffusion coefficient could be determined. The final form of the equation being:

[{D (C ) dCd } m

d

{Dd (Cm)}i

m

(i2)

{A}

(i1) (i1) (Ci C(i2)) 2

]

{dCd } m

i

The authors published formulae for saturated conditions in addition to the unsaturated condition. Experimental programmes are reported in the paper on the determination of the solution content profiles, the relationship between solution concentration and , and the relationship between Cm and . Model for ageing concrete Roelfstra et al. (1996) presented a numerical model for chloride ion penetration in concrete structures with particular reference to the effect of moisture migration. The model is specifically intended for application in the case of ageing concrete. The model considers ionic diffusion and the effect of convection by means of moisture migration and cement hydration. This model is a development of a chloride diffusion model by Seatta et al. (1993) and incorporates earlier research by Roelfstra on modelling hydration processes. A system of equations is solved for combined heat and moisture transport. The chloride ion penetration is then computed through a formula of the form: b,t w e (b 1) w,t e b w e,t Dc w ,xx e v,x e 0 where w moisture content e free chloride ion concentration coefficient Dc diffusion coefficient for chloride ions v moisture flux ,x nabla operator (spatial gradients) and b

Ct 1 (1 p) Cf

where Ct total chloride ion concentration Cf free chloride ion concentration

Chloride ingress 121 p concrete porosity ratio between free and physically bound chloride ions.

Application of models to service life prediction Service life prediction that involves modelling the rate of chloride ingress is complicated by the variety of complex mechanisms involved in both the uptake of chlorides from external sources and the processes that influence penetration through the cover. Nevertheless it is recognised that the best parameter for models currently available is the apparent chloride diffusion coefficient based on Crank’s solution to Fick’s second law of diffusion: Cx x 1 erf Cs 2(Dca t)0.5 This is particularly so in the case of concretes which are wet most of the time, such as marine and coastal structures. Despite its shortcomings the solution by Crank allows prediction of future chloride profiles from which a design estimate may be made of the time taken for the critical chloride threshold to build up at the level of the reinforcement. It is recognised that the formula does not perfectly model the case of chloride ingress into concrete but it can be usefully employed in an empirical manner. The input parameters can be derived from data gained from structures in service and, perhaps, an additional allowance of about 10 mm can be added to the calculated cover to allow for the additional effect of absorption. The input parameters required, other than the design cover, are notional surface chloride level, apparent diffusion coefficient and the corrosion threshold level. The main focus is on the time to depassivation for a specific cover. A simplified form of the equation for design is: t

x2 A Dca

where t time to depassivation x depth of cover A constant dependent on the surface chloride level (Cs) and the critical chloride corrosion threshold (Ccr) Dca diffusion coefficient The constant (A) is determined through Fick’s second law. Siemes et al. (1998) demonstrated the use of this equation with the following values of the constant, assuming a critical corrosion threshold of 1 per cent by weight of cement: A 1.8, for Cs 3 per cent (by weight of cement) A 2.65, for Cs 4 per cent.

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Notional surface chloride levels Use of Crank’s solution in design would involve the assumption that the surface chloride level remains constant over time. It has been found (Bamforth 1996) that this is a reasonable assumption using values of the order of those observed in practice shortly after exposure. Notional surface chloride levels can be determined either from extrapolated values in particular cases or from the literature. In the severe case of the tidal and splash zone, Helland (1999) reported that a Cs value of 0.9 per cent by weight of concrete is a reasonable design estimate. This would equate to a value of approximately 6.0 per cent by weight of cement. Bamforth (1996) proposed the somewhat lower value of 0.75 per cent on the basis of eight-year exposure trials. Design values of 0.50 and 0.75 per cent were proposed for the spray zone and marine atmosphere respectively. He also distinguished between values for Portland cement concretes and those containing pfa or slag. The values for Portland cement concretes were about 20 per cent lower than for blended cement concretes. Berke and Hicks (1993) cite a consultant’s report by Hartt in reporting typical values of surface concentration applicable to the marine splash and tidal zone. A value of 17.8 kg/m3 is quoted, which would equate to about 4 to 5 per cent by weight of cement for a typical structural concrete. In the case of car park decks similar values are quoted. Siemes et al. (1998) used Cs values of 3 and 4 per cent by weight of cement in the case of a tunnel exposed to saline groundwaters. The rate of chloride build-up in bridge decks is thought to be lower than that in the marine environment. Berke and Hicks (1993) quote a rate of 0.7 kg/m3 per year due to the effect of rain and the surface concentration builds up to about 14.8 kg/m3. This would equate to about 3.5 to 4.5 per cent by weight of cement for typical structural concretes. Surface concentrations of 2 to 4 per cent by weight of cement were found by Polder and Hug (2000) in a thirty-year old bridge subject to de-icing salt application at an annual dosage rate of about 250 grams of chloride per square metre. Apparent diffusion coefficient A review of a wide range of published values of diffusion coefficients determined for Portland cement concretes are of the order of 1012 m2/s. Values for pfa and slag concretes are generally lower by a factor of 10. For example Polder and Hug (2000) reported a value of 3.0 1012 m2/s for Portland cement concrete in a thirty-year-old bridge. Siemes et al. (1998) employed the following values of diffusion coefficient which they described as typical: 1.50 1012 m2/s for a dense Portland cement concrete; 0.75

1012 m2/s for a high slag content ggbs concrete; 0.3 1012 m2/s for a high quality ash pfa concrete. The value of diffusion coefficient decreases with time and therefore service

Chloride ingress 123 life prediction models may need to incorporate an age factor to take account of this. It is particularly significant in the case of pfa and slag concretes. The diffusion coefficient based on penetration tests or monitoring of structures in service would therefore need to be referenced to a value calibrated to a standard age at time of test. For example Maage et al. (1996) observed that the decrease in diffusion coefficient obeyed the mathematical expression: D(t) t 0 D0 t

()

with where D(t) diffusion coefficient at time t D0 reference diffusion coefficient at age t0 parameter determined by regression analysis of test results parameter representing the effect of continued hydration parameter representing the pore blocking effect at the surface layer of magnesium and potassium ion exchange between sea water and concrete The Brite-Euram project, DuraCrete, proposed a modification to the Crank solution of Fick’s second law to take account of both the age factor which effects diffusion coefficient and the other factors which differentiate the diffusion coefficient as measured and that which applies in practice over the full service life. The form of the equation may be presented as follows, charting the ingress of the critical chloride level:

(

x(t) 2 erf 1 1

C(Crit) CSN

) √k D t

RCM,0

ke kc

n

(tt ) t 0

where x(t) depth of chloride ingress at critical level C(Crit) critical chloride content CSN notional surface chloride level DRCM,0 chloride migration coefficient measured at reference time t0 kt, ke, kc constants to take account of method of test, environment and curing on the value of the effective chloride diffusion (D0) at time t0 n age exponent t exposure period The constant parameter kt takes account of the method of testing the diffusivity and is used to convert the test value into a diffusion coefficient. The constants ke and kc take account of the influence of environment and curing on the diffusion coefficient so that a value (D0) may be determined. The value

124


of D0 is then a calibrated figure representing the effective chloride diffusion coefficient at a reference time (t0) at defined compaction, curing, and environmental conditions. The age factor (n) takes account of the influence of age on the initially determined diffusion coefficient. An example of values used in a design case was presented by Breitenbucher et al. (1999), reporting on the specification for the Western Scheldt Tunnel in the Netherlands. Diffusivity was initially determined by a rapid chloride migration method (Gehlan and Ludwig 1999) at a reference time (t0) of 28 days. This was converted to a diffusion coefficient (D0) at a defined level of compaction, curing and environment by the application of factors with mean values of 0.85 (kt), 1.00 (ke), and 1.00 (kc). The mean value of the age exponent was 0.60. A strong dependence of the diffusion coefficient value on temperature was commented on by Berke and Hicks (1993). A 10°C increase in temperature results in an approximate doubling of the value. Thus diffusion coefficients determined in laboratory tests at room temperature (T1) may need to be corrected if the average yearly in-service temperature (T2) is markedly different. A correction based on an Arrhenius type equation and the NernstEinstein equation is of the form: D2 D1

(TT ) exp [k (T1 T1 )] 2

1

1

2

where k is the activation energy divided by the gas constant. It has been estimated to vary between 3850°K and 6000°K for concrete with water/ cement ratios of 0.6 to 0.4 respectively. It may also prove possible to determine a design value of diffusion coefficient based on the mix proportions. Possible relationships between diffusion coefficient and water/cement or water/binder ratio have been demonstrated by, for example, Berke and Hicks (1993) and by Hobbs and Matthews (1998a). Corrosion threshold value Values of corrosion threshold levels for design purposes would typically lie in the range applicable to the possible risk of corrosion. Thus values of 0.4 per cent to 1.0 per cent are typically used as design values. The CEB (1989) design guides can be used to determine a more precise value depending on relative humidity, concrete quality, and degree of carbonation.

Chlorides: limitations and exposure categories in EN 206–1 European standard EN 206–1 covers the durability of concrete from the viewpoint of both initial internal chlorides and the resistance to ingress of external chlorides. The threat of corrosion from external chlorides is considered under two sets of exposure classes: chlorides from seawater (XS)

Chloride ingress 125 and chlorides other than from seawater (XD). Clearly the matter of exposure to de-icing salts is the significant aspect of the second set. Cather (2000) indicates that the distinction originated in an effort to isolate chemical effects particular to seawater and the view that de-icing salts might be more aggressive than chlorides in seawater. Three exposure classes have been identified in each set: XS1, XS2, XS3 and XD1, XD2, XD3. The informative annex in EN 206–1 presents indicative limits on concrete composition (maximum water/cement ratio, minimum cement content, and minimum strength class) based on an intended working life of fifty years. The informative values are merely based on the mean of values currently representative of European practice for CEM I concretes and are not therefore necessarily appropriate for use in all regions, or for all intended working lives. They are informative but specifiers must consult the relevant limiting values in national annexes or complementary standards valid in the place of use. Initial internal chlorides The need for limits on the chloride level in the original mix is emphasised by Neville (1995) who cites Lambert, Page and Vassie in noting that the action of such chlorides is more aggressive than the action of the same amount of chloride ingressing from external sources. The limit on initial chlorides in European standard EN 206–1 limits the maximum total chloride level in fresh concrete to 1.0 per cent by mass of cement for concretes that do not contain steel reinforcement. Lower limits of 0.40 per cent and 0.20 per cent apply for concrete with steel reinforcement; lower limits of 0.20 per cent and 0.10 per cent apply in the case of concrete containing prestressing steel. The precise limits to be applied depend upon the provisions valid in the place of use. Typical limits are 0.40 per cent for concrete with steel reinforcement and 0.10 per cent for prestressed or heat cured concrete. The European limits on initial chlorides are generally higher than those applicable in the United States of America. However calcium chloride and chloride-based admixtures are not permitted by EN 206–1 where metal is embedded in concrete. The use of calcium chloride accelerator is permitted in the USA. External chlorides: exposure class XS1 Exposure class XS1 covers reinforced and prestressed concrete that is exposed to airborne salt from seawater but is not in direct contact with the seawater. The degree of exposure to chlorides may not be as significant as in the other chloride exposure classes, nevertheless the ever-present exposure to airborne salt must be taken into account considering the lengthy lifetime of a typical structure. Examples of such structures include those situated along the coast beyond the spray zone. The sub-group that drew up recommendations

126


in respect of chlorides for the United Kingdom complementary standard to EN 206–1 attempted to define the zone influenced by spray. Recognising that the value is a variable dependent on sea state, wind direction, and topography they nevertheless suggested that a distance of up to 100 m from the sea could be used as a guide. Thus the zone influenced by airborne salt may be regarded as extending 100 m from the sea to a line several kilometres inland. The informative annex to EN 206–1 shows that in the case of XS1 the mean values in European practice for CEM I concretes indicate a requirement for a minimum cement content of 300 kg/m3 and a maximum water/cement ratio of 0.50. The study by Hobbs and Matthews (1998a) concluded that a similar limit on maximum water/cement ratio would be appropriate. Slightly more generous limits on concrete composition may be found in national annexes and complementary standards, reflecting the balance to be achieved between caution and pragmatism. In relation to a longer intended working life Hobbs and Matthews (1998a) recommend increased cover rather than more onerous restrictions on mix composition. An indicative increase of 15 mm is tentatively proposed in BS 8500. External chlorides: exposure class XS2 Exposure class XS2 covers reinforced and prestressed concrete that is permanently submerged in seawater. The dominant chloride penetration mechanism will be diffusion but the effect of hydrostatic head may also be significant. Thus chloride ingress may be considerable leading to high chloride concentrations at the reinforcement. The durability threat is tempered by the likelihood that the amount of oxygen reaching the reinforcement will be low with correspondingly low rates of corrosion. Examples of concrete covered by the category include parts of marine and coastal structures that lie sufficiently below the lowest low water mark to remain submerged. The distance below the lowest low water mark should allow for the effects of wave action under predominant wind and sea state conditions. A distance of approximately one metre would probably be appropriate. Hobbs and Matthews (1998a) concluded that if all faces of the member were submerged the specification should be based on providing resistance to seawater attack rather than resistance to chloride-induced corrosion. They drew attention, however, to the different case of concrete exposed to sea water on one side only and suggest that this significant durability threat to the reinforcement be distinguished by a separate exposure classification, for example XS2B. A comparison of requirements for XS2 and XS1, through the informative annex to EN 206–1, shows a decrease in maximum water/cement ratio of 0.05 and an increase in minimum cement content of 20 kg/m3. Although the informative annex in EN 206–1 deals only with CEM I concretes, the benefits of pfa and slag mixes in chloride environments is worthy of consideration.

Chloride ingress 127 Advice on such alternative mixes is provided in some national documents, for example BS 8500 in the United Kingdom. External chlorides: exposure class XS3 Exposure class XS3 covers reinforced and prestressed concrete that is in the tidal and splash zone and the case of reinforced and prestressed concrete subject to seawater spray. The conditions covered by this exposure class must be recognised as those representing the greatest threat to the durability of reinforced concrete structures. The uptake of chlorides by capillary absorption can be considerable because of the effect of wetting and drying cycles. Diffusion can then lead to the development of high chloride concentrations at the reinforcement. Significant corrosion damage could occur at an early stage in the lifetime of the structure if the cover concrete is inadequately specified or executed. Examples of structures subject to these exposure conditions include marine and coastal structures located above the lowest low water mark, less an allowance for typical wave height under normal weather conditions, and those structures subject to spray. As noted earlier the spray zone may be considered to extend approximately 100 m from the sea. Recognising the significant threat in the XS3 class it is not surprising that the informative annex to EN 206–1 indicates a maximum water/cement ratio of 0.45 and the highest strength class of those referred to in the table. This would lead to high cement contents, the value of which would depend on provisions valid in the place of use. Given the grave concerns expressed about the adequacy of concrete in these environments the use of the lowest possible water/cement ratio is obviously prudent. The study by Hobbs (1999a) on the influence of aggregate on chloride resistance is also worthy of note in structures required to have a long service life in a chloride environment. The informative annex in EN 206–1 deals only with CEM I concretes but consideration should also be given to the enhanced resistance of concretes manufactured with Portland-slag cement, Portland-silica fume cement, Portland-fly ash cement, blastfurnace cement, and appropriate combinations of Portland cement and fly ash or ggbs. External chloride: exposure class XD1 Exposure class XD1 covers the case of reinforced and prestressed concrete of moderate humidity and in contact with airborne chlorides from sources other than seawater. Examples of concrete covered by the exposure class could include structures in proximity to de-iced highways leading to airborne chlorides generated from the spray associated with significant volumes of high-speed traffic. Hobbs and Matthews (1998a) suggest that such structures could be considered as those located more than ten metres from the highway. They consider that there is insufficient evidence of durability problems

128


caused by such airborne chlorides to warrant the specification of more than a nominal minimum quality of concrete. The informative annex to EN 206–1 shows mean values for XD1 in European practice for CEM I concretes to have a minimum cement content of 300 kg/m3 and a maximum water/cement ratio of 0.55. This is more onerous than that suggested by the Hobbs and Matthews (1998a) study but it is to be expected that standards documents will err on the cautious side. External chloride: exposure class XD2 Exposure class XD2 covers the case of reinforced and prestressed concrete that is wet, rarely dry, in contact with water containing chlorides from sources other than seawater. The dominant transport mechanism would thus be diffusion. The risk of corrosion is present since high chloride concentrations are possible but the rate of corrosion will be low if the amount of dissolved oxygen reaching the steel is low. An example of such a structure is a swimming pool which has semi-permanent contact with water containing chlorides, other than seawater. Such conditions could arise with parts of structures exposed to brackish groundwaters or in particular industrial situations. Given the assumption that the corrosion rate will be low the limiting composition parameters do not need to be onerous. The informative annex to EN 206–1 shows mean values for XD2 in European practice for CEM I concretes to be the same as for XD1. However, many national annexes and complementary standards, such as BS 8500, DIN 1045–2 and IS EN 206–1, reduce the water/cement ratio requirement by 0.05. External chlorides: exposure class XD3 Exposure class XD3 covers reinforced and prestressed concrete which is cyclically wet and dry, in contact with water containing chlorides from sources other than sea water. The effects of both capillary suction and diffusion will be experienced in such conditions. Release of bound chlorides through carbonation could also be a possibility in certain circumstances. Examples of such structures include parts of highway structures subject to de-icing salt application or spray from water containing de-icing salts. Concrete car park structures may also experience such exposure conditions from runoff if de-icing salt use on the adjoining roads is significant. The informative annex to EN 206–1 shows mean values for XD3 in European practice for CEM I concretes to be similar to the requirement for XS3 but with a 20 kg/m3 decrease in minimum cement content. A low value of maximum water/cement ratio, at 0.45, is understandably called for. The water/cement ratio requirement accords with the recommendations of Hobbs and Matthews (1998a) based on a number of assumptions, including a mean surface chloride concentration of 0.1 per cent by mass of concrete. Even lower water/cement ratio restrictions may be found in national guidance documents.


Specification by performance Despite its shortcomings, there is general recognition that the potential rate of chloride ingress can best be characterised by an apparent chloride diffusion coefficient. Performance-based specifications could employ such a coefficient as the relevant parameter. The development of suitable test methods for characterising chloride penetration is being pursued by RILEM for consideration by CEN. Approaches to specification of durability by performance are illustrated by the work of Dhir, Jones and Ahmed (1991b) for chloride exposure and, in a more general way, by Alexander et al. (1999). The former produced a nomogram for determining the performance requirement in respect of chloride resistance. The latter proposed a system based on durability indexes that take into account chloride resistance among other relevant properties. A software program, ‘Life-365’, is being developed to determine life cycle costs of concrete structures and the initial version deals with structures exposed to de-icing salts. It is being further developed by American Concrete Institute Committee 365. Chloride diffusion resistance The concept of specification by performance in respect of chloride resistance is illustrated in the nomogram of Dhir, Jones and Ahmed (1991b). The nomogram relates the intended service life, environmental chloride content, cover, chloride threshold value for corrosion and the required coefficient of chloride diffusion. The format of the nomogram is illustrated in Figure 6.6. The nomogram is based on Fick’s second law of diffusion. It relates diffusion coefficient (D) and the period required to reach a specific chloride concentration at the reinforcement (t1). The initial relationship is based on an assumption of a 50 mm cover, a surface chloride level of 0.5 M, and an initial chloride content of zero. The effect of different values of cover on the period may be assessed by use of the southeastern quadrant of the nomogram to determine the alternative period (t2). Should the effect of a different surface chloride level be of interest the southwestern quadrant can be used to determine the relevant period (t3). The northwestern quadrant includes a graph to yield a multiplication factor to be applied if the initial chloride content differs from the assumed value of zero. A nomogram was also employed by Bamforth (1994) to highlight the significant difference in performance of pfa and slag concretes in comparison with Portland cement concretes. Durability indexes including chloride resistance The approach proposed by Alexander et al. (1999) embraces matters other than chloride resistance but is discussed here because of the significance of the chloride threat in specification for durability. Performance, it is proposed,

130


Figure 6.6 Format of Dhir, Jones and Ahmed (1991b) nomogram for determination of required chloride resistance

would be assured through achievement of a certain minimum requirement in respect of three indexes. In the same way that the cube test may be regarded as an index which characterises the potential of a structure to resist applied stress, Alexander et al. (1999) propose a series of index tests to characterise durability-related transport mechanisms. Three index tests are proposed: oxygen permeability for permeation; water sorptivity for absorption; and chloride conductivity for diffusion. A matrix of index values is being developed to characterise the required standard for durable concrete. In the case of chloride ingress, for example, excellent resistance is characterised by a conductance value of less than 0.75 mS/cm in combination with an oxygen permeability index greater than 10 and a sorptivity less than 6 mm/ √hr. Test methods Research on determining the coefficient of chloride diffusion has been conducted in laboratory studies using immersed samples, diffusion cells comprising two chambers separated by the test specimen, and accelerated penetration using electrical current.

Chloride ingress 131 Diffusion cells commonly utilise two chambers separated by the concrete sample. One chamber would contain chlorides whereas the other would be chloride-free. Diffusion of chloride ions through the sample can then be monitored. This method measures intrinsic diffusion but the time taken to achieve a steady-state condition is very long. Another method accelerates the process by the application of an electrical field. The electrical conductivity of the pore solution is dependent on the mobility of the ions in the solution and may thus be related to chloride diffusion. Diffusivity of chloride ions has been correlated with the coulombs passing or resistivity values. For example Berke and Hicks (1992) proposed the following correlation for concrete without admixtures: Deff 0.0103 108 (Rapid permeability)0.84 cm2/s where the rapid chloride permeability is in coulombs. Application of the results to practice would involve a further calculation to take account of differences between the diffusion coefficient at the temperature of test and the in-service temperature range. A rapid test based on potential difference was developed by Dhir et al. (1990) based on a single-sided diffusion cell bounded on one side by the concrete sample. The cell was immersed in a 5 M sodium chloride solution and a potential difference of 10 V was applied through a graphite electrode immersed in the cell reservoir. The test duration was seven to twelve days. Streicher and Alexander (1995) developed a test that measures ionic flux through conduction only without the additional effect of diffusion. Two cells expose each face of a concrete sample to 5 M sodium chloride solution. The sample is pre-conditioned to standardise the pore water solution by oven drying at 50°C followed by 24 hours vacuum saturation in a 5 M sodium chloride solution. A 10 V potential difference is applied and the conductance is determined from the current flow through the concrete:

it VA

where chloride conductance (mS/cm) i current (mA) V voltage (V) t specimen thickness (cm) A cross-sectional area (cm2) A possible drawback of the rapid tests involving electrical fields is that the chloride ion mobility in such tests is so rapid that the binding effect of the cementitious component of the concrete is somewhat over-ridden. The performance requirements based on such tests may benefit from a modification that would take this into account.

132


Summary The problem posed by chloride ion ingress represents the greatest threat to the durability of the world’s concrete infrastructure. The long service life requirement of coastal structures and highway bridges demand a level of chloride diffusion resistance which can only be achieved by concrete of the highest quality. This at least involves the use of low water/cement ratio concretes but the benefits of secondary cementitious materials has also come to the fore in this context. Mathematical modelling of chloride ingress for use in service-life prediction and performance-based specifications has to date centred on the diffusion coefficient in Crank’s solution to Fick’s second law of diffusion. Recognising that this is an imperfect approach, others have attempted to produce enhanced models. The complex issues to be addressed include the significance of absorption, the estimation of typical surface chloride level for a given environment, the effect of chloride binding, the influence of seasonal temperatures, the change in diffusion coefficient with time, and the corrosion threshold level for a particular mix. Nevertheless it is recognised that Crank’s solution to Fick’s second law provides a sound engineering basis from which to build the first generation of mathematical models. Equally the test methods used to evaluate the diffusion coefficient are still at development stage with several approaches possible including accelerated testing. Development of a test as a pan-European standard and calibration of the test with in-service behaviour will form an important part of research in the next decade. Meanwhile the introduction of national guidance limits on concrete composition for the XS and XD chloride exposure classes of European standard EN 206–1 should alert specifiers to the need for considered judgement – perhaps more so than would have been the case in the past based on the advice in earlier national codes and standards.

7

Alkali–silica reaction

Background Aggregates in concrete are generally considered to be inert. Under certain conditions, however, some components in aggregates, especially siliceous minerals, may react with the sodium and potassium hydroxides in the alkaline pore solution of concrete. A gel is formed which can imbibe water. The reaction products have a molar volume in excess of the reactants, which leads to swelling. The resultant pressures cause expansion of the hardened concrete and, if the tensile strength of the concrete is exceeded, extensive cracking may become apparent on the surface. This phenomenon is known as alkali– aggregate reaction (AAR). Forms of reaction The term ‘alkali–aggregate reaction’ is used to encompass a number of forms of reaction. The predominant of these throughout the world is alkali–silica reaction (ASR). Other forms are alkali–carbonate reaction and alkali–silicate reaction. The alkali–carbonate reaction occurs when alkalis react with finegrained argillaceous dolomitic limestone aggregate containing calcite and clay. The alkali–carbonate reaction has been a greater cause for concern in Canada and China than elsewhere. The alkali–silicate reaction is significantly less prevalent and less well researched. Attention in most continents, including Europe, is focused on alkali–silica reaction and therefore ‘ASR’ is more universally referred to than ‘AAR’, almost to the extent of excluding consideration of the other forms. The problem is widespread geographically but the number of deleterious occurrences in any one country is very minor compared to other causes of durability failure. The reaction occurs to a limited extent in many concretes but the incidence of damage due to ASR is exceedingly rare. The reaction may become apparent at some time in the life of a structure but later exhausts itself, leaving the structure in a deteriorated but serviceable condition.

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Historical and geographical aspects The first reference to the problem was probably that by Poulsen (1914) to the Institution of Danish Civil Engineers in January 1914. Wider interest in the phenomenon is regarded as having been initiated by researchers in the United States of America in the 1930s, exemplified by the publication of a paper by Stanton (1940). Expansion and cracking had been observed at the time in several Californian structures. Stanton demonstrated through tests on mortars that the high alkali cement used locally could react expansively with opaline silica in the fine aggregate sand. The problem has since progressively appeared in many countries although it was several decades before the extent of the phenomenon became fully apparent. Notwithstanding Stanton’s work and extensive research in Australia, the United Kingdom and Denmark, the aggregate sources of many countries still appeared to be innocuous despite siliceous components but in time circumstances conspired to cause reaction and diagnosis of a first case in a country soon led to the identification of several others. For example the prevalence of opaline flint or chalcedonic flint in Danish aggregates gave rise to problems that were diagnosed in the 1950s. Across the border in northern Germany severe damage due to ASR was identified in a bridge in the late 1960s and further ASR-induced damage was identified in structures in the mid-1970s. Interestingly the problems were greatest in northern Germany due to the use of a local opaline sandstone. This regional aspect of the problem was later encountered elsewhere. For example occurrences in Sweden have been primarily in the Scania region. Interest in the problem grew in the 1970s, fuelling media coverage which overstated the scale of the problem. In the United Kingdom ASR damage was identified in a dam in Jersey in 1971 and in three power stations in southwest England in 1976. Further cases were identified in about 200 structures, primarily in the southwest and the Midlands. In Iceland a combination of highly reactive volcanic aggregates and the introduction of a high alkali level (1.5 per cent) home-produced cement caused ASR problems in 1960s structures and these problems were first diagnosed in 1976. The first positive identification of ASR damage in South Africa was also in 1976 and many cases have since been identified in the Western Cape region. In France ASR damage was identified in dams in the late 1970s, and a later survey of 140 bridges in northern France revealed that 29 per cent exhibited AAR with 5 per cent of the total showing signs of significant degradation. In the 1980s a survey in the south of Norway of over 400 dams, hydropower plants and road bridges identified ASR in thirty-one structures. Equally, a survey of over 400 New Zealand bridges commenced in 1989 and of over 100 suspect cases AAR was positively identified in about one-third of these. Five of these structures required major repair work, including replacement of one structure. Meanwhile ASR damage was identified in some Italian industrial


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structures and pavements, mainly on the Adriatic coast, and ASR problems were also identified in Japan. First instances of ASR damage continue to be reported in other countries as late as the 1990s. Damage due to ASR was identified in a thirty-year old viaduct in the Netherlands in 1990 and other cases have since come to light where an indigenous flint aggregate was used. ASR damage has been identified in six Portuguese dams and in a viaduct, while pavement cracking due in part to alkali–silicate reaction was identified in 1993 at the airport of Santa Maria, Azores. Thus the problem continues to appear in countries previously thought to be unaffected. To date no cases have been reported in certain countries including Austria, Finland and the Republic of Ireland. The geographical and chronological pattern of recognising ASR damage to structures indicates that the reaction probably occurred in specific countries during slight changes in concrete practice. These changes unwittingly brought together hitherto uncombined material and environmental conditions conducive to ASR at a time before the reaction was fully understood. Examples of change include higher alkali levels consequent on change in cement chemistry or the rise in cement contents of concrete. The use of new aggregate sources, such as sea-dredged aggregates, or the more extensive use of de-icing salts, may also have been a trigger. Guidelines for minimising the risk of deleterious ASR have been progressively introduced and updated in most countries in the 1980s and 1990s.

Manifestation of the problem The reaction may occur in many concretes but does not always cause damage. The aggregate particle decomposes at the reaction site and the gel is generated. The gel may then replace the decomposed aggregate, permeate into voids, or migrate into microcracks caused by expansion. The reaction sites can be at the aggregate–cement interface or within the aggregate particle. Damage due to ASR is a more rare occurrence than ASR itself. Indeed some cases of damage previously attributed to ASR have subsequently been diagnosed as delayed ettringite formation or freeze/thaw damage. ASR has been detected in deteriorated concrete structures that have required demolition but other factors, such as corrosion due to chloride ingress, were more significant. The reactive nature of Danish aggregates is such, for example, that ASR gel will be found in cracks of most deteriorated structures but ASR will not necessarily have been the cause of the cracking. Damage from ASR most likely takes the form of cracking or differential movement. Some commentators also feel that there are wider durability effects of ASR cracking besides the aesthetic problem. ASR-induced cracking will slightly reduce the strength of the material in compression, tension and flexure.

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Figure 7.1 Map cracking pattern associated with ASR in unreinforced zones

Figure 7.2 ASR crack pattern associated with influence of reinforcement

Visible damage The crack pattern depends on the level of stress in the member and the disposition of reinforcement. Crack depths are normally in the range of 25 mm to 50 mm but can be higher. Cracks initially develop in a three-pronged manner to produce characteristic ‘Manx’ cracks (Hobbs 1988). In an unreinforced zone the cracks further develop to join up with one another and display the characteristics of map cracking illustrated in Figure 7.1. This pattern results from the differential movement of the interior and exterior of the concrete member. The crack pattern is more rectangular where reinforcement lies close to the surface, as illustrated in Figure 7.2, due to the influence of the main direction of stress. The greater the amount of reinforcement present in a member the smaller are the induced strains in the steel due to ASR expansion. This may restrain the amount of expansion and will also influence the crack pattern. An Institution of Structural Engineers (1992) report


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Figure 7.3 Example of a structure damaged by ASR

emphasised the need for three-dimensional well-anchored reinforcement detailing if severe ASR expansions are to be limited. An example of a structure affected by ASR is illustrated in Figure 7.3. Impact on serviceability The expansion and cracking associated with ASR does not usually degrade its structural adequacy. Hobbs (1988) reported that the loss of compressive, tensile and flexural strength due to ASR is of the order of 10 to 30 per cent. The reduction in elastic modulus consequent on expansion ranges from 20 to 50 per cent. The performance of ASR-damaged structures in full-scale load tests has been satisfactory. The Institution of Structural Engineers (1992) issued guidance on appraisal of existing structures. The report contains a flow chart for appraisal of existing structures and guidance is given on strength estimation dependent on the level of expansion to date or predicted in the time frame of interest. Stresses may be induced by ASR expansion if restraints exist. Differential movement may lead to misalignment of members and differential expansion may take place in separate pours of the same member. The expansion of individual members may not be uniform even if cast in one pour. Asymmetrically reinforced members may hog as a result of the bending moment induced by differential restraint.

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The durability question is a difficult one to address. It has been suggested that the introduction of cracks due to ASR in the cover zone of reinforced concrete members should render the reinforcement more vulnerable to corrosion. This is because the permeability of the concrete will rise and depassivating influences such as carbon dioxide and chlorides will have easier access. However the debate on the relationship between corrosion rate and cracking is still unresolved. This is because the corrosion rate is influenced by the location of anodic and cathodic sites on the bar. The latter will not necessarily occur at the worst location (i.e. at a crack). This is discussed further in Chapter 11. A related problem is that of freeze/thaw behaviour where ASR-induced cracks may allow greater water ingress, which then has the capacity to cause disruption if it freezes. Confirmation of ASR as the source of damage Confirmation that ASR is the cause of cracking in a structure requires expert investigation as it involves detection of three features: • • •

presence of gel in cracks; presence of aggregate particles petrologically identifiable as belonging to a ‘potentially reactive’ classification; presence of internal cracks in one of two patterns recognisable to experts in the field of ASR diagnosis.

Diagnosis of ASR is difficult because many of the features of ASR cracking are common to other causes such as drying shrinkage; freeze/thaw damage; leaching of calcium hydroxide and efflorescence. The potential for confusion with efflorescence arises even though the gel produced by ASR exudes with a transparent or brownish colour because on exposure to the atmosphere the gel carbonates and turns white. A commercially available kit for the detection of ASR gel is available but on-site testing alone is not generally adequate to conclusively prove ASR. A number of points are worthy of consideration in relation to ASR laboratory testing. Laboratory staff of geologists, scientists, technicians and engineers should be experienced in the diagnosis of ASR. Some tests being used are still at the developmental stage, in the sense that their applicability to native aggregates in different countries is still in doubt. Tests involving elevated temperatures and humidity to accelerate expansion may not reliably model in-service performance of every aggregate combination. Finally, the costs of a thorough examination may not be economically justifiable. An excellent reference on the topic of diagnosis has been published by the British Cement Association (1992). The procedure may be considered under four distinct headings: preliminary investigation; site investigation; sampling; and laboratory testing. An overview of the diagnosis procedure can be useful background knowledge for the specifier of new works.


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The BCA procedures envisage that the preliminary investigation would involve the determination of pertinent matters from records covering age of the structure; sources of aggregate; mix design; exposure conditions including rainfall and exposure to salts. The site investigation would involve examination of the crack patterns and checks for discoloration in the form of a light coloured zone either side of crack with a dark colour along the edges of the crack itself. Gel exudation would be checked for. The presence of any popouts would be examined to see if the base of the spalled area coincided with a reactive aggregate particle. Differential movement would be surveyed. Sampling would involve recovery of cores representing sound, typical and damaged concrete. The samples would be wrapped in clingfilm for despatch to the laboratory. Laboratory testing would first involve wetting the surface of the core under running water and wrapping it in clingfilm for overnight storage. The next morning it would be unwrapped and allowed to dry. The presence of any sweaty patches may indicate local ASR reaction sites. Microscopical examination would be used to examine the internal crack pattern. Other tests common to aggregate assessment for reactivity prior to use in construction would then follow. Details of these assessment tests are presented later in the chapter.

Mechanism of expansion and reaction The mechanism of expansion is accepted as being the adsorption of water by the gel produced in the reaction. The water in the gel has a lower free energy than the water in the pore solution surrounding it and so a flow into the gel results. The reaction itself is less well understood but a number of key aspects have been identified (Diamond 1976, Dent Glasser and Kataoka 1981, Chatterji et al. 1987, Hobbs 1988, Chatterji 1989a, 1989b, Diamond 1989, Chatterji and Thaulow 2000). The reaction may be idealised as follows: 4SiO2 2NaOH → Na2Si4O9 H2O The reaction is more complex but the formula above gives a rough guide to the compounds involved. In reality the chemistry involves stages including the formation of negatively charged SiO ions from the aggregate that attract positively charged ions including Na and K from the pore fluid. The alkali ions that diffuse into the aggregate grains exceed the counter diffusion of silica (SiO2) and so expansive pressures build up. The driving force is the concentration of alkali hydroxides. Diamond showed that, for a given water/cement ratio, there is a linear relationship between hydroxyl ion concentration and the alkali content of cement, expressed as sodium oxide equivalent. This is significant in two ways. First, the concentration of alkali hydroxides in the pore solution is related to the hydroxyl ion concentration and this correlates with the finding that ASR

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occurs in conditions of high pH. Second, sodium and hydroxyl ions penetrate reactive silica more easily than calcium ions. High alkali cements result in pore solutions that contain significant amounts of dissolved sodium hydroxide (NaOH) and potassium hydroxide (KOH) compared to calcium hydroxide (Ca(OH)2). The solution is capable of attacking reactive aggregates to form dissolved silica. The structure of quartz involves a framework of silicon–oxygen tetrahedra. Each oxygen is shared between two silicons, while each silicon is bonded to four oxygen atoms. If the framework is well-crystallised the attack takes place on the surface and the reaction is slow. However if the material is poorly crystalline the metal alkalis penetrate and rupture the silicon–oxygen bonds (Si-O-Si) through hydroxyl attack. This may be represented as: Si-O-Si OH Si-OH Si-O Chatterji and Thaulow argue that the silica grain attracts cations more strongly because of the surface charge density on the grain. They postulate that the cement–aggregate–pore solution system results in negatively charged cement particles and silica grains surrounded by positively charged cations. The silica grains are larger and attract the Na, K, Ca2 cations more strongly. The negative Si-O attract more Na and OH ions into the silica further loosening the structure of the silica lattice. The penetrating hydroxyl and sodium ions cause a breakdown of the Si-O-Si bonds. This liberates silica but also opens the grains to penetration by further ions including larger ones such as calcium. The diffusion of silica out of the reacting grains is controlled by the calcium ion content and expansion results as the ions penetrating the aggregate exceed the outward diffusion. This generates expansive forces. In the presence of calcium an insoluble gel is formed and over time the residual concentration of alkali hydroxide reduces. The gel forms as the Na ions join the negative oxygen on the silica surface: Si-O-Si 2Na 2OH 2Si-O-Na 2H2O The essential role of calcium in the reaction was highlighted by Chatterji et al. (1987) and in research by Struble, cited by Diamond (1989). First, the presence of calcium hydroxide is essential for sodium and other ions to penetrate into the reacting grain. Second, the reaction products would not form a gel unless there was a source of calcium – the silica would otherwise simply remain in solution. The gel contains substantial amounts of calcium. The resultant complex then imbibes water and causes swelling. Dent Glasser and Kataoka (1981) observed that the maximum volume expansion of silica gel in sodium hydroxide solution occurs at an intermediate level of total SiO2 /Na2O mole ratio. This may account for the pessimum effect observed in some concretes whereby damaging expansion occurs at an intermediate silica level.


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Primary factors influencing the reaction Three components are required for an alkali–silica reaction to occur. These are: • • •

an adequate supply of moisture (usually from the environment); reactive silica (usually from the aggregate); a sufficiently alkaline pore solution (primarily derived from the cement, sometimes increased by admixtures, external salts etc.).

All three must be present simultaneously. Development to the point of damaging expansion requires a significant quantity of reactive silica. Moisture Moisture plays a dual role in the reaction. First, it acts as a transport route for the reactive ions. Second, the gel produced by the reaction will imbibe water if sufficient moisture is available. This leads to expansion that may cause damage. Thus the level of moisture controls the difference between the occurrence of ASR as such and the occurrence of damaging ASR. Damaging ASR has generally been observed in cases where an external supply of moisture was available. Opinions vary on the minimum relative humidity value for damaging expansion but a figure of 80 per cent is the most commonly quoted. Reactive silica Most aggregates used in the production of concrete are not pure in the sense that they may contain a proportion of other mineral components. These other components may include a small percentage of reactive silica. This is especially true of fine aggregate. Indeed it has been estimated by Glasser (1991) that the average chemical composition of the accessible portion of the earth’s crust is about 65 per cent silica by weight. Silica, besides occurring as a solid crystalline oxide such as quartz, also occurs in the form of a silicate whereby silicon and oxygen are combined with other elements such as calcium, potassium, and sodium. The silica in the reaction usually comes from the aggregates but it must be noted that not all forms of silica are reactive. The degree of reactivity of silica minerals depends on the crystal structure with disordered structures being the most vulnerable. This is because the more disordered the structure the greater the surface area available for reaction. The crystalline structure will depend on the method of formation and geological history of the aggregates. Vulnerable silica structures include amorphous, glassy, crypto-crystalline and microcrystalline. Unstrained quartz, the most abundant form of silica, is well ordered, stable and is unlikely to react.

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Particular caution is required where petrographic reports indicate the presence of potentially reactive constituents as follows: opaline silica, crystobalite, tridymite, microcrystalline quartz, crypto-crystalline quartz, chalcedony, volcanic glass. Petrographic reports may indicate the presence of flint, chert, or greywacke. Damaging ASR in some countries has been linked to these. Crypto-crystalline quartz and chalcedony may be present in flint and chert. Microcrystalline quartz may be present in greywackes. A useful review of the nature of these reactive forms of silica was published by Diamond (1976). It is worth noting that it is the aggregate combination – fine and coarse – that must ultimately be assessed. This is because the proportion of reactive silica in the combined aggregate determines the degree of reactivity. In certain cases reaction will only occur if the proportions lie within a critical band. Proportions less than and more than the critical amount will not lead to reaction (the ‘pessimum effect’). One must be conscious of the potential abuse of the term ‘reactive aggregate’, commonly used to describe one that contains a potentially reactive component. The term gives the impression that the aggregates are in some way substandard but the reaction will only occur if sufficient moisture and a high level of alkalis prevail. Even ‘reactive aggregates’ will not react unless the other components are in place, yet one never finds reference to ‘reactive cement’ and ‘reactive moisture’! Alkali level The reaction will only proceed if a certain level of alkalinity in the pore fluid is reached. Even at very low water/cement ratios water may be found in micropores in the cement paste and at interfaces of aggregate and hydrated cement particles. This pore fluid is distinct from the large diameter pore structure characteristic of higher water/cement ratio concretes. The alkalinity (hydroxyl ion concentration) of the pore fluid is primarily influenced by the sodium and potassium alkali metals in the cement and by the cement content. The alkali metals primarily derive from the characteristics of the raw feed materials used in the manufacture of cement. Secondary influences, such as external sources of salt and admixtures, are also contributory. The alkali content of concrete is determined in kilogrammes per cubic metre through the summation of alkalis contributed from the cement, aggregates, admixtures and additions: [A]concrete [a]cement [a]aggregates [a]admixtures [a]additions where [A]concrete alkali content of concrete (kg/m3) [a]x alkali contribution to total of component ‘x’ (kg/m3) In the context of ASR, the reactive alkali content of a Portland cement is represented by the ‘acid soluble equivalent sodium oxide content’. It is a


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Table 7.1 Determination of alkali content of cement Element or compound

Atomic mass (amu)

Sodium (Na) Potassium (K) Oxygen (O) Sodium oxide (Na2O) Potassium oxide (K2O)

22.9898 39.102 15.9994 61.979 94.203

Ratio Na2O/K2O 0.658

convention to express the alkali metal concentrations in terms of their oxides. This value, commonly referred to as the alkali content of cement, is calculated through a summation of the sodium oxide (Na2O) and potassium oxide (K2O), taking account of their relative atomic masses. Thus the alkali content of cement (per cent) is: per cent(Na2O)equiv [Weight per cent Na2O] 0.658[Weight per cent K2O] where (Na2O)equiv acid soluble equivalent sodium oxide content The value of 0.658 derives from the relative atomic masses of the two compounds as shown in Table 7.1. The reason for this approach is related to the significant role of the hydroxyl ions in the reaction. The concentration of hydroxyl ions produced by an alkali metal oxide is proportional to the molecular weight. Thus the mass of sodium oxide required to produce a specific value of hydroxyl ion concentration in a solution is less than that required for potassium oxide. The alkali content contributed by the cement is calculated as follows: [a]cement {per cent(Na2O)equiv cement content}/100 where the cement content is expressed in kilogrammes per cubic metre. The acid soluble equivalent sodium oxide content can be reported by the cement manufacturer in a number of ways and may or may not include an allowance for variability. A common method would be to report the ‘certified average alkali content’ defined as the average of the last 25 determinations of alkali content carried out on consecutive daily samples. Should an allowance for variability be required this may be included by adding a factored number of standard deviations (typically 1.64). The value may safely be regarded as a maximum as not all the alkalis are necessarily available for reaction. Alternative approaches also exist. For example a cement manufacturer may use a ‘declared mean’ approach, whereby the alkali level would not be continuously reported but the user would be notified if alkali levels crept above a predetermined value.

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The alkali content of Portland cements produced throughout the world lies in the range 0.3 to 1.5 per cent and awareness of the ASR phenomenon has generally led to a decrease in the average value. In the Republic of Ireland, for example, the average values for Portland cement have progressively declined from values in excess of 1 per cent to about 0.7 per cent. Cements with alkali values of 0.6 per cent and less are described as ‘low alkali’ cements. It has been the practice in many countries to assume that concretes made with ‘reactive aggregates’ would not suffer ASR-induced damage if used with a ‘low alkali’ cement because the hydroxyl ion concentration may be one-tenth of that found in concrete made with a high alkali cement. The alkali contribution from aggregates needs to be taken into account if they contribute to the alkalis through, for example, sodium chloride content or feldspar. Sea-dredged aggregates would be an obvious source of concern but careful monitoring would also be required in the case of coarse recycled aggregate and coarse recycled concrete aggregate. Some national guidelines state that the contribution from this source may be ignored if the chloride ion content is less than 0.01 per cent. The contribution, if any, may be taken into account by determining the percentage by weight of chloride ion and factoring this by 0.76. The value of 0.76 derives from a consideration of the composition of seawater. This figure may then be multiplied by the mass of the aggregate to yield the alkali contribution of the aggregate. Thus: [a]aggregate [a]fine [a]coarse where [a]fine 0.76 {per cent(Cl)fine aggregate fine aggregate content}/100 [a]coarse 0.76 { per cent(Cl)coarse aggregate coarse aggregate content }/100 where per cent(Cl) chloride ion content as a percentage by mass of dry aggregate and the aggregate content is expressed in kilogrammes per cubic metre of concrete. Potential problems with the release of alkalis from feldspar have been highlighted by Poulsen et al. (2000). The contribution of alkalis from admixtures must be included by reference to its sodium oxide equivalent and the dosage rate. Thus for example if the admixture was used at the recommended dosage rate the expression might be as follows: [a]admixtures {(Na2O)equiv}admixture (cement content/100)


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where {(Na2O)equiv}admixture sodium oxide equivalent of admixture in kilograms per 100 kilogrammes of cement when used at the recommended dosage rate and the cement content is expressed in kilogrammes per cubic metre of concrete. The issue of alkalis contributed by additions ([a]additions) has been a source of difference in guidelines for minimising the risk of damaging ASR. In the past some guidelines recommended that the alkali contribution of pulverised fuel ash and slag be ignored; others that it be taken as the water-soluble alkali content. Most current guidelines require that it be determined from the total acid soluble alkali content of the addition. The precise fraction of the total acid soluble alkali content of the addition depends on various circumstances such as the proportion of the addition relative to Portland cement. Examples of the fraction to be used are discussed further in a later section on specifications.

Other factors influencing ASR occurrence Besides the three fundamental issues of alkalis from the cement, reactive silica from the aggregate, and the moisture level, the following factors are of note: • • • • •

the pessimum effect; the effect of fly ash, slag, microsilica, metakaolin and lithium salts; external salts; temperature effects; aggregate porosity.

Pessimum effect The pessimum effect is exhibited by expansions caused by ASR. At low and at high reactive silica concentrations the expansions may be low. The maximum expansion occurs at an intermediate concentration. This phenomenon is illustrated in Figure 7.4. There are a number of possible explanations. One is that at low concentrations the volume of gel produced is too low to exhibit significant expansion whereas at high concentrations the reaction is vigorous but exhausts itself before the concrete has fully hardened. The pessimum proportion arises from the combination of coarse and fine aggregate. Thus an inert aggregate may contribute to the problem if it is combined with a reactive aggregate such that the combination results in a pessimum level. The avoidance of damaging ASR in cases where aggregate contains significant quantities of chert or flint may be to use a combination of coarse and fine aggregate that possesses a chert content high enough to exceed the pessimum. Effect of fly ash, slag, microsilica, metakaolin and lithium salts Secondary cementitious materials, such as pulverised fuel ash (pfa), ground granulated blast furnace slag (ggbs), silica fume, metakaolin and lithium salts

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Figure 7.4 Pessimum effect

can be beneficial in controlling expansion due to ASR if used in specific proportions. The effect of mineral and chemical admixtures was reviewed by Hobbs (1989) and by the Building Research Establishment (2002). Recent research has influenced the consideration of pfa and slag in the United Kingdom guidelines on avoidance of damaging ASR (Blackwell 1997). The silica in the pfa and ggbs can react with alkalis in cement with consequent lowering of the alkalinity, which in turn reduces the risk of damaging ASR. These materials can contribute some alkali to the pore fluid but much is too tightly bound and will not be released. Chatterji (1989b) stated that the effect of the pozzalan was to combine with the calcium hydroxide thus lowering the OH and Ca2 ion concentrations and consequently decrease ASR activity and expansion. Furthermore, such additions would lower the calcium hydroxide content and allow more silica to diffuse out of the reactive grains. The total alkali content of fly ashes range from about 0.7 to 7.8 per cent expressed as equivalent sodium oxide. At low replacement levels the effective alkali contribution from a fly ash can be high and could have a detrimental effect on expansion. However at intermediate replacement levels, of 20 to approximately 25 per cent, the effective alkali contribution is about one-sixth. At higher levels there may be no effective alkali contribution but research at present is limited to pfa with a maximum equivalent sodium oxide level of about 5 per cent. The coarser the pfa the less effective it is, requiring minimum replacement levels of 30% for coarse material. The alkali content of slags range from about 0.3 to 2.6 per cent expressed as equivalent sodium oxide. At low replacement levels the effective alkali contribution from a slag can be high enough to have a detrimental effect on


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expansion. At intermediate replacement levels, of 25 to approximately 40 per cent, the effective alkali contribution is about one half. At higher levels there may be no effective alkali contribution but research knowledge is biased toward slags with equivalent sodium oxide levels below 1 per cent. The equivalent sodium oxide content of microsilica ranges from 0.3 to 5.5 per cent and the material contains between 60 and 98 per cent by mass of amorphous silica. This has the beneficial effect of promoting a rapid reaction with the formation of a calcium–alkali–silicate hydrate, if used at a content of 5 to 20 per cent. The particular difficulties of ASR in Iceland were overcome by the inclusion of silica fume in Icelandic cement, initially at a level of 5 per cent and later increased to 7.5 per cent. Current research indicates that at least 8 per cent silica fume is required to provide a beneficial effect. Metakaolin, a purpose made pozzolana, when used with Portland cement in concrete reacts with the calcium hydroxide to produce extra calcium silicate hydrates and calcium alumino–silicate hydrates. Jones et al. (1992) demonstrated that the reduction in calcium hydroxide content would suppress damaging expansion due to ASR. The replacement level would typically be at least 10 per cent but up to 15 per cent can be used to beneficial effect. Evidence of the beneficial influence of lithium nitrate (LiNO3) and lithium hydroxide monohydrate (LiOH.H2O) additions exists where the molar ratio of lithium to sodium plus potasium is in the range 0.7 to 1.2. External salts Alkalis from external sources may be introduced into concrete. Nixon et al. (1987) confirmed that addition of sodium chloride to cement paste rapidly converts to alkali hydroxide. This adds to the alkali load and so exacerbates the threat of ASR. Hobbs (1988) notes that sodium chloride in the mix reacts in one of two ways that increases the hydroxyl ion concentration. Sources of external alkalis include road de-icing salts, seawater, and industrial alkalis. Foundations and other buried concrete may be subjected to alkalis in groundwater. Chatterji (1989a) describes the reaction as being of the following type: Reactive silica 2Na 2Cl Ca2 2OH aq. solid alkali–silica hydroxyl complex Ca2 2Cl aq. where aq. aqueous solution The smaller Na ions and OH ions penetrate the reactive aggregate and breakdown the Si-O-Si bonds thereby opening the grains to penetration by more Na, OH and the larger Ca2 ions allowing part of the reactive silica to go into solution. It may be seen that the presence of sodium chloride (NaCl) contributes to the process.

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Figure 7.5 Influence of temperature on ASR-induced expansion

Temperature effects At elevated temperatures reaction rates and expansion rates are high but decline with time. At low temperatures the rates are slower but total expansion may eventually reach or exceed that at higher temperatures (Figure 7.5). The accelerating effect of elevated temperature is a feature of tests used in aggregate assessment. Hobbs (1992) found that the reaction occurred seven times faster for specimens stored at 38°C than for those stored externally at an average temperature of 9°C. The rate was four times faster than for samples stored at 20°C. The reaction generally tends to mature and cease in about twenty years but longer periods may be expected in colder climates and less in hot countries. Aggregate porosity The porosity of the aggregate can be significant. If a potentially reactive situation occurs through a combination that is in the pessimum proportion it has been found (BRE 1999a) that low porosity coarse aggregate will produce more damaging expansion than a high porosity material.

Modelling and service life prediction Modelling the rate of alkali–silica reaction and the form of the expansion curve has been studied by, for example, Hobbs and Gutteridge (1979) and Hobbs (1981). However the application of mathematical models of ASR to specifications and service life prediction is not seen as being appropriate at this stage. The approach envisaged in the medium to long term remains pre-


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scriptive. Indeed even the prospect of harmonising the prescriptive approach in Europe, let alone internationally, is thought to be unrealistic at present.

Specifications to minimise the risk of ASR Alkali–silica reaction has not caused damage in all countries to date but given the current state of knowledge there is clearly an onus on all specifiers to take the phenomenon into account. However, unlike many other issues of concrete technology, the ASR problem requires regional issues to be taken into account due to varying geologic and climatic conditions. Specification clauses to minimise the risk of alkali–silica reaction are not therefore readily amenable to harmonisation in Europe. National guidance is required – otherwise some countries would be exposed to excessive risk of ASR occurrence while others would be forced to comply with unnecessarily restrictive limits. Most countries have now introduced guidance documents for specifiers and materials suppliers to minimise the risk of damage due to ASR in future works. Although national guidance is not harmonised, common principles are identifiable. The relevant task group of CEN (TC104/SC1/TG9) have, for the moment, accepted that the best approach is to publish national guidelines that would be of use not only to specifiers using local materials but also to countries that import aggregates and to contractors who work in several countries. European standard EN 206–1 therefore does not attempt to harmonise the approach. Specifiers are referred to precautions valid in the country of use. Specifically it draws attention to the survey of regional specifications and recommendations for avoidance of damaging ASR published by CEN (1995a) as Report CR1901. Inevitably however some national guidelines have been updated since publication of the report. Specifiers should therefore confirm the validity of the guidance by reference to the latest national documents. RILEM established an international technical committee, TC106-AAR, in 1988 to develop accelerated tests for universal application in the assessment of aggregates. The committee produced an integrated assessment scheme (Sims and Nixon 2001), which is being further developed by a successor technical committee TC-ARP. Common principles in specifications A review of national specifications reveals general consensus on a number of guiding principles. The first issue is consideration of the acceptability of the consequences of alkali–silica reaction. The reaction requires three factors to be present simultaneously: an adequate supply of moisture; a sufficient quantity of reactive silica; and a sufficiently high alkali level. Control of any one of these factors will control the likelihood of ASR occurring. Thus national guidance documents consider to greater and lesser degrees the following issues:

150 • • • •

Fundamentals of durable reinforced concrete consideration of level of risk, control of moisture, use of aggregates unlikely to be reactive, limiting the alkali content.

The degree of dependence on any one aspect varies from country-tocountry depending on the economic consequences of meeting the restrictions. The overall issues are summarised in Figure 7.6 in the form of a flow chart and variations of such a chart are commonly found in many national guideline documents. The issue of alkali–silica reaction has attracted considerable interest but specifiers need to be mindful that other aspects of concrete durability must be considered in tandem when assessing the suitability of a design mix and its constituents. Short-circuiting the assessment Some of the stages involved in assessment of compliance with ASR guidance, for example aggregate testing, can be time-consuming and expensive. It is interesting therefore to note a clause in the first edition of the Irish national specification guidance document (Institution of Engineers of Ireland/Irish Concrete Society 1991) which may be used to short-circuit the procedure. Recognising that, in the final analysis, an aggregate of unknown reactivity could be used if the alkali content of concrete was controlled, the clause is of the following general form: ‘If the alkali content of the cement does not exceed W per cent and the quantity of cement used does not exceed X kg/m3, alkali–silica reaction does not require further attention in specifying the concrete’. Such a clause covers the case where the following conditions apply: W X/100 Y Z where Y (kg/m3) is the appropriate national limit on the alkali content of the concrete and Z (kg/m3) is an allowance as a margin of safety. It will be recognised that a suitably conservative value of X, such as 360 kg/m3, would remove the onus for further assessment of ASR in a very large proportion of cases employing structural concrete. This general recommendation will frequently be appropriate. In the specific case of the Irish document the recommendation does not apply where secondary cementitious materials are added at the mixer, nor where the concrete is subjected to the extensive use of de-icing salt or salts introduced in industrial processes. In determining the value of X allowance was made for the use of admixtures in proportions usually specified and for alkali input from washed properly processed aggregates. The alkali level (W) may include an allowance for variability of 1.64 standard deviations. Another example of attempts to simplify the process of assessing


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Figure 7.6 Structure of a flow chart encompassing the most common routes to control the risk of damaging ASR in a country or region

compliance with national guidance is that published in the United Kingdom. The full details are published by the Building Research Establishment (1999b). However it was recognised that most concrete in the United Kingdom would be made from aggregate combinations classified as being of ‘normal reactivity’ and from cement with a declared mean alkali content of less than or equal to 0.75 per cent. Therefore it was possible to produce a shorter

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version of the guidance document specifically for these conditions (Building Research Establishment, 1999d). Consideration of level of risk A key issue in the sphere of design and specification of engineering works is the minimisation of risk. Recognising that specifications are used in the management of risk the first question to be addressed is whether the consequences of the possible occurrence of ASR are acceptable in the structural element under consideration. If the consequences are acceptable then no further precautions regarding ASR are necessary in specifying that concrete. The balance between economy and acceptable level of risk must be addressed. Damage from ASR is a rare form of serviceability failure and the consequences of the possible occurrence of ASR are more likely to be related to the aesthetics of the building or structure rather than loss of structural integrity. Alternatively it may be acceptable to allow ASR occurrence if the concrete can remain serviceable through the use of specific control measures. For example, Guo et al. (1999) demonstrated that the inclusion of steel fibres at a dosage rate of 1.0 to 1.5 per cent by volume could restrain expansions caused by the reaction such that reactive aggregates could be used at high alkali loads. Control of moisture Circumstances may arise in rare cases where the specifier is confident that low relative humidity levels will be maintained or where the concrete is sufficiently protected from moisture. No further precautions against ASR would be necessary if the relative humidity remained below, say, 80 per cent except for short periods or where the structure would be protected from severe exposure to moisture or the elements. The control of moisture during the entire service life, however, is unlikely to be a realistic option under most climatic conditions. Even indoors the possibility of high relative humidity levels during the service life of a building could not always be discounted. Aggregate assessment The assessment of aggregate reactivity and any consequent control measures such as limitations on alkali level, vary from country to country. However the need for an initial petrographical examination is clear. Subsequent assessment may be required in the form of testing for expansion. A framework for assessment has been produced by an international committee of RILEM and progress has been reported by Nixon and Sims (2000). The aggregate combination must first be subjected to petrographic examination to identify the presence and concentration of any potentially reactive constituents. Methods commonly used are in accordance with standards such as ASTM C295 or BS 812: Part 104. The importance of using


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a petrographer experienced in concrete technology and alkali–aggregate reaction, in particular, has often been emphasised. Guidance has also been prepared by the RILEM committee as Recommended Test Method AAR-1. The procedures are based on visual examination of the sample to identify and quantify rock and mineral constituents that might be alkali-reactive. Two principal techniques are employed. The choice of technique is usually dependent on the particle size and complexity of the sample but both may be used in many cases. The first technique involves visual inspection of the sample and the second involves point counting within a thin section. In the case of aggregate where not all of the mineral constituents can be identified by the two techniques further testing may be carried out, for example by chemical analysis or x-ray diffraction analysis. The RILEM procedure proposes that the aggregate combination be assigned to one of three categories: • • •

Class I: Unlikely to be alkali-reactive Class II: Potentially alkali-reactive Class III: Very likely to be alkali-reactive

Further categorisation would be based on the extent to which the material was siliceous (II-S, III-S), carbonate (II-C, III-C), or a significant combination of each (II-SC, III-SC). The next step in the assessment procedure would depend on the category. Class II aggregates would require testing to establish the degree of reactivity, if any. Some Class III aggregates would not require confirmatory testing if the presence of reactive material had been clearly established. The categorisation of aggregates in the United Kingdom follows a slightly different route. The aggregates are categorised based on reactivity, defined in terms of the threshold alkali level. Three categories have been defined based on experience in service: • • •

low reactivity, normal reactivity, high reactivity.

A listing of rocks and minerals by reactivity class is presented in a Building Research Establishment document (BRE 1999b). In the past chemical analysis was used to categorise aggregates. The quick chemical test ASTM C289 was widely used. The test method involves grinding the aggregates and treating them with a normal solution of sodium hydroxide at 80°C for 24 hours. Measurements are made of the dissolved silica (Sc) and the reduction in alkalinity of the sodium hydroxide solution (Rc). These parameters are plotted against one another and the location of the result compared against standard curves that delineated zones of reactivity. The general form of such a plot is shown in Figure 7.7. The

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Figure 7.7 General form of chart used to differentiate innocuous and deleterious aggregates by chemical means

applicability of the standard zones internationally was questionable and so other tests have been developed rather than calibrating the charts on a regionby-region basis. RILEM Class I aggregates could include those with a proven record of satisfactory use. Many sources of aggregate have been satisfactorily used over a period when alkali contents of cement were higher than today with no cases of ASR damage being reported to date. Thus an aggregate source may be deemed ‘unlikely to be alkali-reactive’ if it has a known satisfactory history of use. A specifier reliant on satisfactory history of use as a criterion for acceptance of an aggregate would need to consider the following issues: • • •

the combination of fine and coarse aggregate must have a satisfactory history; the cement content proposed for new works must not exceed the cement contents on which the satisfactory history is based; the exposure conditions must not be more severe than those on which the satisfactory history is based.

Thus the ‘satisfactory history of use’ route would not apply to new quarried sources, untried combinations of aggregates nor for structures with higher strength concretes than those that have a proven record of satisfactory service.


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Figure 7.8 Indicative limits, concrete prism test

Aggregates which petrography indicates are RILEM Class II (and perhaps Class III) require further testing to assess the degree of reactivity. Long-term reference tests are now available but interest in a rapid screening test has also been addressed for siliceous aggregates. Development of rapid tests for carbonate aggregates is in train. The long-term reference tests involve monitoring the expansion of concrete prisms which are stored for a year in an atmosphere conducive to alkali– aggregate reaction. The tests are of necessity long term but may become the definitive route for aggregate acceptance. The method is covered in standards such as BS812: Part 123 and RILEM Recommended Test Method AAR-3 (Nixon 2000b). Prisms of 75 75 250 mm are made using the coarse and fine aggregate in a standard mix combination, which leads to a high alkali level of about 7 kg/m3. The specimens are wrapped in damp cloth and stored in containers to maintain high humidity. Temperature is maintained at 38°C. The RILEM committee have surveyed experience of the test carried out internationally using aggregate combinations of known performance in service. Definitive interpretation limits have not yet been agreed but the following indications, illustrated in Figure 7.8, have been presented (Sims and Nixon 2001): 0.05 per cent expansion at twelve months: non-expansive; 0.15 per cent expansion at twelve months: expansive.

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For the moment intermediate results would be regarded as potentially alkali-reactive. A variation of the test in the United Kingdom involving a period of storage of two years is recommended for assessment of greywackes. A protocol to the standard method has been issued by the British Cement Association (1999). This is related to the imposition of appropriate limits on alkali content of concrete and is discussed later. An expansion test on mortar bar prisms, ASTM C227, was used internationally for many years in the absence of other tests. It is now less popular internationally because it does not take account of any potential pessimum effect but it did inspire a variation that became known as the ultra-accelerated test. The ASTM C227 test involves crushing the aggregate and forming 25 25 250 mm mortar bars, which are stored at 38°C (plus or minus 2°C) and a relative humidity in excess of 95 per cent. Expansion is monitored and the standard sets acceptance criteria by maximum limits for expansion at three months of 0.05 per cent and at six months, 0.10 per cent. The variation on this test involves immersing the mortar bars in hot alkali solution and monitoring expansion up to fourteen days. Research by Oberholster (1983) in South Africa has led to the development of such an accelerated test and versions of it have been developed as ASTM C1260, BS DD249 and RILEM Recommended Test Method AAR-2 (Nixon 2000a). In summary the accelerated mortar bar test involves constructing mortar bars to a prescribed recipe. Fine aggregate is prepared for test by washing but coarse aggregate is crushed to yield defined percentages of different sizes. The mortar bars are cast to a defined mix recipe. Following demoulding they are brought to a temperature of 80°C by water immersion for 24 hours before being submerged in sodium hydroxide solution at 80°C. Expansion is monitored over the following fourteen days. Limits for interpreting the results are currently being calibrated by international trials. International comparisons (Sims and Nixon 2001) have suggested the following limits at fourteen days, illustrated in Figure 7.9: 0.10 per cent expansion at fourteen days: non-expansive; 0.20 per cent expansion at fourteen days: expansive. For the moment intermediate results would be regarded as potentially alkali-reactive. It is intended that the mortar bar tests would serve as screening tests. Failure to pass the mortar bar test would prompt consideration of testing by the longer-term concrete prism test. One note of caution has been highlighted – the test fails to detect the reactivity of aggregates containing more than 2 per cent porous flint. The lack of expansion may be due to the accommodation of the reaction product within the unexpanded mortar bar. The testing of combinations is important and the existence of a possible pessimum content may need to be explored. The purpose of determining


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Figure 7.9 Indicative limits, accelerated mortar bar test

concentrations of any potentially reactive constituents during the petrographic examination was related to the importance of the pessimum effect. The pessimum could be explored by combining the test aggregate with an unreactive one in proportions ranging from 5 to 100 per cent. A proposal for an ultra-accelerated concrete prism test is being explored by the RILEM committee as RILEM Recommended Test Method AAR-4, based on work by French researchers. This would involve monitoring the expansion of concrete prisms stored at 60°C and could be used as a screening test; a test for establishing the alkali threshold; or as a performance test. Limiting the alkali content It may not always prove possible to make economic use of an innocuous aggregate and so consideration may need to be given to control of the alkali content of the mix. Guidance documents usually provide two routes for satisfying requirements in respect of alkali control: • •

limit the alkali content of concrete; limit the alkali content of cement.

The approach of limiting the alkali content of concrete is related to the potential existence of a threshold value at which an alkali–aggregate reaction

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will be triggered. A review of guidance documents from various European countries reveals that limits on the alkali content of concrete range from 2.5 to 5.5 kg/m3. At first sight this seems to be a very wide range but closer examination of the conditions attaching to the calculations reveal some convergence towards a narrower band for the majority of concretes. The reason for the wide range is due to local differences in conditions attaching to the calculation, concern about highly reactive aggregates, and the beneficial influence of high levels of pfa or slag. A specifier who is used to the rules of one country would need to be mindful of certain differences in calculating the alkali content of the concrete or in establishing the appropriate limit if applying the rules of another country. Essentially the alkali content of the concrete is computed as the product of the alkali content of the cement and the cement content. Additional alkalis from other sources must then be added to take account of: • • • •

the effective sodium oxide equivalent of any pfa, typically 17 to 20 per cent; the effective sodium oxide equivalent of any slag, typically 50 per cent; 0.76 times the chloride ion content of the aggregate, taken as the percentage by weight; the sodium oxide equivalent of any admixtures.

However a number of differences in conditions from country to country need to be highlighted. The first point is that comparisons of national alkali limits of concrete are complicated by the fact that some require the addition of an allowance for variability while others do not. Variability in the alkali content of cement, pfa and slag is discounted in some countries but others require the addition of 1.64 standard deviations to the mean value. One specification for highway structures required the addition of a multiple of two standard deviations. The second point is that some countries quote limits on the alkali content of concrete which include an allowance of 0.20 kg/m3 for alkalis from external sources. The limit is then revised downwards if the summation of external alkalis exceeds this figure. Third, some guidance documents do not require inclusion of any sodium oxide equivalent if pfa or slag is used at levels in excess of 25 or 40 per cent respectively. Fourth, some specifications may require the addition of a figure to the cement content, for example ten kilogrammes for each cubic metre of concrete, to allow for batching tolerances. Fifth, guidance documents may have an upper bound on the cement content within which the guidance is valid. Finally, one guidance document includes a matrix of limits on the alkali content of concrete depending on the combination of aggregate reactivity (categorised as ‘low’, ‘normal’ or ‘high reactivity’) and the alkali level of the cement (categorised as ‘low-alkali’, ‘moderate-alkali’ or ‘high-alkali’). The limit for aggregates of high reactivity in the United Kingdom (Building Research Establishment 1999b) is 2.5 kg/m3 and this must be further reduced by the amount of alkali contributed from other sources


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(aggregates, admixtures etc.) if appropriate. Such a limit appears to be quite onerous but a testing protocol has been prepared (British Cement Association 1999) whereby the limit may increase, to a maximum of 3.5 kg/m3, if expansion limits are not exceeded. The test method involves a modified version of the British Standard Concrete Prism Test (BSI 1999a: BS 812-123). The limits quoted at the upper end of the European range (5.5 kg/m3) typically apply to the use of pfa or slag concretes with minimum percentages required. The limits quoted at the lower end of the scale (2.5 kg/m3) apply to the particular case of the highly reactive greywackes. Consideration of the variations applicable from country to country leads to a band for most concretes of 3.0 to 3.5 kg/m3. Thus an alkali limit of about 3.5 kg/m3, with no allowance for variability, is typical for CEM I concretes in Europe. The other approach available in limiting the alkali load is to limit the alkali content of the cement. It is widely acknowledged that damaging ASR has not been observed in concretes made with low alkali cement. Cement with an alkali value of 0.6 per cent or less is regarded as qualifying. Some countries include an allowance for variability of 1.64 standard deviations to the 0.6 figure in national guidance documents. Others use the value without such an allowance for variability but impose the additional condition that the alkalis contributed from sources other than the cement must not exceed 0.60 kg/m3.

Summary Damage due to alkali–silica reaction is a rare occurrence worldwide but many countries have established guidelines to minimise the risk of its occurrence in future works. The designer must, however, be mindful that the implementation of these guidelines must not compromise other durability considerations. The topic has been extensively researched but international variations in materials, geology, and climate has frustrated efforts to synthesise the data. Nevertheless concrete technologists and scientists continue to research the topic and a clearer focus on universal test methods is emerging. Criteria for the interpretation of the test results may, however, have to remain a matter for national consideration for some time. This is recognised in the absence of a harmonised approach in European Standard EN 206–1. However a common principle may generally be discerned in the diverse national guidance documents. The principle is that the risk of damaging ASR may be minimised by one of the following: control of the availability of moisture to the member; the use of aggregate combinations which are unlikely to be reactive; or limiting the alkali content through use of a suitable mix design or use of a low-alkali cement. The degree of dependence on any one of these depends in large measure on the economic consequences of meeting the restrictions in any particular region or on a specific project. The prospect of convergence of national approaches is likely as further research yields a greater understanding of the phenomenon but local geological and climatic conditions will probably preclude total harmonisation.

8

Freeze/thaw effects

Background Concrete exposed to low temperatures on a cyclical basis is at great risk of durability failure if appropriate measures are not taken. The problems stem from the freezing of a proportion of the pore water in the cement paste or from ice formation in susceptible aggregates. Freezing of water is accompanied by expansion and in concrete this has the potential to be disruptive. Ice will wish to occupy a volume approximately 8 per cent greater than the water from which it was formed. This leads to movement when ice is formed within concrete. The ice forms a moving front and pushes water ahead of it. The concrete will be durable if the movement can be accommodated until relief is provided from a thaw. However expansive pressures build up if movement is restrained due to insufficient pore volume. Disruption will occur if the tensile strength of the concrete – its Achilles heel – is exceeded by the expansive forces. An enlarged pore network and microcracking results and this may permit entry of more water, which in turn can lead to greater ice build-up. The effect is thus cumulative. Concrete subjected to a significant number of freeze/thaw cycles permits penetration of increasing amounts of water following a thaw leading to increased demands on the volume of the pore network during freeze. Durability failure may ensue. Figure 8.1 illustrates the state of a cube after repeated freeze/thaw cycles in laboratory tests. The most vulnerable structures are those which have a high degree of saturation during cold weather. These principally include kerbs, parapets, copings, ledges, dams and other hydraulic structures. The number of cycles is more significant than the absolute lowest temperature. Thus concrete in Arctic regions, frozen for months at a time, is less at risk than concrete located further south which experiences higher average temperatures but more frequent cycles above and below freezing point. The degradation process is thought to be more complex than mere accommodation of the expanded volume. Ice formation begins in the capillary pores. An ice body forms locally in a cavity from pure water until the cavity is filled. The excess unfrozen water and air is pushed out into the surrounding

Freeze/thaw effects

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Figure 8.1 Freeze/thaw damage to a concrete cube following cyclical temperature variation in laboratory tests

network. Resistance to this flow develops a hydraulic pressure. Furthermore the expelled water contains not only existing solutes but also those driven from the now frozen pure water. This increase in solute level leads to a concentration gradient and osmotic pressure. Diffusion of water, including gel pore water, towards a developing ice body results in further ice growth and expansive pressures. The solute concentration aspect may be exacerbated if road slabs are subjected to de-icing salt application in cold weather. Damage may be manifest in one of three ways: • • •

internal cracking with subsequent D-line cracking or disintegration; scaling; pop-outs.

Internal cracking and disintegration results from repeated exposure of the cement paste to the damaging effects of expansive ice formation and related pressures, both hydraulic and osmotic. D-line cracking may be observed in the zone adjacent to free edges and joints in pavement slabs. It is caused by expansion of coarse aggregate particles at depth. The cracks initially appear near to and parallel to the edge or joint. Subsequent deterioration cycles cause additional cracking progressively further from the joint. Surface scaling is characterised by delamination of the surface due to the layered effect of frozen zones. Harrisson (1995) described the problems that

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occur in pavements where the ground under the slab may remain frozen for days at a time. The concrete at depth will remain frozen but the surface layers may thaw during the day. Fall in temperature at night leads to the development of an ice front at the surface that moves towards the lower frozen layers. A lens of water is trapped between the two frozen zones. If the temperature continues to fall the lens may expand and delaminate the surface. Scaling associated with de-icing salt application is another significant problem. First, the melting of thin ice sheets on the surface by the application of salt requires energy which is drawn from the concrete. This leads to a rapid temperature drop near the surface. The consequent thermal shock may cause cracking and subsequent scaling. Second, the melt water which contains de-icing salts penetrates the surface. One consequence of this is the depression of the freezing point of the water near the surface. Rainwater in the uppermost surface layers will therefore freeze in advance of that below. This can lead to differential expansion of the surface layers with consequent scaling. Pop-outs are caused where freeze/thaw-susceptible coarse aggregate is used. Certain coarse aggregate particles may contain significant amounts of water when saturated and swell on freezing. Expansive pressures may be high enough when ice forms for the aggregate to fracture. Particles close to the surface may disrupt the cover concrete sufficiently for conical shaped pieces of concrete to detach from the surface, exposing fractured aggregate at the base of the pop-out. The durability of concrete subject to significant freeze/thaw cycles is a source of particular concern to specifiers in countries where the risk of serviceability failure is high. The problem is being actively researched and the interested reader is referred to research papers presented to RILEM Technical Committee TC117 compiled by Marchand et al. (1997).

Primary factors of influence The factors of influence on freeze/thaw behaviour and damage are the degree of saturation, available water volume, pore structure, concrete age, climatic conditions, aggregate characteristics, and the effect of de-icing salts. Degree of saturation Saturation can occur in concretes that are permanently exposed to water and in concrete occasionally exposed to water under pressure. Concrete exposed to de-icing salts may also become saturated due to the effect of freezing point depression. The volume expansion of about 8 per cent occurring on conversion of water to ice cannot be accommodated in fully saturated concrete. Clearly the more air available in the pores the less damaging the effect. The critical threshold of degree of saturation, based on a Belgian study cited by Neville (1995) and others, lies close to 90 per cent. If the relative

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humidity of the atmosphere is less than 97 per cent (CEB 1989), some evaporation of water is possible during freezing which may allow sufficient space to be created for expansion without deleterious effects. Available water volume The greater the volume of water available the greater the potential expansion and hence the greater the risk of exceeding a concrete’s tensile strength. The total water volume is influenced by both the water present in the concrete after hydration and the subsequent availability of water from external sources. Low water/cement ratio concretes are beneficial in two ways in this context. First, the hydrated concrete contains less free water to begin with. Second, the impermeability of the pore network reduces the amount of external water capable of being introduced during the service life from external sources. The availability of water from external sources is influenced by the environment and by the detailed geometry of the structure. Features that promote a decrease in the contact time between water and concrete are thus beneficial. Details worthy of consideration are well-drained slabs, sloping wall tops, use of drips, and so forth. Pore structure The size of pore influences the temperature at which water will freeze. The water in the largest pores is the first to form ice blocks at a given temperature. Further drops in temperature cause the water in the capillary network to freeze but the gel pore water remains unfrozen. The Comite Euro-International du Beton (1989) design guide reports that due to the wide range of pore sizes in concrete only two-thirds of the pore water will be frozen at 60°C. The depression of the freezing point is due to the surplus energy at the pore surface and a thin film of water at the pore surface remains unfrozen even after ice is formed in the pore. The volume and proximity of spaces into which the expelled water may escape greatly influences the degree of resistance to damage. Thus a balance is required between high porosity and low permeability. For example Dhir et al. (2000) reported that the benefits of reducing the free water content, to lower expansive forces on freezing, may be partly offset by the restriction to movement caused by the improved microstructure. Air entrainment provides a controlled method of accommodating the water movement caused during ice formation. On thawing the water in the entrained voids is drawn back out into the narrower pore network by capillary forces. Concrete age Concrete is at its most vulnerable to frost damage at early ages due to its relatively high capillary pore water content and its low strength. Protection of

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Figure 8.2 Air entrainment: overlap of protected zones

immature concrete from the first freezing cycle is thus essential. The period of greatest risk is during the first week after a pour. Climatic conditions The absolute lowest temperature, period of freezing, and frequency of freeze/thaw cycles are all influential contributors to durability risk. The cumulative effect of freeze/thaw makes the frequency of cycle particularly significant but the other factors are noteworthy. The lower the temperature and the longer the freezing period the greater the volume of pores exposed to potentially damaging conditions. Concrete has a low rate of heat transfer and therefore significant penetration of the ice front requires very low temperatures, long periods of freezing, or a combination of both. It is worth reiterating however that the number of freeze/thaw cycles per annum is more significant than the period spent at low temperature. Aggregate characteristics Aggregates with high water absorption characteristics are at risk of contributing to freeze/thaw damage. Such aggregates may expand disruptively if they have a high water content in coincidence with freezing temperatures. Fracturing of shale, for example, is possible on freezing. The precise relationship between aggregate porosity and freeze/thaw resistance is not clearly established because porous aggregate can also contribute valuable space to accommodate frozen product. Aggregate size also has an influence on freeze/thaw behaviour. The smaller the maximum aggregate size the greater the fraction of cement paste per unit

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volume. This potentially increases the volume of pore water, the expansion of which would have to be accommodated. De-icing salts De-icing salts can influence the degree of saturation and may also subject concrete surfaces to damaging thermal shock. The degree of saturation of pores in service depends on both the exposure to water and the exposure to drying conditions. The smaller the pore the greater the surface tension holding in the water. Thus for typical service drying conditions the smaller pores remain more saturated. One effect of de-icing salts is to increase the surface tension forces thus limiting the drying effect and increasing the likelihood of saturation. The degree of saturation at a given temperature is therefore higher in salt-contaminated pore water than in concrete free from the influence of de-icing salts.

Air entrainment The concept of enhanced freeze/thaw durability by air-entrainment involves consideration of both the content of entrained air and its distribution. Entrained air bubbles protect the surrounding spheres of concrete. Effective use of air entrainment involves the achievement of a distribution of bubbles such that the protected spheres overlap as illustrated in Figure 8.2. Thus a large number of small air voids is effective whereas an equivalent air volume in the form of a small number of large voids is not. The air content required depends on the volume of frozen water to be accommodated. The volume of water is a function of the permeability and porosity and can be minimised through specification of a suitably low water/cement ratio. Air entrainment is achieved by adding an admixture to the concrete that is capable of distributing discrete air bubbles. The admixture is a surfactant, a surface-active agent, which stabilises the air bubbles formed during mixing and distributes them uniformly through repellent forces. Air entrainment is a common feature of American concrete practice and the required characteristics of entrained air voids have long been studied by the American Concrete Institute. Three parameters are relevant – air content, average spacing and specific surface. The mean air content required is approximately 9 per cent of the cement paste volume. This equates to about 5 per cent of the concrete volume in the case of concrete with a maximum aggregate size of 20 mm. Concretes with lower maximum aggregate size have higher cement paste volumes and consequently higher air content requirements. The converse applies in the case of larger aggregate. Typical values are presented in Table 8.1 using values from a national standard (BS 5328: Part 1: 1997) and a national annex to EN 206–1 (National Standards Authority of Ireland 2002). This table also serves to highlight a developing change in practice whereby minimum air content is

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Table 8.1 Variation of air content with maximum aggregate size Maximum nominal upper aggregate size (mm)

10 14 20 40

Recommended air content BS 5328: Part 1: 1997 (Mean total air content %)

National Annex to IS/EN 206–1: 2002 (Minimum air content %)

7.5 6.5 5.5 4.5

5.5 4.5 3.5 3.0

replacing target mean air content in European practice. European standard EN 206–1 places an upper limit on air content at the specified minimum value plus 4 per cent absolute. Thus the change in specification method does not imply any significant alteration to current practice. Although the total air content is significant it is of equal importance that the path length to an air bubble for the expelled water should be as short as possible. Thus average spacing of air bubbles is important and a maximum of 200 m is recommended. Specific surface is the surface area to volume ratio. Thus the higher the specific surface the smaller the air bubbles. It is recommended that the entrained and entrapped air should have a minimum specific surface of 25. The entrained air bubbles have typical diameters of 50 m. Research to date on the interaction of air-entraining admixtures with additions is rather limited. However a recent study of pulverised fuel ash (PFA) concretes by Dhir et al. (1999) yielded interesting results. The chemical characteristics of the admixture had a significant effect, implying that not all would be suitable in PFA concretes. PFA concretes with vinsol resin admixtures performed satisfactorily in freeze/thaw trials but the admixture demand was higher than for comparable Portland cement concretes. The demand increased by a factor of two for air contents of up to 4.5 per cent. A further increase in demand occurred with higher air contents. Demand increase was also demonstrated for pulverised fuel ashes with high loss on ignition. Finally, it is well established that air entrainment leads to a certain sacrifice in concrete strength. This factor has to be taken into account when framing specifications that include requirements both in respect of air content and minimum strength class.

Developments in specification and design practice Scientific models of freeze/thaw behaviour are not yet well developed. Thus the manipulation of significant variables in designing concrete for freeze/thaw durability through the use of mathematical models in service life prediction is

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Figure 8.3 Schematic representation of the event-driven nature of freeze/thaw damage

not possible in the immediate future. Therefore the use of prescribed rules will continue to be a feature of European concrete practice in respect of freeze/thaw durability. Nevertheless Fagerlund (1997) has outlined two possible approaches for durability design based on the degree of saturation. Difficulties in modelling freeze/thaw behaviour Two immediate hurdles are apparent in attempts to model freeze/thaw. First, there is the unpredictable pattern of damaging cycles. Second, it is difficult to verify the apparent influence of mix parameters in laboratory tests with experience of structures in service. Freeze/thaw durability failure is event-driven, as illustrated in Figure 8.3, and does not follow a predictable time-related pattern. The level of damage per cycle could vary from minor growth of internal cracks (for example damage ‘x1’) to delamination of a piece of concrete of unpredictable size (for example damage ‘x2’). Equally, the rate of damage will vary from year to year depending on local climatic conditions. Although average temperatures are reasonably predictable, the number of freeze/thaw cycles in a period is much less certain. Additionally in many locations the dwell period between series of cycles (for example period ‘td’) could vary from a few months to over a year – a harsh winter versus a mild winter – such is the unpredictability of the weather. The second problem relates to translation of laboratory research findings with practice. Laboratory trials are generally conducted with daily cycles in a range from 20°C to 20°C. Although comparative performance of different concretes in laboratory trials is informative the absolute performance

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in these trials cannot be readily translated to field experience due to climatic variability, which influences both temperature and degree of saturation. Durability design and service life modelling The possible direction of models for durability design and service life prediction has been outlined by Fagerlund (1997). The models are based on the concept that the concrete will fracture at a critical degree of saturation. This critical degree of saturation is a function of the material and is a consequence of the distance between a freezing site and the nearest air-filled space. The service life is also a function of the wetness of the environment. The capillary water uptake and long term water absorption of the air pore system are the significant issues and are modelled by a time relationship. The potentially damaging stress on the concrete pores due to freezing increases with increasing distance between the freezing site and the nearest air-filled space. This gives rise to the concept of a critical distance (DCR) in respect of freeze/thaw damage. This can be envisioned, for example, as the thickest possible water saturated cement paste zone around an air-filled void, such as a sphere. This leads to the concept of a critical spacing factor (LCR). These parameters are related by the following equation: DCR 2 LCR

[2L9

1

CR

]

0.5

where specific area of the enclosed air void. It has been determined experimentally that the mean critical thickness (DCR) is 1.2 mm for freeze/thaw in pure water and 1.8 mm in a 3 per cent NaCl solution. These values yield critical spacing factors (LCR) of 0.40 mm and 0.54 mm respectively, assuming a specific surface () of 15 mm1. The values are valid for water/cement ratios in excess of 0.45. The spacing factor (L) of a system of spherical pores in a material matrix may be determined from Power’s equation: L

[ (

)

3 V 1.4 P 1 a

0.3

1

]

where specific area a volume of the pore system, VP material volume, excluding the pore volume. The equation may be rewritten in the form: a

VP L 0.364 1 3

[

]

3

1

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Fagerlund (1997) demonstrated that the equation must be modified to take account of the fact that the air pores will not be completely dry. Therefore the required air content (a0) is as follows: a0 aw aCR ab where aw water-filled air pore volume aCR air pore volume needed to prevent the spacing LCR being exceeded when aw is reached ab safety margin The minimum air content required is therefore: (a0)min aw aCR and the value may be determined from the following formulae: aw Sa (a0)min where Sa is the degree of water-filling of the air pore system and aCR

VP LCR CR 0.364 1 3

[

]

3

1

The safety margin ab may then be added at a level that reflects the required service life. An alternative approach is to determine a potential service life based on a model of the capillary degree of saturation as a function of the suction time. The potential service life (tp) is defined by the critical level of saturation (SCR) where: SCR SCAP (tp) and the capillary degree of saturation (SCAP) is modelled by the formula: SCAP (t)

[1] [ C 0

D

0

a0(t) E ]

where total porosity 0 porosity exclusive of air pores C, D, E constants 0 specific area of the air pore system a0 total air content diffusivity of dissolved air

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Developments in prescriptive practice and testing Despite the slow progress in respect of mathematical modelling, a refinement of prescriptive practice has been made through the introduction of the deterioration-specific exposure classes in EN 206–1. This is likely to yield enhanced durability in future works by specifically focusing attention on the issue. Efforts to harmonise performance tests for freeze/thaw resistance are also bearing fruit. Harmonised procedures for testing scaling resistance, by monitoring material loss, are well advanced. Test methods for detecting internal deterioration, for example by a decrease in dynamic modulus of elasticity, are less well established. Additionally the appropriateness of the test methods for high performance concrete is open to question. Further research on these topics is clearly of great interest, particularly in the context of performance-related specifications.

Freeze/thaw attack: exposure categories in EN 206–1 Resistance to the adverse effect of freeze/thaw cycles is achieved through controlling the pore distribution in concrete. Two approaches are available in respect of the pore distribution. One approach involves reducing the permeability and porosity of the concrete to such an extent that the free water content will be so low that the amount of expansion will not be deleterious. The second approach involves increasing the porosity of the concrete in a controlled manner through air-entrainment so that the pores can act as safety valves, providing space for the freezing water to expand without stressing the concrete. The benefit of restricting the amount of free water in the concrete, from either excess mix water or later ingress from external sources, applies in both cases. Thus even the second approach involves the need for specification of moderately low water/cement ratio concrete. The exposure classes in EN 206–1 take account of the degree of saturation and the presence, if any, of external sources of salt from de-icing agents and seawater. The presence of salt increases the risk of damage due to the detrimental influence of solute concentration gradient. The parameters for limiting values of concrete composition include minimum cement content but some authors have questioned the need for this. Dhir et al. (2000), for example, have failed to correlate cement content and freeze/thaw resistance in non-air entrained Portland cement mixes and state that no data are available on cement content’s role in the resistance of air-entrained concrete. A matrix of four exposure classes have been distinguished in respect of freeze/thaw attack and these are designated as XF1, XF2, XF3, and XF4. The matrix is based on the two degrees of saturation and the absence or presence of salt. The degrees of saturation are ‘moderate’ and ‘high’. At first sight the use of such terms for differentiating exposure classes may cause some confusion. Moderately saturated concrete should not be at risk during a

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freeze/thaw cycle because the water would have room to expand without deleterious effect. Harrison (2000) explained that CEN intended the term ‘moderate saturation’ to imply a moderate risk of damage. This differentiates the moderate risk classes (XF1 and XF2) from the higher risk classes (XF3 and XF4). Hobbs et al. (1998a) noted that the differentiation could be on the basis of a lower number of freeze/thaw cycles per annum or a lower risk of freeze/thaw when saturated. Such an interpretation, they note, would be in accord with the examples given in the European standard. The informative annex in EN 206–1 presents indicative limits for maximum water/cement ratio, minimum cement content, minimum strength class, minimum air content (except for class XF1) and a requirement for freeze/ thaw resisting aggregates for each exposure condition, based on an intended working life of fifty years. It must be emphasised that these are informative values only, serving as useful comparative benchmarks. The values are based on the mean of values currently representative of European practice for CEM I concrete. They are not appropriate for all countries, nor for intended working lives that differ significantly from fifty years. Specifiers must consult the appropriate advice on limiting values in national annexes or complementary standards valid in the place of use. Regarding working life, Hobbs et al. (1998a) argue that the limiting values for a 100-year life should be the same as for fifty years. This is due to the eventdependence of the process and the fact that concrete capable of resisting freeze/thaw events on an on-going basis should be durable, irrespective of age. Exposure class XF1 Exposure class XF1 covers concrete exposed to significant attack by freeze/thaw cycles whilst wet and described as having ‘moderate water saturation, without de-icing agent’. Examples of concrete at moderate risk of damage in this context include the vertical surfaces of members exposed to rain and freezing. The risk to durability posed by freeze/thaw in these conditions may be regarded as the lowest of the four freeze/thaw classes. Thus the limiting values on concrete composition may reflect this through values which extend to the least onerous boundary of satisfactory experience. In the case of XF1, the informative annex to EN 206–1 shows a non-air-entrained concrete with a minimum cement content of 300 kg/m3 and the maximum water/cement ratio as 0.55. The latter value is of greater interest given the lack of agreement on the influence of cement content on freeze/thaw resistance. Experience in some countries indicates that water/cement ratios up to 0.60 would provide protection to the concrete. More onerous requirements, such as limiting water/cement ratio to 0.55, may apply where freeze/thaw cycles are sufficiently numerous to threaten the integrity of the cover to reinforcement. The informative annex to EN 206–1 shows a requirement for freeze/ thaw resisting aggregates in accordance with the recommendations of the

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European aggregate standard. Unanimity on this requirement is not apparent in national annexes and complementary standards in respect of exposure class XF1. Exposure class XF2 Exposure class XF2 covers concrete exposed to significant attack by freeze/thaw cycles whilst wet and described as having ‘moderate water saturation, with de-icing agent’. The inclusion of de-icing agents in class XF2 leads to examples of concrete in this context as the vertical surfaces of road structures exposed to freezing and airborne de-icing agents. The informative values for class XF2 in EN 206–1 indicates a typical requirement of the same cement content, 300 kg/m3, and maximum water/ cement ratio, 0.55, as in class XF1 but with the inclusion of a minimum air content of 4.0 per cent. Consequently the minimum strength class decreases. A requirement for freeze/thaw resisting aggregates in accordance with the recommendations of the European aggregate standard is also included. National annexes and complementary standards contain appropriate values for the place of use. A maximum water/cement ratio of 0.55 is typical but high minimum strength classes are also presented in some national guidance documents as an alternative to air-entrained mixes. The use of non-airentrained mixes is also provided for in the informative example of EN 206–1 through comparative testing using the European freeze/thaw test which is under development. Exposure class XF3 Exposure class XF3 covers concrete exposed to significant attack by freeze/thaw cycles whilst wet and described as having ‘high water saturation, without de-icing agent’. Examples of concrete in this context include horizontal surfaces of members exposed to rain and freezing. The informative values for class XF3 in EN 206–1 indicates a typical requirement for an increase in cement content of 20 kg/m3 and a decrease in the maximum water/cement ratio by 0.05 in comparison with class XF1 but with the additional requirement for air entrainment at 4.0 per cent. A requirement for freeze/thaw resisting aggregates and provision for non-air entrained mixes, through testing, is also included. Experience in some countries indicates that the use of non-air entrained mixes of minimum strength class C40/50 can provide adequate resistance. Exposure class XF4 Exposure class XF4 covers concrete exposed to significant attack by freeze/thaw cycles whilst wet and described as having ‘high water saturation, with de-icing agent or seawater’. Examples of concrete covered by class XF4

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include bridge decks, surfaces exposed to spray containing de-icing agents, and surfaces in the splash zone of marine structures. Some difficulty may be experienced in distinguishing between class XF2 and XF4 for some surfaces of highway structures. It may help to recognise that XF4 is the severest class in the context of freeze/thaw exposure. The informative values for class XF4 in EN 206–1 show a 40 kg/m3 increase in the minimum cement content and a decrease in maximum water/cement ratio of 0.10 in comparison with class XF1 but with a requirement for air entrainment at 4.0 per cent. More pertinently this represents a 20 kg/m3 increase in the minimum cement content and a decrease in maximum water/cement ratio of 0.05 (to 0.45) in comparison with class XF3. Not surprisingly there is a requirement for freeze/thaw resisting aggregates. Provision for non-air entrained mixes, through testing, is included. As a generality it may be observed that exposure class XF4 should demand the use of a concrete composition which has demonstrated satisfactory performance under the harshest freeze/thaw conditions of a given region. As always, the specifier is referred to appropriate national guidance documentation.

Developments in testing Tests methods related to freeze/thaw can be categorised into two groups: • •

tests on fresh concrete to verify air entrainment; tests on hardened concrete.

The first group of tests is concerned with the testing of concrete prior to use on site to verify that the specified level of air entrainment has been achieved. The second group is related to the assessment of anticipated performance in service. Testing of fresh concrete Current specifications commonly state the mean volume of air to be entrained as a percentage of the concrete. Although this is the specification requirement, what is intended in service is a certain volume of air as a percentage of the cement paste and that the air is properly distributed through a set of bubbles which conform to a limited diameter range and spacing. A common test method on fresh air-entrained concrete is to check the total air content using a pressure method which, through Boyle’s Law, can be directly related to air content. Such a test is useful for checking production consistency but does not provide any indication of the manner of distribution of the air bubbles. An alternative method has recently been developed in Denmark that can measure air content, spacing factor and specific surface in about twenty minutes. The Danish method has been described by Price (1996). It is based on the principle that the rate of rise of an air bubble in water is related to its size. The

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Figure 8.4 Test set-up for freeze/thaw resistance: Scandinavian method

technique involves sampling paste from fresh concrete and injecting it into a viscous liquid at the base of a column of water. The air bubbles are released and rise through the column where they strike a plate. The change in buoyancy of the plate with time can be used to determine air content, specific surface and spacing factor. The test method allowed inclusion of air content limits (entrained and entrapped) and minimum specific surface value in the specification of concrete for the Storebaelt West Bridge project. Testing of hardened concrete Various test methods are available for comparative studies of hardened concrete in the context of freeze/thaw resistance. These include an American method using concrete prisms or cylinders, a Scandinavian method using 50 mm thick slabs cut from cubes, a German method using cubes, and a RILEM method based on 70 mm thick specimens cast in cube moulds. The American method (ASTM C666) involves exposing the test specimens to about 300 freeze/thaw cycles in water in a temperature range of 4.4°C to 17.8°C. Alternatively the specimens may be frozen in air and thawed in water. The dynamic elastic modulus is monitored after various numbers of cycles to yield a ‘durability factor’. A European standard is being developed based on Swedish Standard SS 13 72 44 (Swedish Standards Organisation 1995). This involves preparation of the test set-up illustrated in Figure 8.4. Specimens 50 mm thick are sawn from 150 mm cubes. Rubber is glued to the faces other than the sawn test face. The

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Figure 8.5 Test set-up for freeze/thaw resistance: German cube test

specimen is sealed into an insulated mould and covered by 3 mm of water or 3 per cent sodium chloride solution. A polyethylene sheet is used to prevent evaporation. A daily freeze/thaw cycle is achieved by exposing the specimen to a temperature range of 20°C (4°C) to 20°C (2°C) in accordance with a specified time-temperature curve. The amount of scaled material is assessed after 7, 14, 28, 42 and 56 cycles. The cumulative value after 56 cycles is used for evaluating the scaling resistance. The German cube test involves immersing two pairs of 100 100 100 mm cubes in a freezing medium of de-mineralised water or 3 per cent sodium chloride solution. Each pair of cubes is stored in brass or stainless steel containers as illustrated in Figure 8.5. Fifty-six daily freeze/thaw cycles are conducted to a specified time-temperature curve. The temperature range is 20°C (2°C) to 15°C (2°C). Scaling is monitored at 7, 14, 28, 42 and 56 days. Scaling resistance is assessed by determining the cumulative scaled material after 56 cycles. Provision is made for a break in the daily cycle of tests for periods such as weekends when a cycle can be extended by leaving the specimens in a frozen state at 15°C (2°C). A RILEM Technical Committee are developing tests currently designated CF (Capillary suction and Freeze thaw test) and CDF (Capillary suction of Deicing chemicals and Freeze thaw test). The RILEM CF/CDF method involves casting two specimens per 150 mm cube mould by inserting a teflon plate vertically at the midpoint of the sides. This yields 70 mm thick specimens with 140 150 mm faces. The test surface, that cast against the teflon, is immersed in a freezing medium which in the case of the CF test is de-mineralized water and in the case of the CDF test is 3 per cent sodium

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Figure 8.6 Test set-up for freeze/thaw resistance: CF/CDF method

chloride solution. Five specimens are employed per test. Capillary suction is encouraged by immersing the test faces to a depth of approximately 5 mm in the freezing medium. The test set-up is illustrated in Figure 8.6. The specimens are subjected to two freeze/thaw cycles per day by cycling from 20°C to 20°C in each twelve hour period. Scaled material is detached from the specimens by use of an ultrasonic bath after 4, 6, 14 and 28 freeze/thaw cycles in the case of sodium chloride immersion, and after 14, 28, 48 and 56 freeze/thaw cycles in the case of de-mineralised water immersion. Scaling resistance is evaluated by determination of the mean values after 28 cycles (CDF Test) and 56 cycles (CF Test).

Specification by performance Durability failure in concrete due to freeze/thaw effects may take the form of surface scaling or disintegration due to internal deterioration. Freeze/thaw tests for both situations are being developed to serve as performance measures as an alternative to specification by limitations on mix composition. A note of caution has been sounded by the Concrete Society (1996) who noted that many existing freeze/thaw tests on hardened concrete are very severe and, in particular, non-air entrained mixes which have good service records in the United Kingdom may fail such tests. This emphasises the need to take account of service history when setting performance criteria. Further-

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more Siebel (1999) noted that the measured scaling in proposed European standardised methods has not yet been calibrated with behaviour in service and so only comparative testing is possible. Interest in a harmonised test for European practice has initially focused on developing a test for scaling resistance. This is currently referred to as EN FFF–1, entitled ‘Test methods for the freeze–thaw resistance of concrete – tests with water or with sodium chloride solution, Part 1: Scaling’. The reference method is based on the Scandinavian slab test but alternative methods based on the German cube test and RILEM CF/CDF test are also available subject to agreement between the parties involved. The reference method must be used in cases where agreement is not forthcoming or where there is doubt. Reference to the future use of the European scaling test, currently prEN FFF–1, is specifically made in Clause 5.3.3 of EN 206–1. It can be used to prove the adequacy of a non-air-entrained mix as an alternative to the limiting values for composition and properties of air-entrained mixes recommended for exposure classes XF2, XF3, and XF4. Freeze/thaw testing may serve as an introduction in some countries to performance-related design methods with respect to durability. For example the performance test for exposure classes XF1 and XF3 could be EN FFF–1 using water as the freezing medium; testing for exposure classes XF2 and XF4 could be EN FFF–1 with 3 per cent sodium chloride solution as the medium. The criteria for acceptance of an unknown concrete would be the loss of mass equal to or less than a reference concrete and no adverse indicators. The reference concrete would be produced containing aggregates with a local tradition of providing freeze/thaw resisting concrete, CEM I–42.5 cement, and the concrete would conform to a specified slump class such as S2 (50–90 mm). Reference concrete for testing air entrained mixes would also conform to specified values of water/cement ratio, cement content, and air content, dependent on the exposure class. Reference concrete for testing nonair-entrained mixes would also conform to specified values of minimum cube strength, maximum water/cement ratio, and minimum cement content, dependent on the exposure class. While this provides a clear route for introducing performance-related design methods with respect to durability, Harrison (2000) reports that it is not expected to be a route widely adopted in the near future.

Summary Freeze/thaw damage represents a significant proportion of concrete durability failure in countries where the annual number of freeze/thaw cycles can be significant. The parts of structures most at risk are those which can be saturated at time of freezing. Appropriate measures, such as the use of freeze/thaw resisting aggregates and air-entrainment can provide adequate resistance.

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Testing of fresh concrete tended to concentrate on verifying that the correct air content was present but it was not possible to verify that the distribution of the air content was appropriate. Recent advances in Denmark on testing fresh concrete are overcoming these difficulties. Mathematical modelling of freeze/thaw behaviour has not yet advanced and so it will be a long time before a probabilistic approach to design for freeze/thaw durability will be a reality. On the other hand a comparative test for freeze/thaw performance is leading the way for the introduction of performance-related design methods with respect to durability. The harmonised test methods will cover scaling and disintegration due to internal deterioration. The test methodology for the former is well advanced but it is uncertain how willing the industry will be to embrace change. Therefore limiting values for composition and properties of concrete will be the norm. Guidance based on local experience, published in national annexes or complementary standards to EN 206–1, will form the basis for most specifications. Research on the test methods for disintegration due to internal deterioration, mathematical modelling of the freeze/thaw phenomenon, and the interaction of air-entraining admixtures with additions, such as pulverised fuel ash, is required to advance concrete technology in this area.

9

Chemical attack: sulfates

Introduction Concrete may be damaged in a number of ways through contact with salts. Salts are chemical compounds typically formed by reactions between acids and bases and which dissociate in water into their constituent ions. The problems of reinforcement corrosion associated with sodium chloride are well documented but deterioration of the concrete itself may also arise through contact with sulfate salts. The term ‘sulfate attack’ has traditionally been associated with durability failure through disruptive expansion of concrete in contact with sulfatebearing soils or groundwaters. However the effect of sulfate may manifest itself in a number of ways. This chapter considers three phenomena involving sulfate and concrete interaction that can be differentiated through their chemical or chronological characteristics: • • •

‘classical’ sulfate attack; thaumasite sulfate attack; delayed ettringite formation.

The term ‘classical’ is used arbitrarily in this chapter to differentiate the most common form of attack from the two others. A fourth phenomenon, seawater attack, is also related to sulfate action and could be included here but since it also involves the action of acids it is discussed in Chapter 10. The classical form of sulfate attack is a problem generally associated with buried concrete exposed to soils or groundwater containing soluble sulfates. Deleterious conditions may also arise in sewers. Significant durability problems can occur where concrete elements are exposed to such salts in solution above a critical concentration level. Durability failure occurs through a combination of expansive disruption and deterioration of the cement paste. Tricalcium aluminate hydrate and calcium hydroxide can react with sulfate ions in solution to form solid products that occupy a larger volume than the source constituents. Ettringite and gypsum are the principal compounds formed. The disruptive expansion can be accompanied by strength loss

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consequent on the chemical deterioration of cement paste and damage to the aggregate interface bond. Concrete may crack parallel to the surface or become friable. The related but rarer phenomenon of thaumasite formation has received increased attention lately. Its potential deleterious effect has been known for some decades but recent cases of damage in some United Kingdom motorway bridges has heightened awareness. Thaumasite is formed through a sulfate reaction involving limestone in aggregate, filler, or groundwater. It is a complex mineral that can attack the vital calcium silicate hydrates thus rendering the concrete soft, weak, or mushy. Delayed ettringite formation is a form of internal sulfate attack which may occur at an advanced age in particular concretes. Portland cements contain internal sulfates and added gypsum to influence setting and early strength characteristics. Without the gypsum the reaction between tricalcium aluminate and water would lead to a flash set. Ettringite, an expansive compound involving tricalcium aluminate and gypsum, is normally formed at the hydration stage while the material is plastic and can accommodate the resultant strains. In particular conditions this internal sulfate may cause a phenomenon later in service. This phenomenon is quite distinct from the classical and thaumasite forms of sulfate attack. The problem arises if ettringite is not allowed to develop at the plastic stage due to high temperature curing conditions. Subsequent wet conditions in service may encourage ettringite formation in mature concrete with consequent expansion of the cement paste and cracking. High temperatures during curing may arise either through steam curing of pre-cast concrete products or through the significant effect of the heat of hydration in large pours. The European standard EN 206–1 includes sulfate attack as a subset of the multifaceted phenomenon of chemical attack. The standard includes three exposure classes in respect of chemical attack – slightly, moderately and highly aggressive – based on the wider chemical composition of the environment. The specific topic of sulfate action is considered in this chapter. Other forms of chemical attack, covered by the same exposure classes, are considered in Chapter 10.

Physico-chemical aspects Classical form of sulfate attack In the classical form of attack the source of the sulfate ion is typically sodium sulfate, calcium sulfate, magnesium sulfate, or potassium sulfate. Ammonium sulfate may also be encountered. These sulfates may be naturally occurring in soils, particularly clays, and groundwaters. Industrial processes, including those arising through the use of land for agricultural purposes, may also introduce sulfates. Sulfate attack may also occur if the aggregate contains accessible gypsum. Additionally, BRE Digest 363 (Building Research


181

Establishment 1996) notes that disturbance of clays bearing pyrite (iron sulfide, FeS2) may raise the sulfate and acid concentrations in the groundwater through oxidation. 14FeS2 (s) 14Fe3 8H2O 2SO42 15Fe2 16H Aggregates with sulfur contents and potential sulfate contents of less than 1.3 per cent and 4.0 per cent approximately, by mass of cement, respectively were reported as being unlikely to give rise to abnormal expansion when subject to moisture exposure (Hobbs 2000). Sulfates are also found in seawater and contribute to a form of attack considered separately in Chapter 10. The sulfate ion concentration depends on the solubility of the salts. Sodium and magnesium sulfates are highly soluble while calcium sulfate is not. The reaction occurs between the salts in solution and the products of hydrated cement. Salts in solution are ionised, that is, dissociated into their constituent ions. For example, potassium sulfate dissociates as follows: K2SO4 2K SO42 The sulfate ion is generally more significant in the attack but the magnesium ion plays an important role in the case of magnesium sulfate attack. Calcium hydroxide and calcium aluminate hydrates are the most vulnerable products of hydration but calcium silicate hydrates may also be affected if the calcium hydroxide is depleted. Sulfate ions react with calcium hydroxide to form gypsum (calcium sulfate, CaSO4.2H2O) while the reaction with the calcium aluminate hydrates form calcium sulfoaluminate hydrates. Calcium sulfoaluminate hydrates are formed at the hydration stage and are found as ettringite (3CaO.Al2O3.3CaSO4.32H2O) and monosulfates (3CaO.Al2O3.CaSO4.12H2O). The formation of ettringite may be described by the reaction: 3CaO.Al2O3 3CaSO4 (aq) 3CaO.Al2O3.3CaSO4.32H2O The ettringite formed may not be pure and its phase may be what is referred to as ‘AFt’, or ‘alumino-ferrite-tri’. The formation of the monosulfate occurs through reaction with the ettringite if more tricalcium aluminate is available than sulfate. Thus: 3CaO.Al2O3.3CaSO4.32H2O 2(3CaO.Al2O3) 3(3CaO.Al2O3.CaSO4.12H2O) The monosulfate is referred to as ‘AFm’, or ‘alumino-ferrite-mono’. If external sulfates are introduced any unreacted tricalcium aluminates may react to form ettringite while AFm phases may form AFt. The deterioration mechanism is generally considered (Neville 1995) to involve the generation of disruptive forces due first to the fact that the

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reaction products occupy a greater volume than the components causing the reaction and second due to the possible adsorption of water. The reactionproduct volume increase may be by a factor of two to five with the higher values being associated with the effect of ettringite formation. However Marchand and Skalny (1999) point out that the mechanism is a complex sequence of physical and chemical processes. The proportion of ettringite formed may not relate to the amount of expansion. Ettringite and gypsum crystal growth has been observed in cracks of sulfate-damaged concrete. It is not clear whether pressure from the crystal growth causes the cracks or if the cracks are sites for the deposition of ettringite and gypsum from solution. The ettringite first formed may be in a colloidal form and subsequently expands when it imbibes water. Other hypotheses have also been proposed. Reactions with sulfate ions may be one of equilibrium or may go to completion. For example the reaction with sodium sulfate could reach equilibrium when only a portion of the calcium hydroxide has changed to calcium sulfate. In simplified terms: Ca(OH)2 Na2SO4 ⇔ CaSO4.2H2O 2NaOH The intense attack that occurs when a reaction can go to completion is illustrated by the effect of magnesium sulfate. In simplified terms: Ca(OH)2 MgSO4 → CaSO4.2H2O Mg(OH)2 Lea (1998) provides an example of ettringite formation: 2(3CaO.Al2O3.12H2O) 3(Na2SO4.10H2O) → 3CaO.Al2O3.3CaSO4.31H2O 2Al(OH)3 6NaOH 17H2O Sodium sulfate (Glauber’s salt, Na2SO4.10H2O) reacts with calcium hydroxide and with calcium aluminate hydrate to form gypsum and ettringite, respectively. The calcium silicate hydrates may also be attacked. Ammonium sulfate, (NH4)2SO4, can also react, forming gypsum. Calcium sulfate (gypsum, CaSO4.2H2O) reacts with calcium aluminate hydrate to form ettringite. The solubility of calcium sulfate is considerably lower than the solubility of other sulfates and so it can be relatively harmless in soils. The sulfate reaction may however be initiated by calcium sulfate formed from reaction with calcium hydroxide. Magnesium sulfate (Epsom salt, MgSO4) reacts with calcium hydroxide, calcium aluminate hydrates and calcium silicate hydrates to form gypsum, ettringite, brucite (magnesium hydroxide, Mg(OH)2) and magnesium silicates. It is an intense attack because the brucite is insoluble in water and therefore the reaction may go to completion. Hobbs (2001) notes that the attack on the calcium silicate hydrates may be significant due to the effect of the brucite formation in lowering the pH of the pore solution.


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Thaumasite sulfate attack Thaumasite formation results from a reaction between sulfates, calcium silicates in the cement, and calcium carbonate. The carbonate may originate in limestone aggregates, limestone filler in cement, or groundwater percolation. The sulfates typically derive from external sources. Alternatively it may form from a reaction involving ettringite, calcium silicate hydrates and calcium carbonate. If this is the case ettringite formation is a necessary precursor implying that normal sulfate attack may be a prerequisite. Initial reactive alumina is another necessary condition for the reaction. Involvement of the calcium silicate hydrates in the reaction leads to loss of strength and cohesion in the concrete leading to disintegration. Thaumasite has been described (Hartshorn and Sims 1998) as a complex sulfate-bearing mineral with the composition CaSiO3.CaCO3.CaSO4.15H2O. The material has a similar crystal structure to ettringite but thaumasite is a silica-bearing phase as opposed to an aluminate. The composition has been described in an expert group report (Department of the Environment, Transport and the Regions 1999) as a calcium silicate sulfate carbonate hydrate. It may additionally be described as Ca6 (Si(OH)6) 2(SO4)2(CO3) 2.24H2O. Thaumasite formation appears to occur more readily in the presence of magnesium sulfate than sodium sulfate (Hartshorn et al. 1999) but is less likely to form in calcium sulfate solutions. Samples of deteriorated concrete may clearly show white haloes of thaumasite around the affected aggregate particles (Wallace, 1999). In the case of some of the deteriorated concrete bridge foundations in the United Kingdom, Hobbs and Taylor (2000) postulated that oxidation of pyrites in excavated clay backfill led to a pH reduction in the groundwater through formation of sulfuric acid. The sulfuric acid could become depleted by neutralisation within the backfill, by contact with concrete, and by washout. However, the acid’s sulfate ions may remain in solution in the clay’s interstices unless sufficient carbonate is present to both neutralise the acid and precipitate its sulfate: MgCO3(s) H2SO4(aq) H2O → MgSO4(s) 2H2O CO2(g) The groundwater sulfate level could rise, depending on the sulfates formed, leading to gypsum formation. It is possible that further reactions could include reaction of gypsum with calcium silicate hydrates and calcium carbonate to form thaumasite. Most of the thaumasite cases in the United Kingdom motorway bridges involved the presence of pyrite. Delayed ettringite formation Ettringite is formed during hydration in plastic concrete except in high temperature situations. It is unstable at temperatures in excess of about 70°C.

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Higher temperatures can be experienced during the curing period of large pours or through steam curing of pre-cast units. In these particular cases ettringite is not formed. The sulfate is instead absorbed by the calcium silicate hydrate, forms small amounts of monosulfate and some may go into solution in the pore fluid. The alumina content of the calcium silicate hydrates also becomes elevated. These phases are not stable at ambient temperatures. Thus when the concrete cools in service ettringite crystals can begin to form in the paste structure. This only occurs in wet concretes. Expansion of the cement paste may result followed by associated cracking. Hobbs (1999b) reports that the cracks are often of uniform width and proportional in width to the aggregate particle that they surround. The cracks may fill with ettringite.

Factors influencing sulfate attack Long term tests by the United States Department of Agriculture, United States Bureau of Reclamation, the United Kingdom Building Research Station/Building Research Establishment and the British Cement Association have contributed significantly to advances in knowledge and experience on resistance to sulfate attack. The American work pioneered the interest in low tricalcium aluminate sulfate resisting cements while the British research helped to broaden the range of factors which may be taken into account in design and specification. For example, a twenty-five year exposure trial of concrete specimens buried below the water table of a sulfate-bearing soil in west London was completed in the 1990s. The average SO42 content of the groundwater was about 3100 mg/l. Interim results were influential in the revision of earlier BRE guidance documents which were replaced by Digest No. 363 (BRE 1996). The document has been further revised as Special Digest No. 1 (BRE 2001a) to take account of the thaumasite form of sulfate attack and to clarify the ground sulfate classification system, which takes account of factors other than just the sulfate content. Influences on classical forms of sulfate attack Research has identified the following factors that have an influence on the intensity of sulfate attack: • • • • • • •

sulfate concentration; solubility of sulfates; groundwater mobility; concrete permeability; wetting and drying cycles; evaporation; degree of carbonation prior to exposure. A threshold value of sulfate concentration is required to initiate attack.


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Figure 9.1 Influence of groundwater sulfate concentration on relative resistance requirements

Attack intensity increases with increasing concentration up to a certain level above which the rate of increase diminishes. This may be illustrated by the relative minimum cement content requirement that applied in the United Kingdom during the 1990s for resistance to sulfate in groundwater (Figure 9.1). The solubility of sulfate salts in groundwater contributes to the critical sulfate concentration. Sulfate concentration in service, however, may on occasion differ from that determined in tests to classify the site. This is due to the fact that the method of determination, by acid extraction, yields a value for the total sulfates. Thus the method does not distinguish between relatively innocuous calcium sulfate, for example, and the highly aggressive magnesium sulfate. Water-soluble sulfates can provide a better guide to the risk of attack on a particular site. This is reflected in United Kingdom practice that includes a water/soil extract test. Groundwater mobility influences the extent of sulfate supply in a reaction. Static groundwater conditions are therefore less problematic than those in which groundwater movement leads to refreshment of sulfate supply. Equally a site experiencing groundwater movement could be classified as low in sulfates at one time but experience elevated sulfate levels at a later stage. European Standard EN 206–1 differentiates static and mobile groundwater through a threshold soil permeability value of 105 m/s. The ingress of sulfate ions depends on the quality of the concrete. Resistance to sulfate attack is critically dependent on reducing the rate of sulfate ion ingress. Poorly compacted concretes are not sufficiently impermeable to provide protection. Highly impermeable concrete is especially required where the surfaces are in contact with sulfates in solution under the influence of

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a hydrostatic head. It is unsurprising that sulfate-resisting concretes are generally characterised by low water/cement ratios. Concrete in sulfate conditions subject to water table level fluctuation may experience higher rates of deterioration if wetting and drying cycles are experienced. Wetting and drying promotes greater sulfate ingress than conditions in which the concrete is constantly saturated. Sulfate attack is unlikely to be significant if the water table remains predominantly below the underside level of the concrete during the service life. The rate of attack is higher in members where moisture is lost through evaporation. Thick members are therefore more robust than thin ones. Elements with exposure to sulfates on all sides will deteriorate less than those which allow moisture loss from one or more faces. Ground floor slabs can allow evaporation from the top surface encouraging further ingress of sulfates from the fill below. Carbonated concrete is not subject to the same intensity of sulfate attack as non-carbonated concrete. The reason for this is that carbonation of the surface layers reduces the calcium hydroxide content through conversion to calcium carbonate. The insoluble calcium carbonate does not react with the sulfates. Thus if concrete is carbonated prior to sulfate exposure the rate of attack can be lowered. This may be of significance in practice where pre-cast concrete units or concrete blocks are used in contact with sulfate-bearing soils or groundwater. Similarly exposure of in situ concrete to the air for a period after moist curing but before contact with sulfates should be beneficial. Influences on thaumasite sulfate attack The primary factors that must simultaneously be present for thaumasite sulfate attack are as follows (Hartshorn and Sims 1998, DETR 1999): • • • •

sulfates and/or sulfides in the ground; mobile groundwater; presence of carbonate; low temperature.

The sulfates may pre-exist in the soil but thaumasite may also occur through the oxidation of pyrite. An abundant supply of water is required. Cold conditions are required with temperatures at least below 15°C. An alumina content is required, even if only at a low level. Indeed alumina contents encourage ettringite formation whereas lower amounts may facilitate reaction with the carbonate and calcium silicates. Influences on delayed ettringite formation The potential occurrence of delayed ettringite formation is predicated on the temperature during hydration and the availability of moisture in service. The


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principal influences on delayed ettringite formation are: • • • •

air temperature during hydration; size and geometry of pour; cement content; cement chemistry and fineness.

Delayed ettringite formation will not occur if the concrete can be kept below a temperature of 70°C during hydration. Air temperature will obviously contribute to the temperature value and steam-curing causes particular risks in this regard. Hot-weather concreting also raises the risk of reaching the threshold temperature limit. Large concrete pours are at risk if the geometry of the formed element is such that the heat loss during hydration is less than the rate of rise. The effect of heat of hydration in these conditions is such that the temperature within the element may be considerable and thereby increase the risk. This may conflict with one of the approaches to the control of early thermal shrinkage cracking that involves temporary insulation. This is sometimes referred to as the ‘if you cannot keep it cool, keep it hot’ approach. Concern about crack control must be balanced against the risk of inducing delayed ettringite formation. Cement content influences heat of hydration at a contributory rate of about 13°C for each 100 kilograms of cement per cubic metre of concrete. Thus high cement content mixes in large pours and/or during hot-weather concreting operations may push the absolute concrete temperature above the critical threshold for delayed ettringite formation. The cement chemistry and fineness is influential, both in terms of heat rise during hydration and its effect on the microstructure of the paste. Blastfurnace slag and pfa have obvious benefits over Portland cements in this regard. Sulfate resisting cement also appears to have a lower risk of delayed ettringite formation in comparison with Portland cement, due to the lower tricalcium aluminate level. The effects of individual parameters and fineness were studied by Lawrence (1993) and Kelham (1996). Hobbs (1999b) reports that the following equation gave a reasonable fit: Ultimate expansion (per cent) k1( f.MgO) k2SO3 k3Na2O k4 where SO3 total sulfate ( per cent by mass of cement) MgO magnesium oxide ( per cent by mass of cement) Na2O sodium oxide ( per cent by mass of cement) f cement fineness (m2/kg) k1, k2, k3, k4 coefficients (0.00085, 0.30, 0.56, 1.4, respectively) However not all cements low in the parameters identified are immune to delayed ettringite formation, nor are cements high in two of the compositional parameters necessarily susceptible.

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Approaches in specification and design practice The approaches to specification and design practice for sulfate-resistant concrete elements involves consideration of the severity of the environment, the permeability of the concrete, and the choice of cement type. Severity of the environment Specification requirements and design practice are based on an assessment of the risk to durability as reflected in the severity of the sulfate environment. The severity of the environment may be determined by reference to the concentration of sulfate (SO42) or magnesium (Mg2) in water or soil. Analysis reports may record the SO3 concentration rather than SO4 but values can be converted based on the SO4 /SO3 ratio of 1.2. Groundwater values expressed in milligrammes per litre are the most convenient for classification to EN 206–1 limits, which range from about 200 to 6000 mg/l. Values in grammes per litre are traditionally quoted in some countries while values in parts per million may be found in American publications, for example 1.5 g/l and 1500 ppm, respectively. Concentrations in soils may be expressed in milligrammes per kilogramme, which allows direct comparison with the class limits in EN 206–1, which range from about 2000 to 24000 mg/kg. Values in milligrammes per litre or in percentage terms may also be encountered with the latter in a range typically from 0.1 to 0.2 per cent. Sulfate concentration in soils may be determined by hydrochloric acid extraction. A 2:1 water:soil extract test, British Standard BS 1377 (BSI 1990a) is in common use in the United Kingdom which yields values requiring different interpretation limits. It is understood (Mure 2000) that the 2:1 test was specifically designed for soils frequently encountered in United Kingdom clays where calcium sulfate is the main constituent. Concrete permeability An essential aspect of producing durable concrete in sulfate-rich environments is the achievement of low permeability concrete to hinder the ingress of sulfates and thereby limit the attack to surface layers. Concretes with water/ cement ratios below 0.4 are thought to be particularly beneficial. The role of cement in minimising the water/cement ratio is more significant in this context than the absolute value of the cement content. This is because the constituents of the hydrated cement are an integral part of the fuel for the reaction and increasing cement content eventually leads to diminishing returns. Cement type Considerable attention is paid to the use of a type of cement which yields optimum performance in the aggressive conditions created by sulfates. These


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include sulfate-resisting Portland cement and blended cement mixes containing blastfurnace slag or pozzalana, such as pulverised fuel ash. Sulfate-resisting Portland cements are characterised by limits on the oxide and compound composition in respect of the SO3 content and C3A content. The values vary across Europe but typically the SO3 limits range from 2.3 to 4.0 per cent and the C3A limits range from 3 to 5 per cent. In the United States of America cements with tricalcium aluminate contents of less than 5 per cent and 8 per cent are considered to be highly resistant and moderately resistant respectively. The difference with European practice is apparent rather than real – the methods of calculating the tricalcium aluminate content being dissimilar. Sulfate-resisting cements eliminate the risk of damage from formation of ettringite. The concrete may however still be vulnerable to attack on the calcium hydroxide content and potentially on the calcium silicate hydrates. The use of blastfurnace slag cements has been found to be beneficial at high slag contents. Blastfurnace slag content concretes are not in themselves chemically resistant to sulfates but may be sufficiently impermeable to provide an effective defence. A minimum slag content of about 70 per cent, by mass of cement, is required. Low slag levels decrease the sulfate resistance. The tricalcium aluminate content of the clinker and the alumina content (Al2O3) of the slag are relevant. Slags of high alumina content are less resistant, particularly if combined with high C3A content Portland cements. Therefore in the case of Portland cement and slag combinations it is necessary to limit the tricalcium aluminate content of the Portland cement if the alumina content of the slag exceeds a certain value. For example in United Kingdom practice British Standard BS 5328 imposed a limit of 10 per cent where the slag exceeded 14 per cent. Hobbs and Matthews (1998b) drew attention to the possible difference in the sulfate resistance of factory-produced slag cements from blends produced at the mixer. The former may produce more resistant concrete. This is due to the enhanced SO3 level typical of factory-produced cement compared to a blend. Research indicates that increased SO3 levels enhance the sulfate resistance of slag cements. Another approach is to limit the potential production of calcium hydroxide through appropriate pozzolanic mixes. This can be achieved, for example, through the use of a Portland pulverised-fuel ash cement (25–40 per cent pfa by mass of the cement), silica fume, metakaolin, or suitable combinations with Portland cement. However the concrete needs to be mature before exposure to the aggressive medium. Ryle (1999) quotes results of a laboratory study that showed good sulfate resistance in mortar prisms containing 15 per cent metakaolin. Hobbs and Matthews (1998b) cite Sellevold and Nilsen in reporting that similar levels of silica fume have demonstrated good performance in Norway. Specific approach in respect of thaumasite attack The DETR (1999) Expert Group report on thaumasite sulfate attack provides recommendations in respect of practice to minimise the risk. Interim

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conservative recommendations cover modification of the existing advice in the case of classic sulfate conditions in excess of 0.24 per cent total sulfur. This involves implementation of additional measures including designed drainage, surface protection, and use of aggregates of limited carbonate content. Guidance on the practical application of the Group’s recommendations has also been published by the Quarry Products Association (Harrison 1999). In relation to Portland limestone cements, Bensted (2000) urged continuation of the precautionary principle that such cements would be restricted to non-sulfate conditions, pending further research. Specific approach in respect of delayed ettringite formation Consideration of the factors influencing delayed ettringite formation indicate that the risk of its occurrence can be minimised by keeping the concrete temperature below 70°C during the hydration period. This may be achieved through a number of measures including consideration of the combined influences of ambient temperature at time of curing, effect of geometry of pour on heat loss during hydration, cement content, and cement type. Thus pre-cast concrete curing temperatures should be limited at all times below 70°C. The cement content of large pours should be selected with a view to limiting the absolute concrete temperature to a value below 70°C during hydration and cure. In this regard Hobbs (1999b) recommends limits on maximum cement content related to fresh concrete temperature as presented in Table 9.1. Further information and guidance is available in a Building Research Establishment (2001b) information paper.

Sulfate attack: exposure categories in EN 206–1 Sulfate attack is considered under the general environmental action of ‘Chemical attack’ in EN 206–1. Three exposure classifications are differentiated: XA1: Slightly aggressive chemical environment; XA2: Moderately aggressive chemical environment; XA3: Highly aggressive chemical environment. The classification for a particular situation is based on the chemical characteristics of the groundwater, soil, or seawater. Limiting values for each class are presented based on an assumption of water/soil temperature between 5°C and 25°C and a water velocity slow enough to approximate to static conditions. The analysis of groundwater may include testing for sulfate ion (SO42), free ammonium ion (to determine NH4 ), magnesium ion (Mg2), and aggressive carbon dioxide concentration (mg/l), in addition to pH value. Analysis of soil


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Table 9.1 Recommended maximum cement content to minimise the risk of delayed ettringite formation (after Hobbs 1999b) Maximum fresh concrete temperature (°C)

Maximum cement content (kg/m3)

10 15 20 25 30

500 470 440 400 360

may require determination of sulfate ion (SO42) concentration (mg/kg) and soil acidity (ml/kg). The reference test method for sulfate concentration is the acid extraction method in EN 196–2 (CEN 1995c) but a water extraction method is permitted if experience is available in the place of use. This is particularly significant for United Kingdom practice. For example, the Concrete Society (1996) note that the acid extraction test would classify London clay above the upper limit of class XA3. This overestimates the aggressiveness of the environment because the sulfates present are mainly calcium sulfate. The limiting values for exposure classes in EN 206–1, Table 2, in respect of sulfates extend to 6000 mg/l for groundwater and 24 000 mg/kg for soil. The standard indicates that a special study may be needed to establish the relevant exposure condition where the limits are outside those presented in Table 2 of the standard. The threshold limit between class XA1 and XA2 in respect of sulfate ion (SO42) in soil (3000 mg/kg) reduces to 2000 mg/kg where there is a risk of sulfate accumulation in concrete due to cyclical wetting and drying or capillary suction. The class is generally determined by the most onerous value for any single chemical characteristic. However where two or more characteristics lead to the same class, for example sulfate ion concentration and pH, the environment is classified into the next higher class. In the case of a class determined on the basis of the sulfate ion content of the soil, the use of a lower class is permitted if it is a clay soil with permeability below 105 m/s. Indicative values for limiting values for concrete composition (maximum water/cement ratio, minimum cement content) and properties (minimum strength class) are presented in the informative annex to EN 206–1. These show that average European practice points to a requirement for moderate or high sulfate-resisting cement in the case of exposure classes XA2 (600 to 3000 mg/l SO42 in groundwater or 3000 to 12 000 mg/kg SO42 in soil) and that high sulfate-resisting cement is required in the case of exposure classes XA3 (3000 to 6000 mg/l in groundwater or 12 000 to 24 000 mg/kg in soil). These trends are informative but of course the specifier must consult the national complementary standard or national annex in the place of use of the concrete for definitive advice.

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The national complementary standard or national annex in the place of use of the concrete may not necessarily present a straightforward set of limiting values within the EN 206–1 framework of exposure classes XA1, XA2, XA3. For example in the United Kingdom complementary standard BS 8500 does not favour the classification system used in Table 2 of BS EN 206–1 in this regard. The requirements are based on a different set of sulfate exposure classes, which reflect the extensive national research effort invested in the topic. For example the maximum solubility of calcium sulfate, 1400 mg/l, has widespread practical significance in the United Kingdom. However this is not reflected in the division of exposure classes in EN 206–1. Additionally, no minimum strength class is given in BS 8500 in order that the greater significance of water/cement ratio be emphasised. Detailed guidance on the appropriate approach to classification of chemical classes and concrete composition requirements for use in UK practice is presented in a Building Research Establishment (2001a) guide which takes account of recent research. In the severest conditions encountered in practice the concrete is likely to exceed the maximum limits of EN 206–1 class XA3 and thereby necessitate specialist advice. This may lead to the specification of physical protection but consideration of other solutions is permitted.

Developments in testing and specification by performance Testing of the sulfate resistance of concretes has traditionally been based on monitoring the performance of the cement paste fraction through mortar prisms stored in sulfate solutions. The possible approaches available to assess resistance include expansion, strength loss, mass loss, change in dynamic modulus, and change in appearance. Expansion measurement is thought to offer the most reliable basis of assessment. Testing the change in appearance can be rather subjective and the other tests are either destructive and necessitate a high number of samples, or can become problematic as the specimens deteriorate. Standard tests for expansion include the American Standard ASTM C1012, a German flat prism test, and a French standard test (NF-P18-837). The French expansion test (Association Française de Normalisation 1993) is being developed by CEN Technical Committee TC51: WG12 as a possible European sulfate resistance performance test for cements. The objective is to develop a test that could discriminate between sulfate resisting and non-sulfate resisting cements within 91 days. The test involves immersion of mortar prisms in sulfate solution. The prisms, of dimensions 20 20 160 mm, are prepared as for strength testing to European standard EN 196–1 (CEN 1995b). The prisms are cured for 28 days before being placed in sodium sulfate solution (1.6 per cent SO3). Expansion is monitored. Moir (1999) reported that a series of four test programmes had been carried out to examine repeatability, reproducibility, and discrimination between cements with different tricalcium aluminate contents. Further research is required


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before recommendation of the methodology as a performance test is possible.

Summary Sulfates may lead to concrete durability failure through attack from external sources (classical sulfate attack or thaumasite sulfate attack) or from internal sources (delayed ettringite formation). A prescriptive approach to the problem of external sulfate attack is the norm with particular attention to limiting maximum water/cement ratio so that impermeable concrete will result. Cement type is another significant consideration in severe sulfate environments. Delayed ettringite formation is a rare phenomenon that can be avoided by keeping the concrete temperature below 70°C during hydration and cure. Sulfate attack is considered in EN 206–1 as a subset of the general topic of chemical attack. Three classes of chemical attack are defined in EN 206–1 but specifiers need to be aware that national guidance documents may take a different approach to the threshold limits between classes and consequently even the number of exposure classes may exceed three.

10 Chemical attack: acid and seawater attack

Introduction Well-cured and compacted concretes of low to moderate water/cement ratio are generally capable of resisting chemical attack caused by acidic solutions because the rate of attack is usually very slow. However it is a reality that all of the main compounds produced following hydration may be dissolved by acids. Durability failure through contact with acids occurs by dissolution of the cement paste and certain aggregate types. The cement paste compounds affected include the calcium silicate hydrates, calcium hydroxide, tricalcium aluminate hydrate and ettringite. Concrete in contact with chemicals in a dry state are not usually affected but acid attack occurs if aggressive chemicals are in solution and contain acidic ions above a critical concentration level. The effect of dissolution may be manifested through surface erosion or complete disintegration of the concrete. Concrete in contact with flowing acids is more vulnerable than that in contact with static fluids. A specific form of chemical attack is that caused by seawater. It is distinguished from other sources of attack in that it involves processes other than dissolution. These include chemical reaction with magnesium and sulfate ions, physical erosion and freeze/thaw attack. Examples abound of chemical environments that may promote attack. Certain industrial situations obviously give rise to aggressive environments through a requirement to use specific chemicals in the production process. A range of potentially deleterious chemical environments can also be found on farms (Braam and Frénay 1997). Agricultural examples include acidic effluent produced during silage making; naturally occurring waste from housed animals with a chemical composition deleterious to concrete; potatoes in storage sprayed with fungicide containing acids. The implications of chemical attack must particularly be considered in the redevelopment of contaminated sites such as old gas works. Sewers may contain hydrogen sulfide which oxides to sulfuric acid. Sulfuric acid may also occur in soils and groundwater through the oxidation of iron sulfide minerals in the form of pyrites or marcasite (Harrison 1987). Carbonated water may also cause attack. Water streams can become acidic through dissolved

Chemical attack: acid and seawater attack

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Table 10.1 Effect of selected chemicals on concrete Category

Acids

Salts and Alkalis

Effect on good quality concrete Disintegration

Low level of attack

Hydrochloric; Hydrofluoric; Muriatic; Nitric; Sulfuric; Sulfurous Ammonium nitrate

Acetic; Carbonic; Carbolic; Humic; Lactic; Phosphoric; Tannic Chlorides of ammonia, copper, iron, magnesium, mercury, zinc

carbon dioxide and such waters are described as ‘soft’. The carbon dioxide is available from the atmosphere and may dissolve to form carbonic acid with a pH of about 5.6. Animal and vegetable oils contain free acids that can increase with exposure to air, leading to problems in the event of frequent spillage on concrete floors. Robery (1988) also drew attention to bacterially active soils containing acid-forming bacteria such as thiobacillus concretivorous. It is not possible to produce a definitive list of the potential chemical effects of every potentially harmful compound but useful summaries have been published. Table 10.1 gives a brief overview of the primary harmful agents. More comprehensive listings may be found in references such as Lea (1998) and the American Concrete Institute (1985) manual. The European Standard EN 206–1 classifies three exposure classes relevant to chemical attack: slightly, moderately and highly aggressive. The potential susceptibility of a concrete member to chemical attack may be determined by comparison of selected environmental chemical characteristics with the limits in Table 2 of EN 206–1. These classifications relate solely to natural ground conditions and seawater. In general terms, specification of concrete that is to be in contact with acids of pH 6.5 and lower requires consideration. Very harsh environments may involve chemical conditions beyond the scope of the recommendations for limiting concrete composition. Special measures may be required, for example: surface protection by external tanking or acid-resistant coating; improvement of site drainage; or use of a neutralising calcium carbonate backfill (limestone/chalk) to be preferentially attacked. In broad terms the specifier needs to give consideration to potential chemical attack in the following cases: • • • • •

concrete in acidic soils and groundwaters; concrete in industrial projects specifically involving chemical processes; concrete for use on farms; concrete sewers; structures in seawater.

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The specific case of chemical attack from sulfates in soils and groundwater is considered separately in Chapter 9.

Physico-chemical aspects Acidic attack By definition acids are substances that yield hydrogen ions in solution by dissociation. The rate of attack on concrete depends on the amount of hydrogen ions formed. This in turn depends on the nature of the acid (strong or weak) and the concentration of the acid in solution (concentrated or dilute). Strong acids, for example hydrochloric (HCl) and sulfuric acid (H2SO4), have a high degree of dissociation whereas weak acids, for example acetic acid (CH3COOH), do not. The externally introduced acids react with the bases in concrete to form a calcium salt and water. The salt may be soluble and solid product is therefore lost. The reaction of calcium hydroxide, for example, with hydrochloric and sulfuric acid respectively is as follows: Ca(OH)2 2HCl → CaCl2 2H2O Ca(OH)2 H2SO4 → CaSO4.2H2O In the second example sulfuric acid reacts to form gypsum, which can be easily removed from the member surface exposing lower layers which are progressively attacked. Ammonium cations exchange with calcium ions, initially from calcium hydroxide but later from the calcium silicate hydrate, for example: 2NH4NO3 Ca(OH)2 → Ca(NO3)2 2NH3 2H2O xNH4 xCaO.ySiO2.nH2O → xNH3 xCa2 xOH ySi(OH)4 (n 2y)H2O Similarly for magnesium: MgCl2 Ca(OH)2 → Mg(OH)2 CaCl2 xMg2 xCaO.ySiO2.nH2O → xMg(OH)2 xCa2 ySi(OH)4 (n x 2y)H2O Attack by nitrates can form nitroaluminates that disrupt the concrete. Concentrated chloride solutions can form chloroaluminates that may also cause disruption. Concrete in use in agricultural applications may be attacked by silage effluent consisting mainly of lactic acid and acetic acid. These form calcium


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lactate and calcium acetate which are largely soluble. For example with acetic acid: 2CH3COOH Ca(OH)2 → Ca(CH3COO)2 2H2O 2CH3COOH C-S-H → SiO2 Ca(CH3COO)2 H2O Soft water containing free carbon dioxide may strip off the surface layers of concrete exposing the aggregate. The reaction then tends to slow down. The carbon dioxide dissolves to form carbonic acid: CO2 H2O H2CO3 H2CO3 H2O H3O HCO 3 The amount of dissolved carbon dioxide which forms carbonic acid is small but when the HCO 3 is removed by reaction more carbonic acid is formed and so on. The attack initially involves production of insoluble calcium carbonate (CaCO3) from hydration products but this further reacts to form soluble calcium hydrogen carbonate: CaCO3 H2CO3 → Ca(HCO3)2 The attack associated with oxidation of iron sulfide minerals in soils has been described by Hobbs (2000). Two possible mechanisms are presented: 2FeS2 2H2O 7O2 → 2FeSO4 2H2SO4 or 4FeS2 15O2 8H2O → 2Fe2O3 8H2SO4 The mechanism of attack in sewers has been described by van Mechelen and Polder (1990). There is anaerobic production of hydrogen sulfide (H2S) from soluble sulfides. This can become dissolved in moisture above the sewage level in pipes and manholes. It is oxidised by aerobic bacteria to produce sulfuric acid. Harrisson (1995) reports that this type of attack can lead to complete loss of concrete strength. Seawater attack Seawater is a mildly aggressive solution, principally due to high levels of sulfate from highly soluble magnesium sulfate. Chemical reactions in seawater are different to those occurring in sulfate-bearing groundwater due to the presence in seawater of chlorides. Physico-chemical effects caused by contact with seawater include magnesium ion and carbonic acid attack on the

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Figure 10.1 Example of structure damaged by seawater attack

products of hydration, leaching, and pressure from salt crystallisation within the concrete. An example of sea water attack is illustrated in Figure 10.1. Magnesium sulfate reacts with the cement matrix to form calcium sulfate and magnesium hydroxide: MgSO4 Ca(OH)2 → CaSO4 Mg(OH)2 The insoluble magnesium hydroxide can provide protection below low water level but higher up it is lost through wave action. Below low water level ettringite may also be formed by reaction of sulfate with tricalcium aluminate and calcium silicate hydrate. It does not develop to the point of damaging expansion because the ettringite is soluble in the chloride-rich environment of seawater. Concrete just above the high water mark can absorb salts by capillary attraction. The effect of evaporation is to concentrate the salts with


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eventual crystallisation on rewetting. This leads to disruptive pressures in the concrete. The phenomenon is known as salt weathering.

Factors influencing attack Factors that influence the rate of chemical attack include: • • • • • • •

concrete permeability; pH; solubility of reaction products; flow rate of acid; characteristics of cementitious components; aggregate type; temperature.

Concrete permeability The rate of attack depends on the rate of penetration of the acid. Therefore a prerequisite for durable structures in acidic environments is that the concrete be dense and impermeable. Chemical attack in concrete involves acids dissolved in solution. Impermeable concrete resists the ingress of acidic solutions and restricts damage to the surface layers. Obviously low water/ cement ratios are required to achieve low permeability concrete. In the context of acid attack the use of superplasticisers represents a better approach than increasing the cement content since extra cement merely adds fuel to the reaction. The rate of attack by some acids has also been found to decay with time due to a beneficial effect of insoluble reaction products. These products may act as pore blockers and reduce permeability. For example: C-S-H HCl → CaCl2 H2SiO3 The silicic acid (H2SiO3) may polymerise to form silica gel (SiO2) and restrict further acid ingress. pH The pH value may be used as an initial screening test of potential acid attack. If the pH of the acid exceeds 6.5 the possibility of attack is remote. Although each unit decrease in pH represents a ten-fold increase in acidity the degree of dissociation is more significant. For example concrete exposed to contact with oxalic acid of pH 3 reacts to form calcium salts that are insoluble and durability is not impaired. The concentration of acid and the quantity in contact with concrete also influences the degree of attack.

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Solubility of reaction products The effect of the attack on concrete durability depends on the solubility of the reaction products. Calcium sulfate, for example, is less soluble than calcium chloride. Reaction of concrete with humic acid – a complex of organic acids generated by the decay of organic matter in soil – can lead to the deposition of insoluble calcium humate with beneficial effects in the case of acid attack in static conditions. Flow rate of acid Flowing acids are more harmful than static fluids. Where concrete is in contact with flowing acid the option of neutralising the acid by the cementitious components is not available and both soluble and insoluble reaction products are continuously carried away. The loss of insoluble reaction products prevents the formation of a protective skin. Aggressive water in impermeable clay soils may not attack concrete in contact with it because the static acidic solution is neutralised at the concrete boundary. Characteristics of cementitious components In general the cement type does not seem to influence the resistance of the concrete but good results have been reported in the resistance of concretes made with silica fume and with metakaolin. An early study by Halstead (1954) showed some slight benefit in the use of high alumina cement in long term tests of concrete in soft water of pH 4.4 – lower than average. Inclusion of pozzolanic materials is generally beneficial in two ways. First they can be used to reduce the permeability of the concrete; second they lead to a reduction in the content of calcium hydroxide Ca(OH)2 thus reducing the quantity of vulnerable material. However it appears that in the case of slag or pfa concretes the extent of any improvement over Portland cement concretes of comparable water/binder ratio is not significant (Hobbs and Matthews 1998b). Silica fume has been found to be advantageous (Berke 1989; Durning and Hicks 1991). In addition to reducing permeability and the calcium hydroxide content, it produces a highly polymerized calcium silicate hydrate paste that is more stable and resistant to acid attack. The extent of the advantages gained through use of silica fume depends on the microsilica content and the type of acid. Improvement has been demonstrated with contents from 7.5 to 30 per cent. Resistance to attack by acetic acid and formic acid has been more pronounced than against phosphoric acid and sulfuric acid. Mehta (1985) reported good results in tests with lactic acid. Knutsen and Obuchowicz (1997) reported optimum results at a level of 10 per cent by weight of cement, in agricultural applications. Metakaolin, the purpose-made pozzalana made from kaolinitic clay, reacts


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with calcium hydroxide during hydration to form stronger and less soluble hydrates with enhanced acid resistance. Martin (1997) studied the influence of metakaolin in tests with silage effluent. The effluent mainly consisted of lactic acid with acetic acid also present. He reported that inclusion of 15 per cent metakaolin resulted in one-third less damage compared to Portland cement concrete. Further improvement resulted from the use of ternary blends of Portland cement, slag, and metakaolin. The inclusion of metakaolin and latex co-polymer was also very effective. Silverlock (1999) reported that the use of metakaolin was specified in an industrial water diversion scheme handling large volumes of acidic effluent. Aggregate type Aggregate type influences the rate of attack and the effect on serviceability. Limestone aggregate is particularly vulnerable to acid attack but this has potential advantages. First it can contribute to the neutralisation reservoir in the case of static acidic waters in contact with concrete. Second, if used in pipes it decays in tandem with the cementitious component thus maintaining a hydraulically efficient smooth surface. The use of other types of aggregate in pipes not only leads to rough surfaces but the aggregate becomes detached over time, leaving debris in the pipe that could impede serviceability. Protruding acid-resistant aggregate in the floor slabs of horizontal silage silos is undesirable from the viewpoint of animals using the facility in a self-feed manner. Limestone aggregate is preferable to maintain a serviceable, if deteriorated, surface. Temperature High temperature may compound problems of acidic attack because of the well-established relationship between rate of reaction and temperature. For example acid attack in sewers is more pronounced in countries with hot climates.

Mathematical modelling of acid attack The rate of acid attack often decays with time but in other cases the deterioration–time relationship is linear. The reduction of attack rate may often be related to the square root of time. The trends are illustrated in Figure 10.2. The reduction in decay of attack rate with time is due to the formation of a protective layer of silica gel in place of the dissolved concrete. This impedes access by the acid to the soluble hydrated components. This protective layer may remain intact in the case of static acidic waters but otherwise may be lost. A mathematical model proposed by Grube and Rechenberg (1989) tracks the development of the affected layer with time through Fick’s first law.

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Figure 10.2 Typical patterns of acid attack rate

Concrete in contact with a solution containing dissolved carbon dioxide forms, in the first instance, a thin layer of calcium carbonate, CaCO3, but further reaction produces soluble calcium hydrogen carbonate, Ca(HCO3)2. This soluble material diffuses out into the water. In the mathematical model the thickness of the affected layer is estimated as a function of the following: time; the concentration difference of calcium hydrogen carbonate in the concrete and surrounding solution (which is an indicator of the aggressiveness of the environment); the composition of the concrete (mass of soluble matter per unit volume and relative proportion of insoluble matter); and the diffusivity of the protective layer with respect to the calcium hydrogen carbonate. These parameters are represented in the model as follows: x thickness of deteriorated or removed surface layer (cm); D diffusion coefficient, for Ca(HCO3)2 or Ca2 ions, of the gel layer (cm2/s) as /at ratio of soluble matter area to total area ms mass of soluble matter (g) of CaO per cm3 of concrete c*s Ca(HCO3)2 concentration, or highest concentration of Ca2 ions, in the solution at the face of intact concrete (g/cm3) cm Ca(HCO3)2 concentration, or highest concentration of Ca2 ions, in the solution unaffected by the concrete (g/cm3) t time (s) The quantity of material diffusing may be stated as:

Chemical attack: acid and seawater attack dl D as

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c*s cm dt x

From a chemical reaction viewpoint the quantity may be stated as: dl ms at dx Hence ms at dx D as

c*s cm dt x

and x dx

D as * (c cm )dt ms at s

By integration the thickness of the deteriorated or removed layer may be determined at time ‘t’: x

√

– 2 D as * (c cm ) √t ms at s

Grube and Rechenberg estimate the value of the diffusion coefficient (D) at 1.8 106 cm2/s for SiO2 gel of hardened cement paste. The ratio as /at can be determined from the ratio of the volume of soluble constituents to the total volume. If insoluble aggregates are used the volume of soluble constituents is equal to the volume of hardened cement paste. The relationship may be restated as: – x a √t where ‘a’ is a constant for a particular case, thus leading to a parabolic curve. However if the protective layer is removed by mechanical action, for example by turbulent flowing acid, the removal rate becomes linear: xt The value of can be determined in short term tests.

Approaches in specification and design practice The approaches to specification and design practice for chemical attack by acids recognises that no concrete is completely immune to attack. Therefore

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the objective is to reduce the rate of attack to an acceptable level. In most cases this involves the specification of concrete of adequate composition but in the severest environments other protective measures may be required. The approach in EN 206–1 is through definition of exposure classes with upper and lower bounds in respect of selected chemical characteristics. There are underlying assumptions in respect of temperature range and water velocity within which the classes apply. Qualifying exposure conditions generally have associated guidance in national documents on limiting values of concrete composition. In harsher conditions the concrete may need to be protected by other means. Examples from practice to date include surface treatment or coating; provision of a sacrificial layer through the use of section sizes which are above the minimum required for structural resistance; and the use of limestone aggregate or backfill to neutralise the acid through reaction. In the case of seawater attack a 1960s report by American Concrete Institute Committee 201 (1962) recommended use of a Portland cement with not more than 8 per cent tricalcium aluminate. More recently Osborne (1994) once again drew attention to the somewhat inferior performance of high tricalcium aluminate concretes. Hobbs (2001) reviewed the limited published data available to date. He noted that freeze/thaw resistant concretes should give adequate resistance to seawater attack if the water/binder ratio was 0.45 or lower and the tricalcium aluminate content was below 12 per cent. He extended the range of resistant concretes to include the following combinations with Portland cement: 20 to 40 per cent fly ash, 70 to 80 per cent slag, and less than 10 per cent silica fume.

Acid and seawater attack: exposure categories in EN 206–1 Acid and seawater attack are considered under the general environmental action of ‘Chemical attack’ in EN 206–1. Three exposure classifications are differentiated by quantifying the degree of aggressivity: XA1: Slightly aggressive chemical environment; XA2: Moderately aggressive chemical environment; XA3: Highly aggressive chemical environment. The classification for a particular situation is based on the chemical characteristics of the groundwater, soil, or seawater. Limiting values for each class are presented based on an assumption of water/soil temperature between 5°C and 25°C and a water velocity slow enough to approximate to static conditions. Five chemical characteristics are listed in respect of groundwater and two in the case of soils. The classifications approximate to the quantified characteristics of weak, moderate and strong attack proposed by de Sitter in a CEB (1984) bulletin. The class is determined by the most onerous value for any


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single chemical characteristic. The characteristics used in the case of groundwater are pH, and the concentration (mg/l) of sulfate (SO42), ammonium (NH 42), magnesium (Mg2) and aggressive CO2. Soil is classified according to its sulfate content (SO42 mg/kg) and degree of acidity (ml/kg) according to Baumann Gully. National annexes and national complementary standards provide definitive guidance for their jurisdictions in respect of limiting concrete compositions to ensure durability in conditions of potential chemical attack. It should be noted however that not all national documents use exposure classes defined by the same limits as those in Table 2 of EN 206–1. For example United Kingdom complementary standard BS 8500 does not include the EN 206–1 classification system in respect of chemical attack. Detailed guidance on the appropriate ‘design chemical classes’ and concrete composition requirements for use in UK practice are included in the Building Research Establishment (2001a) Special Digest 1. The British complementary standard includes specific advice in respect of concrete in contact with seawater. pH The three classes may be differentiated on the basis of maximum pH level. Ten fold increases in acidity differentiate classes XA1, XA2, and XA3 with maximum values of 6.5, 5.5, and 4.5 respectively. A lower bound of pH 4 is included in Class XA3 and so a special study would be required in the specification of concrete for lower pH conditions. However it is worth emphasising that pH value alone is not an adequate measure of aggressivity although Harrison (1987) suggests that concrete may not be a suitable material in conditions where the pH value is below 2.5. Aggressive carbon dioxide, ammonium, magnesium The XA classes delineated by the aggressive CO2 level have a lower threshold value of 15 mg/l. The lower limit for classes XA2 and XA3 are 40 mg/l and 100 mg/l respectively. Delineation by ammonium ion level (NH 42) commences with a lower threshold value of 15 mg/l and the lower limit for classes XA2 and XA3 are 30 mg/l and 60 mg/l respectively. The upper limit on class XA3 is 100 mg/l. Delineation by magnesium ion level (Mg2) commences with a lower threshold value of 300 mg/l and the lower limit for classes XA2 and XA3 are 1000 mg/l and 3000 mg/l respectively. Soil acidity Soil acidity is used in the potential classification of a condition corresponding to exposure class XA1. The reference test method is DIN 4030–2 (Deutsches Institut für Normung 1991). The threshold is a value in excess of 200 ml/kg Baumann Gully.

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Specification by performance Acid attack Performance-based design requires both a predictive model and a test method. An approach for developing a mathematical model for acid attack, as noted earlier, has been published by Grube and Rechenberg (1989). However no standard test method exists for testing the acid resistance of concrete. This is due in part to the difficulty of accelerating in laboratory tests the deterioration that may take place in the field. Attempted acceleration by the use of unrepresentatively high concentration levels in laboratory tests would fail concretes that may give adequate service under field conditions. Seawater attack Testing the sulfate resistance of concrete was discussed in Chapter 9. The favoured method in developing a European standard test involves monitoring the expansion of mortar prisms stored in sulfate solutions. The potential adoption of this method as a European sulfate resistance performance test for cements is being examined by CEN Technical Committee TC51: WG12. Use of the method for testing resistance to seawater attack could possibly be achieved by substituting the sulfate solution by artificial seawater.

Summary Concrete is not resistant to acid attack. However structures in acidic environments may achieve their required length of service through concrete compositions which have sufficiently slow rates of deterioration. In the harshest environments concrete may need additional surface protection. Exposure to seawater also leads to chemical attack but the process is complicated by the physical effects of wave action and abrasion. Tentative mathematical formulations have been advanced for modelling acid attack. Development of performance-based specifications is somewhat hampered by the difficulty of developing standard test methods for acid resistance. Performance tests for durability often require laboratory tests using harsh conditions to accelerate deterioration. This approach may not provide a reliable basis on which to predict field performance based on laboratory findings in the case of acid attack. Tests for resistance to seawater attack may be developed in tandem with the efforts to produce a European standard test for sulfate resistance. Acid attack and seawater attack are considered in EN 206–1 under three exposure classes defined by limits of selected chemical characteristics. Specifiers should be mindful however that not all national complementary


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documents to EN 206–1 use the same threshold limits to define exposure classes and consequently the framework of exposure classes may differ. There is universal agreement that the achievement of dense impermeable concrete is a prerequisite requirement for durability in acidic environments.

11 Cracking in reinforced concrete structures

Introduction Cracking in reinforced concrete structures is an unpredictable occurrence. Nevertheless, the unintentional appearance of cracks may be regarded as a serviceability failure and perhaps, if it reduces the acceptable service life, a durability failure. The extent of this failure varies from case to case but cracking is generally a cause for initial concern – cracks are guilty until proven innocent. The appearance of cracks in a building or structure, particularly after it has entered service, can cause disquiet for all those involved – its designers, owners, occupants and members of the public who use the structure or travel past it. On occasion, whether justified or not, cracking may lead to a loss of confidence in the building. Cracks, whether structurally significant or not, that deeply disturb the occupants of a building may shorten its economic life. The consequences of cracking range from trivial to catastrophic. Physically, the consequences may be durability and aesthetic problems. Psychologically, cracking affects the attitude to a building or structure. To a client, widespread cracking may undermine his confidence in the professional skills of the designers or builders. The owner of a building would be anxious about the possible loss of structural integrity, unease amongst staff working in the building, and the potential loss of investment. The designer of durable reinforced concrete structures can learn much from the study of case histories of cracking. This leads to a better understanding of how buildings behave and the suitability or otherwise of specific details. In essence an intuitive understanding can be built up through a store of knowledge of the following aspects of the topic: how cracking occurs, the form it takes, the significance of timing, and the underlying causes. The topic is complex in that individual aspects may aggravate or offset the effect of each other. The various aspects mentioned are dealt with in greater detail in this review but a number of essential points are noteworthy. Cracking is caused by tension. Cracks will appear in reinforced concrete when the imposed tensile force exceeds the tensile strength of the concrete. Cracking appears as a

Cracking in reinforced concrete structures

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solitary or a pattern phenomenon. Each crack can be characterised by features that help to identify the cause. These include width, depth, length, taper, direction and size. In the case of pattern cracking, the spacing of the cracks is also significant. In the context of this review, cracking is considered to originate at the design, construction, or service phase of a building’s life. In broad terms the possible reason, or reasons, for cracking may lie in the design history of the building; the mix constituents and environment; the construction history; plastic behaviour of the fresh concrete; subsoil effects; elastic behaviour under load; and the restraint of movement.

Mechanism of cracking From the time of man’s earliest use of concrete it has been recognised that the material had excellent load carrying capacity in compression but very little in tension. Cracking is caused by tension. The tensile strength of concrete is usually ignored in design for flexural resistance and this assumption therefore automatically introduces the risk of cracking in the cover concrete. A significant exception is in the design of crack-free structures such as liquid retaining structures where the tensile resistance of concrete in bending is taken into account. Reinforced concrete structures are subjected to a combination of stresses during the design life. Some of these stresses can arise within days of the concrete being placed and before it can achieve even a modest tensile strength. The concrete is initially in a quasi-liquid phase and is capable of large deformations but the strain capacity drops as the cement paste stiffens before rising slightly as hydration develops. The tensile strain capacity is considered to be the maximum strain that the concrete can withstand without the formation of a continuous crack. A laboratory study of the stress–strain characteristics of reinforced concrete in tension was carried out by Morita and Kaku (1975) using centrally reinforced square prisms loaded in tension. The results showed the extent to which concrete contributed to reducing the average steel strain and that cracking occurred at about 0.01 per cent strain. Harrison (1992) found that the tensile strain capacity of concrete was of the order 130 106 to 400 106. Crushed rock aggregate concrete had a greater strain capacity than rounded aggregate concrete. The initiation of cracking in reinforced concrete is preceded by the development of tensile stress in both the concrete and steel due to external loads or restraint of movement. At first the two materials behave in accordance with elastic theory and act compositely. When the stresses exceed the tensile strength of concrete cracking occurs and a transfer of stress takes place from the concrete to the steel. Morita and Kaku (1975) noted that in direct tension tests the cracking always occurred approximately halfway between existing cracks during a chain reaction of crack formation. Illston and Stevens (1972) observed that cracks first appear over certain lengths of bar in the zone

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Figure 11.1 Influence of reinforcement spacing on crack width and orientation

around the steel due to the breakdown of the adhesion bond between steel and concrete. This partial breakdown in bond leads to high stress levels in the concrete and stable transverse cracks then appear within the body of the concrete. In normal strength concrete these cracks occur at the aggregate– matrix interfaces. Partial bond breakdown leads to high stress conditions near the ends of the bond lengths remaining. It is suspected that in direct tension the next stage involves the propagation of a crack from one of these points by the linking of existing cracks. In bending it may propagate from the highly stressed member surface. Thus the first continuous crack appears, often with much branching. As this crack develops there is a rapid transfer of load from concrete to steel and the crack joins up with the cracked zone around the bar. The resultant breakdown in adhesion in the area of this crack leads to some opening of the crack at the bar surface. A stress relief mechanism is thus set up and a chain reaction may ensue with the development of other continuous cracks along the member. Loss of adhesion along ribbed bars causes a transfer of load to the ribs of the bar. Cracking then occurs within the concrete instead of the failure occurring along the bar itself. Tepfers (1979) showed that the bond forces radiate out into the surrounding concrete. Initially the forces are taken on a concrete ring surrounding the bar determined by the minimum cover


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Figure 11.2 Development of diagonal shear cracks from flexural cracks

dimension. Radial cracks first extend a short distance from the bars. This reduces the uncracked area in the ring and the concrete stress increases leading to further crack propagation until eventually the cracks extend to the edge of the member. Some secondary cracking may be noted as cracks propagate from the zone around the steel to join the primary cracks. The formation of cracks in two-way spanning slabs was studied by Nawy (1975). It was found that the main criterion affecting the process is the spacing of the reinforcement. If the spacing is small then local high stress points are formed at the intersection of the bars. Stress paths between the points join along the paths of least resistance – the planes of discontinuity at the interface of concrete and steel. Small narrow cracks then run along the reinforcement grid. However if the spacing is large then the energy absorbed per unit grid is too small to cause cracking along the bars and consequently wide yield line cracks appear running through the grid, as illustrated in Figure 11.1. In the case of reinforced concrete beams without shear reinforcement, Taylor (1959) demonstrated that diagonal shear cracks extend from flexural cracks. The stages of crack development are shown in Figure 11.2. Vertical flexural cracks begin to curve but the rotation is arrested by aggregate interlock. This sets up a force across the crack that ultimately leads to the diagonal crack classically associated with shear failure. Schiessl (1975) noted that in the case of indirect loading the deformation that leads to the appearance of a crack decreases after the crack occurrence. It would have to increase again before another crack would occur. Although it is not possible to conclusively attribute the cause of cracking on the basis of visual observation alone, certain characteristics may be associated with the nature of the underlying stress. Cracks caused by direct tension are of uniform width. Flexural cracks usually originate at the tensile

212


Figure 11.3 Chronological characteristics of crack-inducing phenomena

face of the member and continue into, and often beyond, the reinforcement. These cracks would normally be at right angles to the tensile face of the member. Haldane (1976) noted that the crack density corresponds to the bending moment diagram. A salient feature of shear cracking is that the cracks are invariably inclined to the tensile face at an angle of approximately 45°. Shear cracks will often originate and stop within the boundary of the cracked face. A frequent location of shear cracking in beams is at the inner face of a support. Torsional distress may take the form of irregular jagged cracks. They are induced by deformations in attached members. In the case of torsion in two-way spanning slabs the cracks would typically occur at the corners and take the form of parallel cracks normal to the diagonals.

Chronological aspects of cracking The occurrence of cracking in a material indicates that the material has been stressed beyond its strain capacity. Reinforced concrete is stressed through


213

the support of external loads and also experiences stresses induced by the reaction of the material to the environment and imposed deformations. The response of reinforced concrete to these effects depends on its age. Freshly mixed concrete is in a quasi-liquid state when placed and is capable of withstanding large deformations. Rapidly it begins to harden and the tensile strain capacity decreases before making a slight recovery with time. The strength of the concrete is then generally stable. Some of the main causes of cracking can be related to a time-scale. These causes are considered in detail in subsequent sections but an overview of the time-scale applicable to significant phenomena is illustrated in Figure 11.3. The causes considered are plastic settlement; plastic shrinkage; early thermal behaviour; crazing; restrained drying shrinkage; elastic deformation under load; settlement; restrained thermal movement; and corrosion-related cracking. This diagram shows that the risk of cracking is present at all ages but that the greatest risk occurs in the first few years. This is due to the long-term behaviour of the concrete and the relief of stress through cracking. The effects of drying shrinkage may be offset by creep, which follows immediate elastic deformation under load. Prolonged settlement may cause the opening of existing cracks rather than the formation of new ones. Stresses due to restrained thermal movement may be relieved through cracking, allowing the subsequent thermal movement to be accommodated by the opening and closing of existing cracks. Another way of viewing the origins of cracking is to view their chronological roots. This may be done by viewing three distinct phases in the life of a structure or building: the design phase, construction phase, and the service phase. An overview of the phenomena and significant parameters at each phase is presented in Figure 11.4 and discussed in the next section. It will be recognised that some of the chronological divisions are arbitrary in that there are interactions, for example between design decisions and construction practice, which can lead to cracking in a particular instance.

Cracking and the design phase Sources of distress Design decisions at the general arrangement stage of a project are critical to the avoidance of cracking. The clearest overall view of the structure will be obtained at the earlier stages of design evolution. The project engineer can draw on knowledge, experience and intuition to arrange the assemblage of structural members that can act together and with the building fabric in a manner which does not cause an unintended stress build-up. The adverse effect of restraining movement can be reviewed, formally or informally, by the project engineer when advising the architect on the preliminary scheme. Later, however, the project engineer will probably have to delegate detailed design of individual structural members to others. Occasionally these less experienced

Figure 11.4 Crack-inducing phenomena associated with three phases of a building: design, construction and service phase


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Figure 11.5 Example of a structural element which could be loaded above its typical design level due to its particular location

Figure 11.6 Formation of a diagonal crack in an inappropriately detailed ‘fixed’ connection

engineers may agree to a succession of seemingly minor changes to the scheme, due to new architectural or services engineer’s constraints, that may lead to unforeseen restraint and thereby promote cracking. During the detailed design stage engineers guard against cracking from design errors rooted in arithmetical errors. However less easy to eradicate is cracking rooted in the inaccurate determination of anticipated loading. For example, Feld (1964) demonstrated that snow loading on roofs could be underestimated in conditions such as on an entrance canopy set into an angle of the building (Figure 11.5). Roof falls requiring drainage points at mid-span could lead to cracking from over-stressing if drains get clogged or frozen – ponding leads to deflection that causes more ponding and so on. Reinforcement detailing may occasionally lead to cracking if each member is considered as a separate entity. Ryan (1977) has stressed the need for careful detailing of supports that are assumed to be ‘fixed’ because diagonal cracking is prevalent in such connections, as illustrated in Figure 11.6. Detailing of individual beams and columns often fail to account for the inherent distortion and forces. Standard detailing of column and beam reinforcement would not include reinforcement across the resultant crack. Problems may also occur

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Figure 11.7 Load case that could generate cracking due to unanticipated positive moments on both sides of a connection

in ‘T’ connections if side sway is large. A case that could be overlooked is the need to detail ‘X’ connections for positive movements on both sides if a cantilever span is subjected to wind uplift, as shown in Figure 11.7. In addition to detailing errors, an inexperienced detailer may at times produce reinforcement details that are inherently crack inducing. Examples include congested reinforcement details leading to poorly compacted concrete; inadequate trimming of opes; an absence of staggered laps; and inadequate anchorage, especially in cantilevers. Cracking may also be rooted in architectural details inappropriate to reinforced concrete. Students of architecture are impressed by the supposed limitless fluidity of form achievable in concrete. It would not be unreasonable therefore for an inexperienced architect to design intricate sections that may not be readily achievable in practice. Regrettably some of these details are improperly built with consequent inadequate cover or compaction leading to cracking or spalling. It is also important that the engineer recognises and eliminates architectural features that inadvertently act as stress raisers. For example the specification of tie-bolt hole locations close to kicker level in exposed concrete walls may induce cracking during the initial thermal contraction phase. Design tactics for control of cracking The function of a building or structure will dictate the constraints that govern the degree of complexity required in crack control. Liquid-retaining structures or those used for shielding harmful contaminants, for example, must remain effectively crack-free. ‘Crack-free’ structures are those in which any cracks that do occur are so narrow that they would not interfere with serviceability. Three approaches to crack control may be identified in respect of the designer’s toolkit for meeting the serviceability state of cracking. First,


217

Figure 11.8 Examples of pre-determined crack-inducing planes of weakness in concrete elements

structures that are effectively crack-free can be achieved by conservative design limits on stresses. A second approach utilises concrete’s greater structural potential by allowing crack propagation but limits the crack width by calculation or prescriptive detailing rules. A third approach is to limit the crack locations to acceptable pre-determined positions while accepting the possibility of the cracks being relatively wide. The design of crack-free structures could be carried out in one of two ways. The crack widths could be estimated by calculation to ensure that rigorous serviceability requirements, published in design codes are met. Alternatively conservative limitations could be imposed on permissible concrete and steel stresses and strains. Control of crack width by detailing is based on prescriptive rules derived from decades of experience in the field. These rules influence bar diameter, spacing, minimum cross-sectional area of reinforcement, and cage geometry. The reinforcement required for structural resistance is supplemented where necessary to ensure compliance with best detailing practice. Control of crack location is achieved by detailing joints and reinforcement in a manner that introduces a deliberate plane of weakness. It is based on the observation that cracks preferentially propagate at highly stressed sections, such as at discontinuities, opes or imperfections in structural members. Two examples of this technique, illustrated in Figure 11.8, are the incorporation of chases at regular intervals in long retaining walls and the use of crackinducing joints in pavements and ground floor slabs. These features introduce a deliberate weakness into the member so that one large crack occurs at the preordained position rather than several smaller cracks appearing at random throughout the member. Large cracks in chases in retaining walls are not visible from even moderate distances. Cracks induced in pavements and

218


ground floor slabs can be located at positions where they can be efficiently sealed or covered by permanent fittings. Control of crack width by calculation involves checking the predicted width against design code recommendations. Many formulae have been proposed to model crack width in reinforced members. An underlying principle is that the cumulative total of crack widths in a member is related to the strain and that the width of any one crack depends on the number of cracks sharing the total width. Albandar and Mills (1974), for example, proposed the following relationship in respect of a reinforced concrete beam: mean (1.5acr 0.4 A0.5)m where mean mean crack width acr distance from crack to nearest bar surface A effective area of concrete surrounding bar surface m average concrete strain at the level where cracking is considered The Concrete Society (1982) published a formula for the maximum crack width based on a consideration of the minimum crack spacing. It highlights the relevant parameters and may be stated as follows: max 2Sm and Sm k1 C k2

p

where max maximum crack width Sm minimum crack spacing strain k1 constant C cover k2 constant dependent on bond strength bar diameter p steel ratio of concrete section The formula is based on the fact that the stress at a crack is zero and therefore another crack cannot form until the stress builds up to a level that exceeds the tensile strength of the concrete, a minimum distance Sm away (Figure 11.9). Thus the maximum spacing of cracks is 2Sm. The development of crack width formulae for use in design codes can be exemplified by the formulae in a bridge design standard, BS 5400 (BSI 1990b):

3acr m 1 2(acr cnom)/(h dc)


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Figure 11.9 Linear development of stress with distance from a crack

and m 1

h(a' d ) M 1 10 ] [3.8b A (h d ) ] [( M ) t

s

s

c

c

q

9

g

where design crack width acr distance from crack to nearest bar surface which controls width m calculated strain at the level where cracking is considered, allowing for the stiffening effect of the concrete in the tension zone cnom cover to the outermost reinforcement h overall depth of section dc depth of concrete in compression bt width of section at the level of the centroid of tension reinforcement a' distance from compression face to point at which crack width is being calculated s calculated strain in the tension reinforcement, ignoring the stiffening effect of the concrete in the tension zone As area of tension reinforcement Mq moment at the section under consideration due to live loads Mg moment at the section under consideration due to permanent loads

220


Design code recommendations in respect of crack width have generally fallen into the range of 0.10 mm to 0.25 mm for outdoor exposure. The lower end of the range applies to certain types of structure in particularly hostile environments but a value at the upper limit may even be acceptable, from a leakage viewpoint, in water-retaining structures due to the beneficial effects of autogenous healing. Interpretation of the limits in practice can be a source of dispute. For example American Concrete Institute Committee 224 (1990) produced tolerable crack widths for reinforced concrete structures under various exposure conditions. The Committee noted that a portion of the cracks in a structure would be expected to exceed the published values even when recommendations for limiting width are followed. The durability debate in respect of influence on corrosion activity is discussed in a later section but there is general agreement regarding aesthetic parameters. The minimum crack width visible to the naked eye is 0.13 mm (Neville 1995). A study by Haldane (1976) found that the public perception of acceptable crack width was influenced by the location of the cracks. Cracks in nibs and corbels caused most disquiet because the structural purpose of the elements was self-evident. Those surveyed regarded crack widths in excess of 0.2 mm as unacceptable. It is worth noting however that cracks narrower than 0.2 mm can be highlighted by the ingress of dirt and moisture. Cracks on smooth surfaces disturbed the public more readily than cracks of the same width on rough textured finishes. Engineers viewed subconsciously for collapse and would accept crack widths up to 0.1 mm greater than those acceptable to the public.

Cracking during the construction phase The focus of this book is on design and specification aspects of durability but in reinforced concrete, perhaps more than any other form of construction, it is important to realise the responsibility that rests on the team constructing the works. The science and art of reinforced concrete design and construction technique has progressed significantly in recent decades but despite (or because of) technological developments the responsibility is undiminished. The practical nature and production pressures inherent in construction work provide many opportunities for departures from ideal conditions and sources of cracking may arise. Structural concrete pours are usually accompanied by much activity as the operatives ensure that the concrete is placed, compacted, and finished efficiently. However danger lurks in the understandable air of relaxation after the last load has been placed and the vibrator has been switched off. Concrete requires monitoring in the hours immediately after placing while it is in a plastic state because of the phenomena of plastic settlement and plastic shrinkage cracking. These forms of cracking may be eliminated if prompt remedial action is taken. A further problem in the days immediately after a pour is that of early thermal cracking. The temperature of concrete can be


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Figure 11.10 Examples of plastic settlement cracking

highest during the first few days after casting due to the heat of hydration. The concrete therefore has an expanded volume when it sets. During cooling to ambient temperature the concrete will contract but may be frustrated from doing so by internal or external restraint. A variety of other crack-inducing stresses may also arise during construction, most notably from early removal of shuttering and accidental overloading. Plastic settlement Plastic settlement cracks are caused by the inability of the concrete to settle uniformly. Thus the problem may be encountered where there is a variation in the depth of the member or where there is significant restraint from reinforcement near the surface. Four examples are illustrated in Figure 11.10. In slabs an insufficient amount of cover to large diameter top steel bars can cause the concrete to break its back over the reinforcement. In these cases plastic settlement cracks will occur along the reinforcement and their depth will extend from the slab surface to the top of the reinforcement. The durability threat in bridge decks subject to de-icing salts is self-evident. In

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columns and walls, the horizontal steel may locally restrict the downward flow. This leads to cracking as the concrete inside the cage settles at a faster rate. Turton (1978) noted the examples of cracks appearing at changes of section in columns with flared heads where the concrete in the head is prevented from settling at a compatible rate due to arching action. In trough and waffle floor slabs settlement in the trough differs from the settlement in the topping and this may encourage cracking at the change in section. These cracks will usually extend throughout the depth of the topping. Measures to reduce the risk of plastic settlement cracking are related to reducing the bleeding. It has been found that the use of an air-entrainment admixture can reduce bleeding. Hobbs (2001) noted that the use of fly ash reduces bleeding but that the benefit is offset by a reduction in failure strain capacity. Slag concretes are at greater risk in this context due to the likelihood of increased bleeding. Vigilance on site can eliminate the problem because the cracks can be noticed while the concrete is still in a plastic state and amenable to revibration. Although the final responsibility for eliminating plastic cracking rests with the site staff, the designer and specifier should be mindful of the inter-related influence of cover, bar diameter, and slump. The bigger the cover, the smaller the bar diameter, and the lower the slump, the less risk there is of plastic settlement cracking. Cover is singularly the most important of the three factors. Plastic shrinkage Plastic shrinkage is a phenomenon rooted in the fact that the volume of hydration products is less than the volume of the constituents used to make concrete. This self-desiccation is a significant problem in high strength concrete. Further plastic shrinkage occurs, particularly in slabs, when the environmental conditions are such that the rate of evaporation exceeds the rate of bleeding. Cracking may occur if the shrinkage is restrained. The high rate of evaporation may be due to a high ambient temperature or strong drying winds. Turton (1978) states that the cracking is probably caused by surface tension forces or the effects of flocculation of ultra-fine solid particles in the mix. The direction of plastic shrinkage cracks depends on the layout of restraint and wind direction. The cracks are typically parallel and approximately at 45° to the slab edges. The spacing of cracks can vary but are usually in the range of 0.3 m to 2 m. Sometimes the cracks occur as a random map pattern. Plastic shrinkage cracks of both patterns rarely extend to the slab edges if these are unrestrained but may pass through the full depth of the slab. The cracks are usually wide at the surface but taper rapidly. Proper curing procedures will often prevent plastic shrinkage cracks. The rate of evaporation is central to the problem. It has been found that covering the freshly placed concrete with polythene sheets is particularly effective if done within ten minutes of the completion of the pour. Sometimes it is


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necessary to suspend the polythene over the slab, to avoid damage to the finish, in which case the edges must be tied down to prevent drying winds funnelling through the covered area. The use of sprayed-on curing membranes are useful for long term curing but usually cannot be applied to concrete while in the plastic state and so will not prevent plastic shrinkage cracking. This issue is significant in the case of fly ash concrete due to the extended plastic period and reduction in failure strain capacity. Early thermal contraction The heat of hydration leads to a temperature rise which will ordinarily bring the concrete temperature above ambient. This heat is dissipated at a rate that depends on a number of factors, not least of which is the thickness of the member. The concrete may therefore set at an elevated temperature and volume. Cracking may occur when the concrete attempts to shrink on cooling if it is restrained by internal or external restraint. For example, the exposed surface of a deep pour will contract at a higher rate than the core (internal restraint); a wall when cooling will wish to contract more than its kicker (external restraint). The tensile strain capacity of concrete in respect of cracking is about 200 106 and the coefficient of thermal expansion of concrete ranges from 6 106 to 13 106 per degree Centigrade depending on the aggregate. Thus the differential temperature must be kept below 16°C to 33°C and a figure of 20°C is typical. Fitzgibbon (1976) proposed a rule-of-thumb for the maximum temperature reached by the concrete during hydration as follows: Tmax Tp

(100w ) (12)

where Tmax maximum temp. of concrete during hydration (°C) Tp temperature of concrete at point of placing (°C) w cement content of mix (kg/m3) He proposed that the formula was valid for cement contents of 300 to 600 kg/m3 and gave a worst case result. Hobbs (2001) estimated the temperature rise at 14°C per 100 kg/m3 for a 52.5R Portland cement and 12.5°C per 100 kg/m3 for a 42.5R Portland cement. Various approaches may be used where the temperature differential between the core and the surface must be limited. For example the formwork could be insulated to keep the surface warm or the aggregates could be cooled to reduce the peak temperature. The use of fly ash and slag can reduce temperature rise but the benefit may be offset by the reduction in tensile strain capacity. Cracking due to external restraint may be controlled by limiting the pour size or by the use of crack-inducing joints. Reinforcement in itself may not prevent the occurrence of cracking but investigators have noted that the crack

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Figure 11.11 Influence of base flexibility on early thermal crack pattern

width is inversely proportional to the amount of reinforcement provided. Harrison (1992) noted that the more flexible the base the more concentrated the crack pattern. The effect is illustrated in Figure 11.11. Formwork aspects The early removal of shuttering is a common cause of crack formation and Smolira (1972) attributed up to 60 per cent of cracking to this cause. Thermal shock may occur when concrete surfaces are first exposed to the atmosphere. Striking times should be temperature related as the formwork provides insulation and aids maturity. Maturity monitoring may be employed or minimum striking time in equivalent hours may be specified. For example the striking time for wall shutters might be specified as 24 equivalent hours. The hours during which the air temperature was above 5°C would be counted fully, hours during which the temperature was between zero and 5°C would be counted as equivalent half hours, and any time when the temperature was below freezing level would be discounted. Cracking may also occur if formwork moves before the concrete is mature enough to support its self-weight and any incidental construction loads. Accidental overloading Overstressing can occur due to accidental overloading. This is liable to occur at any time but is most likely during the construction phase. For example suspended slabs in multi-storey structures have been subjected to the vehicle loading from dumpers laden with concrete blocks; and cantilever canopies have been used to store significant quantities of construction materials. Designers need to be mindful of construction loads, particularly on slabs, but the main responsibility for the prevention of overloading lies with the site staff. An allied problem is the premature removal of propping, even temporarily. Proprietary formwork systems have been patented which allow the removal of


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shutters without disturbing the props. However traditional arrangements are required on occasion and the props may be removed temporarily to allow formwork striking. Cracking may result if an incomplete set of props are reinstated, especially when headers are not used. Another aspect of the propping problem is the need for adequate back-propping in multi-storey work. Even mature concrete can be overstressed by the point loads transmitted through props from construction work overhead if back-propping is inadequate.

Cracking during the service phase Structures and buildings undergo measurable change in position, shape and volume during the service phase. This results from the reaction of the artefact with its surroundings. Chronologically the dominant movements are caused by settlement, response to load, drying shrinkage, and the response over the long term to cyclical climatic patterns, both daily and seasonable. Restraint of the movement associated with any of these factors may cause cracking. Cracking during the service phase may also occur due to the phenomena of alkali–silica reaction, freeze/thaw effects, sulfate attack, and delayed ettringite formation. These topics have been covered in earlier chapters. One further issue is that of fire. Cracking may occur in structures damaged by fire due to the elevated temperatures. Internal and surface cracking may occur as a result of the elevated temperature of the concrete and by stresses induced through restraint of thermal movement of members. The topic is comprehensively covered in a Concrete Society (1990) report. Settlement An important aspect of structural design is allowance for movement caused by settlement. Settlement in itself may not be harmful but differential settlement can be problematic. Differential settlement of monolithic structural elements invariably leads to cracking due to the excessive strain on joints. Two significant factors in design are allowance for the reaction of the founding strata during load transfer and the effect of the structure or building on the local water table. Subsidence may also be a particular problem in mining areas. The incidence and degree of cracking in new construction from excessive or differential settlement can be reduced by careful planning and interpretation of the site investigation. It is an aspect of practice where experience and judgement are invaluable allies. Ideally, the site investigation would be sufficiently comprehensive to pick up any significant strata variations or unusual features of the water table. The designer and contractor also need to consider the implications of construction sequence and technique on both the new and existing structures. Underpinning existing structures must be carried out in such a sequence that

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minimises the risk of settlement cracking. The new construction may alter the local water table permanently or temporarily. This could have significant implications on the bearing capacity of the strata carrying adjacent existing buildings. Temporary dewatering by well-pointing is very efficient but the effect on neighbouring buildings needs to be considered. The construction technique or building function may lead to the transmission of vibrations through the strata. This is particularly noticeable in the case of piling operations involving driven precast piles. A misleading source of cracking in concrete structures can be heave. If part of a building moves upwards it is likely that the stationary section would at first be thought to be settling. Wasteful and ineffective remedial measures might ensue. Heave is ordinarily avoided by placing the foundations sufficiently below ground level that they would not be affected by frost. However, Lossier (1955) documented an interesting instance of heave in a French multi-storey building. The building had a tall central section with shallower wings attached. Cracks appeared in service at the junctions of the central section and the wings. Settlement was suspected but the crack pattern indicated that the shallower wings were moving down relative to the core. The wings were underpinned but the problem persisted. Later it was found that the central section was moving up due to frost heave. The basement of the central section contained a cold chamber operating at 30°C. The chamber was poorly insulated and the water in the surrounding soil froze. The demands of minimising the period from project start-up to commissioning a building sometimes involves some overlap of the design and construction phases. The structural designer needs to be mindful of the effects of seemingly minor architectural revisions after the foundations are cast. Pad and strip footings would have been sized for specific loads. Local increases or decreases in load, or alterations to the load path, may be sufficient to initiate settlement cracking. Response to load Tension reinforcement in an economically designed member will lengthen to a greater extent than the elastic strain capacity of concrete. Thus cracking may be expected but is generally kept within accepted limits through prudent design and detailing. Clearly serious cracking could occur in cases of accidental overloading but even within design load limits the nature of the loading may lead to unanticipated stress that could cause serious cracking. Features that promote stress concentration are also an obvious source of crack initiation and examples include the concrete under bends in reinforcement, angular junctions of members, and at the corners of opes. The response of a reinforced concrete structure to dynamic loading requires careful consideration. Concrete can transmit vibration but is not a good dissipater of shock waves. An initial compressive shock wave may give rise to a reflected tensile shock wave. Although the initial stress may not be


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significant Radomski (1975) reported that the frequency of loading might lead to significant tensile stress through superposition of shock waves. The isolation of vibrating machinery by dampers or movement joints may be required. Repeated loads may cause a fatigue response in both concrete and reinforcement. Crack initiation and stable growth in reinforcement is well understood but the more complex effect in concrete has not been fully researched. Smolira (1972) suggested that the modulus of rupture of concrete might be reduced by up to half at one million oscillations of a stress within its working value. Cracks caused by vibration, shock, or fatigue usually occur between the aggregate and the cement matrix due to their different transmission rates. Where bond failure is suspected of having led to cracking confirmatory evidence may be gained from the fact that the cracks should not originate at the edge of the member. Cracks may also be caused by creep. Creep is the slow deformation of concrete under sustained load. Fortunately the effects of creep are often counterbalanced by the effects of drying shrinkage and cracking is avoided. Drying shrinkage Cracking during the service phase is caused by overstressing of the tensile zone or inducement of tensile stresses in the zones designed for compression. These stress conditions may arise if drying shrinkage is restrained. Drying shrinkage is the decrease in size of a reinforced concrete member due, primarily, to moisture loss. Shrinkage is encouraged by factors such as low environmental relative humidity, large surface area, high water content, and poor curing. Drying shrinkage cracks may be conclusively diagnosed if they occur before the dead load is applied. However, they are sometimes confused with plastic cracks. The distinction is easily drawn if it is known when the cracking occurred. Shrinkage may be 33 per cent complete after six months and 90 per cent complete after twenty months, by which stage the effect of creep may counterbalance the shrinkage. The surface layers of a member lose water before the inner layers and thus the rates of shrinkage vary. The interior therefore acts as a restraint and may induce cracking. The influence of reinforcement may be adverse during the early stages of drying shrinkage because it can act as an internal restraint. The cracks may occur as isolated surface cracks in ground floor slabs or may take a similar pattern to the early thermal cracks illustrated in Figure 11.11. Mackey and Wan (1975) identified drying shrinkage cracks in a field study of deep beams. The cracks occurred as shallow vertical hairline cracks on the side of the beams with no variation in width over the length of the cracks. Drying shrinkage cracking primarily results from a failure to accommodate movement. The primary method of designing for movement in buildings

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involves the provision of adequate movement joints. Movement joints are best considered as predetermined crack positions where appropriate details can be incorporated in design so that the building as a whole functions as intended, with a minimum of maintenance. The final location of these joints is determined by the architect but the structural engineer will advise on the need for joints at intervals of thirty to sixty metres and at changes of building mass or construction type. It is not always feasible to provide twin supports at every movement joint, in which case movement is accommodated on sliding bearings. A poorly constructed bearing surface, or inadequate performance of a friction pad, could cause significant cracking in the supporting structure due to the introduction of a horizontal force, resistance for which may not be provided. Cracking will also result if the structure is forced to complete an incomplete movement joint. The use of incomplete movement joints can be prudent in certain circumstances. For example a greater number of movement joints are required in a long canopy than in the building to which it is attached. However, the designer must be satisfied that such incomplete movement joints may be terminated at the location proposed. The validity of this practice is the observed fan action of structures, whereby the movement at joints decreases with proximity to major restraint. Cyclical movement One of the causes of movement in reinforced concrete buildings and structures is the variable temperature and moisture conditions to which it is subjected by both its internal and external environment. It is well understood that the heating and cooling of materials leads to an expansion and contraction process which may or may not restore the material to its original size. In the case of unrestrained reinforced concrete it has been noted that a daily cycle pertains whereby the structure will expand during the day and contract to the same extent at night. Superimposed on this diurnal pattern many observers have detected a seasonal pattern of progressive contraction throughout the autumn and winter. Dimensional changes may also take place due to moisture content variation with changes in relative humidity. The consequence of restraining these movements would be the generation of secondary stresses in the structure, which it may be unable to contain. Stresses may also be induced by a variation of body temperature throughout the structure. This may be caused by variation in the exposure of the structure to solar radiation due to its aspect and shelter. The thermal inertia of concrete results in different thermal conditions existing in the structure at any given time. This thermal behaviour may change during the service life of the building if shelter from trees or adjacent buildings is varied by growth, demise, construction or demolition. Shelter will also influence the degree to which the structure is affected by rain and wind and therefore the distribution


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of moisture content. Thus the appearance of cracks many years after a building has been commissioned may be due to differential thermal or moisture regimes not hitherto experienced. Differential response of materials to environmental conditions in service may cause cracking. A building may be regarded as a combination of a structure and a variety of non-structural elements. The materials forming structural and non-structural elements will behave differently when subjected to changes in temperature, stress or saturation condition. The migration of moisture from one material to another may also give rise to differential effects. The interaction of the elements may involve restraint on the movement of one element by another. If the differential response of materials induces stress by restraint in reinforced concrete the risk of cracking is increased. Interestingly at least one failure has been recorded due to differential shrinkage between precast concrete and insitu concrete. The Institution of Structural Engineers (1971) documented the case in which precast floor slabs were supported on insitu concrete beams. Higher shrinkage in the insitu work induced horizontal forces in the restrained precast slabs that cracked and ultimately collapsed. Investigators found that the widely differing shrinkage rates were due to the different aggregates in the two mixes. Thus the avoidance of cracking caused by inappropriate restraint involves the design of suitable connections or joints between different materials.

Cracking and corrosion of reinforcement A cracked reinforced concrete structure is not necessarily less durable than an uncracked structure. However corrosion of reinforcement in a structure will often be first detected through the appearance of cracks. A view therefore developed that cracking influenced the rate of corrosion. The nature of corrosion leads to expansive forces that lead to crack formation, widening of cracks and eventual spalling. It may be difficult to differentiate cause and effect and conceive that the opening of the cracks accelerates the corrosion process. This has led to two common but diverse views on the relationship between cracking and corrosion rate. One view is that cracks promote corrosion and significantly reduce service life by allowing ease of access to carbon dioxide and chlorides during the initiation phase, and to oxygen and moisture during the propagation phase. The other view is that the initiation phase is facilitated by cracking but that the propagation phase is not significantly effected and therefore the overall influence on service life is not significant. The diverse views have been reconciled in publications such as those by the Building Research Establishment (1993), the Swedish Concrete Association (1994), and the Concrete Society (1995). It is argued that both views are valid but that their applicability is dependent on the geometric orientation of the cracks in relation to reinforcement. The ability of the cracks to seal in the zone around the reinforcement is also significant.

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Crack width limits in codes of practice were related to durability but Beeby (1978a, 1978b) questioned the validity of the relationship. He noted that the crack width limits in the codes were based on short-term research, the findings of which were not borne out by tests carried out over ten years. Beeby concluded that cracks can increase the rate of corrosion initiation by allowing ingress of depassivating agents. The corrosion rate, however, is controlled by the equality of cathodic and anodic reactions. Assuming that cathodic sites lie predominantly in crack-free zones the corrosion rate would be much more influenced by the permeability of the uncracked concrete than the crack width at anodic sites. There is now general support for Beeby’s findings in so far as they relate to relatively isolated cracks transverse to the direction of reinforcement, known as ‘intersecting cracks’. Although Beeby’s research highlighted that service life is not necessarily directly related to crack width there is a relationship with corroded length. The wider the surface crack width the greater the corroded length of bar at the junction of the crack and the reinforcement. Thus durability may be compromised if there are a large number of intersecting cracks. Therefore prudent design and detailing involves limiting the cracks to the smallest number possible. Given that a tensile stress may be relieved by a large number of narrow cracks or by a small number of wide cracks, it seems that the latter option is better from the durability viewpoint. A serviceability requirement may however limit the crack widths to a value dictated by aesthetic considerations or liquid retention. A serious durability concern relates to ‘coincident cracks’. These are cracks that align with reinforcing bars and occur over them. Clearly coincident cracks allow ease of access for moisture and oxygen to cathodic sites. Both initiation and propagation phases of corrosion would be adversely influenced and service life reduced. A secondary factor in assessing the durability threat is the ability of cracks to seal themselves and thereby impede access to reinforcement of depassivating agents or oxygen and moisture. Autogenous healing involves the deposition of material within the crack. The deposits may be rust from the corroding reinforcement or calcium carbonate formed from calcium hydroxide through carbonation. Hydroxyl ions migrate to the site and re-alkalise the concrete leading to passive conditions. The likelihood of autogenous healing depends on whether the cracks are either intersecting or coincident, and either dormant or live. The width of a dormant crack is constant whereas the width of a live crack varies with time. Clearly intersecting and dormant cracks are the most likely to self heal.

Summary Cracking is an integral aspect of the use of reinforced concrete. The occurrence of cracking in a structure, whether specifically anticipated or not, does not necessarily imply that service life will be diminished. However the durability


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of concrete structures may be compromised by cracking in certain circumstances. These primarily involve the occurrence of cracks whose orientation, frequency, and status increases the exposure of cathodic zones of corroding reinforcement to the atmosphere. Although it is difficult to avoid cracking, the limitation of its occurrence or its extent can help to ensure durability. The threat of structural cracking can be eliminated by satisfactory design calculation but non-structural cracking is more difficult to avoid. A thorough understanding of the roots of nonstructural cracking combined with experience gained from reviewing case studies will provide a formidable force with which to counter the threat.

12 Abrasion, erosion and cavitation

Surface deterioration Surface deterioration can be a feature of concrete in contact with traffic or moving materials including fluids. The term ‘abrasion’ is typically used in the context of deterioration through friction between dry surfaces. Two forms of damage are distinguishable where concrete is in contact with flowing water: erosion and cavitation. In one case the surface deteriorates at an even rate over its surface area. In the other, local pitting of the surface results at positions where the flow of water is disturbed by a change in direction or velocity. The first case is classified as erosion whereas the second is a result of cavitation. Erosion may also occur from wind-borne sand particles striking concrete surfaces. All concrete surfaces may be expected to have a degree of resistance to wear but particular uses of buildings or infrastructure demand the explicit specification of abrasion-resistance. As an aside it may be worth noting three circumstances that give rise to soft floor surfaces that perform badly under even light traffic. The first is through carbonation of immature concrete surfaces, typically during cold-weather construction through elevation of carbon dioxide levels from unventilated heaters. The second arises from use of excessively wet mixes and where the bleed water is mopped up by the sprinkling of cement during finishing operations. The third circumstance is where the surface is excessively wet through over-trowelling. Some national codes hitherto included abrasion resistance through inclusion of abrasive conditions in one of the exposure classifications. An exposure category based on a series of abrasion resistance classes (the ‘XW’ series) was considered for inclusion in European Standard EN 206–1 but did not form part of the final draft. However, the German national application document, DIN 1045–2, includes guidance on concrete composition for three exposure classes, designated XM1, XM2 and XM3. Clause 6.2.3 of Standard EN 206–1 makes provision for additional requirements in specification of concrete through the use of performance requirements and test methods where they are appropriate. Resistance to abrasion is one of the items that may be specified. The United Kingdom national complementary standard

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BS 8500 includes guidance on additional requirements for aggregates to be used in wearing surfaces or those subject to abrasion. Wear resistance specified by performance is referred to in a developing European standard (EN 13813) ‘Screed materials and floor screeds – Properties and requirements of screed materials’. Performance criteria, based on a test method developed in the United Kingdom, were also included in a Concrete Society (1994) technical report. Further research has been used to develop performance criteria incorporated in a British Standard, BS 8204: Part 2 (BSI 1999c).

Abrasion Abrasion of concrete surfaces results from friction, which may cause a grinding action, or by repetitive impact and overloading, which causes local crushing. The sources of friction include pedestrian or vehicular traffic, materials dragged across floors or pavements, and wind-borne particles impacting on façades. Normal vehicular traffic is not generally problematic but abrasion can result from exposure to studded tyres and chains. Points of highest friction are most effected such as at turns and other points subject to braking action. Research by Sadegzadeh et al. (1987) indicated that the key factor in the abrasive resistance of concrete is the porosity of the top millimetre of the slab. The research explained the significance of parameters previously identified in empirical research. Later research highlighted other important characteristics. It has been found that the production of abrasion-resistant surfaces primarily involves consideration of surface finishing technique, curing regime, aggregate characteristics, and cement content. Concrete compressive strength may also provide a relative indication of resistance. Abrasion-resistant floors and pavements may by achieved either in single layer or double layer construction. The latter is sometimes used where the upper layer is constructed with especially resistant, but more expensive, materials. This may be justified in circumstances where abrasion-resistance is a significant serviceability requirement. Surface finishing technique has a critical influence on the porosity of the surface layer and therefore on surface hardness. Early or over-trowelled surfaces perform poorly due to the excessive water content of the surface layers. Floating and trowelling should be delayed until the concrete has lost its surface water sheen. A particularly hardwearing surface for heavy-duty applications may be achieved through the ‘dry-shake’ technique. This involves the use of a properly proportioned mix including specially graded metallic aggregate. Durable surfaces in conditions where abrasion is expected involves particular attention to the achievement of well-hydrated surfaces. This is best achieved by prolonged curing. Extending the curing period from three days to a week apparently doubles the abrasion resistance while prolonging the period to two weeks may be required to achieve the quality specified. Specifications

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in Germany required a doubling of curing time compared with normal exposure conditions without abrasion (Marsh 1993). Surfaces that need to be kept damp for periods of up to 14 days may be trafficked if surface protection can be set up. Abrasion is resisted by the coarse aggregate near the surface as it protects the mortar from wear. Cement matrix is worn down quicker than aggregate. Specification of hard aggregate may be justified in certain projects and Halvorsen (1993) recommends quartz, traprock or emery. Deterioration may also be experienced where the abrasive effect is one of pulling aggregate particles out of the surface. Large particles survive better than small ones. Webb et al. (1996) found that floors constructed with lower grade coarse aggregate and exposed to light or medium traffic can have satisfactory abrasion resistance if the fine aggregate is of good quality. The important role of aggregate leads to an optimum cement content level. Dhir et al. (1991a) tested four mixes at a fixed free water/cement ratio of 0.55 and cement contents of 300, 340, 350, 365 kg/m3. Highest resistance was found at the lowest cement content although there was little difference between the 300 kg/m3 and 340 kg/m3 mixes. Neville (1995) states that a cement content of 350 kg/m3 should probably be regarded as a maximum. This reflects the need for an adequately high aggregate content and close proximity of aggregate to the surface. The tests also demonstrated the benefits of reduced water content for a given water/cement ratio and the possibility of less bleeding. Concretes of high compressive strength perform better than weaker ones. Gjorv et al. (1990) noted that high strength concretes had better abrasive resistance than normal strength ones. High strength in this context refers to a subset of the new generation of high performance concretes and covers strengths in excess of 80 MPa. The use of fly ash up to 30 per cent has been found to increase abrasion resistance. Mitchell (2000) notes that the inclusion of polypropylene fibres in a mix also enhances abrasion resistance. On the other hand some difficulties may be experienced in achieving the required surface resistance if high levels of air-entrainment are used. Numerous test methods exist for assessing abrasion resistance. Typically the surface is exposed to rotating or rolling wheels for a period and the mean depth of wear is measured. A new European Standard, to be EN 13892, is being developed on floor screeds and in-situ floorings. The drafts of prEN 13892 include three test methods. These are the wear resistance test developed in the United Kingdom by the Cement and Concrete Association and Aston University, described in British Standard BS 8204: Part 2 (BSI 1999c); a German test which uses a standard cube and a grinding wheel; and a slightly modified version of a Swedish test which utilises a cylindrical rolling wheel test. There are also several ASTM standard tests that provide a measure of abrasion resistance. These include C418, C779, and C944. Test method C418 (ASTM 1990), involves blasting with air-driven silica sand. Procedures are

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described in C779 (ASTM 1995a), for inducing wear through rotating devices. Cores may be tested by method C944 (ASTM 1995b). Performance criteria are exemplified by those in British Standard BS 8204: Part 2: (BSI 1999c). Maximum wear depth is specified for various duties. At one end of the scale, surfaces subject to severe abrasion require special concrete mixes, dry-shake or sprinkle finish, and a limit of 0.05 mm maximum wear depth. At the other end, surfaces subject to moderate abrasion require a specified concrete grade and minimum cement content, and a limit of 0.4 mm maximum wear depth.

Erosion Erosion is caused by heavy particles suspended in water flowing across concrete surfaces or by wind-borne sand particles impacting on façades. The degree of water-borne damage is influenced by the velocity of flow and is exacerbated by the presence of eddies. The quantity of the particles being transported also contributes as does the size and shape. It is difficult to achieve permanent resistance to erosion unless the source of degradation can be eliminated. Ballim and Basson (2001) recommend exclusion of as much abrasive material as possible, for example through the construction of sand traps or the use of stilling basins to self-cleanse the water. Measures to reduce the velocity of flow would also be beneficial. Resistance to erosion from a concrete technology viewpoint is achieved through the use of measures similar to those discussed for general abrasion resistance. A test method that specifically simulates abrasion by water-borne solids is ASTM (1989b) C1138. The test involves wearing down a concrete surface using steel balls moving at high speed in a water tank.

Cavitation Pitting of surfaces arises where local pressure changes in flowing water leads to the collapse of vapour bubbles. This phenomenon – known as cavitation – may occur in the same location over a piece of concrete due to the layout of a hydraulic channel or pipe. A liquid flowing at high velocity develops a zone of sub-atmospheric pressure immediately downstream of an abrupt change in velocity or direction. Pockets of vapour form but collapse by implosion on entering the next zone of higher pressure. This leads to an impact effect and pressure waves strike the concrete surface. This continual effect may eventually cause local loosening of the concrete. If the surface becomes pitted the disturbance to the concrete flow will worsen and may cause an acceleration of damage rate. Once a large area is damaged the rate of deterioration reduces. Elimination of a potential location of cavitation at the design stage is the most effective method of managing the problem. If this is not possible,

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Graham and Lindholm (1978) point out that aeration of the water in the area of damage can help to reduce the degradation by a cushioning effect. Production of concrete surfaces resistant to cavitation is difficult but a number of approaches are possible. Resistance is offered by the fine-grained mortar. Compressive strength may be significant from the viewpoint of bond, which helps prevent aggregate detaching from the surface. Controlled permeability formwork has been used on dam spillways to enhance the surface properties.

13 Weathering and efflorescence

The brief for major infrastructural and building projects often includes the requirement that the surface finish should be of a high standard. There is an implication, particularly on public works projects, that the surface should remain visually attractive over a prolonged life with little or no expectation of costly maintenance. Such structures and buildings should, at the very least, grow old gracefully. A common public perception of concrete as something to be taken for granted rather than valued, can quickly lead to a perception that any new building with a concrete façade is to be tolerated rather than admired. Worse still, if the façade loses its uniformity of appearance through weathering or efflorescence any prejudice against concrete is reinforced.

Weathering The failure of a concrete surface to maintain a specified or implied level of aesthetic appeal during its intended service life is usually due to patchy discolouration or algal growth. The former is invariably unsightly whereas concern about the visual effect of the latter is a matter of subjective opinion. Failure through discolouration generally takes the form of a lack of uniformity in the colour of the façade. Thus buildings in polluted environments whose surfaces progressively discolour over time may not be problematic if the change is uniform across the surface area. This is difficult to achieve but control of the water runoff on façades can minimise the effects of weathering. Negative effects of weathering may generally be traced to the effect of different rates of flow of water on a façade. Differing vertical path lengths over differing materials tend to segregate a building façade into sets of vertical panels that weather at different rates. The rainwater flow over a panel consisting of glass and concrete, for example, will be more rapid than one of concrete alone. The flow over a low permeability concrete will be quicker than over one of high permeability. Flow rate over smooth concrete surfaces exceeds that over rough ones. The effect of rain may be either to wash dirt off surfaces or to streak dirt from lodgement points on horizontal surfaces. Both effects can be detrimental since the former can lead to clean streaks on otherwise dirty surfaces.

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Weathering is influenced by four principal aspects of an environment. The first is the degree of environmental pollution. Structures and buildings in urban sites require more considered detailing than those in rural settings. The second aspect is the wind pattern. Locations exposed to prevailing winds are subject to differential weathering of the most exposed face from the others. The average wind strength will also influence weathering through the driving rain index. High driving rain indices may lead to a well-washed surface on the façade facing the prevailing wind. The other faces get rain orientated such that it mainly falls on sills and string courses, washing dirt accumulations down the surface in streaks. This is also related to the third aspect, the rain pattern. Clearly the total amount of annual rainfall is significant but so too is the annual number of days on which rain falls. This latter point gives an indication of the duration of dry periods. Finally the acidity of the environment, reflected in the sulfur dioxide levels, is significant. Control of the problem involves consideration, at the design stage, of water flow patterns in service. Appropriate details can then be incorporated to pre-empt as many sources of uneven weathering as possible. The primary concern is the avoidance of concentrated flows on panels. Vertical panelling and horizontal banding may assist. Fenestration, leading to rapid flow, could be positioned above rough concrete surfaces and thereby contrast with neighbouring smooth panels so that differential weathering patterns would not be as noticeable. Horizontal bands, used to reduce flow path length, could be clearly delineated by projections. These projections would need to incorporate drip grooves. The variety of concrete finishes available can also be used to differentiate panels or faces likely to weather at different rates. Rough surfaces disguise the detrimental effects better than smooth ones. Where smooth surfaces are used those with highly impermeable concrete, manufactured with non-absorbent aggregates, weather best. Control of differential water flow rates by tilting some panels may also be beneficial. It is difficult to be prescriptive in the control of weathering since each project is a unique combination of a structural entity and a location. However, much can be learned from a study of existing buildings, particularly those in the vicinity of the proposed new works. Guidance, based primarily on research by Huberty (1980), may be found in the design guide of Comite Euro-International du Beton (1989). Reviews of existing buildings, in publications such as the British Cement Association’s Concrete Quarterly, are also most informative.

Efflorescence Efflorescence – the appearance of whitish deposits on the surface of concrete – may be considered a durability failure if appearance is critical but is usually amenable to remedial action. Efflorescence is caused by the capillary transport of calcium hydroxide to the surface where it is leached out and deposited on the surface, or by soluble salts that migrate to the surface zone and are

Weathering and efflorescence 239 deposited by evaporation. In time the calcium hydroxide may carbonate and stain the concrete surface. The phenomenon is most prevalent where cyclical weather patterns are experienced of dry, warm spells followed by wet, cool periods. Three forms of efflorescence are distinguishable (Higgins 1982): lime bloom, lime weeping, and salt crystallisation. Lime bloom occurs when calcium hydroxide migrates to patches of the surface or the entire surface area where it reacts to form a thin film of calcium carbonate. The film may be so thin that it becomes transparent on wetting but reappears on drying. Equally it may give the appearance of fading where it occurs on coloured concrete. Lime bloom generally appears early in the life of a structure, if at all, and disappears by natural weathering as the calcium carbonate, CaCO3, changes to the more soluble calcium hydrogen carbonate, Ca(HCO3)2, which is washed off. In some cases it may be necessary to give nature a helping hand by washing off the bloom with water or acid. Bensted (1994) cautions that remedial treatment involving acid is best undertaken by experts to ensure that the acid solutions have very limited contact time with the concrete and are fully washed off. Dilute hydrochloric acid (HCl), for example, even used as 10 per cent solution in water, can lead to corrosion of embedded metals if left in contact too long. Lime bloom should therefore be of little consequence in a long-term durability context. Lime weeping is related to lime bloom but takes the form of thick white deposits, which originate from points of water leakage from the structure, such as from cracks or joints. In severe cases stalactites may grow where the water drips off a free edge. An example of lime weeping is illustrated in Figure 13.1. It may be observed, for example, in old water-retaining structures, earth-retaining walls, bridges and multi-storey car parks, where concentrated flows occur. The encrustation is usually too thick to disappear by natural weathering and would require mechanical removal. The problem reoccurs however if the source of the problem is not tackled. Salt crystallisation occurs where soluble salts, not normally found in concrete, migrate through the concrete to the surface zone. The salts usually originate from external sources, such as seawater and groundwater, but could also originate from contaminated mix constituents. The salts migrate to the surface zone where evaporation leads to their deposition on the surface. More worryingly, in porous concretes the evaporation may take place before the salts appear on the surface leading to saturation and crystallisation below the surface. Restraint of crystal growth by the narrowness of the concrete pores and the volume expansion consequent on the hydrate reaction may lead to disruptive stresses. This form of efflorescence is uncommon in European practice but significant problems have been reported in California and in Adelaide, Australia. Efflorescence is thus rarely a source of structural durability failure but it should be noted that the products of hydration are soluble. However the quantity of water required for dissolution is relatively high even for the most soluble product, calcium hydroxide. A litre of water is required to dissolve

240


Figure 13.1 Example of weathering due to lime weeping

1.85 grams of calcium hydroxide at a temperature of 0°C (Swedish Concrete Association 1994). Impermeable concrete subject to rainfall is therefore not at risk but problems could arise in structures such as dams where the quantity of flow could be high enough to cause loss of monosulfate and ettringite and decalcification of calcium silicate hydrates. Prevention of efflorescence involves hindering the flow of water through the concrete. Drainage has an effective role to play, such as by the incorporation of weep holes in retaining walls. The production of dense impermeable concrete is obviously important. The incorporation of ggbs, pfa, microsilica, or metakaolin can be helpful. These materials not only aid impermeability but also lower the calcium hydroxide content by reacting to form calcium silicate hydrate. However Suprenant (1992) notes that the reaction with calcium hydroxide may be too slow with some pozzalans to pre-empt the problem and advice is given on testing suitability. Surface treatments may also be effective.

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Index

Abrasion of concrete, 232 effect: of aggregate, 234; of air entrainment, 234; of cement content, 234; of curing, 233; of fly ash, 234; of porosity, 233 exposure class XM1, XM2, XM3, 232 exposure class XW, 232 testing, 234 Absorption, 39, 45, 112, 114, 127, 164 Accelerated mortar bar test, 156 Acid attack contaminated sites, 194 effect: of aggregate, 201; of cementitious properties, 200; of evaporation, 198; of flow rate, 200; of fly ash, 204; of ggbs, 204; of metakaolin, 200; of permeability, 199; of pH, 199, 205; of silica fume, 200, 204; of solubility of product, 200; of specific acids, 195, 196; of specific alkalies, 195; of specific salts, 195; of temperature, 201 exposure class XA1, XA2, XA3, 204 groundwater analysis, 205 mathematical modelling, 201 seawater, see Seawater Attack sewers, 194, 197, silage, 194, 196, soil analysis, 205, specification, 203, 206, testing, 206 Acetic acid, 195, 196, 200 Acid soluble equivalent sodium oxide content, 143 Adsorption, 39, 104 Aggregate absorption, 164

–alkali reactivity assessment, 152 greywacke, 156 influence: on abrasion resistance, 234; on acid attack, 201; on chloride ingress, 113; on freeze/thaw attack, 164 petrographic classification, 153 petrographic examination, 152 porosity, 148 sea-dredged, 135, 144 voids, 43 Air entrainment, 9, 40, 165, 234 Alkali–aggregate reaction, 133 Alkali–carbonate reaction, 133 Alkali–silica reaction acid soluble equivalent sodium oxide content, 143 alkali content of additions, 145 alkali content of admixtures, 144 alkali content of aggregates, 144 alkali content of cement, 143 alkali content of concrete, 142 cracking, 136 effect: of alkali level, 142, 159; of fly ash, 145, 158; of ggbs, 145, 158; of lithium hydroxide monohydrate, 147; of lithium nitrate, 147; of metakaolin, 147; of reactive silica, 141; of salts, 147; of temperature, 148; of silica fume, 147; of water, 141, 152 greywacke, 157, mathematical modelling, 148, mechanism, 139 pessimum effect, 140, 142, 145 specification, 148 Alkali–silicate reaction, 133, 135 Alumino-ferrite-mono, 181

256

Index

Alumino-ferrite-tri, 181 Ammonium ion level, 205 Ammonium sulfate, 180 Anion, 55 Anode, 53, 63, 75, 138, 230 Aragonite, 78 Arrhenius equation, 124 ‘Autoclam’ test, 48 Autogenous healing, 69, 220 Blastfurnace slag, see Ground granulated blastfurnace slag Boltzmann’s Transformation, 119 Brucite, 183 Bubble spacing, 166 Calcite, 78 Calcium aluminate hydrates, 105 Calcium carbonate, 69, 77, 202, 239 Calcium chloride, 101, 102, 125 Calcium chloroaluminates, 105, 109 Calcium chloroferrites, 105 Calcium humate, 200 Calcium hydrogen carbonate, 202, 239 Calcium hydroxide, 40, 77, 89, 140, 194 Calcium oxide, 96 Calcium silicate hydrates, 40, 78, 89, 194 Calcium sulfate, 180, 182 Calcium sulfoaluminate hydrates, 181 Capillary rise, 39 Carbon dioxide, 77, 89 Carbonated water, 194 Carbonic acid, 197 Carbonation of concrete, 77 effect: of cement content, 97; of cement type, 82, 96; of chlorides, 104; of cover, 86, 95; of curing, 82; of diffusivity, 81; of environmental CO2 level, 83; of fly ash, 82; of hydration, 89; of permeability, 81; of porosity, 89; of relative humidity, 84, 95; of temperature, 82; of water/cement ratio, 82, 86, 90; of wetting/drying, 93 exposure class: XC1, 97; XC2, 98; XC3, 98; XC4, 99 mathematical modelling, 85 measurement, 79 performance-based specification, 99 reaction engineering, 89 surface effect, fresh concrete, 232 testing, 79

Capillary absorption, 111, 127, 128, 198 Cathode, 53, 62, 75, 138, 230 Cation, 55, 140 Cavitation, 232, 235 Cement content, context, 4 Chalcedonic flint, 134 Chalcedony, 142 Chemical attack by acid, see Acid attack by seawater, see Seawater attack by sulfates, see Sulfate attack Chert, 142 Chloride ingress admixtures, 125 binding, 109, 112 corrosion, 101, 110 effect: of aggregate diffusivity, 113; of carbonation, 104, 106, 114; of curing, 109; of fly ash, 122, 127, 129; of ggbs, 122, 127, 129; of hydrostatic head, 114; of metakaolin, 111; of silica fume, 127; of temperature, 114; of water/cement ratio, 113 effective diffusion coefficient, 105, 111, 113, 122, 129 exposure class; XD1, 127; XD2, 128; XD3, 128; XS1, 125; XS2, 126; XS3, 127 initial internal level, 125 mathematical modelling, 114, 123 measurement, 105 surface chloride level, 115, 122 temperature, 108 testing, 105 threshold for corrosion, 107 Capillary attraction, 38 Cold weather concreting, 232 Concreting in cold weather, 232 in hot weather, 187 Condensed silica fume, see Silica fume Controlled permeability formwork, 12, 112, 236 Corrosion current, 57, 60, 75 depassivation, 51 effect: of carbonation, 53, 62, 70; of chlorides, 53, 63, 101; of cracking, 67, 229; of oxygen supply, 65; of relative humidity, 66; of temperature, 67 electrochemistry, 54

Index inhibitors, 12 mathematical modelling, 69 rate, 57, 70 significance, 1 Cover to reinforcement, 8, 11, 12, 14, 77, 126, 222 Crack, see Cracking Cracking corrosion, 52, 229 crazing, 213 D-line cracking, 161 design aspects, 214 due to alkali–silica reaction, 136 drying shrinkage, 213, 227 early thermal cracking, 187, 212 effect: of age, 212; of aggregate interlock, 211; of bleeding, 222; of corrosion, 213; of cover, 210; of creep, 227; of curing, 222; of evaporation, 222; of fly ash, 223; of ggbs, 222, 223; of load, 213, 226; of moisture gradients, 228; of reinforcement spacing, 211; of tensile strain capacity, 209; of thermal behaviour, 213, 223, 228; of torsion, 212 plastic settlement, 213, 221 plastic shrinkage, 213, 222 settlement cracking, 213, 225 shear cracking, 211 width, 217, 218, 230 Crank, error function, 44, 114, 121 Crazing, 213 Creep, 227 Crypto-crystalline quartz, 142 Crystobalite, 142 Curing, 14, 40 D’Arcy’s law, 45 De-icing agents and salts, 101, 109, 113, 122, 128, 135, 147, 165, 221 Delamination, 52, 161 Delayed ettringite formation, 135, 180, 183, 191 Depassivation, 38, 56, 60, 78 Design life, 3, 20 Dicalcium silicate, 40 Differential thermal analysis, 79 Diffusion chloride diffusivity, 105 gaseous diffusion, 43 influence of moisture, 44 ionic, 39, 44

257

Dirac delta function, 118 Dry-shake, 233 Drying shrinkage, 138 Durability design, introduction, 35 Durability grades, 5, 6, 11, 12 Durability indexes, 129 Early thermal cracking, 187 Efflorescence, 138, 239, 240 Electrochemistry of corrosion, 51, 54 Electrode potential, 59, 108 Electrolyte, 56, 60, 77 Emery, 234 Entrained air, 42 Entrapped air, 42 Epsom salt, 182 Equivalent sodium oxide content, 143 Erosion, 232, 235 Error function, 44 Ettringite, 179, 181, 182, 194, 198, 240 Evan’s diagram, 60, 103 Exposure classification general framework, 9, 15 qualitative system, 6, 7, 12 XA, 190, 204 XC, 97 XD, 127 XF, 171 XM, 232 XS, 125 XW, 232 Faraday’s laws, 57 Ferric hydroxide, 52 Ferrous hydroxide, 52, 63 Ferrous oxide, 78 Fick’s laws, 36, 44, 88, 114, 117, 118, 121, 123, 201 ‘Figg-type’ permeability test, 48 Flint, 142 Fly ash influence: on abrasion resistance, 234; on acid attack, 204; on air entrainment, 166; on alkali–silica reaction, 158; on carbonation, 82; on chloride ingress, 122, 127, 129; on efflorescence, 240; on sulfate resistance, 187, 189 Formic acid, 200 Freeze/thaw action, 160 air entrainment, 163, 165 damage, 135, 161

258

Index

Freeze/thaw (cont.) effect: of admixtures, 166; of age, 163; of aggregate characteristics, 164; of alkali–silica reaction, 138; of climate, 164; of de-icing salt, 161, 162, 165; of fly ash, 166; of porosity, 163; of saturation, 162 exposure class; XF1, 171; XF2, 172; XF3, 172; XF4, 172 mathematical modelling, 166 performance-based specification, 176 tests: on fresh concrete, 173; on hardened concrete, 175 Frost heave, 226 Gel pores, 42 ggbs, see Ground granulated blastfurnace slag Glauber’s salt, 182 Grade, national durability, 5, 6, 11, 12 Greywacke, 157 Ground granulated blastfurnace slag (ggbs), influence: on acid attack, 204; on alkali–silica reaction, 158; on chloride ingress, 127, 129; on cracking, 222; on efflorescence, 240; on sulfate resistance, 187, 189 Groundwater, saline, 122 Gypsum, 179, 182 Heave, 226 Hot-weather concreting, 187 Humic acid, 200 Hydration, influence on permeability, 40 Hydrogen chloride, 64 Hydrogen sulfide, 194, 197 Hydronium ions, 64 Ice, expansive effect, 160 Impermeability, see Permeability Infra-red spectroscopy, 79 Intended service period design method, 29 Internal restraint, 212 Intrinsic permeability, 45 Iron sulfide, 181, 197 Lactic acid, 195, 196, 200 Lifetime design method, 31 Lifetime safety factor method, 27 Lightweight aggregate, 43

Lightweight concrete, 40 Lime bloom, 239 Lime weeping, 239 Linear polarisation resistance, 59, 74 Low-alkali cement, 159 Magnesium hydroxide, 182 Magnesium ion level, 205 Magnesium sulfate, 180, 182, 198 Manx-cracking, 136 Marcasite, 194 Metakaolin influence: on acid attack, 200; on alkali–silica reaction, 147; on chloride ingress, 111; on efflorescence, 240; on sulfate resistance, 189 Microcrystalline quartz, 142 Microsilica, see silica fume Mixed electrode, 56 Modelling context, 10, 17 degradation model, 27 deterministic, 20 Dirac delta function, 118 finite difference model, 117 Green element method, 118 Intended service period design method, 29 Lifetime design method, 31 Lifetime safety factor method, 27 Nernst-Einstein equation, 124 Nernst-Planck equation, 118 performance model, 27 rate: of acid attack, 201; of carbonation, 85; of chloride ingress, 110, 114; of corrosion, 69; of freezing resistance, 167 stochastic, 20 Monosulfate, 181, 240 National durability grade, 5, 6, 11, 12 Nernst-Einstein equation, 124 Nernst-Planck equation, 118 Neutron radiography, 79 Nitrates, 196 Opaline flint, 134 Opaline sandstone, 134 Opaline silica, 142 Ostwald’s law, 79 Oxalic acid, 199, Oxidation, 55, 64

Index Oxide film theory, 104 Oxychloride compounds, 64 Passivity, 61, 102 Performance-based specification introduction, 9, 10, 15 acid attack, 206 carbonation, 99 chloride, 128 freeze/thaw, 176 methodology, 17 sulfate resistance, 192 Permeability D’Arcy coefficient, 46 durability role, 5, 38 effect: of curing, 49; of entrapped air, 50; of hydration, 49; of water/cement ratio, 50 testing, 48 Pessimum effect, 140, 142, 145 pfa, see Fly ash pH, 62, 77, 105, 108, 183 Phenolphthalein, 79 Phosphoric acid, 200 Pitting corrosion, 52, 64, 103 Plastic settlement, 68, 221 Plastic shrinkage, 222 Polarisation curves, 60 Pop-outs, 52, 138, 161 Pore size distribution, 38 Porosity, 38, 40, 169 Potassium hydroxide, 61, 140 Potassium oxide, 62 Potassium sulfate, 180 Potential mapping, 71 Predictive modelling, see Modelling Prescriptive approach to specification, 11 Probability, 3, 10, 15, 20 Pulverised-fuel ash, see Fly ash Pyrite, 181, 183, 194 Quartz, 234 Reduction, 55 Reinforcement corrosion, see Corrosion Resistivity, 59 Reliability index, 3, 34 Risk analysis, 3, 18, 22 Salt crystallisation, 198, 239 Scaling, 161 Seawater, 147, 181

259

Seawater attack, 197, 206 Service life, 3, 28, 30, 32, 71, 94, 121, 123 Silica, 140 Silica fume influence: on acid attack, 200, 204; on alkali–silica reaction, 147; on chloride ingress, 108, 111, 127; on efflorescence, 240; on sulfate resistance, 189 Silicic acid, 199 Silver nitrate, 107 Sodium hydroxide, 61, 140 Sodium oxide, 62 Sodium sulfate, 180, 182 Sorptivity, 38, 45, 111 Sorptivity test, 47 Spalling, 52 Specific surface, 166 Specification acid attack, 203, 206 alkali–silica reaction, 148 carbonation, 99 chloride, 128 framework, 1 freeze/thaw, 176 performance-based, 17 prescriptive, 11 seawater attack, 204 sulfate attack, 188 Steel corrosion, see Corrosion Sulfate attack effect: of carbonate, 186; of carbonation, 186; of cement properties, 188; of evaporation, 186; of fly ash, 187, 189; of ggbs, 187, 189; of groundwater mobility, 185, 186; of metakaolin, 189; of permeability, 185, 188; of silica fume, 189; of sulfate concentration, 184, 186, 188; of sulfate solubility, 185; of temperature, 186, 187 exposure class XA1, XA2, XA3, 190 mechanism, 180 specification: general European practice, 188; delayed ettringite formation, 190; thaumasite attack, 189; United Kingdom practice, 192; testing, 185, 192; thaumasite, 180, 182, 183 Sulfate-resisting cement, 9, 64, 109, 184, 187, 189 Sulfuric acid, 194, 196, 197, 200

260

Index

Surface absorption test, 47 Surface area, see Specific surface Temperature influence: on acid attack, 201; on alkali–silica reaction, 148; on chloride ingress, 124; on thaumasite formation, 180 Testing alkali–silica reaction: in concrete prism test, 155; in mortar bar test, 156; in petrographic examination, 152; in quick chemical test, 153 carbonation, 79 chloride ingress: in conduction, 131; in diffusion cells, 130 context, 3 corrosion, 71 freezing resistance, 173, 174 linear polarisation resistance, 74 permeability, 48 potential mapping, 71 resistivity, 73 seawater attack, 206 sorptivity, 47 sulfate concentration, 185 sulfate resistance, 192 surface absorption, 47 water absorption, 47 Thaumasite, 180, 183, 189 Thermal shock, 162, 165 Thermogravimetry, 79 Thiobacillus concretivorous, 195

Tortuosity, 38, 91 Transitory complex theory, 104 Traprock, 234 Tricalcium aluminate, 109 Tricalcium aluminate hydrate, 179, 194 Tricalcium silicate, 40 Tridymite, 142 Valence, 58 Vaterite, 78 Volcanic glass, 142 Volhardt titration method, 107 Water absorption test, 47 Water/binder ratio, water/cement ratio durability role, 4 influence: on acid resistance, 199; on carbonation, 97; on chloride ingress, 113, 125; on sulfate resistance, 185, 192 Water/soil extract test, 185 Weathering effect: of climate, 238; of concrete permeability, 237; of environmental acidity, 238 salt weathering, 199 Wenner four point apparatus, 73 Wetting and drying, 39, 93, 101, 112, 116, 127, 128, 186 Working life, 9,98 X-ray diffraction, 79 X-ray fluorescence spectrometry, 107

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