Frontiers in Offshore Geotechnics II

PROCEEDINGS OF THE 2ND INTERNATIONAL SYMPOSIUM ON FRONTIERS IN OFFSHORE GEOTECHNICS, PERTH, AUSTRALIA, 8–10 NOVEMBER 20...

Author: Susan Gourvenec | David White

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PROCEEDINGS OF THE 2ND INTERNATIONAL SYMPOSIUM ON FRONTIERS IN OFFSHORE GEOTECHNICS, PERTH, AUSTRALIA, 8–10 NOVEMBER 2010

Frontiers in Offshore Geotechnics II Editors Susan Gourvenec & David White Centre for Offshore Foundation Systems, University of Western Australia

© 2011 by Taylor & Francis Group, LLC

Cover photo credit © Norske Hydro

CRC Press/Balkema is an imprint of the Taylor & Francis Group, an informa business © 2011 Taylor & Francis Group, London, UK Typeset by MPS Limited, a Macmillan Company, Chennai, India Printed and bound in Great Britain by Antony Rowe (a CPI Group Company), Chippenham, Wiltshire All rights reserved. No part of this publication or the information contained herein may be reproduced, stored in a retrieval system, or transmitted in any form or by any means, electronic, mechanical, by photocopying, recording or otherwise, without written prior permission from the publisher. Although all care is taken to ensure integrity and the quality of this publication and the information herein, no responsibility is assumed by the publishers nor the author for any damage to the property or persons as a result of operation or use of this publication and/or the information contained herein. Published by:

CRC Press/Balkema P.O. Box 447, 2300 AK Leiden, The Netherlands e-mail: [email protected] www.crcpress.com – www.taylorandfrancis.co.uk – www.balkema.nl

ISBN: 978-0-415-58480-7 (Hbk + CD-ROM) ISBN: 978-0-203-83007-9 (ebook) © 2011 by Taylor & Francis Group, LLC

Frontiers in Offshore Geotechnics II – Gourvenec & White (eds) © 2011 Taylor & Francis Group, London, ISBN 978-0-415-58480-7

Table of Contents

Preface

XIII

Committees

XV

Reviewers

XVII

1 Keynotes A systematic approach to offshore engineering for multiple-project developments in geohazardous areas T.G. Evans Recommended best practice for geotechnical site characterisation of cohesive offshore sediments D.J. DeGroot, T. Lunne & T.I. Tjelta Gulf of Guinea deepwater sediments: Geotechnical properties, design issues and installation experiences J.-L. Colliat, H. Dendani, A. Puech & J.-F. Nauroy Geotechnics for subsea pipelines D.J. White & D.N. Cathie

3 33

59 87

Axial and lateral pile design in carbonate soils C.T. Erbrich, M.P. O’Neill, P. Clancy & M.F. Randolph

125

New frontiers for centrifuge modelling in offshore geotechnics C. Gaudin, E.C. Clukey, J. Garnier & R. Phillips

155

Risk and reliability on the frontier of offshore geotechnics R.B. Gilbert, J.D. Murff & E.C. Clukey

189

2 Geohazards and gas hydrates Neotectonic deformation of northwestern Australia: Implications for oil and gas development J.V. Hengesh, K. Wyrwoll & B.B. Whitney

203

Deepwater Angola part I: Geohazard mitigation A.J. Hill, J.G. Southgate, P.R. Fish & S. Thomas

209

Deepwater Angola part II: Geotechnical challenges A.J. Hill, T.G. Evans, B. Mackenzie & G. Thompson

215

Shallow gas hazard linked to worldwide delta environments S. Kortekaas, E. Sens & B. Sarata

221

Analysis of submarine flow slides in fine silty sand P.V. Lade & J.A. Yamamuro

227

Hydrate dissociation around oil exploration infrastructure A.K. Sultaniya, J.A. Priest & C.R.I. Clayton

233

An investigation of past mass movement events in the West Nile Delta S. Thomas, L. Bell, K. Ticehurst & P.S. Dimmock

239

Deformation of seabed due to exploitation of methane hydrate reservoir J. Yoneda, M. Hyodo, Y. Nakata, N. Yoshimoto & R. Orense

245


V

3 In situ site characterisation and pore pressure measurement A site investigation strategy to obtain fast-track shear strength design parameters in deep water soils D. Borel, A. Puech & S. Po

253

Enhancement of the ball penetrometer test with pore pressure measurements N. Boylan, M.F. Randolph & H.E. Low

259

Laboratory free falling penetrometer test into clay S.H. Chow & D.W. Airey

265

Offshore sediment overpressures: Overview of mechanisms, measurement and modeling B. Dugan, T.C. Sheahan, J.M. Thibault & T.G. Evans

271

Angolan deepwater soil conditions: GIS technology development for sediment characterization P. Enjaume, M. Hamon, K. Epalanga & B. Mackenzie

277

Strength measurement in very soft upper seabed sediments P. Kelleher, H.E. Low, C. Jones, T. Lunne, S. Strandvik & T.I. Tjelta

283

CPT in polar snow – preliminary observations A.B. McCallum, A. Barwise & R. Santos

289

Parametric study of a free-falling penetrometer in clay-like soils M. Nazem & J.P. Carter

293

The future of deepwater site investigation: Seabed drilling technology? J.J. Osborne, A.G. Yetginer, T. Halliday & T.I. Tjelta

299

Mini T-bar testing at shallow penetration A. Puech, M. Orozco-Calderón & P. Foray

305

Piezometer installation in deepwater Norwegian Sea T.I. Tjelta & J. Strout

311

Luva deepwater site investigation programme and findings T.I. Tjelta & A.G. Yetginer

315

Investigations into novel shallow penetrometers for fine-grained soils Y. Yan, D.J. White & M.F. Randolph

321

Seabed drilling vs surface drilling – a comparison A.G. Yetginer & T.I. Tjelta

327

4 Soil characterisation and modelling Rheological behaviour of soft clays P.E.L. de Santa Maria, I.S.M. Martins & F.C.M. de Santa Maria

335

A three-dimensional finite element study of the direct simple shear test J.P. Doherty & M. Fahey

341

Repeated loading and unloading of the seabed H.J.E. Hu, K.K. Tho, C.T. Gan, A.C. Palmer & C.F. Leung

347

A new interpretation of the simple shear test H.A. Joer, C.T. Erbrich & S.S. Sharma

353

Physical modelling of the crushing behaviour of granular materials H.A. Joer & S.S. Sharma

359

New evidence for the origin and behaviour of deep ocean ‘crusts’ M.Y-H. Kuo, M.D. Bolton, A.J. Hill & M.J. Rattley

365

Soil unit weight estimated from CPTu in offshore soils P.W. Mayne, J. Peuchen & D. Bouwmeester

371


VI

Strain rate dependent simple shear behaviour of deepwater sediments in offshore Angola M.J. Rattley, A.J. Hill, S. Thomas & B. Sampurno Simplified calibration procedure for a high-cycle accumulation model based on cyclic triaxial tests on 22 sands T. Wichtmann, A. Niemunis & Th. Triantafyllidis Understanding cyclic loading behavior of soil for offshore applications J. Yang

377

383 389

5 Shallow foundations Observations of shallow skirted foundations under transient and sustained uplift H.E. Acosta-Martinez, S. Gourvenec & M.F. Randolph

397

Numerical study of grillage foundations on sand under combined VHM loading M. Banimahd, A. Maconochie & J. Oliphant

403

The vertical bearing capacity of grillage foundations in sand M.F. Bransby, P. Hudacsek, J.A. Knappett, M.J. Brown, N. Morgan, D.N. Cathie, R. Egborge, A. Maconochie, G.J. Yun, N. Brown & A. Ripley

409

Behaviour of skirted footings on sand overlying clay C.T. Gan, K.L. Teh, C.F. Leung, Y.K. Chow & S. Swee

415

Numerical study of piping limits for suction installation of offshore skirted foundations and anchors in layered sand L.B. Ibsen & C.L. Thilsted

421

Shallow foundation performance in a calcareous sand B.M. Lehane

427

A numerical study of the vertical bearing capacity of skirted foundations D.S.K. Mana, S. Gourvenec & M.F. Randolph

433

The effect of torsion on the sliding resistance of rectangular foundations J.D. Murff, C.P. Aubeny & M. Yang

439

Foundation design challenges of the MCR-A skirted gravity platform L. Tapper, C. Humpheson & B.M. Lehane

445

Constructing breakwater with prefabricated caissons on soft clay S. Yan, X. Feng & J. Chu

451

6 Piled foundations Simplified analysis of laterally loaded pile groups F.M. Abdrabbo & K.E. Gaaver

459

Behavior of piles under combined lateral and axial loading M. Achmus & K. Thieken

465

Investigations on the behavior of large diameter piles under cyclic lateral loading M. Achmus, J. Albiker & K. Abdel-Rahman

471

BP Clair phase 1 – Pile driveability and capacity in extremely hard till T.R. Aldridge, T.M. Carrington, R.J. Jardine, R. Little, T.G. Evans & I. Finnie

477

Photoelastic investigation into plugging of open ended piles J. Dijkstra, E.A. Alderlieste & W. Broere

483

Soil-pile interaction during extrusion of an initially deformed pile C.T. Erbrich, E. Barbosa-Cruz & R. Barbour

489

BP Clair phase 1 – Geotechnical assurance of driven piled foundations in extremely hard till T.G. Evans, I. Finnie, R. Little, R.J. Jardine & T.R. Aldridge

495


VII

Pile driving experiences in Persian Gulf calcareous sands K. Fakharian & I.H. Attar

501

FLAC3D analysis on soil moving through piles E.H. Ghee & W.D. Guo

507

Cyclic loading of barrettes in soft calcareous rock using Osterberg cells C.M. Haberfield, D.R. Paul, M.C. Ervin & G.A. Chapman

513

Shaft capacity of drilled and grouted piles in calcareous sandstone B.M. Lehane

519

Numerical analysis of mudmat contribution to capacity of piled offshore platforms L.S.D. Lorenti, M.A. Ismail & B.M. Lehane

525

Simplified numerical model for analysis of offshore piles under cyclic lateral loading M.M. Memarpour, M. Kimiaei & M. Shayanfar

531

Centrifuge modelling of rapid load tests with piles in silt and sand C.T. Nguyen, H. van Lottum, P. Hölscher & A.F. van Tol

537

Field measurements on monopile Dolphins A. Sadeghi-Hokmabadi & A. Fakher

543

Behaviour of driven tubular steel piles in calcarenite for a marine jetty in Fujairah, United Arab Emirates J. Thomas, M. van den Berg, F. Chow & N. Maas CPT-Based design method for axial capacity of offshore piles in clays B.F.J. Van Dijk & H.J. Kolk

549 555

7 Foundations for renewable energy Evaluation of pile capacity approaches with respect to piles for wind energy foundations in the North Sea M. Achmus & M. Müller

563

Installation of suction caissons for offshore renewable energy structures O.J. Cotter, B.W. Byrne & G.T. Houlsby

569

Lateral behaviour of large diameter monopiles at Sheringham Shoal Wind Farm L. Hamre, S. Feizi Khankandi, P.J. Strøm & C. Athanasiu

575

Centrifuge modelling of offshore monopile foundation R.T. Klinkvort & O. Hededal

581

Gravity based foundations for the Rødsand 2 offshore wind farm, Denmark L. Krogh, J.H. Lyngs & J.S. Steenfelt

587

Geotechnics for developing offshore renewable energy in the US M. Landon Maynard & J.A. Schneider

593

Engineering issues for fixed offshore wind turbines on Lake Michigan Mid Lake Plateau, USA P.J. Lang, J.A. Schneider, K. Smith & T. McNeilan

599

Centrifuge model tests on piled footings in clay for offshore wind turbines B.M. Lehane, W. Powrie & J.P. Doherty

605

Design of monopile foundations in sand for offshore windfarms M. Saue, T.E. Langford & N. Mortensen

611

Experimental evaluation of backfill in scour holes around offshore monopiles S.P.H. Sørensen, L.B. Ibsen & P. Frigaard

617

An investigation of the use of a bearing plate to enhance the lateral capacity of monopile foundations K.J.L. Stone, T.A. Newson, M. El Marassi, H. El Naggar, R.N. Taylor & R.J. Goodey

623


VIII

Optimizing site investigations and pile design for wind farms using geostatistical methods: A case study B. Stuyts, V. Vissers, D.N. Cathie, C. Jaeck & S. Dörfeldt

629

Towards the FE prediction of permanent deformations of offshore wind power plant foundations using a high-cycle accumulation model T. Wichtmann, A. Niemunis & Th. Triantafyllidis

635

Cyclic accumulation effects at foundations for offshore wind turbines H. Wienbroer, H. Zachert, G. Huber, P. Kudella & Th. Triantafyllidis

641

Study on soil-structure interaction of suction caisson by large-scale model tests B. Zhu, D.Q. Kong, L.G. Kong, R.P. Chen & Y.M. Chen

647

8 Jack-up units Simplified VH equations for foundation punch-through sand into clay J.-C. Ballard, P. Delvosal, P.H. Yonatan, A. Holeyman & S. Kay

655

Characterisation of undrained shear strength using statistical methods B. Bienen, M.J. Cassidy, M.F. Randolph & K.L. Teh

661

Centrifuge modelling of spudcan deep penetration in multi-layered soils M.S. Hossain, M.F. Randolph & Y.N. Saunier

667

A probabilistic approach to the prediction of spudcan penetration of jack-up units G.T. Houlsby

673

An assessment of jackup spudcan extraction O.A. Purwana, H. Krisdani, X.Y. Zheng, M. Quah & K.S. Foo

679

3D FE analysis of the installation process of spudcan foundations G. Qiu, S. Henke & J. Grabe

685

Undrained bearing capacity of deeply embedded foundations under general loading Y. Zhang, B. Bienen, M.J. Cassidy & S. Gourvenec

691

9 Anchoring systems Trajectory prediction for drag embedment anchors under out of plane loading C.P. Aubeny & C.-M. Chi

699

Setup following keying of plate anchors assessed through centrifuge tests in kaolin clay A.P. Blake, C.D. O’Loughlin & C. Gaudin

705

Seismically-induced displacements of a suction caisson in soft clay A.J. Brennan, S.P.G. Madabhushi & P. Cooper

711

SEPLA keying prediction method based on full-scale offshore tests R.P. Brown, P.C. Wong & J.M. Audibert

717

Set-up of suction piles in deepwater Gulf of Guinea clays J.-L. Colliat & D. Colliard

723

Centrifuge testing of suction piles in deepwater Nigeria clay – Effect of stiffeners and set-up time J.-L. Colliat, H. Dendani, H.P. Jostad, K.H. Andersen, L. Thorel, J. Garnier & G. Rault

729

Numerical FEM and laboratory study of the bearing capacity factor Nc for plate anchors L.N. Equihua-Anguiano, M. Orozco-Calderón, P. Foray & M. Boulon

735

Caisson capacity in clay: VHM resistance envelope – Part 2: VHM envelope equation and design procedures S. Kay & E. Palix Installation and in-place assessment of drag anchors in carbonate soil M.P. O’Neill, S.R. Neubecker & C.T. Erbrich © 2011 by Taylor & Francis Group, LLC

IX

741 747

Caisson capacity in clay: VHM resistance envelope – Part 1: 3D FEM numerical study E. Palix, T. Willems & S. Kay

753

Numerical investigation of the behaviour of suction caissons in structured clays S. Panayides, M. Rouainia & A. Osman

759

Cyclic moment loading of suction caissons in sand B. Zhu, B.W. Byrne & G.T. Houlsby

765

10 Pipelines and risers Multidirectional analysis of pipeline-soil interaction in clay R.G. Borges & J.R.M.S. Oliveira

773

Geotechnical challenges for deepwater pipeline design – SAFEBUCK JIP D.A.S. Bruton, M. Carr & F. Sinclair

779

Large deformation finite element analysis of vertical penetration of pipelines in seabed S. Chatterjee, M.F. Randolph, D.J. White & D. Wang

785

Implementation of geotechnical techniques in the analysis of pipeline response G. Cumming & N. Brown

791

Lateral soil resistance to an untrenched pipeline under the action of ocean currents F.P. Gao, S.M. Yan, E.Y. Zhang, Y.X. Wu & X. Jia

797

Vertical cyclic testing of model steel catenary riser at large scale T.E. Langford & V.M. Meyer

803

Kupe gas project pipeline – optimisation of discrete rock berm design shore approach B.L. Larsson

809

Model test studies on soil restraint to pipelines buried in sand R. Liu, S.W. Yan & J. Chu

815

Pipe-soil interaction on clay with a variable shear strength profile D.R. Morrow & M.F. Bransby

821

Sweeping behaviour of shallowly-embedded pipeline during cyclic lateral movement T. Takatani

827

Advanced nonlinear hysteretic seabed model for dynamic fatigue analysis of steel catenary risers I.H.Y. Ting, M. Kimiaei & M.F. Randolph

833

Mobilization distance in uplift resistance modeling of pipelines J. Wang, S.K. Haigh, N.I. Thusyanthan & S. Mesmar

839

Theoretical, numerical and field studies of offshore pipeline sleeper crossings Z.J. Westgate, M.F. Randolph, D.J. White & P. Brunning

845

Observations of pipe-soil response from the first deep water deployment of the SMARTPIPE® D.J. White, A.J. Hill, Z.J. Westgate & J.-C. Ballard

851

11 Trenching, ploughing, excavation and burial Influence of object geometry on penetration into the seabed A. Ivanovi´c, R.D. Neilson, G. Giuliani & M.F. Bransby

859

Investigation into the effect of forecutters on plough performance K.D. Lauder, M.J. Brown, M.F. Bransby & J. Pyrah

865

State-of-the-art jet trenching analysis in stiff clays J.B. Machin & P.A. Allan

871

Numerical modelling of soil around offshore pipeline plough shares W. Peng & M.F. Bransby

877


X

Anchor–chain–rockfill–soil interaction: Evolution of design methods H. van Lottum & H.J. Luger

883

Development of a jet trenching model in sand J.-F. Vanden Berghe, J. Pyrah, S. Gooding & H. Capart

889

12 Design and risk Structural factors affecting the system capacity of jacket pile foundations J.Y. Chen, R.B. Gilbert, J.D. Murff, A.G. Young & F.J. Puskar

897

The new API Recommended Practice for Geotechnical Engineering: RP 2GEO P. Jeanjean, P.G. Watson, H.J. Kolk & S. Lacasse

903

Comparison of ISO 19901-2 and API RP 2A seismic design criteria for a site in the Caspian Sea, Turkmenistan Z.A. Lubkowski, J.E. Alarcon & Z.A. Razak

909

Offshore geotechnics – safe and sustainable J. Peuchen & J. Haas

915

Author Index

921


XI



Preface

The International Symposia on Frontiers in Offshore Geotechnics (ISFOG), hosted by the Centre for Offshore Foundation Systems (COFS) at the University of Western Australia (UWA), provide a platform for academics and practitioners to discuss the emerging challenges in offshore geotechnical engineering and present recent practice and research. Offshore design and construction presents unique challenges to geotechnical engineers. Many of the challenges that are routinely encountered today have persisted for decades and continue to be gradually overcome with advances in technology and analysis. Meanwhile, new challenges are faced as the industry moves to harness resources in deeper waters, harsher environmental conditions and new frontier regions with previously-uncharacterized seabed conditions. The inaugural ISFOG was hosted by COFS in September 2005 and arose from discussions with colleagues from around the world which highlighted a consensus that there was a niche for a new specialist offshore geotechnical conference. That symposium was attended by 182 delegates from 22 countries and the proceedings contained 7 keynote papers and 127 peer-reviewed general papers that reflected the state-of-the-art of practice and research (Gourvenec and Cassidy, Eds, Taylor & Francis, ISBN 041539063X). Immediately afterwards and over the years following the symposium, we received extremely positive feedback from many of the participants, who had found the event to be both valuable and enjoyable. As a result, in the latter part of 2007 we canvassed the International Advisory Committee (IAC) of the inaugural ISFOG – a 30-strong team of leading academics and practitioners – for their views on a second ISFOG. They offered their unreserved support for a repeat event and it was agreed that a 5-year gap would be appropriate between symposia. The original Australian Research Council grant that led to the establishment of COFS expired in 2005. In some respects the first ISFOG was a chance to mark the achievements of COFS during its original tenure, at what might have been its zenith. However, COFS has grown since 2005 and is now more active and larger than ever. We therefore felt able to organise a 2nd International Symposium on Frontiers in Offshore Geotechnics, inviting friends and colleagues, old and new, to visit Perth again in November 2010. On advice from the IAC, we retained the single session format of the inaugural ISFOG, which kept participants together throughout the event, allowing continuity of discussions and avoiding clashes between concurrent presentations. Emphasis on the poster sessions and general reports offers exposure to the papers that inevitably do not receive full oral presentation given the constraints of the single session format. Key themes of the inaugural ISFOG are still as relevant 5 years on and therefore remain unchanged for the second ISFOG – these include geohazards, site investigation techniques and foundations for renewable energy, as well as anchoring solutions for deep water and pipeline geotechnics. Greater emphasis has been placed on in situ soil testing techniques and pore pressure measurement, and the characteristics of unusual seabed deposits found in frontier regions. The keynote papers reflect the key stages of an offshore project. The first three keynotes cover the assessment and interpretation of offshore geohazards, in situ site characterisation and pore pressure measurement, and the behaviour of West African clays – West Africa being a critical frontier region with particular challenges; two further keynotes describe state-of-the-art design considerations for piled foundations and pipelines; a further keynote describes recent advances in centrifuge modelling and the final keynote examines risk and reliability in offshore geotechnics. The keynotes alone contain many man-years of accumulated experience. We are sincerely grateful to the keynote authors, particularly those from industry who have generously found time to contribute their expertise to these comprehensive papers. The papers collected in these proceedings include the 7 keynotes and a further 117 peer-reviewed general papers that represent the current state-of-the-art in offshore geotechnics. These provide an invaluable resource to all those working in offshore construction, design and research. Each of the papers has been peer-reviewed by at least 2 reviewers, drawn from COFS and from academic institutions and industry within Australia and around the world. We are indebted to the reviewers for all their efforts, which have ensured that the papers are of a very high standard. We also thank Divya Mana and Santiram Chatterjee of COFS for their editorial assistance whilst assembling these proceedings. Stephanie Boroughs and Monica Mackman of COFS have provided administrative support throughout the organisation of this second


XIII

ISFOG, and we acknowledge in particular Stephanie’s tireless work corresponding with the authors and reviewers to compile the proceedings. The assistance of the Local Organising Committee is also acknowledged, in particular Noel Boylan of COFS who co-ordinated our sponsorship programme. Finally, COFS is grateful for the support for ISFOG provided by the following organisations and individuals: the Australian Department of Innovation, Industry, Science and Research through the International Science Linkages Programme, our industry sponsors, the members of our International Advisory Committee and the International Society for Soil Mechanics and Geotechnical Engineering (ISSMGE) under whose auspices the ISFOG symposia are held. Susan Gourvenec and David White July 2010


XIV


Committees

LOCAL ORGANIZING COMMITTEE Susan Gourvenec (Chair)

Centre for Offshore Foundation Systems (COFS), UWA Stephanie Boroughs (Sec.) COFS, UWA Monica Mackman (Sec.) COFS, UWA Britta Bienen COFS, UWA Nathalie Boukpeti COFS, UWA Noel Boylan COFS, UWA Mark Cassidy COFS, UWA Liang Cheng School of Civil and Resource Engineering (SCRE), UWA Fiona Chow Chair, Australian Geomechanics Society (AGS), WA Chapter James Doherty SCRE, UWA Sarah Elkhatib Arup Martin Fahey COFS, UWA Ian Finnie Advanced Geomechanics

Andy Fourie SCRE, UWA Christophe Gaudin COFS, UWA Andrew Grime Arup Jim Hengesh COFS, UWA Shazzad Hossain COFS, UWA Yuxia Hu SCRE, UWA Mehrdad Kimiaei COFS, UWA Barry Lehane SCRE, UWA Mark Randolph COFS, UWA Yinghui Tian COFS, UWA Dong Wang COFS, UWA Phil Watson Advanced Geomechanics David White COFS, UWA Long Yu COFS, UWA Hongxia Zhu COFS, UWA

INTERNATIONAL ADVISORY COMMITTEE Susan Gourvenec (Chair) David Airey Peter Allan Marcio Almeida Knut Andersen Charles Aubeny Fraser Bransby Nick Brown Byron Byrne Tim Carrington John Carter David Cathie Johnny Cheuk Chris Clayton Ed Clukey Jean-Louis Colliat Don DeGroot Jason DeJong Earl Doyle Carl Erbrich Trevor Evans Albert Griffith


Australia Australia UK Brazil Norway USA Australia Australia UK UK Australia Belgium Hong Kong UK USA France USA USA USA Australia UK Saudi Arabia

Andy Hill Phil Hogan Jacques Garnier Bob Gilbert Harry Kolk Richard Jardine Andy Lane Colin Leung Jayme Mello Roger Moore Don Murff Julian Osborne Andrew Palmer Alain Puech Matthew Quah Richard Raines Roderick Ruinen Marc Senders Dan Spikula Jørgen Steenfelt Tor Inge Tjelta Shuwang Yan

XV

UK USA France USA Netherlands UK Australia Singapore Brazil UK USA UK Singapore France Singapore USA Netherlands Australia USA Denmark Norway China



Reviewers

Tony Abbs Martin Achmus Hugo Acosta-Martinez David Airey Tom Aldridge Marcio Almeida Knut Andersen Senthil Arasu Charles Aubeny Jean-Christophe Ballard Edgard Barbosa-Cruz Britta Bienen David Bonjean Nathalie Boukpeti Noel Boylan Fraser Bransby Andrew Brennan David Bruton Byron Byrne Tim Carrington Mark Cassidy Santiram Chatterjee Liang Cheng Johnny Cheuk Fiona Chow Patrick Clancy Ed Clukey Richard Dean Andrew Deeks Jason DeJong James Doherty David Edwards Clarence Ehlers Sarah Elkhatib Ed Ellis Carl Erbrich Martin Fahey Ian Finnie Andy Fourie Christophe Gaudin


Guido Gottardi Susan Gourvenec Jürgen Grabe Andrew Grime Ole Hededal James Hengesh Andy Hill Matt Hodder Phil Hogan Shazzad Hossain Guy Houlsby Yuxia Hu Hans Hugel Mostafa Ismail Ana Ivanovic Christophe Jaeck Richard Jardine Eric Jas Hackmet Joer Steve Kay Lindita Kellezi Feizi Khankandi Yoshiaki Kikuchi Mehrdad Kimiaei Melissa Landon Thomas Langford Kok Kuen Lee Barry Lehane Colin Leung Nina Levy Tom Lunne Jon Machin Chris Martin Jayme Mello Richard Merifield Vaughan Meyer Neil Morgan Damian Morrow Don Murff Tejas Murthy

XVII

Andrzej Niemunis Steve Neubecker Tim Newson Elio Novello Conleth O’Loughlin Michael O’Neill Julian Osborne Ashraf Osman Andrew Palmer Jeff Priest Richard Raines Mark Randolph Mike Rattley Oliver Reul David Richards James Schneider Marc Senders Shambhu Sharma Tom Sheahan Dan Spikula Doug Stewart Kevin Stone Kar Lu Teh Luc Thorel Yinghui Tian Manh Tran Jean-Francois Vanden Berghe Alastair Walker Dong Wang Phil Watson Zack Westgate David White Nobutaka Yamamoto Jun Yang Gulin Yetginer Long Yu George Zhang Hongxia Zhu

1 Keynotes



A systematic approach to offshore engineering for multiple-project developments in geohazardous areas T.G. Evans BP Exploration Operating Company, Sunbury-on-Thames, UK

ABSTRACT: The development of economic and safe oil and gas facilities in geotechnically challenging and geohazard-prone areas calls for inter alia early engagement of geospecialists, fit-for-purpose data acquisition, geological understanding, engineering pragmatism and innovation and perhaps above all, multi-disciplinary teamwork. Jeanjean et al. (2005) described some of the strategies and practices developed by BP that embody these basic principles and this paper updates and expands these ideas and concepts with special reference to the investment sharing and planning opportunities afforded by programmes of multiple projects. 1

INTRODUCTION

centralised Geohazard Assessment Teams (GATs) and regional ground modelling. This paper develops the geohazard engineering risk management themes and methods described by previous authors, with particular reference to the opportunities for investment sharing and learning from BP’s multiple-project development programmes in deep water offshore Angola and Egypt. The paper is in two parts: Part 1 (Section 2) is a review of the some of the main engineering challenges faced by operators such as BP in geohazardous environments and Part 2 (Section 3) describes some of the ways that BP is tackling these challenges in Angola and Egypt as part of its asset management process.

In their keynote paper to ISFOG 2005, Jeanjean et al. (2005) described the challenges faced by oil and gas operators when developing offshore facilities in deep water frontier areas and environmentally and geologically difficult settings. The main theme of the paper was geotechnical and geological risks faced by projects operated by BP in geohazard-prone areas in the Caspian Sea, West Nile Delta (WND) and the Gulf of Mexico (GoM), with particular reference to the lessons learnt from the Mad Dog and Atlantis fields that are located along the Sigsbee Escarpment in GoM. The underlying message of the keynote was that geohazards do not necessarily preclude safe offshore developments provided operators are proactive and invest sufficient resources and planning in the geohazard management process. Some of the principles and best practices described in the paper include the use of multidisciplinary geospecialists, integrated and phased field-wide geophysical surveys and geotechnical surveys, development of new fit-for-purpose survey tools and use of project-specific engineering analyses to inform risk assessments. From about 2003 the focus of BP’s new deepwater projects shifted from the GoM towards the Caspian Sea, Angola and Egypt West Nile Delta (WND). Each of these areas has its own specific geotechnical and geohazard challenges that require some form of rigorous management. Significantly, BP’s developments in Angola and Egypt involve rolling programmes of projects that have provided systematic learning and investment leveraging opportunities in geohazard risk management, integrity management and value engineering not necessarily available for single projects. Evans et al. (2007) and Moore et al. (2007) introduced some of the multiple-project investment concepts that have been applied in the WND, including the use of © 2011 by Taylor & Francis Group, LLC

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OFFSHORE GEOTECHNICAL ENGINEERING CHALLENGES

2.1 The geotechnical maze The ever-increasing geotechnical challenges faced by offshore operators in pursuit of hydrocarbons in frontier areas have been widely reported in the past decade: Power and Clayton (2003), Kvalstad (2007), Jeanjean et al. (2003, 2005), Evans et al. (2007), Evans et al. (2010) and Hadley et al. (2008) and many others. However, as discussed by Evans et al (2007), offshore exploration and development policy is rarely, if ever, dictated by shallow subsurface and geotechnical considerations. Rather, the shallow geo-specialists are expected to work as part of a bigger team to: (1) ensure that shallow-subsurface issues do not compromise project-life safety and function (loss prevention) and (2) offer cost-effective geotechnical solutions. These two goals are particularly onerous in frontier areas where the challenges for geotechnical and foundation engineering are especially labyrinthine, as illustrated on Figure 1. The chaotic nature of this figure shows

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Figure 1. Geotechnical challenges.

how offshore geotechnical engineering challenges can arise from the interactions of one or more negative factors that can result in increased costs and risks to exploration, development and production. For the purpose of this paper the geotechnical challenges are assumed to have eight root causes:

seabed-founded structures. Offshore developments, in water depths of 500 m to 1500 m, are now relatively commonplace and a number of projects are underway in over 2000 m of water. Within five years it is expected that development will extend beyond 3000 m of water. The costs of geotechnical operations in such water depths are high, especially in remote and harsh conditions. Consequently, despite the obvious technical advantages of early geotechnical data for supporting exploration and development, it is often difficult to justify spending money on such activities until there has been some exploration success. Engineers therefore rarely get the early geotechnical insight that is actually needed for frontier developments where there are no ‘offset’ data.

1 2 3 4 5 6 7 8

Water Depth Harsh environments Remoteness Geological Complexity Development Scale Data Acquisition Constraints Structural Complexity Inexperience and Uncertainty Each of these root causes is discussed in the remaining parts of Section 2.

2.3

In an unpublished review of literature on submarine geologies and environments Brunsden (2010) confirmed the complex morphological, geomorphological and geological legacies which geo-specialists and engineers need to understand and resolve to support offshore exploration and development. The main purpose of this review was to summarise conventional wisdom on landslide controls and processes. The main findings of the review were:

2.2 Water depth, environment and remoteness – (Root Causes 1 to 3) The first three root causes listed relate to access difficulties; geographical, subsea and subsurface. The logistical problems of working in frontier areas away from established centres of industry and support infrastructure are clear and will have a significant affect on the efficiency and cost of even routine offshore operations. Deep water and rough weather can restrict access further and cause additional problems for geophysical and geotechnical data acquisition and for the installation of foundations, anchors and other © 2011 by Taylor & Francis Group, LLC

Complex geology (Root cause 4)

– Offshore depositional centres are largely determined by plate tectonics and they are subject to repeated sedimentation and landslide processes.

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Table 1.

Offshore geological and geotechnical controls.

Origin

Type

Cause

Effect/Factor

Natural

Inherited quasistatic conditions

Past geological & environmental processes

Preconditioning – structural relief/seabed ruggedness – slope geomorphology & geomorphology – sediment fabric and structure – sediment stress-strain history – sediment mechanical properties

Inherited transient Equilibration of conditions Aggravating processes arising from past geological & environmental processes or events Future processes Ongoing or new slow Aggravating geological and environmental processes Future episodic Random short-duration Triggering events natural phenomena Man-made Future processes or events

Episodic or progressive offshore activities

Triggering

– fluid flow – swelling – strain softening – physiochemical changes – sedimentation at slope crest – slope toe erosion – uplift from diapirism – earthquakes – storm waves and currents – tsunamis Project-life operations: Exploration, e.g. – exploration well drilling – rig anchoring – geotechnical investigation Construction/Installation, e.g. – development well drilling and anchoring – installation of gravity structures – pipe laying – pile driving Hydrocarbon Production, e.g. – operational loadings on facilities – fluid and gas injection – fluid withdrawal and subsidence

complex interactions between inherited characteristics and future processes, events or activities that may precondition, aggravate or trigger geohazard incidents that threaten projects. A good understanding of the geological past and project-life future is important since damaging geohazard events, such as submarine slides, may arise during the lifetime of a project due to ongoing geological processes alone, or in combination with shorter duration triggers. Geological history also has an important influence on the physical properties of marine sediments. Chandler (2000) and Cotecchia and Chandler (2000) describe how the mechanical behaviour of a clay is controlled by the ‘soil structure’ that it develops during deposition solely as a result of one-dimensional consolidation and by natural post-depositional geological processes such as mechanical unloading, creep, diagenesis and tectonic shearing. The following subsections discuss some of the sedimentological and post-sedimentological processes that can affect the physical behaviour of marine clays in areas such as Egypt and Angola. Soil Structure. A challenge for the offshore geotechnical engineer is to understand the physical characteristics of shallow marine sediments and how these may affect their behaviour when disturbed by natural events or by exploration and development activities. Marine sediments of most interest to engineers

– Terrestrial, hemipelagic, contourite and shelf sediment sources provide soils at variable rates and in relatively definable packages. – Structural controls are determined by plate tectonic history, fabric, structural relief, sedimentary and basin architectures, surface morphology and residual accumulated strains are the results of a chequered inheritance. – Mass movements are much larger than onshore but the seabed often exhibits extraordinary freshness that may reflect long periods of inactivity between episodic events and/or our inability to resolve slow sedimentological or geomorphological changes with current geophysical imaging technology. Leroueil (2001) describes how geological and geomorphological inheritances and future aggravating or triggering factors are the major controls for terrestrial and submarine landslides. Locat and Lee (2000) and Locat (2001) present similar ideas in connection with instabilities along ocean margins which can be equally applied to geohazards in general. The main natural and man-made controls on seabed instability and other geohazards in the context of offshore exploration and development are summarised on Table 1. A significant challenge for those responsible for planning and engineering offshore developments is to unravel the © 2011 by Taylor & Francis Group, LLC

Examples

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in deep water are predominantly clays and Burland (1990), Chandler (2000) and Cotecchia and Chandler (2000) describe how such soils have significant geotechnical inheritances that condition their response under mechanical loading. Burland (1990) used the inherent compressibility and strength characteristics of reconstituted clays, called intrinsic properties, as a reference for interpreting the in situ state and physical characteristics of inorganic natural non-carbonate clays. He performed one dimensional compression tests on clay slurries and by plotting a normalising parameter called the Void Index , Iv, against the vertical effective stress σv , derived an Intrinsic Compression Line (ICL) as a datum for the in situ stress states for natural clays. Iv is defined as: Figure 2. Normally consolidated clays with sedimentation structures – after Cotecchia and Chandler (2000).

deposited slowly in still water will have more random open structures with Iv values that may lie above the SCL, Point’C’. Highly flocculated clays deposited in seawater with higher than average salt concentrations and siliceous and carbonate oozes would also be expected to plot above the SCL. When natural clays are loaded by foundations in the field, or in one dimensional compression tests (oedometer tests) in the laboratory, the rate of loading is sufficiently high to disrupt the inter-particle bonding and fabric of the clays. The corresponding compression lines are relatively steep compared to the natural SCL and fall towards the ICL, also shown on Figure 2. It is therefore particularly important to detect and characterise clays with natural states that plot above the SCL since they are especially brittle and compressible and therefore more susceptible to progressive failure and possibly liquefaction under external loading. Post-sedimentation Effects. All naturally sedimented clays are altered to some degree by environmental, geological and physio-chemical processes, or other ageing effects. As described previously, many large-scale post-sedimentation changes can occur in submarine environments and these can be expected to alter the stress states and mechanical properties of sediments significantly. However, even in the most benign environments, soils will experience postsedimentation changes due to one or more of the following processes:

In this expression e0 is current in situ voids ratio, e∗100 is the voids ratio of a reconstituted sample of the same soil compressed one dimensionally to a vertical effective stress, σv , of 100 kPa and C∗c is the compression index of the reconstituted soil measured between σv of 100 kPa and 1000 kPa. The asterisk denotes an intrinsic property of a reconstituted soil prepared from a slurry mixed at a water contents of between 1.25 and 1.5 times the liquid limit. According to Chandler (2000), Iv is a useful normalising parameter for different soil types because C∗c is defined uniquely by the one-dimensional compression procedure, the soil mineralogy and the pore water chemistry. Burland also established an Iv compression line, for marine clays in their natural state, called the Sedimentation Compression Line (SCL). The distance between the SCL and ICL in Iv-σv space is a measure of the sensitivity and brittleness of the natural clay. The SCL represents the average condition for natural marine clays with sensitivities of between about 2 and 9 and soils that plot on the idealised SCL have sensitivities of about 5 (Cotecchia and Chandler 2000). Sedimentation Structure. Mitchell (1976) defined soil ‘structure’ as a state that reflects the arrangement of soil particles, or soil fabric, and interparticle bonding. As shown on Figure 2, the ICL and SCL provide a simple and convenient framework in for comparing the in situ stress states or ‘structures’ and the mechanical behaviours of natural and reconstituted clays (Chandler 2000). The structure that clay has inherited during sedimentation depends on its mineralogy and the depositional conditions and it is not easily changed by further burial. The natural stress states of most marine clays lie on or around the SCL, Point ‘A’ in Figure 2. Clays that are deposited relatively rapidly from dense suspensions in high energy environments such as hyperpycnal flows, tend to have more orientated compact structures with Iv’s closer to the ICL, Point ‘B’. Debrites, contourites and turbidites may also have these characteristics. Conversely, hemipelagic clays that are © 2011 by Taylor & Francis Group, LLC

– Ageing – Creep./Secondary Consolidation – Unloading and swelling (Mechanical Overconsolidation) – Diagenesis – Physio-chemical changes including bonding and cementation, – Biological activity (Ehlers et al. 2005, Kuo et al. 2010) Such soils will exhibit gross yield, a threshold stress state beyond which soil stiffness and strength fall significantly (Hight et al. 1992, Cotecchia and Chandler 2000). Mechanical behaviour pre- and post-yield will be fundamentally different, so the identification of

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Figure 4. Normally consolidated and mechanically over-consolidated clays with diagenetically altered structures – after Cotecchia and Chandler (2000).

Figure 3. ‘Aged’ normally consolidated and mechanically overconsolidated clays with sedimentation structures – after Cotecchia and Chandler (2000).

and slopes and (2) the interpretation of insitu tests that are used to infer these mechanical properties (Randolph et al. 2005). In addition soils with natural stress states close to or above the SCL and/or with significant post-sedimentation structures that do not fit well in to normalised stress history frameworks such as SHANSEP that require consolidation beyond gross yield (Ladd & Foot 1974). Burland (1990) indicates that the peak strength and brittleness of such soils are underestimated using the SHANSEP test procedure. The detection, identification and characterisation of atypically or unusually sensitive soils are therefore important in offshore geotechnics. The natural soil structure framework described by researchers like Burland (1990), Cotecchia and Chandler (2000) and De Gennnaro et al. (2005) can be used with other soil frameworks such as Critical State Soil Mechanics to predict the mechanical behaviour of natural soils affected, for example, by ageing, desiccation, ice-loading, erosion and mass wasting. However, BP and other offshore operators are increasingly encountering post-sedimentation processes that are more difficult to characterise using traditional methods and which may require different, possibly bespoke, soil mechanics approaches, as suggested by Gens (2010). Four examples of effects that have posed additional challenges during BP’s exploration and development activities are:

gross yield stress states in shear and compression is also important for modelling soils correctly in offshore foundation design and geohazard assessments. Aged normally consolidated clays and clays that are over-consolidated due to mechanical unloading but which are unaffected by diagenesis or biological activity, would be expected to yield close to their natural sedimentation compression lines; the SCL for most natural marine clays. The oedometer com pression lines and vertical yield stresses, σvy , for an aged normally consolidated clay and a mechanically over-consolidated clay are shown schematically on Figure 3. The vertical yield stress for an over consolidated soil, σvy , will be coincident with the . geological preconsolidation stress of the soil, σvc The ratio of geological preconsolidation stress and /σv is traditionthe in situ vertical effective stress σvc ally termed the over-consolidation ratio (OCR). The ratio of the yield stress and in situ vertical effective stress σvy / σv is the called the yield stress ratio, YSR. Conversely, as shown on Figure 4, many normally consolidated clays will have been affected by diagenesis and will have diagenetically altered postsedimentation structures that will result in yield to the right of the SCL. Some over-consolidated soils that have retained diagenetic characteristics will also yield to the right of the SCL above the preconsolidation pres sure, σvc . In this case the YSR is greater than the OCR. Oedometer compression curves for normally consolidated and over-consolidated clays with diagenetically altered structures are shown schematically on Figure 4. The natural in situ states of soils condition their responses to mechanical loading and disturbance and are therefore important geotechnical controls. Grossyield in compression and shear and the post-yield brittleness are important characteristics in offshore geotechnical engineering since they affect: (1) in situ operational shear strengths and stiffnesses of soils, and consequently the behaviour of foundations, anchors © 2011 by Taylor & Francis Group, LLC

– – – –

Shallow Gas Gas Hydrates Salt Diapirism Hydrocarbon Migration Products

Shallow Gas. Gas-charged sediments, largely of biogenic (bacterial activity) and petrogenic (thermally altered) origins, are commonplace offshore. The presence of gas, in the form of free bubbles, dissolved gas or gas hydrates can change the shear strength and compressibility characteristics of sediments under static and cyclic loading. Unlike unsaturated soils, gassy

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soils encountered offshore contain large amount of dissolved gas that can exsolve. The degree of saturation of offshore gassy sediments is generally above 90% and free gas is generally in the form of discrete bubbles much larger than the pore spaces between individual soil particles (Sultan 2009). Shallow gas per se is a ubiquitous drilling hazard and shallow gas hazard assessments are obligatory for hydrocarbon exploration, development and production operations in most parts of the world. For example, in the GoM, the requirements for assessing shallow gas are specified by MMS (1998) and in the UK sector North Sea they are covered by guidelines prepared by UKOOA (1997). Gas also has important influences on the mechanical properties of sediments. These are less well understood but have potentially negative effects on offshore wells and structures too. Sands are especially vulnerable since the difference in pore water pressure and pore gas pressure is small, so high gas pressures equate to low effective stresses and reduced strength, as well as increased compressibility. However, the effects of gas on the physical properties of clays with low free gas concentrations, typically encountered offshore, have been the subject of limited research and are consequently less predictable. Research by Wheeler (1988) showed that the effects of gas on the mass undrained shear strength of clay depend on the initial effective and total stresses and the volume fraction of gas voids, and that these effects may be detrimental or beneficial. For example, Wheeler’s model suggests that the undrained shear strengths of gassy soils in triaxial compression at shallow depths in deep water, which have high initial total stresses and low initial effective stresses, may be less than the comparable strength of the saturated clay. Conversely, gassy clays with low initial total stresses and high initial effective stresses such as deep sediments in shallow water are estimated to be stronger than the saturated clay matrix. Both effects are predicted to increase with increasing gas saturation. However, Wheeler’s research was limited and it is clear that the mass strength of gassy clay depends, amongst other things, on the degree of soil de-structuring caused by the formation of free gas (Hight & Leroueil 2003, Sultan et al. 2009), and the direction and sense of loading. Wheeler’s soil model may provide useful upper and lower bound modifying factors for the effects of gas on undrained shear strength but the industry’s capabilities of inferring the operational properties of gassy soils from laboratory and in situ tests and for modelling their behaviour in routine geotechnical engineering are still very limited. Gas Hydrates. Gas hydrates are crystalline solids, physically resembling ice, that form when gas molecules are encaged by hydrogen-bonded water molecules. The main factors affecting hydrate formation and stability are temperature, pore pressure, gas chemistry and pore-water salinity and they can occur in the pore spaces of marine sediments under the right conditions.A change in any of these parameters such as a reduced pressure or an increased temperature could © 2011 by Taylor & Francis Group, LLC

Figure 5. Bottom simulating reflector (BSR).

affect the system equilibrium and reduce the stability of hydrates, resulting in decomposition, dissociation and dissolution, and the release of free gas and water (Sultan 2007). Many gases of low molecular weights can form gas hydrates but in marine conditions the gas is generally methane of biogenic origin. Marine gas hydrates are therefore common in organic-rich sediments at depocentres, especially where deposition has been rapid. They are therefore of relevance to most offshore hydrocarbon developments in relatively deep water where gas has infiltrated shallow sediments. Figure 5 is a seismic section from an offshore location which shows high acoustic amplitude traces below a bottom simulating reflector (BSR) that is indicative of possible gas-charged soils below a stable zone of gas hydrate-bearing sediments. A joint industry project (JIP) carried out on behalf of West African offshore operators (Sultan 2007), concluded that cementing action of gas hydrates could inhibit the normal compaction processes, resulting in sediments with more open structures and higher voids indices than would otherwise be the case. Hydratebearing sediments would therefore be expected to have higher yield stresses, elastic moduli and peak strengths, and to be more brittle, than comparable hydrate-free sediments. Sultan (2007) describes the potential physical collapse of such soils and the development of excess pore pressures, leading to loss of strength, increased compressibility and possible hydraulic fracture. The threats posed by gas hydrates to offshore exploration and production are still uncertain and speculative and views on this subject are sometimes controversial. Although problems have been reported (Nimblett et al. 2005), it is generally agreed that well drilling in hydrate-bearing sediments is possible with good planning and close control of well bore pressures and temperatures. However, experience of hydrocarbon production in hydrate-prone areas is limited so there is more uncertainty about interactions arising from longer-term offshore operations, for example raised temperatures around hot production wells. Therefore, for the present, the industry tends to avoid producing in areas where there is direct evidence or strong indirect evidence of hydrate-bearing sediments.

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Figure 6. Salt dome.

soil overlying salt. These observations were based on the results of preliminary large strain numerical analyses performed by Imperial College/GCG (2007, unpublished) and reflect conditions often encountered offshore in regions such as Angola. Further generic and project-specific research work is clearly needed to support developments in areas where salt effects are unavoidable. Hydrocarbon Migration Products. Hill et al. (2010a, 2010b) describe areas of persistent hydrocarbon seepage in deep water Angola where the shallow predominantly clayey sediments contain beds/layers with anomalously high acoustic impedances, Figure 8a. These seismically reflective zones often have a surface expression and locally extrude above the seabed, as shown on Figures 8b and 8c. Investigations have shown that these atypical soils comprise sequences of hard carbonate-rich claystone beds with variable amounts of heavy hydrocarbons that have the consistency of thick oil and asphalt (bitumen) at room

Salt Diapirism. Hill et al. (2010a, 2010b) describe how salt diapirism in deep water Angola has uplifted older sediments which has resulted in strong soils close to the seabed that can cause installation problems for suction-installed foundations and anchors and increased seabed slopes that can lead to pipeline routing difficulties and instabilities. The seismic section and schematic shown in Figure 6 indicate how the overburden can be severely distorted by salt movement resulting in large changes to the depositional stress states of cover sediments. These changes may affect the behaviour of foundations such as suction-installed caissons that are designed using empirical methods developed from field tests in areas well away from such influences, or from centrifuge or physical models that replicate these far-field conditions. This point is illustrated in Figure 7, which summarises some of the observations made about the possible combined effects of uplift and subsequent erosion on a normally consolidated © 2011 by Taylor & Francis Group, LLC

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Figure 7. Effects of salt uplift and surface erosion.

temperature when recovered in geotechnical samples, Figure 8d. Sediments that have been modified by hydrocarbon seepage over geological timescales are, unsurprisingly, relatively common in areas of interest for oil and gas exploration and development. These soils are inevitably highly variable and are difficult to characterise geotechnically. Hard beds arising from authigenic mineral cementation can clearly pose problems for foundation and anchor installations that will probably warrant location-specific geotechnical investigations and bespoke engineering solutions. Sediments contaminated with hydrocarbons may pose even greater engineering challenges and arguably open up a new branch of soil mechanics that requires unconventional engineering thinking and new geotechnical solutions, as suggested by Randolph et al. (2005) and Gens (2010). 2.4

Figure 8. Hydrocarbon migration products. © 2011 by Taylor & Francis Group, LLC

Development scale (Root Cause 5)

Offshore oil and gas developments range from discrete bottom-founded shallow water structures with relatively small seabed footprints at locations that are generally fixed early in the development cycle, to floating or subsea-tieback developments in deepwater that cover large areas of the seafloor and which are constantly evolving. For example, the total footprint area of the three subsea developments currently planned by BP in the WND amounts to about 1400 km2 . The footprint of a typical BP Angolan project is about 750 km2 . Scale is therefore a major challenge for deepwater geospecialists who are required to define the geotechnical conditions and geohazards for a large number of structures spread across the seabed with sufficient flexibility to guide, or accommodate, inevitable changes in scheme layout. For logistical and economic reasons, it may not be possible or practical to acquire geotechnical data at every structure location and it is often necessary to infer local design conditions from regional geotechnical data; a methodology described in some construction industry standards such as Eurocode 7 (2004). Eurocode 7 classifies structures in to three geotechnical risk categories, according to structural

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parameter selection and geotechnical design reliability assessments.

complexity, the ground conditions, type of load, and the consequences of failure. Small and relatively simple low-risk structures constructed on uniform ground are designated Category 1. Large or conventional structures which are supported on unexceptional ground and have well defined loads and normal risks fall in to Category 2. Structures that are unusually large or complex, or which are associated with abnormal loads, ground conditions or risks are classified as Category 3. Eurocode 7 advice is that it may be acceptable to use regional (offset) geotechnical data for the detailed design of Category 1 structures but that location-specific data are needed for Categories 2 and 3. Many offshore structures fall in to Categories 2 and 3 but for the reasons alluded to earlier it may be overly constraining or uneconomic to apply Eurocode 7 advice rigidly in deep water. However, the concept may be extended by improving the reliability of the regional database through informed ground modeling and applying risk-based design to manage the residual geotechnical uncertainties. Probabilistic-based geotechnical design methods using offset geotechnical data have been discussed and presented by various authors including Gilbert and Gambino (1999), Clukey et al (2000), Jeanjean et al. (2005) and Griffiths et al. (2009). These methods generally use normal or log normal soil property distributions that are best suited for areas with relatively uniform geology and soils that have similar sedimentation and post-sedimentation histories. They are less suitable for use in areas with complex geologies or numerous geohazards. The approach may also be unconservative in uniform conditions if mechanical behaviour of the soil mass is dominated by local weak or brittle seams, layers or zones that have gone undetected. The SHANSEP normalization procedure is often used to infer the undrained shear strengths of clays of similar origins and compositions over large areas. The method is proposed in ISO 19901-4 (2003) to achieve consistent shear strength profiles and has been used successfully for geotechnical engineering and geohazard assessments in Norway (Kvalstad 2007) and in the Gulf of Mexico (Jeanjean et al. 2005). Burland (1990) and Chandler (2000) indicate that the procedure may be reliable for relatively insensitive clays that lie below the SCL. However, for the reasons described earlier and discussed by Le et al. (2008), they propose that the procedure may be unreliable for more structured clays. Jardine et al. (2005) also suggest that the procedure may be unconservative for plastic clays that develop brittle shear bands that suppress dilation, Reconsolidation to the in situ effective stresses as performed in CAU tests results in less sample compression than SHANSEP. However, this technique may also cause some ‘destructuring’ of some highly structured clays resulting in underestimates of undrained shear strengths and sensitivities. It is therefore important to recognize the significance of soil structure when preparing or ‘conditioning’ laboratory test samples and to consider factoring this into geotechnical © 2011 by Taylor & Francis Group, LLC

2.5

Data acquisition constraints (Root Cause 6)

There are no hard and fast accepted strategies for geophysical and geotechnical data acquisition in deep water. Rather acquisition plans are invariably a compromise between budget and technical requirements, and are often constrained by an industry-wide shortage of survey vessels, equipment and know-how to carry out the work. Data Acquisition Costs. Shallow geological and geotechnical characterization of the shallow subsurface, including geohazards, is generally achieved by a combination of geophysical surveying and intrusive geotechnical investigation. The precise scope of work is defined by the area, line density and resolution of the geophysical survey and by the numbers and locations of in situ tests and samples. A range of acquisition strategies is possible as illustrated on Table 2. One extreme, Strategy 1, would be to spend very little on location-specific data until the scheme layout has been finalised and to rely on regional or offset data for planning and design during the early stages of the project. The other extreme would be to carry out fully calibrated area-wide high density ultra high resolution surveys at the beginning of a project as described for Strategy 3. The delayed approach is most likely to be suitable for single projects in mature areas with uniform well understood geology and few geohazards, whereas multiple projects in geologically complex frontier areas with numerous geohazards may justify the significant front-end investment implied by Strategy 3. A phased approach somewhere between these two extremes is generally adequate for most deep water projects. The amount of money spent on shallow subsurface investigations should ideally reflect positive and negative risks, i.e. the opportunities or benefits of investing the right amount wisely and the threats from not spending enough, or spending it wastefully. The UK Institution of Civil Engineers UK indicate that the largest element of technical and financial risk on construction projects is normally in the ground (ICE 1991), and that the costs overruns and delays arising from inadequate geotechnical investigations can effect 18 to 50% of construction projects. For underground structures with significant soil interfaces such as tunnels, up to 85% of projects could be affected (USNCTT 1984). The amount spent on geotechnical investigations for onshore construction projects in the UK is typically between 1% and 5% of the capital cost but geotechnical cost overruns can be up to 50% of the project value (AGS 2003). There are few comparable statistics for offshore projects but industry surveys have shown that offshore geotechnical risks may be underestimated and that associated installation problems arise frequently in delays and cost overruns (Clayton & Power 2002). There is also little published about the costs of offshore

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Table 2.

Data acquisition strategies.

Mature Areas/ Less Geohazards ⇔ Frontier Areas/ More Geohazards Increasing Front-end Uncertainty ⇔ Increasing Front-end Cost Purpose

Strategy 1

Strategy 2

For Concept Selection

Use exploration seismic data, Medium-density grid of 2D AUV & offset data & regional 2D UHR survey lines calibrated knowledge with geotechnical cores, boreholes & insitu tests

For Preliminary Medium to high-density design (FEED) AUV & UHR surveys with selective geotechnical calibration For Detailed Targeted geotechnical survey Design when development fixed

Strategy 3 High-density 2D AUV & 3D UHR blanket survey calibrated with geotechnical cores, boreholes & insitu tests

Additional selective ‘infill’ AUV & UHR surveys and geotechnical calibration

Possibly no additional acquisition

Selective location-specific geotechnical investigations

Selective location-specific geotechnical investigations

Table 3. The cost of data acquisition – Influencing Factors. 0.1% Capex ⇐

⇒ 1.5% Capex

– Shallow water – Mature areas – Simple geology – Few geohazards – Uniform soil conditions – Compact developments

– Deep and ultra-deep water – Frontier areas – Complex geology – Many geohazards – Highly variable soils – Laterally extensive developments – New unproven technologies

– Routine proven offshore technologies – Limited well understood – Many poorly understood soil-structure interactions soil-structure interactions – Financial risk only – Societal, Environmental and Financial Risks – 1st Party risks only – 3rd and 1st Party Risks Figure 9. Optimum spend on data acquisition.

10% loss would be 10−4 . The probabilities of a 10% loss for intermediate spends would be between 10−4 and 1.0. No attempt has been made to discount the costs in Figure 9 and consequential losses such as reduced production are not considered. Available Resources. The offshore geotechnical industry’s capabilities and resources for supporting the relatively rapid advance of hydrocarbon exploration and development in to deep water are sorely stretched. One of the main constraints for deepwater operators is the limited availability of shallow geophysical survey and geotechnical investigation vessels that can operate in these conditions and the dominance of one or two main contractors in these markets. The development of survey class autonomous underwater vehicles (AUV) (Bingham et al. 2002), has significantly improved the offshore industry’s capabilities for acquiring engineering quality ultra high resolution data (bathymetry, sidescan sonar and subbottom profiling) in deep water. However, at present only three companies offer AUV surveying services on a commercial basis in water depths greater than

geophysical and geotechnical surveys in the oil and gas industry. From BP’s experience these may range from as little as 0.1% of the capital cost for simple fixed shallow water platforms in uniform predictable soils, to 1.5% for large deep water projects in geotechnically difficult areas. Although there are no strict rules about the level of spend on site characterization, it is possible to identify those factors that may influence it, as shown on Table 3. The optimum spend on a project would be that which minimises the overall geotechnical cost, namely the front-end cost of the geophysical and geotechnical investigations plus the risked cost of geotechnical construction problems. A purely hypothetical example for a large deepwater project in a geologically complex area is shown on Figure 9. The optimal spend on investigations for this illustrative case is 1.5% of capital cost (Capex). This is based on the simple premise that: (1) if no investigations are performed the equivalent financial loss would be 10% of Capex and this would occur with certainty (probability of 1.0) during the project life and (2) at the other extreme, if 10% of Capex was spent on investigations the probability of a © 2011 by Taylor & Francis Group, LLC

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about 500 m. Geotechnical investigation in deep water is still a significant technical challenge, time consuming and expensive, and requires significant investment by contractors. Only one contractor, Fugro, currently provides a full range of geotechnical services in deep water. The world’s fleet of deep water geotechnical vessels is also severely limited with Fugro operating the only three dedicated geotechnical drillships capable of working in water depths over about 1500 m. This shortage can mean that some deepwater basins may not have access to such vessels for years; a situation that is likely to worsen as exploration and production moves in to even deeper water and the technical, operational and safety challenges with conventional geotechnical drillships increase and acquisition costs rise. The market capability for geotechnical laboratory testing, engineering and reporting is slightly better but the numbers of companies offering independently accredited laboratories is still very small and the pool of deepwater geotechnical engineering specialists is tiny compared to that available to clients in the onshore construction business. For operators like BP, the lack of competition and resources reduces programming flexibility and often leads to geotechnical acquisition as a critical path activity. Sole-source generally means additional costs to the operator and in busy periods, poses extra burdens on the sole-provider with greater risk of log -jams and delays in post-field activities such as laboratory testing and reporting. The use of consortia as an alternative to a ‘one-stop’ service can work well but involves the management of a number of interfaces that can pose extra contractual, technical and HSSE pressures on operators’geotechnical staff. In the long term the deepwater oil and gas business would clearly benefit from the emergence of new geotechnical site investigation contractors and the use of quicker and remotely-operated investigative tools and equipment. However, these would require significant investments by contractors that ultimately can only be justified by market forces and a sustainable industry workload. Nevertheless, there has been encouraging progress in recent years; for instance through the use of Fugro’s SMARTPIPE® and SMARTSURF™ Systems (Hill & Jacob 2008; Denis & De Brier 2010), and breakthroughs made by Benthic Geotech Pty with their multi-functional seabed drilling system, PROD (Tjelta 2010). These developments have been achieved with the encouragement and support of one or more operators. Guidelines and Standards. Currently, there is a lack of international guidelines and standards for the performance of marine geophysical and geotechnical investigations, including vessel and equipment audits and best practices for data acquisition, laboratory testing and reporting. As a consequence operators often produce diverse job specifications that may have led to global inconsistencies and inefficiency, especially in the geotechnical investigation contracting business. The lack of best-practice industry guidelines has also hampered the integration of the two branches of © 2011 by Taylor & Francis Group, LLC

Figure 10. Subsea structures.

geoscience that underpin deep water site characterization; geophysics and geotechnics. Advice on offshore site investigation practices has been published by some national bodies Norsok (2004), or specialist groups such as SUT (2003 & 2004) and ISSMGE (2005), but these efforts are largely uncoordinated and are generally focused on shallow water. This industry standards deficiency is currently being addressed by the joint API RG7/ISO and SC7/WG10 committee for offshore foundations with the aim of publishing two new standards; ISO 19901-8-1, ‘Marine Soil Investigations’ and ISO 19901-8-2 ‘Geophysical Site Investigations within the next few years (Jeanjean et al. 2010). Ultimately, it is hoped that a third standard will be published that will provide guidance and encouragement for a multidisciplinary approach to shallow geohazard risk assessments and engineering for offshore developments. 2.6

Structural complexity (Root Cause 7)

Offshore Facilities. Randolph et al. (2005) provide a comprehensive review of geotechnical design practices for offshore facilities and how these have diverged from onshore practices because of the nature of marine soils, and the sizes, loads and performance characteristics of offshore foundations. As explained by Randolph et al. exploration and development in frontier regions has meant engineering in soils that are significantly different to those encountered onshore and which require new geotechnical design approaches. BP and others have reported many cases of unusual soil conditions, including extremely high plasticity structured clays in Angola (Gennaro et al. 2005; Le et al. 2010; Hill et al. 2010b), highly saline clays in the Caspian Sea (Kay et al. 2005), extremely dense sands in Norway (Alm et al. 2004) and exceptionally hard clays West of Shetlands (Evans et al. 2010). Structure Types. Figure 10 shows some typical subsea structures, ranging from relatively light surfacelaid in-field flowlines to large manifolds weighing

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several hundred tonnes. The foundations for the heavy structures can be much larger than those built onshore. They are also required to resist much larger lateral loads and overturning moments and are often subjected to combined loads in all six degrees of freedom, including significant torsion. Large Strain Behaviour. Operationally subsea structures can generally tolerate relatively large displacements, although settlement may be a controlling factor for heavy structures such as manifolds. Many seabed interactions also involve repeated loads that can result in large cumulative strains and failures. A good understanding of large strain soil behaviour under cyclic and monotonic loading and the potential for progressive failure is therefore important for design reliability. In addition, the ability to model such behaviour accurately increases the potential for cost savings through the use of performance-based design methods that are described later. Low Stress Behaviour. Conversely, knowledge of the compressibilities and shear strengths of soils at very low stresses is also important for deep sea projects. For example, the operational normal total stresses applied by unburied in-field flowlines are generally in the range of 1 to 10 kPa, which is significantly below the range considered in onshore and shallow water engineering. The study of ‘low stress soil mechanics’ problems is difficult using conventional laboratory soils testing machines and often requires bespoke equipment to produce reliable results (Bruton et al. 2007). Uniqueness. Some deepwater structures have interactions with soil that are distinctive by having few onshore analogues (Evans et al. 2007). Examples of behaviour that are difficult to predict and continually pose engineering challenges include: (1) lateral buckling, walking and vibration of surface-laid pipelines that transport multiphase products under high pressures and temperatures, often on undulating seabeds (Figure 11a), (2) the random motions of compliant steel catenary risers (SCRs) that dangle on the seabed (Figure 11b) and flow-induced vibrations of bottom-founded flexible spools (Figure 11c). Design Codes. The background to present-day offshore foundation design practices is discussed by Jeanjean et al. (2010). Foundation design procedures in offshore codes and recommended practices are still largely based on the traditional bearing capacity methods that have been developed for onshore foundations; ISO 19904-1 (2003), API RP2A, (2000) and DNV (1992). These methods generally give safe designs for relatively simple foundations under fully undrained or fully drained conditions but provide little advice about designs under partially drained conditions or with torsional loads, and do not cover structures with more complex load-displacement characteristics such as described earlier. New Methods. The development of new reliable user-friendly geotechnical solutions for offshore engineering is proving difficult. According to Randolph et al. (2005), there are only a few theoretically exact © 2011 by Taylor & Francis Group, LLC

Figure 11. Soil-structure interactions.

plastic solutions and limit equilibrium methods are generally unconservative. Numerical methods based on finite elements have been shown to be useful for improving fundamental understanding of soil structure interactions but are difficult to adapt for routine design and, according to Randolph et al. (2005), may also be unconservative in some cases. ISO 19901-4 recommends the use of large scale load tests, model tests and field instrumentation to reduce residual uncertainties with foundation behaviour. Geotechnical engineering in geohazard-prone areas is potentially even more challenging due to more complex geology and greater natural soil variability that increases the ‘aleatory’ design uncertainties, Jeanjean et al. (2005) and possible interactions arising from active geological processes such as submarine slides. The goal for offshore operators such as BP is to develop reliable, cost-effective and robust, preferably codified, design methods that can be used with confidence by all their design and installation contractors. However, at the present time some design procedures are still under development, often through proprietary research, with very little field validation. The theoretical and empirical bases and expertise behind these methods is often exclusive to academic institutions or specialists contractors and some of the design

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Table 4. Industry and International Guidelines – Coverage of Foundation Design and Geohazards – after Jeanjean (2010).

solutions are not readily assimilated into standard structural computer codes. Well Geotechnics. The commentaries above relate to offshore facilities but years of experience suggest that the geotechnical engineering of wells also needs serious attention. The top section of a well, or tophole, is drilled through soils and poorly lithified weak rocks. The depth of the tophole is typically between 500 m and 1500 m and it is drilled without pressure containment. Geotechnical problems are often encountered within well topholes during well conductor installations and drilling. These include wellbore instability and hydraulic fracture and drilling fluid loss in to the shallow formation. Drilling difficulties in the shallow section cause costly delays and may require expensive remedial measures. Uncontrolled shallow well drilling can also affect structures in proximity, as discussed in ISO19901-4 (2003), ISO 19902 (2007), HSE (1997) and by Hobbs and Senner (1998). Tophole problems can arise from a combination of difficult subsurface conditions and poor drilling practices. Despite many years of experience they still occur (Schroeder 2007), and the situation will only improve through increased collaboration between the offshore well drilling and geotechnical communities. Some loss-prevention initiatives have already been taken in BP with the geotechnical and drilling groups jointly developing internal best practices for tophole drilling and completions. At industry level, the Offshore Soil Investigation Forum (OSIF) and the SUT’s Offshore Site Investigation Group (OSIG) are currently preparing a guidance document on this subject.

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API RP2A ISO 19901-4 ISO 19902 API RP2T API RP2SK ISO 19901-7 ISO 19904-1 ISO19905-1

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36p (16%) 34p (100%) 42p (6.5%) 6p (4.2%) 38p (21%) 10p (8%) 0.1p (0.05%) 46p (16%)

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validated. However, it is likely that these will still be limited to relatively simple structure-soil combinations. The lack of industry guidelines on geotechnical engineering in geohazardous areas is even starker. This is illustrated qualitatively on Table 4, Jeanjean (2010, unpublished), which shows that current API Practices and ISOs give little textual coverage to geohazards. This is clearly something that needs to be redressed but it is unlikely that any future code or standard will be able to give anything but general advice for engineering in extremely variable conditions. Engineering in such areas will therefore continue to call for geotechnical rigour. Contractual Responsibilities. The responsibility for geotechnical and geohazard risks should ideally be with those that are best able to manage them. At the present time, operators are probably in the best positions to manage geohazards since they employ the necessary specialists and have more time to investigate them and assess their impact. Similarly, experienced offshore design and installation contractors are better placed to manage geotechnical risks. However, the responsibility for geotechnical risks is not always clear, even in so-called non-geohazardous areas. For example, unproven, often bespoke, designs naturally carry greater than normal construction and operational risks for operators and their contractors. These risks increase for EPC contractors when they are required to adopt or adapt research-based design methods that are the intellectual properties of operators or niche specialists. In the short term, these additional risks are likely to be reflected in bid prices and in the numbers of contractual exclusions and claims but in time it is hoped that the new methods will be tested and calibrated by field monitoring and will evolve in to more reliable and routine solutions that attract less uncertainty and risks. In the meantime, the challenge for offshore operators, their specialist advisors and their contractors is to find ways of working together to manage the transition from evolving best practices to established design methods, while protecting the interests of all the stakeholders.

2.7 Inexperience and uncertainty (Root Cause 8) One of the biggest issues currently faced by offshore industry in general is working in ever deeper water in new geographical areas and with new technology. The geotechnical discipline is no different and is on a fast-track learning programme that is most evident at the delivery end of projects, namely in engineering procurement and installation. As described earlier, the ground conditions are a major source of uncertainty on all construction projects and are a common cause of delays and cost overruns. Although, great advances have been made in offshore construction there are still a lot of challenges associated with the engineering and installation subsea structures in deep water, particularly in geohazard-prone areas. Codes and Guidelines for Geohazards and Foundations. The relative lack of engineering experience is also evident by the shortage of guidelines, standards and recommended practices for the design of deep water structures. Jeanjean et al. (2005) suggested that the paucity of formal guidelines reflects the newness and uniqueness of many deepwater structures and the natural time lag required for field validation of new design practices before code acceptance. The implication is that best practices and new code-based procedures will emerge, as methods mature and are © 2011 by Taylor & Francis Group, LLC

Standard

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As the offshore industry moves in to areas with increasing geological complexities and uncertainties the roles and responsibilities for geotechnical engineering and geotechnical risks need to be clear and equitable. In this respect, the industry may benefit by adapting the risk management practices developed in the underground construction industry where projects invariably have significant geotechnical exposure. The tunneling industry has over many decades developed best practices that centre on the production of Geotechnical Baseline Reports (GBRs) that define the interpreted or assumed reference geotechnical conditions and risks at the beginning of the contract (Essex 2007).The main purpose of GBRs is that they provide a set of geotechnical conditions against which the actual conditions encountered can be assessed and compared and therefore provide clarity on risk sharing. 2.8

Communication of geotechnical risks and costs

Figure 12. Benchmarking risk severity and manageability.

For the reasons discussed above, shallow subsurface risks in deepwater frontier areas, and therefore the time and costs of investigating and mitigating them, are often much greater than for traditional shallow water projects. Possibly the biggest challenge of all for shallow geospecialists is therefore to convey clear matter-of-fact messages about these new challenges to non-specialists who may lack this insight. The ultimate objective would be to provide timely fit-forpurpose geotechnical guidance that would allow the project team to understand and manage the geotechnical elements of the development efficiently and systematically, ideally as non-critical activities. Table 1 summarises the main natural and manmade controls over shallow subsurface conditions for offshore oil and gas developments. The main hazards are natural variable and/or weak soils and dormant or active geological features and processes, and the field activities that may aggravate these conditions. These subsurface conditions may pose significant threats to the integrity and safety of offshore developments during and after installation and construction. Equally, there are opportunities and strategies by which these threats can be managed and mitigated. It is important that all these negative and positive factors are communicated and discussed with the various stakeholders in the right ways and at the right times, so that they are interpreted correctly for inclusion in risk registers and are used effectively in budget planning, design and construction. Communication of geotechnical and geological risks (threats and opportunities) can be a major asset to the project when done properly but if done poorly, or at the wrong time, may be misleading and counterproductive. At BP advice on the shallow subsurface is generally tailored to meet the needs of six distinct project stages: Access: Geologic appraisal and prospect evaluation Appraise: Project feasibility Select: Choose the preferred project(s) Define: Develop the preferred project and fund Execute: Detailed design and build Operate: Production

Access. When considering acquisitions, investors and financial planners are most concerned in asset ‘prospectivity’ and value and the shallow subsurface has little influence on investment decisions. However, interest increases once access has been gained, especially with respect to the risks and costs associated with drilling exploration wells. The production of shortterm shallow drilling hazard risk assessments is the major activity for shallow geospecialists at this time. However, longer-term risks may also need to assessed even at this very early stage should exploration or appraisal wells be considered for use in production, ie so-called ‘Keeper’ wells. Appraise. Following exploration success the information of immediate use to project planners and managers is an assessment of the main threats to development, benchmarked against other projects, and the levels of investment that may be needed to manage these threats. A simple way of benchmarking levels of risk and complexity/manageability is illustrated on Figure 12 (HSE 2006). This figure also provides guidance on specific methodologies might be needed to assess risk, including Qualitative (Q), Semiquantitative and Quantified risk assessment (QRA) methods. Early Appraise is also the stage at which the data acquisition and risk assessment strategy should be discussed and agreed with project funders. Ultimately, at the end of the Appraise stage the shallow geo-specialists need to assure the project that there are no showstoppers and that there are unlikely to be major geotechnical surprises; in effect a ‘declaration of geotechnical feasibility’. Select and Define. During the Select and Define stages the main role of geotechnical engineers and the rest of the shallow subsurface team, is to build partnerships with project planners and facilities and well engineers to help develop economic and reliable geotechnical designs and pragmatic risk reduction solutions. Risk mitigation can be achieved by avoidance through optimised layout planning or


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Execute. Offshore engineering and construction are usually, although not always, carried out through lump sum EPC contracts. Whatever, the type of contract it is important that roles and responsibilities for geotechnical engineering and the associated risks are fully understood by the operators and main contractors. Good geotechnical communication is needed between the operator and the main contractors and Geotechnical Baseline Reports supported by fit-for-purpose factual Geotechnical Data Reports (Essex 2007) may be the key to this; especially if the main contractors’ views can be incorporated in the GBRs before contract award. Operate. The geotechnical conditions encountered during construction are often different to those assumed beforehand. It is important that these differences and any design changes associated with them are recorded and included in an updated as-built geotechnical report that could be used as a reference for future field expansion and maintenance, and for decommissioning. The preparation of as-built geotechnical reports is not an industry norm at present but would appear to be a natural extension of current best practices.

3

FACING THE CHALLENGES

3.1 Evolving practices The number of oil and gas facilities that have been built by BP and other operators in geohazardous and geotechnically difficult areas has increased significantly over the past 15 years. During this time BP has developed strategies and methods to maintain project reliability and cost-effectiveness in such environments. Many of these practices are basic engineering common sense while others have evolved to meet new challenges and opportunities. In their ISFOG 2005 keynote, Jeanjean et al. (2005) shared some of the best practices and methodologies that BP had developed for engineering in geohazardprone areas over the previous decade, with special reference to pioneering projects in the GoM. The authors championed the use of integrated multidisciplinary teams of geospecialists and phased geophysical and geotechnical surveys and supported the development of new enabling technology for these surveys. Amongst other things, they described (1) the use of shallow seismic data and the SHANSEP normalisation technique for inferring the undrained shear strengths of soils over large areas, (2) deterministic and probabilistic methods for assessing seabed stability, (3) model testing to calibrate non-code based geotechnical designs and (4) reliability-based geotechnical designs. The authors also recommended a consistent and systematic approach to geotechnical and geohazard risk assessment so that the geo-risks could be ranked alongside other project risks. No clear step changes have emerged in geotechnical and geohazard risk management in the past five

Figure 13. Risk assessment methods.

by geohazard-resistant design and economies can be made by exploring value engineering options such as risk based design, performance based design and observational engineering. Ideally, the relative costbenefits of the geohazard-optimised options would be assessed using probability based techniques that account for increased Capex and reduced Opex that may accrue compared to the ‘do nothing’ case. The level of residual risks should be tracked continually during this period and assessed using methodologies that are compatible with those used by the project for assessing risks from other sources. These may include Qualitative (Q), Semi-quantitative methods (SQ), in which frequency and severity/impact are quantified approximately within ranges and Quantified risk assessment (QRA) in which both are fully quantified. Examples of Q, SQ and QRA are shown in Figure 13. The objective at the end of Select is a technically feasible geohazard-tolerant field layout. The goal at the end of Define is an optimized reference scheme that is suitable for taking forward into Execute. © 2011 by Taylor & Francis Group, LLC

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years and shallow subsurface conditions cotinue to pose difficulties for offshore exploration and development and many projects are still dealing with complex poorly defined conditions with no prescriptive solutions. However, BP’s experiences continue to show that an effective way of managing the risks and costs is through clear-ownership and teamwork, and through the early and continuous engagement of geospecialists with the BP project teams and with the contractors who plan, engineer and install offshore wells and facilities. In about 2004 prospects for multiple sequential projects emerged in Angola and Egypt that allowed BP to learn from GoM experiences and to take advantage of some important longer-term investment and learning opportunities for geotechnical engineering and geohazard risk management in these two areas. The next sections develop some of the geotechnical engineering best-practice themes described by Jeanjean et al. (2005) and describe how these have evolved and been adapted to meet new circumstances, lessons learnt and fresh challenges. Particular reference is made to the use of dedicated regional geoteams, regional ground modeling and possible methodologies for engineering facilities. 3.2

Figure 14. Centralised geotechnical and geohazard teams.

Figure 15. Shallow geospecialists – Project-life support.

Shallow geotechnical and geohazard assessment teams

Consultancy Group (GCG), Imperial College, URS Corp, Fugro William Lettis & Associates and AtkinsBoreas. The GAT approach is particularly cost-effective for multiple-project programmes by providing critical mass and know-how and the means to capture and transfer lessons learnt and develop standard design and risk assessment practices and engineering solutions. The co-location of GATs from different regions provides additional advantages in the form of economies of scale, cost-sharing, technology transfer and consistent and standardised corporate engineering practices.

Evans et al. (2007) describe how the industry-wide shortage of offshore shallow geohazard specialists, especially in engineering contracting, had motivated BP Exploration Technology Group (EPT) to set up dedicated multidisciplinary shallow geotechnical and geohazard assessment teams, GATs, to support a rolling programmes of offshore projects in Egypt and Angola. The GATs for these two regions are co-located with each other and close to their respective planning and engineering teams, as illustrated on Figure 14. The rationale behind this concept is that a GAT has the critical mass and competencies to provide sustained specialist support that is tailored to the project’s whole-life needs and activity levels as shown on Figure 15. The discipline bias of the GAT is geophysics and geology at the start of a project, gradually changing to geotechnical engineering as work progresses in to development planning and engineering. A key benefit of co-located GATs is that they encourage continuous dialogue between the geospecialists and BP’s project team and contractors so that they can manage the shallow subsurface engineering risks and costs together. BP’s Egypt and Angola GATs were established in 2004 and are presently supporting projects at various stages of development from Appraise to Execute. The teams are formed from a central core of shallow geophysicists, geologists, geomorphologists, sedimentologists and geotechnical engineers from within BP and from resident consultants Halcrow Group Limited and Fugro GeoConsulting. Additional external support is provided by other consultants and contractors, including the Norwegian Geotechnical Institute, D’Appolonia, Senergy, Geotechnical © 2011 by Taylor & Francis Group, LLC

3.3

Geotechnical and geohazard (G&G) mitigation strategy

The shallow geotechnical and geohazard (G&G) mitigation strategies BP is adopting for its rolling programmes of projects in Angola and Egypt are similar and in each case was developed early in the programme cycle following a desk-top geotechnical review or ‘Health Check’. The desk studies were performed using in-house or published regional and local geophysical and geotechnical (G&G) data and hydrocarbon exploration seismic quality geophysical data. Desk studies were the important first steps in the geotechnical and geohazard risk management process since they provided immediate insights in to the levels of risk, the resources and investment likely to be needed to manage these risks, and the best strategy for acquiring data to help this process. The results of the desk study were used to develop a fit-for-purpose geotechnical and geohazard risk management plan to investigate, characterize and mitigate geohazards and to provide geotechnical engineering support to the

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Figure 17. Geophysical data acquisition equipment.

3 if necessary, to carry out supplementary geotechnical investigations to support detailed design and construction when the scheme layouts and architectures are defined better. Figure 16. Geotechnical and geohazard mitigation process.

The prospect of multiple projects allowed the Egypt and Angola Business Units to take a far-sighted view of surveys and make longer term arrangements with suppliers, as a hedge against the shortage of worldwide survey vessels. Integrated Surveys. As described earlier, deep water subsea tie-back schemes like those being developed by BP in the WND and others based on floating systems similar to those being built in Angola cover 100’s km2 of the seabed and huge volumes of soil. For cases like these, it is clearly not feasible or practical to investigate every seam, layer or stratum of soil or geological feature within the development footprints. Spatial interpolation is unavoidable and needs to be carried out systematically by experienced geospecialists to ensure the main geotechnical objectives are achieved at least cost. The integrated survey and ground model approaches favoured by BP in Angola and Egypt help this optimisation process by providing an initial regional view with gradual focus on areas or soils that ultimately have the greatest influence on project safety and costs. The geophysical data acquisition technologies used for imaging the seabed and shallow subsurface to provide the skeletal ground models in Egypt and Angola are shown in Figure 17, which is reproduced from Moore et al. (2007). The geotechnical acquisition priorities for a ground model strategy are to provide data for calibration of the shallow geophysics and characterisation of geohazards and to get representative geotechnical data for designing and installing wells and facilities. An important feature of BP’s optimisation strategy in Angola and Egypt is to achieve these goals with least effort and at lowest cost, by maximising the multi-functional site investigation activities, as shown on the Venn diagram in Figure 18. BP’s deepwater geotechnical investigation methods include the use of sampling and in situ testing in seabed and drilling modes. In the seabed mode BP frequently investigates the soils at a single location using a combination of sampling and testing tools, including

projects. The Geotechnical and Geohazard Mitigation (GGM) plans were based on the overall project plans, the inferred complexity of the shallow subsurface conditions and the perceived challenges these pose to developments. The prospect of multiple projects in both Egypt and Angola had a significant influence on the planning strategy and was instrumental in the decision to set up regional GATs and to adopt the regional ground model methodology. The key steps in the GGM process are shown in Figure 16. They comprise an integrated sequence of Data Acquisition, Ground Modeling, Risk Assessment and Layout Planning and Engineering designed to help the projects develop geohazard-tolerant and geotechnically robust reference schemes that can be taken forward in to EPC contracts for detailed design and construction. 3.4 Data acquisition Phasing. Although the actual data acquisition strategy and scope will depend on the nature and timing of the offshore development, it is good practice to take a staged approach. This involves phases of geohazard and geotechnical engineering quality geophysical surveying followed by geotechnical investigations to calibrate the geophysics and to provide geotechnical data for engineering. BP adopted acquisition campaigns based around Strategies 2 and 3 in Table 2 for subsea developments offshore Egypt and subsea and floating developments offshore Angola. The aim in each case was to acquire sufficient data to build up a development-wide predictive ground model. The main steps in this process were: 1 To carry out front-end development wide 2D ultra high resolution and Autonomous Underwater Vehicle (AUV) surveys, 2 to perform geotechnical reconnaissance investigations to fully calibrate the geophysics for preliminary design, and © 2011 by Taylor & Francis Group, LLC

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Figure 18. Optimised geotechnical data acquisition.

Figure 19. Geotechnical investigations in seabed mode.

box cores, gravity cores, gravity piston cores, ‘T’bar testing and piezocone penetration tests (PCPTs) as illustrated on Figure 19. This combination, or ‘Set’, of tools is generally sufficient to fully characterise the shallow subsurface conditions for assessing shallow geohazards and for designing flowlines, pipelines, shallow foundations and anchors, including suction installed caissons. However, in recent years BP has also actively supported the development of other seabed in situ testing tools such as Fugro’s SMARTPIPE® and SMARTSURF™ systems (Figures 20a and 20b) to improve the understanding of soil conditions and pipe-soil interactions at or very close to the mudline (Denis & De Brier © 2011 by Taylor & Francis Group, LLC

Figure 20. New equipment.

2010). BP is also considering Benthic Geotech’s remotely drilling system Prod shown in Figure 20c, that has already been used by other operators such as Woodside, Total and Statoil (Tjelta 2010).

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Figure 21. Ground model.

One innovation favoured by BP in Egypt is the acquisition of 3-m to 20-m long gravity piston cores exclusively for geological, sedimentological and ichnological logging and geochronological testing, as described by Thomas et al. (2010). The continuous geological cores were taken alongside geotechnical cores of similar length and proved to be a sound investment. The core logging significantly improved the WND GAT’s understanding of the types, frequencies and magnitude of historical submarine slides, and helped them optimise the geotechnical testing programmes for engineering shallow foundations and assessing the potential of future shallow slide activity. A key advantage of continuous core logging is that it can help to detect discontinuities and pre-existing shear surfaces, and possibly identify soils that may be especially susceptible to mechanical disturbance. Boreholes were used selectively in Angola and Egypt to perform in situ tests, and to provide samples for geological logging and geotechnical testing. The borehole data were used for calibrating regional seismostratigraphic models, exploring deeper geohazards, inferring in situ pore pressures and providing geotechnical data for engineering well topholes and deep foundations such as piles. The downhole work included high resolution geophysical logging, piezocone and temperature cone testing and piezorobe testing.

subsurface conditions anywhere within it, (2) to use the interpreted data for geohazard risk assessment, field layout planning and engineering at most locations and (3) limit site-specific geotechnical investigations to areas or structures that pose the biggest risks or offer the best value engineering opportunities to the project. The integrated geological modelling is discussed in ISO 19901-4 and the advantages of the approach are largely self evident. However, the success of the method is highly dependent on having sufficient data of the right quality to populate the model, a competent interpretation team and purposeful deliverables that are communicated clearly with end-users. The challenge is particularly onerous in geologically and geotechnically complex areas but the multidisciplinary GAT and phased data acquisition concepts described earlier have proven to be well suited to the task. The ground modelling methodology adopted by BP is illustrated on Figure 21. The procedure involves the integration of environmental, geophysical and geotechnical data and the detailed mapping, interpretation and calibration of these data to develop a seismostratigraphic-soil model of the area of interest, and to capture this in GIS-format for interfacing with the project team and its engineering contractors. The main outputs from the regional model are shown on Figure 21 and in an idealised flow diagram on Figure 22. These may be summarised as:

3.5

– Terrain Units. These are developed from the geomorphological interpretation of Multibeam Echo Sounder, Side Scan Sonar, 2D UHR and AUV geophysical data. Each Terrain Unit has a distinctive geological setting and depositional history with similar geological forms and processes. – Geohazards. Terrain units that are associated with past geohazard events such as submarine slides, and therefore require more detailed assessments, are generally distinguished from those that have

Ground modelling

The predictive ground model approach that BP is adopting for its deep water projects offshore Egypt and Angola is centered on the creation of a 3-D block model that captures the geomorphology, main stratigraphic units, geological features, geohazards and representative geotechnical conditions across the development footprint. The main objectives of the method are to: (1) build a model that is reliable enough to infer the shallow © 2011 by Taylor & Francis Group, LLC

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Figure 22. Ground model outputs for geohazard risk assessment and engineering.

sediment types, one layered and the other more disturbed, are shown Figure 23(b). The hemipelagites (Set 1) have high Iv values that are at, or above, the SCL and they are relatively compressible. Conversely, Set 2, interpreted from geological evidence as mass flow deposits, have Iv’s close to the ICL and they are less compressible in the stress ranges of interest for most offshore foundations. These findings are consistent with the inferred geological stress histories. However, a second group of soils, Set 3, also interpreted as mass flow sediments, have higher Iv values closer to the SCL. This suggests that these soils may have retained some of their sedimentation structures during deformation. Alternatively the facies interpretation may be imprecise and may be need to be reconsidered and refined. Figures 24 shows similar plots for recently deposited shallow sediments from two regions in deepwater Angola, Regions 1 and 2. The data envelope shown in Figure 24(a) is for oedometer tests performed on samples of hemipelagic soils from Region 1. These soils are interpreted to be unaffected by geological processes such as salt diapirism and fluid expulsion and the test data broadly straddle the SCL. Figure 24(b) shows the results of oedometer tests on four samples of shallow soils from Region 2, in an area influenced by salt and erosion. These results suggest that the effects of uplift and erosion may include ‘destructuring’ and overconsolidation as implied in Figure 7. The results of the oedometer tests in Figures 23 and 24 clearly show the potential insights in to the stress histories and mechanical properties of marine soils that may be gained from using frameworks such as proposed by Chandler (2000). These frameworks offer the means to calibrate terrain units

more benign origins. These geohazard terrain units are subject to further detailed sedimentological, ichnological and geochronological assessment and geomechanical back analyses to estimate the frequency and magnitude of these historical events (Thomas et al. 2010). The results of this work are fed in to the future geohazard risk assessment. – Soil Provinces. Soil Provinces are linked closely to Terrain Units and are idealised areas of the seabed and shallow subsurface that have broadly similar geotechnical characteristics and properties to depths of interest for engineering. Each Terrain Unit may comprise one or more Soil Provinces and some Soil Provinces may be common to a number of Terrain Units. – Soil Units. These are the main soil layers and strata within each Soil Province that are interpreted to have broadly similar mechanical properties. Some Soil Units are common to more than one Soil Province and they are interpreted by dividing the ‘geophysical’ column into significant seismostratigraphic units separated by regional unconformities and defining the sediments that make up these units as one or more distinct soil facies. The soil facies represent different depositional and post-depositional processes and histories and they are interpreted from sedimentological, lithological, ichnological and chronological logging and analyses and geotechnical data obtained by selective sampling and in situ testing. As described earlier, it can be instructive to consider the results of incremental oedometer tests on some of the interpreted soil facies in void index – vertical effective stress space. Figure 23(a) shows envelopes of oedometer results from tests performed on WND normally consolidated hemipelagic clays and others inferred to be mass flow deposits (debrites, mudflows and slides). The contrasting soil fabrics for these two © 2011 by Taylor & Francis Group, LLC

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Figure 23. One-dimensional consolidation tests – WND Soils.

Figure 24. One-dimensional consolidation tests – Angolan soils.

and soil units and may provide the geologicalgeomechanical link that is required for a truly integrated approach to geotechnical engineering, particularly in geologically complex areas. – Geotechnical Modifiers. Some of the Soil Units may have local depositional fabrics or structures or have undergone local post-depositional changes © 2011 by Taylor & Francis Group, LLC

similar to those discussed earlier in this paper. These features and characteristics may not necessarily be resolvable by geophysics but nevertheless may have a significant effect on soil behaviour. Examples include sediments with high void indices and soils affected by cementation, hydrocarbon

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contamination, bioturbation and pre-existing shear surfaces – Regional Soil Parameters. Each of the main Soil Units is assigned representative mass soil parameters, such as undrained shear strengths and unit weights, which are inferred from the results of the geotechnical calibration programmes. These regional soil parameters are defined as credible ranges to reflect the levels of uncertainty with the data interpretation. The SHANSEP technique was not used to derive regional soil parameters in Angola and Egypt since the soils of greatest interest for engineering were considered to be too brittle or too variable or complex to apply the method reliably over large areas. However, the technique has been used selectively to infer operational undrained shear strengths for specific geotechnical designs or geohazard assessments where there are abrupt changes in OCR/YSR. – Operational Geotechnical Parameters.The geotechnical parameters used for engineering wells and facilities and carrying out geohazard assessments are necessarily problem-specific since they need to account for: (1) the mass properties of the soils, including the effects of geotechnical modifiers, (2) the type of structure or geohazard, (3) the imposed loads, and (4) the methods of analysis or design (code-based, reliability-based etc). For most of BP’s work in Egypt and Angola the regional soil parameters derived from the ground model databases have been used as the reference data for deriving operational design parameters. Location-specific data are used where local conditions are significantly different to the regional model and/or when geotechnical conditions may have an important bearing on safety or cost. Operational geotechnical parameters are key input to Geotechnical Baseline Reports used as reference for detailed design and construction contracts and as such, should be selected with support from the project engineering teams. The use of regional geotechnical data as the basis for design generally leads to higher factors of safety to achieve that same level of reliability as obtained using location-specific data. However, the strategy is pragmatic in largely uniform marine conditions and is broadly in line with one of the design approaches described in Eurocode 7 (2004). It also provides the flexibility to progress the geotechnical planning and design productively despite inevitable scheme changes. Nevertheless, significant effort is needed to identify atypical conditions that may control the mechanical behaviour of soil and invalidate the regional model. 3.6

Scheme definition – geohazard screening and risk assessment, and scheme layout planning

The geohazard management strategy being adopted in Angola and Egypt may be summarised as: – Assess the zones of influence of potentially active geohazards and where possible avoid them. © 2011 by Taylor & Francis Group, LLC

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– Where, for overriding reasons, ‘geohazardous’ areas are unavoidable, assess the risks in terms of probability of occurrence and impact and consider special geohazard-resistant engineering measures to reduce the risks to tolerable levels. Geohazard Screening and Risk Assessment. The first step in this process is to screen the development area for geohazard ‘hotspots’ that denote major risk drivers.The method involves considering potential event scenarios associated each of the geohazards in the ground model database and eliminating those that are not significant because they are very unlikely to occur during the project life, would have little impact on the development or could be accommodated by routine design. The geohazard scenarios considered in this study include natural incidents similar to those in the past, reactivated geohazards, completely new events and man-made effects. The magnitudes and frequencies of the historical events, such as submarine slides, were assessed by multidisciplinary expert consensus opinion informed by the ground model terrain evaluation work, as illustrated on Figure 25. This process was supported by geomechanical back analysis and forecasting using simplified inhouse methods that are suitable for screening large areas. For example, the potential for shallow submarine slides was investigated using an undrained sliding block model (Newmark, 1965) that has been adapted to account for the effects of residual excess pore pressures, earthquake loading and soil brittleness. Similarly, the return periods for delayed failures of scarps (Potts et al. 1997, Leroueil 2001), formed in previous slides were estimated by a simple pore pressure equilibration model. Regional geotechnical parameters were generally used in the screening process but the local effects of exceptionally weak or brittle layers, or pre-existing shear surfaces (geotechnical modifiers) were also considered. Complementary engineering studies were performed to assess the potential impacts of geohazards such as submarine slides on typical seabed infrastructure (Parker et al. 2008, 2009). The risk assessment process is being facilitated by a series interactive multidisciplinary workshop sessions and the results are integrated in to the overall project risk assessment using the Qualitative, Semi-quantitative and Quantitative Risk Assessment methods shown on Figure 13. Scheme Layout Planning. The results of the geohazard screening and risk assessment are used with other criteria to optimise the scheme layouts. This is a multidisciplinary task that depends on a number of factors such as reservoir access, flow assurance and cost, as well as geohazard threats. The ideal objective for the GAT is to avoid all the buffer zones associated with the major geohazards. However, compromises are inevitable, and for technical or economic reasons it may be necessary to encroach some of the buffer zones. In such cases more detailed location- and structure-specific assessments are performed to define the geohazard risks more accurately.

Figure 25. Assessment of frequencies and magnitudes of submarine slides – Fully-developed slab slides.

Specialised geohazard-resistant engineering measures are also being considered to reduce the risks in these areas. Optimised Reference Scheme. The layout planning is an iterative process performed throughout the Appraise to Define stages of a project with the ultimate objective of developing a geohazard-tolerant reference scheme that can be taken forward in to detailed engineering and construction, generally performed through an EPC contract. 3.7

solutions, often derived from first principles, are being sought for complex structures for which there is little or no experience. Geohazardous Areas. Similar design approaches are also being applied in geohazardous areas, although additional special geohazard-resistant design measures are also considered. Although sometimes difficult to achieve, the overall objective is to develop safe and functional standard engineering solutions in both non-geohazardous and geohazardous areas and to limit the use of bespoke designs. Code-based Designs. As discussed earlier industry and international guidance for the design of shallow and deep offshore foundations has improved considerably in recent years and is beginning to address the complex behaviour of many subsea structures (Jeanjean et al. 2010). However, codes still provide solutions for relatively simple structures designed to avoid failure, with less consideration for serviceability criteria such as deflections. BP’s normal practice is to use published codes such as ISO 19901-4 (2003), ISO 19902 (2007) and API RP2A (2000) where possible but to consider alternative methods should these standard methods not be applicable, or if they are known to be excessively conservative. Value-engineering Solutions. Value engineering is a process by which design function and/or design reliability are increased at no additional cost, or cost

Engineering

The design strategy being followed by BP in their West Nile Delta and Angolan deep-water programmes is shown on Figure 26. The methods are based on the nature of the seabed and shallow subsurface conditions and the types of structures, and are driven by the scheme layout and architecture. Non-geohazardous Areas. In areas designated as non-geohazardous, simple structures with relatively predictable performance are being designed using codified methods and where such methods do not exist or are uneconomic or unsuitable, more pragmatic value engineering solutions are being adopted. The use of centrifuge tests to calibrate the design of suction anchors in layered soils for the GoM Mad Dog project described by Jeanjean et al. (2005) is an example of such a pragmatic design approach. New design © 2011 by Taylor & Francis Group, LLC

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Figure 26. Design strategies for non-geohazardous and geohazardous areas.

longer term it is important to validate and/or refine these methods through field monitoring. The idealised flow diagrams in Figures 27 and 28 illustrate just two possible options for developing and calibrating these new designs methods. Figure 27 shows an approach that is based on the observational method. In this case a structuresoil interaction behavioural model is developed and used with project-specific soil and structures data to produce a job-specific prototype design to meet the project’s performance objectives. Provision is then made to monitor the structural response during field operations. If observed behaviour matches expectations the design may be verified. Otherwise, preplanned intervention and remedial measures may have to be implemented to assure acceptable life-of-field performance. Data from the field monitoring is also used to update and improve the behavioural model for future designs. Monitoring of flowline buckling and walking in Angola and the design and assurance process applied for the West of Shetland Clair Phase 1 Platform piles (Evans et al. 2010), are examples of the use of the observational method in BP. The method shown in Figure 28 is often used in onshore geotechnics to confirm design assumptions at sites or areas where there is little or no previous experience and is one of the ways recommended in ISO 19901-4 to manage geotechnical uncertainty Typically, it involves carrying out a field trial in advance of construction contract to validate or calibrate design assumptions for a specific site or types of soil and to update these accordingly. An example would be preliminary trials to confirm the design methodologies for predicting the axial capacities and load-deflection responses of piles which would then be used in projectspecific designs. It would be very rare for a single operator to do this in a frontier deep water area for a single project because of the high cost. However, the idea is more appealing when multiple projects are planned and trials can be performed opportunistically

is reduced without affecting function and reliability. Keaton and Eckhoff (1990) describe the approach in geologic hazard risk management terms as eliminating or reducing those aspects of a system that add cost without reducing risk. In BP the application of value engineering includes such things as challenging conventional thinking, doing things differently, design innovation, lateral thinking, adapting existing practices, extending the performance envelope and risk-based design with rigorous validation. In recent years BP has spent considerable effort investigating and developing cost-effective and safe engineering solutions based on these principles. Installation is a major cost driver and there is a big incentive to reduce the size and weight of subsea foundations and anchors. BP is tackling this challenge by developing reliability- and function-based (performance-based) designs that require the project to set performance objectives, such as failure probability, system displacements and tolerable levels of damage. These methods may also involve performance monitoring to manage the residual uncertainties, as practised in the observational engineering approach (Eurocode 7 2004). Other cost-saving initiatives presently being pursued include the development of composite foundations, the adaptation of wellconductor jetting techniques for installing piles and the use of numerical and physical modeling to challenge conventional design practices such as the use beamcolumn analyses with API RP2A p-y springs (Jeanjean 2009). New Design Methods. Many subsea structures have complex foundation behaviours and seabed interactions for which there is very little or no operational experience. In such cases it is necessary, at least in the short term, to develop design methods largely from first principles using mathematical and physical modelling and the results analogous structures or research together with expert judgement. In the © 2011 by Taylor & Francis Group, LLC

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Figure 27. New design solutions using observational methods.

Figure 28. New design solutions using field trials.

by increasing component flexibility and/or strength. The key conclusions drawn from these studies are:

in the early projects to benefit later ones. For example in Angola BP is planning on leveraging selective deepwater pile design and installation experiences on a later project currently in Execute to extend the foundation design options on later projects. Investment in frontier trials would also more attractive when performed through joint industry projects (JIPs) sponsored by a number of operators, contractors and other interested parties with common interests. As engineering experience with new behavioural models increases and design confidence grows the plan would be to adopt them as standard company practices. In the longer term, this field experience may also provide the bases for industry best-practices, particularly if it could be pooled with similar data from other offshore operators. Geohazard-resistant Design. Preliminary studies for BP (Parker et al. 2008, 2009), have shown that the susceptibility of wells and bottom-founded facilities to geohazards such as submarine slides may be reduced © 2011 by Taylor & Francis Group, LLC

– Pipelines. The vulnerability of pipelines in landslide areas may be reduced by laying them in curves, a common practice in the GoM. – Sub-seabed soil movements. Subsea structures may be made more resistant to sub-seabed soil movements (landslides, erosive run-outs, fault movement etc) by increasing foundation capacities; driven piles being more effective for this purpose than suction-installed caissons due to the smaller area exposed to soil forces. Conversely, structural connections may be made more tolerant of this type of movement by increasing their flexibilities rather than strength. – Supra-seabed soil movements. Stronger foundations would also increase resistance to impact forces above the mudline (turbidity currents, mud volcano outflows, fluid non-erosive debris flows)

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Procurement and Construction (EPC) contracting. In the short term, it may even be necessary for Operators to accept more responsibility than normal for the operational performance of designs based on methods that are not field-calibrated. Irrespective, EPC tender and contract documents will need to be adapted to ensure that the risks for new design methods are fully defined and roles and responsibilities for the work are clearly understood. As mentioned earlier, one possible way of defining and agreeing contractual risks for non-standard design solutions is through the use of Geotechnical Baseline Reports (GBRs) which are discussed further below. 3.8

Earlier in this paper, the communication of the shallow subsurface conditions to stakeholders was identified as a big challenge for geo-specialists. The approach taken by the Egypt and Angola GATs has been to merge the results of their work in a single document called the Geotechnical Engineering and Geohazard Mitigation (GGM) Report that is used to assist project decision making and to underpin more purposeful deliverables that support engineering and construction, as shown on Figure 30. The GGM Report evolves as more geophysical and geotechnical data are gathered and the project’s development plans emerge and represents the consensus state-of-awareness of the shallow subsurface conditions and how these affect the planned development. The GGM Report is updated whenever there is a significant change in the state-of-knowledge or project status, triggered for example by a programme decision gate, additional survey data or major scheme changes. The GGM Report embodies the ground model and amongst other things, describes the interpreted geological and geotechnical conditions and geohazards and, with input from the engineering teams, provides general guidance on geohazard avoidance, layout planning and engineering.The GGM database is captured in GIS format for ease of communication and interfacing with others in the project team. The GGM Report also provides the platform for the production of more engineering-focused and contractual documents such as the Geotechnical Basis of Design (BoD) that is specifically tailored to the scheme taken forward through FEED in to an EPC contract. The two key elements of the Geotechnical BoD are the Geotechnical Data Report (GDR) that includes all the factual data gathered to support the project and the Geotechnical Baseline Report (GBR) that contains interpreted or assumed ‘baseline, or ‘reference’ statements about the ground conditions and risks. EPC tenderers are encouraged to propose supplementary surveys and geotechnical investigations to support the BoD should they consider the data provided at the bid stage to be insufficient to support their specific design and installation plans. The GBR represents the project’s consensus assessment of the ground conditions and geohazards and how these will be taken in to account in design and for

Figure 29. Geohazard impact framework.

and piles and caissons are equally suitable for this purpose. Structural connections also need to be strengthened but this is a design challenge since it conflicts with the need for increased connector compliance to accommodate sub-seabed ground movements – ‘Weak-link’ Design. When safety and environmental considerations are accounted for, the most pragmatic geohazard-resistant design strategy may be to provide weak links that are relatively easy to repair. Equipment, structures, foundations and anchors used routinely offshore will have some inherent reserve capacities to resist geohazard-related loads However, the resistance thresholds may be increased and risks reduced by additional engineering as illustrated on Figure 29. The investment in geohazard-resistant design would be justified if it is less than the additional operating costs and production losses that would arise if the measures were not taken. Probability-based cost-benefit analyses would support this decision. Additional Considerations. The motives behind value-engineering and new design solutions are to enhance design by increasing reliability or reducing costs, or to enable developments to proceed in controlled and responsible ways in the absence of significant field experience. However, the methods are not universally applicable and therefore need to be used selectively and with care. For example, the observational method may not be suitable for safetycritical structures or where soil-structure interactions are insufficiently ductile to give warning and time to implement planned interventions. Performance-based methods may also need to be avoided if there is potential for brittle behaviour. There may also be some contractual considerations since many of the non-routine design methods are not ideally suited to traditional offshore Engineering, © 2011 by Taylor & Francis Group, LLC

Deliverables

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Figure 30. Geotechnical documentation.

that may not be justified or practical for a single project. These include opportunities for:

construction. It also provides reference conditions for tendering or setting cost targets and for comparing with actual conditions encountered to support the interpretation of any ‘Differing or Unforeseen Site Condition’ clauses included in contracts. GBRs are most effective when they are prepared jointly by the Operator’s design team and EPC contractors and are agreed before contract award.

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– Long term agreements with geophysical and geotechnical survey contractors. – Early geophysical and geotechnical data acquisition – The use of regional ground models to support the systematic assessment of geotechnical and geohazard risks and the geotechnical engineering for wells and facilities. – Use of observational methods and field trials to support value engineering solutions or new design methods for foundation and anchor design and for predicting complex soil-structure interactions. – Applying best-practices from onshore civil engineering projects with high geotechnical exposures, such as tunnelling, to develop strategies for managing offshore geotechnical risks within an EPC contracting framework that satisfy all parties.

CONCLUDING REMARKS

This paper describes the problems faced by offshore oil and gas companies operating in geohazardous and geotechnically-challenging areas and some of the ways that BP’s geotechnical engineers and other geospecialists are tackling them. Jeanjean et al. (2005) described some guiding principles and best-practices from the company’s pioneering deep water developments in geohazard-prone areas in the Gulf of Mexico. These included: (1) operator whole-life ownership, (2) multidisciplinary geoteams, (3) phased development-wide geophysical surveys and geotechnical site investigations and (4) the development of new engineering solutions using numerical and physical modeling and reliability-based design. This paper develops these ideas and concepts with particular reference to strategies and practices that are unfolding from deep-water development programmes in the West Nile Delta and Angola. Multiple projects are planned in both these regions which allowed BP to take a long-term holistic approach to geotechnical and geohazard risk management and to take advantage of a number of investment and leveraging opportunities © 2011 by Taylor & Francis Group, LLC

The approach that BP is taking to manage and mitigate geotechnical and geohazard risks in Egypt and Angola is clearly not the only way and may not necessarily be suitable or practical for all projects. Other, equally valid, management and mitigation strategies that embody the best practices and principles described by Jeanjean et al. (2005) and in this paper are being adopted by the company in other regions such as GoM and in the Caspian Sea. Some of the ideas and practices discussed in this paper are work-in-progress and others are still aspirational. However, to date the methodology adopted has been relatively successful. Arguably, the most significant steps were: (1) the establishment of semipermanent geotechnical and geohazard assessment

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teams, GATs, for both regions to help provide projectlife geotechnical support and (2) to co-locate the GATs centrally, close to their respective project teams. The centralised GATs have technical and commercial advantages over more ad-hoc arrangements generally made for single projects, including: critical mass, proximity to project teams, technology transfer, shared best-practices and learning, standardisation, training opportunities and economies of scale. Perhaps the most important benefit of the GAT approach is that it has improved the interactions between the shallow geospecialists, deep geospecialists and the engineering teams and encouraged the holistic teamwork that is so often the hallmark of a successful project.

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ACKNOWLEDGEMENTS The author is grateful to BP Exploration and Production Technology (BP EPT) for permission to publish this paper, although the views expressed are wholly his own and not necessarily those of BP. The author wishes to thank his BP geotechnical colleagues, Junius Allen, Jim Clarke, Ed Clukey, Paul Dimmock, James Hansen, Kevin Hampson, Andy Hill, Hugo Galanes-Alvarez, Philippe Jeanjean,Attasit Korchaiyapruk, Eric Liedtke and Tony Lusted for their views and support, and Bryn Austin and Mike Fiske for leading the GAT geophysical interpretation work. Special thanks to Mike Sweeney for his support and encouragement. Also, thanks to Giles Thompson and his colleagues at Senergy Survey & Geoengineering, Tore Kvalstad of NGI, Eric Parker and his colleagues at D’Appolonia and Alan Niedoroda of URS Corp for strengthening the team when it matters. Finally, this paper would not have been possible if had not been for the dedication, hard work and professionalism of a diverse team of specialists from Halcrow Group Limited and Fugro GeoConsulting Limited who have underpinned BP’s Angola and Egypt GATs over the past 6 years. The individuals are too many to mention here but I will single out three; Roger Moore and Denys Brunsden of Halcrow and Steve Thomas of Fugro, who have been here since the beginning. Many thanks. REFERENCES AGS. 2003. Response to UK Government Minister for Construction. Association of Geotechnical and Geoenvironmental Specialists. Aldridge, T.R., Carrington, T.M., Jardine R.J., Little, R., Evans, T.G. & Finnie, I. BP Clair Phase, 2010. ‘Offshore foundation design in extremely hard till’, Proceedings Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia. Alm,T., Snell, R.O., Hampson, K.M.& Olaussen, A. 2004. Design and Installation of the Valhall piggyback structures. Proceedings, Offshore Technology Conference, Houston, Texas, OTC 16294. API. 2000. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – working © 2011 by Taylor & Francis Group, LLC

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Sultan, N. 2007. Gas hydrates stability Law, Hydrate Fraction and acoustic Velocities of Gas-Hydrate Bearing Sediments: Theoretical Study and empirical Expressions, Gas Hydrate Research Project. West Africa Deepwater Operators, (WADO), Deliverable 1. Sultan, N., Adam, S., De Gennaro, V., Lakshmikantha, M.R. & Puech, A. 2009. Acoustic properties and mechanical behaviour of marine sediments partially saturated by gas. IFREMER Scientific Report Task 2, Joint Industry Funded Project. Thomas, S., Bell, L., Ticehurst, K. & Dimmock, P. S. 2010. ‘An investigation of past mass movement events in the West Nile Delta’. Proceedings Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia. Tjelta,T. I. 2010. ‘Prod probes Statoil’s seabed soils’. Offshore Engineer, February 2010. UKOOA. 1997. UK Offshore Operators Association. ‘Guidelines for the Conduct of Mobile Drilling Rig Site Surveys’, Volume 1, March 1997 and Volume 2, April 1996. USNCTT. 1984. U.S National Committee on Tunneling Technology, Geotechnical Site Investigations for Underground Projects, National Research Council, Washington D.C. (2 Volumes) Wheeler, S.J. 1988. The undrained shear strength of soils containing large gas bubbles. Géotechnique, 38, No 3, 399–413.

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Recommended best practice for geotechnical site characterisation of cohesive offshore sediments D.J. DeGroot University of Massachusetts Amherst, Amherst, MA, USA

T. Lunne Norwegian Geotechnical Institute, Oslo, Norway

T.I. Tjelta Statoil, Norway

ABSTRACT: The paper presents recommendations for conducting site characterisation programs to determine the geotechnical properties of cohesive offshore sediments with the primary focus being on clays. The engineering behaviour of clays is complex and characterisation of their in situ properties is magnified by additional challenges presented by offshore environments. Site investigation programs are best conducted using well calibrated in situ tests and laboratory testing of high quality undisturbed samples. Results from these test programs should be coupled with geophysical data and collectively evaluated in the context of a regional geological framework. The paper reviews clay behaviour, describes the unique conditions of offshore environments, and lists key cohesive soil parameters required for infrastructure design and geohazards analysis. Best practice recommendations founded on these fundamentals are then presented including drilling methods, in situ testing and instrumentation, soil sampling, laboratory testing and evaluation of test data. The paper concludes with an assessment of present and future challenges. 1

INTRODUCTION

regional geologic framework that encompasses the past and current geologic states. The paper discusses characterisation of cohesive offshore sediments, which can encompass clays, plastic silts and organic soils, although the main focus is on soft to medium consistency clays. This is in part due to the authors’ experience is primarily with clay deposits, but also because a large portion of offshore sites consist of soft clay deposits, especially in deeper waters. There are equally important challenges with characterisation of other offshore deposits such as sands, calcareous soils, and non-plastic silts. While some of the site characterisation methods described are to varying degrees relevant to such soils, they are not the focus of the paper. The paper begins with a review of clay behaviour since such knowledge is essential to designing and implementing an appropriate site characterisation program. The unique and complex challenges presented by offshore environments such as over pressured zones, ultradeep waters, high salinity, ice gouging, etc. are described. Specific soil parameters required for design of offshore infrastructure such as anchor systems, pipelines, conductor installation, and geohazards analysis are listed. These background sections provide the context for the main objective of the paper which is to present recommended best practice for geotechnical site characterisation programs to determine reliable

Offshore geotechnical site characterisation programs are used to determine soil stratigraphy, in situ pore water pressure, and soil parameters for foundation design of infrastructure and geohazards analysis. Cohesive soils can have highly varied geologic histories making systematic quantification of their stressstrain behaviour complex. They exhibit significant stress history effects, can have a high degree of anisotropy, and exhibit strain rate effects. They are difficult to sample and test without causing excessive and irreversible disturbance. These challenges are magnified for offshore site characterisation programs since many offshore regions have complex geologic environments and site investigations are increasingly being conducted in deeper waters. Geotechnical site characterisation programs ideally combine in situ testing, in situ instrumentation, collection of high quality samples and follow-on laboratory testing. Assessment of sample quality is essential for evaluating the accuracy of laboratorymeasured mechanical properties. Coupling laboratory data with results from well calibrated in situ tests greatly enhances the reliability of site characterisation programs. The resulting geotechnical data should be studied concurrently with geophysical data and the collective data set should be evaluated within a

33 © 2011 by Taylor & Francis Group, LLC

soil parameters. Recommendations are given for all geotechnical aspects of the site characterisation process including drilling methods, in situ testing and instrumentation, soil sampling, laboratory testing, and evaluation and selection of design profiles. This paper follows other related overview papers on offshore site characterisation, notably Randolph (2004), Kolk & Wegerif (2005), and Andersen et al. (2008). The paper stresses the importance of geophysical and geological work as part of a comprehensive site characterisation program but these topics are not within the authors’ expertise. ISSMGE (2005) reviews offshore geophysical methods.

2

FUNDAMENTAL CLAY BEHAVIOUR

An understanding of clay behaviour is necessary to formulate and conduct a successful site characterisation program. It is important for the planning and execution of the in situ testing, soil sampling, and laboratory test programs so that the most relevant tools and test methods are utilized. It is also valuable during the process of comparing and analyzing the measured data sets and ultimately in development of design profiles for key soil properties such as in situ vertical effective stress (σv0 ), preconsolidation stress (σp ), consolidation behaviour, undrained shear strength (su ), and stiffness. This section reviews the fundamentals of intact (i.e. undisturbed) clay behaviour, the influence of sample disturbance on this behaviour, and the behaviour of remoulded clays. Leroueil & Hight (2003) and Ladd & DeGroot (2003) each provide a detailed treatise on clay behaviour; Lunne & Andersen (2007) emphasize behaviour of soft, deepwater clays. The focus herein is on monotonic testing behaviour; Andersen (2004, 2009) provides details on cyclic behaviour of clays.

Figure 1. Fundamentals of 1-D consolidation behaviour: compressibility, hydraulic conductivity, coefficient of consolidation vs. vertical effective stress (after Ladd & DeGroot 2003).

Key aspects of intact clay behaviour are stress history, consolidation, strength anisotropy, non-linear stressstrain behaviour, and strain rate effects.

Figure 2 plots su versus σp data for 27 low to medium OCR clays worldwide (22 from offshore locations) and with a majority having a plasticity index (PI) between 10 to 50% (NGI 2002). The su values were determined from triaxial compression tests conducted on good to excellent quality samples that were anisotropically consolidated to estimated in situ effective stresses prior to undrained shear (CAUC). The σp values were determined from 1-D consolidation tests conducted on companion test specimens taken from the same sample. The strong link between su and σp implied in Figure 2 is often expressed in terms of normalised soil parameters such that

2.1.1 Stress history All significant aspects of clay behaviour are influenced by stress history. It is quantified through the preconsolidation stress and the corresponding overconsolidation ratio OCR = σp /σv0 . The preconsolidation stress is essentially a yield stress, which separates small, mostly elastic vertical strains (εv ) from large, mostly plastic strains, as illustrated in Figure 1. It is more appropri ately referred to as the vertical yield stress (σvy ) but the familiar σp notation is used in this paper. In terms of consolidation behaviour, Figure 1 also illustrates the significant changes in the coefficient of consolidation (cv ) as a function of stress levels relative to σp . The hydraulic conductivity (kv ) decreases with an in increase in σv with an approximate linear relationship between void ratio (e) and logkv .

with a common value of m = 0.8 ± 0.1 and S depending on the particular su measurement method/mode of shear being investigated (e.g. Ladd 1991, Ladd & DeGroot 2003). Mesri (1975) and Terzaghi et al. (1996) reason that su is directly proportional to σp (i.e. m = 1) such as for example, in the case of stability problems with su (ave) = 0.22σp , where su (ave) = the average mobilized su . Linear regression to the data set shown in Figure 2, which assumes m = 1, results in su (CAUC) = 0.28σp . Jamiolkowski et al. (1985) describe several mechanisms that cause an OCR = 1 soil deposit to become overconsolidated including vertical stress relief, desiccation, ageing/drained creep (often referred to

2.1 Intact clay behaviour


Figure 4. su (CAUE) vs. σp for low to medium OCR and plasticity index offshore clays (from NGI 2002).

Figure 2. su (CAUC) vs. σp for low to medium OCR and plasticity index offshore clays (from NGI 2002).

Figure 5. Schematic of shear modulus versus shear strain (modified after Jardine 1992, Leroueil & Hight 2003). same database as the CAUC (σ1f = 0◦ ) data plotted in Figure 2. Linear regression to these data results in su (DSS) = 0.23σp and su (CAUE) = 0.17σp . The regression coefficients imply su anisotropy ratios of Ks = su (DSS)/su (CAUC) = 0.82 and su (CAUE)/ su (CAUC) = 0.61. While there is scatter in the data sets presented in Figures 2-4, the trends are evident and similar to that presented for terrestrial soils (e.g. Ladd 1991, Terzaghi et al. 1996) showing significant su anisotropy for these low to moderate plasticity soils.

Figure 3. su (DSS) vs. σp for low to medium OCR and plasticity index offshore clays (from NGI 2002).

as apparent preconsolidation), and physicochemical effects (e.g. cementation). This highlights the importance of developing an understanding of the geologic history of a site so that stress history data from in situ and/or laboratory testing can be properly evaluated and understood.

2.1.3 Non-linear stress-strain behaviour The stress-strain behaviour of clays is highly nonlinear. Clay stiffness, expressed as Young’s modulus (E) or the shear modulus (G), is greatest at small strain levels (less than ∼ 0.01% for clays) and degrades with increasing strain. Jardine (1992) presented a conceptual model and examples of soils in which the stiffness-strain degradation curve is divided into several zones, as shown schematically in Figure 5. This includes linear-elastic behaviour (up to Pt. A), followed by non-linear elastic behaviour (Pts. A to

2.1.2 Anisotropy Undrained shear strength is said to be anisotropic when its magnitude depends on the orientation of the major principal stress at failure (σ1f ), with su decreasing as σ1f rotates from vertical (same direction as deposition) to horizontal. Figures 3 and 4 plot results from consolidated direct simple shear ∼ 45◦ ) and anisotropically consolidated (DSS; σ1f triaxial extension tests (CAUE, σ1f = 90◦ ) for the


B), plastic straining (Pts. B to C), and large plastic straining (beyond Pt. C). The shear stiffness within the region up to Pt. A is the maximum value (nearly constant) and is commonly referred to as the small strain shear modulus Gmax (or G0 ). It is often determined using elastic shear wave measurements such that

where ρt = total density and Vs = shear wave velocity. Like su , Gmax is also anisotropic and thus it should be reported with appropriate subscripts Gij to explicitly indicate the direction of propagation (i) and polarization (j) of the shear wave. For example, downhole seismic testing results in a measurement of Vvh and a computed Gvh while cross-hole testing yields either Ghh or Ghv . Figure 6. Normalised static undrained shear strength versus rate of shear strain for clays (from Lunne & Andersen 2007).

2.1.4 Strain rate effects Clays are sensitive to the rate of strain or loading and can exhibit a significant increase in su at fast rates of undrained shearing. Figure 6 shows an example of the influence of shear strain rate on su of several clays. The data show not just an increase in su with an increase in strain rate, but furthermore that the change in su per log cycle increases as the strain rate increases. The magnitude of the effect can depend on many factors including clay type, OCR, stress path, and may depend on whether the load is strain- or stresscontrolled and whether the shearing is static or cyclic (Lunne & Andersen 2007). Nevertheless, the effect is present for all such conditions and must be considered when evaluating laboratory and in situ test data relative to field conditions, all of which typically have significant differences in the rate and duration of loading. The rate effect shown in Figure 6 is often modeled using a simple log function with most data showing an increase in su between 5 to 20% per log cycle increase in strain rate. The effect can also be modeled using an inverse hyperbolic sine function (e.g. Randolph 2004), which allows for decaying strain rate effects (as is evident in Figure 6) below a specified threshold rate.

Figure 7. Hypothetical stress path for a low OCR clay element during tube sampling, specimen preparation and undrained shear (from Ladd and DeGroot 2003).

Each stage of the sampling process, from initiation of drilling to preparation of laboratory specimens, causes potential disturbance. The most important effect of sample disturbance in low to moderate OCR clays is a destructuring of the soil, which is accompanied by a significant reduction in the sample effective stress (σs ). For example, Figure 7 shows how the reality of sampling and testing can vary unpredictably from ). This figure shows the anticithe perfect sample (σps pated stress path for a low overconsolidated clay as stresses change from the in situ stress state (Point 1) to the stress state at laboratory testing (Point 9) as a result of disturbance caused by sampling, storage and handling. There are similar impacts on the small strain shear modulus, as depicted schematically in Figure 8 in terms of the shear wave velocity. Stress relief alone results in a reduction in Vvh , which is presumably recoverable during laboratory reconsolidation, however, destructuring results in an irreversible decrease in Vvh . It is common in geotechnical engineering practice to rely on the behaviour measured from laboratory

2.2 Influence of sample disturbance Sample disturbance is the most significant issue affecting the quality and reliability of laboratory test data for clays. It causes changes in the natural soil state and structure and as a result all key design parameters such as E, G, σp and su are adversely influenced by sample disturbance. Figure 1 schematically shows this for the one-dimensional consolidation properties with sample disturbance resulting in a more rounded compression curve with greater εv at all stress levels. This tends to obscure and usually lower σp , especially for more structured soils. The only parameters not significantly affected by sample disturbance are cv well beyond σp and the e–logkv relationship, unless there is severe disturbance.


Figure 8. Hypothetical reduction in sample effective stress and shear wave velocity of a low OCR clay for the perfect sample and a disturbed sample.

index strength tests (e.g. unconfined compression tests, UUC). In such tests, the specimen is at an effective stress state represented by Point 9 in Figure 7 when, in fact, the in situ behaviour under field loading starts at Point 1. Moreover, not only is the effective stress state incorrect, but destructuring that occurs during poor sampling and handling further magnifies differences between measured laboratory and in situ behaviour. As an example of these effects, Figure 9 plots data showing the influence of different samplers on the undrained shear behaviour of the Onsøy, Norway, clay. The results are from CAUC tests conducted on Sherbrooke Block, 76 mm tube and 54 mm tube samples. The Sherbrooke Block sampler generally collects high quality samples followed by decreasing sample quality for the 76 mm tube sampler and especially for the 54 mm tube sampler (e.g. Lunne et al. 1997a, 2006). There are significant differences in the strain-strain-strength behaviour as sample disturbance increases, including: decrease in su , decrease in rate of strain softening, and increase in large strain shear strength. Lunne et al. (2006) show additional examples from DSS and CAUE tests for the Onsøy clay and several other clays.

Figure 9. CAUC recompression tests conducted on samples of Onsøy, Norway, clay from 14.5 m: a) stress-stress curves, and b) effective stress paths (from DeGroot et al. 2007).

been performed using the field vane and more recently full-flow penetrometers. As an example of some of these issues, Figure 10 presents results of motorized laboratory vane (MLV) tests conducted on an intact sample of Troll clay from the Norwegian sector of the North Sea. The sample was first tested to measure the intact su and the post peak residual shear strength. It was subsequently remoulded through multiple rotations of the vane (VR) and again tested. Thereafter the sample was thoroughly remoulded by hand (HR), tested, and followed by a final remoulding through multiple vane rotations (HRVR). Different values of sur also resulted from other methods including that of the fall cone (FC). In fact, for the FC, there are different calibration factors in current use (e.g. Norwegian Standard, Swedish Standard, ISO) and different values result from these various methods. The resulting sensitivity values St = su /sur depend on which measure of sur , is used to compute the sensitivity, and in the case presented in Figure 10, ranges from 2 to 12 for the MLV. Lunne & Andersen (2007) present data from UUC tests conducted on remoulded clays at different strain rates. The results indicate that the rate effect clearly observed for intact clays (Figure 6) also occurs for remoulded clays. Preliminary tests conducted at NGI also showed sur may be anisotropic.

2.3 Remoulded clay behaviour Some design solutions (as outlined in Section 4.0) require information on the remoulded undrained shear strength (sur ) and also its strength and stiffness evolution with time after remoulding. Laboratory measurement of sur has the significant advantage over measurement of su because sample disturbance is generally not an issue. However, it remains a relatively complex parameter to evaluate because results can vary greatly depending on the degree of remoulding and the measurement method used. Some test devices require a remoulded test specimen to be prepared (e.g. fall cone) while other devices can be used to remould an intact soil (i.e. device remoulding) and also directly measure sur (e.g. laboratory or in situ vane test). In situ methods of estimating sur , which all employ device remoulding by default, have long


Table 1.

Summary of key aspects of clay behaviour.

Clay behaviour

Significance

Stress history (σv0 , σp , OCR) su anisotropy

Most important factor, controls all significant aspects of clay behaviour Very significant for low to medium PI clays, su (CAUC) > su (DSS) > su (CAUE)† Significant degradation in stiffness (E or G) with increasing strain Increase in su with increase in strain rate Value depends significantly on how measured and method of remoulding. Exhibits rate effects and anisotropy Most significant issue affecting quality and reliability of laboratory test data

Non-linear stress-strain Rate effects Remoulded sur Sample disturbance

Note: † except for varved clays where su (DSS) < su (CAUE)

should always be considered when selecting equipment and test methods to conduct a geotechnical site investigation. Figure 10. Shear stress vs. angular rotation for motorized laboratory vane tests conducted on Troll clay from 4.1 m depth.

3

UNIQUE CONDITIONS IN OFFSHORE ENVIRONMENTS

The various aspects of fundamental clay behaviour presented in the previous section challenge any site investigation program. However, these challenges are magnified in the offshore environment because of numerous unique logistical issues and geologic conditions that are often present. This section presents some of these conditions and describes how they add another layer of complexity to characterising the engineering properties of offshore sediments. Many of these conditions are classified as offshore geohazards, a topic for which keynote papers were presented at ISFOG 2005 by Jeanjean et al. (2005) and at ISFOG 2010 by Evans (2010). 3.1

Figure 11. Fall cone undrained shear strength (left axis) and shear modulus (right axis) vs. time for thixotropic test conducted on remoulded Troll clay.

Remoulded clays, if left undisturbed at constant water content and temperature will typically exhibit significant thixotropic hardening. Figure 11 plots results for the same Troll clay in Figure 10 showing the significant gain in su and Gvh (as measured using bender elements) with time after thorough hand remoulding. The rate in gain in su and Gvh are similar implying a near constant rigidity index IR = Gvh /su during thixotropic hardening. 2.4

Summary

Table 1 lists a summary of the fundamental aspects of clay behaviour described in this section. It is evident that clay behaviour is complex and this complexity


Non-hydrostatic pore pressures

Many offshore regions have in situ pore pressures that are in excess of hydrostatic (i.e. overpressures). These overpressure zones can be caused by rapid sedimentation, mud volcanoes, pore fluid flow, manmade activities, and other mechanisms as summarized by Dugan et al. (2010). Given that the in situ vertical effective stress profile is critical to determining soil design parameters, detection of overpressured zones and measurement of the overpressure is an essential part of a site characterisation program. Overpressures are believed to be a contributing factor to the massive Storegga submarine landslide in the Norwegian Sea (Kvalstad et al. 2005) and have caused major difficulties during drilling operations in the Caspian Sea (Allen et al. 2005). Figure 12 presents an example of overpressures for the West Azeri field in the AzeriChirag-Gunashli (ACG) development in the Caspian Sea as measured using pore pressure dissipation tests

Figure 13. Schematic of reduction in effective stress state for consolidated OCR = 1 clay subject to post deposition influx of pressurised pore water.

platform. There are many such cases in SE Asia where poor cementing results in gas charging of shallow permeable layers. 3.2 Figure 12. Pore water pressure versus depth, West Azeri, Caspian Sea (after Allen et al. 2005).

(Section 5.5). The magnitude of overpressure equals about 800 kPa at 230 m below the seabed, and over the profile is supporting 40 to 50% of the soil buoyant unit weight. Determining the genesis of overpressures is also important. If it is a result from ongoing self-weight consolidation in a rapid sedimentation environment then the deposit is considered “under-consolidated” relative to the eventual final equilibrium σv0 . In such cases the deposit will be relatively weak and assuming hydrostatic pore pressures could lead to significant errors in selecting consolidation stresses for laboratory strength testing and evaluating final su design profiles. Conversely, a deposit that has excess pore pressures that evolved after self-weight deposition pore pressures have come to equilibrium should not be categorized as under-consolidated. One mechanism that can cause this is upward or lateral fluid flow from adjacent overpressure regions. If the onset of fluid flow excess pore pressure occurs after a previous equilibrium effective stress state, then the deposit will be overconsolidated (Figure 13). In some regions excess pore pressure from both rapid sedimentation and fluid flow from adjacent overpressure zones cans occur concurrently which makes for a complex in situ state to measure and evaluate. Although Dugan et al. (2010) note that recent advances in modeling overpressures have better defined different mechanisms and spatial scales. Overpressures can also result from manmade activities. For example, Lunne et al. (1996) report on gas charging of permeable layers from deep gas leaking upwards through weak cement around conductors in the Duyong B field, Malaysia. It was discovered during geotechnical drilling for a new nearby platform when a blow-out occurred. The gas charging built up pore pressures in adjacent clay layers threatening the safety of the foundation system of the existing

3.3

Highly irregular seafloor topography

There are numerous past and present geologic events that create highly irregular seafloor topography. Examples include mud volcanism, seabed slumping, small to mega-submarine landslides, debris flows, channel erosion, faulting, salt diapirism, and ice gouging. Such complexity in the seafloor topography presents obvious challenges to locating and installing offshore infrastructure, especially for routing of pipelines. It also implies that there can be significant spatial variability in soil properties, which requires careful planning and layout of the geotechnical site exploration program. 3.4

Gas exsolution

Stress relief from sampling, especially in deep waters, can cause exsolution of even small amounts of gas dissolved in the pore water. The subsequent expansion, if undrained, can damage the soil structure which in turn will impact the quality of the measured soil behaviour


Low effective stress state

For flowlines and pipelines, the largest category of offshore infrastructure, soil-structure interaction takes place in the upper few decimeter of soil where there are very low effective stress (e.g. σv0 ∼ 2 to 3 kPa at 0.5 m depth). This, along with temperature induced cyclic loading effects (from episodic flows causing pipe contraction and expansion), complicates characterisation of the in situ soil properties such as modulus and shear strength. Laboratory measurement using consolidated test specimens (e.g. CAUC) is difficult to do reliably at such low effective stresses. For in situ testing methods, such as the piezocone and full-flow penetrometers, the tip force acting on the device from hydrostatic water pressure at the sea bottom can be very large relative to the penetration resistance of the upper few decimeters of soil. This creates measurement inaccuracy issues; a problem that is especially significant for soft soils and/or as water depth increases (e.g. Lunne 2010).

and design parameters. Lunne et al. (2001) conducted a laboratory test program that simulated deepwater sampling of clay with various amounts of dissolved gas. The measured CAUC stress-strain-strength behaviour was influenced by the degree of gas saturation, with a large reduction in su with increasing gas saturation. When disturbance caused by gas exsolution was coupled with simulated tube sampling disturbance, the combined effect was significant. CAUC reconsolidation volumetric strains were larger implying greater disturbance from the combined effect. But the stressstrain curves tended to strain harden, presumably due to the large volumetric strains and reduction in void ratio prior to undrained shear. The results indicate that disturbance caused by gas exsolution depends on the state of soil structure prior to exsolution with an intact structure being better able to resist gas exsolution than an initially disturbed structure.

3.5

Surficial crust

Figure 14. Evidence of surficial crust zone based on CPTU data, offshore Nigeria (from Ehlers et al. 2005).

Ehlers et al. (2005) describe the presence of a thin crust zone near the seabed in several deepwater regions as shown for example in Figure 14. This su profile, with the relatively high strength within the upper meter, is unusual compared to the more familiar deepwater soft clay profile which starts with the lowest values at the seabed and increases near linearly with depth. It is hypothesized that these crusts are a product of bioturbation and geochemical transformation, primarily into pyrite (Ehlers et al. 2005). Detecting the presence of these thin surficial crusts and properly characterizing the strength and stiffness properties is especially important to design and installation of shallow foundation systems, flowlines and pipelines (Section 3.2).

3.6 High salinity Some offshore regions have been found to have extremely high pore fluid salinity. Figure 15 plots data for the West Azeri field in the Caspian Sea (same site as Figure 12). The pore fluid salinity reaches values as high as 250 g/kg in comparison with Caspian Sea water of 12 g/kg. This high pore fluid salinity is believed to be associated with mud volcano activity as there appears to be higher pore water salinity at locations that are close to mud volcanoes (Kay et al. 2005). The upward migration of materials within the mud volcanoes includes pore water from greater depths and may also form conduits for further fluid flow. Excess pore water pressure in the vicinity of the mud volcanoes provides a mechanism for migration of dense brine vertically and laterally. High pore water salinity needs to be taken into account for computation of in situ pore water pressure. It also impacts many aspects of laboratory testing including measurement of index properties that rely on an oven dry mass and the need to match water used for saturation and testing of consolidated specimens to the pore fluid salinity.

Figure 15. Pore fluid salinity vs. depth below seabed for West Azeri region, Caspian Sea (after van Paassen & Gareau 2004).

In the Caspian Sea ACG development, there are high plasticity clay units that were found to be slickensided (Allen et al. 2005). This indicates that the sediments were subjected to failure conditions with repeated shearing. These soil units were not exposed to subaerial conditions and thus have not been desiccated. The likely mechanisms are mechanical, such as deformation of the soil units over the crest of the underlying Apsheron Ridge, or physio-chemical. Synaeresis cracks can occur in submerged sediments due to volume changes induced by changes in pore fluid salinity (Collinson & Thompson 1982). These post-deposition salinity changes are present at several of the ACG fields where the original pore fluid has been displaced by high salinity pore fluid migrating


Figure 16. Normalised preconsolidation stress versus temperature (from Leroueil & Marques 1996).

from adjacent overpressure zones. It is likely that the fluid migration is cyclic, and in combination with the high plasticity swelling clay minerals present, is a condition conducive to synaeresis. Figure 17. Preconsolidation stress for ice-gouged and non ice-gouged sites in Smith Bay, Beaufort Sea (modified after Young 1986).

3.7 Temperature changes In offshore polar regions and in deep water, the seabed temperature often approaches 0◦ C (and even slightly below) with the pore fluid freezing prevented by the salt concentration. For example, Bugge (1983) reported that on the Norwegian continental margin at latitude of 64◦ N, the sea water temperature equals 0◦ C at 850 m and reaches an equilibrium temperature of −0.9◦ C at 1300 m. In the Mexican part of the Gulf of Mexico, Vidal et al. (1994) report temperatures lower than 5◦ C in 1200 m water depth. Collecting soil samples from a 0◦ C degree environment and testing them at room temperature (∼20◦ C), which is the common practice, raises the issue of temperature effects on the measured behaviour. Another temperature effect is that during production, hot gas or oil in pipelines can cause heating of surrounding soil. Leroueil & Marques (1996) studied the influence of temperature on σp and found a nearly 1% decrease in σp per ◦ C increase for temperatures ranging from 5 to 40◦ C, as shown in Figure 16. Since su is linked to σp , as described in Section 2.1.1, a similar temperature effect on su is expected (Perkins & Sjursen 2009). 3.8

out to sea at approximately 20 m water depth. Both sites involve the same Pleistocene deposit with Site W being heavily overconsolidated with nearly constant σp suggesting mechanical precompression due to erosion. Site T is at a greater water depth and in a zone of extensive ice gouging, which was determined in part from geophysical data. The repeated reworking at Site T caused significant softening within the upper 2 meters of the initially heavily overconsolidated deposit. Below 2 meters the σp values rapidly increase with depth as the effects of ice gouging diminish and coincide with those of Site W. Another interesting feature in offshore arctic regions is the presence of subsea relict permafrost below existing unfrozen sediment. For example, in regions of the Beaufort Sea, the surficial soils in shallow waters are often soft sediments that have a temperature just below freezing, but are unfrozen due to higher pore fluid salinity. These deposits were found to be underlain by permafrost, which formed during the last ice age when the sea level was much lower than present and the then subaerial deposits were subject to extreme freezing temperatures. During subsequent sea level rise, some of the top portion of the permafrost thawed and in some regions was covered with Holocene sediments. As a result, a veneer of soft sediment exists over a semi-rigid layer of permafrost. Foundation analysis for gravity based structures in such circumstances will involve modeling squeezing of a relatively thin layer of soft sediments. Additional problems, such as settlement, can result if oil and gas production melts the permafrost. In polar regions, ice loading can impose large lateral forces on offshore infrastructure such as gravity platforms. Ice loading depends on the nature of the ice encountered (i.e. landfast, transition or polar pack) and will have both mean and cyclic components. When

Ice gouging, subsea relict permafrost, and ice loading

Offshore regions that have been, and continue to be, influenced by the presence of sea ice often have several unique geologic features. Sea ice keels that interact with the seabed can gouge the surficial sediments. Repeated ice gouging reworks the seabed and can significantly reduce σp and su . It can also result in significant variations in seafloor topography. Figure 17 plots an example for the Smith Bay region of the Beaufort Sea, offshoreAlaska.The Sites W andT are located about 17 km from each other with Site W closer to land at approximately 10 m water depth and Site W further


Table 2. Some unique offshore conditions and implications for conduct of site characterisation programs.

combined with gravity forces imposed by the structure, it creates a complex stress condition within the foundation soil. In some cases it can include significant shear stress reversals that are difficult to characterise (e.g. DeGroot et al. 1996).

Condition

Site characterisation remarks (details in Section 5)

Overpressures

3.9

Essential to detect and measure, use in situ dissipation tests and/or install piezometers Low effective Use in situ tools with high sensitivity stress conditions or hydrostatically compensated CPTU, collect high quality surficial samples and test on the vessel deck (e.g. box coring) Highly irregular Select in situ testing and sampling topography borehole locations to ground truth geophysical data and check for anomalous soil properties, carefully assess soil spatial variability Gas exsolution Use water sampling probe (e.g., BAT) to measure dissolved gas, consider use of pressure sampler Surficial crust Conduct accurate shallow depth in situ testing, collect thick (∼1m) box cores (if possible) and test on vessel deck High salinity Use corrected γw for computation σv0 , correct measured soil properties for salt content, could be indicator of overpressures Temperature Research ongoing, results may lead to changes development of data correction procedure Ice gouging, Select in situ test and soil sampling subsea relict borehole locations to ground truth permafrost geophysical data, carefully assess soil spatial variability Shallow gas and Use soil sampling boreholes to ground gas hydrates truth geophysical data, consider use of pressure samplers Cyclic loading Must collect undisturbed samples and conduct laboratory cyclic test program

Shallow gas and gas hydrates

Shallow gas and gas hydrates are hazards that can be encountered during site investigation and production drilling. Shallow gas can cause dangerous blow-outs during drilling operations (e.g. Lunne et al. 1996). The presence of gas hydrates, which have been found in many deep water regions, is a potential hazard because of the large volume expansion that can occur if they melt or dissociate. Geophysical methods are typically used during a site investigation program to detect the presence of shallow gas and gas hydrates. Ideally follow-on soil sampling should be conducted to ground truth findings from the geophysical surveys. This may require special pressure sampling equipment that can maintain in situ conditions in samples during their retrieval and storage (e.g. Kolk & Wegerif 2005). 3.10 Cyclic loading Cycling loading of offshore infrastructure can result from wind, waves, ice, and earthquakes. In production facilities such as pipelines, cyclic loading can be induced by temperature changes due to episodic flow of product. In soft cohesive soils, cyclic loading causes a reduction in effective stress due to pore pressure generation and hence a reduction in su . The behavioural response during cycling loading depends on the stress path and the combination of average and cyclic shear stresses. It is another level of complexity beyond that of the monotonic shear response of cohesive soils as covered in Section 2. Andersen (2004, 2009) presents comprehensive reviews of the cyclic behaviour of soils and provides guidance on testing and design applications.

3.12

Table 2 lists the unique and demanding offshore conditions discussed in this section. It also provides some comments that are a preview of the recommendations given in Section 5 for conduct of geotechnical site characterisation programs to evaluate these conditions.

3.11 Deep to ultra deep waters Site investigations in deep (>500 m) to ultra deep (>2000 m) water depths pose a multitude of logistical and soil testing and behaviour issues. Many of the challenges noted in the previous sections are exacerbated when working in deep to ultradeep waters. Some of these include: the regions are usually remote and thus have limited support from onshore facilities, handling of equipment can take a long time due to the large water depth, large hydrostatic water pressures may limit the accuracy of some in situ tools (Section 3.2), the degree of stress relief for samples is very large and there is a higher potential for gas exsolution (Section 3.4) and dissociation of gas hydrates (Section 3.9), the soil units are typically very soft (although sometimes a thin surficial crust exists Section 3.5), and the seabed temperature is often near 0◦ C in some regions (Section 3.7).

4

REQUIRED SOIL PARAMETERS FOR DESIGN AND ANALYSIS

This section lists soil parameters required for design and analysis of offshore infrastructure and assessment of geohazards involving cohesive soils. This information when coupled with a fundamental knowledge of soil behaviour and an awareness of the unique conditions that can often exist at offshore sites should be used to formulate the required outcomes from a geotechnical site characterisation program. Table 3 provides a list of soil parameters and examples of design concepts and geohazards for which these


Summary

Table 3. Summary of cohesive soil properties required for various design concepts and geohazard analyses (modified after Andersen et al. 2008). Soil property

Requirements

Soil classification Index strength (e.g. FC, MLV, UUC, torvane, etc.)

– required in all cases – not required in many cases but almost always performed because tests are quick and easy to conduct – used in empirical design methods – required in all cases /σv0 for some – need K0 = σh0 applications (e.g. allowable drilling pressure) but difficult to measure – CR for set-up analysis of some anchor systems and skirted seabed structures; analysis of pipelines and risers – CR and Cαε for settlement computations (e.g. gravity based structures) – set-up analysis of some anchor systems, skirted seabed structures, piles; analysis of pipelines and risers – settlement time rate computations (e.g. gravity based structures) – anisotropic su required for most applications (e.g. tip resistance of anchors, pile shaft friction, shallow foundation bearing capacity, jack-up spudcans, slope stability) – Gmax and anisotropic stress-strain behaviour required for slope stability, skirted seabed structures – sur required for penetration resistance of suction anchors; capacity of anchor systems and skirted seabed structures; analysis of flowlines, pipelines, risers and debris flow – thixotropy and reconsolidated sur required when calculating set-up effects (e.g. anchor systems, skirted seabed structures) – set-up analysis of some anchor systems (e.g. suction, torpedo) and skirted seabed structures

In situ stress state – σv0 – u0 – σp Compressibility – consolidation CR = ε/logσv – creep Cαε = ε/ logσv

Intact soil flow properties – kv and cv – rk = kh /kv

Undrained shear strength – su anisotropy – Gmax and anisotropic stress-strain behaviour (including strain softening) Remoulded undrained shear strength – sur or St = su /sur – thixotropy – reconsolidated sur

Remoulded compressibility and flow properties – CR, kv and cv Cyclic properties – permanent/cyclic shear strains, cyclic pore pressures Dynamic properties – strength, modulus and damping Rate effects Temperature effects

Figure 18. Map of soil tests and design applications (from Randolph et al. 2007).

parameters are required. The information presented in this table is abstracted from the comprehensive design focused papers by Randolph et al. (2005) on challenges of offshore geotechnical engineering and Andersen et al. (2008) on deepwater geotechnical engineering. Clearly all site characterisation programs will involve tests necessary to classify the various soil units encountered. In terms of specific soil parameters, the in situ stress state (σv0 , u0 , and σp ) is the most important and required in all cases. For cohesive soils, su is also required for almost all cases. The need for other parameters depends on the specific design concept and geohazard being analyzed. Some of the factors that influence clay behaviour have a compensating effect, although to variable and typically unknown degrees. It must be considered when linking results from laboratory and in situ tests with soil response conditions anticipated for a design application. Figure 18 illustrates this interplay for the undrained shear strength. It maps the relationship between the influence of remoulding (decreases su ) versus the influence of strain rate (increases su ) for lab and in situ testing relative to that of various offshore systems.

5

– required of all applications involving cyclic loading

Competent site investigation programs utilize an integrated approach that engages a multidisciplinary team of geo-specialists. This is especially important for assessment of complex geohazards as described for example by Jeanjean (2005), Kvalstad (2007), and Evans (2010). Geophysical, geological and geotechnical investigation methods are often conducted in a phased approach to take advantage of the best that each method offers and as project requirements evolve. A collective evaluation of the resulting data sets by a multidisciplinary team provides the greatest opportunity

– required for all applications involving earthquake loading – consideration required in most cases – required for all cases with soil temperature change > about 10 to 20◦ C


BEST PRACTICE RECOMMENDATIONS FOR GEOTECHNICAL SITE CHARACTERISATION

detailing the requirements for the project. It is also important that the field and laboratory factual reports explicitly describe the equipment and test methods used.This documentation is especially important when multidisciplinary teams, which are often composed of members from different organizations and far removed from the work conducted, need to use the factual reports to interpret the results and develop design parameters. In terms of notation for soil parameters, examples of poor practice in site characterisation reports persist. One such example is reporting of su values. As discussed in Section 2, su is not a unique parameter and its value depends on how it is measured. It is therefore essential to explicitly indentify the measurement method used when reporting su values, be it laboratory measured, e.g. su (FC), su (CAUC), etc. or interpreted from in situ test data using N factors relative to a reference su measurement, e.g. Nkt,CAUC , NT-bar,CAUC , etc. (see Section 5.8 for definitions of these N factors). Recommendations: While publication of the proposed ISO standard on Marine Soil Investigations will not be a panacea, and it may not necessarily be thoroughly adopted worldwide, it will provide a muchneeded reference framework. Project specifications should be developed detailing required site investigation activities. Field and laboratory factual reports should thoroughly document equipment, procedures, and data processing methods used. Notation used for soil parameters should be more explicit than is often the case to avoid ambiguity.

for an effective and reliable site characterisation outcome. In the sections to follow, best practice recommendations are given for the geotechnical component of site characterisation. The focus is primarily on presentation of key recommendations with many of the details provided in the references cited. These recommendations should be tailored to the scope of the investigation (i.e. preliminary, intermediate or final), anticipated challenges (Table 2), applications being considered (Table 3), schedule, and budget. 5.1 Work scope Several non-engineering factors such as availability of equipment and personnel, time constraints, and budget, often dictate the final work scope. Nevertheless, proper planning of the investigation involves making decisions on the type, depth, and number of in situ tests and/or borings that should be planned for. Most often this is governed by the depth of stress influence of the design concept and soil spatial variability. In terms of depth, the extremes include pipelines for which only the upper few meters are of interest to analysis of large submarine landslides that may require borehole depths of 100 m or more. Recommendations: Specific recommendations on number and depth of boreholes are not given here because it is highly project dependent (i.e. application, location, soil variability, anticipated geohazards present, etc.). In principle, it is recommended that both in situ tests and high quality sampling, followed by laboratory testing, should be performed. SUT-OSIG (2000, 2004) and ISSMGE (2005) provide general guidance on developing exploration plans. 5.2

5.3 Vessels and deployment modes ISSMGE (2005) gives an overview of the type of platforms and vessels that can be used for offshore investigations. In shallow waters, the common options are anchored barges, jack-up rigs, and anchored vessels, which all must provide a stable platform. In deeper waters, options include specialized soil drilling vessels with a moon pool or geophysical survey vessels that handle geotechnical tools (i.e. soil sampling and in situ testing devices) using an A-frame, crane or winch. Use of dynamic positioning is more efficient than vessel anchoring systems in shallow to moderate water depths, and is required for deep water investigations. The draft ISO standard on Marine Soil Investigations (ISO 2010) defines deployment modes as being either “drilling mode” or “non-drilling mode” (traditionally referred to as downhole and seabed mode, respectively). In drilling mode, a borehole is advanced into the seabed using rotary drilling. At selected depths a geotechnical tool is lowered into the drill string and advanced into the seabed from the bottom of the borehole. The drilling can be performed from a vessel or stable platform (“vessel based drilling”) or from a seabed system (“seabed drilling”), which involves a remotely operated drill rig that is placed on the seabed. Non-drilling mode involves using geotechnical tools that are advanced from the seabed. This includes simple free fall devices or more sophisticated equipment

Standards, reporting and notation

There is a large body of standards that cover geotechnical site investigations with most of these developed for in situ and laboratory testing of terrestrial soils. These standards have been developed by various countries and organizations (e.g. ISO, BS, ASTM, etc.) and in some cases there is a significant difference amoung such standards for conduct of the same test. This creates opportunities for confusion and mistakes in using the resulting data. At present the only comprehensive standard that was specifically prepared for offshore investigations is NORSOK G-001 (2004). This standard was originally developed in the 1980s for projects offshore Norway but evolved over time with input from an international group of North Sea operators. In 2007, ISO formed an international committee to develop a new standard on Marine Soil Investigations (ISO 2010), using NORSOK G-001 as a basis. At the time this paper was written (May 2010) a draft of the full standard was completed and under review. In spite of standardization efforts, some standards leave considerable latitude on how to conduct a test and in some cases no standards exist. For each soil investigation, project specifications should be prepared


Table 4.

that is landed on the seabed prior to commencement of testing. The important distinction here is that no borehole drilling takes place and thus the depth of penetration is limited. There is a large variety of nondrilling mode equipment being used and the quality of data obtained can vary significantly. More details and recommendations on this are given in Section 5.4 for in situ testing and Section 5.6 for sampling methods. The technology for seabed drilling systems continues to evolve with the development of increasingly versatile and sophisticated robotic units. For example, the Portable Remotely Operated Drill (PROD) as originally described by Carter et al. (1999), is currently used in commercial applications. Upon landing of the system on the seabed, rotary drilling, in situ testing, and sampling are controlled via robotic assembly of the drill string. Casing, rods, in situ tools, and samplers are stored on a carousel. In situ tools can be deployed directly from the seabed or at the bottom of a drilled borehole. Borehole depths in excess of 100 m have thus far been achieved using this equipment. In deep water, this type of seabed drilling system is more efficient than vessel drilling due to less handling of the drill string. Recommendations: Depth accuracy during drilling, in situ testing, and sampling is critical. For vessel based drilling, it is important to utilize a heave compensation system to stabilize the drill string. Good systems use the so-called “hard-tie” method that stabilizes the drill string against surface motions by using an installed seabed template as the reference point. Set-down of seabed systems and seabed templates can disturb the upper seabed and also apply a stress field to it. To minimize these problems, the set-down must be carefully controlled and the equipment should be designed to minimize imposed seabed stresses within the region of in situ testing and soil sampling (see Lunne et al. 2010).

Deployment method Vessel based drilling Seabed drilling Non-drilling mode – seabed system ROV mounted Free fall

Rod type and potential penetration depth (m) drop-in or wireline (>100 m) carousel w/ casing and drill string (75 m) or wireline (150 m) straight (40–50 m), carousel (40–50 m) or coiled (15–30 m) straight, coiled or split (all 3 m+) 15–20 (?) m

Figure 19. Schematic of recommended sequence for recording deck-to-deck reference readings for seabed in situ testing (from Lunne et al. 2010).

not necessary for coiled rod systems. Remotely operated vehicles (ROV) have the significant advantage of being able to maneuver to specific locations. The obvious disadvantage is limited thrust capacity and hence penetration depth; adding suction anchors to the ROV will increase capacity but such a system has not yet been deployed in practice. Free fall penetrometers are either expendable or retrievable with a cable system and are relatively easy to deploy. Recommendations: Long continuous push strokes are preferred for push-in tools such as the CPTU and full-flow penetrometers. It is also important when deploying these tools to monitor zero reference readings at all stages of the test as shown in Figure 19 (i.e. on deck prior to deployment, at sea bottom before and after push, back on deck). This deck-to-deck logging is considered an important quality control procedure and should always be conducted (Lunne et al. 2010).

5.4 In situ testing 5.4.1 In situ testing deployment systems Table 4 lists the various deployment systems that are available for in situ testing. Vessel based drilling with downhole in situ testing is conducted using pressurized drilling mud or a hydraulic system to advance the tool (the depth limitation is normally 500 to 600 m in the latter case). For tools such as the piezocone (CPTU), stroke lengths of up to 3 meters are possible. A disadvantage of using pressurized mud that does not use a data umbilical line is that data are stored on a memory module and cannot be viewed in real time. Seabed drilling systems that use pushing by a drill string have the advantage of being able to push an in situ tool 10 m or more from the seabed, or the bottom of the borehole. Seabed drilling systems that use wire line deployment are in development (e.g. Boggess & Robertson 2010) and should be able to test at greater depths, although the push stroke for in situ tools will likely be limited to about 3 m. Non-drilling mode seabed systems that use straight rods require the use of a constant tension winch or a tower to support the rods, which is

5.4.2 In situ tools Table 5 lists the main in situ tools available for site investigations with the CPTU being the most common tool. There is a large body of experience in using the


Deployment systems for in situ testing tools.

Table 5.

In situ testing tools.

In Situ Test/ Measurements

Comments

Piezocone (CPTU) – tip resistance (qc ), sleeve friction (fs ), pore pressure (u), inclination

– most common tool – best to use differential measurement but equipment still in development

Seismic CPTU – same as CPTU + Vvh

– requires a seabed shear wave velocity source

Full-flow penetrometers T-bar, ball – intact and remoulded penetration resistance

– higher resolution than CPTU, minor overburden correction, cyclic testing can be used to measure shear strength degradation

Field vane – intact and remoulded shear resistance

– remoulded shear strength depends on number of rotations used to remould soil

Piezoprobe – in situ pore water pressure

– data often difficult to inter-pret (covered in Section 5.5) – collects pore water sample to analyze for dissolved gas

Deep water gas probe – water sample, temperature

standard CPTU it is a more sensitive device (for same capacity load cell). Full-flow probes are deployed in the same manner as the CPTU using the same penetration rate of 20 mm/s. Cyclic testing using short strokes (e.g. ± ∼0.5 m) at selected intervals during the penetration phase of the test allows for measurement of shear resistance degradation upon remoulding. The field vane test (FVT) has been used extensively in soft Gulf of Mexico soils. For terrestrial soils, it has been calibrated for use in analyzing stability problems to account for anisotropy, rate effects and insertion disturbance (Bjerrum 1973). However, in offshore practice, the Bjerrum correction factor is typically not applied (Kolk et al. 1988, Randolph 2004) and the measured results are directly used to report su (FVT). This is justified in part because the reduction in shear resistance due to insertion disturbance is partially compensated by the high strain rates applied by the test (Randolph 2004). Expendable penetrometers and recoverable free-fall penetrometers continue to be studied (e.g. Aubeny & Shi 2006, Mosher et al. 2007) with some promising results. However, more research is needed before it can be recommended as a reliable tool. Recommendations: Lunne et al. (2010) provide guidelines for use of the CPTU, full-flow penetrometers, and the FVT in deepwater soft clays. Most of these recommendations are also relevant to their use in shallow water depths. DeJong et al. (2010a) outline recommended practice for onshore full-flow penetrometer testing, much of which is also relevant for offshore practice.

CPTU and in interpreting the measured data to determine soil parameters (e.g. Lunne et al. 1997b). Saturation of the pore pressure measurement system is not a problem offshore, in contrast to terrestrial applications. However, there are accuracy problems when testing in deep waters because of the high hydrostatic pore water pressure acting on the sensors at the seabed – a problem that is exacerbated when testing soft sediments. Measuring the differential cone resistance and pore pressure would solve this problem but no such cone has yet been deployed in routine practice. Lunne (2010) points out that due to differences in design, the sleeve friction depends on the type of cone penetrometer used. This makes sleeve friction values less reliable compared to the tip resistance and pore pressure. The seismic CPTU uses geophones or accelerometers to measure shear wave velocity arrival times which can be used to estimate the small strain shear modulus via Equation 2. The shear wave source is usually deployed at the seabed. Most seismic CPTU systems use a single set of sensors and the depth specific travel time is estimated as the difference between the arrival times for successive push intervals (pseudo-interval). However, a more accurate set-up is to use two sets of sensors located 1 m apart in the cone to record the shear wave arrive at both locations simultaneously (true-interval). In recent years, full-flow penetrometers such as the T-bar and ball have started to be used in offshore practice. Because they allow the soil to “flow” around the device, the vertical stress correction (as required for full displacement devices such as the CPTU) is minimal and with an area typically 10 times that of the

– The CPTU is recommended as the main tool to be used for in situ testing. It is the best tool for soil profiling and identification of soil behaviour type. It can be used for estimating σp and su but is not reliable for estimating sur . – The seismic CPTU should be used more often to measure Vvh (ideally using true-interval equipment), which can be used to estimate Gvh . – Full-flow penetrometers have proved to be as reliable as the CPTU for estimating su of soft clays and in some circumstances are better, e.g. shallow test depths and/or very soft sediments. They are the best option for measuring soil shear degradation and estimating sur . Both penetration and extraction resistance should always be measured when conducting full-flow tests. – The FVT remains a good tool for estimating su and sur (provided the soil is remoulded through multiple fast-rotations of the vane). However, the CPTU and full-flow penetrometer tests are quicker to perform and give a continuous profile of penetration resistance. 5.5

Determining the in situ pore water pressure condition is critical to conduct of a reliable site characterisation program. As described in Section 3.1, there are a number of offshore regions where overpressured zones


Measurement of in situ pore water pressure

exist and have caused significant problems. Simply assuming hydrostatic pore water pressure conditions without verification is a risk. There are two options available at present for directly measuring in situ pore water pressure: piezoprobes and piezometers. 5.5.1 CPTU and piezoprobes The CPTU can be used to measure in situ equilibrium pore water pressure (u0 ) during a pause in penetration. Excess pore pressures that are generated during penetration will eventually dissipate back to u0 with the rate depending on cv . In clays with a low cv , this can take several days for the standard diameter CPTU (d = 36 mm) which is not practical for most site investigations. Since the rate of dissipation is proportional to d2 , small diameter piezoprobes have been developed to speed up dissipation. These probes consist of a tapered section that narrows down to a diameter of about 5 to 8 mm with a filter element near the tip. However, Whittle et al. (2001) showed that the pore pressure developed by the regular diameter push rod, to which the piezoprobes are attached, eventually reaches the smaller diameter tip and influences the dissipation time (and it can take as long as with a regular diameter CPTU). Whittle et al. (2001) thus developed an interpretation method that uses data from a dual element piezoprobe which consists of two pore pressure filters, one at the small diameter tip and one at the normal location on the shoulder of the cone (Figure 20a). The simultaneously recorded dissipation data are interpreted to estimate u0 . Piezoprobes have successfully been used in site investigations but unresolved issues remain. The lengthy, small diameter extension, which ideally should be as long as possible to reduce influence of the larger diameter push rod, is vulnerable to damage during penetration. When deployed in vessel drilling mode, imperfect heave compensation of the drill string can produce erratic results that are difficult to interpret. The use of a dual-element piezoprobe and the corresponding interpretation theory of Whittle et al. (2001) are still been researched (e.g. Flemings et al. 2008) and have not yet been validated for use in commercial practice.

Figure 20. a) T2P dual element piezoprobe (from Flemings et al. 2008), and b) example of multi-piezometer string installed in borehole (from Strout & Tjelta 2007).

Figure 21. Pore water pressure measurement options for piezometers (from Strout & Tjelta 2007).

and the soil formation. This is especially important for shallow depth installations. Three methods can be used to measure the in situ pore pressure at the depth of installation of the filter element (Figure 21, Strout & Tjelta 2007): a) direct measurement of u0 at the filter; b) differential pressure measurement at the filter relative to the seabed; and c) differential pressure measurement at the seabed. Practical considerations include saturation of the connecting fluid column for differential pressure measurements and the density of this fluid column needs to match that of the in situ pore water, sensors must be protected against corrosion, and the logistics of collecting data from the instrumentation package. Recommendations: Strout & Tjelta (2007) conclude that the most reliable method for measuring in situ pore pressure in low permeability sediments is to use piezometers. The key advantage is that they provide time history data. The cost is not trivial and new

5.5.2 Piezometers Piezometers can now be installed as part of a soil investigation in a predrilled borehole (i.e. drilling mode) or via non-drilling mode using a seabed push frame. In drilling mode, the creation of a borehole allows for the installation of a single piezometer or a string of piezometers. The piezometers are grouted in place and an instrumentation package is connected to the top of the string at the seabed (Figure 20b). An important practical consideration is the effectiveness of the grout, which must create a stable hydraulic seal around the piezometers with a hydraulic conductivity that is equal to or less than that of the formation soil. For push-in piezometers, the push stroke must be straight and concentric so as to not potentially create an open void, and thus a short circuit between the piezometer rise pipe


Table 6.

solutions are being developed and this effort needs to continue. Seabed sediment overpressures was the focus of a 2009 workshop (Sheahan & DeGroot 2009) that identified challenges in dealing with overpressures. For current practice, workshop recommendations were made for measurement and installation options as follows:

Soil sampling equipment.

Deployment

Sampler and diameter (mm)

Non-drilling, DWS 110 seabed mode

– for shallow depth (50

Piston relative to vessel, problems with recovery and sample quality

Box corer typically 500 mm square

0.5–1.0

Miniature in situ tests can be conducted in box upon retrieval, subsamples can be collected

Seabed drilling

Piston or push, 44/75 Percussion or rotary, 75

75–150

Depth depends on system casing/rod storage capacity with wireline systems thus far being able to go deepest

Vessel based drilling

Piston, push, >100 or percussion 75

Advanced by push, drive or percussion using mud pressure or hydraulics with power from an umbilical.

5.6 Soil sampling, preservation and storage 5.6.1 Sampling methods Table 6 lists the different types of soil sampling equipment presently available. Section 2.2 discussed the issue of sample disturbance and its impact on the reliability of laboratory measurement of design parameters such as σp and su . When such advanced laboratory tests are to be conducted it must be a key objective of the site characterisation program to use a sampler that can collect good quality samples. Lunne & Long (2006) outlined the key features that an offshore sampling system should have in order to have the potential to collect good quality samples. These include (Figure 22): sharp cutting shoe, stationary piston relative to the seabed, minimal inside and outside friction, small area ratio, sample diameter of 100 to 120 mm, and a retractable core retainer. In terms of deployment, the sampler should be pushed at a steady rate of ∼20 mm/sec, the push rate should be measured, and the underpressure developed behind the piston should also be measured. The Deep Water Sampler (DWS) was developed with the features displayed in Figure 22 as a goal for its design and deployment (Lunne et al. 2008). It has thus far been used on several onshore and offshore investigations, including the Troll site (Norwegian North Sea), and has shown a capability to collect good to


Figure 22. Design features of ideal sampler for soft to medium-stiff clays (from Lunne & Long 2006).

high quality samples (Lunne at al. 2008). The depth of these investigations was less than 10 m and its good performance remains to be verified for the design full penetration depth of 20 to 25 m. The STACOR gravity sampler (Fay et al. 1985) uses a piston that is fixed relative to a small frame that sets-down on the seabed during deployment. Thus if the frame remains stationary during sampling then so

48

does the piston. Penetration is not directly measured. Numerous cases have reported collecting good quality samples using the STACOR sampler (e.g. Borel et al. 2005). The Kullenberg type of sampler also penetrates the seabed by gravity and can collect very long samples of up to 50 m or more, which can be an efficient way of collecting long sample cores. However, there are problems with this type of sampler including uncertain penetration depth and low recovery, and it has been found that sample quality tends to decrease with depth. These latter two problems are largely a result of the use of a thick walled, blunt tipped, cutting shoe and the fact that the piston is fixed with a wire to the vessel and not positioned relative to the sea bottom. Downhole sampling in vessel based drilling mode has long been shown to be able to collect good quality samples providing a fixed piston sampler is used in conjunction with a thin walled sample tube having a sharp cutting edge (e.g. 5◦ for soft clays), small area ratio and zero inside clearance ratio, and furthermore that the drilling is carefully conducted with an efficient heave compensation system. As noted in Section 5.3, several multipurpose seabed drilling systems have been developed. For the PROD system, which uses a drill string to deploy a sampler, potential sampling depth depends on the number of rods the carousel can hold (which is a function of the diameter casing being used). Box cores are good for characterizing the upper 0.5 m of the seabed. Common box cores normally collect a 0.5×0.5×0.5 m sample and can be tested immediately after retrieval to the vessel. For soft soils, miniature laboratory vane and full-flow probes can be used to test the sample while it is still in the box (e.g. Low et al. 2008). Because the effective stress of these surficial sediments is quite low, it is generally not possible to conduct reliable advanced laboratory tests on these samples; thus, the use of these miniature probes is a good, practical solution. Figure 23 shows an example of such data recorded for a soft seabed sediment. Subsamples can be collected for follow-on onshore laboratory testing. The large volume of collected soil can be saved from multiple box cores and resedimented in the laboratory for scale model testing of shallow soil structure interaction problems as pipelines and risers (e.g. 1g model tests described by Andersen et al. 2008). Recommendations: Handling of long recovery samples is difficult and the vessel must be equipped to handle such samples without disturbance. Box core testing using miniature full-flow probes and MLV is recommended for characterizing the upper (∼0.5 m) sediments for shallow depth investigations. Tube samplers are ranked as follows relative to their potential to obtain as good a quality sample and as high a recovery ratio as possible:

Figure 23. a) Penetration, extraction and cyclic resistance versus depth for miniature T-bar test conducted in box core sample of a soft seabed sediment. b) degradation plot for cycle 2.

– For non-drilling mode seabed based sampling: 1) piston sampler fixed relative to seabed, favorable geometry, steady penetration with real time measurement of penetration, and accurate measurement of recovery (e.g. Figure 22, DWS), 2) gravity sampler with piston fixed relative to seabed (e.g. STACOR), 3) gravity sampler but piston fixed relative to vessel (e.g. Kullenberg), and 4) gravity sampler without fixed piston. 5.6.2 Sample handling and storage Sample handling and storage must avoid handling disturbance and must maintain sample moisture and temperature conditions. Samples should not be subject to excessive temperature changes from warm climates, overheated wax and especially freezing temperatures. This is especially important for all samples that may be used for advanced laboratory testing (Section 5.7) Evidence suggests that soft clay samples can survive commercial transport when properly protected (and not subjected to catastrophic impact such as dropping of the sample box). For example, Figure 24 plots results of constant rate of strain consolidation (CRS) tests performed on specimens trimmed from Sherbrooke block samples of the Onsøy clay at NGI in Oslo, Norway and again at UMass Amherst, USA after commercial air transport.

– For downhole sampling in drilling mode (vessel or seabed based): 1) thin walled fixed piston sampler with favorable geometry as noted above; 2) thin walled push sampler (although sampling without a fixed piston is not recommended).


49

Table 7.

Figure 24. Comparison of CRS tests conducted on Sherbrooke Block samples of Onsøy clay at NGI, Oslo, Norway and again after air transport to UMass Amherst, USA (from DeGroot et al. 2009).

Recommendations: Sample tubes/liners should be sealed with a mechanical seal and/or with wax and capped. The samples should be stored vertically and should not be able to move inside the sample tube or liner. Soft clays should be kept in the tube or liner for transport to the onshore laboratory. For stiffer clays, the sample may be extruded in the field and sealed in a cardboard cylinder using plastic, aluminum foil and wax. Samples should as quickly as possible be placed in a constant temperature controlled container that has been set at the estimated in situ temperature. Samples should be transported upright in containers that damp shock and vibration. Once at the onshore laboratory, samples should be stored in a container or room with humidity and temperature control.

Test Category

Test Types

Comments

Classification and Basic Index Testing

– – – – –

– simple and relatively quick to perform – necessary part of any site characterisation program – cannot provide design parameters

Express/Index Strength Testing Parameters measured: su , sur

– pocket penetrometer – fall cone – torvane – laboratory vane, – unconfined compression – unconsolidated undrained triaxial

– simple and relatively quick to perform – equipment commonly available – often gives scattered results – successful use in design requires soil/ site specific correlations

Advanced Laboratory Testing Parameters measured: CR, Cαε , σp , cv , kv , su , sur , St c , φ , Gmax

– IL and CRS Oedometer – CAUC, CAUE & DSS – ring shear – permeability – resonant column – cyclic CAUC & DSS – bender elements

– provides best control of soil state and conditions in the laboratory – relies on good quality samples – provides direct measure of design parameters – equipment more complex but automation very efficient

water content density Atterberg Limits grain size specific gravity

Advanced laboratory tests, as listed inTable 7, are an essential component of any site characterisation program requiring design parameters for compressibility, consolidation/flow, and strength behaviour. The need for such testing is highlighted in Table 3 for the various design concepts and geohazards listed. Section 5.7.3 to follow gives an overview of recommended scope and methods for conduct of advanced laboratory test programs. For some projects, geochemical an dating tests may also be required for assessing the geologic origin and history of sediments.

5.7 Laboratory testing Table 7 divides laboratory tests into three primary categories: classification and basic index testing, index strength tests, and advanced laboratory tests. Every design concept will first and foremost require classification of the various soil units encountered and such testing should always be conducted as part of any site investigation. The express/index strength tests listed in Table 7 are popular because they are generally quick, easy to perform and can readily be conducted offshore. However, it is important to note that these tests generally use fast shear rates, different modes of shear, and test small soil volumes, and the results are greatly affected by sample disturbance. Undrained shear strength profiles developed using these devices often show significant scatter. Therefore, the data from these devices represent, at best, relative strengths rather than values suitable directly for design. They should be relied upon only to indicate the general consistency of soil layers. Reliable determination of su values for design should focus on appropriate in situ tests and use of laboratory equipment that can conduct consolidated-undrained (CU) tests.

5.7.1 Field laboratory For logistical reasons, offshore laboratory testing is usually limited to soil description, classification and index strength tests. In some cases it can be advantageous to conduct CRS tests offshore in order to get an early indication of stress history, which provides a preliminary indication of the soil state and aids in planning the advanced laboratory test programme. The CRS data can also be used to evaluate sample quality as described in Section 5.7.2 to follow. If box cores are collected they should also be tested offshore as described in Section 5.6.1. 5.7.2 Evaluation of sample quality While no definitive method exists for determining sample quality, valuable information can be obtained


Laboratory tests for fine grained soils.

Table 8. Evaluation of sample quality for low to medium OCR clays using Lunne et al. (2006). e/e0 at in situ stresses for sample qualities 1 to 4 OCR

1 (very good to excellent)

2 (fair to good)

3 (poor)

4 (very poor)

1 to 2 2 to 4

< 0.04 < 0.03

0.04 – 0.07 0.03 – 0.05

0.07 – 0.14 0.05 – 0.10

> 0.14 > 0.10

from making use of both qualitative and quantitative methods. Visual observations on the vessel during sample processing and testing in the offshore laboratory can provide useful qualitative information on sample quality (e.g. evidence of cracking). Qualitative (visual) assessment of sample quality is best made by examination of sample X-rays. Radiographic methods allow for non-destructive visual evaluation of sample quality, layering and the presence/quantity of stones, cobbles and other inclusions. It is also possible to obtain computerized tomography (CT) scans of soil samples that are contained in non-metallic sampling tubes or liners. CT scans provide a visual image along the length of a sample and cross-sectional images.The sample radiographs and/or CT scans should be used to select the location within a tube or liner sample of the most critical advanced laboratory tests. Quantitative assessment of sample quality for intact, low to medium OCR clays, can be done by measuring volume change during laboratory recon solidation to the estimated in situ stress state (σv0 , σh0 ). The normalised sample quality parameter e/e0 of Lunne et al. (2006) is computed as

Figure 25. Example application of e/e0 sample quality evaluation for three samplers used at Troll field (from Lunne et al. 2008).

where e = change in void ratio during reconsolidation to in situ stresses, e0 = initial void ratio, and εvol = volumetric strain (= V/V0 ) from reconsolidation to in situ stresses. These data are available from 1-D consolidation tests (IL and CRS) and the consolidation phase of CU strength tests (triaxial and DSS). The corresponding sample quality is determined using Table 8. Figure 25 plots an example application for samples collected from the Troll field using the DWS, borehole drilling with tube sampling, and a gravity core sampler. The technique of evaluating sample quality using shear wave velocity continues to be developed (e.g. Landon et al. 2007, Donohue & Long 2007). The method involves measuring Vvh of a short sample section using bender elements mounted in a jig such as the one shown in Figure 26. The advantage of this method is that by using small strain shear waves generated by the bender elements, the measurement is nondestructive and can be performed immediately after sample collection. Measurements can also be repeated on the same samples again in the onshore laboratory to potentially track any additional disturbance from sample sealing, transport and storage.

Figure 26. Schematic of device for combined sample shear wave velocity and soil suction measurements.

Figure 27 shows the proof of concept for terrestrial soils by comparing Vvh normalised by the in situ Vvh , in this case measured with downhole seismic CPTU, versus e/e0 . Ideally this method could be used to quickly and non-destructively screen samples on the vessel and again later on in the onshore laboratory prior to making decisions on selection of samples for


that are important to proper and reliable conduct of these tests. Lunne & Andersen (2007) describe an approach used at NGI for large testing programs to optimize the use of in situ and laboratory testing, and when su anisotropy data are required. A brief summary is as follows: 1) select a key representative borehole and conduct sufficient number of CAUC tests on the best quality samples to establish a reliable su (CAUC) profile, 2) in the same borehole conduct some CAUE and DSS tests to establish anisotropy ratios, 3) correlate su (CAUC) values with the results of CPTU and/or full-flow tests to develop a site specific correlation for the continuous in situ data (generic correlations for these in situ tests are discussed in Section 5.8), and 4) if possible conduct a few CAUC and DSS tests in other borings. When selecting the location of a test specimen, soil from within 1 to 1.5 times the tube/liner diameter from the tube top and bottom should not be used because of greater disturbance near the sample ends (Lacasse & Berre 1988). Sample sides should be trimmed during specimen preparation to remove potentially disturbed material. Sub-sampling should not be used for preparation of final specimen dimensions. Very soft clays will need to be supported during trimming and mounting without touching the specimen by hand during preparation. The best test to measure compressibility, flow and σp is the CRS. The CRS test is preferred over incremental loading (IL) consolidation tests because of the ability for continuous measurement of deformation, vertical load, and pore pressure for direct calculation of the stress-strain curve, kv and cv . The loading strain rate should be selected such that normalised base excess pore pressure (ub /σv ) is less than about 15% in the normally consolidated stress range. Sandbækken et al. (1986) report that rates of 0.5 to 1%/hr are adequate for most clays. Experience at NGI indicates that σp from CRS tests gives a value about 10 to 15% greater than that determined from IL tests that use 24 hr load increments. The CRS tests should be conducted as early as possible. Stress history information is important for preliminary assessment of various design concepts and is also for conduct of the CAUC/E and DSS strength tests. The CRS results will also give a direct indication of sample quality via the e/e0 method described in Section 5.7.2 and Table 8. Evaluation of undrained shear strength (su ) anisotropy is often a critical aspect of many offshore design concepts as discussed in Section 2.1.2 and listed in Table 3. In the laboratory, this is best evaluated using a combination of CAUC, DSS and CAUE test. This is especially important for low to medium plasticity soft clays for which the undrained shear strength anisotropy is typically the most significant (e.g. Figures 2–4). The rate of strain for undrained shear should be selected to account for strain rate sensitivity of clays and typical field loading rates. Recommended rates include 0.5 to 2.0% per hour for triaxial tests and 5%/hr shear strain for DSS (Germaine & Ladd 1988, Lacasse & Berre 1988).

Figure 27. Normalised sample shear wave velocity versus CRS measured e/e0 for samples of three soft clays (from DeGroot et al. 2007).

advanced laboratory testing. Once samples are tested, confirmation e/e0 data can be collected during the consolidation phase. The method does require in situ Vvh , with downhole seismic CPTU being the best option (Section 5.4.2). Measurement of sample suction may be used as an additional refinement of this non-destructive sample evaluation scheme. Suction in a clay sample is the negative equivalent of the sampling effective stress (σs in Figure 7) and its value relative to σv0 is an indication of sample disturbance. Measuring both Vvh and σs could be used in the framework shown in Figure 8 or that presented by Donohue & Long (2007) to assess sample quality. Sample suction can be measured using a portable suction probe such as that described by Ridley & Burland (1993). The suction probe in Figure 26 at the base of the bender element jig is based on that of Poirier & DeGroot (2010). While these non-destructive methods of evaluating sample quality show promise, more research and development is needed before it can be recommended for regular use in offshore practice. 5.7.3 Advanced laboratory test procedures The most important advanced laboratory tests for characterisation of clays include the CRS test for measurement of compressibility, flow and stress history, and CAUC, DSS, and CAUE for measurement of stress-strain-strength behaviour and anisotropy. Other measurements listed in Table 3 such as small-strain behaviour, remoulded behaviour, and cyclic response are also important in many cases but the CRS and CAUC/E and DSS tests are critical to almost all applications. The following discusses procedure issues


Ideally triaxial and DSS specimens should be anisotropically consolidated to the estimated in situ effective stress state (σv0 , σh0 ), i.e. the Recompression method (Bjerrum 1973). In the absence of measured K0 data, the Brooker & Ireland (1965) correlation amoung K0 , PI and OCR can be used to estimate K0 and compute the laboratory consolidation stress σhc ∼ σh0 . The quality of su data from Recompression tests depends on sample quality. Increasing sample disturbance can destroy structural bonds and subsequent loss in measured su ; however, large volume changes during reconsolidation can result in an increase in measured su . If sample quality is found to be poor to very poor then consideration should be given to conducting SHANSEP (Ladd 1991) tests, which involves laboratory K0 consolidation to a OCR = 1 state of stress and then to various mechanically overconsolidated stress states. Otherwise the SHANSEP testing protocol (e.g. evaluation of anisotropy, shear rates) is the same as that of the Recompression method. The advent of reliable automated triaxial stress path cell systems greatly enhances the efficiency of conducting K0 consolidation. SHANSEP triaxial and DSS test data for each major soil unit are used to evaluate the S and m parameters for Equation 1 and the best estimate OCR profile is used in Equation 1 to compute the in situ su profile. For under-consolidated and normally consolidated clays, OCR = 1 SHANSEP testing is the only viable option. Ladd (1991) and Ladd & DeGroot (2003) provide further details on the Recompression and SHANSEP methods and describe the advantages and disadvantages of both. For laboratory measurement of sur , the motorized laboratory vane (MLV) is the most versatile device. It covers a large range of shear strength, can use different rates, and can measure shear resistance of samples in various states. Several measures of the sur are possible using the MLV as discussed in Section 2.3 and it is important that the method(s) required or used be clearly documented. The fall cone (FC) is also recommended to measure sur . While not as versatile as the MLV, there is a large amount of experience with the FC, it can cover a large range of shear strength (through the use of different size cones), and uses only a relatively small volume of soil. However, calibration factors for converting FC penetration to sur differ among standards (e.g. ISO, Norwegian, and Swedish standards use different factors), and thus the calibration factor used should be documented. Measurement of sur using UUC tests is not recommended if sur is less than about 5 kPa because the tests are difficult to perform and the membrane and area corrections are significant. The UUC test is better suited for stiffer soils and has the advantages of being suitable for investigating rate effects and to investigate sur anisotropy. Recommendations: Major recommendations for conduct of advanced laboratory tests include:

– it is essential to evaluate sample quality (Table 8) and the e/e0 value should be reported for all CRS (or IL) and consolidated strength tests – 1-D compressibility, flow, and σp are best determined using CRS tests – CAUC/E and DSS tests should be used for measurement of stress-strain-strength behaviour and su anisotropy. For good quality samples, specimens should be consolidated to the estimated in situ effective stress state. – for poor to very poor quality samples the triaxial and DSS test program should involve conduct of SHANSEP tests. SHANSEP testing should be used for under-consolidated and OCR = 1 soils. – sur is best measured using the MLV, and the method of remoulding must be specified. Companion tests may be run using the FC. – Andersen et al. (2008) give recommendations for cyclic, thixotropy, creep, and reconsolidated remoulded testing and present example results. 5.8

5.8.1 Stress history recommendations profiles are developed using soil total unit In situ σv0 weight measurements and estimated in situ pore pressures. In the early stages of a site investigation, it is common to assume that the in situ pore pressure profile is hydrostatic with a sea water unit weight γw = 10.1 kN/m3 . However, measurement of in situ pore pressure (Section 5.5) and pore fluid salinity will be required in regions with non-hydrostatic pore pressures and high salinity. There are many methods of estimating σp using CRS or IL data and the K0 consolidation phase of SHANSEP tests with no method being definitively better than any other. The authors typically use two to three methods including Casagrande, constrained modulus (Janbu 1969) and strain energy (Becker et al. 1987) and report values from all methods in the factual laboratory report. The Andresen et al. (1979) su /σv0 , PI and OCR correlation is also often used. CPTU data can be used to estimate σp using the correlation (Lunne et al. 2007b).

where qnet = CPTU net penetration resistance and k is typically within the range of 0.25 to 0.35. These data should be used with the laboratory σp values and geologic history information to develop a best estimated in situ σp profile. This profile should then be used to compute the corresponding OCR profile. If the

– samples should be X-rayed to check for features such as layering and visual signs of disturbance


Data evaluation and interpretation

The following presents recommendations for data evaluation and interpretation with a focus on stress history and su . The evaluation of sample quality should be used to determine if σp and su . from the advanced laboratory tests are reliable. Generally it is desired to use data from tests conducted on samples of e/e0 quality rating 1 or 2 (Table 8).

Table 9. Recommended soft clay CPTU and T-bar N-factors for evaluation of su and sur (modified after Low et al. 2010). N-factor N factor, Nrem factor Nkt,CAUC Nkt,suave Nu NT-bar,CAUC NT-bar,suave NT-bar,rem,UU NT-bar,rem,fc NT-bar,rem,MLV

Definition

Mean

Range

qnet /suCAUC qnet /suave or qnet /suDSS (u2 – u0 )/suCAUC qT-bar /suCAUC qT-bar /suave or Tbar /suDSS qT-bar,rem /sur,UU qT-bar,rem /sur,fc qT-bar,rem /sur,MLV

12.0 13.5

10.0–14.0 11.5–15.5

6.0 10.5 12.0

4.0–9.0 8.5–12.5 10.0–14.0

20.0 14.5 14.0

13.0–27.0 12.5–16.5 12.0–16.0

Notes: su N values for clays with St < 8. Use NBall = NT-bar . Table 10. Typical SHANSEP S and m values for su (ave) (modified after Ladd and DeGroot 2003). Soil Description

S

m

Sensitive cemented marine clays (PI < 30%, LI > 1.5) Homogeneous sedimentary clays of low to moderate sensitivity (PI = 20 to 80%) Sedimentary silts and organic soils (Atterberg Limits plot below A-line) and clays with shells Lacustrine varved clays

0.20

1.0

0.22

0.8

0.25

0.8

0.16

0.75

Figure 28. Undrained shear strength data for low OCR soft clay site, Haltenbanken area of the Norwegian Sea.

as those presented in Figures 2–4, are also useful in evaluating the data. For low OCR soils, reference lines of minimum expected su ratios should be plotted, i.e. assume OCR = 1 in Equation 1 and select an S value for the su of interest [e.g. Table 10 presents typical S and m values for su (ave)]. The recommendations given in this paragraph are for evaluating su data and not intended to be a substitute for in situ and laboratory testing. Figure 28 plots an example data set for a lightly overconsolidated, soft, clay site in the Haltenbanken area of the Norwegian Sea. Site and soil layer specific CPTU Nkt factors were developed in reference to the su (CAUC) tests which were conducted on good quality samples. The most notable aspect of the data is the scattered su values from the strength index tests and that many of these values are very low. This result is represents a lower a common occurrence. The 0.28σv0 bound su (CAUC) reference line assuming OCR = 1 ). (i.e. Figure 2 with σp = σv0

Note: PI = plasticity index, LI = liquidity index

σp data are considered reliable, then a site specific k coefficient should be developed. 5.8.2 Undrained shear strength recommendations The CPTU and full-flow penetrometer data can be converted to su using an appropriate N factor

N values are specific to the reference su and in situ test conducted. Low et al. (2010) and DeJong et al. (2010b) present recommended N factors for interpretation of CPTU, T-bar, and ball data for soft clays. Table 9 presents the Low et al. (2010) factors, which were developed using a database that included several offshore sites. The laboratory su data should be plotted together with that estimated from the in situ tests using the N factors in Table 9. Depending on the quality of the laboratory su data, site specific N values should be developed. The su profiles should always be evaluated in conjunction with the stress history data. As described in Section 2.1.1, su and σp are strongly linked and Equation 1 can be used to help in evaluating the consistency of the su profiles relative to the σp profile. Andresen et al. (1979) present correlations amoung su /σv0 , PI and OCR based on Equation 1. Database correlations, such

6

Many innovative systems and test methods have been developed for geotechnical characterisation of offshore sediments. Significant progress has been made in implementing improved deployment systems, in situ tools, and samplers in practice. However, challenges remain, some of which are a matter of convincing practice to implement more widely what are already known best practice methods. For example, sample quality is


PRESENT AND FUTURE CHALLENGES

so critical to the reliability of laboratory data that the use of proper sampling tools must be made a priority. Another example is project specifications, which must be developed with an understanding of appropriate procedures and standards to specify. This largely involves clear communication amoung all parties and use of proper QA/QC programs. Universal adaptation of the proposed ISO standard on Marine Soil Investigations will provide a valuable reference framework. At the other end of the spectrum, some challenges are more daunting, such as for example developing geophysical techniques that yield reliable soils parameters. Other important challenges and related questions include:

the quality and reliability of in situ and laboratory test data. The paper describes these phases and gives best practice recommendations for each. Some of the most important recommendations include: accurate depth control relative to the seabed during deployment of in situ and sampling tools, CPTU and full-flow penetrometers for in situ testing, piezometers for measurement of in situ pore pressure, non-drilling seabed sampling with a piston sampler that is fixed relative to the seabed and with an appropriate geometry for sampling of cohesive sediments, evaluation of sample quality, and conduct of CRS, CAUC/E and DSS tests to measure consolidation, stress history and anisotropic stress-strain-strength behaviour.

– implementation of new seabed based drilling systems is an exciting development and the innovation needs to continue. Their use should be more widespread, especially for deep waters. – implementation of hydrostatically compensated CPTU for deep water investigations and increased use of seismic CPTU to obtain small strain shear modulus profiles. – in situ pore pressure u0 : 1) piezometers should become standard practice for projects where knowledge of u0 is essential, 2) develop inexpensive reliable drop-in piezometers, 3) develop reliable indirect indicators of overpressures, e.g. could geophysical measurements be used? – characterisation of the upper 1 to 2 m of sediment, especially in deep waters. Is box core testing good enough and how to evaluate the quality of a box core sample? How accurate are in situ tools in characterizing this surficial zone? – in situ testing of intermediate soils such as silts that can undergo partial drainage during testing. Can variable rate testing become a reliable and practical approach for offshore investigations? – do ambient pressure and temperature samplers need to be used more often or can we compensate for stress relief and temperature change in laboratory measurements? 7

ACKNOWLEDGEMENTS The authors thank their colleagues at UMass Amherst, NGI and Statoil for their contribution to the numerous research and consulting projects from which results and findings are presented in the paper, and Thomas Sheahan who reviewed the manuscript.The first author thanks NGI and the University of Western Australia for providing sabbatical opportunities to work on offshore site characterisation and acknowledges the US National Science Foundation for its support on grant OISE-0530151.

REFERENCES Allen, J.D., Hampson, K., Clausen, C.J.F., & Vermeijden, C. 2005. Well deformations at West Azeri, Caspian Sea. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis, 999–1004. Andersen, K.H. 2004. Cyclic clay data for foundation design of structures subjected to wave loading. Proc. Int. Conf. on Cyclic Behaviour of Soils and Liquefaction Phenomena, CBS04, Bochum, Germany, 371–387. Andersen, K.H. 2009. Bearing capacity under cyclic loading – offshore, along the coast and on land. 21st Bjerrum Lecture presented in Oslo, 23 November 2007. Can. Geotechnical J. 46: 513–535. Andersen, K.H.A., Lunne,T., Kvalstad, T. & Forsberg, C.F. 2008. Deep water geotechnical engineering. Proc. XXIV Nat. Conf. of the Mexican Soc. of Soil Mechanics, Aguascalientes, 26–29 November, 2008. Andresen, A., Berre, T., Kleven, A. & Lunne, T. 1979. Procedures to obtain soil parameters for foundation engineering in the North Sea. Marine Geotechnology, 3(3): 201–66. Aubeny, C.P. & Shi, H. 2006. Interpretation of impact penetration measurements in soft clays. J. Geotech. and Geoenvironmental Eng. 132(6): 770–777. Becker, D. E., Crooks, J.H.A., Been, K. & Jefferies, M.G. 1987. Work as a criterion for determining in situ yield stresses in clays. Can. Geotechnical J. 24: 549–564. Bjerrum, L. 1973. Problems of soil mechanics and construction on soft clays. Proc. 8th Int. Conf. on Soil Mech. and Foundation Eng., Moscow, 3: 111–159. Boggess, R. & Robertson, P.K. 2010. CPT for soft sediments and deepwater investigations. Proc. 2nd Int. Symp. on Cone Penetration Testing, Los Angeles. 9–11 May 2010.

CONCLUDING REMARKS

This paper presented best practice recommendations for geotechnical characterisation of offshore cohesive sediments. The particular focus has been on clays which have a complex mechanical behaviour, are not easily characterized, and are vulnerable to irreversible disturbance when sampled. This complexity is exacerbated for offshore investigations because of the many unique and challenging conditions that are present offshore. Many tools and methods have been developed to meet these challenges. Competent geotechnical site characterisation programs should combine the best of in situ testing and laboratory testing, i.e. using well calibrated in situ tools and collection of high quality undisturbed samples for advanced laboratory testing. Each phase of the site characterisation process, from drilling to evaluation of laboratory data, can influence


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Borel, D., Pench, A., Dendani, H. & Colliat, J.L. 2005. Deepwater geotechnical site investigation practice in the Gulf of Guinea. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis, 921–927. Brooker, E.W. & Ireland. H.O. 1965. Earth pressures at rest related to stress history. Can. Geotechnical J. 2(1): 1–15. Bugge, T. 1983. Submarine Slides on the Norwegian Continental Margin, with special emphasis on the Storegga area. IKU, Publication 110, 152 p. Carter, J.P., Davies, P.J., & Krasnostein, P. 1999. The future of offshore site investigation – robotic drilling on the sea-bed. Australian Geomechancis. 34(3): 77–85. Collinson, J. D. & Thompson, D. B. 1982. Sedimentary Structures. George Allen & Unwin, London, U. K. DeGroot, D.J., DeJong, J.T., Yafrate, N.J., Landon, M.M. & Sheahan, T.C. 2007. Application of recent developments in terrestrial soft sediment characterization methods to offshore environments. Proc. OTC, Houston, Paper18737MS. DeGroot, D.J., Ladd, C.C. & Germaine, J.T. 1996. Undrained multidirectional direct simple shear behavior of a cohesive soil. J. of Geotechnical Eng. 122(2): 91–98. DeGroot, D.J., Landon, M.M. & Lunne, T. 2008. Synopsis of recommended practice for sampling and handling of soft clays to minimize sample disturbance. Proc. 3rd Int. Conf. on Site Characterization, Taipei, Taiwan. DeJong, J.T., Yafrate, N.J. & DeGroot, D.J. 2010b. Evaluation of undrained shear strength using full-flow penetrometers. J. of Geotechnical and Geoenvironmental Eng. In press. DeJong, J.T., Yafrate, N., DeGroot, D.J., Low, H.E., & Randolph, M. 2010a. Recommended practice for full-flow penetrometer testing and analysis. Geotechnical Testing J. 33(2): DOI: 10.1520/GTJ102468. Donohue, S. & Long, M. 2007. Rapid determination of sample quality using shear wave velocity and suction measurements. Proc. Conf. on Soil Invest. and Found. Behaviour, SUT, London, Sept. 2007, 63–72. Dugan, B. Sheahan, T.C., Thibault, J.M. & Evans, T.G., 2010. Offshore sediment overpressures: overview of mechanisms, measurement and modeling. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. In press. Ehlers, C. J., Chen, J. Roberts, H. H., & Lee, Y. C. 2005. The origin of near-seafloor “crust zones” in deepwater. In Gourvenec S. & Cassidy M (eds). Proc. Int. Symp. on Frontiers in Offshore Geotechnics. Taylor & Francis. 927–933. Evans, T. 2010. A systematic approach to offshore engineering for multiple-project developments in geohazardous areas. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. ISFOG 2010. In press. Fay, J.B., Montargès, R., Le Tirant, P., & Brucy, F. 1985. Use of the PAM self-boring pressuremeter and the STACOR large sized fixed piston corer for deep seabed surveying. Proc. Conf. Offshore Site Invest., SUT, London, 187–199. Flemings, P.B., Long, H., Dugan, B., Germaine, J., John, C., Behrmann, J.H., Sawyer, D., & Expedition 308 Scientists, 2008. Pore pressure penetrometers document high overpressure near the seafloor where multiple submarine landslides have occurred on the continental slope, offshore Louisiana, Gulf of Mexico. Earth and Planetary Science Letters, 269(3-4): 309–32. Germaine, J.T. & Ladd C.C. 1988. State-of-the-Art: Triaxial testing of saturated cohesive soils. Advanced Triaxial Testing of Soil and Rock, ASTM STP 977, 421–459. ISSMGE 2005. Geotechnical and geophysical investigations for offshore and nearshore developments. Int. Society for Soil Mechanics & Geotechnical Engineering. Handbook.


Lunne, T. & Andersen, K.H. 2007. Soft clay shear strength parameters for deepwater geotechnical design. Proc., 6th OSIG, SUT, London, UK, 151–176. Lunne. T., Andersen, K.H., Low, H.E., Randolph, M. & Sjursen, M. 2010. Guidelines for offshore in situ testing and interpretation in deepwater soft clays. Can. Geotechnical J., In press. Lunne, T., Berre, T., Andersen, K.H., Strandvik, S. & Sjursen, M. 2006. Effects of sample disturbance and consolidation procedures on measured shear strength of soft marine Norwegian clays. Can. Geotechnical J., 43: 726–750. Lunne, T., Berre, T. & Strandvik, S. 1997a. Sample disturbance effects in soft low plastic Norwegian clay. Proc. Recent Developments in Soil and Pavement Mechanics. Rio de Janeiro, Brazil, 81–102. Lunne, T., T. Berre, S. Strandvik, K.H. Andersen & T.I. Tjelta 2001. Deepwater sample disturbance due to stress relief. Proc. Int. Conf. on Geotechnical, Geological and Geophysical Properties of deepwater Sediments. College Station, Texas, 64–85. Lunne, T., Isa, O. & Tan, M. 1996. Shallow gas problem at Duyong B offshore Malaysia. Proc. 11th Offshore South East Asia Conference. Singapore. Lunne,T. & Long, M. 2006. Review of long seabed samplers and criteria for new sampler design. Marine Geology. 226: 145–165. Lunne, T., Robertson, P.K., & Powell, J.J.M. 1997b. Cone Penetration Testing in Geotechnical Practice. Spon Press, London. Lunne, T., Tjelta, T.I., Walta, A. & Barwise, A. 2008. Design and testing out of deepwater seabed sampler. Proc. Offshore Technology Conf. Houston, USA, Paper 19290. Mesri, G. (1975). Discussion of ‘new design procedure for stability of soft clays. J. of Geotech. Eng., 101(GT4): 409–412. Mosher, D.C., Christian, H., Cunningham, D., MacKillop, K., Furlong, A. & Jarrett. K. 2007. The Harpoon free fall cone penetrometer for rapid offshore geotechnical assessment. Proc. 6th OSIG, SUT, London, UK, 81–90. NORSOK Standard 2004. Marine soil investigations. G-001, Rev. 2, October 2004. Norwegian Geotechnical Institute 2002. Early soil investigations for fast track projects: Assessment of soil design parameters from index measurements in clays. Summary Rep./Manual, NGI Report 521553-3, 15 Jan. 2002. Perkins, S. & Sjursen, M. 2009. Effect of cold temperature of unfrozen Troll clay. Can. Geotech. J. 46(12): 1473–1481. Poirier, S.E. & DeGroot, D.J. 2010. Development of a portable probe for field and laboratory measurement of low to medium values of soil suction. Geotech. Testing J. 33:3.


Randolph, M.F. 2004. Characterisation of soft sediments for offshore applications. Proc. 2nd Int. Conf. on Site Characterization, Porto, Portugal, 1 209–231. Randolph, M. F., M. J. Cassidy, S. Gourvenec & C. J. Erbrich 2005. Challenges of offshore geotechnical engineering, Invited State of the Art paper. 16th Int. Conf. on Soil Mechanics and Foundation Eng., Osaka, Japan, 123–176. Randolph, M.F., Low, H.E. & Zhou, H. 2007. In situ testing for design of pipeline and anchoring systems. Keynote lecture. Proc. of the 6th OSIG, SUT, London, UK, 251–262. Ridley, A.M. & Burland, J.B. 1993. A new instrument for the measurement of soil suction. Géotechnique 43(2): 321–324. Sandbeakken, G., Berre, T., & Lacasse, S. 1986. Oedometer testing at the Norwegian Geotechnical Institute. Consolidation of Soils: Testing & Evaluation, ASTM STP 892, 329–353. Sheahan, T.C. & DeGroot, D.J. 2009. Seabed Sediment Pore Pressure: Genesis, Measurement and Implications for Design/Analysis. Final Report Int. Workshop for US Natl. Sci. Fdn. Project, www.offshoregeohazards.org, 114 pp. Strout, J.M. & Tjelta, T.I. 2007. Excess pore pressure measurements and monitoring for offshore instability problems. Proc. OTC, Houston, TX. Paper18706. SUT-OSIG 2000. Guidance Notes on Geotechnical Investigations for Subsea Structures. OSIG-Rev. 2, Prep. by Subsea Working Group of Offshore. Soil Invest. Forum. SUT-OSIG 2004. Guidance Notes on Geotechnical Investigations for Marine Pipelines. OSIG-Rev. 3, Prep. by Pipeline Working Group of Offshore. Soil Invest. Forum. Terzaghi, K., Peck, R.B., & Mesri, G. 1996. Soil Mechanics in Engineering Practice. John Wiley and Sons, New York. Tjelta, T.I. & Strout, J. 2010. Piezometer installation in deepwater Norwegian Sea. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics. ISFOG 2010. In press. van Paassen, L.A. & Gareau, L.F. 2004. Effect of pore fluid salinity on compressibility and shear strength development of clayey soils. In Hack et al. (eds). Eng. Geology for Infrastructure Planning in Europe. Springer-Verlag. 327–340. Vidal, V.M.V., Vidal, F.V., Hernandez, A.F., Meza, E. & Zambrano, L. 1994. Winter water mass distributions in the western gulf of mexico affected by a colliding anticyclonic ring, J. of Oceanography, 50: 559–588. Whittle, AJ, Sutabutr, T, Germaine, JT, & Varney, 2001. A Prediction and interpretation of pore pressure dissipation for a tapered piezoprobe. Géotechnique. 51(7): 601–617. Young, G.A. 1986. The strength deformation properties of Smith Bay arctic silts. SM Thesis, MIT, Cambridge, MA.

57


Gulf of Guinea deepwater sediments: Geotechnical properties, design issues and installation experiences J.-L. Colliat & H. Dendani Total, Pau and Paris, France

A. Puech Fugro France, Nanterre, France

J.-F. Nauroy Institut Français du Pétrole, Rueil-Malmaison, France

ABSTRACT: The paper describes the geotechnical properties of deepwater sediments encountered on the continental slopes of the Gulf of Guinea, in water depth ranging from about 400 m to 2000 m. After more than 10 years of site investigations, a large database on the behaviour of these sediments is now available and is tentatively summarized, with illustrations from various Gulf of Guinea sites located between Nigeria and Angola. The main characteristics of the deepwater West Africa clays are addressed, comprising plasticity, carbonate and organic contents, particle size distribution, mineralogy, sensitivity and thixotropy, compressibility and shearing behaviour in relation to the clay microstructure. It is shown that specific laboratory and in-situ testing procedures are required for proper determination of fine particle sizes and for the measurement of the relatively high soil sensitivity. The presence and origin of a near seabed “crust” is highlighted. Its origin is questioned, and potential implications on the design of soil-pipeline interactions are emphasized. Typical results of installation of suction piles, driven piles and VLA anchors are presented, which further illustrates the clay behaviour.

1

INTRODUCTION

(CLub pour les Actions de Recherche sur les Ouvrages en Mer) and with the financial support of CITEPH (Programme de Concertation pour l’Innovation Technologique dans l’Exploration et la Production des Hydrocarbures). As a key regional characteristic, the Gulf of Guinea is a “geotechnically remote” area, meaning that no soil investigation drilling vessel is permanently present in West Africa. In 1998, the first deepwater Gulf of Guinea soil investigation, at the Girassol site in 1300 m of water offshore Angola, was done in conventional drilling mode. Following this, the soil investigation strategy adopted for the Gulf of Guinea was based on the use of 30–40 m penetration seabed-based in-situ testing tools and 15–25 m long gravity piston corers. In soft deepwater sediments, such seabed-based equipments are limited by their stroke range, not thrust capacity. The requirements for geotechnical engineering purposes for suction piles and anchors (main type of foundation in deep offshore West Africa clays) are then covered. The drilling of deep soil borings (needed for e.g. driven piles for TLPs) currently still represents a small minority of cases in the Gulf of Guinea, which is a major difference with the Gulf of Mexico practice. The physical and geotechnical properties of the Gulf of Guinea deepwater sediments are described in detail in this paper. The main characteristics of these soft clays are addressed, comprising index properties

In about ten years, the deep offshore Gulf of Guinea has become a mature oil province with rapid development of numerous oilfields, like in the Gulf of Mexico or offshore Brazil. Compared to the Gulf of Mexico, the Gulf of Guinea is known to present benign environmental conditions. If subsea currents are indeed generally low and maximum storm waves limited to less than 8–10 m, strong squalls may impose severe design loads for the moorings of floaters (i.e. for temporary moorings of MODUs, or for permanent moorings of FPUs or FPSOs). The Gulf of Guinea is also known to present relatively gentle seafloor conditions, but the full list of geohazards can be encountered off Angola and Nigeria, such as faults, pockmarks, shallow gas, gas hydrates, salt and mud diapirs, seabed slope instabilities, and sub-seabed mass transport deposits. These deepwater geohazards, which may represent serious challenges in terms of geotechnical design, have been studied, principally by Total and Ifremer, since the early 1990’s, with several scientific cruises in Nigeria, Gabon, Congo and Angola (e.g. Cochonat et al. 1996, Sultan et al. 2007 & 2010). Geotechnical characterisation of the Gulf of Guinea deepwater sediments has been the topic of various collaborative R&D projects, carried out under the umbrella of CLAROM


(with the noticeable exceptions of Nigeria and ultra deepwater Angola, see Sultan et al. 2007, Hill et al. 2010, or Evans 2010), a new soil investigation strategy has emerged with the use of 30–40 m stroke seabedbased in-situ testing tools and large-diameter gravity piston corers allowing the recovery of 15–25 m long samples (Borel et al. 2002 & 2005). A first application for the Kuito FPSO anchors in 430 m of water offshore Cabinda is described in Alhayari et al. (2000). Adopting this survey strategy (also applicable to other remote deepwater provinces, such as the Australasia area) may allow a reduction of mobilisation times by performing the soil investigation from vessels of opportunity. Although still rarely used, vessels of opportunity may become important in fast track development projects. Deep soil borings are of course still required, either for the design of long driven piles (e.g. for the Kizomba TLP or the Benguela-Belize CT in Angola, Labbe & Perinet 2004, Will et al. 2006), for the installation of jetted well conductors (Evans et al. 2002), or in areas where more sophisticated testing and sampling might be needed in relation to specific geohazards such as in ultra deepwater offshore Angola (Hill et al. 2010, Evans 2010). But the drilling of 100 m+ deep soil borings represents a small minority of cases in the Gulf of Guinea, which is a key difference with the Gulf of Mexico practice. Currently, three geotechnical drilling vessels are capable of drilling in up to 2000 m of water, and two more can reach 3000 m of water. The PROD (Portable Remotely Operated Drill) is an available alternative, with a seabed-based system that combines the ability to take samples (piston sampling or rotary rock coring) and perform penetration testing (CPTs BPTs). The PROD1 and PROD2 systems are capable to operate in up to 2000 m and 3000 m of water, respectively (Kelleher & Hull 2008). By saving the time for tripping the drill-string, such a seabed-based system can be more efficient at deepwater sites than conventional drilling tools deployed from a vessel deck at the sea surface (Osborne et al. 2010). In 2009, the PROD1 was applied for the first time in the Gulf of Guinea for a soil investigation in water depths of 1100–1400 m offshore Angola.

(water content and plasticity, carbonate and organic contents), particle size distribution, mineralogy, sensitivity and thixotropy, compressibility and shearing behaviour in relation to the clay microstructure. Having possible implications for the design of pipelines, the near seabed “crust” of stiffer clay, locally encountered at several sites in Nigeria and Angola, has been the subject of several studies. Some key results are discussed, but the paper shows that the origin of this “crust” remains largely unexplained. Where relevant, specific laboratory and in-situ testing procedures are emphasised and illustrated by case examples taken from a number of West Africa deepwater sites located between Nigeria and Angola. The examples presented in the paper, covering water depths in the range of 400 to 2200 m, come from several sites operated by Total (containing both published and proprietary data) and from data published by the other three operators actively involved in the Gulf of Guinea, i.e. BP, Chevron and ExxonMobil. Some data presented in the paper were previously published but are completed by a larger database. Other unpublished data come from recent R&D studies. Some preliminary results from ongoing work on new topics of interest, such as strain rate effects, interface friction and modelling, are also presented. The design issues related to soil-pipeline interactions are briefly discussed. Finally, typical results of installation experiences are presented, covering suction piles, driven piles and VLA plate anchors. Such installation case studies provide a further illustration of the behaviour of Gulf of Guinea deepwater sediments, in particular with respect to friction degradation (i.e. clay sensitivity) and regain of strength with time (i.e. thixotropy and set-up). 2

SOIL INVESTIGATION STRATEGY

2.1 A strategy adapted to regional requirements The first deepwater soil investigation in the Gulf of Guinea was done in 1998 for the Girassol site at a water depth of 1300 m offshore Angola. It was carried out in conventional mode, i.e. from a specialised geotechnical drilling vessel and with down-hole sampling and in-situ CPT tools operated through the drill-string. Extending the shallow water soil investigation strategy to deep water (as was done in the Gulf of Mexico too) presented operational efficiency drawbacks, like the long time required for tripping the drill-string and the use of a 1.5 m stroke CPT tool, thus making the acquisition of data in deep borings rather time consuming. With FPSO and subsea wells being the preferred option for deepwater oilfield developments in West Africa, most of the foundations and anchors to be designed and installed only require a detailed investigation of the first 20–30 m of sediments, with emphasis on the top 1–2 m for the soil-pipelines or soil-risers interaction. Coupled with relatively gentle seafloor conditions in many Gulf of Guinea development areas

2.2

Large-diameter (100 mm) piston corers have been shown to provide quality of soil cores similar to 75 mm (3 ) thin-walled pushed-in piston samplers with the conventional down-hole method from a specialised geotechnical drilling vessel (Young et al. 2000). However, gravity piston corers may by-pass some of the top 1.5–2 m of very soft seafloor sediments. This may also be the case in extremely soft seabed sediments for the STACOR® which uses a base plate and a truly fixed piston (Wong et al. 2008). In its current configuration, the STACOR® corer has a 25 m long core barrel and will generally recover 12–25 m long samples, depending on the water depth or the clay strength gradient (Borel et al. 2002, Enjaume et al. 2010). Longer gravity piston corers may be able to recover up to 30–35 m long


Scope to 30–40 m for foundations or anchors

cores in specific conditions, i.e. with a heavier ballast weight and in extremely soft sediments or highly sensitive clays (Montarges et al. 1987), but the operation of such long corers is not a routine practice and can be detrimental to the sampling quality. Using accelerometers for accurate monitoring of the coring operation is a significant improvement for better understanding the corer behaviour and improving core recovery and quality (Buckley et al. 1994, Bourillet et al. 2007). In addition to the CPT, a number of in-situ testing tools can be deployed from seabed-based systems, the most commonly used being the field vane (VST), and the full-flow penetrometers (T-bar TPT and ball BPT; Randolph et al. 1998 & 2005). In soft sediments, the maximum investigation depth is not limited by the thrust capacity but by the maximum achievable stroke (generally 30–40 m). Based on years of practice, the undrained shear strength of the clay is determined by correlation with the cone resistance (Lunne & Andersen 2007). Other in-situ testing tools have the advantage of either a direct measurement (VSTs), or an increased accuracy in very soft sediments (TPTs and BPTs). Typical examples of Gulf of Guinea measurements are provided in subsequent sections of the paper. The use of various in-situ testing tools offers a superior data quality and a better definition of the soil strength intercept at the seafloor. Borel et al. (2010) propose a soil investigation strategy, where “clusters” of in-situ tests (CPT, VST and cyclic TPT or BPT) are performed at a number of selected locations, in combination with high quality sampling for advanced laboratory testing and calibration of the Nk factors. At other locations, the scope can be limited to penetrometer (CPT or full flow) testing and a reduced number of sampling for assessing the variability in soil conditions over the site. An accurate calibration of the Nk (alternatively Nt ) factor then allows a relatively precise determination of the clay strength, and the main geotechnical design parameters can be defined shortly after the soil investigation on the basis of the penetrometer test results.

Figure 1. Evidence of seabed crust from lightweight CPT testing and laboratory testing in box core (after Borel et al. 2005).

soil disturbance and frame penetration in the soil. An example of results obtained with the SEASCOUT® is shown in Figure 1. Piston gravity corers have been extensively used for pipeline soil investigations in deep waters. The analysis of the top decimetres of samples however often discloses partial loss or remoulding of the material which cannot be used for accurate measurements at very low stresses. Concerns have also arisen with giant piston corers regarding their capacity to capture subtle changes in the geotechnical parameters in the top decimetres below seabed, due to possible washing-out of the extremely soft seabed materials during the initial phase of free-fall penetration. Until recently, box corer sampling was the best option for recovering intact seabed samples. The standard box corer, typically weighing about 200–300 kg, penetrates 40–50 cm into the seabed sediments under its own weight. A trigger mechanism then releases a latch which causes the swivel base to close-off the captured sample, before the whole unit is recovered on board the vessel. The recovered samples are of high quality, but the extremely soft nature of seabed soils cannot allow any sub-sampling without disturbing the material. “In situ” testing inside the box therefore appears as the only reasonable way to obtain relevant shear strength data. This can be done with a minivane, but using a mini T-bar has the advantages of (a) obtaining continuous profiles and (b) measuring the soil sensitivity by performing quick cyclic tests. A manually operated mini T-bar (called DMS) was developed at the Centre for Offshore Foundation Systems (COFS, Perth, Australia) for strength profiling in box cores (Randolph et al. 2007). In 2008, Fugro developed the DECKSCOUT™ system (Fig. 2), in which the tool can be fitted with full flow penetrometers (T-bar or ball probes) with an actuator which can apply constant rates of penetration in the range of 0.01 to 2 cm/s (Puech et al. 2010); cyclic testing for sensitivity measurement can be automatically monitored.

2.3 Scope to 1–2 m depth for pipelines or risers Soils encountered in deep waters generally present shear strength profiles linearly increasing with depth (strength gradient between 1.0 and 1.5 kPa/m), starting at very low values at seabed (typically 1 to 2 kPa). In these very soft soils, assessing the longitudinal and lateral restraints of pipelines or flowlines laid on the sea bottom is a challenging issue for the industry. This requires both a very accurate measurement of the intact and remoulded undrained shear strengths over the first decimetres (accuracy in the order of or better than 1 kPa) and a precise soil-pipe interaction model. In the Gulf of Guinea the random presence of the so-called seabed “crust” (see details in section 4.9) increases difficulties in assessing accurate shallow shear strength profiles and impact on pipe penetration. In-situ tests (CPTU and T-bar tests) have been performed from lightweight frames in order to reduce


Figure 3. SMARTPIPE® and SMARTSURF™ modules. Table 1. Country

Site

Water depth (m)

Nigeria

A K O Q∗ U M Z

350–650 1250–1500 1500–2000 1400–1500 750–800 550–1100 1300–1400 400 1300–1400 700–1100 650–700 800–1400 1400–2250 1200–1500 1350 1300–1400

Figure 2. Fugro DECKSCOUT™ system in operation on box corer.

It is now widely accepted that soil-pipe interaction laws in very soft soils should better be obtained directly on the sea bottom. The Fugro SMARTPIPE® system measures directly in-situ the penetration of an instrumented pipe section and the longitudinal and lateral restraints resulting from static or cyclic loads (Hill & Wintgens 2009, White et al. 2010). The system is equipped with a 1 m stroke mini T-bar to characterise the seabed sediments at the test location. The second version of the equipment was successfully used offshore Angola at the end of 2009. As the SMARTPIPE® will necessarily be used at a limited number of discrete locations to obtain detailed information on the soil-pipe behaviour, a companion and complementary tool called SMARTSURF™ has been designed (Puech et al. 2010). It is aimed at providing accurate information on seabed soil properties required at a large number of locations regularly distributed all along the pipeline routes to (a) assess whether the chosen test locations give representative results and (b) extrapolate data with confidence to the entire pipeline network. The SMARTSURF™ is equipped with a 3 m stroke standard T-bar or CPTU, a 1 m stroke mini-T-bar and a pushed-in stationary piston sampler specifically designed to recover 2 m long cores of soft undisturbed soil. The SMARTPIPE® and SMARTSURF™ can be operated with the same launch and recovery system (LARS), allowing safe handling from any vessel of opportunity fitted with a convenient A-frame (Fig. 3). The SMARTPIPE® and the SMARTSURF™ can also be deployed from a specialist drilling vessel. The equipment is rated for up to 3500 m water depth. A site investigation strategy to obtain fast and reliable design parameters for soil-pipe interaction assessment entirely based on in-situ testing is proposed by Borel et al. (2010).

Congo Angola

∗

3

B∗∗ D F G I L P R S

Ehlers et al. (2005) ∗∗ Dutt & Ehlers (2009)

PHYSICAL PROPERTIES

Physical and geotechnical properties described in sections 3 and 4 were derived from a review of industrial or R&D data from sixteen sites. Details are given in Table 1. 3.1 Water content and submerged unit weight Deepwater sediments in the Gulf of Guinea are characterised by very high water contents (Fig. 4). At the seafloor, w is typically comprised between 150 and 250%, and decreases to 100–200% over the very first metres of penetration (typically 6–8 m). Over these upper 6–8 m the water contents are close to or over the liquid limit (giving 1 < LI < 1.2). Beyond that depth the rate of decrease becomes very slow over the depth of interest for the engineering of most deepwater structures (i.e. 20–30 m below seabed). Most of the sites exhibit water contents between 80 and 150%. The highest values are mainly (but not exclusively) representative of the deepest sites. A few deep penetration boreholes are available. They confirm a decreasing trend (Fig. 5), but water


Gulf of Guinea sites considered in paper.

Figure 4. Water contents versus penetration in top 20 m (from STACOR® samples).

Figure 5. Water contents versus penetration depth (deep boreholes).

contents still present relatively high values (80–120% at about 120 m of penetration in the deepest boreholes). The site B borehole appears as an exception but it is worth noting that site B is in only 400 m water depth (Dutt & Ehlers 2009). There is a relatively large scatter in the values of water contents at a particular depth. This observation is not only valid for the global data set but also for a particular site or field, and even for a given borehole as highlighted in Figure 4. The bulk unit weights are very low, starting at 12–13 kN/m3 at seabed and reaching 13–15 kN/m3 below 6–8 m. Submerged unit weights of Gulf of Guinea soils are significantly lower than in the Gulf of Mexico where submerged unit weights become rapidly higher than 6 kN/m3 (e.g. Quiros & Little 2003). 3.2 Carbonate content Carbonate contents are generally lower than 20% of the total soil weight (Fig. 6). Variations with depth are mostly erratic even along the same profile (see for example site M). Scanning electron microscopy (SEM) observations indicate that carbonates are mainly shell debris or foraminifers randomly distributed in the clay matrix. The proportion of calcite detected in the finer fractions (20% and up to 40%).

Figure 14. Atterberg limits versus penetration depth.

3.6 Plasticity The plasticity index (PI) of Gulf of Guinea deepwater clays is typically between 70 and 130 but can reach values as high as 150 near the seabed. These values are much higher than those frequently encountered in the “highly plastic” Gulf of Mexico clays with PI typically in the range 30–70. The plastic limit (WP) is quasi constant with depth, in the range of 50 ± 10% whereas the liquid limit (WL) typically decreases from 150–200% at the seafloor to 130–170% below 6–10 m of penetration (Fig. 14). When represented in the Casagrande diagram, the soils plot close to the A-line and classify as highly plastic clays (CH) to highly plastic silts (MH), as shown in Figure 15. 4

Figure 15. Gulf of Guinea clays in Casagrande diagram.

to define the stress history of the Gulf of Guinea soils, since no implicit assumption regarding the past overburden pressure is introduced. The YSR is determined from the ratio of the vertical yield stress σvy to the actual overburden pressure σv0 :

The vertical yield pressures derived from a number of standard oedometer tests performed on several typical deepwater sites are plotted versus depth in Figure 16 and are used to compute the yield stress ratio. is higher than the vertical effective The yield stress σvy stress σv0 . The distance between the profiles expressed − σv0 ) is quasi constant by the difference (σ = σvy with depth, typically in the range of 15–40 kPa for most sediments. The corresponding YSR values decrease

GEOTECHNICAL PROPERTIES

4.1 Yield stress ratio The term yield stress ratio (YSR) is preferred to the commonly used term over-consolidation ratio (OCR)


Figure 17. Sensitivity framework for site Z clay. Figure 16. Vertical yield pressure and YSR versus penetration depth for typical Gulf of Guinea sites.

The properties of a natural clay differ basically from its intrinsic properties due to the influence of the soil structure (fabric and/or bonding). The structure of a natural clay depends on many factors, such as depositional conditions, ageing, cementation and leaching. The evolution of the void ratio with penetration depth, or with the effective vertical pressure σvo in a natural clay deposit, is called the sedimentation compression curve (SCC). The majority of normally consolidated natural clays have sedimentation compression curves which (after normalisation) lie in a narrow band above the ICL. The regression line through this band has been called the sedimentation compression line (SCL) by Burland. The SCL is not a fundamental line but represents the trend followed by a majority of natural sedimentary clays. Temporal variations may result in “saw-toothed” curves instead of smooth regular curves. The fact that the SCL lies to the right of the ICL implies that, for a given void ratio, a natural clay is capable of carrying an effective overburden pressure higher than the corresponding reconstituted clay. Cotecchia & Chandler (2000) have generalised Burland’s approach to sensitive clays. They have shown that the SCL is not unique (as implicitly assumed in Burland’s work) but depends on the soil sensitivity. Typically, for site Z clay, the sensitivity framework approach is illustrated in Figure 17 by plotting the following data:

from about 3 at 2–3 m of penetration to 1.1–1.3 at depth. Higher values of YSR can be found within the first 2 m below seabed, depending on the presence of a so-called “crust” (see sections 2.3 and 4.9), but this particular point is not discussed here. Yield stress ratios in excess of 1 are not the result of over-consolidation in the geological sense (i.e. no past overloading of the material). De Gennaro et al. (2005) have shown that these soils exhibit a significant structural effect, and that the (σvy − σv0 ) difference is a quantitative measurement of the “extra-strength” due to the material structure. 4.2 Compressibility De Gennaro et al. (2005) and Le et al. (2008) have highlighted the interest of interpreting the compressibility data of Gulf of Guinea deepwater sediments in the light of the conceptual framework introduced by Burland (1990) for comparing and interpreting the compressibility of natural sedimentary clays and reconstituted clays. A normalising parameter called void index Iv is introduced to aid in correlating the compression characteristics of various clays:

– the ICL, as proposed by Burland (1990); – the ICL obtained for Gulf of Guinea deepwater sediments (Le 2008). For effective vertical stresses in excess of 10 kPa (i.e. soils below 3 m penetration), a very good fit is observed; – the SCL proposed by Burland, which corresponds to a sensitivity St of about 4 to 5; – the SCL proposed by Cotecchia & Chandler (2000) for a sensitivity of 6; – the envelope and mean value of the SCCs obtained from the water content measurements on the site Z samples.

where e∗100 and e∗1000 = intrinsic void ratios corresponding to an effective vertical stress σv of 100 kPa and 1000 kPa respectively (determined by oedometer tests on reconstituted sample); and C∗c = intrinsic compression index. The properties of reconstituted clays are called “intrinsic” properties since they are inherent to the soil and independent of its natural state. The compressibility of reconstituted clays can be described by a reasonably unique line called the intrinsic compression line (ICL) passing by the points (Iv = 0, σv = 100 kPa and Iv = −1, σv = 1000 kPa). For a particular clay, e∗100 and C∗c are related to the value of the void ratio at the liquid limit eL . © 2011 by Taylor & Francis Group, LLC

Based on the trend of the sedimentation compression curves, this analysis suggests that the clay has a marked

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structure and a sensitivity in the order of 4 to 6. It is worth mentioning that the sensitivity measured from in situ cyclic T-bar testing was found close to 6. According to the sensitivity framework, a normally consolidated soil with a post-sedimentation structure is characterised by compressibility curves that cross the SCL at their in-situ stress (σv = σvo ) before show ing an abrupt increase in compressibility at σv = σvy (yield stress). This is exactly what is observed in Fig ure 17 for site Z. At stresses above the yield stress σvy , the compression curves plunge and reach the experimental ICL at higher stresses (about 200–300 kPa). The compressibility is maximal after the yield point and values of compression index Cc as high as 2 can be observed. In summary, the Gulf of Guinea deepwater clays (a) are normally consolidated, (b) have a marked post sedimentation structure, and (c) have a sensitivity in the range of 3 to 6, and this sensitivity can be interpreted as a quantitative measurement of the structural effect. 4.3

Figure 18. SEM micrographs of site D clay samples from 0.5 m and 14.0 m penetration.

Microstructure

The microstructure of intact samples has been investigated at ENPC/CERMES (Le 2008) by using scanning electron microscopy (SEM) and mercury intrusion porosimetry (MIP) in parallel, as suggested by Delage & Lefebvre (1984).To preserve the microstructure during dehydration, samples were frozen as quickly as possible so as to get a crypto-crystalline structure of ice with no volumetric expansion due to freezing according to the methodology proposed by Delage et al. (2006). Figure 18 shows two SEM pictures of surface (0.5 m) and deeper (14 m) samples. At both depths, a clear evidence of large voids can be observed, with diameters equal or larger than 1 µm, typical of soft soils. Voids appear to be as frequent as the solid phase, corresponding to an approximate porosity of 50%. Connecting clayey bridges are well observed, giving to the assemblage the aspect of a honeycomb microstructure. Clay minerals are easily observable with an average thickness smaller than 0.1 µm and an average diameter close to 0.5 µm. In Figure 19, the SEM photos of the 0.5 m sample submitted to vertical stresses of 50, 200, 800 and 1600 kPa emphasise the changes in microstructure during compression.The progressive collapse of the material structure, principally due to a decrease of the pore sizes, is responsible for the high compressibility of the clay. Mercury intrusion porosity (MIP) is based on the principle that a non-wetting fluid (here mercury) cannot enter a porous medium unless a pressure is applied. Assimilating pores to cylindrical capillaries, the pressure p can be related to an equivalent entrance pore radius r by Laplace’s law:

Figure 19. SEM micrographs of site D clay samples (0.5 m penetration) after compression to 50–1600 kPa (from Le 2008).

As the pressure p is increased, pores of smaller and smaller radius are filled with the intruding liquid. Results are plotted under the form of a cumulative curve giving the pore size distribution (eHg ) of the porous medium. In Figure 20, the MIP pore size distribution curves of intact samples from various penetration depths define a particular pore entrance value of about 0.2 µm. This radius is compatible with the intra-aggregate pore sizes observed through SEM. Below 0.2 µm, the curves are superimposed and this domain is known to be typical of the inter-aggregate porosity and to be poorly sensitive to macroscopic changes in void ratio. 4.4

4.4.1 CPT cone resistance and T-bar profiles Typical cone and T-bar resistance profiles are presented in Figure 21. They are representative of the large number of CPTs performed on West Africa continental slopes. Below 2 m of penetration and down to 30–40 m, the penetration resistance increases quasilinearly with depth. The gradient in net cone resistance lies between 10 and 30 kPa/m with a general tendency

where σ = surface tension of the intruding liquid; and θ = contact angle.


Undrained shear strength

Figure 22. Gradient of net cone resistance versus water depth.

Figure 20. MIP pore size distribution curves for site D clay (from Le 2008).

Figure 23. T-bar and VST test results with Ntv = 11.5 (from Borel et al. 2005).

4.4.2 Undrained shear strength and sensitivity Vane shear tests (VST) provide a measurement of the in-situ peak shear strength Suv at typically 1 m or 1.5 m intervals. The net cone resistance and the vane shear strength can be related by a cone factor Nkv such as qn = Nkv . Suv . The Nkv factor is in the range of 10–15. More detailed analyses performed on a small number of sites where 30 to 40 m profiles of VST and CPT tests could be closely correlated suggest an Nkv factor increasing with depth from about 11 near the seabed to 14–15 at depth (Puech et al. 2005). The Ntv factor applicable for T-bar data has been found close to 11.5 either for in situ standard T-bar testing (Fig. 23) or for mini T-bar testing (Puech et al. 2010). For water depth in excess of 700–800 m, the average gradient in shear strength ranges from 1.0 to 1.5 kPa/m. Slightly higher values can be locally encountered at shallower sites. Ultra deep water sites investigated so far (in a water depth of about 2000 m) have gradients on the lower bound. However, gradients of the same order have been found at sites in only 1400 m of water are depth. Normalised shear strength ratios SuVST /σvc high, with values in excess of 1 over the very first

Figure 21. CPT net cone resistance and T-bar profiles (site K).

to decrease with increasing water depth as shown in Figure 22. In areas unaffected by the presence of geohazards, resistance profiles are highly repetitive (additional examples of CPT profiles are given in Colliat & Colliard 2010). Over the top 2 m of penetration two different types of profiles may be encountered i.e. profiles linearly increasing with depth or profiles presenting a peak in resistance corresponding to the seabed crust. Comparisons are presented later in Figure 33. The origin and consequences of such peaks in geotechnical engineering are discussed in sections 4.9 and 5.


Figure 24. Composite plot showing intact shear strength profiles (CPT and VST) and residual shear strength profiles (VST) compared to CPT sleeve friction and cyclic T-bar tests (cycles 10–30 shown). Figure 25. Design remoulded shear strength profiles (assuming St = 3 to 4) compared to CPT sleeve friction, VST tests (remoulded strength after 3 rotations) and cyclic BPT test results (cycles 1–10 shown).

1–2 m of penetration, decreasing to about 0.4–0.5 below 10 m. The latter is compatible with a ratio = 0.36 mentioned by Andersen & Jostad SuDSS /σvc (2004), assuming SuDSS = 0.75 SuVST (Fig. 27). The sensitivity St is expressed as the ratio of the intact undrained shear strength Su,i to the remoulded undrained shear strength Su,r, measured using the same instrument. The measurement of clay sensitivity is a topic that attracted specific attention, in particular since installation experiences suggested that it tended to be under-estimated by the measurements provided by both laboratory tests (UU) and in situ tests (see section 6.1). Farrar et al. (2008, cited in Robertson 2009) claim that the CPT sleeve friction is often close to the remoulded undrained shear strength measured by the VST in soft NC clays. But Lunne & Andersen (2007) have shown that the CPT sleeve friction measurements do not agree with other remoulded shear strength measurements. Figures 24 and 25 present a comparison of data obtained with the VST, the CPT and full flow penetrometers (Tbar and ball) at two different sites in Angola. In Figure 24 are shown:

Figure 25 present results from another site, showing: – the CPT sleeve friction; – the remoulded shear strength as obtained with the VST but after only 3 rotations of the blades instead of 10 as per ASTM recommendations (limitations due to time constraints); – the remoulded shear strength derived from the ball tests (Nb = 12) after 10 cycles; – the design remoulded shear strength profiles based on labvane tests and past installation experiences in the area (assuming a clay sensitivity of 3 to 4). These data illustrate typical results obtained in the normally consolidated but structured clays of the Gulf of Guinea, i.e.: • The CPT sleeve friction clearly compares with the

residual (not remoulded) VST shear strength values; • The VST remoulded shear strength measurements

(when the number of blade rotations is limited to 2 or 3 as is often the case in offshore practice) plot below the residual shear strengths but provide underestimated values of the clay sensitivity (i.e. in the range of 2 to 3); • The cyclic T-bar or ball tests provide higher values of sensitivity, generally consistent and less scattered

– the cone resistance (divided by a cone factor of 13) and the peak VST shear strength; – the CPT sleeve friction and the residual shear strength obtained with the VST; – the remoulded shear strength as derived from cyclic T-bar with Nt = 11.5 (cycles 10 to 30 are shown). © 2011 by Taylor & Francis Group, LLC

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Figure 27. Correlation between VST and DSS shear strengths.

value representative of normally consolidated conditions (K0nc ) measured in the post-yield domain is close to 0.5 (in the range of 0.45–0.55 for the small number of specimens tested). Higher values are obtained in the pre-yield domain. 4.5.3 Shear strength anisotropy Series of direct simple shear (DSS) tests and anisotropically consolidated (CAUc and CAUe ) triaxial tests were performed. The shear strengths obtained are respectively noted SuDSS , Suc and Sue . The ratio Suc /SuDSS has been found close to 1.2. The values obtained for the Sue /SuDSS ratio are more scattered, due to the difficulty in controlling extension tests on very soft soils and at very low confining pressures, but a value of about 0.8 seems appropriate on the basis of the most reliable sets of data. A fair correlation has been found between the shear strengths obtained from VST and DSS tests, of the type SuDSS # 0.75 SuVST (Fig. 27). This result may be interesting for fast track engineering when only insitu data are available. It is noted that the data imply relatively high conversion factors between penetrometer resistances and average laboratory strengths (e.g. NtDSS # 15). This may be explained by both the high rate sensitivity of Gulf of Guinea clays and the robustness of their fabric (reduced softening with first pass of penetrometer).

Figure 26. Typical results from cyclic T-bar tests in site Z clay.

than labvane measurements. Sensitivities are often in the range of 3 to 4 but can reach 6 for site Z as shown in Figure 26. These values are in good agreement with the interpretation of compressibility data in the sensitivity framework (section 4.2) and were confirmed by back-analysing suction pile installation data at the respective sites (see section 6.1). Recent (unpublished) data obtained with the mini T-bar indicate that sensitivity values in excess of 10 may be obtained within the “crust”. It is generally admitted (e.g. Randolph et al. 2005) that 10 cycles are sufficient to reach stabilisation and derive a sensitivity value from full flow penetrometers. In the structured clays of the Gulf of Guinea, Borel et al. (2010) and Puech et al. (2010) emphasise the need to perform at least 25–30 cycles to reach stabilisation (Fig. 26).

4.6 Thixotropy 4.5 Advanced laboratory testing

Thixotropy is measured as the regain in undrained shear strength with time of remoulded samples maintained at constant water content and sheared at increasing time intervals. All available results (21 tests from 8 different sites) were compiled to provide the envelope curves presented in Figure 28. These results are consistent with those presented by Andersen & Jostad (2002). They are also in agreement with the set-up behaviour of suction piles at various Gulf of Guinea sites described by Colliat & Colliard (2010). The results obtained for site Z may suggest that the higher the sensitivity the faster the undrained

4.5.1 Sample quality A large majority of samples recovered in the deep waters of the Gulf of Guinea were taken using the STACOR® giant piston corer. All results referred to in this paper were derived from high quality STACOR® samples with the Lunne et al. (1998) quality index generally rating fair to excellent. 4.5.2 Coefficient of lateral earth pressure at rest K0 The coefficient of lateral pressure K0 can be obtained in drained triaxial tests under zero radial strain. The


Figure 28. Results from thixotropy tests.

shear strength is regained, but data are too scarce to draw definite conclusions. 4.7

Consolidation properties

The coefficient of vertical consolidation Cv of Gulf of Guinea clays measured in incremental oedometer tests is generally found between 1 and 10 m2 /year in the virgin compression zone. The coefficient of consolidation Cv,r corresponding to recompression in the stress range of 0.5 σvo to σvo (often used to evaluate reconsolidation around suction emplaced piles) lies in the same order of magnitude. These values apply in the range 2 to 20 m with a slightly decreasing trend with depth. An interesting point is that similar results are obtained on oedometer tests carried out on horizontally or vertically trimmed specimen, which is consistent with SEM observations showing identical microstructure in both directions. Unfortunately the authors are not aware of experimental field data (settlement of structures, pore pressure dissipation around instrumented pile walls) which could be used to confirm the validity of these values at large (foundation) scale. Thorel et al. (2010) have performed centrifuge tests on reconstituted Gulf of Guinea clays. The Cv values calculated from the observed settlements by the Asaoka method are of about 30 m2 /year, i.e. 5 to 6 times higher than Cv values measured from oedometer tests on in situ samples of the same clay. Of particular interest is the value of Cv at mudline for pipeline applications. Laboratory testing on the extremely soft surface material is highly challenging and poorly reliable. Hill & Wintgens (2009) and White et al. (2010) report consolidation measurements performed with the SMARTPIPE® system (Fig. 3) where an instrumented pipe section was allowed to settle on the seabed. The pore pressures dissipate rapidly showing that lay-induced pore pressures under a real pipeline would dissipate within days, not weeks. A satisfactory match between field observations and Finite Element Analyses is obtained for Cv values of about 75–100 m2 /year.

Figure 29. Results of triaxial shearing tests on site M clay (from Le et al. 2008).

4.8


Effective parameters from triaxial testing

The shearing behaviour is largely explained by the structure of the clays. Le (2008) and Le et al. (2008) performed two types of tests on a natural clay: (a) triaxial compression and extension tests, in which the samples were consolidated to their estimated in-situ effective stress state, then sheared in undrained condition, and (b) SHANSEP type tests, in which the samples were consolidated isotropically (or anisotropically) beyond their in-situ states of stress before being sheared in undrained condition. Typical results are presented in Figure 29. The TN4 (compression) and TN7 (extension) tests performed at in-situ stress state present a non-plastic response in the q − p space (p nearly constant before yield). Failure is characterised by high effective friction angles of 40◦ or even more (M > 1.6), and such results are typical of all sites investigated so far (Fig. 30). The non-plastic pre-yield response is related to the strong structure of the clay. The high effective friction angles may be explained by a “sand-like” behaviour of the material where the aggregates of fine particles play the role of grain-sized elements. Friction angles in excess of 40◦ have been found in other clays, e.g. Mexico City clays (Diaz-Rodriguez et al. 1992).

Figure 31. Interface shear test results: peak shear strengths from ring shear tests on remoulded clay compared to Cam-shear tests on undisturbed clay.

Figure 30. Composite plot of triaxial test results in the q − p space.

smooth interfaces simulating different pipeline coatings (Kuo et al. 2010). To simulate interface conditions at shallow pipeline embedment, the stress level was very low ( 25).

6.1.3 Sensitivity estimated from self-weight penetrations During penetration, either by self-weight or when suction is applied, the friction resistance along the outside and inside pile wall is assumed to be governed by the remoulded clay strength at the soil-pile interface, i.e. by the soil sensitivity. From the back-analysis of the self-weight penetration of a number of suction piles at various deepwater West Africa sites, Colliard & Wallerand (2008) suggest a clay sensitivity larger than expected.

6.1.4 Installation behaviour Figure 40 summarises the installation behaviour of three sets of FPSO piles, with a self-weight penetration equal to 50–67% of the final embedment depth, which is typical of suction piles in soft deepwater sediments. At final penetration, the suction ranges from 70–95 kPa (at 16.5 m for site S) to 100–135 kPa (at 20.5 m and 24.0 m for site D and site K, respectively). Required suction reduces with increased pile diameter (for a given penetration) since it acts over an increased internal pile cross section. With the suction piles from

Figure 40. Typical suction versus depth curves for installation of FPSO suction piles.


site D and site K having similar diameter and submerged weight, together with similar undrained shear strength profiles at both sites, the lower suction measured for the K piles suggests a higher sensitivity of the deepwater clays from this site. The suction curves given in Figure 40 are typical of the penetration resistance of suction piles with a diameter in the order of 5 m and a height-to-diameter ratio of about 4 to 5 (see Table 2). Colliat (2006) showed that similar installation behaviour is obtained for suction piles in the Gulf of Mexico when the gradient in undrained shear strength is similar (i.e. of about 1.0 to 1.5 kPa/m). However, it should be noted that the suction curves shown in Figure 40 were not always measured directly inside the pile and have not been corrected for venturi effects or hydraulic losses in the pumping system; this may be the reason for the step increase of suction immediately after self-weight penetration for the D and K piles. 6.1.5 Effect of ring stiffeners The lack of heavy lift vessels in the Gulf of Guinea initially promoted fabrication solutions in favour of reducing the suction pile weight by using internal ring stiffeners. This was the case for the Nkossa FPU or the Girassol FPSO piles (Colliat et al. 1997 & 2007). When such ring stiffeners are relatively wide and close to the pile tip, it is now well known that they have a direct influence on the soil-pile interface friction. The inside friction is reduced, in relation to an increased remoulding effect and down-dragging of a mixture of seafloor soil and water, whereas the outside friction may be increased by displacing a larger volume of the inside soil plug out of the pile during penetration. For the latter effect, Andersen & Jostad (2004) describe how the normal stresses along the external pile wall are initially taken by the pore pressures, then give increased effective stresses and friction resistance as pore pressures dissipate. Pull-out tests of model piles with and without stiffeners in the centrifuge show that the net pull-out capacity of stiffened piles is 10–35% higher than for piles without ring stiffeners (see Fig. 43 and Colliat et al. 2010).

Figure 41. Examples of installation and retrieval behaviour.

Gulf of Guinea sites is described in Colliat & Colliard (2010), with set-up periods ranging between 1 day and 3½ years. The results obtained suggest a rapid increase in friction resistance, from 35–45% in one week to about 70% in one month, but little gain between one month and 3½ years. This is in relatively good agreement with (a) the results of thixotropy tests in high plasticity clays (Andersen & Jostad 2002, Dendani 2003, Fig. 28), and (b) similar data obtained at deepwater Gulf of Mexico sites (Jeanjean 2006). The interpreted interface friction factor ranges from 0.25 to 0.35 at time of installation, and increases to 0.40–0.56 in the longer term. Despite some inevitable uncertainties related to too scarce a database and other operational issues (in particular, when the measurements of the retrieval over-pressure and submerged pile weight are inaccurate), obtaining a long term friction factor below the generally accepted design value of 0.65 for high plasticity clays is a concern (where the friction factor is equal to 1.95/St with St = 3, from Andersen & Jostad 2002). A thorough confirmation or verification would require a larger set-up database, or (preferably) the performance of specific field tests.

6.1.6 Set-up behaviour Suction piles might have to be retrieved when they are installed out of tolerances (generally within one or two days after installation), or in the event of other operational constraints such as the replacement of a mooring line (i.e. several months or years after installation). In both cases, retrieval of the suction pile by over-pressurising allows a direct measurement of the increase in friction resistance with consolidation time (set-up). One example of installation and retrieval behaviour of two FPSO piles is given in Figure 41, showing an installation suction of about 90 kPa at 16.5 m depth, and an over-pressure of 180 kPa and 165 kPa at time of retrieval one and two days later, respectively. Although limited to a database of six retrieval cases, the set-up behaviour of suction piles at three different

6.2

The case of the Girassol RTA piles experiencing an abnormally low friction resistance due to the painted outside wall (contractor’s choice for protection against corrosion) has been published in detail (Dendani & Colliat 2002, Dendani 2003, Colliat et al. 2007) and is not repeated here. But, as a direct detrimental


Suction piles for riser towers

Figure 42. Installation predictions and results for large RTA piles (from Colliat et al. 2007).

consequence, the design of following RTA piles was more conservative and did not rely on passive suction. For the Girassol RTA piles, the design shear strength (determined from DSS tests) was multiplied by 0.70 for the effect of constant tension loading on the undrained clay strength (thus increasing the material factor from 1.30 to 1.86), and passive suction (REB) was considered only for the short term loading case. After installation, the reduced friction capacity of the painted piles was balanced by doubling the RTA submerged weight with ballast, thus making it a kind of “gravity friction pile”. On the other hand, the RTA pile for site R was originally designed as a gravity friction pile, with the submerged weight of the RTA equal to 1.25 times the service tension load case, in order to have the RTA working under compression loading. A similar type of foundation with counter-weights is also mentioned by Zimmermann et al. (2009) for the Greater Plutonio riser tower in Angola. Figure 42 shows a comparison of the installation results and predictions for both the site S and site R RTA piles. Being the first suction piles ever installed in soft deepwater Gulf of Guinea sediments, a rather conservative prediction was done for the site S RTA piles, but the much lower penetration resistance is mainly related to the paint effect (Colliat et al. 2007). The good installation prediction for the site R RTA pile confirms that the design model is correct. However, more field experience will be required before the industry regains confidence in REB capacity for RTA piles under tension service loading.

Figure 43. Centrifuge pull-out testing of suction piles with and without stiffeners (from Colliat et al. 2010).

This RTA experience was also studied by model pull-out tests in the centrifuge, specifically performed with similar deepwater clay samples reconstituted in the centrifuge, and with emphasis on the effect of ring stiffeners on both penetration and pull-out resistance (Thorel et al. 2010). It was observed that significant passive suction is mobilised in rapid pull-out loading of both the stiffened (S) and un-stiffened (U) piles, although it is necessary to displace the pile in excess of one metre to develop 100% of the REB capacity (Fig. 43). Similar results were obtained by Raines et al. (2005) with centrifuge tests carried out with Speswhite clay samples. 6.3

With a penetration of 150 m, the foundation piles of the Benguela-Belize compliant tower (together with those of the Tombua-Landana CT in the same area) are within the longest piles ever driven in soft deepwater sediments. The 2.74 m (108 ) diameter CT piles were driven with a Menck MHU 2100 hammer, and relatively low driving blow-counts (i.e. 22–24 blows/25 cm at final depth) are reported by Will et al. (2006). On another hand, with a penetration of 123 m, the 2.14 m (84 ) diameter piles of the Kizomba A TLP are within the range of pile diameters and penetrations for equivalent Gulf of Mexico TLPs (Doyle 1999).


Driven piles

Despite a slight difference in pile diameter (i.e. 96 for Kizomba, versus 72 –96 in the Gulf of Mexico), a fair comparison of the Kizomba pile driving behaviour is possible since the same type of hydraulic underwater hammer (Menck MHU 500T) is used in the Gulf of Mexico. For the Kizomba TLP piles, Labbe & Perinet (2004) report a lower SRD than expected and lower driving blow-counts at final depth (i.e. 28 blows/25 cm instead of a predicted value of 45 blows/25 cm). In the Authors’ opinion, this lower than expected SRD is due to the use of the Stevens et al. (1982) method, which assumes implicitly a clay sensitivity of 2 and will then provide conservative drivability predictions in soft clays with a larger sensitivity. For the Kizomba TLP piles, a better drivability prediction would have been obtained by using an alternative SRD calculation method with a friction degradation model (e.g. Puech et al. 1990, Colliat et al. 1993, or Dutt et al. 1995). The largest difference between the Kizomba and Gulf of Mexico sites is in the self-weight penetrations of the TLP piles, with 51–55 m for Kizomba (for a prediction ranging between 34 m and 52 m), as compared to a typical range of 30–40 m for similar pile sizes in the Gulf of Mexico (Doyle 1999). At Kizomba, Labbe & Perinet (2004) also report the performance of a re-drive test, showing a rapid gain of pile shaft friction resistance with time. Interestingly, similar rapid set-up of large diameter driven piles is described by Dutt & Ehlers (2009), also comparing two case studies from two deepwater sites in West Africa and in the Gulf of Mexico. These pile driving results from various Gulf of Guinea sites suggest a sensitivity of the clay higher than considered in standard driveability prediction methods and a rapid regain of resistance with time. This is in good agreement with the installation behaviour of suction piles described above.

deepwater clay, the required UHC is typically in the order of 4–5 MN for temporary MODU moorings, as compared to about 8–9 MN for semi-permanent or permanent moorings (Colliat 2006). In a recent application for three stand-by moorings for supply boats, 15 m2 Stevmanta anchors were installed in 1300 m of water offshore Nigeria, reaching an embedment depth of 28–32 m. Published Stevmanta or Dennla application cases (Ruinen & Degenkamp 2001, Murff et al. 2005, Magne 2008) mention embedment depths of 15–25 m for 10–15 m2 VLA anchors (with the smallest values of penetration and plate area corresponding to VLAs installed into stiffer clays offshore Brazil). In comparison, the result obtained with this Nigeria example suggests a lower penetration resistance in the Gulf of Guinea sediments, in a similar way to suction piles or driven piles. 7

Originally considered as an area with relatively gentle soil conditions, the deep offshore Gulf of Guinea has proven to be unique in certain aspects, highlighted by a number of differences with the Gulf of Mexico characteristics. After more than 10 years of investigations in the Gulf of Guinea, a large database on the behaviour of these deepwater sediments is now available, and has been tentatively summarised. A soil investigation strategy adapted to the regional requirements of this remote area is proposed, with emphasis on the importance of in-situ testing. Geotechnical characterisation is based on the use of conventional CPTs and VSTs, with the need for performing cyclic full-flow penetrometer tests (T-bar or ball tests) with at least 25–30 cycles for the determination of the soil sensitivity. Specific laboratory testing procedures were developed to address some key physical and geotechnical properties of these unusually high plastic clays, e.g. for the measurement of organic content and the determination of fine particle sizes. Since the Gulf of Guinea deepwater sediments have a high degree of structure, triaxial testing of high quality samples under in-situ stresses should be preferred to the SHANSEP test procedure usually applied for Gulf of Mexico clays. This high degree of structure is also the cause for measuring an apparent overconsolidation ratio of about 1.5–2 in the top 10–20 m of sediments, but the Gulf of Guinea deepwater sediments are generally normally consolidated clays (away from topographic highs created by diapirism). Two important properties of the Gulf of Guinea deepwater sediments are confirmed by the results of installation experiences, i.e. a relatively high sensitivity (St = 4 to 6), giving a lower than expected resistance to penetration of suction piles or driven piles from friction degradation, and a rapid and important regain in friction resistance with time from thixotropy and set-up effects. Some geotechnical issues still require further studies, in particular: (a) origin and characterisation of the near seabed “crust”, (b) interface soil-pipeline friction

6.4 Vertically loaded plate anchors In 1995, conventional drag anchors were used for the moorings of the Nkossa oil FSO in a water depth of 125 m offshore Congo (Colliat et al. 1997). To the Authors’knowledge, the first application of VLA plate anchors in TLM moorings in the Gulf of Guinea was for the semi permanent mooring systems of MODUs, for the drilling of production wells at the Kizomba field offshore Angola. SEPLA anchors were installed by means of a suction follower, and the same type of SEPLA anchors has then been used at the TombuaLandana field, offshore Cabinda, for the same kind of application. On another hand, both the Vryhof Stevmanta and Bruce Dennla VLAs have been used in a number of pre-installed temporary moorings for exploration drillling MODUs. For permanent moorings, VLA plates of about 25–30 m2 in surface area are generally used, whereas smaller anchors, with areas of the order of 12–15 m2 , are used in the latter case of temporary moorings. The plate area is directly related to the required anchor UHC (Murff et al. 2005). For an estimated embedment depth of about 25 m in soft


CONCLUSIONS

Prof. P. Delage (ENPC-CERMES), Prof. F. Thomas (INPL/LEM), N. Sultan and J. Meunier (Ifremer) and H. Manh Le (Fugro).

resistance, (c) effect of gas and gas hydrates in the sediments, and (d) definition of a single constitutive model to reproduce the effects of stress history, structure and strain rate. These issues are targeted by ongoing R&D studies.

REFERENCES 8 ABBREVIATIONS AOCR BPT CT CPT DSS FPDSO FPSO FPU FSO ICL MIP MODU NC OCR OLT REB RTA SCL SCC SEM SEPLA SHANSEP SRD TOC TPT TLM TLP UHC VLA VST YSR

Alhayari, S., Chen, J.H. & Maerten, B. 2000. Kuito project: Successful suction pile design and installation based on innovative soil investigation. Proc. Offshore West Africa Conference, OWA, Abidjan. Andersen, K.H. & Jostad, H.P. 2002. Shear strength along outside wall of suction anchors in clay after installation. Proc. Int. Seminar on Offshore and Polar Engineering, ISOPE, Kyushu. Andersen, K.H. & Jostad, H.P. 2004. Shear strength along inside of suction anchor skirt wall in clay. Proc. Offshore Technology Conference, OTC paper 16844, Houston. Andersen, K.H., Murff, J.D., Randolph, M.F., Clukey, E.C., Erbrich, C., Jostad, H.P., Hansen, B., Aubeny, C., Sharma, P. & Supachawarote, C. 2005. Suction anchors for deepwater applications. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Baudet, B.A. & Ho, E.W.L. 2005. Modelling the effects of structure in deep-ocean sediments. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Baudet B.A. & Stallebrass S.E. (2004). A constitutive model for structured clays. Géotechnique 54 (4), 269–278. Borel, D., Puech,A., Dendani, H. & De Ruijter, M. 2002. High quality sampling for deep water geotechnical engineering: the STACOR experience. Proc. Int. Conference on Ultra Deep Engineering Technology, UDET, Brest. Borel, D., Puech, A., Dendani, H. & Colliat, J.L. 2005. Deepwater geotechnical site investigation practice in the Gulf of Guinea. Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Borel, D., Puech, A. & Po, S. 2010. A site investigation strategy to obtain fast-track shear strength design parameters in deep water soils. Proc. 2nd Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG, Perth. Bourillet, J.F., Damy, G., Dussud, D., Sultan, N. & Woerther, P. 2007. Behaviour of a piston corer from accelerometers and new insights on quality of the recovery. Proc. Int. Conference on Offshore Site Investigation and Geotechnics, SUT OSIG, London. Broadbent, F.E. 1953. The soil organic fraction. Adv. Argon, 5, 153–183. Brausse, M. 2001. Underconsolidation and apparent overconsolidation in marine sediments: a study of stress states in the northwestern Gulf of Mexico. Civil Engineering MsC Thesis, University of Rhode Island. Bruton, D., White, D., Cheuk, C., Bolton, M. & Carr, M. 2006. Pipe-soil interaction behavior during lateral buckling, including large amplitude cyclic displacement test by the Safebuck JIP. Proc. Offshore Technology Conference, OTC paper 17944, Houston. Bruton, D., White, D.J., Langford, T.L. & Hill, A. 2009. Techniques for the assessment of pipe-soil interaction forces for future deepwater developments. Proc. Offshore Technology Conference, OTC paper 20096, Houston. Buckley, D.E., MacKinnon, W.G., Cranston, R.E. & Christian, H.A. 1994. Problems with piston core sampling: Mechanical and geomechanical diagnosis. Marine Geology, 117, 95–106. Burland, J.B. 1990. On the compressibility and shear strength of natural clays. Géotechnique, 40 (3), 329–378.

Apparent Over-Consolidation Ratio Ball Penetration Test Compliant Tower Cone Penetration Test Direct Simple Shear Floating Production Storage Drilling & Off-loading Floating Production Storage & Off-loading Floating Production Unit Floating Storage & Off-loading Intrinsic Compression Line Mercury Intrusion Porosimetry Mobile Offshore Drilling Unit Normally Consolidated Over-Consolidation Ratio Oil off-Loading Terminal Reverse End Bearing Riser Tower Anchor Sedimentation Compression Line Sedimentation Compression Curve Scanning Electron Microscope Suction Embedded Plate Anchor Stress History And Normalized Soil Engineering Properties Soil Resistance to Driving Total Organic Carbon T-bar Penetration Test Taut Leg Mooring Tension Leg Platform Ultimate Holding Capacity Vertically Loaded Anchor Vane Shear Test Yield Stress Ratio

ACKNOWLEDGEMENTS The authors thank Total EP, Fugro France and IFP for the permission to publish this paper. Parts of the work published come from various R&D programmes conducted in the framework of CLAROM (CLub pour les Actions de Recherche sur les Ouvrages en Mer) with financial support from the private programme for the oil & gas E&P industry CITEPH (Programme de Concertation pour l’Innovation Technologique dans l’Exploration et la Production des Hydrocarbures). Projects were carried out with the joint participation of Acergy, Doris Engineering, Ifremer, Saipem, Technip and Total, and in collaboration with several University laboratories. The contribution of colleagues from these organizations is acknowledged. The Authors are particularly indebted to V. De Gennaro and


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Geotechnics for subsea pipelines D.J. White Centre for Offshore Foundation Systems, University of Western Australia, Perth

D.N. Cathie Cathie Associates, Brussels, Belgium

ABSTRACT: The geotechnical analysis performed for subsea pipeline design involves challenges that are not common in conventional foundation engineering. This paper reviews recent research in pipeline geotechnics and shows examples of how this research is being applied in practice. A general theme running through this paper is the twin challenges of the changes in seabed topography and the changes in soil properties that occur through the installation and operating life of a pipeline. Results from in situ and element testing of soils that replicate the loading and disturbance imposed by pipelines are used to show that significant changes in strength are induced. Soil generally weakens during the episodes of remoulding that accompany pipeline laying, buckling, walking and storm loading, and during ploughing and trenching. The soil strength recovers during subsequent episodes of reconsolidation between storms, and between startup and shutdown events. Solutions for incorporating this behaviour into the estimation of axial and lateral pipe-soil resistance, and the assessment of trenching and ploughing operations, are discussed. A unifying theme is the relative magnitude of drained and undrained soil strengths, the evolution of these strengths through cyclic episodes, and the importance of recognising the widelyvarying rates of shearing involved in pipe-soil processes. Pipeline geotechnics can involve drained behaviour in fine-grained clayey soils – for example, during slow axial expansion of pipelines – and undrained behaviour in coarse-grained soils – for example during ploughing. Concepts from critical state soil mechanics often provide a simple framework for clarifying this behaviour.

1

INTRODUCTION

soft fine-grained soils, where the management of thermal and pressure-induced expansion is a critical design issue. The second is the stability of light large-diameter pipelines in shallow water, where primary and secondary stabilization measures represent a significant capital expenditure, and where geotechnical analysis techniques are not well established. The first scenario is relevant to pipeline design in almost all deepwater frontiers globally.The second scenario is particular relevant off the coast of Australia, where many hundreds of kilometers of gas trunkline are currently planned, to bring gas from deepwater fields to onshore LNG facilities. Section 1.2 provides a brief overview of some key pipeline design considerations that have a strong geotechnical influence. This overview is intended to provide a brief introduction to some of the most novel aspects of pipeline geotechnics, for those who are unfamiliar with this area of geotechnics. In Section 1.3 comparisons are made between pipeline geotechnics and conventional foundation engineering. Some of the most relevant aspects of soil behaviour are highlighted in Section 2, using recent experimental observations, including some from unusual forms of penetration testing. These observations provide a backdrop to the mechanisms and analysis techniques that are outlined in Sections 3, 4 and 5 for assessing

1.1 Scope of paper The purpose of this keynote paper is to set out the challenges of pipeline geotechnics and to highlight some recent developments in this area that the authors have been involved with. This is not an exhaustive treatment of the subject, but is intended to – provide an overview of areas in which intensive research has recently been published. – highlight certain novel analysis techniques for design that this research is beginning to permit. More complete introductions to pipeline geotechnics are provided by the relevant chapters of books on offshore geotechnical engineering by Randolph and Gourvenec (2010) and Dean (2010). Cathie et al. (2005) presented a more exhaustive review of research across the whole of pipeline geotechnics. The present paper is intended to be of value to pipeline engineers as well as geotechnical specialists, and includes some basic geotechnical content where we consider this useful. The topics in this paper are generally relevant to one of two design scenarios. The first is high pressure high temperature pipelines laid in deepwater on


Figure 1. Some geotechnical aspects of pipeline design.

pipeline embedment, lateral pipe-soil resistance and axial pipe-soil resistance respectively. Sections 6 and 7 are focussed on the geotechnics of pipeline trench construction by ploughing and jetting respectively. Space limitations preclude discussion of in-trench pipeline stability, and the associated upheaval, backfill liquefaction and flotation issues. The paper finishes with brief conclusions. 1.2

Pipe-soil interaction processes

Many of the areas of pipeline design that have geotechnical aspects are illustrated in Figure 1. Offshore pipelines are often left on the seabed, unburied, if this does not lead to unacceptable instability under hydrodynamic loading. The interaction between the pipeline and the seabed feeds into many aspects of the pipeline design. If the pipeline must be buried, for stability or to avoid fishing gear, the shielding of the pipeline via the construction of a trench (possibly backfilled) requires geotechnical design. On-bottom pipelines are increasingly being designed to allow movement during their operation, either under hydrodynamic loading or under thermal and pressureinduced expansion. Steel catenary risers, which are extensions of pipelines that connect to surface facilities, inevitably move where they touchdown on the seabed, in response to oscillation of the floating facility. Throughout the lay process and during subsequent operating cycles, the pipeline is subjected to geotechnical forces where it is in contact with the


seabed. A pipeline is a forgiving structure, being able to tolerate significant deformation and gross movements across the seabed, except at points of fixity such as end terminations. Such instabilities are unacceptable for the platforms and foundations that conventional geotechnical engineering is equipped to design. Indeed, if a pipeline was not permitted to move across the seabed under thermal loading, this would often induce unacceptable thermally-induced stresses. The forgiving flexibility of a pipeline does not, therefore, alleviate the need to quantify the pipesoil resistance forces to a sufficient accuracy that the robustness of the design is demonstrated. One of the most difficult aspects of pipeline design, which is an increasing challenge as operating temperatures and pressures rise, is the management of thermal and pressure-induced loading. Controlled on-bottom lateral buckling is an attractive design solution but one which requires the pipe-soil responses to be bracketed: both high and low geotechnical resistance can hamper a design (Bruton et al. 2007; AtkinsBoreas 2008). A second and related behaviour that arises from the thermal and pressure-induced loading is the tendency for pipes to ‘walk’ axially over cycles of startup and shutdown (Tornes et al. 2000; Carr et al. 2006). This phenomenon can be driven by the asymmetry of the heat-up and cool-down processes or by the presence of a seabed slope or end-of-line tension (which creates an asymmetry in the mobilized axial pipe-soil resistance).Accurate assessment of the axial pipe-soil resistance forces is required for robust modelling of this process.

Table 1.

Comparison of pipeline geotechnics and conventional foundation engineering (after White and Gaudin 2008). Foundation

On-bottom pipeline

Problem geometry

Known, controlled.

Design criteria for in-service behaviour

To remain fixed, movement uD.

Surrounding soil conditions

Similar to in situ state. Relatively unaffected by installation.

Soil-structure interaction

Usually minimal. Imposed loads are not strongly affected by foundation displacements. Scour and wave-induced liquefaction may require mitigation Usually available. Can assume lowest credible geotechnical capacity

Uncertain. Embedment affected by lay process and metocean conditions. Subsequent pipeline movements disturb seabed topography. May be required to displace significantly, u D, through hundreds of cycles of operation or hydrodynamic loading. Soft soil is significantly affected by installation. Remoulding, heave and reconsolidation affect the local strength. Often significant. Local pipe-soil load-displacement relationship affects overall pipeline response.

Soil-ocean interaction Single conservative design approach

Another significant design issue that is particularly relevant in the shallow waters offshore Australia, is pipeline stability under hydrodynamic loading from storm-induced currents and waves. In this situation a conservative approach is to adopt a low value of soil resistance. However, the cost of stabilization measures such as concrete coating is huge, and there is a strong incentive to refine the geotechnical analysis to remove any unnecessary conservatism in the design seabed resistance. In shallow water a pipeline may require additional – ‘secondary’ – stabilisation for hydrodynamic stability. Secondary stabilisation solutions revolve around reducing the hydrodynamic loading and increasing the available lateral resistance. An open trench provides partial shielding from hydrodynamic load. Burial of the pipe eliminates direct hydrodynamic loading (although soil liquefaction under hydrodynamic loading can destabilise a buried pipe). Geotechnical assessments must be made of the trenching process – which may be by ploughing, cutting, jetting, dredging or a combination. Other secondary stabilisation techniques include continuous rockdumping, or engineered solutions to provide local anchoring at intervals along the pipe. These solutions include flexible concrete mattresses, anchor blocks or saddles placed over the pipeline, or small piles on either side of the pipeline. The stability of these objects must also be assessed in design, taking account of the additional cyclic loading transferred to them by the unstable pipeline.

foundations and piles. This is partly because it is only recently that design codes have permitted gross pipeline movement, and so designers have not needed to explicitly assess the interaction forces as pipelines sweep across the seabed. Also, it is only recently that some of the complexities of the underlying soil behaviour have been recognized. A generally accepted framework for routine analysis has not emerged. The contrasts between pipeline geotechnics and conventional foundation engineering are summarised in Table 1 and illustrated in Figure 2. The designer’s task in the geotechnical design of a pipeline is aided by the structure’s tolerance of movements and mild deformation, but is hampered by the difficulty of assessing the geometry of the scenario and the operative soil properties. The laying of a pipeline and any subsequent lateral or axial movements disturb the topography of the seabed. The changed geometry and the altered soil properties need to be captured in calculations of the available pipe-soil resistance. Even the intact soil properties are difficult to establish at the shallow embedments relevant to pipeline geotechnics. Undisturbed sampling of soft near-surface soils is difficult and penetrometer tests at shallow embedment require particular interpretation techniques (Puech and Foray 2002, White et al. 2010a). Some soil properties and parameters such as friction angles and undrained strength ratios tend to be different at the very low stress levels relevant to pipeline geotechnics. A further complication is that interaction between the ocean and the seabed – leading to scour and liquefaction – can be significant in shallow water. The result is a tripartite interaction between the ocean, the pipeline and the seabed, which is illustrated in Figure 3. Cross-disciplinary design approaches for

1.3 Comparison with foundation engineering Geotechnical design procedures for pipelines and risers are relatively undeveloped compared to


Scour and wave-induced liquefaction can dominate behaviour Often unavailable. Both upper and lower bound geotechnical capacity may adversely effect structural response

Figure 3. Tripartite interaction between the seabed, the ocean and a pipeline in shallow water during storms.

buckles to sweep laterally across the seabed – and storms – which create high hydrodynamic loading. Over the operating life of a pipeline, the surrounding soil may therefore be subjected to a large number of episodes of disturbance followed by recovery and reconsolidation. The soil strength will generally fall and rise with each episode, and the net effect may be an overall increase or decrease in the strength of the soil. Coupled with the associated changes in seabed topography, this may lead to a rise or a fall in the geotechnical restraint on the pipeline. Figure 4a illustrates the case of a buckling pipeline on a fine-grained soil (in deepwater, where hydrodynamic loading is negligible). Each startup or shutdown of the buckle causes gross monotonic remoulding of the surrounding soil, comparable to the high disturbance created during passage of a penetrometer. The startup and shutdown episodes are separated by a period of time which may or may not be sufficient for full dissipation of the excess pore pressures generated during the previous disturbance – depending, obviously, on the consolidation characteristics of the soil relative to the frequency of the startups and shutdowns. Typical pipeline designs require several hundred or even one thousand shutdown and startup events to be considered. Figure 4b illustrates the case of a pipeline on a sandy or silty soil in shallow water, where storms create high hydrodynamic loading, which in turn causes the pipeline to exert cyclic loads on the surrounding soil. A storm loading event is perhaps best considered as a pre-failure cyclic disturbance as distinct from the gross (undrained) monotonic remoulding of the previous example. In this illustration the cyclic loading leads to a weakening, associated with pore pressure buildup, which is subsequently compensated for by reconsolidation and an associated densification of the soil. Storms occur at a frequency such that full dissipation occurs between events, for the soil considered here. In design, it is necessary to consider the effect

Figure 2. Comparison of pipeline geotechnics and conventional foundation engineering (images from Jayson et al. 2008 and Fisher and Cathie 2003).

this interaction are in their infancy. There is not yet a routine basis for assessing pipeline stability under hydrodynamic action that incorporates all three interactions concurrently (Damgaard and Palmer 2001, Cheng et al. 2010).

2 2.1

RELEVANT SOIL MECHANICS Illustrations of soil behaviour near pipelines

The soil close to a pipeline is grossly disturbed as the pipe is laid. If that disturbance happens rapidly enough for excess pore pressure to be generated then the subsequent reconsolidation process generally leads to an increase in the strength and density of the soil. Subsequent events may disturb the pipeline and the surrounding soil further. These events include the startup and shutdown of the pipeline – which cause


Figure 5. Changes in the undrained penetration resistance of fine-grained soils during cyclic penetrometer tests (after Gaudin and White 2009).

remoulded state, with this ratio being termed the sensitivity. Much higher sensitivities can sometimes be found, particularly in carbonate soils. In the analysis of problems involving significant disturbance, it is necessary to identify the relevant soil strength, which may lie somewhere between the intact and remoulded values. Cyclic T-bar or ball penetrometer tests allow this behaviour to be quantified. The progressive reduction in net bearing resistance through cycles of disturbance is shown for various different soils in Figure 5. The strength degrades exponentially with the number of cycles of disturbance, which allows the response to be characterized by two parameters – the sensitivity, St (or its inverse, δrem ) and a parameter related to the ductility, N95 (which is the number of cycles of disturbance after which the resistance has decayed by 95% of the difference between the intact and remoulded values). It is evident that both of these parameters vary significantly between soil types (particularly St ), but the form of the decay is similar. Analysis techniques for design can capture the decaying soil strength by adopting a value that represents the relevant level of disturbance. The general approach is to firstly convert the penetrometer ductility parameter, N95 , to an equivalent strain level (e.g. Zhou and Randolph 2009). The relevant strain level for the problem being considered is then used to deduce the operative undrained strength. Such techniques have been proposed for spudcan penetration (Erbrich 2005, Hossain and Randolph 2009). These methods utilise the type of strength degradation curves shown in Figure 5 to link the strains and operative strength around a spudcan to those around a T-bar. T-bar and spudcan penetration involves a comparable level of disturbance to monotonic pipe embedment. However, the dynamic motions that accompany pipe laying mean that a greater level of disturbance and hence a lower operative soil strength is applicable, compared to that mobilised during initial T-bar penetration. The dramatic reductions in strength evident in Figure 5 are partly due to the generation of positive

Figure 4. Illustrative histories of soil element behaviour near unstable pipelines: episodes of disturbance and recovery.

of these disturbances on the available lateral pipe-soil resistance – which controls the stability. This stability is affected by changes in the pipeline embedment as well as the changes in the strength of the soil around the pipe during disturbance. The same history of disturbance and reconsolidation shown in Figure 4b is applicable to the backfill above a trenched pipeline. These illustrations of the history of soil behaviour are clearly very idealized, and it is rare that the related changes in soil strength are tracked explicitly within a design analysis. However, it is important to recognize these effects, since they have a significant influence on the geotechnical restraint on a pipeline. These illustrations provide a convenient background to the following three examples of soil element behaviour. These examples show the changes in soil strength that can accompany episodes of disturbance and recovery, and also showcase novel testing techniques that may in the future be utilized to quantify this behaviour for design. The examples are: – episodes of undrained disturbance and reconsolidation seen by a T-bar penetrometer; – episodes of undrained disturbance and reconsolidation seen by a vertical rod penetrometer; – drained and undrained failure of different soils; – low stress friction response of fine-grained soils.

2.2 Disturbance and recovery: T-bar tests The strength of soft fine-grained seabed soils typically reduces by a factor of 2–5 from the intact to the fully


91

Figure 6. Undrained strength through episodes of remoulding and reconsolidation (test in lightly overconsolidated kaolin in the UWA beam centrifuge) (White and Hodder 2010).

Figure 7. Critical state interpretation of episodes of remoulding and reconsolidation (White and Hodder 2010).

the attraction of setting this behaviour within an effective stress framework, but it does rely on a very crude simplification of the overall behaviour, in which the response of all the elements of soil around a penetrometer is lumped into a single representative effective stress level and specific volume. More refined models will allow this behaviour to be more accurately quantified and numerical simulations will test the validity of this simplification. The key point, however, is that the rises in soil strength during episodes of reconsolidation can eclipse the reductions in soil strength during the preceding episodes of remoulding.

excess pore pressure in these contractile materials during undrained shearing. As this positive pore pressure dissipates and the effective stress rises back to the geostatic state the material densifies and the subsequent undrained shear strength may be higher. Cyclic T-bar penetrometer tests with periods of reconsolidation between episodes of cycling show this regain in strength. Figure 6 summarises the results of a cyclic T-bar test in kaolin clay reported by White and Hodder (2010), expressing the T-bar strength at a particular depth – 2.25 m – during each cycle. After just three episodes of full remoulding and reconsolidation, the current remoulded strength was comparable to the original intact strength. These results quantify the contrasting effects of disturbance and recovery shown in Figure 4a for this soil and the particular disturbance pattern imposed by a T-bar. This behaviour is easily understood within a critical state-type framework, since this provides an explicit link between moisture content (which reduces as positive pore pressures dissipate) and undrained strength. This interpretation can be extended to a quantitative treatment, expressed in terms of the operative strength averaged over all of the soil near the penetrometer (rather than of a single soil element). An accurate back-analysis of the results shown in Figure 6 can be achieved by defining two failure lines in stressvolume space, which represent the intact and fully remoulded strengths of the soil, as proposed by White and Hodder (2010). As shown in Figure 7, this back-analysis of the T-bar resistance at a depth of 2.25 m (which corresponds to an in situ effective vertical stress of σvo = 12 kPa) is based on the intact strength line (ISL) being reached during the initial T-bar stroke of a episode, and the effective stress point migrating towards the remoulded strength line (RSL) according to an exponential trend (i.e. the reduction in effective stress per T-bar stroke is proportional to the difference between the current effective stress and the effective stress at the remoulded state for the current specific volume). This analysis has

2.3

The example above involves only 3 episodes of reconsolidation. A significantly larger number of episodes of reconsolidation are involved in the second example. A novel vertical rod penetrometer has been used on a recent centrifuge project at UWA, with the aim of quantifying the resistance and strength of surficial material, as the soil is forced to flow past the penetrometer. This device is a cylindrical bar, oriented vertically, 4 mm in diameter. The device is embedded until the tip is typically 5 – 10 diameters below the soil surface. The bar is equipped with multiple levels of strain gauging located above the soil surface, which allow the magnitude and distribution of the pressure on the bar to be derived. In one test the bar was embedded in soft kaolin to a depth of 45 mm then cycled laterally by a distance of 20 mm at a rate of 0.3 mm/s. This rate corresponds to a dimensionless velocity of vD/cv ∼ 10 which is almost fully undrained, based on the limits demonstrated by Finnie and Randolph (1994) (albeit for a different geometry of problem). The elapsed time between the bar passing the mid-point of each lateral stroke was 100 seconds, which corresponds to a dimensionless consolidation time of T = cv t/D2 = 0.5. This value is indicative of significant (∼50%) pore pressure dissipation, based on limits provided by Randolph (2003) (again for a slightly different geometry of problem).


Disturbance and recovery: vertical rod tests

Figure 8. Lateral resistance on a vertical bar penetrometer through episodes of disturbance and reconsolidation.

Figure 9. Drained and undrained resistance during penetration.

The resulting variation in the average lateral resistance on the bar penetrometer is shown in Figure 8. Initially the resistance reduces, as the remoulding damage exceeds the recovery from reconsolidation. However, a steady rise in resistance is soon evident, with the strength after several episodes exceeding the initial (intact) value. After this test the level of the soil surface around the penetrometer had lowered. This is evidence that the soil surrounding the device had densified through the episodes of reconsolidation, which is consistent with the changing strength. This example highlights again the changes in soil strength that can occur through the episodes of disturbance and recovery that can be imposed on the soil surrounding a seabed pipeline. As shown in Figure 8, the strength continued to increase through more than 25 cycles of heavy remoulding, but ultimately should reach a limiting value, when the soil has densified sufficient that there is no longer a tendency to generate positive excess pore pressure when disturbed.

plough share (Peng and Bransby 2010). In all cases the velocity is normalized by the coefficient of consolidation of the seabed and an appropriate drainage distance – generally the size of the foundation. There is some variation in the dimensionless velocities at which fully drained and fully undrained behaviour occurs, although this may be partly due to the difficulties in establishing appropriate values of the coefficient of consolidation. There are more significant differences between the relative magnitude of the drained and undrained resistance. These arise from the state of the soil – and hence its tendency to generate positive or negative pore pressure in undrained conditions – and also the particular boundary value problem. For example, for a soil with a particular undrained strength and angles of friction and dilation, the relative magnitude of the drained and undrained bearing capacity depends on both the applicable bearing factors (i.e. Nq and Nγ for drained conditions and Nc for undrained conditions) as well as the soil strength properties. The relative magnitude of the drained and undrained strengths of an interface is not affected by the geometry of the problem, so provides a more simple differentiation between the drained and undrained behaviour of a particular soil (albeit in combination with a particular interface). Figure 10 shows the steady residual resistance measured during monotonic shearing of normally consolidated kaolin clay over a rough steel surface at different velocities. The kaolin was normally consolidated to a stress of 2.5 kPa prior to shearing. These tests used a direct shear box at UWA that has been modified to operate at the low stress levels relevant to pipeline geotechnics. The four tests show a trend of increasing resistance with reducing velocity. This trend is consistent with a hyperbolic backbone curve of the same form as used in Figure 9, drawn between the fully drained and fully undrained limits. These limits correspond to an interface friction angle of 30◦ and an undrained = 0.25, both of which are constrength ratio of su /σvc sistent with other published results for this stress level

2.4 Drained and undrained soil responses The shear strength of a given soil in a particular state depends on whether drained or undrained conditions are imposed, as well as the mode of shearing. In conditions in which drainage is permitted, the shearing can be imposed at rates that span from fully drained (i.e. in which no excess pore pressure builds up) to fully undrained (i.e. in which effectively no pore pressure dissipation occurs, despite drainage being permitted). Various authors have recently explored the continuous variation in mobilized soil strength and geotechnical resistance between drained and undrained conditions. Results summarized in Figure 9 show the variation in penetration resistance of circular surface foundations (Finnie 1993), and cone penetrometers in soft clay (Randolph and Hope 2004) and dense silt (Silva 2005). Similar relationships have been derived for the uplift resistance of buried pipelines (Bransby and Ireland 2009) and the sliding resistance of a pipeline


Figure 11. High effective stress friction and a non-linear failure envelope during low stress interface tests and axial pipe-soil movement (data from Bruton et al. 2009, White et al. 2010b).

Figure 10. Interface shear resistance at varying rates, from drained to undrained.

(Bolton and Barefoot 1997; Pedersen et al. 2003; Bolton et al. 2009). The movements of an on-bottom pipeline in a given soil may span the ranges of velocities and therefore drainage conditions shown in Figure 9 and Figure 10. Thermal expansions and the associated lateral movements span from zero at ‘virtual anchor points’ to millimetres per second of axial movement near ends and buckles, and even metres per second of lateral movement during initiation of lateral buckles. A further range of velocities arises from pipeline movements driven by hydrodynamic action – which include oscillations in the touchdown zone during laying that are created by vessel motion, or oscillations in response to direct hydrodynamic action on the pipe during storms. A high velocity applies during the ploughing of a pipeline trench. In combination with the large size of a ploughshare, this leads to undrained conditions even in sands. As a consequence of the varying velocities involved in these processes, it is common for the geotechnical analysis for pipeline design in a fine-grained soil to require an assessment of the drained response. Conversely, in coarse-grained soils an assessment of the undrained behaviour can be required. Also, there are often occasions when the actual response involves partial drainage, and it is necessary to tie together drained and undrained assessments in order to predict the most likely behaviour, and the potential range of responses.

2.5

The same trend appears in low stress soil-soil and soil-interface shearing of fine-grained soils (Pedersen et al. 2003, White and Randolph 2007, Hill and Jacob 2008, Bruton et al. 2009, White et al. 2010b). Nonlinear failure envelopes that express the friction angle or limiting stress ratio as a function of effective stress can capture this variation (e.g. Figure 11). It is important to recognise that the friction angles measured at conventional geotechnical stresses may not be appropriate for the assessment of drained pipe-soil resistance. The high friction angles are also reflected in the higher undrained strength ratios found at low stresses – a link highlighted in the approximate expression for normally consolidated undrained strength ratio su /σvc = φ/100 derived from the analysis of Wroth (1984), where φ is the friction angle in is the consolidation stress. Changes in degrees and σvc friction angle affect both the drained and undrained strengths of soil, since the underlying behaviour is principally frictional. 2.6

As well as the conventional aspects of soil behaviour that are considered in the analysis of foundations, pipeline geotechnics is also often concerned with a greater degree of soil disturbance, and intervening periods of recovery and reconsolidation. These processes also take place at lower stress levels compared to conventional geotechnics, and can lead to significant changes in the state and therefore the strength of a soil through the operating life of a pipeline. Also, due to the relevant drainage distances and rates of movement, it is can be necessary to focus on the drained response of fine-grained soils and the undrained response of coarse-grained soils.

Low effective stress friction

A final feature of soil behaviour that is particularly relevant to pipeline geotechnics is the variation in friction angle with stress level. At the low stresses relevant to pipelines, higher friction angles are found compared to more usual geotechnical stress levels. The peak friction angle of sands increases with reducing stress level (Bolton 1986). Results from experiments performed on Earth (Fannin et al. 2005) and onboard the space shuttle (Sture et al. 1998) also show that the critical state or constant volume friction angle is higher at very low stresses.

3 3.1

PIPELINE EMBEDMENT Pipelaying mechanics

The as-laid embedment of a pipeline affects the subsequent pipe-soil resistance, as well as the thermal


Summary of soil behaviour

Figure 12. Pipeline laying notation (Randolph and White 2008a).

Figure 13. Maximum stress concentration factors for pipe laying on an elastic seabed (Randolph and White 2008a).

insulation. The installation process involves soilstructure interaction, since the maximum vertical pipe-soil force during the lay process will exceed the submerged pipe weight, W , by an amount that depends on the seabed stiffness and the geometry of the catenary created by the S-lay or J-lay arrangement. The configuration of a pipeline during laying is shown in Figure 12. A key parameter is the horizontal component of tension, T0 , which is constant through the suspended part of the pipeline and can be assessed from the pipe weight, water depth and hang-off angle. The maximum contact force (per unit length) with the seabed, Vmax , and hence the local force concentration factor, flay = Vmax /W , is a function of the seabed stiffness, k (defined as the secant ratio of force per unit length, V, to embedment, w) in addition to the bending rigidity, EI, and T0 . The force concentration factor reduces with increasing water depth and decreasing seabed stiffness. A characteristic length, which relates to the length over which the bending stiffness moderates the catenary behaviour, is given by λ = (EI/T0 )0.5 . Parametric solutions for the static lay conditions have been presented by Randolph and White (2008a), who showed that for horizontal tension of T0 > 3λW (which holds for most pipelines), results from analytical solutions (Lenci and Callegari 2005) and numerical analysis using OrcaFlex (Orcina 2008) all converge to unique design lines. The value of flay may be expressed approximately as (Figure 13):

Section 3.4, but firstly the significant influence of dynamic pipe movements during the laying process is highlighted. During J-lay or S-lay installation, dynamic movement of the pipe occurs within the touchdown zone, driven by the vessel motion and hydrodynamic loading of the hanging pipe. These loads induce a combination of vertical and horizontal motion of the pipeline at the seabed (Lund 2000, Cathie et al. 2005). In addition to vessel motion due to swell and waves at the sea surface, cyclic changes in pipeline tension may occur (depending on the accuracy of the tensioning system) if the offloading of the pipe is not smoothly coincident with the vessel advancement. This dynamic movement, although often of very small amplitude, leads to local softening of the seabed sediments and can push soil away to either side of the pipe alignment, creating a narrow trench in which the pipe becomes embedded. An illustration of the significant additional embedment that can occur simply due to small amplitude cyclic motions is shown in Figure 14. These results are from a centrifuge model test on lightly overconsolidated kaolin clay (Cheuk and White 2010a). A model pipe was penetrated to a normalised embedment of w/D = 0.1 (point A), when the vertical pipe-soil load was fixed constant (the normalised vertical load, V/su D reduced with pipe embedment, due to the increasing su with depth). A series of packets of horizontal oscillations were then imposed, increasing in amplitude (Figure 14a). The adopted amplitudes of motion reflect ROV observations during laying, although these are dependent on the lay geometry and metocean conditions (Westgate et al. 2010a). The aim in this experiment was to represent dynamic lay motions in an idealised manner. The pipe initially settled at a rapid rate, by an amount that far exceeds that due to the combined vertical and horizontal loading alone. The lateral soil resistance mobilised during the first two cycles, when the embedment doubles (to point B), corresponds to an equivalent friction factor of H/V < 0.25. As the embedment increases, a greater soil resistance is mobilised for a given amplitude of

3.2 Seabed disturbance during pipelaying Equation 1 is derived based on a single value of secant seabed stiffness, V/w, which would be applicable to a purely elastic seabed. The actual seabed response is generally non-linear during vertical penetration, and is stiffer during unloading since the seabed has been plastically deformed. As a result, the actual operative secant stiffness varies along the touchdown zone. Theoretical solutions for monotonic vertical pipe penetration into undrained soil are described in


95

motion, reflecting the increasing constraint on the pipeline. The softening of the surrounding soil is evident in the reductions in lateral resistance at points C and D, with increasing disturbance. Complementary assessments of the level of soil remoulding during pipeline laying can be made using large deformation finite element analysis. A study reported by Wang et al. (2009) replicated the first stage of the model test shown in Figure 14 (with horizontal motions of +/−0.05D) using continuum finite element analysis. The soil constitutive model included softening through a reduction of the undrained strength with accumulated plastic strain, in a manner consistent with the behaviour shown in Figure 5. The resulting patterns of lateral resistance and embedment and the local soil remoulding are shown in Figure 15. Even small lateral motions of just +/−0.05D lead to a remoulded zone that extends by almost one pipe diameter to each side, and the pipe itself rests on fully remoulded soil. These results highlight the importance of assessing pipeline embedment using an appropriately degraded value of soil strength, as well as accounting for the catenary overstress via Equation 1 or some comparable approach. 3.3 As-laid pipeline survey observations Similar conclusions can be drawn from the results of ROV surveys following pipe laying. The variation in embedment along a pipeline varies for a variety of reasons including local variations in the seabed strength. However, consistent variations have also been identified due to wave height, lay rate, downtime events, and changes in lay angle. The first three of these effects influence the level of dynamic motion that the pipe is subjected to in the touchdown zone and the latter effect influences the vertical stress concentration (Westgate et al. 2010a). The resulting range of embedment can be expressed as a statistical variation, and this range can be compared with calculations performed using theoretical solutions for the catenary overstress and the seabed penetration resistance, using both intact and remoulded soil strengths. Figure 16 shows the distribution of embedment from survey measurements taken at 1 m intervals along a 13 km long pipeline (excluding short lengths of pipeline that had a far greater embedment due to downtime events). The pipeline was laid on almost uniform deepwater soil conditions, in a water depth of 1215– 1450 m (Westgate et al. 2010b). The lay process took several days, during which the sea state varied with a significant wave height of between 0.6 m and 1.7 m. The calculated pipeline embedment based on the static catenary overstress (Equation 1) coupled with intact and fully remoulded strengths are highlighted, along with estimates based on a dynamic analysis of the vertical stress concentration performed using Orcaflex (Orcina 2008) (Figure 16). The fully remoulded strength coupled with the static overstress

Figure 14. Effect of lateral motions on pipeline embedment (after Cheuk and White 2010a).

matches well with the most frequent embedment, although the agreement was not as good for some other pipelines at the same site. This result and other comparisons from a limited range of post-installation surveys, suggest that the fully remoulded soil strength leads to reasonable estimates of the average pipe embedment for average lay conditions. However, there remains significant scatter between different pipelines in the same conditions and variations in embedment along a single pipeline (Westgate et al. 2010a, Westgate et al. 2010b). An additional effect appears to be that lighter pipelines are more susceptible to dynamic lay effects – probably due to their reduced inertia – and so a greater degradation in soil strength applies.


Figure 15. LDFE simulation of soil strength after dynamic pipe laying (after Wang et al. 2009).

breakout resistance since it is not immediately obvious what combination of input parameters will lead to upper and lower bound outcomes. It is also attractive to set the geotechnical components of pipeline analysis within a full probabilistic framework. This is consistent with the probabilistic structural reliability analyses that are increasingly performed during the assessment of pipeline on-bottom stability or lateral buckling. 3.4 Solutions for vertical pipe penetration Using an appropriate operative soil strength, su , the shallow embedment of a pipeline in undrained conditions can be calculated via a bearing capacity expression that reflects the appropriate geometry of failure mechanism. Experimental results presented by Dingle et al. (2008) revealed the internal soil failure patterns during pipe penetration (Figure 17a). These detailed observations, coupled with the vertical force-displacement response from other tests, have been used to validate numerical simulations (e.g. Figure 17b, Merifield et al. 2008a) and plasticity limit analyses (Randolph and White 2008b). The resulting bearing capacity, Vult , can be calculated as the superposition of components related to the soil strength at the pipe invert (via a bearing factor, Nc , that varies with embedment, pipe roughness and strength heterogeneity) and the soil buoyancy – enhanced by heave (Randolph and White 2008a, Merifield et al. 2009):

Figure 16. Distribution of as-laid embedment for a single pipeline (Westgate et al. 2010b).

The use of the remoulded soil strength over-predicts the embedment in the case of minimal pipeline motions (for example in calm weather or during lay down of the final catenary section of pipe) and under-predicts embedment during severe weather or downtime events, again based on limited field data (Westgate et al. 2009; Westgate et al. 2010b). Due to this inevitable scatter, including variability due to effects that cannot be predicted in advance of the pipelaying, assessments of pipeline embedment are subject to uncertainty. If only an upper or lower bound embedment is required for design then this uncertainty can be circumvented using conservative assumptions. However, both upper and lower bound embedments are usually critical for different design considerations. It is therefore necessary to perform lower and upper bound assessments using opposite extremes of the input parameters. These can be extended to a full probabilistic assessment, using statistical variations in input parameters, as illustrated in Section 4.4. Such an approach is desirable when assessing the subsequent

The nominally embedded cross-sectional area is denoted A , the soil submerged unit weight is γ and


Figure 19. Undrained failure envelopes for lateral breakout in uniform soil.

4

LATERAL PIPE-SOIL INTERACTION

4.1 Theoretical failure envelopes The same combination of experimental, numerical and analytical techniques has been used to derive failure envelopes in vertical-horizontal load space, which allow the lateral breakout resistance in undrained conditions to be assessed (Dingle et al. 2008; Merifield et al. 2008a, Randolph and White 2008b). Different forms of failure mechanism apply, depending on whether tension can be sustained at the rear of the pipeline (Cheuk et al. 2008; Merifield et al. 2008b) as shown in Figure 19. For the no-tension case (and w/D < 0.5), the failure envelopes pass through the V-H origin, and there is a ‘frictional’ cut-off corresponding to a failure mechanism involving the pipe riding up at an angle shown as θ in Figure 19, with no soil deformation occurring. In drained conditions, experimental results show that the failure envelope has a similar shape to the no-tension undrained case (Zhang et al. 2002a). Centrifuge modelling results from tests on soft clay show that the breakout resistance is well predicted by the unbonded failure envelopes if the pipe is installed with some level of dynamic movement during laying (such as the simulation in Figure 14) (Cheuk and White 2010b). In this case, for very soft clay, the appropriate undrained strength appeared to be the intact value. In this particular case the effects of remoulding and reconsolidation approximately cancel out. Without dynamic movement the breakout resistance is often higher, but is highly brittle, reflecting the opening of a crack behind the pipe prior to mobilisation of a full two-sided mechanism (Dingle et al. 2008). It is thought that any dynamic movement leaves a skin of weak remoulded soil at the pipe surface. This can fail in a local mechanism behind the pipe, even if some tension is sustained, providing negligible extra resistance above the unbonded case. Modified versions of these failure envelopes can be created for other boundary conditions, such as a pipe running parallel to a slope (e.g. Morrow and Bransby

Figure 17. Failure mechanisms during vertical pipe penetration (Dingle et al. 2008, Merifield et al. 2008a).

Figure 18. Bearing factors during vertical penetration into heterogeneous and uniform soil (Chatterjee et al. 2010b).

the factor fb captures the enhancement of soil buoyancy beyond that given by Archimedes’ principle. A value of fb = 1.5 is typical. Solutions for the bearing factor, Nc , in this expression have been provided for uniform and linear soil strength profiles (Aubeny et al. 2005, Merifield et al 2008a), and with modifications to account for heave (Merifield et al. 2009, Chatterjee et al. 2010a, 2010b). A pair of bearing factor profiles derived from large deformation finite element analyses are shown in Figure 18, for two different profiles of soil strength (Chatterjee et al. 2010b).


Figure 21. Effect of load path on equivalent breakout friction.

load path reaches the unbonded failure point marked A, with the equivalent friction factor marked µeq . The load path marked B represents hydrodynamic loading, when the pipe is subjected concurrently to both horizontal drag and upwards lift. In this case the equivalent friction factor at failure is higher (although the constant V path marked A provides a conservative assessment). The theoretical failure envelopes of the form shown in Figure 19 can be re-expressed as equivalent friction factors, varying with load path and load level, W /Vult . For illustration, results have been derived from the unbonded envelopes corresponding to a smooth pipe interface, uniform soil strength and embedments of w/D = 0.25 and 0.45. Load paths varying from dV/dH = 0 to dV/dH =−2 (i.e. a lift force of twice the drag) have been used (Figure 20). This form of presentation clarifies how much the seabed response differs from the simple Coulomb friction that is often assumed. As the pipe weight reduces, the equivalent friction factor rises, particular at low load levels. The effect of load path is shown in Figure 21 which compares µeq for two ratios of dV/dH, using the pure drag case of dV/dH = 0 for normalisation. The load path has a significant influence on the equivalent friction factor at breakout. For the load path with an uplift of dV/dH = −2, the equivalent friction at failure is typically 1.5-2 times higher than pure drag at w/D = 0.25. The difference rises to >2.5 for w/D = 0.45. At low loads, however, the load path does not affect the equivalent breakout friction because the same tangential mechanism applies. This limiting friction factor at low loads that is caused by the tangential mechanism is not captured in conventional ‘friction + passive’ models for breakout resistance. The component of passive resistance in these models is uninfluenced by the vertical load level, leading to a finite breakout resistance under a vertical load of zero. This is incorrect and unconservative for stability design. The tangential failure mechanism leads to zero breakout resistance under zero vertical load for any embedment of w/D < 0.5, if tension cannot be sustained at the pipe-soil interface (regardless of the pipe roughness).

Figure 20. Equivalent friction factors from failure envelopes.

2009). For this case the failure envelope is rotated by the slope angle and changes shape slightly, depending on the ratio of soil strength to weight. Due to the curved shape of the failure envelopes, the equivalent friction factor for lateral failure in a downslope direction is not simply the value for level ground, adjusted by the slope angle according to Coulomb friction (unless failure is via the tangential mechanism). 4.2 Failure envelopes as friction factors Failure envelopes depict the limiting pipe-soil loads in a different manner to the empirical expressions that are commonly used in practice (e.g. DNV 2007a, Verley and Sotberg 1994, Verley and Lund 1995, Bruton et al. 2006). These approaches divide the soil resistance into frictional and passive resistance components that are superposed. It is then common for the resulting resistance to be expressed as an equivalent friction factor, by dividing by the vertical pipe-soil load. In this way, the geotechnical resistance is linked only to the pipe weight (and lift), and any structural analysis software need not include any soil parameters, nor incorporate explicit calculations of the pipe embedment. The theoretical failure envelopes can also be manipulated to express the pipe breakout resistance as an equivalent friction factor, µeq . For a given initial state, the equivalent friction at failure, µeq = H/V depends on the load path in H-V space. The initial state is typically well within the failure envelope (W /Vult qc , the cone tip resistance, can be used. Equations 11-13 can be combined to relate the pressure at the nozzle (p0 ) required to cut a soil of a given shear strength as a function of stand-off distance, considering that Equation 12 also applies to the pressure and velocity at the nozzle (following Bernoulli). As shown also by Machin and Allan (2010), the resulting relationship between standoff distance and the nozzle pressure to cause bearing failure is:

As a first approximation, x may also be considered the depth of cut that can be made for a jet located at the surface of the soil. The relationship is shown on Figure 53. With a jetting pressure of 20 bar from a 20 mm OD nozzle, in a soil of 150 kPa shear strength, a cut depth of about 0.2 m can be anticipated from each jet excluding any stand-off distance. From a practical perspective, considering a jetting sword of the type shown in Figure 48 with multiple jets, it is unlikely that zero stand-off distance can ever be achieved for all jets simultaneously. A minimum average stand-off might be 0.1 m if the trencher is being driven hard into the trench face. Assuming a minimum acceptable cut depth for each jet (of 20 mm OD) of 0.2 m gives x/d of 15. A typical 5 bar low pressure trenching system could be expected to operate acceptably in shear strengths of less than about 20 kPa. This agrees with general experience in the trenching industry. Low pressure (say 5 bar) systems which work effectively in sand can also create a trench in very soft to soft clays. As the bearing capacity failure occurs and the hole is deepened, the actual failure mechanism becomes more complex. The bearing capacity increases with depth, and the flow is no longer as shown on Figure 52 but is constrained inside the hole. The mechanism of entrainment of water leading to the reduction in jet velocity is quite different. However, Equation 14 is considered to provide a good indication of the ability of a static jet to cut the soil and experimental work does seem to confirm that this approach provides a conservative assessment (Machin and Allan 2010). Jet trenching involves the jets traversing rather than remaining static. Atmatzidis and Ferrin (1983) investigated the influence of time and traversing speed in the laboratory with a 1 mm diameter nozzle in clean sand, silty sand, silt and clay. The depth of jet penetration into the soil target and jet effectiveness was measured as a function of exposure time, the degree of saturation of the soil, the dry density of the soil and the traversing velocity of the jet over the soil target. Their research showed that the time required for the jet to approach maximum penetration in the soil target was about 15–20 seconds for a driving pressure

Figure 53. Required jet pressure for cutting.

Figure 54. Cutting depth as a function of jet trenching speed.

of 1000 psi (69 bar), being less for coarse-grained soil than fine-grained. It was found that the jet penetration depth, x, in a given soil is related exponentially to the corresponding time from initial impact. Machin and Allan (2010) indicate that it takes several seconds to reach the full cavity depth in clays. Atmatsidis and Ferrin (1983) also show that the penetration depth, x, could be related exponentially to the jet’s traversing velocity, v:


where xmax is the limiting penetration for a stationary jet after infinite time, and ζ is an empirical constant (with units of velocity) depending on jet and material properties. Combining the static and traversing equations provides an indication of the likely depth of cut that a specific jetting system could provide, if the value of ζ were known. Machin and Allan (2010) suggest that a quasiinstantaneous cavity depth is achieved for translation speeds exceeding 0.1–0.5 m/s (360–1800 m/hr) but slower speeds are required for maximum penetration. Figure 54 shows a typical result for cutting stiff clay (su = 100 kPa) considering that the quasiinstantaneous cavity depth is achieved at a speed of 500 m/hr (assumed).

soil deformations to be observed. Other insights have emerged through finite element analysis – with notable advances being the use of coupled methods in sand to capture undrained and partially-drained ploughing behaviour, and the use of large deformation techniques to capture gross remoulding of fine-grained soils during large lateral movements.

Successful lowering of a cable or pipeline also depends on the orientation of the jets so that the soil is actually cut into blocks which can be educted or transported away from the zone where the product must sink. As far as the authors are aware, the first discussion of this issue in the public domain is given in Machin and Allan (2010). 8

CONCLUSIONS ACKNOWLEDGEMENTS

This paper has reviewed various aspects of pipeline geotechnics, by reference to recent and emerging research activity from both academia and industry. The design challenges in pipeline geotechnics differ somewhat from conventional foundation engineering. Two particular challenges are the changes in both the seabed topography and also the soil properties that can occur throughout the installation and operating life of a pipeline: these are themes that run throughout this paper. Results from novel forms of repetitive in situ testing have been used to illustrate the response of soil to the forms of loading and disturbance that are induced by a pipeline. In soft fine-grained soils, these tests illustrate the balance between the reduction in strength from remoulding and the recovery that accompanies subsequent reconsolidation. Concepts from critical state soil mechanics provide a useful framework for capturing this behaviour. Solutions for incorporating this behaviour into the estimation of axial and lateral pipe-soil resistance, and the assessment of trenching and ploughing operations, are discussed. A common theme is the relative magnitude of drained and undrained soil strengths, the evolution of these strengths, and the importance of recognising the widely-varying rates of shearing involved in pipe-soil processes.The slow rates at which pipes move under thermal loading and the high rates at which trenching machines are driven mean that pipeline geotechnics often involves a drained response in fine-grained soils and undrained behaviour in sands. The conventional laboratory tests used to characterise these soils often require modification in order to extract the properties that are relevant for pipe-soil interaction. Some of the concepts discussed in this paper represent merely a snapshot of an evolving understanding, which will no doubt advance in the coming few years. For example, the changes in soil strength through episodes of disturbance and recovery, and through cycles of partially-drained interface shearing, have only recently been observed in experiments. The underlying mechanisms are not fully established and quantitative calculation methods for design use are in their infancy. New modelling technologies have been recently applied to pipeline geotechnics. The mechanisms of pipe-soil penetration and breakout behaviour have been quantified through sophisticated centrifuge model tests, which allow complex load sequences to be imposed on a model pipe and detailed internal

The work described here forms part of the activities of the Centre for Offshore Foundation Systems (COFS), established at the University of Western Australia in 1997 under the Australian Research Council’s Special Research Centres Program. COFS is now supported by Centre of Excellence funding from the State Government of Western Australia. The first author is supported by an Australian Research Council Future Fellowship (grant FT0991816) The assistance from a number of colleagues at UWA and Cathie Associates during the preparation of this paper is acknowledged. In particular, Fauzan Sahdi kindly provided the data in Section 2.3 from his PhD studies. The interface shear box tests in Section 5.2 were performed by Nat McNab assisted by Binaya Bhattarai. The MCWHIPLASH software (Section 4.4) was written by the first author with David Bonjean of Advanced Geomechanics, Perth. Helpful comments provided by George Zhang, also of Advanced Geomechanics, who reviewed a draft of this paper are also acknowledged. Some of the research described in this report has been guided by the SAFEBUCK Joint Industry Project, which is coordinated by David Bruton of AtkinsBoreas. The support of the SAFEBUCK participants is gratefully acknowledged. REFERENCES AtkinsBoreas. 2008. SAFEBUCK JIP: Safe design of pipelines with lateral buckling; design guideline. Report BR02050/C, AtkinsBoreas 252pp. Atmatzidis, D.K. and Ferrin, F.R. 1983. Laboratory investigation of soil cutting with a water jet, 2nd US Water Jet Conference, Rollo, Missouri, University of Missouri, 101–110. Aubeny, C.P., Shi, H., and Murff, J.D. 2005. Collapse loads for a cylinder embedded in trench in cohesive soil. ASCE Int. J. Geomechanics 5(4):320–325. Bolton, M.D. 1986. The strength and dilatancy of sands. Géotechnique 36(1):65–78. Bolton, M.D. and Barefoot, A.J. 1997. The variation of critical pipeline trench back-fill properties. Proc. of Conference on Risk-Based and Limit State Design and Operation of Pipelines, Aberdeen. Bolton, M.D., Ganesan, S.A. and White, D.J. 2009. SAFEBUCK Phase II: Axial pipe-soil resistance: summary report. Cambridge University Technical Services, Report for Boreas Consultants (SAFEBUCK JIP), ref. SC-CUTS-0705-R01. 54pp. Bransby, M.F. Yun, G.J. Morrow, D.R. and Brunning, P. 2005. The performance of pipeline ploughs in layered


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Axial and lateral pile design in carbonate soils C.T. Erbrich, M.P. O’Neill, P. Clancy & M.F. Randolph Advanced Geomechanics, Perth, Australia

ABSTRACT: Design of piles in carbonate soils remains one of the most challenging geotechnical problems, fraught with traps for the unwary. Whilst carbonate soils generally behave in accordance with the same underlying rules of soil mechanics that are applicable to all soils, they often exhibit characteristics at the extremes of this continuum of behaviour. Carbonate soils themselves straddle a wide range of different material types, varying from soft uncemented silts and muds to dense well cemented calcarenites and limestones. We also find dense uncemented carbonate soils and loose and compressible cemented soft rocks (calcarenites, etc.). All combinations of these are encountered offshore Australia. This paper presents new work that addresses two key aspects of pile design in this array of carbonate material types: axial design of drilled and grouted piles in compressible cemented carbonate soils and lateral design of piles (drilled and grouted or driven) in uncemented carbonate soils, varying from soft muds to dense sands. The new work on axial pile design builds on the extensive work undertaken in the 1980s following the problems with the North Rankin A platform, and which was subsequently put to good use in the design of the nearby Goodwyn A platform. However, more recent developments have identified a need to develop a more generalised approach that can be applied at sites with a variety of characteristics. In addition, more detailed review of some of the data collected during the original work has revealed important new mechanisms that are of considerable significance but have hitherto been neglected. The new work on lateral pile design follows a different tack from any of the existing methodologies used for assessing lateral pile response in such soils, which are essentially of a purely empirical form, with minimal theoretical basis. A new method with a robust theoretical basis is presented for assessing the lateral pile response in carbonate soils, which particularly accounts for their well known strong susceptibility to degradation (liquefaction) when subject to cyclic loading. 1

INTRODUCTION

The axial and lateral response of piles is analysed almost universally in the offshore industry using load transfer approaches. Interaction between the pile and the soil is treated through non-linear springs distributed down the pile shaft, with each horizontal layer of soil treated independently. The load transfer curve thus represents the integrated response of the soil layer extending outwards from the pile. The early development of load transfer curves followed an empirical approach, derived from the results of instrumented pile load tests, with limited guidance on how to scale the curves for different pile diameters or soil stiffness. More rational frameworks have since been established, aided by modern numerical analysis, but there is still an underlying need for experimental data for aspects such as the effects of cyclic loading. The key ingredients of load transfer curves may be summarised as follows:

loading) or net lateral pressure (force per unit projected area, for lateral load transfer). 3. Post-peak softening response, particularly for axial load transfer but also for lateral response in cemented soils. 4. Allowance for the effects of cyclic loading.

1. A pre-failure portion, with initial gradient linked to the shear modulus of the soil and a non-linear transition to a limiting load mobilised at an appropriate local displacement. 2. Limiting loads under monotonic loading, generally expressed as an equivalent shaft friction (axial


As in much offshore design, the last aspect above provides the greatest challenge. The most common approach has been implicit, reducing the limiting load transfer for cyclic conditions and assessing the pile response by means of an equivalent monotonic analysis, rather than attempt to carry out cycle by cycle analyses where degradation of the load transfer response is modelled explicitly. In many soil types, post-peak softening under monotonic loading is relatively minor (the API guidelines suggest a maximum reduction of 30% under axial loading) and the implicit approach for cyclic loading has proved sufficient. However, post-peak softening for piles grouted into cemented carbonate soils can be dramatic, and the effect on the overall pile capacity must be addressed in detail. This is also true in respect of lateral loading of piles driven or grouted into cemented sediments, where cracks in the surrounding sediments can extend to the free surface. In addition, pile tests at laboratory and field scale in carbonate sediments have shown that the

response under axial cyclic loading is complex, and with potential for accumulating damage to lead to progressive failure of the entire pile. A brief background is provided here with respect to the load transfer algorithms that have been developed over the last two or three decades for pile response, and in particular their application to carbonate sediments. 1.1 Axial load transfer Load transfer analysis of the axial response of piles stems from the 1960s (Coyle and Reese 1966). The initial stiffness of the pre-failure load transfer response may be linked directly to the shear modulus of the soil (Randolph and Wroth 1978), and the approach can be extended to allow for non-linear soil response, by integrating the stress-strain response radially outwards from the pile (Kraft et al. 1981). For a typical hyperbolic stress-strain response, the resulting load transfer curve turns out to match closely an inverted parabola, with an initial gradient that is about double the secant gradient at full mobilisation of the ultimate friction. This form of curve is also very similar to the generic shape recommended in the API guidelines (API RP2A 2000). In siliceous clays and sands, ratios of shear modulus to limiting shaft friction result in a load transfer curve where full mobilisation occurs at a local pile displacement of 0.25 to 2% of the pile diameter, with a typical value of around 1% (Jeanjean et al. 2010). In cemented material, it is appropriate to adopt a somewhat stiffer load transfer response, partly because of higher ratios of shear modulus to limiting shaft friction, but also because this is conservative from the point of view of increased stress concentration in the upper part of the pile, hence greater damage during cyclic loading. The load transfer software, RATZ, was developed originally in the mid 1980s (Randolph 1986), and at that stage was calibrated to match the response of laboratory rod shear and constant normal stiffness (CNS) tests, and subsequently field grouted pile tests (Randolph 1988, Randolph & Jewell 1989, Randolph et al. 1996). The program uses an explicit approach, similar to the commercial finite difference code, FLAC, to follow the non-linear response of the pile under either monotonic or cyclic loading. The current version of the software is based on Excel for convenience of data input and viewing results, although the numerical computations are carried out in a Fortran-based DLL (Randolph 2003). Details of the load transfer algorithm are described later, in relation to the CYCLOPS software, which was later developed from RATZ. The load transfer algorithm includes a non-linear pre-failure response, post-failure strain-softening, cyclic degradation calculated through a damage algorithm based on accumulated irreversible relative pile-soil displacement, and a low cyclic residual friction within the current range of plastic pile-soil displacement followed by (partial) recovery for displacements outside that zone (see Figure 1). As discussed later, the low cyclic residual

Figure 1. Schematic of load transfer algorithm in RATZ.

(CFG in Figure 1) is a key feature observed in grouted pile and CNS test responses. An important distinguishing feature of the load transfer analysis in RATZ is the provision for cycle-bycycle analysis of a complete storm loading sequence acting on the pile. This allows simulation of gradual ratcheting, and transfer of load progressively down the pile as the more heavily loaded soil in the upper part of the compressing pile accumulates displacement and softens. 1.2


Lateral load transfer

Forms of load transfer curves for lateral pile response are recommended for siliceous clays and sands in the API and ISO guidelines (API 2000, ISO 2007). These are mostly based on approaches developed in the 1970s by Reese, Matlock and others. At the time of the strut strengthening developed for first generation platforms in Australia’s Bass Strait (Wiltsie et al. 1988), Exxon undertook an extensive suite of centrifuge model and small field tests in order to develop lateral load transfer curves specifically for carbonate sands (Wesselink et al. 1988). The above work led to a generic load transfer curve for monotonic loading, where the lateral load per unit length, P (or net pressure, p = P/D), was expressed as a power law function of the cone resistance, qc , and the normalised displacement, y/d. A normalised version of the load transfer formulation was later proposed, following a more extensive series of centrifuge model tests (Dyson and Randolph 2001), expressed as:

where γ is the effective unit weight of the soil. Best fit values of the parameters, R, n and m, were established as 2.7, 0.72 and 0.58 respectively. A similar expression was proposed by Novello (1999), given by

The inclusion of a depth term, z/D, in this relationship means that, in general, the ratio, p/qc , increases more rapidly with depth than using Eq. (1). For example, if the cone resistance were to vary proportionally with depth, Eq. (2) would result in the lateral load for a given displacement, y, also being proportional to depth, rather than varying with depth to the power of 0.72 using Eq. (1). In other respects though, the two formulations are quite similar. Carbonate sediments tend to have high friction angles, in the region of 40◦ . However, comparing the above relationships with the API load transfer formulations for siliceous sand, the resulting response tends to be bracketed by the API curves for 20◦ and 35◦ , at least for displacements less than 0.1D. At larger displacements, the carbonate soil formulations give lateral resistance in excess of that for the 35◦ siliceous sand (Wesselink et al. 1988). The forms of load transfer curve given by Equations (1) and (2) are not bounded as y increases, and also have an infinite initial gradient (although a finite value of around 4G0 may be adopted for numerical implementation, where G0 is the small strain shear modulus). For practical values of deflection, with y wpeak ) followed by ‘positive’ monotonic loading. During the first cycle the load path follows the monotonic curve through τpeak down to τfail-1 . The sample is then displaced −2wtest and then back to the origin. An important aspect to note at this point is development of the cyclic ‘gap zone’ which, following completion of the first cycle, exists between −wtest and +wtest and represents the maximum extent of the plastic pile-soil relative displacement. The second cycle commences and the sample is loaded back to +wtest . Relative to the previous

cycle when the sample was at this displacement, the total plastic displacement that the sample has been subjected to is approximately 4wtest , which is equivalent to a positive displacement along the ‘backbone’ monotonic curve of (approximately) 4wtest . Hence, the strength on the ‘backbone’ monotonic curve has reduced from τfail-1 to τfail-2 . At the edge of the gap zone the shear stress mobilised is therefore now equal to τwmax-2 in accordance with Equation (4). After another full cycle down to a total negative displacement of −wtest followed by reloading back to +wtest the maximum bias stress reduces to τwmax-3 , and similarly after another full cycle the maximum bias stress at the end of the gap zone is τwmax-4 . As the sample is then monotonically loaded past +wtest


Figure 6. CYCLOPS – Residual shaft friction. Figure 8. CNS test comparison – RATZ and CYCLOPS.

2.2.2 Enhancements specific to CYCLOPS The CYCLOPS t-z algorithm incorporates a number of enhancements compared to the RATZ t-z model, namely: – Inclusion of a variable bias parameter (defined by bi and bf ). – Inclusion of an initial strain softening parameter (i-η). – Modification of the shape of the t-z curve within and immediately outside of the ‘gap zone’ formed during cyclic loading. The following sections discuss in detail the reasons for and implementation of the enhancements outlined above. Figure 7. CYCLOPS – Matching of various parameter sets.

and beyond the gap zone, the shear stress path gradually transitions from the maximum bias stress curve (i.e. τwmax-4 ) to the ‘backbone’monotonic failure curve (i.e. τfail-4 ) over a displacement equal to the cyclic transition width (wcyc ). The displacement controlled cyclic CNS test example described above illustrates an important aspect of the CYCLOPS t-z model, in that the amount of post-τpeak plastic displacement that occurs within the gap zone during cycling corresponds to an equivalent amount of plastic displacement along the monotonic τfail curve, and therefore essentially defines the magnitude of τfail at any point during cycling. It should be noted, however, that in the example test description above, it was assumed that all of the post-τpeak displacement was plastic. CYCLOPS actually assumes that a small component of the displacement is elastic (as determined by the maximum shear modulus Gmax ), and this elastic component is not included in the ‘total’ accumulation of plastic displacement that is used to calculate τfail .

2.2.3 Variable bias parameter As briefly described in Section 2.2.1, the bias parameter (b) defines the maximum bias shear stress (τwmax ) at the edge of the gap zone during cycling, primarily as a function of the monotonic failure stress (τfail ). The importance of the bias parameter can be demonstrated by examining a typical displacement controlled cyclic CNS test response. Such a response is shown plotted as shear stress (τ) versus horizontal displacement on Figure 2 and as the maximum shear stress attained during each cycle (τmax ) versus cycle number on Figure 8. The test comprised 25 cycles of displacement controlled loading at ±5 mm followed by monotonic loading to +12 mm, then to −5 mm and then back to the origin. The two points of interest to this discussion are:


– The rate at which τmax reduces with increasing cycle number. – The recovery of τ during the post-cyclic monotonic loading phase (i.e. Cycle 26 on Figure 8) as the sample is displaced beyond the ±5 mm gap zone. Figure 8 also shows the calibrated CNS test response as determined using RATZ. In RATZ, b has an implied value of 1.0, such that the maximum bias

Figure 10. CNS test and CYCLOPS comparison – Example 1. Figure 9. Variable bias parameter definition.

shear stress curve illustrated on Figure 3 is identical to the corresponding ‘backbone’ monotonic failure curve.As indicated on Figure 8, there is no increase in τ as the sample is displaced beyond the ±5 mm gap zone at cycle 26. In attempting to fit RATZ to this test, one can decide to set τres equal to the fully recovered postcyclic test value as adopted for the response shown on Figure 8, but the drawback of this is that the τmax values obtained in each cycle are significantly (and non-conservatively) greater than the actual test values. Alternatively, one could decide to set τres equal to the value of τmax at the end of cycling. This would result in a better fit to the τmax versus cycle number response, but would fail to capture the post-cyclic recovery in τ to the final τres . Hence, for these reasons it was considered essential to introduce a user-specified bias b into the CYCLOPS model. However, further detailed evaluation of various CNS tests results demonstrated that b not only varied from sample to sample but also often decreased in value over the course of each test. Therefore, CYCLOPS incorporates a non-constant b, which is specified in terms of initial and final values (bi and bf respectively). The manner in which CYCLOPS calculates the value of b at any given stage of cycling can be expressed as:

The parameter Ngap is the number of load cycles where the axial displacement is sufficient to reach either side of the edge of the gap zone, while m is a simple exponent. Figure 9 shows the relationship between b (normalised with respect to bi and bf ) and Ngap for various values of m. The benefits of adopting a non-uniform b are illustrated by two CYCLOPS calibrated responses to the previously considered CNS test. Results from the first CYCLOPS fit (Example 1; uniform b) are presented on Figure 10 (τ versus horizontal displacement) and

Figure 11. CNS test and CYCLOPS comparison – Example 2.

Figure 8 (τmax versus cycle number). Identical parameters to those adopted for the RATZ fit were used, with the exception of i-η (which is not included in RATZ), wcyc (implied value of 0 mm in RATZ) and bi = bf which were both set to 0.21. The resulting CYCLOPS response shows significantly improved agreement with the observed test behaviour compared to RATZ, particularly with regards to τmax during the latter stages of cycling and the post-cyclic recovery to τres . However, it can be seen that during early stages of cycling CYCLOPS still underpredicts τmax , which is somewhat overconservative, and implies that a value of b greater than 0.21 should be adopted for the first few cycles. The second CYCLOPS fit presented on Figure 11 and Figure 8 (Example 2; variable b) adopted identical t-z parameters to Example 1 but with bi and bf set to 0.68 and 0.21 respectively. The resulting predicted response shows generally excellent agreement with the test response, highlighting the benefit of adopting a non-uniform b. 2.2.4 Initial strain softening parameter The ‘standard’ strain softening parameter (η) is an exponent that controls the rate at which the postτpeak monotonic failure stress (τfail ) decreases with


Figure 12. Initial post-peak strain softening parameter.

Figure 13. Shear stress within gap zone.

increasing displacement, as demonstrated on Figure 5. Calcarenites typical of those encountered on the North West Shelf generally exhibit a fairly brittle post-τpeak response, with η values usually less than 0.5. For these relatively low η values, the negative gradient of the τfail versus displacement curve immediately following τpeak can be very large (in absolute terms), and it has been found previously that this large gradient can result in a slight overprediction of the degradation in shaft friction resulting from cyclic loading. In order to avoid this problem an additional parameter, termed the initial strain softening parameter (i-η), was incorporated into the CYCLOPS monotonic τfail algorithm which enables the immediate post-τpeak ductility of the τfail response curve to be increased slightly. The influence of i-η on τfail is illustrated on Figure 12, which plots τ/τpeak versus the post-τpeak displacement normalised by wres , assuming η = 0.4 and for various values of i-η. It should be noted that for a relatively large i-η value the CYCLOPS response matches the ‘baseline’ RATZ response (i.e. as shown by the iη = 1 × 106 curve on Figure 12). As i-η decreases, the post-τpeak displacement at which the CYCLOPS response curve rejoins the baseline (i-η = 1 × 106 ) response curve increases. This is demonstrated on Figure 12, where for i-η values of 1000 and 100 the CYCLOPS response rejoins the baseline response at approximately 0.8% and 7% of the displacement to residual (wres ) respectively. 2.2.5 Shear stress within gap zone As noted earlier the RATZ model gives a poor fit to the stress-displacement response within the cyclic ‘gap’ zone. A new formulation was therefore implemented in CYCLOPS to address this limitation. This modification is illustrated on Figure 13 which shows curves representing τ/τfail versus w/wcyc , where w is the displacement increment relative to the edge of the gap zone. Note that a negative normalised displacement represents behaviour inside the gap zone (i.e. a w/wcyc of zero defines the gap zone edge).

Example curves are presented for b values ranging between 0.05 and 0.95. 2.2.6 Shear stress outside gap zone As discussed earlier RATZ assumes a hardwired value of b equal to 1.0, implying that the maximum bias shear stress (τw max ) developed at the edge of the gap zone is equal to the current monotonic failure stress (τfail ). However, the available CNS test and GST data do not support this assumption since it is generally found that the value of τ at the gap zone edge is less than τfail . A certain displacement must be applied beyond the edge of the gap zone for τ to recover from the value at the gap zone edge to τfail . To address this limitation the CYCLOPS model allows the user to specify the displacement over which τ recovers from τw max (as defined by Equation 4) at the edge of the gap zone to τfail . This transfer displacement is termed the cyclic transition width (wcyc ). Further examination of the available CNS test and GST data indicate that the magnitude of wcyc is correlated with the size of the gap zone; an increase in size of the gap zone is matched by an increase in wcyc but at a decreasing rate, such that wcyc eventually appears to approach a limiting maximum value. Hence the CYCLOPS model incorporates a user specified maximum cyclic transition width (wcyc-max ). The value of wcyc at any stage is then determined according to the size of the gap zone (wgap ) and the exponential relationship presented on Figure 14, which plots wgap against wcyc , all normalised by wcyc-max . The implication of this approach is that wcyc /wgap decreases from close to unity (as implied by the dashed line on Figure 14) for wgap /wcyc-max near zero to approximately 0.2 for wgap /wcyc-max = 5. The shape of the shear stress-displacement curve assumed in the CYCLOPS model immediately outside the gap zone (i.e. w/wcyc = 0 to 1.0) is illustrated on Figure 15 for various values of b ranging between 0.05 and 0.95.


Figure 16. OC GST 400S2 – CYCLOPS comparison.

Figure 14. Cyclic transition width.

Figure 17. OC GST 400L – CYCLOPS comparison.

Figure 15. Shear stress within and outside gap zone.

2.3 CYCLOPS model applied to GSTs The discussion presented to date has focused on the observed response in CNS tests. However, the same characteristic features are also clearly evident from GST data. Typical CYCLOPS fits to some of the OC GST results (Randolph et al, 1996) are presented on Figure 16 and Figure 17. Similar response characteristics have also been observed elsewhere, with the recent full scale pile test results from Haberfield et al. (2010) being a notable example. Despite the ability of CYCLOPS to provide a very good match to the OC GSTs, problems arose when attempting to find a common set of parameters that could realistically model both the OC CNS tests and the monotonic and cyclic GSTs. These problems were of two basic forms, namely: – Use of the optimal set of parameters derived from the monotonic GST responses resulted in premature


failure when the grouted sections were subject to cyclic loading. – The optimal monotonic and cyclic parameters derived from the CNS tests were unable to predict the actual capacity of the grouted sections, giving conservatively low results. This originates from the fact that the observed CNS test τpeak values were low compared to the GST results. Although it is possible that this represents a systematic difference between the τpeak that can be achieved in a CNS test and that which can be achieved for a GST, we believe that this is unlikely to be the case. It is suspected that there may have been some problems in the sampling or testing of the OC cores which have contributed to the observed difference, but this is subject to confirmation. In summary, despite the much more realistic modelling that is possible with CYCLOPS compared to RATZ, the available correlation with GST data still gave a conservatively low prediction of measured GST response.This is of course safe for design purposes and hence is a robust position to occupy into the future. 2.4

Scale effects

In order to try to assess whether the underprediction of the GST response using CYCLOPS might be

due to some kind of scale effect between small scale CNS tests and full scale piles, or due to some other cause, a programme of numerical analyses was conducted using FLAC (Itasca, 2005). In these analyses an axisymmetric continuum model for the pile and soil was adopted, rather than the t-z approach used in CYCLOPS. These analyses focused on comparing responses under monotonic loading for CNS tests and GSTs and specifically addressed the potential role of borehole roughness on the GST response. A particular focus of this study was whether or not a full scale pile might exhibit a thicker interface failure plane (ie. shear band) than obtained in a CNS test, which could in turn lead to a ‘stretching’ of the required displacement to mobilise the residual friction. 2.4.1 Constitutive model The standard ‘Strain Softening Plasticity Model’ provided within FLAC was used for these analyses. Prior to the onset of plastic yielding the soil response was modelled as linear elastic. The soil cementation was modelled through a cohesive strength, which degraded progressively to zero with increasing plastic strain. Following breakdown of the cementation, the soil strength reverted to a frictional material with the strength defined by a friction angle; the friction angle was also variable starting from zero prior to any plastic yielding and then increasing to a maximum value as the true cohesion degraded to zero; this ‘peak friction’ angle reflected the fact that whilst the true cementation might be gone, the soil sample was still likely to be compact and interlocked at this stage. Further plastic shearing led to a gradual reduction of the friction angle until a ‘critical state’ was achieved at larger strain, and thereafter a constant value was assumed. Another key feature of these materials is contraction when subject to plastic shearing. This was incorporated into the soil model through the adoption of a negative dilation angle; this varied from a high initial value, corresponding to the maximum rate of contraction, and thereafter gradually increased (ie. became less negative) at increasing plastic shear strains. At the ‘critical state’ a zero dilation angle was imposed. A ‘cap’ yield surface was not included since the axial pile friction in these soil types is governed by normal stresses well below the cap yield surface and hence it is the shear induced contraction, combined with the effects of cementation degradation, that should be captured for a realistic model. Whilst cap models can simulate the shear induced contraction well, most are inadequate for dealing with the brittle cementation component. 2.4.2 CNS test simulation The shear stress-displacement response predicted in the FLAC CNS test simulation, assuming a normal stiffness applicable to a 2 m diameter pile, is presented on Figure 18. The initial brittle post-peak response reflects the rapid breakdown in cementation while the gradual reduction thereafter is due to the shearing induced compaction.


Figure 18. CNS test simulation – 2 m diameter pile.

2.4.3 Mesh details for GST simulation Following a program of initial analyses, the key components identified for capturing realistic GST behaviour were identified as: – Meshes needed to be vertically aligned to ensure that the dominant and known shear band orientation (ie. vertical) could be captured accurately and within a single element thickness. Borehole roughness was introduced by assigning stiff and strong ‘pile’ properties to some of the elements that were adjacent to the pile but which would ordinarily be assigned soil properties (for a ‘smooth’ pile). At certain locations on the grout-soil interface, where the shear band is forced to propagate from one radial distance to another (invariably where a distinct change in grouted section diameter occurs), a bearing type failure is initiated at the ‘lip’ of the grouted section outstand, and this would, by definition, be smeared over several soil elements in front of the ‘lip’ and would involve imposition of high normal stresses. Strain softening and localisation into a shear band is not a critical feature of such a failure mode and hence the vertical alignment of the mesh is not believed to have a significant adverse influence on these bearing type failure modes. – Very thin soil elements were used immediately adjacent to the grouted section interface since this was essential in order to capture an appropriate degree of borehole roughness. This also offered the advantage that several elements would always be involved in any local bearing type failures that developed around individual hard-points. The minimum element thickness adopted in our analysis was around 4.5 mm at the nominal grout-soil interface, increasing steadily to the radial boundary of the mesh. For the five elements in closest proximity to the groutsoil interface the average element thickness was around 6 mm to 6.5 mm, which was very comparable to the element thickness used in the CNS test simulation (5.8 mm). This similarity of thickness is vital in order to avoid spurious mesh scale effects,

Figure 20. Load displacement response – ‘rough’ and ‘smooth’ 2 m GST.

imposed on the top of the mesh to represent a burial depth of around 45 m. – The grouted section itself was modelled as an elastic solid with a resultant EA that proved to be around double that of the actual OC test sections. A Poisson’s ratio of 0.3 was also assumed.

Figure 19. Grout-soil interface details – 2 m diameter GST analyses.

since the minimum shear band thickness is limited by the element thickness which ultimately controls the displacement rate from peak to residual friction. Analyses were undertaken with a variety of different roughness assumptions – details of two of the considered cases for the 2 m diameter GST are presented on Figure 19. The meshes used for these analyses had the following key features: – The modelled GST length to diameter ratio varied from about 1.5 to 6, which covers the range for most of the actual OC GSTs. No end bearing was included below the grouted section. – The outer radial boundary of the mesh, which was fixed against radial movement, was defined at a radius of 60 times the grouted section radius. – Soil was included above and below the grouted section in order to avoid local effects, such as unrealistic stress concentrations at the base of the grouted section and under-confined mesh distortion at the top of the grouted section. – The in situ effective horizontal stress at the mid depth of the mesh was the same as assumed in the CNS test simulation. A vertical surcharge was

2.4.4 GST analysis results The shear stress-displacement responses predicted in the 2 m diameter GST simulations are presented on Figure 20 for a variety of different roughness conditions. It may be noted that for all cases the peak frictions obtained from the rough GSTs are significantly less than the simulated CNS test peak. Some of the variability in peak friction from the GSTs is due to the slightly different diameters for the differing roughness, but the enhancement for the ‘rougher’ piles was generally significantly greater than the change in diameter would imply. The reason for the lower average peak friction in the GSTs compared to that obtained in the CNS test simulation was found to be due to progressive failure; at peak mobilisation of the GST a sufficiently high strain had developed over a short section at the top and a large section at the bottom of the grouted section such that the initial cohesion had degraded to zero. A purely frictional response therefore applied over these zones, which gave a significantly smaller interface friction than implied by the initial soil cohesion. The residual frictions exhibited a different trend with higher values obtained in all of the GST cases compared to the CNS test simulation. An evaluation of the various results revealed that this was due to several reasons:


– Even at low roughness, the horizontal effective stress averaged along the GST was found to be higher than the normal stress acting at the residual condition in the CNS test simulation.

When considering results from this kind of analysis due caution should be exercised since such analyses are notoriously difficult to perform reliably. Our examination of the GST simulation results did not reveal any obviously faulty features, except (in a few cases) a slightly enhanced residual friction at large head displacements resulting from some over constraint in the system. Overall, we believe that these analyses do provide some evidence for an enhanced residual friction on a full scale grouted section compared to a CNS test, although we recommend that this conclusion should be treated as provisional in the absence of additional confirmatory data. In addition it may only be applicable in certain cases such as for soils subject to shear induced contractancy, as modelled here. Other questions may be asked as to the modelling assumptions made for the shear band. An alternative strategy, adopted in some cases (usually clay soils), is to include some strain rate dependency of the soil strength, which in turn controls the minimum shear band thickness that develops. However, for the cemented soil under consideration here, it is not readily apparent why any significant viscous strain rate effects might arise. Within a continuum framework, it is not apparent that any other assumptions could be sensibly made, in the absence of empirical data. However, it is possible that a more fundamental assessment using a discrete element method (DEM) approach might potentially reveal some different types of failure mechanism (at the micro-mechanical scale) in these kinds of soil, which could suggest a scalability of shear band thickness that is not resolvable in the current work. In any event, in order to further advance our understanding of this complex matter either additional testing at the OC test locations or more full scale GST or pile load tests combined with comprehensive programs of CNS tests from suitable sites elsewhere, will be required.

– Calculation of the ‘equivalent’ CNS spring acting along the GSTs revealed that while this was generally in accordance with the CNS test simulation, strong end effects were apparent which gave rise to the enhanced average horizontal effective stress. – For the rougher interface cases, highly concentrated pressure bulbs were mobilised just above many of the significant outstands on the grout-soil interface. Further detailed inspection revealed that each pressure bulb was associated with a transition of the sliding failure plane from a larger to a smaller diameter (i.e. to become closer to the nominal diameter of the pile). These pressure bulbs were therefore associated with local bearing failures around the outstand rather than pure frictional sliding failures. At these pressure bulbs, highly concentrated horizontal effective stresses were mobilised, which offset the contraction induced reduction in horizontal effective stress that occurred in the CNS test simulation, in turn enhancing the frictional resistance. As can be seen by comparing Figure 18 and Figure 20, the initial stiffness of the simulated CNS test response was found to be very much greater than obtained from the GST simulations. This is readily understood when considering the different configurations of the CNS and GST test responses; the former only required shearing of a single 5.8 mm thick element while the applied shear stresses at the grout-soil interface of a GST propagate throughout the mesh and hence elastic deformations accumulate over a wide area. From the same figures, it can also be seen that none of the GST analyses revealed any difference in the displacement required to transition from the peak to residual friction between a CNS and full scale GST. Hence despite the imposed interface roughness in the GST simulations, the shear band thickness in these analyses was controlled by the minimum element thickness in all cases; the increasing roughness cases did nothing to enhance the shear band thickness, and hence there was no mechanism to enhance the displacement to residual. 2.4.5 Conclusion from FLAC analyses Unfortunately the results of the FLAC analyses did not reveal a definitive explanation as to why the response in a CNS test might differ significantly from that of a full scale GST. In most cases the response of the simulated CNS tests were found to be a close proxy of the numerical GST analysis. However, the analyses did hint at the potential for some differences, both due to end effects (although these might not be important for an actual pile) and due to the influence of a ‘roughened’ interface. Specifically, increasing roughness appeared to enhance the residual friction compared to an equivalent CNS test. None of the analysis results suggested that the displacement required to transition from the peak to residual friction should be any different between a CNS test and a full scale GST.

2.5

The usual process by which the design storm history is incorporated into the cyclic axial capacity analyses for drilled and grouted piles offshore Australia is as follows: 1. A structural model of the platform incorporating all dead, live and environmental loads is used to determine the total pile head axial load occurring at the peak of the maximum design wave (i.e. maximum overturning moment) and at the trough of the same wave (i.e. minimum overturning moment) for the given design condition. 2. The structural model of the jacket is then subjected to waves of various height (H) and period (T) corresponding to the design lifecycle and storm wave H-T histograms. Processing of the model output results in a pile head cyclic axial load ‘transfer function’ which relates H and T to the pile head axial load at each wave peak and trough, thereby defining the loads applicable to each cycle. 3. The pile head cyclic axial load transfer functions, together with the number of cycles of each wave


Influence of storm order

type as specified in the lifecycle and storm histograms, are input into the RATZ or CYCLOPS software in some order. Historically, the lifecycle and storm wave H-T histograms were generally only defined as occurrence matrices rather than specific time histories. Therefore, the lifecycle and storm occurrence matrices were arranged to produce various ‘ideal’ time histories, including ‘increasing’ (i.e. smallest wave to largest wave), ‘decreasing’(i.e. largest wave to smallest wave) and ‘increasing-decreasing’ (i.e. smallest waves at the beginning and end, with the largest wave occurring at the storm midpoint). Generally it was found that the ‘decreasing’ arrangement would result in the highest reduction in axial pile capacity as a result of cyclic storm loading. 2.5.1 ‘Actual’ storm load time history Recently ‘actual’ wave-by-wave storm time histories have been developed for pile capacity analyses. An example of an actual storm profile, in terms of pile head axial load at the wave peak and wave trough, is presented on Figure 21. Surprisingly, it was found that the cyclic degradation inflicted on a pile subjected to an actual storm was greater than that for a pile subjected to any of the idealised storm arrangements that had been used in the past. The reasons for this were not immediately apparent but on further investigation it was found that this was related to: – The arrangement of the larger waves within the profile. – The degree to which each of the larger waves can cause zones along the pile to go post-τpeak . – The degree to which packets of smaller waves, which occur after each of the larger waves, are able to cause further damage to the post-peak zones. Fundamentally, it is a basic characteristic of both the RATZ and CYCLOPS models that the degradation process is governed by the amount of plastic degradation that occurs for soil elements that are loaded at some point during the storm such that their monotonic (peak) strength is exceeded. Cycling at loads that never reach τpeak but which exceed the elastic threshold shaft friction ξτpeak can also cause degradation, but at a much slower rate than for any element that goes post-τpeak at some stage. As an example of the shaft friction degradation process, first consider a pile subjected to an idealised decreasing storm loading profile, where the largest load is applied first and the smallest last. The process is demonstrated schematically in a step-by-step fashion on Figure 22 and described as follows: – At the start of the storm the largest wave causes the soil along the upper portion of the pile to reach its peak shaft friction and to commence along the strain-softening part of the t-z curve (i.e. the post-τpeak zone on Figure 22a). This is due to the flexibility of the pile which results in much larger pile-soil interface movements at the top of the pile


Figure 21. Pile head axial load profile – actual storm.

–

–

–

–

compared to the bottom. Hence, from this point on, these soil elements operate in the post-τpeak domain, as illustrated by the resulting failure shaft friction (τfail ) profile on Figure 22b. Subsequent large waves of decreasing height will cause some additional soil elements to transfer from pre-τpeak to post-τpeak behaviour, which enables the post-τpeak zone to propagate further down the pile (Figure 22c). In addition to this, further damage is inflicted on the initial post-τpeak zone, leading to a steady reduction in the available shaft friction (i.e. τfail ) over this region (Figure 22d). It is important to note at this point that since all the large cycles are lumped together at the beginning of the storm, the total amount of post-τpeak degradation that occurs in the first few cycles is limited. Hence, as the storm progresses and the modelled wave height reduces, the ability of subsequent waves to extend the post-τpeak zone rapidly decreases. Following the first few large cycles, the subsequent very large number of medium to small cycles are only able to continue to degrade the elements that were forced post-τpeak in the first few cycles (Figure 22e). Once full degradation of the post-τpeak zone to the residual shaft friction (τres ) has occurred, which is generally after only a relatively small proportion of the total number of smaller waves in a storm, the large number of remaining small waves are unable to cause any further damage (Figure 22f). Hence, at the end of the storm, the post-τpeak zone has generally transformed almost completely into a zone where τfail = τres . However, the initial postτpeak zone caused during the first large cycle is only extended to a limited degree by the subsequent cycles.

Now consider the same pile subjected to the same storm loads but with the load profile ordered such that the larger waves are spread out over the storm’s duration and separated by packets of smaller waves. An example of such an ‘arranged’ profile is presented on

Figure 23. Pile head axial load profile – arranged storm.

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–

Figure 22. Pile axial capacity degradation – ‘decreasing’ storm.

–

Figure 23, which has been developed from the ‘actual’ storm history presented on Figure 21. The arranged profile comprises five ‘load packets’ with each packet comprising one large wave followed by a series of smaller waves ordered according to decreasing wave peak load. The process is demonstrated schematically in a stepby-step fashion on Figure 24 and described as follows:

–

– After application of the first large wave (Figure 24a), a post-τpeak zone will form along the upper portion of the pile (Figure 24b). This is the same as for the first cycle in the decreasing storm. – A subsequent packet of smaller waves will then work on the post-τpeak zone (Figure 24c). This


post-τpeak zone is not extended, but most or all of it will be transformed into a τfail = τres zone (Figure 24d). Note that once a soil element goes post-τpeak , even very small subsequent loads may propagate the degradation, because any stress that exceeds the (very small) cyclic residual friction (τcyc-res ) leads to plastic strain and hence an accumulation of plastic damage. The small waves are used much more effectively (in terms of causing damage) than in the descending storm analysis, since one large wave followed by a moderate packet of small waves may lead to a τres zone only a little smaller than caused for the entire descending storm. When the large wave from the second load packet is applied (Figure 24e), the post-τpeak zone caused by the first large cycle is unable to offer much resistance to this load, since it has been fully degraded to the τres value. Hence, this load is distributed much further down the pile than would have occurred if this were only the second wave in a descending storm history. This therefore maximises the size of the new post-τpeak zone caused by this second large cycle (Figure 24f). A subsequent package of small waves then works on this new post-τpeak zone (Figure 24g) until it has degraded to τres (Figure 24h). With subsequent ‘packets’ of a relatively large damage triggering wave followed by a set of smaller waves the process continually repeats itself, with each larger wave propagating the post-τpeak zone further down the pile and each packet of smaller waves degrading the strength in this zone towards τres .

In summary, it can be seen that by separating the larger damaging waves by sets of smaller waves, cyclic damage is maximised. In the descending storm only a relatively few of the small to medium cycles make any contribution, whereas in an ‘optimally’designed storm (for maximising damage) all the small to medium cycles will contribute to the damage process.

this model for practical design purposes a significant number of CNS tests need to be performed. For each CNS test a best fit calibration is developed, which requires considerable skill and care, since as mentioned in Section 2.2.1, superficially similar monotonic responses over a limited strain range can result in a very different cyclic response. A variety of stress and displacement controlled CNS tests are required at each tested location in order to properly anchor the calibrated CYCLOPS response. Any excessive brittleness observed in the monotonic CNS test stress strain response also needs to be treated with extreme caution since this may never be apparent in a field situation. A well calibrated view on sensible ranges for the peak skin friction is required to ensure that an unconservative parameter selection is not made. Once the CNS test calibrations are complete the values of the various parameters adopted in each CYCLOPS fit are then plotted as a function of depth in order to ascertain appropriate trends. Design lines can be fitted through the calibrated data as desired. Our experience with this procedure so far has been good, with generally sensible ranges of parameter variability found from these calibration exercises. As more experience is obtained using the CYCLOPS model over time it is hoped that some of the required parameters may generally be found to lie within tight bounds, which may reduce the number of detailed calibrations that are required at each site. Notwithstanding, it is likely that design of drilled and grouted piles under complex cyclic loading conditions will remain a challenging task for many years to come. 2.7

Figure 24. Pile axial capacity degradation – ‘arranged’ storm.

2.6 Assessment of parameters for design The CYCLOPS model outlined in this paper can accurately model measured CNS test and GST responses, but this ability comes at a cost in that many different parameters need to be defined. In order to use

Results of an example analysis are presented here to demonstrate the application of CYCLOPS to a typical full size pile, and are compared to the equivalent results obtained using the standard RATZ algorithm. A steel pile has been adopted for this example with a 2.2 m outer diameter, a constant 90 mm wall thickness and it is assumed to be grouted in a 2.5 m diameter hole. The soil parameters used in the analysis are presented in Table 1. The assumed storm load history has been applied in various ways to illustrate how important this can be: the full actual storm (Figure 21), a ‘5 packet’ storm (Figure 23), a standard decreasing storm and an increasing storm. The pile head load-displacement response predicted using CYCLOPS for the ‘actual’ storm is presented on Figure 25, along with the undegraded monotonic response. It should be noted that the somewhat spikey appearance of the cyclic response, particularly for isolated large cycles, is due to the fact that only the peaks (minimum and maximum) of each cycle are saved in the analysis for later post processing. The t-z curves extracted from the actual storm analysis at a couple of nominal depths (9.8 m and 39.8 m) are presented on Figure 26. The extent of cyclic degradation caused by the storm is readily apparent from


Example analysis

Table 1. analysis.

CYCLOPS & RATZ parameters for example

Parameter

CYCLOPS

RATZ

τpeak τres /τpeak Gmax ξ wres η i-η τcyc /τpeak bi bf m wcyc-max

440 kPa 0.3 1232 MPa 0.4 0.2 m 0.4 300 0.01 0.5 0.2 5 0.1 m

440 kPa 0.3 1232 MPa 0.4 0.2 m 0.4 N/A 0.01 N/A N/A N/A N/A

Figure 25. CYCLOPS pile head response – actual storm.

these figures, leading to an 18.4% reduction in the post-storm monotonic capacity. From the t-z curves it is apparent that the response is affected by cyclic degradation, even at 39.8 m below the mudline. This is further illustrated on Figure 27, which presents the peak mobilisable friction at the end of the storm for each of the different representations of the storm. It can be seen that the actual storm causes the greatest degradation, and is slightly worse than the ‘5 packet’ storm. The increasing storm causes little damage (0.8%), but it can still be observed from Figure 27 that the peak mobilisable friction is less than the prescribed peak friction over the upper 20 m. However, this reduction is principally due to monotonic progressive failure not cyclic loading. The standard decreasing storm that has generally been used for design purposes in the past is shown to be significantly less damaging (10.2%) than either the actual storm or the ‘5 packet’ storm. The key feature evident from Figure 27 is the propagation of damage to increasing depth with increasing disorder in the storm, as anticipated by the process described in Section 2.5. The pile head load-displacement response for the equivalent RATZ analysis of the actual storm is presented on Figure 28. The cyclic degradation (15.0%) is only a little less than obtained with CYCLOPS for this particular case. The RATZ generated t-z curves from the actual storm are shown on Figure 29. By comparison with Figure 26, the more stylised t-z response obtained from RATZ at a depth of 9.8 m is readily apparent. Also it can be seen that cyclic degradation did not propagate as far as 39.8 m in the RATZ model although this did occur with CYCLOPS. The peak mobilisable friction at the end of the storm is presented for the various RATZ analyses on Figure 30. Interestingly, the cyclic degradation obtained with RATZ for an increasing storm was slightly higher, albeit still very small (1.4%) compared to CYCLOPS. However, for all the other storm representations, CYCLOPS gave slightly higher cyclic degradation than RATZ. In summary it can be seen that for the example presented here CYCLOPS gave slightly lower

Figure 26. CYCLOPS t-z response – actual storm.

post-storm capacities than RATZ. However, we have found much larger differences in other practical cases, particularly where complex stratigraphy is encountered. The general rule to date is that CYCLOPS invariably yields lower post-cyclic capacities than obtained with RATZ, all other things being equal. The importance of modelling actual storm histories rather than idealised representations is also clearly demonstrated by both the RATZ and CYCLOPS analysis results. 3 PART 2 – LATERAL PILE DESIGN 3.1

A new model for determining the lateral response of piles in uncemented carbonate soils when subject to undrained cyclic loading is presented herein. The new


Introduction

Figure 29. RATZ t-z response – actual storm. Figure 27. CYCLOPS post-cyclic failure shaft friction profiles.

Figure 28. RATZ pile head response – actual storm.

method retains the basic p-y curve format used as standard for design of laterally loaded piles in the offshore industry but builds on the design approaches used for shallow foundations and suction piles. Design of such structures is now generally performed using a ‘cyclic strength approach’. For these kinds of structures the standard practice is to define a ‘cyclic strength’, which is some reduced proportion of the monotonic strength, and is a function of the number of applied cycles, the degree of cyclic loading (2-way, 1-way, etc.) and the magnitude of strain that is deemed acceptable for the soil to sustain.Appropriate factors of safety are applied to these cyclic strengths and the combined bearing and sliding capacity is then determined in order to define


Figure 30. RATZ post-cyclic failure shaft friction profiles.

the maximum loads that may be safely applied to the foundation, with only limited in-service movements. Unfortunately, modelling of laterally loaded pile behaviour is more complex than design of shallow foundations or suction piles since the ultimate geotechnical capacity is generally not the governing design criteria. For a typical long and flexible pile, an increasing applied pile load will simply mobilise the soil over a greater depth and hence it is the pile bending

moments and pile head displacements that are usually the critical design parameters. To assess these it is therefore necessary to determine the appropriate load displacement response at each depth level in the soil. In addition, whilst shallow foundations and suction piles are usually pure load controlled systems (i.e. applied loads do not generally redistribute in a global sense) the upper part of a laterally loaded pile when subject to cyclic loads will exhibit a partially displacement controlled response, leading to the potential to redistribute soil reaction pressures down the pile as cycling progresses. In order to capture these features we believe that it is first necessary to define the appropriate monotonic p-y response and then to perform a cycle-by cycle analysis to capture the progression of the response over time. 3.2

Model development

Development of the new p-y model comprised four basic stages: – Stage 1: Finite element type analyses were performed to establish the p-y response of a pilesoil slice element for a range of generic stress strain curves representative of uncemented carbonate soils.These were developed for various numbers of applied load cycles. Analyses were also performed to assess the modification required to allow for near-surface effects. – Stage 2: The results from Stage 1 were used to develop an appropriate p-y curve format that can be used for any specific soil strength assumption and required number of cycles. – Stage 3: Using the results from Stage 2 along with an examination of the characteristic response of raw simple shear test data, an algorithm was developed to allow a cycle by cycle analysis to be advanced. – Stage 4: The p-y curve format and the cycle-bycycle algorithm were implemented into a new Excel program, which has been named pCyCOS. To aid in comprehension of the new model, the basic algorithm for the new p-y model is first presented followed by the detailed analyses that were performed to calibrate the various required parameters. 3.3 Envelope p-y response The envelope p-y response is the first basic building block of the p-y model and defines the p-y response of a horizontal slice through the pile-soil system when subjected to load controlled cycling for various numbers of cycles. For any element of soil with a given normalised monotonic undrained shear strength (su /σvo , where su is the monotonic undrained shear strength and σvo is the initial vertical effective stress), an S-N curve may be defined, which is a representation of the normalised cyclic stress ratio (S = τcyc /σvo ) and the number of cycles (N) to mobilise a nominal shear strain level (γ).


Figure 31. Envelope p-y response – basic concepts.

By drawing vertical slices through the S-N curve it is possible to define stress strain curves for any given value of N. This is illustrated on Figure 31a and Figure 31b. By implementing these stress strain curves into a finite element type model of a lateral pile element we can then develop the appropriate p-y curve that applies for the specified number of cycles, N, as illustrated on Figure 31c. It may be noted that a normalised pressure

response is defined from this assessment (p/σvo where p is the soil pressure applied to the pile element). the process described For the specified su /σvo above is repeated for different values of N and the results compiled, as shown on Figure 31d. This latter plot essentially defines the response of a lateral soil element when subject to load controlled cycling. For example, considering a pile element that imposes a constant peak cyclic pressure p1 on the soil, the lateral displacement for that pile element would be equal to y1 after the first cycle, y2 after the second cycle, y3 after the third cycle and so on (Figure 31d). The entire process outlined above is repeated for various values of su /σvo , as required, in order to cover the range of monotonic strengths expected in the field. We have conservatively adopted pure 2-way cyclic loading conditions in the model to date, and hence 2-way S-N data has been used to define the necessary stress strain curves for developing the envelope p-y curves. The model may be readily changed for any other degree of cycling, but would require additional assessment of the various input parameters.

3.4 Cycle by cycle p-y response The envelope p-y curves that define the pile-soil behaviour under pure load controlled cyclic conditions have been defined above. If the problem under consideration were purely load controlled, then the final model would require no further sophistication than to add together all of the load controlled element responses for any given number of cycles, N, in order to give the total response for the entire system after N cycles. However, as noted earlier, the (partially) displacement controlled nature of the pile-soil system means that pile-soil elements near the top will shed load to elements further down, as the upper elements, which are subject to larger cyclic displacement amplitudes, ‘soften’ to a much greater degree than the lower elements. A cycle by cycle model is therefore required to capture this behaviour. 3.4.1 First cycle The lateral soil response used for the first cycle is simply the p-y envelope response for N = 1 as discussed in Section 3.3. By definition this describes the monotonic loading response of a pile-soil element. Monotonic (first cycle loading) is a purely load controlled process and therefore for any given pressure, p1 , applied during the first cycle, the lateral displacement for this element will be y1 (Figure 32a). 3.4.2 Second cycle The p-y response assumed for the second cycle is defined from the envelope p-y response obtained for N = 2, but is not equal to the N = 2 envelope curve. For any given pressure, p1 , applied during the first cycle the lateral displacement obtained during the second


Figure 32. p-y curve format for first and second cycles.

cycle, assuming that the same pressure p1 was still applied, would be equal to y2 (Figure 32b). However, in reloading to the pressure p1 during the second cycle, the pile-soil element is assumed to follow a different

pressure–displacement curve. This reloading curve is formed from 3 parts: Part 1 – Up to an applied pressure of 25% of p1 , the pile-soil element follows a linear path starting from the origin and giving a displacement of y0.25 at P0.25 (=25% of p1 ). (Evaluation of y0.25 is considered later.) Part 2 – For a pressure exceeding 25 % of p1 , the pile-soil element follows a second linear path that passes through the point (p0.25 , y0.25 ) and through the point (p1 , y2 ). Part 3 – Where the pressure defined from Part 2 of the p-y curve exceeds that which would be obtained using the N = 1 (monotonic) envelope curve, then the monotonic envelope curve is adopted. It can be seen that a key feature of this second cycle p-y response is an initial ‘soft’ zone followed by a hardening response. Such behaviour is encountered in all simple shear tests on uncemented carbonate soils where a reversal of the cyclic shear stress is imposed. The selection of an inflection point at 25% of p1 has been made based on inspection of the results from many simple shear tests. It may also be noted that the monotonic response may (eventually) be recovered, partly due to the dilatory soil response as large shear strains are applied following partial liquefaction. This is consistent with the philosophy that the generation of excess pore pressure during cyclic loading mainly affects the soil stiffness rather than its ultimate strength. However, it should be also appreciated that when a load cycle imposes a soil pressure that is close to the ultimate limit defined for cycle N, the degree of softening that occurs for cycle N + 1 may be such that there is a complete loss of stiffness under any further cyclic loads (Figure 32c). The pile will therefore be free to move through the softened soil, but would eventually encounter fresh soil, which would constrain the ultimate displacement somewhat. Large displacement hardening of this type has not been included in the model, but in practical applications it is unlikely to be an important limitation, due to the partially displacement controlled nature of lateral pile loading, since this also serves to limit the increase in displacement from one cycle to the next. 3.4.3 Subsequent cycles The form of the p-y curve used for all subsequent cycles (N = i) depends on whether or not the peak pressure applied to the pile-soil element in cycle N = i − 1 is the maximum pressure that has ever been applied to that element (pmax ), or is less than pmax . Where the peak pressure applied in cycle N = i − 1 (i.e. pi-1 ) is equal to pmax , the p-y curve for the next cycle is defined using the same procedure as used for the second cycle (Section 3.4.2), as illustrated on Figure 33a. This approach includes a degree of oversoftening in the p-y response, since inherently it is assumed that the pile has been subject to i − 1 cycles of pressure, pi-1 , even though it may have been subject to

Figure 33. p-y curve format for subsequent cycles.

a lower pressure during the earlier cycles, which would be expected to lead to a lower degree of softening for cycle N = i. Where the pressure applied in cycle N = i − 1 is less than pmax , two alternative criteria are used to assess the p-y response for cycle N = i, with the criterion that gives the greatest degree of softening adopted for the final p-y response. The first criterion is identical to that presented on Figure 33a: the displacement for cycle N = i is calculated assuming that the pile has been subject to i − 1 cycles of pressure, pi-1 , even though it may have been subject to a greater pressure during the earlier cycles. Inherently this means that the predicted degree of softening will underpredict that which would occur in practice, and indeed if pi-1 is significantly less than the maximum pressure ever applied, it can lead to a p-y curve for cycle N = i that is stiffer than that used for cycle N = i − 1, which is clearly a physically unreasonable outcome. Hence the second criterion is to first determine the p-y response that would be obtained if the pile-soil element had been subject to i − 1 cycles with an applied pressure equal to pmax , (i.e. as per Figure 33a), and then apply a scale back factor to the predicted degree of


softening obtained from this calculation to account for the fact that the pile-soil element is now being cycled at a pressure that is less than pmax . The scale back factor used in this calculation has been derived by consideration of the available S-N data and will be discussed later. This process is illustrated on Figure 33b. 3.4.4 Modifications for surface effects The p-y curves discussed above are defined based on a horizontal slice through the pile and soil. This therefore represents the ‘deep’ failure mode where the soil can flow around the pile but with no vertical movement of the soil. However, near the mudline, vertical movements can occur in the soil giving rise to wedge type failures. To account for this effect a scaling factor (‘p-multiplier’) is applied to the pressures calculated using the ‘deep’ failure p-y curves. The scaling factor is determined as the ratio between the ultimate lateral resistance calculated accounting for the actual embedment depth, to the ultimate lateral resistance that would be obtained for the ‘deep’ failure mode. The ultimate peak lateral resistance has been defined at any depth based on the plasticity model developed by Murff and Hamilton (1993). The validity of this assumption was confirmed through a program of 3D finite element (FE) analyses, using the models discussed in the following section. These analyses showed that the simple ‘p-multiplier’ approach proposed here generally gave results within 10% of those obtained from the more rigorous FE analyses, which is considered acceptable for engineering design purposes. 3.5 Analysis of envelope p-y response The envelope p-y curves required for the new p-y algorithm have been developed from FE analyses performed with ABAQUS (ABAQUS Inc., 2006) and finite difference (FD) analyses performed using FLAC (Itasca, 2005). We used two programs for this work, since they have different strengths and weaknesses. In addition, a few cases were analysed using both programs to enable internal verification of the results obtained. Irrespective of the specific program used, the meshes generated for developing the envelope p-y curves are based on the same basic assumptions: – A plane strain ‘slice’ through the pile and soil was modelled. – Infinite elastic elements were included to model the far field behaviour; these were found to be essential during the equivalent work performed to determine the p-y curves for the CHIPPER model (Erbrich, 2004). – The pile was modelled as having a ‘smooth’ (i.e. frictionless) interface with the soil. – It was assumed that no separation would occur between the pile and the soil, due to the presumed undrained nature of the applied loading. – A small strain formulation was generally adopted (although we also performed some confirmatory analyses using a large strain formulation).

3.5.1 Constitutive model All of the numerical analyses were performed using an undrained total stress approach. For all cases we selected the Mohr-Coulomb soil model, which degenerates to a simple Tresca model when a friction angle of zero is specified, as required for the assumed undrained conditions. To account for the very high degree of non-linearity in the stress-strain response, the cohesive strength was defined with a hardening response as a function of the plastic shear strain. The elastic part of the response was modelled using a simple linear elasticity model, with a Poisson’s ratio of approximately 0.5 to ensure constant volume conditions. 3.5.2 Stress strain response Stress-strain curves characteristic of uncemented carbonate soils were established based on the available test data. From these tests, a variety of stress strain curves were developed for use in the FE/ FD analyses which reflected the full range of different normalised monotonic strengths, su /σvo . For each case, we used the procedure presented in Section 3.3 in order to develop the shear stress versus shear strain curves required as input for the numerical analysis. These curves were defined for pure monotonic loading and for pure 2-way cyclic loading at various numbers of cycles. Typical stress-strain curves for a) soft/loose = 0.25 and b) dense soil with soil with a low su /σvo a high su /σvo = 10 are presented on Figure 34a and Figure 34b respectively. We performed verification analyses in both FLAC and ABAQUS under simple shear loading conditions in order to confirm that the intended stress-strain curves were indeed modelled accurately. Figure 35 presents one such comparison, which confirms that the model operated as expected. 3.5.3 Envelope p-y analysis – results Example analyses showing the resulting p-y curves obtained from the FE/ FD analyses using the input stress-strain curves shown on Figure 34 for a) soft/loose soil with a low su /σvo = 0.25 and b) dense soil with a high su /σvo = 10 are given on Figure 36a and Figure 36b respectively. These p-y curves are presented for a number of different applied load cycles. Similar curves were developed for a range of differ ent su /σvo values, encompassing the relevant range. From the range of results obtained, it was found to be possible to interpolate linearly between the values obtained in order to determine the appropriate p-y curves for any other desired combination of su /σvo and N.


3.6

Parameters for cyclic p-y response

In addition to the envelope p-y response, the two other essential components of the new p-y model are the definition of the inflection point and the stiffness degradation function to account for reducing loads from one cycle to the next.

Figure 34. Typical envelope stress-strain responses. Figure 36. Computed envelope p-y curves.

(i.e. ‘stiffening’) load displacement path. The basic form for this p-y curve was obtained from an evaluation of the stress strain curves observed in cyclic simple shear tests performed on uncemented carbonate soil samples, where N > 1. Typical responses from these tests indicate that the convex nature of the stressstrain response in the first cycle contrasts with the concave response for all subsequent cycles (Figure 37). Furthermore, the test data implies a clear trend for γ0.25 (the shear strain at 25% of the maximum shear stress) to increase with increasing γcyc (the maximum cyclic shear strain in each cycle) and a conservative best fit to the data may be expressed as:

Figure 35. Comparison of stress-strain response.

3.6.1 Inflection point As discussed in Section 3.4.2 all load cycles after the first load cycle have a three part p-y format, with the first two parts describing a linear or concave upwards

Having obtained a relationship between γ0.25 and γcyc , it was necessary to define an equivalent relationship between y0.25 and yapp , where yapp is any given lateral movement of the pile. To assess this, some postprocessing was performed of the various horizontal slice FE/ FD analyses discussed earlier. This postprocessing took the form of evaluating the shear strain


Figure 37. Detailed stress-strain responses.

(i.e. γcyc ) obtained at specific load levels throughout the FE/FD meshes used for these analyses, assessing γ0.25 at each element from Equation (6), and then determining appropriate weighted average values for γcyc and γ0.25 . It was then assumed that y0.25 can be calculated as a function of yapp as:

Figure 38. p-y inflection point data.

Various methods were tried in order to assess appropriate weighted values for (γ0.25 /γcyc )ave and it was found that the best approach, in terms of overall consistency and minimising the dependence of the final result on the size of the mesh used in the numerical analysis, was as follows:

where: Ne = the number of elements in the FE/FD mesh. dA = the area of element i in the FE/FD mesh. γ0.25 and γcyc are the relevant strains for element i in the FE mesh. This method gives greater weight to the values of γ0.25 /γrmcyc obtained in parts of the mesh that experience higher cyclic shear strains compared to those regions that are more distant from the pile and therefore experience lower cyclic shear strains. The weighting by γcyc dA was selected intuitively, but it would also be considered theoretically robust if a linear trend could be demonstrated between yapp and the integral of γcyc dA over the whole mesh. This was there fore checked for a variety of cases (su /σvo = 0.35 to 1.5 and N = 1 to 30) and it was found that an approximately linear response was obtained in all cases, with no more than an 8% (and generally a lot less) deviation from a linear relationship, which is acceptable for practical engineering purposes. The minimal mesh dependency obtained with this approach is illustrated on Figure 38 which presents y0.25 /D versus yapp /D for four test cases, where D is the pile diameter. These results were obtained from four horizontal slice analyses, all of which have meshes that contain an equal number of elements but with the outer boundary steadily increased (i.e. increasing outer

Figure 39. Model tests; gap width vs. cyclic displacement.

diameter (OD) to inner diameter (ID) ratio). Very similar results were obtained irrespective of the magnitude or N, which allowed for a single trendline to of su /σvo be fitted and incorporated into the p-y model, as also shown on Figure 38. It is interesting to compare the results presented on Figure 38 with the results of a series of laterally loaded model pile tests in carbonate soil undertaken at the University of Sydney (Randolph et al., 1988). Under 2-way loading these tests revealed a pronounced ‘S’ shaped hysteresis curve, with low lateral resistance obtained in the central region, which was assumed to be associated with a gap zone or softened soil. The size of this ‘gap zone’ (ug ) was measured and, as shown on Figure 39, was plotted in an identical format to that for y0.25 on Figure 38. Comparing these two figures it may be seen that the design line for y0.25 follows the same trend as the various lines presented for ug , but with a slightly steeper gradient.


3.6.2 Stiffness degradation for reducing loads As discussed in Section 3.4.3, for N > 2 two criteria are used to determine the p-y response when the lateral pressure applied during the previous cycle (pi-1 ) is less than the maximum pressure that has ever been applied, pmax . The second of these criteria first determines the p-y response that would be obtained if the pile-soil element had been subject to i − 1 cycles with an applied pressure equal to pmax , and then imposes a scale back factor to the predicted degree of softening obtained from this calculation to account for the fact that the pile-soil element is now being cycled at a pressure, pi-1 , that is less than pmax . The approach that we have used to assess the magnitude of the scale back factor is to consider the relationship between normalised stress (or strength) and the corresponding equivalent amount of damage (i.e. incremental cyclic shear strain) at varying levels of cyclic strain. The initial step in this assessment was to examine the available data from 2-way strain controlled cyclic simple shear tests performed on uncemented carbonate soil samples. The results of these tests were assessed in terms of the normalised stress ratio τ/τref (where τref is the peak shear stress attained during the first cycle) versus the stiffness degradation ratio Kdeg at selected cycle numbers during each test. The value of Kdeg is in turn defined as γ/γref (where γref is the change in γ between the first and second cycles, while γ is change in γ between any other two cycles). For displacement controlled tests the strain does not change between one cycle and the next, but the stiffness does and hence it is still possible to ascertain γ and γref in an indirect manner as shown on Figure 40. Curves were fitted to the τ/τref versus Kdeg response for each test according to the following relationship:

where the exponent n was found (empirically) to vary as a function of the cyclic shear strain. The experimental results have demonstrated an appropriate form for Kdeg in terms of stresses and strains. However, for application to the lateral pile analysis, we need to transform these results into an equivalent format but with stress replaced by the lateral pressure acting on the pile, and with strain replaced by pile lateral movement. For this exercise we have assumed that pressure may be directly substituted for stress; i.e. P/Ppeak acting on the pile is assumed to be directly proportional to the values of τ/τref acting throughout the soil. We have then interrogated the various FE/FD analyses in order to determine a value for Kdeg-ave , which is the average stiffness degradation factor value applicable to a pile rather than a specific value that applies for a single soil element. For specified values of P/Ppeak , the appropriate values of γcyc and τ/τref are determined on an element by element basis from the FE/ FD analysis results, with γcyc first being used to determine the appropriate value for the exponent n,

Figure 40. Calculation of stiffness degradation from strain controlled tests.

which is then inserted along with τ/τref into Equation (9) in order to determine element specific values of Kdeg . The weighted average value of Kdeg-ave is then obtained following a similar approach to that used for assessing y0.25 , as discussed in the previous section:

Using this approach it was found that the best-fit relationships obtained for any specific value of yapp /D varied by only a modest degree for different soil types and cycle numbers, but there was a systematic variation with varying yapp /D. This is demonstrated on Figure 41, which presents the best fit curves of the form described by Equation (9) for varying yapp /D but with τ/τref replaced by P/Ppeak . As might be expected it can be seen that the stiffness degradation factor is higher (ie. more degradation) when the cyclic displacement amplitude is large. With Equation (9) recast in terms of P/Ppeak , unique values for the exponent n were then determined as a function of yapp /D. This simple equation finalises the determination of the p-y response when the lateral pressure applied during the previous cycle (pi-1 ) is less than the maximum pressure that has ever been applied, pmax . 3.7

A computer program named pCyCOS has been developed to incorporate the new p-y model that is presented here. This program is based on the platform that was developed for the CHIPPER program (Erbrich, 2004) since the new algorithm has many structural similarities to the CHIPPER algorithm. It was therefore


Description of the pCyCOS program

– Pile geometry parameters (diameter, wall thickness, length, stickup, Young’s modulus). – Pile loads (various combinations of lateral load, head moment, lateral head displacement and pile head rotation are permitted). – Whether or not monotonic or cyclic loading is to be considered, and if the latter, how many cycles are to be applied. Post-cyclic monotonic loading can also be specified. Output obtained from pCyCOS includes full information along the pile length after the required number of cycles has been applied (displacement, bending moment, shear force and reaction pressure), and p-y data pairs along the pile. The latter data can be compiled from various analyses at different load levels (with a constant N) to give p-y curves that can be used as input into structural analysis packages such as SACS or SESAM. It should be appreciated that a single unique p-y curve does not exist for any particular depth in the soil, even for a constant N, since it is a function of the pile head fixity. Hence it is necessary to perform full pile analyses for each particular case, and an iterative approach will be required with the structural designer in order to ensure that the finally defined p-y curves are consistent with the pile head fixity arising out of the structural model.

Figure 41. Stiffness degradation for reducing cyclic loads.

relatively straightforward to incorporate the new p-y model into CHIPPER, while retaining much of the pre-existing code. The existing CHIPPER code has been extensively tested and verified and hence this approach also helped to minimise the risk of bugs being introduced. It should be understood that the new program does not perform a full time domain analysis, such as that used in the CYCLOPS program for axial pile design discussed in Part 1 of this paper. Cyclic loading in the pCyCOS program is assessed on a peak-to-peak basis, meaning that the lateral response is only calculated at the peak of each cycle. This approach is consistent with that generally used to determine the undrained cyclic bearing capacity of shallow foundations and suction caissons over the last 30 years (Hoeg, 1976). In principle, analyses of full storm histories could be performed but in practice there are several important reasons why this is not recommended. The most important of these is that the assumptions made about the p-y form are really only appropriate when considering relatively few cycles. The appropriate equivalent number of cycles should therefore be calculated from the full design storm history data using strain accumulation methods similar to those used for these other applications. The input parameters required for pCyCOS include: – Profile of monotonic undrained shear strength, su , as a function of depth and the effective unit weight of the soil, γ . – Envelope p-y curves for various values of su /σvo , sufficient to encompass the full range of monotonic strength ratios encountered in the design profile, and for various numbers of cycles between N = 1 to N = ne + 2 where ne is the number of equivalent cycles that need to be modelled. Linear interpolation is used to assess any intermediate values of su /σvo and N from the input data set. – Parameters to define the stiffness degradation function. – Parameters to define the p-y curve inflection point.

3.8

Two examples are presented here to illustrate the performance of the pCyCOS model. It is unfortunate that there are no suitable prototype scale lateral load tests in uncemented carbonate soil which can be used to validate the model presented in this paper. The limited available historical data are generally for fully drained conditions, which are quite different to the undrained conditions that are applicable for the typical large diameter piles used in offshore construction, even for quite sandy conditions (albeit the equivalent number of cycles for these more sandy soils may be only 1 to 3). The first example presented demonstrates the key characteristics of the cyclic degradation model, and are compared qualitatively with the results of some centrifuge model tests. The second example considers a ‘typical’ offshore pile, and the results from pCyCOS are compared to those predicted with other commonly used methods. 3.8.1 Example 1 The first example illustrates the global stiffness response of a single rigid pile subject to 2-way cyclic loading with fixed displacement amplitude. A rigid 2.83 m diameter by 20 m long pile subject to pure translation is considered for this case. A monotonic undrained strength profile that varies linearly with depth has been assumed, defined as su = 0.5 σvo . The response obtained for this pile in terms of pile head load versus number of applied load cycles is illustrated on Figure 42. Results are presented for a


Example applications

Figure 42. Example 1 – Displacement controlled cycling.

range of pile head displacement amplitudes, varying between 50 mm and 3 m. At the smallest displacement amplitudes it can be seen that the pile head load in the first cycle is modest, but this load hardly degrades with continuing cycles. As the cyclic amplitude increases, the pile head load in the first cycle steadily increases but increasing levels of degradation occur during subsequent cycles. After 10 cycles, the highest head load that can still be sustained is found to be for a 200 mm head displacement, albeit the stiffness has degraded by around 40% from that obtained during the first cycle. As the pile head amplitude is increased even further, the 10 cycle response gradually degrades to a very low value; at 3 m head displacement over 5 MN can be applied during the first cycle but not much more than 100 kN during the 10th cycle. This sort of response is entirely consistent with expectation and implies that the soil is liquefied by the very large stresses imposed during the first cycle. The dramatic degradation of soil stiffness under cyclic loading, as simulated by the pCyCOS model, has also been demonstrated in the laboratory as shown on Figure 43. This figure presents p-y curves extracted from two tests (numbered 1111 and 1131) which are extracted from Dyson (1999). These were obtained from centrifuge model tests of laterally loaded piles in carbonate silt. The p-y curves are neither constant load nor constant displacement amplitude but still show the same mechanisms as revealed on Figure 42. For example, comparing the results at the shallowest depth of 0.5 pile diameters and the deepest depth of 8 diameters it may be noted that that a much higher p/su ratio was mobilised in the first cycle at the shallow depth than at the deep depth. However, after several load cycles, the resistance of the upper soil has degraded to almost nothing (disregarding a small negative offset, which is assumed to be an interpretation/ measurement error), whereas a reasonable resistance is still obtained (albeit reduced from the first cycle) at the deepest depth.

Figure 43. Typical p-y response in carbonate silts (centrifuge).

3.8.2 Example 2 The second example represents a typical large diameter offshore pile, 2.65 m diameter by 75 m long, with a wall thickness varying from 110 mm over the top 20 m, to 45 mm below a depth of 30 m, and 85 mm in between. We have performed analyses using pCyCOS, the API soft clay p-y criteria and the method proposed in Novello (1999). Lateral pile analyses were performed for the extreme cases of fully fixed or fully free pile heads and we have considered both monotonic and cyclic loading. The API soft clay procedure has often been considered for many of the silty and muddy carbonate soils found offshore Australia, particularly under monotonic loading where the complications of liquefaction under cyclic loading do not apply. The monotonic undrained shear strength profile used in both the pCyCOS andAPI analyses is presented on Figure 44, which is representative of a typical uncemented carbonate silt at most depths, interspersed with a few stronger and


Figure 44. Undrained strength for example analysis.

more sandy layers. The API cyclic method is not tied to any specific number of cycles and simply represents an envelope of ‘residual’ cyclic response for clay. However, the pCyCOS response is a direct function of the number of applied load cycles. Hence for this example we have adopted 20 maximum load cycles, which is a typical value to represent a design storm in fine grained soil, where fully undrained conditions can be expected over the duration of a storm. The other parameters required for the API analysis are ε50 and J, for which values of 1% and 0.5 were adopted respectively. As discussed in Section 1.2 the Novello (1999) procedure is formulated in terms of cone resistance, qc , rather than undrained shear strength and hence it is necessary to derive an appropriate profile of qc for the analysis. For the current example we have adopted a cone factor, Nkt , of 12 which is a typical value for undrained carbonate silts. The required profile of qc is then simply determined as 12 times the undrained shear strength. The Novello procedure nominally incorporates cyclic degradation in accordance with Equation (3). It is suggested in Novello (1999) that by adopting a typical K0 of 0.4 for uncemented carbonate soil, Equation (3) can be simplified to pcyclic = (1-0.6U∗ )pstatic . A complication with this approach is that U∗ needs to be determined from a completely separate analysis, and hence retaining internal consistency is rather difficult. We have therefore not performed an explicit cyclic analysis with the Novello procedure. However, it is still important to note that, by definition, the Novello procedure implies a maximum softening of only 60% in a fully liquefied soil (ie. u/σvo = 1), which is not consistent with available cyclic laboratory test data on carbonate soils, nor with the centrifuge tests discussed earlier (Dyson, 1999), and nor with the predictions of pCyCOS, as demonstrated by Example 1. The pile head load displacement responses obtained from the various cases considered are presented on Figure 45. The maximum bending moments at the pile

Figure 45. Comparison of pile head lateral deflection.

head for the fixed head cases are summarised on Figure 46, while the maximum bending moments within the body of the soil are summarised for both fixed and free pile heads on Figure 47. It may be noted that the pCyCOS and API results are remarkably similar for monotonic loading, with the pCyCOS response being marginally stiffer at low load levels and marginally softer at higher load levels. The pile bending moments also exhibit the same trend. This is surprising and probably coincidental given that the API approach is essentially a purely empirical method derived from load tests on soft clays, whereas pCyCOS is a theoretically based model, with the p-y curves determined from FE/ FD analyses using laboratory measured stress-strain curves from carbonate soils. We do not anticipate that such good agreement would always be obtained; for example in denser mate rials with higher su /σvo ratios than considered in this example. The Novello (1999) procedure leads to a generally softer monotonic response than obtained from the other two methods. This is not entirely surprising since drained lateral pile load tests are the empirical basis for the method, and the high compressibility of the carbonate soils in those tests has led to a significantly softer


Figure 46. Pile head bending moments.

response than would be expected under incompressible undrained conditions. Hence while the Novello procedure purports to be able to address undrained loading, with cyclic softening also included, we consider that the method, and all similar methods, are only applicable where fully drained loading conditions are expected to apply. Whereas close correlation was obtained between the pCyCOS and API methods for monotonic loading, a rather different result is obtained under cyclic loading. Except at the lowest load levels, a much softer response and significantly higher bending moments are obtained with pCyCOS compared to the API method. However, this is not unexpected since normal clays do not exhibit the significant cyclic degradation (‘liquefaction’) that occurs with carbonate soils. The softening behaviour can be clearly seen on Figure 48, which compares the pCyCOS reaction force distributions along the pile at 1 and 20 applied load cycles for a fixed head pile subject to a 12 MN head load. It can be seen that as the upper part of the soil softens, the load is shed down the pile. This inevitably increases the pile deflections and the pile bending moment as shown on Figure 49. The change in reaction force is also illustrated on a cycle by cycle basis on Figure 48, at three points on the pile.

Figure 47. Maximum bending moments below mudline.

Figure 48. Reaction forces.

The applied reaction pressure in each cycle is normalised by the monotonic ultimate resistance, pu , applicable at each specific depth on this figure. Once again it can be seen that shedding occurs over the


Figure 49. Deflections and bending moments.

Figure 50. Monotonic and cyclic p-y curves.

upper section of pile, where high reaction pressures are mobilised in the first cycle. These are transferred down the pile into regions where lower, and hence nondegrading reaction pressures are mobilised in the first cycle, which allows the applied reaction pressures to increase with advancing cycles. Example p-y curves extracted from pCyCOS at selected depths for a fixed head pile are presented on Figure 50 for monotonic loading and for N = 20. These demonstrate nicely the substantial difference between monotonic and cyclic loading conditions. The resulting p-y curves can be input directly into structural analysis packages, although some manipulation of the cyclic p-y curves may be required to avoid numerical difficulties associated with their strain softening nature. 4

CONCLUSIONS

This paper has presented and demonstrated application of two new tools for the analysis and design of piles in carbonate soil: CYCLOPS for axial loading and pCyCOS for lateral loading. The procedures used in both


are underpinned by a strong theoretical basis, but also include a number of factors determined from empirical calibrations in order to properly address observed soil behaviour. A key objective in the development of these methodologies was to simulate real behaviour as accurately as possible, but where the available data led to any doubt, conservative assumptions have been adopted. The procedures were also developed to be applicable to a wide variety of general situations, although as with all geotechnical design in carbonate soils, great care and good engineering judgement remain vital ingredients. With respect to axial pile design, we consider that the ‘CNS calibration approach’ embodied in CYCLOPS is an essential process for developing safe designs of cyclically loaded drilled and grouted piles in cemented carbonate soils.Actual sites exhibit highly variable properties; these may differ significantly from the specific characteristics encountered at Overland Corner, which remains the only site where a comprehensive program of cyclic pile tests has been undertaken in cemented carbonate soil. Direct scaling of the OC test results is therefore not recommended. The CNS device is the main tool available for differentiating the behavioural characteristics between different soil types and hence underpins the CYCLOPS method. With respect to lateral pile analysis, the pCyCOS procedure is also based on a rigorous link to laboratory monotonic and cyclic element testing, from which the basic p-y curves are developed. It is unfortunate that at this stage we do not have any equivalent of the OC tests for cyclic undrained lateral loading in uncemented carbonate soils. However, the centrifuge tests reported in Dyson (1999) show the same qualitative response as obtained with pCyCOS. It should also be appreciated that this method only represents a modest philosophical advance on the industry standard procedures that would be adopted for design of a shallow foundation or suction pile, albeit the practical implementation for lateral pile analysis proves to be necessarily complex. The reality of the destructive effects of cyclic loading on shallow foundations in carbonate soils has been amply demonstrated in full scale field situations (Erbrich, 2005). While the approach in pCyCOS may yield significantly more onerous results for cyclic loading than implied by current procedures (and this may have implications for existing structures), the paramount concern has been to ensure safe design, taking due account of the observed behaviour in laboratory testing. To date, the authors have utilised CYCLOPS and pCyCOS for the design of several major new facilities and all these designs have been accepted by third party certification bodies. Finally, it should be appreciated that while the discussions in this paper have centred on carbonate soils, the methodologies developed are of more general applicability. For example, pCyCOS could be used for any kind of uncemented soil, provided suitable laboratory test data is available to enable development of

the required p-y curves. In practice, this would probably only be a worthwhile activity where the soil in question might be subject to significant cyclic degradation under the design loading. Similarly, CYCLOPS has been developed specifically for weakly cemented carbonate rocks, but could also be applied to axial pile design in other kinds of soft rock, provided suitable CNS test data is available to allow determination of the necessary calibrated parameters. ACKNOWLEDGMENTS The authors would like to acknowledge their colleagues who made significant contributions to this work. In particular we would like to thank Professor John Carter who performed detailed independent analyses to verify the pCyCOS program, Edgard Barbosa Cruz, who performed the ABAQUS analyses used in the development of pCyCOS and to Mohamed Khorshid who provided support and encouragement throughout this work. We would also like to thank the many others in both AG and throughout our industry who provided constructive and valuable contributions to the development of these new design tools. REFERENCES ABAQUS Inc. (2006), ABAQUS Version 6.6, User Manual, Dsimuli Abbs, A.F. (1983). Lateral pile analysis in weak carbonate rocks. Proc. Conf. on Geotechnical Practice in Offshore Engineering, ASCE, Austin, 546–556. API (2000). API RP2A – Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms—Working Stress Design, 21st Edition, American Petroleum Institute, Washington. Coyle, H.M. and Reese, L.C. (1996). Load transfer for axially loaded piles in clay. J. Soil Mechanics and Foundations Division, ASCE, 92(SM2), 1–26. Dyson G.J. (1999), Laterally Loaded Piles in Calcareous Sediments, PhD Thesis, UWA. Dyson, G.J. and Randolph, M.F. (2001). Monotonic lateral loading of piles in calcareous sediments. J. Geotech Eng. Div, ASCE, 127(4), 346–352. Erbrich, C.T. (2004). A new method for the design of laterally loaded anchor piles in soft rock. Proc. Offshore Tech. Conf., Houston, Paper OTC 16441. Erbrich, C.T. (2005). Australian Frontiers – Spudcans on the Edge. Proc. International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth.

Haberfield C.M. Paul D.R. Ervin M.C. and Chapman G.A. (2010) Cyclic Loading of Barrettes in Soft Calcareous Rock using Osterberg Cells. Proc. International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth. Hoeg K. (1976), State of the Art – Foundation Engineering for Fixed Offshore Structures, Proc. BOSS76, Trondheim. ISO (2007). ISO 19902 – Petroleum and Natural Gas Industries — Fixed Steel Offshore Structures. International Standards Organisation, Geneva. Itasca (2005), FLAC Fast Lagrangian Analysis of Continua, User Manual, Itasca Consulting Group. Jeanjean, P., Watson, P.G., Kolk, H.J. and Lacasse, S. (2010). RP 2GEO: The new API recommended practice for geotechnical engineering. Proc. Offshore Technology Conf., Houston, Paper OTC20631. Kraft, L.M., Focht, J.A. and Amarasinghe, S.F. (1981). Friction capacity of piles driven into clay. J. Geotechnical Engineering Division, ASCE, 107(GT11), 1521–1541. Murff, J.D. & Hamilton, J.M. (1993), P-Ultimate for Undrained Analysis of Laterally Loaded Piles, Journal of Geotechnical Engineering Div. ASCE, 119(1), pp 91–107. Novello, E. (1999). From static to cyclic p-y data in calcareous sediments. Proc. 2nd Int. Conf. on Engineering for Calcareous Sediments, Bahrain. 1, 17–27. Randolph, M.F. (1986). RATZ: Load transfer analysis of axially loaded piles. Report No. Geo:86033, Department of Civil Engineering, The University of Western Australia. Randolph, M.F. (1988). Evaluation of grouted insert pile performance. Engineering for Calcareous Sediments, Perth, 2, 617–626. Randolph, M.F. (2003), RATZ Version 4.2 – Load Transfer Analysis of Axially Loaded Piles, User Manual. Randolph M.F., Jewell R.J. & Poulos H.G. (1988), Evaluation of Pile Lateral Load Performance, Proc. Eng. for Calcareous Sediments, ed. Jewell R.J. & Khorshid M.S. Randolph, M.F. and Jewell, R.J. (1989). Axial load transfer models for piles in calcareous soil. Proc. 12th Int. Conf. on Soil Mech. and Found. Eng., Rio de Janeiro, 1, 479–484. Randolph M.F., Joer H.A., Khorshid M.S. and Hyden A.M. (1996), Field and laboratory data from pile load tests in calcareous soil’, Proc. 28th Annual Offshore Tech. Conf., Houston, Paper OTC7992. Randolph, M.F. and Wroth, C.P. (1978). Analysis of deformation of vertically loaded piles. J. of Geotechnical Engineering Division, ASCE, 104(GT12), 1465–1488. Wesselink, B.D., Murff, J.D., Randolph, M.F., Nunez, I.L. and Hyden, A.M. (1988). Analysis of centrifuge model test data from laterally loaded piles in calcareous sand. Engineering for Calcareous Sediments, Perth, 1, 261–270. Wiltsie, E.A., Hulett, J.M., Murff, J.D., Brown, J.E., Hyden, A.M. and Abbs, A.F. (1988). Foundation design for external strut strengthening system for Bass Strait first generation platforms. Engineering for Calcareous Sediments, Perth, 1, 321–330.



New frontiers for centrifuge modelling in offshore geotechnics C. Gaudin Centre for Offshore Foundations Systems, UWA, Perth, Australia

E.C. Clukey BP America Inc., Houston, USA

J. Garnier Laboratoire Central des Ponts et Chaussées, Nantes, France

R. Phillips C-CORE, Newfoundland, Canada

ABSTRACT: Today, centrifuge modelling is a recognised tool, both within the academic and the industry community, to help understanding the behaviour of offshore structures and assist in their design. However, although expanding, its real impact on design practices is limited. The paper proposes, therefore, to revisit the perception of the advantages and disadvantages of centrifuge modelling and the technological developments achieved over the last few years, in the light of recent offshore projects undertaken. It aims to provide both academic and industry perspectives. Examples are presented that illustrate the difference between centrifuge testing and centrifuge modelling and which highlight the contribution that centrifuge methods can make in designing offshore structures and in providing new insights into soil-structure interaction. 1

INTRODUCTION

The initial use of centrifuge modelling in offshore geotechnics took place in Manchester University in 1973 where the behaviour and performance of gravity platforms to be used in the Gulf of Mexico was investigated (Rowe & Craig, 1981). The work encompassed a wide range of soil and loading conditions and provided pivotal insights into the failure mechanism taking place (Craig & Al-Saoudi, 1981). Even in the very early days of centrifuge modelling (when the technological capabilities were limited compared to present day), it was yet understood and acknowledged that centrifuge modelling could contribute significantly to design when novel or incompletely understood conditions (both in terms of soil and loading) prevailed (Craig, 1984). Since this pioneering work in 1973, a significant number of projects have been undertaken worldwide, on a wide variety of offshore geotechnical problems. A few centrifuge centres have emerged, in the wake of Manchester University, developing expertise on offshore geotechnics through collaboration with industry, such as Cambridge University in the UK (CUED), C-CORE in Canada, LCPC in France, Deltares (former Geodelft) in Netherlands, University of Colorado, Boulder in the USA and COFS at the University of Western Australia.

The work performed, focusing first on phenomenological and site-specific studies, developed progressively towards more general studies, including the observation of failure mechanisms and the understanding of soil-structure interaction, eventually aiding in the development of predictive design methods in some cases. This transition from what was initially centrifuge modelling to centrifuge testing is discussed further in the paper. As centrifuge modelling was technically and scientifically developing, along with an increasing need for performance data and understanding of offshore soil structure interaction, the acceptance and awareness by the offshore community of the benefits of the use of centrifuge grew significantly. A key milestone in that process was certainly the keynote address given by Don Murff at OTC (Offshore Technology Conference) in 1996 to the wider offshore community, advocating the benefits of centrifuge modelling (Murff, 1996). Since then, the number of offshore projects benefiting from centrifuge modelling input has increased significantly and industry users have developed a strong expertise in analysing centrifuge modelling outcomes and incorporating them into their design approach. Nevertheless, as new needs arise from the increasing complexity of offshore projects, along with new possibilities from the technological development of new modelling techniques, it appears necessary to


revisit the benefit and the contribution centrifuge testing and modelling may provide to the offshore industry in designing offshore structures subsea facilities and pipelines. This paper aims then to answer the fundamental questions any potential user has once facing a challenging offshore project: Why use a geotechnical centrifuge? Is it cheaper or more expensive than alternative investigation technique?; is it more or less reliable?; is it quicker to obtain relevant data?; does it account for behaviour aspects other techniques do not?; should it be replace or complement other techniques such as 1g reduced scale modelling or field testing? How to use a geotechnical centrifuge? Is it better to use natural or artificial soils?; should it investigate a wide range of options or focus on a particular solution?; should it aim at validating a design or assisting in the design; should it be modelling or testing? When to use a geotechnical centrifuge? Should it be considered at the early stages of the project or latter once some key parameters are already established?; at what stage within the design process is centrifuge modelling the most valuable? To answer these three questions, this paper proposes: 1. to revisit the advantages (and disadvantages) of the use of centrifuge in light of the scientific and technological developments of the past decade; 2. to detail the contribution of the use of centrifuge on industry practices, highlighting the difference between centrifuge testing and centrifuge modelling, an; 3. to evaluate the impact of centrifuge modelling/testing in industry practice and eventually to present some perspective for the future.

to obtain qualitative or quantitative information about the problem. The difference between centrifuge testing and modelling follows similar concepts. By performing centrifuge testing, one aims at validating or confirming a general set of assumptions about a geotechnical problem, while by performing centrifuge modelling, one aims at predicting or anticipating the behaviour or response of an actual geotechnical structure. The difference is significant as the requirements, the modelling techniques and the objectives for the modeller are different, as will be the outcomes for the user. 2.1

Focusing on soil-interaction structure, examples are presented from the authors’ experience, providing the perspective of both academic modellers and an industry user. Indeed, this presents a small part of the entire centrifuge work performed world wide. However, the authors believe that the variety of examples is representative of the contribution and possibilities that centrifuge testing and modelling can offer for the offshore industry.

2

ROLES OF CENTRIFUGE TESTING AND CENTRIFUGE MODELLING

The process of designing, performing, analysing centrifuge experiments and eventually integrating the outcomes into the design of offshore structures cannot be undertaken correctly without a clear understanding of the difference between centrifuge testing and centrifuge modelling. In engineering, it is common to refer to in-situ testing and numerical modelling. The former is usually used to validate an assumption (soil properties, design loads, soil response, etc) while the latter resorts to an idealisation of the problem investigated


Centrifuge testing

The aim of centrifuge testing is to create an idealised representation of a problem in order to obtain quantitative and/or qualitative prediction about the mode of behaviour of the structure investigated, (Lee, 2001). It is therefore necessary to interpret the results in light of a model for soil-structure behaviour, and then use that model to simulate the design situation. Being an idealisation, the model does not necessarily replicate all the features of a real structure (commonly called prototype) and consequently, some flexibility is permitted regarding the representativeness of the model as long as the key features of the behaviour which are investigated are modelled correctly. The difficulty is to know which features are simulated exactly, approximately or not at all and to account for these differences in the interpretation of the results. The complementary use of other modelling and testing methods (such as finite element analysis) may help in this interpretative process. The idealisation relates mostly to the soil and the structure as follows: – The soil used in the model may be different from the prototype as long as some pivotal features of the behaviour are reproduced (such as the dilatancy of dense sand or the shear strength of cohesive soil, for instance). Local heterogeneity and non-critical stratigraphy or topography are often not simulated, but may still be accounted for to interpret dissimilar behaviour. – Some particular features of the prototype (geometric, mechanics, etc), which are believed to not have a significant effect on the phenomenon investigated, are not simulated. – Phenomenon aside of the ones investigated, and for which the influence is not believed to be dominant, are not simulated (a specific method of installation, a particular loading sequence, etc). In designing the model and choosing which features to replicate or not, the engineer needs to make assumptions about the behaviour to be modelled, to understand the consequences of the simplifications made and to define clearly the purpose of the model. Centrifuge testing may be conducted to: – Develop an initial understanding of the engineering concern.

– Identify a particular failure mechanism on which an analytical solution can be developed. – Observe a particular mode of soil behaviour (is the response drained or undrained, does the soil flow or collapse? etc). – Gather performance data which can be used to calibrate a numerical model. – Perform a sensitivity study to understand the relative importance of various parameters in the behaviour of the structure. – Perform parametric studies to generate design charts. – Determine soil properties by performing reduced scale testing of in-situ soil properties.

2. The requirements to represent the prototype are much stricter, particularly with regards to the soil characteristics and the loading conditions. Hence, as opposed to centrifuge testing, one, when designing centrifuge modelling should ensure that: – The same soil, featuring the same properties as the prototype, is used in the model. – The model considers all the details of the prototype (e.g. particular attention must be paid to the fidelity of the model). – All key phenomena involved in the response of the structure must respect the similitude principles. – Loading conditions must be the same between the model and the prototype (rate of loading, amplitude and frequency of loadings, etc).

Indeed, the objectives sought may affect the idealisation made as much as the assumptions and some judgement will be required to decide whether or not the outcomes may be directly extrapolated to prototype conditions and whether or not they need to be checked or validated by other means. The objectives presented could sometimes be achieved by other means, such as by numerical modelling, but the centrifuge offers the advantage to replicate specific features of the soil behaviour or the structure loadings. These include soil softening and reconsolidation, large deformation, cyclic loadings, collapse and the installation process. All of these features are particularly relevant for offshore geotechnics. Often, the complementary use of both numerical modelling and centrifuge testing accelerates the understanding of the phenomena, as the two methods are based on different sets of assumptions.

In other words, in centrifuge modelling, all aspects of the model must be properly scaled to that of the prototype. The objectives of centrifuge modelling may be various, although they are limited to the case studied. Extrapolation to other conditions should be undertaken with caution. Centrifuge modelling should help to: – Validate a technical solution (type and geometry of structure). – Validate or evaluate the performance of a final design (design loads, displacements within the limits established, etc). – Validate a mode of behaviour upon which the design was based (i.e. mode of collapse). – Observe the response of a given structure under a particular type of loadings, possibly different from the one used for design (e.g. check the performance of the structure under cyclic loadings).

2.2 Centrifuge modelling The aim of centrifuge modelling is to establish the validity/relevance of certain assumptions, designs or models. It aims to directly replicate the field conditions. The results provide an immediate representation of the design situation. What characterises centrifuge modelling is the uniqueness of the model investigated. It must refer to a specific prototype (as opposed to centrifuge testing) and focus on a particular aspect of the validation. This allows definitive conclusions about the prototype behaviour to be made. Note that the assumptions, design or model to be validated might have arisen from centrifuge testing (Lee, 2001). The reference to a particular prototype results in two fundamental differences with centrifuge testing: 1. The use of similitude principles and associated scaling laws (Garnier et al., 2007) is required to extrapolate the information gathered on the model (typically the measurements of dimensions such as loads, displacements, etc) to the prototype. The use of similitude principles in centrifuge testing is indeed required (this is the justification for the use of the centrifuge), but the scaling laws (e.g. the use of prototypes dimensions) are not necessarily required.

The objective presented here could be also achieved by field testing. However, centrifuge modelling offers the advantage (while ensuring the correct stresses are applied within the soil), of being less costly, provides a better control of the modelling conditions (e.g. known soil characteristics, better quality of data), and allows more data to be collected in a shorter period. It also provides the opportunity to repeat the experiments if required.

2.3 The use of centrifuge outcomes into design The design of an offshore foundation structure follows a series of steps, which are simplified in Figure 1. The soil and loading conditions will dictate a particular type of foundation whose geometry and characteristics will be determined using design methods, either empirical or based on constitutive soil models linking strains and displacements to the stresses and loads. Along with these different steps, the geotechnical engineer will use various tools to help him assess the soil conditions, choose the relevant technical solution, develop or use an existing design method and eventually validate the design.


Figure 1. Steps involved in designing geotechnical structures and potential input of centrifuge modelling and testing (white shapes symbolises design steps, shaded shape symbolises centrifuge modelling or testing).

These tools can be divided in two categories; field testing and numerical/analytical modelling. Field testing is used to characterise the soil and help validate the design, while the modelling methods would be used to develop the design. The objective of this paper is to describe the potential centrifuge methods may have to assist in geotechnical design. As illustrated in Figure 1, by performing both testing and modelling, centrifuge methods may be a valuable tool for geotechnical engineers in every phase of the engineering process. Centrifuge testing may be used to develop a new foundation concept, a new technical solution, or to characterise soil properties. It may also be used to investigate soil behaviour and failure mechanism to establish constitutive models. In contrast, centrifuge modelling may be used to validate a new foundation concept, to validate a design and to calibrate and validate a specific design method. In offshore geotechnical engineering, where guidelines and design rules play a lesser role than onshore, engineers’ judgement is paramount in establishing the most technically sound and economically efficient design. Basing judgment on reliable and quality data from a range of modelling and testing methods may make a significant difference between a satisfactory and an optimal design.

The following sections will present various examples of centrifuge testing and modelling, highlighting the contributions to design. It is not implied here that centrifuge methods should necessarily replace more classical methods. Indeed, for routine design, classical methods are often preferred as they provide a safe and efficient path towards satisfactory solutions. However, for more challenging geotechnical design issues, which frequently occur in offshore engineering, geotechnical engineers should seriously consider the potential centrifuge methods may offer. 2.4

The offshore industry has come to accept the advantages of centrifuge methods as more and more results from various studies became available. ExxonMobil (former Exxon) were initial leaders in this area with a number of programs performed in the early 1980s (Murff, 1996). With the publication of the ExxonMobil results and, as the industry encountered more complex problems where 1-g testing was either difficult and/or expensive to perform, centrifuge testing or modelling became a more viable option for other operators. Joint Industry Programs involving a significant centrifuge testing component have become increasingly more frequent in the past 2 decades, such as


Industry perspective

the SEPLA: suction emplaced plate loading anchor JIP involving 5 operators and 2 consulting groups in early 2000 at C-CORE. The Pressure Ridge Ice Scour Experiment (PRISE) performed at C-CORE attracted over 7 operators and 2 regulatory agencies and developed guidelines for design of buried offshore pipelines subject to ice scouring (Phillips et al., 2005). The INSAFE JIP, gathering 19 offshore operators, jack-up manufacturers, safety regulators and consulting companies, has been using centrifuge testing and modelling outcomes initially from Cambridge University, and more recently from COFS and NUS (National University of Singapore) to provide guidelines for design and best practice of jack-up installation and performance. Other examples of offshore JIPs using centrifuge techniques, in complement of more traditional techniques, includes the SAFEBUCK JIP for the design of on—bottom pipeline subjected to buckling, the MERIWA JIP, involving six major offshore companies, to model submarine landslide and their impact on pipelines. In France also, several centrifuge projects were carried out at LCPC, founded by CLAROM (www.clarom.com), an institution which gathers marine and offshore companies to carry out R&D programs aiming at better understanding offshore structures behaviour and improving design methods. The success of centrifuge techniques outside the offshore community has also helped to convince management to support various centrifuge efforts. The ability, for instance, of centrifuge tests to accurately define the failure mechanism for the levy failures in New Orleans during Hurricane Katrina is a recent example of the usefulness of the technique for defining failure modes. (Sasanakul et al., 2008 and Ubilla et al., 2008). The use of centrifuge testing by numerous earthquake researchers involved in a very large United States program (Wilson et al., 2010) also helped improve the credibility of the methodology which led to testing of large gravity based structures under ice and seismic loads (DeWoolkar et al., 2006 and DeWoolker et al., 2008). More recently the industry has used centrifuge techniques for a large number of deepwater foundation as well as subsea and pipeline issues, based on an operator-testing contractor partnership, rather than based on a JIP (Gaudin et al., 2010a). The timeliness in which results could be obtained was often a major reason why centrifuge methods were selected vs. alternative techniques such as 1g testing. This will be developed further in section 5 and 6.

3

PERCEPTION OF ADVANTAGES AND DISADVANTAGES OF CENTRIFUGE MODELLING AND TESTING

The advantages and disadvantages of centrifuge modelling are well known, and have been listed and commented in many publications, including Murff (1996) for offshore related problems.

However, centrifuge testing and modelling is an advancing science and the advantages and disadvantages are continuously changing with the problems they are investigating. Hence, some previously perceived disadvantages are disappearing due to progress made in centrifuge modelling theory and techniques, while others are appearing as centrifuge methods are applied to new geotechnical problems. This section presents advantages and disadvantages of centrifuge methods, as they relate to offshore geotechnics today. 3.1 Advantages The main advantages are as follows: 1. Centrifuge modelling properly simulates the prototype body forces and, thereby, maintains proper relationships with stresses and strains. 2. Because actual model tests are small, tests can be performed in a relatively short amount of time and at reduced cost, allowing the investigation of a wide range of parameters. 3. The field in-situ conditions are more easily replicated than is usually the case for 1-g model tests. Advancements in in-situ devices used to measure soil properties in the centrifuge, such as the piezocone and T-bar (Stewart and Randolph, 1994) have significantly enhanced the knowledge of the material being tested. 4. Pore pressures dissipate much more rapidly than in the prototype condition. This is clearly an advantage when preparing a model for testing, since it reduces consolidation time, and for testing, as cyclic sequences over several years can be modelled within hours in the centrifuge. For monotonic loading in clays, undrained conditions in clays can usually be maintained by loading to failure at a higher rate and correcting for strain rate effects. 5. Centrifuge test data can be used to validate numerical models. This method is very efficient since the soil characteristics, the boundary conditions and the applied loads are known. This is not always the case in full scale field testing. 3.2

Disadvantages are more difficult to list, as it is necessary to separate the perceived disadvantages resulting from a somewhat limited knowledge of the physical modelling principles and the real disadvantages, which the centrifuge modelling community clearly identifies (Garnier et al., 2007), and for which mitigation or correction measures may be applied. The following presents a list of the disadvantages and the measures to undertake to avoid or mitigate them: 1. Scale effects (different from size effects) were for a long time listed as the main disadvantage of centrifuge methods, supposedly preventing extrapolation of centrifuge modelling and testing outcomes to field problems. As the size of the particles may


Disadvantages

not be negligible when compared to the dimensions of the model, some particular soil behaviour may be generated, that is different to that expected in-situ. This is a scale effect, first identified by Ovesen (1975) who observed different responses for footing of various widths loaded on sand. Ovesen concluded that, for flat footing resting on sand, the width of the footing should be higher than 30 times the mean diameter of the particles. This prevents any scale effects to occur (in this case, an increase of the prototype bearing capacity with a reduction of the footing width). Indeed, the identification of a scale effect as the ratio of a relevant model dimensions to the particle size indicate that this type of problem is more likely to be critical for sand than for clay. Over the years, various research has been undertaken to quantify the minimum model dimension to particle size ratio for a variety of problems. This includes a significant number of bearing related geotechnical problems and, maybe more importantly, interface shearing problems where the scaling of the interface roughness may lead to significant difference in behaviour between the model and the prototype (Garnier & Köenig, 1998). The outcomes of all this work and some significant progress made in establishing similitude principles (notably through dimensional analysis) have been gathered in a scaling law catalogue published by Garnier et al. (2007). This catalogue provides modellers with a set of rules to design valid modelling and testing, free of potential scale effects, or to account for them when extrapolating the results to prototype conditions if they cannot be avoided (e.g. increased shearing forces due to scale effect can be calculated following procedure developed by Lehane et al. (2005). In practice, as already pointed out by Murff (1996), scale effects are rarely an issue in offshore geotechnics, especially as marine soils are mostly constituted of small particles (clay and silt in majority). If scale effects are really a major concern (for instance, if prefailure displacements in sand need to be accurately scaled, Palmer et al., 2003), modelling of models (testing different size models at different accelerations to model the same prototype) is a relevant technique to validate the modelling outcomes. 2. Creep is a complex phenomenon which may not be modelled correctly in the centrifuge. If creep relates to volumetric compaction (e.g., the displacement of a flat footing subjected to sustained vertical loading), the rate of creep is governed by the rate water escapes or moves into a volume of soil. It is therefore scaled by n2 , similarly to consolidation. Consequently, any creep phenomenon would occur n2 time faster in the centrifuge and would be captured by the model within the time frame of the experiment. If creep relates to shear deformations (e.g. the displacement of a pile subjected to sustained tensile loading), the rate of creep is governed by the viscous resistance of the soil structure and displacements in the centrifuge would occur at

the same rate as that in the field (as the strain rate would be equivalent). This result has, for example, been observed in centrifuge loading tests of shallow footing resting on dry sand (Canepa et al., 1988). Consequently, creep phenomena related to shear strains are difficult to capture in a centrifuge model within the usual time frame of an experiment. 3. Inertia effects scale differently than body forces. Therefore, for earthquake problems the loading frequencies need to be significantly increased. Even with these enhanced frequencies the pore pressure response is much faster than the prototype case in cohesionless sandy soils and a more viscous pore fluid is required to properly model this response. This is now a commonly used technique as demonstrated in section 4.4. 4. For problems where shear strain localization issues are important, such as ice gouging of the seafloor, centrifuge test interpretation of discontinuities versus shear zones is important. The use of physical modelling of models and complementary numerical modelling has been found to be beneficial in correctly interpreting the observed behaviour (Phillips et al., 2010). 5. The acceleration field across the models are variable and dependent on the radius from the centre of rotation to any point on the model. Corrections are available to account for this effect. Although some of the potential problems can be overcome or mitigated, as described above, the different scaling relationships for body forces, pore pressures viscosity and inertia can make testing more complex and difficult to perform. It relies then on the experience of the modeller to clearly establish the limits of the validity of the model. 3.3

An implicit advantage of centrifuge testing is that similitude relationships are almost always directly addressed. Conditions where similitude is or is not achieved are more carefully considered. Therefore, the advantages and limitations of the technique are better defined. In contrast, 1-g model tests have significant similitude issues, requiring the modification of the analytical models used, to match the experimental model test data. However, if the experimental model tests do not properly account for all the key factors that influence the prototype behaviour, then the potential for errors when the analytical models are extended to the prototype condition is significant. A potential useful non-geotechnical analogy to illustrate the importance of considering similitude is the model testing performed for deepwater production facilities such as TLPs, SPARs and FPSOs. For almost any large scale offshore project that utilizes one of these facilities, extensive tests are performed in model test basins (Newman, 1977). The modellers and engineers who perform these tests are always aware of the key model parameters, such as the Froude number and


Consideration of similitude

Reynolds number when interpreting results. Because all the key similitude parameters for their problem are usually not satisfied (Lyle Finn, personal communication), the model tests are augmented with extensive field measurements to further enhance measurements used to calibrate analytical and design models. Perhaps regrettably, comparable field measurements are not often performed on offshore foundations. Therefore, there is a greater reliance on the model test data for the foundation design procedures and a greater demand on centrifuge modellers to provide modelling validation. The laws of similitude need to be carefully considered when performing model tests. To assist modellers in validating their model tests, the centrifuge community has been working extensively to produce a scaling law catalogue, listing scale factors and dimensional analysis procedures for a wide range of geotechnical problems (Garnier et al., 2007). A particularly relevant example is the dimensional analysis carried out by Boylan et al. (2010a) to validate the centrifuge modelling of the behaviour of the runout of submarine slides. While 1g laboratory are commonly performed to investigate runout behaviour, the modelling of a phenomena which relies on both soil mechanics and fluid mechanics necessitated detailed similitude analyses. 4

CENTRIFUGE TECHNOLOGY DEVELOPMENT

Over the last 30 years, the progress made in miniaturised electronics, micro-computing, software engineering and digital imagery have been incorporated almost immediately in centrifuge technology. It results in more realistic simulations, more accurate measurements and more detailed observations, subsequently improving the benefit centrifuge modelling can yield to the geotechnical community. This is particularly relevant for offshore geotechnics where complex environmental conditions, emphasis in capacity and saturated soil conditions prevail. Hence, sophisticated motion control to apply environmental cyclic loadings, image acquisition techniques to observe failure mechanisms and enhanced data acquisition systems to measure accurately pore pressure developments and fast responses have been key developments in centrifuge modelling techniques. Some of the most noticeable developments recently undertaken in centrifuge technology are presented below, with examples highlighting the benefits obtained from these improvements presented in sections 5 and 6. 4.1

Data acquisition system

In the early days of centrifuge testing data acquisition was limited to post-testing measurements and observations. With the development of computing and electronic, on-board data acquisition systems appeared. A typical data acquisition system would

Figure 2. New high-speed data logging unit developed at COFS. The unit replaces standard PC-data acquisition card combination (after Gaudin et al., 2009a).

include transducers, a signal amplifier and an analog/digital converter system in a slave computer onboard of the centrifuge (Liu et al., 1988; Garnier & Cottineau, 1988). The signal would be transmitted through electrical slip rings to a master computer inside the control command room. In such a system, electrical slip rings constitute the weak point (Taylor, 1995). They are prone to failure and generate significant electrical and mechanical noise. Over the years, improvements in sampling, signal resolutions and signal conditioning were undertaken, with 128 channels sampled at 100 Hz with a resolution of 16-bits being the norm. The use of either optical slip rings or wireless transmission also contributed to improve the quality to the data acquired. Nevertheless, the overall architecture remained unchanged as one can see in the development of the newest centrifuge facilities (Ng et al., 2006). Recently, some significant breakthroughs in data acquisition technology have been made at COFS. An independent self-powered unit 150x60x40 mm in size, embarking processing and storage units and capable to monitor up to 8 transducers, has been developed for centrifuge application (Gaudin et al., 2009a). Each unit (see Figure 2) is capable of powering and monitoring eight instrument channels at a sampling rate of up to 1 MHz at 16-bit resolution. The data are stored within the logging unit in solid-state memory, but may also be streamed in real-time at low frequency (up to 10 Hz) to the centrifuge control room, via wireless transmission. Unlike PC-based data acquisition solutions, this system performs the full sequence of amplification, conditioning, digitization and storage on a single circuit board via an independent micro-controller allocated to each pair of instrumented channels. It results in a much cleaner signal and the faster sampling rate, coupled with automatic triggering features, proves to be extremely useful in acquiring data during cyclic events or fast events such as submarine landslide (Boylan et al., 2009a). 4.2

Actuation devices have followed an improving trend similar to data acquisition systems. Early devices


Robotic control

such as hydraulic mono-axis actuators developed into bi-axis electric or hydraulic actuators, to eventually multi-axis robots, first in LCPC (Derkx et al., 1998) and then in HKUST (Ng et al., 2002). In parallel, the development of sophisticated motion control systems occurred (De Catania et al., 2010). Complex loading sequences, including varying control modes on various axes are now possible, mimicking more realistically typical offshore loadings and permitting the study of particular aspects of soil structure interaction. One example is the investigation of pipeline embedment during laying process, when the pipe is subjected to a combination of horizontal and vertical cyclic motion with varying amplitude (Gaudin & White, 2009). This example is developed in more detail in section 6. The last ten years have also witnessed the development of actuation devices to generate wave loadings and wave-induced phenomenon. The device commonly includes a wave paddle (Sassa & Sekigushi, 1999) or a plunger moved by a horizontal (Baba et al., 2002). In both cases, frequency and amplitude of the motion of the wave generator device may be controlled in order to generate the appropriate wave spectrum. Particular attention is paid to limit or control the wave refraction against the edges of the centrifuge container and the use of fluids with viscosity higher than water to satisfy the similarity rule with regard to laminar flow (e.g. conservation of the Reynolds number). The drum centrifuge, which features a high developed length, was initially used for such tests (Phillips & Sekiguchi, 1992) and is particularly appropriate to model waves (Gao & Randolph, 2005). The steel catenary riser tests at C-CORE described in section 5.7 utilise 2 synchronised servo-hydraulic actuators to provide the necessary broad band frequency surge and heave motions (Figure 3). The actuators are counter balanced to minimise driving forces to maximise the system frequency response. 4.3

Digital imagery

In the early ages of centrifuge modelling, computer image processing techniques were used in combination with digital cameras (Garnier et al., 1991). The subsequent development of the particle image velocimetry (PIV) and photogrammetry techniques, for soil mechanics, by White et al. (2003) have constituted a major breakthrough for centrifuge technology, expanding considerably the range of investigations. More detailed observations can be gathered and insight into failure mechanisms taking place around foundations can be obtained. This is particularly relevant for offshore foundations where failure often governs the design, as opposed to onshore foundations where designs are governed more by displacements or serviceability. The technique consists in processing digital images of the soil and structure investigated, placed against a transparent observation window (see Figure 4), in order to obtain information about the displacements of the soil and the structure (White et al., 2005).

Figure 3. Servo hydraulic actuators developed at C-CORE.

The use of the PIV technique has been pivotal in a significant number of breakthroughs on the behaviour and performance of offshore structures. One may highlight: 1. The identification of the back flow mechanism, as opposed to the wall collapse mechanism previously assumed, governing the limiting cavity depth during spudcan penetration (Hossain et al., 2005). 2. The identification of the bearing capacity mechanism for spudcans penetrating sand overlaying clay, which showed the changes in overall failure mechanism due to varying geometric and strength conditions of the layered soil (Teh et al., 2008). 3. The identification of the spudcan extraction mechanism and the resulting insight provided about the development of suction at the spudcan invert (Purwana et al., 2006) or the identification of the ground movement generated by spudcan penetration leading to indirect loading of nearby piles (Leung et al., 2008). 4. The identification of the berm creation mechanism and its influence on the soil resistance during lateral pipeline motion (Dingle et al., 2008). 5. The identification of a Hill-type reverse end bearing mechanism during uplift of skirted foundation, due to the development of suction at the foundation invert (Mana et al., 2010).


Other pivotal examples include the identification of the strain path followed by a soil particles nearby the tip of a pile which contributed to the development of theoretical solutions for the assessment of pile

Figure 4. Typical PIV setup in the COFS drum centrifuge (after White et al., 2005).

Figure 5. Soil displacement vectors for skirted foundation with embedment ratio = 0.2 subjected to uplift (Mana et al., 2010).

base resistance (White & Bolton, 2004) and the understanding of the phenomenon of friction fatigue which governs the shaft resistance along piles driven in sand (White & Bolton, 2001).

The PIV technique may provide insights, not only on the soil displacement pattern and the failure mechanism taking place, but may also reveal the particular behaviour of the structure investigated. A


typical example of such insight of offshore structures behaviour is the identification and the understanding of the mechanism governing the behaviour of keying flap of suction embedded plate anchors. Recent experiments clearly demonstrated the inefficiency of the keying flap in reducing the loss of embedment during keying (Gaudin et al., 2010b). All these examples highlight the dramatic improvements in the utility of physical modelling in providing important insights in the mechanisms and behaviours which can subsequently be used to develop, improve or validate design and calculation procedures. 4.4

Soil sample reconstitution and characterisation

Improvements in soils sample reconstitution results more from the experience gathered by modellers and from the better understanding of soil behaviour, than from technological developments. Automatic air pluviation techniques to reconstitute sand samples and in-flight consolidation of normally consolidated clay samples or in-press consolidation of over consolidated clay samples are now well established techniques in every centrifuge centre, resulting in homogeneous and well controlled soil samples (Garnier, 2002). Nevertheless, recent projects have seen a significant increase in sophistication of the soil samples reconstituted, with natural soils sourced in-situ (as opposed to commercially available clay and sand) becoming more common, both in full-size samples, or in small boxes for PIV analysis. This has become common practice at, for example, LCPC (natural clay from North Sea and from West Africa) and at COFS (carbonate sediments from North West Shelf, soft clay from West Africa). The reconstitution of multi layered soils (typically a sand layer within a clay seabed) and the use of alternative pore fluid in granular material (such as silicon oil or aqueous methyl cellulose) to increase the pore fluid viscosity in order to generate and observe liquefaction phenomenon is also becoming common practice. A typical example is the reconstitution of a sample featuring a kaolin clay layer saturated with water overlaying a silica sand layer saturated with silicon oil. This technique has been used to replicate in-situ conditions offshore Hong Kong, where suction caissons to found wind turbine tripod will be installed (Gaudin & Randolph, 2008). 100 cst silicon oil (e.g. 100 times more viscous than water) was used as the pore fluid to model correctly the pore pressure regime resulting from environmental cyclic behaviour whose frequency could not be scaled correctly by a factor of 100. The silicon oil, therefore allowed proper scaling of the pore fluid. Unsaturated conditions may also be simulated in centrifuge models as demonstrated during the European project NECER (Rezzoug et al., 2000). In parallel to the increasing sophistication of reconstitution techniques using natural soils, alternative materials have been developed, replicating the geotechnical properties of natural soil, with the aim to either simplify and increase the reliability of the

reconstitution techniques or provide particular features to maximise testing outcomes. Artificial soil made from mineral ingredients, high boiling liquid and solvent and exhibiting shearing behaviour identical to soil, has been developed in order to overcome the existing limitations of reconstitution techniques (Sarma, 2006). However, the very good control of the reconstitution process of natural and kaolin clays and the improvement in soil characterisation techniques have dramatically limited the need of such artificial soil. Another artificial material which is seeing much use is the “transparent soil” (Iskander et al., 2001). By mixing flumed silica with paraffin oil and white spirit, one can obtain a transparent slurry, which upon consolidation, exhibits a strength profile similar to clay materials. The process permits the visual observation, through a window, of the offshore structure investigated. This is similar to the PIV technique, but it also permits accurate measurements of the loads applied, which is not possible with the PIV technique due to the contact of the structure with the observation window. An example offshore application using observation, through transparent soil, has been the investigation of the behaviour and trajectory of an embedded plate anchor (Figure 6) during keying and pullout (Song et al., 2009). Improved in-flight soil characterisation techniques may also be partly responsible for the improvement in sample reconstitution techniques. Reduced scale versions of cone penetrometer, piezocone and vane apparatus, identical to the in-situ tools, were used successfully since the early day of centrifuge testing to characterise soils. The last two decades have, however, seen the development of new characterisation tools in the centrifuge. This has proved to provide more accurate and more comprehensive information about the soil properties. This is notably the case of the T-bar penetrometer (Stewart & Randolph, 1994), initially developed for the centrifuge, which is now widely used offshore. The T-bar provides an accurate measurement of the soil undrained shear strength and offers a valuable alternative to cone penetrometer devices. The latter requires various corrections and have a disadvantage of a low ratio of the change in resistance to the ambient hydrostatic pressure acting on the cone tip (Randolph et al., 2005). In addition, more information can be extracted from T-bar results as illustrated by recent research. By performing successive cyclic sequences, one may access the strength degradation rate, the soil sensitivity and the post reconsolidation strength (Hodder, et al., 2010, Zhou & Randolph, 2009). By performing tests at various penetration rates, one may access information about the consolidation characteristics of the soil and the strength enhancement due to viscous effects (Chung et al., 2006, Lehane et al., 2009). More recently, a new analysis technique, based on the T-bar, has been developed which allows the accurate determination of the shear strength in the first meter of the


Figure 7. O-bar developed at C-CORE.

Figure 6. Anchor model in transparent soil (after Song et al., 2009).

seabed, accounting for buoyancy of the apparatus and surface effects (White et al., 2010). This analysis technique is particularly relevant to assess the soils shear strength for soil-pipeline interaction. This is developed further in section 6.3. C-CORE has recently developed an O-bar for the assessment of near surface strength by reducing the bar diameter to 2.4 mm while increasing its length to 110 mm using the form of an annular ring (Figure 7). Other full flow penetrometer such as the ball penetrometer (Kelleher & Randolph, 2005) and more recently the piezoball penetrometer (Boylan et al., 2010b) have also been developed and have started to be introduced offshore (Figure 8). The centrifuge environment, with perfect control of the testing conditions, is the appropriate environment to develop and validate new soil characterisation tools. Near future will surely see new devices arising, such as the “doughnut” shearing device developed to measure peak and post peak pipe-soil friction factors and the associated pore pressures (Yan et al., 2010). Soil containers have also benefited from some technological developments. Thermally controlled strongboxes (Phillips et al., 2002) and ice generators (Lau et al., 2002) have been introduced to investigate the behaviour of offshore structures in artic regions. These

Figure 8. Piezoball penetrometer developed at COFS to investigate strength and consolidation properties of soft soils (after Boylan et al., 2010b).

are based on the similitude principles for soil freezing established by Yang & Goodings (1998). 4.5


Models and instrumentation

The general improvement in instrumentation and sensing technology has indeed benefited centrifuge technology. Laser sensors offering high resolution are progressively replacing linear displacement transducers. The miniaturisation of commercially available load cells, pore pressure and total pressure transducers results in more accurate and more numerous measurements. This is giving access to information pivotal to understand detailed phenomena. One relevant example is the use of contact stress transducers to measure the radial stress changes on the outside of the 0.4 mm thick skirt of a suction caisson during installation, consolidation and subsequent extraction (Chen & Randolph, 2004). Measurements provided direct insight onto the skin friction of the caisson skirt, which was previously back calculated from penetration and extraction resistance. Jeanjean et al. (2006) used a double walled suction caisson to separate internal and external skin friction by nesting two stainless steel tubes and measuring the

Figure 10. Example of model suction embedded plate anchor (30 mm high, 53 mm wide) replicating all key features of prototype anchors, including keying flap and flap hinge eccentricity (from COFS internal report).

Figure 9. “Stroud” type load cell with protective sleeve removed (after Daiyan et al., 2010).

loads applied both to the inner wall and on the total system. There was a 0.4 mm gap between the outside wall and the inside wall of the anchor. Specific instrumentation for centrifuge applications has also been developed. For instance, Take & Bolton (2002) and Oung & Bezuijen (2002) have developed pore pressure tensiometers which overcome some of the limitations of the commercially available pore pressure transducer Druck PDCR-81. Daiyan et al., (2010) have used a miniature sealed ‘Stroud’ type load cell to measure the axial, transverse and moment loads transmitted through a pipeline subject to oblique loading in sands and clays (Figure 9). The use of bender element to measure in-flight shear wave velocity of a soil sample is also becoming regular practice (Brandenberg et al., 2006). The manufacturing of models has also improved significantly, with more precise foundation and anchor models now commonly available. An example of this improvement relates to drag and plate anchor models which feature particular details such as shank and fluke geometry and keying flap. The use of devices to replicate suction installation of caissons (Clukey & Phillips, 2002, Chen & Randolph, 2007b, Raines et al., 2005, Jeanjean et al., 2006, Colliat et al., 2010) or suction embedded plate anchors (Gaudin et al., 2006c), has also become common practice (Figure 10).

5 THE CONTRIBUTION OF CENTRIFUGE TESTING/MODELLING The contribution of centrifuge testing and modelling to the design of offshore structures and the understanding of offshore structure soil interaction has been addressed successively by Murff (1996), Garnier (2004) and Gaudin et al. (2006a). With the continuous development of new centrifuge technology and the great state of flux exhibited by

the offshore industry, continuously moving towards deeper waters and unusual soil conditions, it is anticipated that new areas will emerged in the near future. While the offshore industry will most likely use more and more centrifuge modelling techniques and benefits to its outcomes, it is as important to appreciate that the offshore industry will contribute significantly to the development of new modelling and testing techniques. The following provides a review and some additional examples of where centrifuge testing has enhanced offshore foundation engineering. 5.1

As opposed to field tests, centrifuge modelling provides homogeneous and well characterised soil conditions, known boundary conditions, accurate measurements of parameters and repeatable testing conditions. Therefore centrifuge testing offers reliable performance data for a given idealised problem which can be used to calibrate analytical and numerical models. These can subsequently be applied to specific field problems, which by nature are more complex. Some relevant examples, outside those already presented by Murff (1996), Martin (2001) and Gaudin et al. (2006a), include: 1. The calibration of numerical model for the design of the foundations of the Rion-Antirion bridge in Greece (Garnier & Pecker, 1999), a novel type of foundation including a sand raft resting on pilereinforced marine clay and subjected to seismic loading in addition to the more common environmental loadings. 2. The calibration of the geotechnical model used to design the pier foundation of the Confederation bridge in Canada (Phillips et al., 1998). The Confederation Bridge main pier foundation design


Provide performance data to calibrate analytical or numerical models

was carried out using the LRFD approach and a probabilistic assessment (Becker et al., 1998). A reliable assessment of bearing resistance requires that the failure mechanism that is likely to develop in a foundation material be known and understood. There is limited field data regarding the performance of structures founded on ring footings, particularly in an offshore environment where high horizontal loads and moments are generated. Therefore, centrifuge model tests of the main pier foundations were carried out to investigate potential mechanisms of failure, to examine how these mechanisms vary with changes in foundation and loading conditions, and to evaluate the foundation design analysis carried out as described above. This testing also helped to reduce some of the inherent uncertainties associated with the analyses, such as limitations in constitutive models, and to improve the fundamental understanding of the geotechnical aspects of foundation design under high horizontal loads and associated eccentricities. 3. The calibration of the numerical model developed to design the sloping sea wall of the pier foundations of the Gwang Yang bridge in Korea against the ship impacts (Gaudin & Colwill, 2007). The Gwang Yang bridge is a 1.28 km long suspended bridge in South Korea. Its piers are funded on a 60.5 m long by 36 m wide raft resting on rubble, soft marine clay and sand compaction piles. In turn, the raft is supported by a network of 38 piles, 32.6 m long, embedded in the rock bed. The piers are protected against potential ship impact by a sloping sea wall, whose geometry was initially designed from empirical analytical models. In order to calibrate the numerical model used to finalise and validate the design, centrifuge tests were performed on an idealised model, where the rubble, soft marine clay and sand compaction piles were replaced by loose sand. Other features of the prototype such as the embedment of the piles into the rock bed through a layer of soft marine clay and the geometry and velocity of the bow and energy realised by the ship during impact were accurately replicated. The model was heavily instrumented and provided performance data such as the total pressure and the pore pressure in the front and at the invert of the raft, displacements and acceleration of the raft, and 3D profiling of the geometry of the sloping sea wall after impact. These data were used to calibrate a finite element model, which was subsequently improved to account for the properties of the rubble and sand compaction piles. The FE model was then used for design (Figure 11). 5.2 Providing qualitative insights into soil-structure interaction and mechanisms This aspect is particularly important when novel concepts or unusual conditions are encountered. Understanding the structure behaviour, observing the failure mechanism taking place, is a pivotal step into the

Figure 11. Model of the GwangYang bridge pier foundation and protecting seawall. (a) Footprint left by the impact of a ship at 30 degrees, and (b) comparison with the numerical model (b)(after Gaudin & Colwill, 2007).

development of sound design methodology. Some of the most noticeable breakthrough includes: 1. The identification of the shallow failure mechanism for gravity platform under horizontal and vertical cycling loading in comparison to the much deeper mechanism resulting from monotonic loading (Craig & Al-Sauodi, 1981). 2. The identification of the flow failure (Figure 12) taking place during spudcan penetration, as opposed to a cavity wall failure (Hossain et al., 2005). This has resulted in the amendment of the SNAME (2008) guidelines. 3. The observation during pipeline lateral breakout of the growth of a soil berm in front of the pipe and of the suction developed at the rear of the pipe, governing both the trajectory of the pipe and the peak breakout resistance (Dingle et al., 2008). This aspect is developed in more details in section 6.5. 4. The observation of the behaviour of the keying flap equipping suction embedded plate anchors. The flap was designed to rotate during keying and hence limit the loss of embedment. Centrifuge results demonstrated that the flap does not rotate (Figure 13) due to the rotational mechanism of the soil, resulting in the soil applying a bearing pressure at the back of the flap (Gaudin et al., 2010b).


Figure 13. Plate anchor pulled out vertically. Demonstration of the non-activation of the keying flap during anchor keying (after Gaudin et al., 2010b).

with full-scale measurements to provide confidence to the end-users that this is a viable approach.

5.3 Validate the structure design for a particular site or a specific design approach By allowing the use of natural soils, sampled in-situ, by applying complex loading sequences directly relevant to design and by using a models replicating precisely prototypes, centrifuge modelling offers a valuable tool to validate or justify specific offshore structures. Among the relevant examples are:

Figure 12. Observation of the back flow mechanism during spudcan penetration (after Hossain et al., 2005).

Centrifuge modelling has also provided an efficient and cost-effective complement to conventional ice tank modelling to simulate the deformation of level ice and rubble. Lau et al. (2002) presented an overview of the ice-structure interaction research conducted at C-CORE. They discuss the advantages of centrifuge modelling of ice problems in terms of the scalability of ice properties, the controllability of test environments, and the quality of test data. There was an excellent agreement between data obtained from the centrifuge, an analytical algorithm and other conventional refrigerated model basins for a range of ice conditions. These included level ice, ridges and rubbles interacting with structures including conical piers with level ice sheets; unconsolidated ice rubble and ridges with cylindrical structures; and consolidated rubble ice with conical structures. Due to the small size of the model required, centrifuge modelling also has the advantage of simulating and maintaining a controlled test environment at a fraction of the cost of conventional tanks. Centrifuge modelling is relatively new to the ice engineering community and additional comparisons are still needed

1. The investigation of the response under cyclic horizontal loading of the Maari platform in New Zealand. Operated by OMV, the Maari platform is founded on a rectangular skirted mat (Figure 14a), installed in sand overlaid by impermeable silt (Gaudin et al., 2006b). Tests were used to investigate both the feasibility of suction installation through layered soil, and the performance of the platform under cyclic horizontal loading. Results demonstrated that suction installation was achievable, with limited silt plug uplift, provided that sufficient direct load is applied and the pumping rate is high enough to result in a fast installation, limiting the amount of plug uplift. Similar observations were made on another offshore project in Gulf of Mexico where suction installation on sand overlaid by clay was required (Watson et al., 2006). Results for the Maari platform, under horizontal cyclic loading highlighted the very stiff response of the top silt layer and the very limited vertical displacements experienced by the platform, even after several packages of cyclic loading (Figure 14b). 2. The investigation of the performance of a dolphin made of sheet pile walls for the protection of the Korena Incheon bridge foundation against ships


Figure 14. (a) The model Maari platform and (b) Accumulation of vertical displacements under cyclic horizontal loading (after Gaudin et al., 2006b).

impact (Bezuijen et al., 2007). A total of 31 tests were performed at 200g investigating the response of the dolphin under static and dynamic tests loadings. Other parameters investigated included the density of the soil, the diameter of the dolphin and the location of impact. The results show that for dynamic tests (as opposed to static tests) suction was generated underneath the dolphin increasing its stability. 3. The validation of design of the Genesis Spar platform installed in the Gulf of Mexico, and operated by Chevron. The 70 m long piles anchoring the platform were modelled in a series of centrifuge tests performed at LCPC to investigate notably the performance of the piles under combined vertical and horizontal loading, and cyclic storm loading. The bending moment profiles (see Figure 15a) along the piles were monitored to optimise the pile thickness and to develop p-y reaction curves (see Figure 15b) these were subsequently used to validate the pile design.

Figure 15. (a) Bending moment (MN.m) vs. depth (m) recorded during the Genesis pile loading tests ranging from 10 to 23 MN, and (b) experimental P-y reaction curves derived from the bending moment profiles (depth ranging from 21 m to 58.8 m). (From internal LCPC report).

5.4

The very good control of the testing conditions, the possibility of measuring a significant number of parameters, such as displacement, loads, pressures and even strains from digital imagery and the relatively low cost of centrifuge testing compared to field testing (even in reduced scale) make the centrifuge an ideal tool to develop new concepts and investigate the feasibility of particular foundations, which have not been tested in-situ yet. The modelling does not intend to replicate precisely a prototype structure, but rather puts emphasis on a particular aspect which requires validation or deeper understanding. Clukey & Morrison (1993) investigated the efficiency of multi-cell suction caissons as anchoring system for tension leg platforms (Figure


Investigating the feasibility or developing new foundation concept

Figure 16. Model multi-cell suction caisson (source LCPC).

16). Tests were performed on kaolin clay and particular attention was paid on the potential for suction to be developed at the caisson lid invert to resist monotonic and cyclic tensile loads. Gaudin & Cassidy (2007) investigated the feasibility of suction-induced preloading of a skirted mat as an alternative to spudcans as a foundation for jack-up platform in intermediate water depth, where ballast tank may not provide enough preloading. Tests did not model any particular type of foundations, but focused on the preload level which could be achieved by suction and the drainage conditions within the foundation. Results demonstrated the feasibility of the concept but also the mechanism governing suction-induced preloading. It was notably observed, the absence of drained plug failure during the application of suction-induced preload, and the dominant effect of the consolidation time versus the preload level on the post preloading bearing capacity of the foundation (Figure 17). Another typical example of a novel design that has greatly benefited from centrifuge methods is the development of dynamic anchors, commonly called torpedo anchors. Torpedo anchors are usually 1 to 1.2 m in diameter and 10 to 15 m high. They are released from a height of 50 to 100 m above the seabed, achieving a free fall a velocity of 10 to 30 m/s, before impacting soft seabed sediments and embedding by 2 to 3 times their length. Key uncertainties relate to (i) the embedment depth resulting from the dynamic penetration and (ii) the soil setup after installation and the resulting holding capacity. Centrifuge techniques were particularly appropriate to conduct model tests, using the high g environment to achieve a fast free fall velocity from limited drop height. They played a significant role in both providing key insight about the soil-anchor interaction and establishing techniques to predict anchor embedment and anchor capacity. Early centrifuge work, reported notably by O’Loughlin et al. (2004) and Richardson et al. (2006) focused on anchor installation and highlighted the effect of the very high strain rate generated by the

Figure 17. (a) Hybrid foundation model with the suction caisson in the centre and the skirted mat around the caisson (b) Increase of bearing capacity due to suction induced preloading (after Gaudin & Cassidy, 2007).

penetration high velocity. This results in an increase in shear strength due to viscous effect, potential entrainment of water at the soil anchor interface and a decrease of skin friction. The necessity to account for a drag factor, in addition to the soil bearing factor, in calculating the anchor embedment depth was also demonstrated. More recent work, reported by Richardson et al. (2009) focused on the soil strength recovery after installation and the subsequent holding capacity (Figure 18). Results demonstrated the significant increase of capacity with time (by a factor of 5 from immediate extraction to full consolidation) and the significant time required to achieve full consolidation, compare to piles or suction caisson (up to 7 years in the case presented here). 5.5

As highlighted by previous examples, the range of benefits from centrifuge application to offshore


Developing new design methods

the pile head lateral displacement may be estimated by:

Figure 18. (a) Typical model torpedo anchors, without fluke and with four flukes (b) Illustration of the gain in capacity with consolidation time after penetration. QS refers to static installation, DY to dynamic installation (after Richardson et al., 2009).

geotechnics offers the opportunity of providing valuable information at all stages of foundation design. Centrifuge modelling can also be an appropriate means to develop empirical design methods for some particular cases, where the phenomenon taking place are too complex to be captured by numerical or analytical models. One example is the development of a new design method for piles under lateral cyclic loading, which was poorly accounted for in common design guidelines. Centrifuge tests have been carried out by Rosquöet et al. (2007) and Rakotonindriana (2009), where up to 75000 lateral cycles were applied to model piles installed in dry sand. By normalising these cyclic loads by the maximum applied load Fmax and by introducing the amplitude of the load variation DF, Rosquoet et al., 2007 demonstrated that, for service conditions, the effect of the number of cycles n on

where y1 and yn are the pile head displacement under the maximum load Fmax , at respectively the first loading and at the nth cycle. Cyclic p-y curves were determined from bending moment measurements during loading (Figure 19). The degradation of the soil lateral resistance at different depths was then evaluated and P-multiplier coefficients r were established. These permits the determination of the cyclic p-y curves from the static ones (Table 1). Another typical example is the empirical method developed by Bienen et al. (2009) to determine the flow rate required to successfully extract spudcans in soft clay using on-bottom jetting (Figure 20a). The method was based on insights provided by a series of centrifuge tests, investigating various parameters such as the extraction rate, the jetting flow rate and the jetting pressure (Bienen et al., 2009; Gaudin et al., 2010c) and the demonstration that the efficiency of the jetting related to its capability to fill the gap at the spudcan invert created by its uplift. Hence, jetting flow was shown to be a dominant factor over jetting pressure, in contrast to generally accepted belief in the jack-up industry. A jetting methodology was then established based on (i) the pullout force resulting from the buoyancy of the jack-up hull, and a filling ratio f, defined as the ratio of the jetting flow rate to the product of the extraction rate to the spudcan area (i.e., the volume at the spudcan invert created by the extraction) (Figure 20b). The Pressure Ridge Ice Scour Experiment (PRISE) developed the capability to design pipelines and other seabed installations in regions gouged by ice, taking into account the soil deformations and stress changes, which may be caused during a gouge event (Phillips et al., 2005). This capability has been used to design offshore pipelines in regions such as the US Beaufort Sea and off Sakhalin Island, Russia. One of the main activities for the PRISE program was to assess the significance of subgouge deformations imposed on buried pipelines during ice gouge events. The PRISE joint industry research program conducted analytical, numerical, experimental and field studies. Comparison of field investigations on relict gouge events with centrifuge experiments demonstrated that subgouge deformations should be considered. Analytical and numerical models were developed to assess the gouge forces in sands and clays, based on the experimental failure mechanism observations. The implications of the models on ice keel ablation, spoil heap development and the development of subgouge deformations are discussed by Phillips et al. (2005). Eulerian based finite element models for clay have also been validated against the centrifuge model test results (Phillips et al., 2010).


Figure 19. Cyclic p-y curves at different depths obtained from one-way lateral loading tests (after Rosquoet et al., 2007).

Table 1. P-multipliers r at different depths for determining the cyclic p-y reaction curves (after Rosquoet et al., 2007). Depth z

P-multiplier r

0 < z < 1.5B 1.5B < z < 3B 3B < z < 5B

r = 0.7 − 0.12 DF/Fmax r = 0.94 − 0.058 DF/Fmax r = 0.97 − 0.029 DF/Fmax

5.6

5.7

Characterising in-situ soils

The improvement of soils reconstitution methods allows the replication of complex soil stratigraphy using natural soils exhibiting the same properties (such as permeability, compressibility, void ratio, water content, unit weight and shear strength) as found in the field (see development in section 1.3). The centrifuge offers the possibility of performing a large number of tests at a reduced cost, using characterisation tools identical to the ones used in-situ. Also, in addition, the techniques provides extra information by using newly developed characterisation tools not yet commonly used in-situ. One such example is the recently developed piezo-ball penetrometer presented in section 4.4, for characterisation of soft soils. The test, in addition to the measurement of the penetration resistance (and hence undrained shear strength), provides information on the drainage characteristics of the soil and potentially the consolidation behaviour (Boylan et al., 2010b). With the acquisition of large soil samples taken with box core becoming more standard practice, it


is also possible to perform typical soil characterisation tests in a relative undisturbed state (White & Gaudin, 2007). Centrifuge testing helps augment these tests. Providing data for assessment of fatigue for conductors and SCRs

Designing for fatigue of conductors (uppermost outer pipe in a well) and Steel Catenary Risers (SCRs) is an important design challenge in deepwater. The interaction of the conductor and riser with the soil is an important part of both these fatigue assessments. Jeanjean (2009) describes a case where the API p-y curves were modified based on finite element analyses and centrifuge testing to satisfy fatigue requirements for a number of offshore conductors. For this particular case the small displacement lateral soil response was critical for predicting the expected fatigue life. Figure 21 shows a comparison between theAPI recommended lateral soil response and the response derived from FE analyses and centrifuge testing. In this problem a stiffer soil response reduced fatigue life. Cyclic centrifuge tests did show some reduction in the soil stiffness. However, based on the design methods used to assess fatigue, during the unloadreload soil response appropriate to cyclic loading, the soil stiffness was never less than the tangent stiffness from the monotonic test results used in the analyses (Jeanjean, 2009). The fatigue of an SCR in the soil touchdown point region (Figure 22) depends on both small and large amplitude displacements. Hodder et al. (2008) describes a series of centrifuge tests to investigating

Figure 20. (a) Model spudcan with jetting in action and (b) conceptual chart of jetting extraction efficiency. The chart indicates, for a given extraction load, the required flow rate for a successful jetted extraction (after Gaudin et al., 2010c).

the response of a small portion of pipe in the soil touchdown point region. The tests show that the soil stiffness can be dramatically reduced by the large displacements incurred from the riser motions, especially when


the riser pipe separates from the seafloor. However, a significant portion of the stiffness loss is regained when the motions stop and the soil has an opportunity to recover from thixotropic and consolidation effects

Figure 21. Comparison of API p-y curves versus p-y curves obtained with centrifuge testing and FE analyses (after Jeanjean, 2009).

Figure 22. Schematic of offshore facility and SCR interacting with soil in the touch down area.

(Hodder et al., 2009). The large motions of the SCR will also enhance trench formation which will also affect fatigue life. Centrifuge tests to investigate the response of a complete riser through the touchdown point region are now being performed at C-CORE. A special actuator has been developed for these tests to simultaneously provide both heave and surge motions to the riser. The motions are applied to both these motions at two frequencies to capture the overall riser motions expected in the field. Comparisons of trenches observed in the field and in the centrifuge are shown on Figure 23 and Figure 24. 5.8

Figure 23. SCR trench developed in centrifuge test.

distribution of loads. One such example is the work performed by Gaudin & Landon (2008) and Gaudin et al., 2009b who investigated the efficiency of a rock cover to protect a buried offshore pipeline against the dragging of anchors. The assessment of the clearance between the anchor and the pipeline over a series of centrifuge tests resulted in the determination of frequency of failure of the pipeline which was subsequently introduced into a probabilistic analysis on the frequency of hazard affecting the pipeline. 6

Providing data for probabilistic analyses

One recognised advantage of centrifuge modelling is the ability to conduct parametric studies in a well controlled environment. It is therefore possible to focus the study on a particular parameter, such as the natural variation of soil characteristics or the statistical

6.1

Suction caisson technology development

The history of the development of suction caisson as anchoring system for deep water solutions illustrates particularly well the contribution of centrifuge


EXAMPLE OF CONTRIBUTIONS

Figure 26. Evolution of penetration resistance and external radial total stress of a suction caisson in centrifuge during jacking and suction installation (After Chen & Randolph, 2007a).

Figure 24. SCR trench in field.

Figure 25. Results of static pullout tests. Comparisons between in-situ 1g tests and centrifuge tests performed on natural (Lysaker) and artificial (kaolin) clays. (after Morrison et al., 1994).

testing and modelling in investigating behaviour and validating design procedures throughout their development. Centrifuge results have impacted installation procedure and assessment of the capacity, first under monotonic vertical loading and then under combined cyclic loading. In contrast to the development of driven piles technology for offshore structures, which was primarily based on a series of 1-g model tests, suction caisson technology has achieved significant advancements through centrifuge testing. The initial tests, however, for suction caissons used in deepwater were in-situ 1-g tests for the Snorre TLP in the North Sea. (Dyvik et al., 1993). These tests were subsequently replicated with centrifuge tests at LCPC (Morrison et al., 1994). When the undrained shear strength profile is well simulated, the load-displacement behaviour between the in-situ and centrifuge tests are very close, both in peak resistance and stiffness (Figure 25), and provided added confidence in the use of centrifuge testing for suction caisson applications.

Additional testing for TLP applications (Clukey & Morrison, 1993, Clukey et al., 1995) provided additional information on suction caisson behaviour under both monotonic and cyclic loads for TLPs, notably about the cyclic ratios resulting in failure and about the increase of post-cyclic static capacity. With ever increasing water depths, the design trends moved from multi-cell configurations systems, to single cell arrangements with the mooring line attached about 2/3 down the side of the caisson to enhance the holding capacity. The utilization of these single cell suction caissons expanded considerably as the industry moved into water depths too deep for conventional offshore pile driving. Centrifuge testing at a number of facilities then began to provide important information on the behaviour of these foundations both during installation and under a variety of loading conditions. Andersen et al. (2003) showed that suction caissons could potentially be installed to over ten times their diameter without failing the internal soil plug. Clukey & Phillips (2002) described monotonic tests which showed a potential for the reduction in external skin friction for suction installed versus jacked in caissons. Subsequent tests by Raines et al. (2005), however, showed no difference in external skin friction for suction versus jacked installed suction caissons. Similar results were obtained by Chen & Randolph (2007a) (see Figure 26) and Jeanjean et al. (2006). The latter used a double walled suction caisson to separate internal and external skin friction. Their results did show, however, a reduction in external skin friction versus API recommended values for driven piles. The reverse end bearing observed by Jeanjean et al. (2006) had a peak bearing capacity factor (Nc ) of 12 at large displacements and 9 when the external skin friction and total holding capacity reached peak values. Similar magnitudes of design parameters were obtained from research on suction caisson response carried out by Chen & Randolph (2007a). More recently, the effect of stiffeners in both the installation resistance and the pullout capacity has been investigated by Westgate et al. (2009) and Colliat


verified the effectiveness of design tools in predicting suction caisson holding capacity in stiff overconsolidated soils in the Gulf of Mexico. It is most noteworthy that the results were obtained in a time frame, at a cost and with a level of details that could not have been matched by field testing alone. 6.2

Figure 27. Effects of stiffeners and consolidation on normalized pull-out capacity (after Colliat et al., 2010).

et al. (2010) at LCPC. The former investigated various stiffener configurations in kaolin clay, while the latter, using in-situ deepwater Nigeria clay, focused on the effect of consolidation time following installation, for caissons with and without stiffeners. Results, from both studies, demonstrated the significant contribution of the stiffeners to the pullout capacity. For undrained pullout, the increase ranges from 20% shortly after installation, up to 33% when full consolidation is permitted (Figure 27). Combined loading was investigated by Randolph et al. (1998), who described tests that demonstrated the potential for gapping behind the caisson, in soft carbonate sediments, due to interactions between horizontal and vertical loadings. Clukey et al. (2003), further investigated combined loading and demonstrated that limit analysis techniques could be used to predict these interactions. Clukey et al. (2004) also described a combination of independent centrifuge test programs at UWA and C-CORE that showed that reverse end bearing could be maintained for several months without reduction in capacity from pore water pressure dissipation. These results were also further corroborated with finite element analyses (Clukey et al., 2004). These results were important in assessing the performance of suction caisson for long term loop current load. As demonstrated above, the use of centrifuge techniques from various centres in the world, in collaboration with offshore operators have resulted in significant breakthrough in understanding the behaviour of suction caissons, both during installation and under various conditions of loadings. These various research programmes had significant impact in operators’ practice, both validating and optimising innovative designs. Hence, centrifuge model tests formed part of the design validation process for applications in Australia, off the west coast of Africa (Randolph et al., 1998), as well as in the Gulf of Mexico. Jeanjean et al. (2006) also described results obtained at the University of Colorado, Boulder, which

1. Accelerated time frame. Centrifuge modelling and testing require a limited volume of soil, accelerating sample preparation. For soft soil, the process may be further accelerated by in-flight self-weight consolidation. Similarly, testing sequences are considerably shortened while ensuring the correct drainage conditions. This allows collection of a significant number of data in a short time frame and at a reasonable cost. 2. Accurate loading sequences. New motion control techniques permit accurate replication of complex loading sequences, including cyclic vertical and/or horizontal motion under either load or displacement control. This is particularly relevant when mimicking the complex motion of the pipe at the touchdown zone during laying, or the large deformations resulting from lateral buckling.


Pipeline-soil interaction

Offshore pipelines are laid on the seabed and used as flowlines or trunklines tied back to shore. They are becoming more and more common as oil and gas extraction is taking place in deeper and more remote areas. Geotechnical design procedures for offshore pipeline and risers have not yet reached the maturity exhibited by design procedures for piles, shallow foundations and anchoring systems. Reasons are multiple and relate mainly to the unknown geometry of the problem (the final embedment of the pipe is uncertain), the uncertainties of the soil conditions (both difficult to assess at very shallow depth and significantly affected by the installation process) and the very large deformations (and associated post-failure behaviour) the pipeline may experience (notably during controlled or unexpected lateral buckling). Whilst the on-bottom stability of pipelines (i.e. its stability against hydrodynamic forces) has been extensively investigated since the early 60s, notably in the USA, lateral buckling and axial walking (due to changes in internal pressure and temperature in the pipeline) have only emerged recently as a major research topic. Significant advances have been made over the last 5 years, using data from 1g model tests (performed notably in Cambridge, Cheuk et al., 2007, and NGI, Dendani & Jaeck, 2007), in-situ tests (notably using the SMARTPIPE device developed by Fugro, Jacob & Looijen, 2008) and also centrifuge tests. While each method has its own advantages and disadvantages (see Hill & Jacob, 2008 for a comparison of each method as applied to pipelines), centrifuge tests have certainly boosted knowledge of pipeline-soil interaction and generated significant breakthroughs. The main reasons are:

3. Use of natural soils. By requiring a limited volume of soil, it is possible to use in-situ soil with reasonable supply costs. The use of natural soils for model tests is maybe more important for pipelines than for other geotechnical structures, because of the heavy remoulding and the large deformations experienced by the soil. This may trigger specific behaviour that would not be captured by common artificial laboratory soils. 4. Accurate seabed characterisation. The good control of the soil reconstitution process results in a homogeneous sample. By using standard soil characterisation tools (such as the T-bar) or dedicated ones (such as the O-bar) in controlled conditions, it is possible to determine the soil characteristics accurately. 5. Enhanced instrumentation. By using image acquisition systems and pore pressure measurements, in addition to the measurements of load and displacements, one gains access to particular features of the pipe-soil interaction, such as the development of lateral berms during buckling, the creation of a trench during dynamic laying and the drainage conditions around the pipe during the pipe motion. All this information is pivotal in understanding and describing pipe-soil interaction. 6. Case-specific study. Centrifuge testing can be used to provide insight in particular issues related to pipeline design. This includes site-specific storm loading conditions or geohazard conditions, such as scarp crossings. The rest of this section aims to illustrate by a few relevant examples, some of the insights and breakthroughs, obtained from both centrifuge testing performed to gain insight into specific aspects of pipeline interaction, and from centrifuge modelling performed to assist in the design of specific pipeline projects. It is not a state-of-the-art review of pipeline behaviour and design. Such reviews have been presented by Cathie et al. (2005) and White & Cathie (2010).

6.3 Seabed characterisation Improving the reliability of design predictions of pipeline embedment during laying and lateral buckling relies on accurate measurements of the seabed characteristics in the upper 0.5 m of the seabed, which is the zone relevant to pipeline-soil interaction. The T-bar penetrometer, originally developed for the geotechnical centrifuge (Stewart & Randolph, 1994) is increasingly favoured in the field for characterising soft soils. The interpretation of T-bar penetration tests to assess the strength of soft seabed soils in the upper 0.5 m can be refined beyond the use of a constant NT-bar = 10.5 factor (Stewart & Randolph, 1994). These refinements have been described by White et al. (2010) and account for the soil buoyancy and the reduced bearing factor arising from the shallow failure mechanism mobilised prior to full flow of soil


Figure 28. Example the influence of near surface and buoyancy correction on the determination of the strength at shallow embedment of lightly overconsolidated clay. (after White et al., 2010).

around the bar. These corrections, which are marginal for assessing the strength of clay at moderate depths (>1 m) can be significant at shallow depth in soft materials. White et al. (2010) demonstrated that the omission of these effects may result in an underestimation of the shear strength in the depth range relevant to pipeline analysis (Figure 28). More recently, the use of T-bar penetrometers has been extended to investigate in the centrifuge phenomena relevant to pipeline behaviour, including the successive cycles of consolidation and remoulding experienced by the soil (Hodder et al., 2010) and the potential entrainment of water (and consequent swelling) during cyclic events. Figure 29 illustrates the loss of strength (expressed as the ratio of the remoulded shear strength su,r to the intact shear strength su,int ) experienced by different soils during cycles of remoulding for artificial clay and naturals clayey silts typical of the North West Shelf region, off the coast of Australia. The most striking feature of this cyclic test is the ten-fold reduction in strength during cycling for the Western Australian carbonate silt A in contrast to the two-fold reduction in strength for normally consolidated Kaolin. Further strength loss may be experienced by the soil as the pipeline remoulds and softens the soil, and water is entrained, increasing the moisture content of the soil. This is illustrated in Figure 30, which compares the strength loss for normally consolidated Kaolin clay resulting from deep cycles, from cycles 22 mm deep, breaking at the surface and from cycles 50 mm deep, breaking at the surface (using model scale units). The deep cycles do not break the surface, so water entrainment does not occur. For the kaolin test to 50 mm depth, the strength loss has been quantified at four different depths, from 10 to 40 mm, while for the kaolin

177

test to 22 mm, the loss of strength has been quantified at a depth of 20 mm. This test clearly indicates that the entrainment of water contributes to a continuous strength loss down to a value 5 times lower than the fully remoulded shear strength determined from deep cycles (without water entrainment). For typical normally consolidated kaolin, this strength reduction corresponds to an increase of water content by about 30%. The cycles to 50 mm depth indicate a similar trend with the magnitude of strength loss decreasing with depth. At 10 mm depth, the strength loss reaches a value similar to the one exhibited by the 22 mm cycles, but at a much faster rate. At deeper depths the degree of strength degradation reduces to eventually reach a value similar to the one obtained from deep cycles. As the T-bar penetrates deeper, the backflow of soil gradually limits the water entrainment. Even at very shallow depth, the effect of the water entrainment seems to reach a maximum, resulting in the soil still exhibiting some shear strength, even after a significant number of cycles. The simple examples presented above illustrate (i) the wide range of sensitivity exhibited by different types of offshore soils and (ii) the significant loss of strength resulting from water entrainment – a phenomenon likely to occur during pipe-soil interaction but which is not accounted in current design

approaches. In both cases, simple characterisation tests performed on in-situ reconstituted samples in the centrifuge may provide valuable and reliable information for design. 6.4

The knowledge of the pipe embedment is pivotal for subsequent on-bottom stability and lateral buckling design. The effect of dynamic laying on pipe embedment has been first observed by Lund (2000) who concluded on the necessity to account for pipeline laying history in subsequent on-bottom stability design, as it results in excessive embedment compared to a static laying process. During the lay process, an element of pipe moves through the touchdown zone, from an initial contact with the seabed to a stationary position, where the pipe weight is supported by an equal upwards seabed reaction force. The dynamic behaviour of the pipe through this process is complex and difficult to replicate. It is a function of the following parameters:

Figure 29. Degradation of strength during cyclic T-bar penetrometer tests (after Gaudin & White, 2009).

Figure 30. Strength loss at different depths in normally consolidated clay during deep and surface-breaking cyclic T-bar penetrometer tests (after Gaudin & White, 2009).


Dynamic lay embedment

– The profile of stress concentration in the touchdown zone due to the catenary shape, resulting in a vertical load on the soil varying along the pipe and higher, at some locations, than the as-laid pipeline weight. These pipe stresses can be estimated, based on a structural analysis of the hanging pipeline using the planned lay tension and hang-off angle. – The horizontal and vertical oscillation due to the motion of the laying vessel, resulting in a sweeping and damping of the soil at the touchdown zone. It can be estimated from the analysis of the motion of the vessel under for the wave motion considered. – The number of oscillations during the entire lay process. It can be assessed, based on the estimated pipeline laying rate, the touchdown zone length and the wave-induced oscillation frequency. Figure 31 presents a typical pipeline setup used to model dynamic laying and lateral buckling. The pipe is modelled as a short section (with an aspect ratio of at least 6 to prevent any end effect) and may be sandblasted to achieve a particular roughness. It is connected to a VHM loading arm, used to measure the horizontal loads, via a shear load cell, insensitive to bending moment. The purpose of this load cell is to provide a very accurate reading of the vertical load applied (permitting a resolution of 1 N at model scale) that is essential for pipelines, which impose stresses on the seabed that are orders of magnitude lower than typical geotechnical structures. Additional instrumentation may include pore pressure transducers at the pipe invert, as shown in Figure 31. A purposely developed motion control system (De Catania et al., 2010) is used at COFS to (i) apply a targeted vertical load depending on the stress concentration profile in the touchdown zone, via a feedback loop on the axial load cell and (ii) apply lateral oscillation of varying amplitude, linked via a second feedback loop to the pipe embedment.

The following example presents centrifuge modelling performed recently to determine the final embedment of a flowline to be installed offshore Australia (Gaudin & White, 2009). In-situ soil was used and the dynamic motion applied accurately replicated the motions determined from a numerical analysis of the wave, vessel and pipeline motion. The pipeline was divided into six sections, from the initial contact point to a far away position when the pipe was considered to be unaffected by the laying motion. The six sections featured six different loading sequences. Each sequence included a cyclic vertical loading of specified amplitude (from 0 to 1.55 times

Figure 31. Typical pipeline model setup.

the as-laid pipeline weight Vlay ) concurrent with horizontal cyclic motion, also of specified amplitude, as presented in Figure 32. The first sequence, modelling the first segment of the pipe at the touchdown zone, features a specific cyclic motion in which the pipe was pushed into the soil to a specified load of 0.7Vlay before being lifted up until separation between the soil and the pipe occurred. The resulting accumulation of pipe embedment is presented in Figure 33 for each sequence as a function of the cumulative number of imposed cycles. The key observation from Figure 32 and Figure 33 is the much larger pipe embedment (about 0.24 m) compared to a static embedment (about 44 mm using the formulation presented by Randolph & White, 2008a), even accounting for the stress concentration. Other important observations are the suction developed at the pipe invert during the first sequence of loading, where the pipe was lifted up from the soil (which potentially affects the stresses in the pipeline), and the dominant effect of the lateral motion amplitude compared to the cyclic vertical load. The first sequence, which featured the largest amplitude of lateral motion, resulted in a deeper embedment that the subsequent sequences, which featured a higher maximum vertical load but smaller lateral motion amplitudes. In other words, for these lay conditions and this soil type, the action of pushing the soil to either side of the pipeline during lateral motion (and concurrently remoulding it) has a dominant effect on

Figure 32. Dynamic lay simulation. Cyclic vertical loading and the associated pipe invert trajectory (after Gaudin & White, 2009).


Figure 33. Accumulation of pipe embedment resulting from the dynamic lay simulation of Figure 32 (after Gaudin & White, 2009).

the pipe embedment in comparison to any remoulding or penetration of the soil resulting from the vertical cyclic loading. These results stress the necessity to model accurately dynamic laying motion, in order to determine the pipe embedment, but also the remoulding of the soil resulting from the laying process, which are going to govern both on-bottom stability and lateral buckling. Further, the results demonstrate the capability of centrifuge modelling for simulating accurate laying motions, capturing particular soil behaviour features and delivering useful performance results. 6.5

Lateral buckling

To reduce the structural load resulting from thermal expansion, pipelines are permitted to buckle at targeted location, resulting in lateral displacements of the order of 10–20 times the pipeline diameter (Bruton et al., 2006). Whilst lateral bucking reduces the axial load on the pipeline, it also generates bending moment in the buckling zone, which can ultimately lead to local bending failure. The structural analyses used to assess lateral buckling in design require as input the soil resistance when the pipeline moves laterally. Unlike on-bottom stability design, overestimating the lateral resistance is not necessarily conservative as a high ‘friction’ factor can result in a more onerous structural load. The determination of an accurate friction factor is therefore pivotal for a safe and sound design. This requires the understanding of the soil behaviour at large displacements and through many cycles of loading, well beyond the point of failure, all features that can be well investigated by centrifuge methods. Two different examples are presented of centrifuge studies of lateral buckling. One of centrifuge

testing performed to understand mechanisms taking place during lateral breakout, and one of centrifuge modelling to provide friction factors for a particular soil and loading conditions, to assist in the design of a pipeline offshore Australia. In the first example a specific device was developed to observe and analyse, using PIV techniques, the deformation of the soil around the pipe by placing it behind a Perspex window (Dingle et al., 2008). In order to accommodate the large displacements occurring during lateral breakout, the camera was placed on a second actuator, which followed the displacement of the pipe, so the pipe remained constantly at the centre of the image. The corresponding setup is presented in Figure 34. Lightly over consolidated kaolin was used, and the model pipe was lowered onto the seabed to a targeted depth, hence ignoring any dynamic process and consequent soil remoulding. The test setup aimed at replicating the lateral breakout from known initial conditions. It deliberately ignored side aspects, which although important, are not believed to affect the qualitative response of the soil. Hence it qualified as testing rather than modelling. Dingle et al. (2008) presented a thorough analysis of the testing results, covering notably, the soil heave resulting from penetration and the assessment of the mobilised soil shear strength during lateral breakout. Two pivotal observations for heavy pipes (i.e. penetrating significantly into the seabed) were made, unknown at the time of project, regarding lateral breakout. First, the soil behaviour exhibited a brittle response, characterised by a sudden drop in resistance once suction at the rear of the pipe was lost. (Figure 35a) and resulting in the pipe rising up to shallower embedment. Hence, the peak lateral resistance is governed by the available tensile resistance at the rear of the pipe. Note that theoretical solutions for the two-sided mechanism


Figure 34. Centrifuge testing of pipe lateral breakout (after Dingle et al., 2008).

and for the one-sided mechanism were derived from upper bound plasticity limit analysis by Cheuk et al. (2008) and Randolph & White (2008b), respectively, following the observations from the centrifuge testing. Second, the post peak lateral response is governed by the growth of a soil berm at the front of the pipe (see Figure 35b; this can also be seen in Figure 31 and Figure 34). Further analysis presented by White & Cheuk (2008) indicated that the lateral resistance could be expressed as a conventional frictional term plus a berm resistance term proportional to the volume of the berm. This finding has critical consequences for design as it can lead to an overall friction factor, combining both phenomena, higher than unity while conventional friction factors used for design lied in the range 0.2–0.8. The second example presents centrifuge modelling performed to provide design data (dynamic embedment and friction factors) to assist in the design of a pipeline offshore Australia (White & Gaudin, 2008). Natural soils sampled from the site were used and reconstituted so they exhibited the same strength characteristics and index parameters as the in-situ soil. The dynamic laying process was accounted for, replicating the change of vertical contact force along the touchdown zone during laying. It was followed by lateral breakout and a series of lateral sweeps. By replicating soil conditions and key phenomena of the soil-pipe interaction, and by aiming to provide data directly applicable for design, the project qualifies as modelling rather than testing. Detailed results are presented by White & Gaudin (2008). Some key observations, related to initial breakout in fine grained soils are presented in Figure 36. Load developments, pore pressure measurements at the pipe invert and vertical displacement with pipe lateral displacements from Figure 36, and the knowledge obtained from centrifuge testing from Dingle et al.


Figure 35. Instantaneous velocity fields during pipe lateral breakout. Illustration of the tensile crack at the rear of the pipe (left) and formation of the soil berm at the front of the pipe (right) (after Dingle et al., 2008).

(2008), indicate the mechanism operating during initial breakout. Prior to separation the failure mechanism is two-sided, with soil being pulled behind the pipe in addition to being pushed ahead. The opening of a crack

Figure 36. Centrifuge modelling of lateral breakout (replotted from White & Gaudin, 2008).

at the rear of the pipe was demonstrated by the pore pressure measurements. After separation, a one-sided mechanism (e.g. with a berm forming at the front of the pipe) is operative. After breakout, the pipe rises to a shallow embedment. During this phase the lateral resistance also drops, reflecting the shallower failure mechanism that is mobilised. Once the pipe reaches a steady elevation the lateral resistance remains approximately constant and is created by shear at the base of the pipe and passive resistance against the berm of soil being pushed ahead. It is noteworthy that the pipe-soil interaction features observed during centrifuge testing, from which a design method has been refined (Bruton et al., 2008) were also observed during centrifuge modelling. Hence, the friction factor extracted from modelling and applicable to the specific site conditions could be confidently used for design using enhanced design methods developed from centrifuge testing. The two examples illustrate particularly well the combined contribution of centrifuge modelling and testing. The knowledge about a particular feature of soil-pipe interaction obtained from well controlled centrifuge testing helped interpret the data obtained from centrifuge modelling on specific soil and loading conditions. 6.6

Integration into design

The integration of physical model methods within the design process for geotechnical aspects of pipeline design such as lateral buckling has been discussed by White & Gaudin (2008) and is summarised in Figure 37. Several recent projects worldwide have included


centrifuge testing of pipe-soil interaction as part of the design process, making pipeline design the offshore area where centrifuge methods are the most commonly used. As, for any other offshore site specific project, centrifuge modelling needs to be planned in the early stages of the project so sufficient volume of soil can be recovered from the seabed during the geotechnical survey. The modelling programme may include specific centrifuge soil characterisation test to complement results from the geotechnical survey and/or validate the soil sample reconstitution process. A preliminary design of the pipelines will have been undertaken, from which the likely pipe weights, laying motion and in-service behaviour can be extracted in order to provide the required parameters to develop the model and instrumentation and to specify the number of tests needed to provide the required information for design. The centrifuge modelling programme should then be tailored to replicate the anticipated lay process and the in-service pipe movements (with an appropriate range to capture uncertainties) aiming for the results to provide a suitable representation of the design situation. Best practice is for the reporting to follow the same pattern as the conventional geotechnical investigation, with a sequence of factual and interpretive reports. The factual report provides a full record of the modelling activity, typically presenting pipe load-displacement responses during laying, lateral breakout and lateral sweeping, as well as the development of excess pore pressure if measured. The interpretive report analyses the results in light of the calculation model that the design team is anticipating to use for the pipe-soil behaviour. Key values from the testing, such as the as-laid embedment and

Figure 37. Integration of pipe-soil centrifuge modelling within the pipeline design process (after White & Gaudin, 2008).

the friction factors at key stages, should be extracted from the load displacement responses, allowing the uncertain aspects of the calculation model to be calibrated for the particular site. The calibrated calculation model, which may have been originally developed from centrifuge testing, is then used to simulate the particular design situation. This generalisation is usually required because the scope of the centrifuge modelling cannot encompass all patterns of pipe movement during operation. These vary longitudinally along the pipeline. It is therefore necessary to interpolate between the patterns of pipe movement simulated in the model tests. Similarly, the project may involve various weights and diameter of pipeline, which cannot all be simulated in the centrifuge study. Depending on the complexity of the project, outcomes from centrifuge modelling may be completed using finite element analysis calibrated from the centrifuge results.

7

DIRECTIONS FOR THE FUTURE

7.1 Hybrid modelling Real time substructure testing is an emerging technology within civil engineering modelling (Blakeborough et al., 2001). In this approach, the physical model represents only part of a larger dynamic system, with the remainder being simulated numerically in real time, during the experiment.

This permits the modelling of geotechnical structures under realistic and complex loading sequences which will be updated in real time from numerical analysis looped to the motion control system. This is particularly relevant for offshore structures, such as pipelines or jack-up, where the structural response and the geotechnical response are often dependant. This will surely constitute a major breakthrough in physical modelling techniques in the near future. One such example is the hybrid actuator, currently in development at COFS for centrifuge application, to investigate the behaviour of jack-up legs and spudcan when penetrated close to an existing footprint. The actuator models a single leg of a jack-up unit and can apply or measure independent or combined vertical V, horizontal H and moment M loading. In order to include the structural response of the jack-up unit, a stiffness matrix is incorporated at the top of the actuator leg, governing the leg displacement response as a function of the loads developed within the leg. The stiffness matrix is updated in real time, depending on the structural response of the jack-up unit, using a finite element programme. This provides a more realistic modelling of the footprint interaction problem. An even more advanced development is the UKNEES project (Madabhushi et al., 2010). The UKNEES project is a distributed testing network, where testing facilities in one site can be controlled by numerical or physical models located in a different site. In addition, the network features tele-participation with video feed, allowing participants to communicate


Figure 38. A pile featuring a wireless transmission module at its head: Mounted on the centrifuge actuator (left), details of the wireless transmission module (right) (after Deeks et al., 2010).

and share information simultaneously. In the example provide by Madabhushi et al. (2010), different parts of a bridge structure with shallow foundations were tested at different locations. The shallow foundation was tested in the Cambridge centrifuge facility. Loading on the foundation was achieved by an actuator, whose input came from physical and analytical models for the pier and sub-structure of the bridge deck tested at Bristol and Oxford. This type of network opens new possibilities for soil-structure interaction, with potential applications in the field of offshore geotechnics. A distributed testing network could for instance integrate the real-time numerical calculation of environmental loadings in one facility, which would feed the demand of a motion control system of a physical model in another facility. With the rapid development of networking facilities and the need for more realistic model testing, such networks will likely become more common in the future. 7.2

Enhanced instrumentation

The continuous development in computing and sensing technology will likely result in improved modelling techniques and new types of investigations. Two trends are identified, which will augment the benefit provided by centrifuge methods, namely image acquisition and intelligent sensors. As described in section 4.3, the development of PIV techniques has already constituted a major

breakthrough in physical modelling. Further improvements are expected with the increasing resolution and miniaturisation of cameras, leading to enhanced soilstructure investigations. High definition high speed cameras have started to be used more commonly for model testing at 1-g (Chow et al., 2010) and in the centrifuge (Boylan et al., 2009b), providing valuable information on high speed events. A good example is provided by Boylan et al. (2010a), where the usual geotechnical analysis is augmented by a geomorphologic analysis of submarine landslide, providing key information about the runout behaviour, and notably about the fluidisation of material and the transition from debris flow to turbidity current. This information could not be captured by usual data monitoring devices. The second trend relates to the development of intelligent sensors. One such example is the system presented by Deeks et al. (2010) to model pile installation by rotary jacking. The challenge associated with the continuous rotation of the pile during installation required the development of a wireless miniaturised data acquisition system to be located within the pile (Figure 38). The system is approximately 90 mm × 40 mm × 40 mm in size, and provides logging, conditioning and wireless transmission of up to 8 channels of data, eliminating the need of a data acquisition card on-board a computer. Such development opens new possibilities in term of data monitoring opening the field for new type of investigations.


8

CONCLUSIONS

The paper presented a review of typical offshore centrifuge applications, with a particular emphasis on the contribution of centrifuge methods to offshore design and the benefit of collaboration with industry. The paper lists the advantages and disadvantages of the use of centrifuge methods. It details the difference between centrifuge testing and modelling, with an aim to assist potential centrifuge users and existing modellers in optimising the outcomes of their model tests by defining why, how and when centrifuge methods should be used. The authors advocate the use of centrifuge methods as a valuable tool to assist the offshore industry in developing and designing solutions for a wide range of geotechnical problems. Indeed, centrifuge modelling and testing should not been considered as the unique tool to be used, but as a particular one whose outcomes may be maximised if integrated in a global approach. It is believed that while technological and scientific developments will, without doubt, increase the utility of centrifuge methods, the most spectacular improvement will be in understanding centrifuge methods contribution and their integration at key stages into a global design procedure, incoporating in-situ, numerical and analytical methods. ACKNOWLEDGEMENTS Centrifuge modelling and testing are experimental techniques which rely upon the expertise and dedication of technical staff. The authors would like to gratefully acknowledge the contribution of the technical teams at COFS, C-CORE and LCPC which were invaluable in carrying out successfully the projects presented in this paper. The authors would also like to thank BP for allowing the 2nd author to participate on this paper and funding several studies discussed herein. Progress of centrifuge techniques have greatly benefited from industry incentive and support. The contribution and financial support of the various industry partners involved in the projects presented in the paper are gratefully acknowledged. Fruitful and thought provoking discussions with various colleagues have helped shape this paper. The authors would like to thank particularly Prof. Mark Cassidy, Prof. Dave White and Dr Philippe Jeanjean for their contribution. REFERENCES Andersen, K., Jeanjean, P., Luger, D., Jostad, H.P. 2003. Centrifuge tests on installation of suction anchors in soft clays”. Proc., Intern. Symposium, Deepwater Mooring Systems, Houston, TX. Baba, S., Miyake, M., Tsurugasaki, K., Kim, H. 2002. Development of wave generation system in a drum centrifuge. Proc. Intern. Conf. on Phy. Model. in Geotechnics, St Johns, Canada, 265–270.

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Risk and reliability on the frontier of offshore geotechnics R.B. Gilbert The University of Texas at Austin

J.D. Murff Consultant

E.C. Clukey BP America

ABSTRACT: Operating in the frontiers of offshore energy production poses significant risk. The objective of this paper is to present lessons learned in managing risk on the frontier of offshore geotechnics. Case histories are described to underscore the following themes: (1) Achieving an appropriate risk requires balancing risk and conservatism; (2) Managing risk requires understanding the loads on our designs as well as the capacities; (3) Maximizing the value of data used to make design decisions requires considering the potential the data have to affect the decisions; and (4) Developing effective geotechnical designs requires understanding how these designs fit into the larger systems they support. The paper concludes with guidance to improve the state of practice. Better communication between all parties involved in design, construction and operation and earlier application of risk and reliability principles in the life cycle of a project will enhance the practical value of these principles.

1

INTRODUCTION

Operating in the frontiers of offshore energy production poses significant risk. The consequences of a failure are severe and the costs to mitigate risks are enormous. The offshore industry has been a leader in explicitly considering risk in design and decision making. One of the first reliability-based design guidance documents was developed for offshore facilities. In addition, this industry has been at the forefront in articulating and managing risk levels. Most importantly, due to the extreme conditions under which we operate, we have had the opportunity to learn from experience to innovate and improve practice to better manage risk. The objective of this paper is to present lessons learned in managing risk. Case histories are described to underscore the following themes: 1. Achieving an appropriate risk requires balancing risk and conservatism. 2. Managing risk requires understanding the loads on our designs as well as the capacities. 3. Maximizing the value of data used to make design decisions requires considering the potential the data have to affect the decisions. 4. Developing effective geotechnical designs requires understanding how these designs fit into the larger systems they support. The case histories presented represent a wide range of facilities, operators and locations. While what

is described is derived from real projects, it only represents a small and geotechnically-biased window into the numerous decisions and factors being considered in making those decisions. The paper concludes with guidance for improving the state of offshore geotechnical practice to better manage risk in the future. 2 ACHIEVING APPROPRIATE RISK Risk is the possibility of a loss. An appropriate risk is one that balances the cost of reducing the risk against that of accepting the risk.Appropriate risk is never zero risk, and excessive conservatism can be as troublesome as excessive risk. Risk is quantified as the expected consequence of loss. In the simplest case, if the possibility of loss is a binary event that either does or does not occur, then the risk is equal to the probability of the loss multiplied by its consequence. Therefore, risk is commonly represented by the probability and consequence of a loss. Consequences can include human safety, environmental and economic losses. General guidance for striking the balance between reducing versus accepting risk has been developed by different governments and industries. Examples include USNRC (1975) for nuclear power plants, AIChE (1989) for chemical process facilities, ANCOLD (1998) and USBR (2003) for dams, and Bea (1991), Stahl et al. (1998) Goodwin et al. (2002) and Gilbert et al. (2001) for offshore production facilities.


ballast would provide an acceptable reliability, even if the riser was installed immediately after pile driving. 2.2

Figure 1. Required ballast versus set-up time to achieve target probabilities of overload during installation for a riser tower foundation.

or on the capacity side of the design check (increasing the capacity),

The following case histories illustrate how the cost of reducing risk against that of accepting risk can be balanced in practice. 2.1

where rn is the nominal design capacity of the suction caisson in uplift due to side shear and reverse end bearing, sn is the nominal design load in uplift, and ϕR and γS are resistance and load factors, respectively. The ratio γS /ϕR is greater than 1.0, meaning that a larger nominal design capacity would be required if the weight of the caisson were considered to be a component of the capacity (Equation 1b). This case underscores the arbitrariness implied typically in different design formulations. The design-build contractor proposed using Equation (1a) versus Equation (1b). The basis for their proposal was that there was relatively less uncertainty in the weight of the caisson than the components of side shear and reverse end bearing. In addition, this proposal provided for a less expensive foundation. In order to evaluate the appropriateness of this proposal, a reliability analysis was performed. This analysis has the advantage that the results do not depend on a particular formulation for the design check. The event of interest was failure of the suction caisson in uplift under the long-term sustained load from the riser tower (short-term dynamic loads were very small in comparison to the total uplift load). The calculated probability of uplift failure in the 20-year design life was between 0.01 and 0.1. This probability of failure was considered to be intolerably high, and the design standard was subsequently clarified and revised in order to provide for a higher level of reliability.

Driven pile foundations for riser towers

The riser towers for an oil production system in deep water were to be held in place with driven pile foundations. Since the preliminary design was based on suction caissons, the soil borings were relatively shallow. The driven pile lengths were subsequently constrained by the depth of the deepest soil boring due to concerns about driveability. In addition, the largest load would be applied during riser installation, which potentially could occur before significant set-up after pile driving. Based on a design check using nominal loads and capacities, the axial uplift capacity of the piles alone was not sufficient. In order to increase the capacity, a ballasting system was developed, with boxes at the head of the piles that could be filled with steel ballast to increase the uplift capacity. A reliability analysis was performed to assess the need for the ballast boxes and the optimal amount of ballast (Fig. 1). The event of concern was overload during riser installation when the uplift force would be greatest because the riser is empty and the pile capacity would not necessarily have reached its maximum value due to full set-up. Based on a consideration of the consequences of a failure (primarily economic), the target threshold for the probability of overload was set between 0.001 and 0.0001. The results in Figure 1 provide a range of possible means to achieve an appropriate level of risk, including using larger ballast with a shorter set-up time or smaller ballast with a longer set-up time. Based on the proposed check, in which nominal or conservative values were used both for the uplift load and pile capacity, a ballast weight that was nearly 70 percent of the design load was selected (Fig. 1). The reliability analysis indicated that less than half that amount of

3

UNDERSTANDING LOADS

The focus in offshore geotechnics is generally on the capacity of geotechnical systems. However, the reliability of these systems can be influenced as much by the loads they experience as by their capacity. The


Suction caisson foundations for riser towers

A suction caisson was designed to anchor a bundle of risers for an oil production system in deep water. This design was unconventional in that the maximum load was a sustained tension load and not a storm load and the capacity was primarily from the weight of the foundation and not the shear strength of the soil. This particular situation meant that conventional factors of safety in the design standard did not necessarily apply. Specifically, the question was whether to include the weight of the foundation on the load side of the design check (reducing the total uplift load),

Figure 2. Schematic of wave-induced mudslides.

loading

mechanism

for

following case histories highlight the importance in understanding loads.

Figure 3. Results of slope stability analyses for offshore site in 130 m of water.

3.1 Wave-induced mudslides Wave-induced mudslides have caused considerable damage to offshore facilities in the Gulf of Mexico, particularly to pipelines that are an essential link in the supply of oil and gas. In Hurricanes Ivan (2004) and Katrina (2005), more than 50 pipelines and one platform were damaged or destroyed by mudslides in the Mississippi Delta (OTRC 2008). A schematic of the loading mechanism for a wave-induced mudslide is shown in Figure 2. The differential pressure on the sea floor underneath the waves (Fig. 3) imparts shear stresses in the soil. If these shear stresses exceed the shear strength of the soil, then a slope failure will occur. Results of stability analyses for a site subjected to large waves in Hurricanes Ivan and Katrina are shown in Figure 3 (OTRC 2008). The analysis assumes that the soft clays are sheared under undrained conditions by the waves and uses a profile of undrained shear strength versus depth from a nearby soil boring. Mudslides occurred here in both hurricanes, which is consistent with the stability analyses (Fig. 3). While the maximum wave heights in Hurricane Ivan at this location were nearly 20 percent smaller than those in Hurricane Katrina, the bottom pressure loading was actually higher in Hurricane Ivan and the factor of safety was smaller due to the longer wave period (Fig. 3). This result is significant because the eye for Hurricane Ivan was nearly 150 km to the east of the Delta (in contrast to Katrina’s eye, which passed over the Delta); the relatively small wave heights for Ivan in the Delta would not have been expected to induce mudslides, particularly in water depths of 130 m. However, the magnitude of bottom pressures, and therefore the risk of mudslides, depends both on the wave height and period. A comparison of wave heights and periods in the Delta for these two storms is shown in Figure 4. The conventional approach to characterize sea states in a hurricane is to relate wave period to wave height

Figure 4. Comparisons of wave periods and wave heights for waves in Mississippi Delta during recent hurricanes (from OTRC 2008).

based on empirical data collected from the largest waves in a storm (i.e., those waves near the eye). The curve labeled “Average” in Figure 4 is typically used to associate wave periods with wave heights for design purposes, such as in designing a production platform. However, the average curve under-represents the potential for long period waves away from the eye; the periods for the waves in the Delta during Ivan were similar to those near the eye, 150 km to the east, even though the wave heights were smaller (Fig. 4). The risk for mudslides in the Delta would be underestimated if this potential for longer period waves were not considered. Since platform loads are not very sensitive to the wave period, this issue was not recognized by and is not captured by the standard design guidance documents developed for platforms. However, it is an important consideration for mudslides. This case history underscores the significance of having a careful understanding of loads and not using information from design guidance documents without knowing its source and limitations.


Figure 6. Conceptual relationship between design conservatism, information and acceptable reliability.

of designers to apply conservatism at every decision point, it can clearly result in costly designs.

Figure 5. Annual probability that the actual load on the foundation will exceed the design load.

4

3.2 Tendon loads There can be significant conservatism embedded in the design loads that geotechnical engineers use to size foundations. To illustrate this point, the following conservative assumptions were made in establishing the design tendon load for a driven pile foundation of a Tension Leg Platform. First, an unlikely combination of wave height and period was assumed to establish the maximum tendon load. The 100-year design load corresponded to a significant wave height having a 0.01 annual probability of exceedance (the “100-year” wave height) combined with an associated peak spectral period with only a 5-percent probability of exceedance. Since the maximum tendon load increases with decreasing wave period due to the natural period of the structure, the resulting design load can be significantly greater than the load that will be exceeded with an annual probability of 0.01 (the “100-year load”). In addition, a conservative estimate of the tide and the worst possible wave loading direction were used for design, giving the maximum possible tendon load for a sea state. However, this combination of tide and wave loading direction was relatively unlikely in the conditions expected to produce the largest waves. Second, conservative assumptions were made concerning pre-tension measurement error in the tendon, mis-positioning of the pile head, and subsidence of the sea floor. The result of these compounded conservatisms was a design load with an extremely small probability of being exceeded. Figure 5 shows a probability distribution for the maximum annual tendon load on the pile. The annual probability that it will be exceeded is less than 1 in 100,000 (Fig. 5). Therefore, the “100-year” design load was really a “100,000-year” design load, which was subsequently factored up to establish the required design capacity. Therefore, the probability of failure for this design was many orders of magnitude smaller than what would generally be considered tolerable. This approach, selecting a series of conservative parameters in a design, is often referred as “double dipping.” Although it may be the natural tendency

A common decision point in offshore design is whether or not additional data are needed or would be beneficial enough to justify the cost of acquiring the data. A conceptual schematic illustrating how reliabilitybased design can be used to guide this decision is shown in Figure 6. This figure shows the cost of a foundation versus the cost of the information that is used to design the foundation. An acceptable level of reliability can be achieved by a variety of combinations of design information to reduce uncertainty in the foundation performance and design conservatism to limit the effect of uncertainty on the performance of the design. Design conservatism is reflected in the load and resistance factors or factors of safety as well as the nominal values used for loads and capacities. Increasing the cost of design entails increasing the difference between the expected capacity and the expected load. The level of information used in developing a design is rarely expressed explicitly in a code; however, it is typically implied through conventional practice. For example, design codes for foundations typically assume that a site-specific geotechnical investigation has been conducted to develop the design. Once a reliability-based design code is calibrated, the combination of the design information and the design conservatism presumably gives foundations with an acceptable level of reliability. The value of additional information depends on how much that information is expected to reduce design conservatism. This potential cost savings can then be compared against the cost of obtaining additional information, which is often governed by the time it will take to acquire it and the subsequent delay in development. Maximizing the value of information entails finding the optimal combination of design information and conservatism, such as the combination that will minimize the total expected cost. The following case histories illustrate projects where risk and reliability principles were used to improve the value of information.


MAXIMIZING THE VALUE OF INFORMATION

4.1 Design without site-specific soil boring A mature field had a rather substantial amount of geotechnical data, with more than 50 soil borings and geophysical survey lines. In addition, the cost and time required to obtain site-specific borings had been steadily increasing. Therefore, the question was whether or not a site-specific soil boring was worth its cost for every new facility in this field. The conceptual approach shown in Figure 6 can be captured in the design checking equation for a reliability-based code as follows:

where rn is the nominal design capacity and sn is the nominal design load; ϕR and γS are the conventional load and resistance factors, respectively, to account for the uncertainty in these quantities even if a new soil boring were available at the site location; and ϕspatial is a partial resistance factor to account for the added uncertainty if the amount of design information is less than the convention. As the magnitude of added uncertainty increases due to having less design data, the quantity (1/ϕspatial ) increases, meaning that a more conservative design is required to achieve the conventional reliability. In order to relate the value of (1/ϕspatial ) to the amount of design information, a geostatistical model was developed for this offshore field and calibrated with the available geotechnical data. The details of this model are described elsewhere (Gambino and Gilbert 1999 and Gilbert et al. 2008), and its important features are summarized here: • The model accounts for horizontal and vertical

correlation in the pile capacity. • The model incorporates data from modern borings (pushed sampling methods) as well as data from older borings where samples were obtained using wire-line percussion methods and were more disturbed. The effect of the method of sampling was included both in the expected pile capacity as well as the standard deviation in the capacity. • The model predictions reflect uncertainty both due to spatial variations as well as systematic uncertainty in the model due to the limited amount of data available to calibrate it. The output from this model is shown in Figure 7 for a small section in this field. The mean capacity at the location of an available, modern boring is equal to the design capacity obtained directly from that boring (Fig. 7a). The mean capacity at the location of an available, older boring is adjusted from the boring data to reflect what would be expected if a modern boring were drilled at that location. The unconditional mean or average for the field is 30 MN (Fig. 7a); the data at both the modern and the older boring suggest that the pile capacity within this section is higher than that for the field on average. As the location of the structure moves away from the location of the borings, the

Figure 7. (a) Expected value (b) and coefficient of variation (c.o.v.) for estimated pile capacity versus location for a 100-m long, 1-m diameter steel pipe pile (from Gilbert et al. 2008).

expected value for the design capacity approaches the unconditional mean (Fig. 7a). The added uncertainty in the capacity, expressed as the coefficient of variation or standard deviation divided by the mean, is zero at the location of the modern boring because this data point represents the conventional practice (Fig. 7b). The added uncertainty in the capacity is greater at the location of the older boring, reflecting the greater uncertainty associated with these data compared to modern practice. The ceiling of Figure 7b corresponds to the unconditional variance for the field. As the location of the structure moves away from the boring locations, the added variance approaches the unconditional variance (Fig. 7b). The required value for (1/ϕspatial ) is related to the added uncertainty due to not having a modern boring at the site location. Figure 8 shows this relationship for this field, where the added uncertainty is expressed as a coefficient of variation. In this application, the coefficient of variation for the added uncertainty was typically between 0.05 and 0.08 (Figure 7b); therefore, the required safety margin is about 10 percent greater than for the conventional case (i.e., 1/ϕspatial is about 1.1 in Fig. 8). These results provided the owner of the structure with guidance in making a decision between drilling a


Figure 8. Partial resistance factor versus coefficient of variation for spatial variability.

Figure 9. Probability distributions for tendon load and tensile pile capacity.

new boring or relying on the existing data and increasing the safety margin by about 10 percent. If the cost of increasing the pile length was less than the cost of a site-specific boring, then the alternative of moving ahead without another soil boring (i.e., “Less Data” on Fig. 6) would be preferred. The value of information from a site-specific boring depends on the geology; the greater the spatial variability, the greater the penalty to be paid by not having a site-specific boring (Fig. 8). The value of information also depends on the use of the information. For example, decisions about pile driveability may be more or less sensitive than those about pile capacity to the information from a site-specific boring. 4.2

Updated design using pile driving data

The pile foundation for a tension leg platform in a frontier area was designed based on a preliminary analysis of the site investigation data. The geotechnical properties of the site were treated in design as if the soil conditions were similar to other offshore areas where the experience base was large. The steel was then ordered. Subsequently, a more detailed analysis of the geotechnical properties showed that the soil conditions were rather unusual, calling into question how the properties should be used in design and leading to relatively large uncertainty in the estimated pile capacity. With three different commonly-used methods for estimating the tensile capacity of the piles, the factor of safety ranged from less than 2.5 to more than 4 (Fig. 9). For reference, the target factor of safety was 3. The owner was faced with a series of decisions. First, should they stay with the original pile design or change it given the uncertainty in the pile capacity, considering that changing the pile design after the steel had been ordered would substantially impact the cost and schedule of the project? Second, if they decided to stay with the original pile design, should they monitor the installation to confirm the capacity was acceptable, considering that this approach required a flexible contract where the pile design may need to be updated after the pile is installed?


Figure 10. Expanded view of probability distributions for tendon load and tensile pile capacity.

A reliability analysis was performed to provide guidance in answering these questions. One of the first findings from this analysis was that the uncertainty in the maximum tendon load was small compared to that in the pile capacity (Fig. 9). The second finding from this analysis was that the reliability of the pile foundation was governed by the possibility of the capacity being far below what was predicted by any of the design methods being considered. Figure 10 shows how the most probable combination of load and capacity leading to failure (the “Most Probable Failure Point”) is well below the estimated axial capacity assuming that the side shear is equal to the remolded undrained shear strength, which is a reasonable lower-bound on the available capacity. To support this concept of a lower-bound capacity, the pile load database that was used to develop and calibrate the API Design Method is shown in Figure 11 together with the calculated lower-bound for each data point. In all cases, the measured capacity exceeds the calculated lower-bound.

Figure 12. Probability of foundation failure in design life versus a lower-bound estimate of the pile capacity.

Figure 11. Comparison of measured capacity for driven piles in clay soils with estimated lower-bound capacity calculated assuming the side shear is equal to the remolded undrained shear strength of the clay; measurements are shown as a range to account for uncertainty in interpreting the load tests results (adapted from Najjar 2005).

A reliability analysis that explicitly accounted for the existence of a lower-bound or minimum value of the pile capacity was then performed. The lower bound was modeled as uncertain, with a normal distribution having a coefficient of variation equal to 0.2. The results of this analysis are shown in Figure 12, where the estimated value for the lower bound represents the most likely lower-bound value. If the estimated lower bound was greater than 0.25 times the estimated (or median) pile capacity after set-up, then the probability of foundation failure in the design life was smaller than the target of 0.001 (Fig. 12). The estimated value for the lower bound, obtained using the side shear equal to the remolded undrained shear strength, was about 0.4 times the estimated capacity after set-up, meaning that the existing design would have an acceptable reliability. Pile driving data during and after driving were used to update the reliability of this foundation. In general, pile monitoring information for piles driven into normally to slightly overconsolidated marine clays cannot easily be related to the ultimate pile capacity due to the effects of set-up following installation and uncertainties in the pile driving model. The capacity measured at the time of installation may only be 20 to 30 percent of the capacity after set-up. However, the pile capacity during installation does arguably provide a lower bound on the ultimate pile capacity, which governed the reliability of the foundation in this case. The estimated capacity based on the soil resistance to driving indicated that the reliability of the foundation was acceptable (Fig. 12). A re-tap after several days of set-up further supported this conclusion (Fig. 12). 4.3 Communication of uncertainty due to lack of information An important aspect in evaluating information is to clearly communicate uncertainty. As geotechnical

Figure 13. Stratigraphic cross-section with locations of soil borings and proposed platform (adapted from Gilbert et al. 2006).

engineers, we are rarely in the position of making decisions directly; rather we are conveying what we do and do not know as guidance to the decision makers. If a decision maker does not understand the magnitude of uncertainty, then they may not appreciate the potential value of additional information. A common offshore example is shown in Figure 13 where the final location of the platform does not correspond to the locations of the available soil borings. The geologic setting is a normally consolidated marine clay that is interbedded about every 30 m with overconsolidated clay crusts. In addition, there are buried alluvial channels that are filled with a variety of clay, silt and/or sand. A channel filled with dense sand was encountered at the location of Boring B. The risk is whether the driven piles at the platform location will encounter a thick enough layer of dense sand to cause driving problems. The conventional means to convey the available information is shown in Figure 13. A cross-section is drawn between the borings that includes geophysical data together with the data from the borings. The difficulty is that this cross-section conveys what is


Figure 14. Uncertainty multiples showing uncertainty in straigraphy at platform location (adapted from Gilbert et al. 2006).

expected, not what might occur. Even though there are question marks on the channel location away from the boring, a decision maker will tend to assume that there is no possibility of sand at the platform location because it is not shown below the platform location in the cross-section (Fig. 13). However, there is considerable uncertainty in whether or not and how much dense sand lies below the platform location. The available geophysical survey lines do not match up directly with the location of the platform. In addition, there is considerable uncertainty in the interpretation of the geophysical data, time-slice (seiscrop) plots from high resolution, surface-towed boomers. Comparisons of geophysical interpretations with subsequent boring data in this field indicated that the geophysical interpretations were not very accurate at identifying and locating channels buried with sand. An alternative depiction of the uncertainty in this information is shown in Figure 14. This depiction makes use of multiples, which are small-scale images positioned within the eye span on a single page or screen (Tufte 1990). In this case, “uncertainty multiples” (Gilbert et al. 2006) are used to convey the range of possible interpretations of the information based on the collective judgment of the geotechnical engineers and geologists, with each image representing an equally probable possibility. Figure 14 shows that there is a reasonable possibility (4/9 probability) of encountering sand at the platform location. It shows that if sand is encountered, its thickness could range from several meters to tens of meters. It shows that if a channel does lie below the platform location, it may or may not be the same channel or filled with the same material as the channel encountered at Boring B. It even shows that the thickness of the sand layer at the location of Boring B is uncertain; there is a 1/9 probability that it is thinner due to sand running into the borehole when the layer was encountered. The intent of Figure 14 is that a decision maker, who is most likely not a geotechnical engineer or a geologist, will be readily able to understand the uncertainty in the subsurface conditions. Whether or not a new site-specific soil boring is warranted will depend

Figure 15. Schematic plan view of mooring system for study spar.

on the cost of the boring, the cost of being prepared to drive through a dense sand layer if it is encountered, and the expected cost (or risk) of pile driving difficulty. 5

Most design considerations for foundations are based on individual components. However, these components generally perform within a system of foundation and structural elements. Ultimately, it is the performance of the system that is of most interest and concern. The following case histories underscore the importance of considering the system as well as the components. 5.1

Mooring system

The mooring system for a spar is shown in Figure 15. This spar design was developed by industry to be representative of typical practice for the purposes of studying and assessing that practice (OTRC 2006). There were fourteen lines, each composed of a section of chain connecting the line to the hull, a section of rope from the upper chain to just above the mudline, a section of lower chain connecting the rope to the suction caisson anchor at a padeye below the mudline. Three different water depths were considered for this system: 1,000 m with a semi-taut mooring spread and wire rope lines; 2,000 m with a taut mooring spread and polyester lines; and 3,000 m with a taut mooring spread and polyester lines. The probabilities of failure for individual components in the most-heavily-loaded line are shown in Figure 16. For each design (water depth), the probability of failure for the anchor is more than two orders of magnitude smaller than that for the rope and chain segments of the line; hence, failure of a mooring line during a storm is expected to be a break in the line itself versus a pull-out of the anchor. There is a potential to make the anchor designs more efficient (e.g., using a lower factor of safety) without jeopardizing the performance of the mooring system. Second, the consequence of a failure may depend on how the


CONSIDERING SYSTEMS AS WELL AS COMPONENTS

Figure 16. Probability of failure for components in most-heavily-loaded line.

failure occurs. Failure in the lines means that the hull could move off station by hundreds of kilometers during a storm and collide with other offshore facilities or coastal facilities. Failure by pull-out of the anchors means that the hull may not move off station as far due to the restoring force provided by the weight and dragging resistance of the anchors. However, the dragging anchors could damage seafloor facilities such as well heads and pipelines. The results on Figure 16 also show that the reliabilities of the components (chains, rope and anchor) and of the system of components that make up a single line (total) depend on the type of mooring system. The probabilities of failure are notably larger for the semi-taut system in 1,000 m of water compared to the taut systems in 2,000 and 3,000 m of water. These differences in reliability are due to using the same design recipe for all mooring systems, which is the standard of practice. Uncertainty in the maximum line load for a taut system is smaller than that for a semi-taut system because a greater proportion of the total load is pre-tension and controlled. Hence, the probability of failure for a taut system is smaller since it is less likely that the actual load will exceed the design capacity when the same factors of safety or load and resistance factors are used. Redundancy in the mooring system is shown in Figure 17 as the probability that the mooring system will fail (i.e., loss of station keeping for the spar) in the event that a single line has failed during a hurricane. A measure of redundancy is the inverse of the probability of system failure given component failure; for example, this redundancy factor for the semi-taut system is about 1.6 (or 1/0.6). Again, there is a significant difference between the different designs (water depths). For the semi-taut system, there is relatively little redundancy between failure of a single line and failure of the system. Conversely, the taut system distributes the environmental load more evenly between lines, and the corresponding redundancy factor is between 15 and 20. For comparison, the redundancy for an eight-leg fixed jacket in shallow water is between 30 and 125, depending on the loading direction (Tang and Gilbert 1993).

Figure 17. Probability of failure of the mooring system in the event that one line fails during a hurricane.

5.2


Jacket pile system

System effects can be even more pronounced for fixed jackets compared to floating production systems. To illustrate interactions between the jacket and the foundation system, the capacity of the foundation system for an 8-pile jacket that experienced large loads in Hurricane Ike is shown in Figure 18. The curves in Figure 18 represent capacity envelopes for various combinations of base shear and overturning moment applied to the foundation system; a combination of shear and moment that is outside of the envelope is expected to cause system failure. These system capacity curves were obtained using an upperbound, plasticity model of the pile system (Murff and Wesselink 1986 and Gilbert et al. 2010). The case study platform in Figure 18 is of particular interest because while the maximum load in Hurricane Ike was greater than the estimated capacity for the foundation system, the foundation survived the hurricane. The structural engineers assessing this structure after the hurricane increased the undrained shear strength of clay layers along the pile sides by a factor of three times in order to “explain” its survival. This practice is common place, and has led to a perception that foundation designs are excessively conservative (e.g., Gilbert et al. 2010). This case also points out the fallacy of trying to adjust an overall result by dramatically varying a parameter that has a relatively weak effect. However, the survival of this foundation system can more plausibly be explained by considering the entire structural system. The ultimate capacity of the system will be reached at large lateral deformations of the individual piles; therefore, any degradation of lateral soil resistance due to cyclic loading at lower load levels are not expected to affect the ultimate lateral capacity. In addition, the steel members in the structure, including the piles and jacket legs, are expected to have a higher yield strength than the nominal value used in design. Using expected lateral soil resistance and steel yield strength alone are enough to explain the

Figure 18. Increase in foundation system capacity of jacket platform in the hurricane loading direction due to higher lateral soil resistance (static versus cyclic p-y curves), increased steel yield strength (fY ) and modeling jacket leg stubs (adapted from Gilbert et al. 2010).

survival of this foundation in Hurricane Ike (Fig. 18). In addition, the jacket legs actually extend 3 to 4 m below the mudline, which increases the shear capacity of the system by forcing the plastic hinges in the piles deeper below the mudline. Accounting for this system effect further increases the capacity of the foundation system beyond the loading experienced in Hurricane Ike. Therefore, the performance this platform does not necessarily suggest that foundation designs are excessively conservative when the entire structural system is considered. A comparison of system effects for two jacket pile systems is shown in Figure 19. These figures were developed by assessing the sensitivity of system capacity to variations in the capacity of individual piles. The axial and lateral capacities of each pile were varied by ±30 percent from the design capacities to reflect possible variations in soil properties, installation conditions and structural capacities between piles. The maximum and minimum capacities obtained from this variation in any single pile are shown in Figure 19 to assess robustness of the pile system (i.e. the ability of the system to accommodate variations in component strengths). Both jacket pile systems are robust to variations in individual pile capacities for a shear-dominated failure (i.e., smaller overturning moments). These results complement those on Figure 18, where the pile system is more sensitive to structural factors than geotechnical factors for a shear-dominated failure. Conversely, both jackets are less robust to variations in individual pile capacities for an overturning-dominated failure (Fig. 19).Therefore, these jacket pile systems are much more sensitive to uncertainties in axial pile capacity than to uncertainties in lateral pile capacity. The three-pile system (Fig. 19a) is less robust in overturning than the six-pile system (Fig. 19b). For the same ±30-percent variation in axial capacity for individual piles, the three-pile system exhibits


Figure 19. Robustness in pile system capacity for (a) a three-pile jacket and (b) a 6-pile jacket (b) (adapted from Chen et al. 2010).

about a ±25-percent variation in system overturning capacity while the six-pile system exhibits less than ±10-percent variation in system overturning capacity. This difference in system behavior is real. Both pile systems experienced hurricane loads near their system capacities. The three-pile system failed in overturning while the six-pile system survived in shear without detectable damage. This difference in system behavior is not explicitly reflected in design practice. The same design recipe is used to calculate the loads and required design capacities for individual piles, regardless of whether the piles are in a single-pile system (like a caisson or a riser tower), a three-pile system, or a six, eight or more pile system. Therefore, the resulting reliability for these various systems will not necessarily be consistent. 6 THE FUTURE The keys to better managing risk in the future on the frontier of offshore geotechnics are embodied in the preceding lessons and case histories. Achieving an appropriate level of risk requires that we pay as much attention to the consequences of failures as to the probabilities of failures and costs associated with reducing the probabilities of failure.

198

A thoughtful examination of the consequences of failures can also highlight useful means to reduce risks. It may be more cost-effective to reduce risk by mitigating consequences compared to increasing reliability. A fundamental understanding of what affects the performance of a geotechnical system, including both the loads and the capacities, is far more effective in engineering appropriate solutions than simply being conservative to manage risk. Realistic estimates for loads and capacities are needed as well as the “nominal” values used in design. Realistic estimates include what is most likely, how much uncertainty or variation there is, what the sources of uncertainty are (e.g., temporal variations, spatial variations, inference errors, model uncertainties, installation tolerances, human errors, etc.), and any lower or upper bounds that physically limit what is possible. Furthermore, being “conservative” can have unintended consequences that result in taking on greater risk. Difficulty installing a “conservatively” designed foundation that is larger or longer than usual may leave a facility with less capacity in the foundation than if a “less conservative” design had been used. There is ample potential to improve the benefitto-cost ratio for the information used in offshore geotechnics. Realizing this potential will mean focusing on acquiring data that will be most informative or have the greatest impact on project-specific decisions. This approach is in contrast to always getting similar information for each project. It requires a frontend investment to develop preliminary decisions and designs based on available information so that the value of additional information can be assessed. It requires developing decisions and designs that are adaptable so that they can be updated throughout their lifetime based on new information. It requires considering how information from an existing project may provide value in future projects. Adopting a system-wide perspective can pay large dividends in better managing risks. Designs need to implicitly consider how the performance of geotechnical components will affect the overall performance of the system. These system considerations are not necessarily captured in design codes and guidelines, and they are project specific. These system considerations extend beyond the physical facilities to the factors driving decisions on the project. An elegant and economical foundation design using new technology may not be useful if the contractor selected to install it is unwilling to use different technology. These system considerations require that geotechnical engineers interact closely and communicate clearly with other disciplines, from conceptual design through construction and operation. Finally, humans are an integral part of managing risk. Humans collect and interpret the available information, make decisions and develop designs, construct and operate facilities, and are impacted directly and indirectly when failures occur. Since geotechnical engineers are not formally trained in psychology, sociology or communication, we need to reach out to

professionals in these non-technical areas in order to more effectively manage risk. 7

The intent of this paper was to provide case histories to guide practitioners in managing risk on the frontier of offshore geotechnics. The major themes are: 1. Achieving an appropriate risk requires balancing risk and conservatism. We need to become as involved in assessing and mitigating the consequences as we are assessing and minimizing the probabilities of failures, particularly in the context of the larger systems. 2. Managing risk requires understanding the loads on our designs as well as the capacities. We need to be careful to convey realistic ranges of possibilities and to avoid inadvertently compounding conservatisms. 3. Maximizing the value of data used to make design decisions requires considering the potential the data have to affect the decisions. We need to identify where information will be and will not be most valuable on a project-specific basis, develop designs and decisions that can readily be adapted based on new information, and convey uncertainty as clearly as possible to the decision makers. 4. Developing effective geotechnical designs requires understanding how these designs fit into the larger systems they support. We need to interact closely with other disciplines, from conceptual design through construction and operation. The use of risk and reliability principles has become an important part of offshore geotechnics, especially on the frontier. Better communication between all parties involved in design, construction and operation and earlier application of these principles in the life cycle of a project will enhance the practical value of the principles. ACKNOWLEDGMENTS We would like to acknowledge the following organizations that have directly or indirectly supported the material in this paper: US Minerals Management Service, American Petroleum Institute, National Science Foundation, Offshore Technology Research Center, BP, ExxonMobil, Sonaghal, Chevron, Shell, and numerous other operators and contractors that constitute the offshore oil industry.The views and opinions presented herein are ours alone and do not necessarily reflect those of any sponsor or employer. REFERENCES AIChE. 1989. Guidelines for Chemical Process Quantitative Risk Analysis. Center for Chemical Process Safety of the American Institute of Chemical Engineers. New York, New York.


CONCLUSIONS

ANCOLD. 1998. Guidelines on Risk Assessment, Working Group on Risk Assessment. Australian National Committee on Large Dams. Sydney, New South Wales, Australia. Bea, R.G. 1991. Offshore platform reliability acceptance criteria. Drilling Engineering, Society of Petroleum Engineers. June Issue, 131–136. Chen, J-Y., Gilbert, R.B., Murff, J.D.,Young, A. and Puskar, F. 2010. Structural factors affecting the system capacity of jacket pile foundations. Proceedings of International Symposium on Frontiers in Offshore Geotechnics. in press. Gambino, S.J. and Gilbert, R.B. 1999. Modeling spatial variability in pile capacity for reliability-based design. Analysis, Design, Construction and Testing of Deep Foundations, Roesset Ed., ASCE Geotechnical Special Publication No. 88, 135–149. Gilbert, R.B., Najjar, S.S., Choi, Y.J. and Gambino, S.J. 2008. Practical application of reliability-based design in decision making, Book chapter in Reliability-Based Design in Geotechnical Engineering: Computations and Applications, Phoon Ed., Taylor & Francis Books Ltd., London. Gilbert, R.B., Chen, J.Y., Materek, B., Puskar, F., Verret, S., Carpenter, J., Young, A. and Murff, J.D. 2010. Comparison of observed and predicted performance for jacket pile foundations in hurricanes. Proceedings of Offshore Technology Conference. OTC 20861. Gilbert, R.B., Tonon, F., Freire, J., Silva, C.T. and Maidment, D.R., 2006. Visualizing uncertainty with uncertainty multiples. Proceedings, GeoCongress 2006, ASCE, Reston, Virginia. Gilbert, R.B., Ward, E.G. and Wolford, A.J. 2001. A comparative risk analysis of FPSO’s with other deepwater production systems in the Gulf of Mexico. Proceedings of Offshore Technology Conference. OTC 13173.

Goodwin, P., Ahilan, R.V., Kavanagh, K. and Connaire, A. 2000. Integrated mooring and riser design: target reliabilities and safety factors. Proceedings, Conference on Offshore Mechanics and Arctic Engineering. 185–792. Murff, J.D. & Wesselink, B.D. 1986. Collapse analysis of pile foundations. 3rd International Conference on Numerical Methods in Offshore Piling, Nantes, France: 445–459. Najjar, S.S. 2005. The Importance of Lower-Bound Capacities in Geotechnical Reliability Assessments. Ph.D. Dissertation, The University of Texas at Austin, 347 pp. OTRC. 2006. Reliability of Mooring Systems for Floating Production Systems. Final Report for Minerals Management Service, Offshore Technology Research Center, College Station, Texas, 90 pp. OTRC. 2008. Mudslides During Hurricane Ivan and an Assessment of the Potential for Future Mudslides in the Gulf of Mexico. Final Project Report, Offshore Technology Research Center. Prepared for US Minerals Management Service. 190 pp. Stahl, B., Aune, S., Gebara, J.M. and Cornell, C.A. 1998. Acceptance criteria for offshore platforms. Proceedings of Conference on Offshore Mechanics and Arctic Engineering. OMAE98-1463. Tang, W.H. and Gilbert, R.B. 1993. Case study of offshore pile system reliability. Proceedings of Offshore Technology Conference. 677-686. USBR. 2003. Guidelines for Achieving Public Protection in Dam Safety Decision Making. Dam Safety Office, United States Bureau of Reclamation. Denver, Colorado. USNRC. 1975. Reactor Safety Study: An Assessment of Accident Risks in US Commercial Nuclear Power Plants. United States Nuclear Regulatory Commission, NUREG75/014. Washington, D. C. Tufte, E.R. 1990. Envisioning Information, Graphics Press. Cheshire, Connecticut.


2 Geohazards and gas hydrates



Neotectonic deformation of northwestern Australia: Implications for oil and gas development J.V. Hengesh, K. Wyrwoll & B.B. Whitney The University of Western Australia, Perth, Australia

ABSTRACT: Although Western Australia is commonly viewed as a Stable Continental Region with low rates of earthquake activity, geological and geomorphological evidence indicates that active tectonic processes are occurring: (1) the north coast is accommodating crustal flexure due to the collision with the Banda Arc; (2) the central west coast exhibits evidence of active fold growth and reverse movement on reactivated normal faults; and (3) the Murchison region has clear evidence of Quaternary tectonic deformation, repeated surface rupturing events, and a record of two large magnitude historical earthquakes. The evidence of Neotectonic deformation in northwestern Australia indicates that a number of seismic sources are present and these sources have the potential to produce moderate to large magnitude earthquakes such as the Mw 7.1 Meeberrie event. Future seismic hazard assessments should implement refined seismic source models that treat distinct seismic sources and the epistemic uncertainty associated with those sources. These seismic sources should be considered in the selection and engineering of seabed infrastructure such as manifolds, flowlines, export pipelines, and anchorage systems, as well as onshore LNG plants and port facilities. 1

INTRODUCTION

Western Australia is commonly viewed as a “Stable Continental Region” (SCR). It is largely composed of Archean age terranes, such as the Yilgarn and Pilbara cratons, Proterozoic and Phanerozoic age basins such as the Kimberly, Canning, Carnarvon, and Eucla Basins, and intervening deformed belts such as the Albany Frazier, Capricorn, King Leopold, and Halls Creek orogens. There is no orogenesis (active mountain building) occurring in WesternAustralia (WA) and the landscape is severely weathered with deep regolith attesting to long term stability on a regional scale (Anand and Paine, 2002). However, the occurrence of large magnitude historical earthquakes such as the 1885 ML 6.6 Mt. Narryer, 1941 ML 7.1 Meeberrie, and 1967 ML 6.7 Meckering events, and geomorphic evidence of crustal deformation and fault scarps from previous earthquakes indicate that parts of WA are being actively deformed. EPRI (1994) established criteria to define SCRs, which include: (1) evidence for no tectonic activity younger than early Cretaceous (∼100 million years before present {Ma}); (2) no deformed forelands or orogenic belts younger than Cretaceous (∼65 Ma); (3) no anorogenic intrusions younger than Cretaceous; and (4), no rifting or significant extension younger than Paleogene (∼35 Ma). SCRs are divided into domains composed of extended and non-extended crust, and domains underlain by extended crust are further divided into continental margins and failed rifts. According to EPRI (1994), SCRs that are underlain by extended crust have greater seismogenic potential than

those underlain by non-extended crust.Although Western Australia broadly meets these criteria, regional tectonic warping as evidenced by coastal submergence in the north, folding, the presence of numerous fault scarps, and the occurrence of several large magnitude earthquakes suggest that tectonic deformation is occurring across parts of WA. Active crustal dynamic processes occurring in SCR’s remain an enigma within the earth sciences. Understanding these processes is important for characterizing the location, severity, and frequency of earthquake occurrence, and assessing potential triggers for submarine landslides, liquefaction, and site response for structural design of both offshore and onshore facilities. 2

The northern margin of the Australian Plate is involved in a complex collision with the Sunda and Philippine Sea plates (Figure 1). Relative motion of the Australian, Sunda, and Philippine Sea plates is constrained from repeated geodetic surveys that use Global Positioning System (GPS) satellites to measure the precise positions of survey points located throughout a region. The repeated surveys provide direct measurements of the rates and directions of motion of points on different tectonic plates. Reoccupation of GPS sites in the Pacific, Australia, Indonesia, and Southeast Asia between 1991 and 2003 indicate that: (1) theAustralian Plate is moving along an azimuth of 015◦ and is converging with the Sunda Plate/Banda Arc at a rate of


REGIONAL TECTONIC SETTING

Figure 1. Regional tectonic setting showing major tectonic structures and relative motion vectors of tectonic plates.

67 to 75 mm/yr relative to a fixed Eurasian reference frame (Figure 1) (Bock et al., 2003; Nugroho et al., 2009); (2) the Philippine Sea plate is moving northwestward at a rate of approximately 110 mm/yr relative to Eurasia (DeMets, et al., 1994); and (3) the Sunda Plate is moving east-northeastward at a rate of about 7 to 11 mm/yr relative to Eurasia (Simons et al., 2007). The northern boundary of the Australian Plate follows the Sunda Arc subduction zone and the Banda Arc collision zone (Figure 1). The Australian Plate consists of two main parts including Australian continental crust and oceanic crust of the Indian Ocean, and the differences in crustal type control the nature of processes occurring along the plate boundary. Subduction, in the past, has occurred along the entire Sunda and Banda plate boundary, but now is limited to that part of the plate boundary where oceanic crust of the Indian Ocean is colliding with continental crust of the Sunda Plate (McCaffrey and Nabelek, 1984). Here, the thinner, denser oceanic crust is subducted northward beneath the thicker, less dense continental crust. Subduction extends from the Andaman Islands in the northwest to approximately 120–121◦ east longitude near the island of Flores (Audrey-Charles, 1975; Karig et al., 1987). East of this location, the oceanic crust of the Indian Ocean already has been consumed and subduction of oceanic lithosphere has ceased (Silver et al., 1983; Genrich et al., 1996). The plate boundary from Flores to East Timor is now characterized by collision of the Australian continental crust with fragments of the former island arc (e.g. Sumba, Rote, and Timor islands) and accretion of those fragments to the Australian Plate. The main deformation front now occurs along the northern side of the island arc on a system of south-dipping north-verging reverse faults that are referred to as the Bali-Flores and Wetar thrusts (McCaffrey and Nabelek, 1984). The formation of a south dipping thrust system on the north side of the island arc and similarities in motion vectors for both Australia and Timor (Genrich et al., 1996), indicates that the former island arc is being accreted to the

Australian Plate and the subduction zone has reversed polarity (compared to the Sunda Arc to the west). The collision of the Australian continental crust along the southern Banda Arc has caused profound changes in the style of deformation including cessation of north-directed oceanic subduction, accretion of the former island arc, and reversal of subduction polarity along the Flores and Wetar thrusts (east of 120◦ east). The collision of the Australian continental crust with the South Banda Arc also is causing warping and deformation of northern and northwestern Australia. Therefore, although the main active plate boundary lies on the north side of the Banda Arc, northern Australia is responding to the effects of the collision. We speculate that the transition from oceanic subduction to continental collision at approximately 120◦ east longitude is generating stresses in the Australian crust that are a source of strain energy for earthquakes observed in northern WA. Furthermore, although the main deformation front now lies on the north side of Timor, structures along the Timor trough may still be active earthquake sources. 3 TECTONIC DEFORMATION IN A STABLE CONTINENTAL REGION 3.1 Tectonic flexure Northeast directed convergence of the Australian Plate with the Banda Arc is causing downward flexure of the continental lithosphere in response to the collision. The flexure is most evident along the Kimberly and has formed an anomalously wide continental shelf and sinuous shoreline morphology consistent with crustal subsidence and coastal submergence. The continental shelf, shown on Figure 2, is over 500-km wide (at the position of East Timor) and decreases to less than 100-km wide near Exmouth (Cape Range). From Exmouth southward along the West Australian coast the continental shelf is typically between 40- and 100-km wide.


Figure 2. Digital elevation model of the northwest Australian continental shelf. Note wide shelf between Australia and Timor and narrowing of the shelf to the southwest. Elevation data from Geosceince Australia (2002).

elevations to sea-level high and low-stands. Figure 3 shows a strong correlation of surface elevations to the sea-level curve and using the maximum ages predicts minimum subsidence rates of 0.22 to 0.25 mm/yr for the shelf area south of the Timor trough (Table 1 and Figure 4). The hinge line of this flexure lies south of Broome, where coastal morphology changes from wave erosional (stable) to sinuous (drowning), and last interglacial (Marine Isotope Stage 5e) shoreline deposits (circa 120–130 ka) are present to the south and appear to be absent to the north. If the last interglacial deposits are now submerged the minimum coastal subsidence rate in the vicinity of Broome would be approximately 0.05 mm/yr. This indicates that the tectonic flexure associated with Australia’s collision with the Banda Arc diminishes from north to south and reaches zero north of the Cape Range where coastal uplift is observed and Last interglacial marine units are at their expected heights (Kendrick et al., 1991). The tectonic flexure documented by subsidence of geomorphic surfaces indicates that the continental lithosphere of northwestern Australia is being actively deformed due to the collision with the Banda Arc. This deformation will cause strain in the crust, and therefore the area of tectonic flexure may have greater seismogenic potential than areas not undergoing flexure.

Figure 3. Sea-level prediction diagram showing correlation between surface elevations and sea-level stages. From top to bottom, geomorphic surface elevations are −25 m, −40 m, −50 m, −70 m, −85 m, and −125 m. Elevation from Van Andel & Veevers, 1965. Sea-level curve from Lambeck and Chappell, 2001. Upper and lower curves indicate uncertainty.

The anomalously wide shelf in the north is inferred to be the result of marine erosion during progressive tectonic subsidence. Van Andel & Veevers (1965) recognized a series of submergent shoals, shelfs and terraces and postulated that these could have formed through a combination of tectonic subsidence and sealevel fluctuations. However, at the time, the theory of plate tectonics had not yet emerged, data on sea-level fluctuations were sparse, and thus it was not possible to document the nature and rate of deformation. Van Andel & Veevers (1965) recognized six main submergent surfaces at elevations of −120 to 140 m (at the shelf break), −85 m, −70 m, −50 m, and −40 m (along the continental shelf), and −25 m at a series of submergent atolls on the shelf break (Karmt Shoals) (Figure 2).The geomorphic surface elevations are plotted on a sea-level prediction plot (Figure 3), which is tentatively used to correlate the geomorphic surface

3.2 Folding and faulting in the Cape Range The 320-km long section of coast from the Cape Range to Cape Cuvier exhibits evidence of Neogene uplift and folding.The folding and uplift have strong regional expression, giving rise to the Cape Range, Cape Rough and Giralia Ranges, as well as the Lake MacCleod basin. Each of these ranges is mapped as an anticlinal


Table 1.

Correlation of geomorphic surface elevations and ages.

OIS stage

Sea-level (m)

Surface elevation (m)

Subsidence (m)

Age min. (ybp)

Age max. (ybp)

5e 5d 5b 3 3 2

3 −16 −30 −57 −75 −118

−25 −40 −50 −70 −85 −125

28 24 20 13 10 7

118000 106000 82000 54000 31000 19000

123000 108000 92000 60000 43000 28000

OIS = Oxygen Isotope Stage; ybp = years before present; m = metre

Figure 4. Subsidence rate diagram showing net subsidence and age values used in the rate calculation, and linear regression.

fold with intervening west-dipping reverse reactivated normal faults (Myers and Hawking, 1998). Evidence of uplift is spatially associated with the anticlinal structures. A series of at least four marine terraces occur along the west coast of the 120-km long N-NE trending Cape Range anticline. The presence of these elevated shoreline deposits indicates long-term emergence of the coast and active fold growth. The growth of the Cape Range anticline implies crustal shortening and movement along associated thrust faults. Coastal uplift also is expressed in the anomalous height of Pleistocene marine deposits exposed in coastal sections west of Lake MacLeod. Uplift of this section of coast appears associated with anticlinal structures, which has prevented a number of rivers from reaching the coast, and resulted in formation of the Lake MacLeod evaporate basin. In the Lake McLeod area, folding of Miocene and younger sedimentary deposits has formed the Gnargoo Range and diverted the Lyndon and Minilya rivers around the resulting anticlines (Figure 5). Development of these supercedent streams may indicate Quaternary uplift rates in excess of incision rates and can provide constraints on rates of fold growth and associated fault slip rates.

Figure 5. Neogene folding at Lake MacCleod. Black lines are fold axes. Streams flow around the noses of folds (supercedent) indicating fold growth controlled stream position. Modified from GSWA, 1985.

The crustal shortening and fold growth is likely associated with reverse movement along former extensional structures and is more pronounced in this region than elsewhere in the Carnarvon Basin. The geomorphic evidence for reactivation of former extensional structures indicates that these are potential seismogenic sources that should be explicitly considered in seismic hazard assessments. These structures follow the continental shelf break and cross much of the Exmouth Plateau (Myers and Hawking, 1998) and therefore may be near-field sources of ground shaking for offshore facilities. 3.3

The Mt. Narryer fault zone in central WA is located near the northwest margin of the Yilgarn craton and the eastern margin of the Southern Carnarvon Basin


Surface faulting and historical earthquakes

Figure 6. Digital terrain model (30 m digital elevation data reproduced by permission of the Western Australian Land Information Authority, 2010) showing oblique view to northeast along the Mt. Narryer fault scarp at the Roderick River. Arrows point to fault scarp and line shows position of topographic profile on Figure 7.

Figure 7. Topographic profile across the Mt. Narryer fault showing offset of alluvial valley surface (2009 data reproduced by permission of Western Australian Land Information Authority, 2010).

(Williams, 1979). The fault is approximately 120-km long and strikes in a northeast direction. Historical reports of earthquake strong ground shaking suggest that the fault zone may have produced the 1885 ML 6.6 Mt. Narryer earthquake (Clark, 2006). The epicenter for the 1941 ML 7.1 earthquakes also is located near the fault zone and so it is likely that the Mt. Narryer fault zone has produced two historical large magnitude earthquakes. The 1941 ML 7.1 Meeberrie earthquake is the largest earthquake to have been recorded in Australia. The Mt. Narryer fault zone includes at least four left-stepping en-echelon fault segments. From north to south the segment lengths are 11 km, 33 km, 40 km, and 35 km. The northern fault segments are expressed by strong vegetation alignments and fault scarps on the order of 1 to 1.5 m high (Clark, 2006).The linear nature and subvertical dip of the northern scarps suggests a significant strike slip component of motion. The two southern segments of the fault zone are expressed by a west-side up reverse sense of displacement and have formed east facing scarps across the Roderick and Sanford river alluvial valley deposits (Williams et al., 1983; Myers, 1997; Clark, 2006). Analysis of imagery and digital terrain models indicate that the scarps across the Sanford and Roderick rivers have

captured and diverted active stream flow, formed sag ponds, and impounded Lake Wooleen.The alluvial surfaces in both valleys are uplifted, warped and incised (Figures 6 and 7). Folding in the hanging wall of the fault zone has caused uplift and abandonment of the main river channel and incision of the river through the fold. Where the river cuts through the fold the channel pattern changes from braided to incised. Scarp heights of 3 to 8 m (Figure 7) suggest that the fault has experienced multiple surface rupturing events in Quaternary time. Preliminary analysis of drainage patterns and stream profiles west of the Mt. Narryer fault zone suggest that additional fault scarps may be present. 4

Although Western Australia is commonly viewed as a Stable Continental Region with low rates of earthquake activity, geological and geomorphological evidence indicates that active tectonic processes are occurring: (1) the north coast is accommodating crustal flexure due to the collision with the Banda Arc; (2) the central west coast exhibits evidence of active fold growth and movement on reactivated normal


IMPLICATIONS FOR OIL AND GAS DEVELOPMENT

faults; and (3) the Murchison region has clear evidence of Quaternary tectonic deformation, repeated surface rupturing events, and a record of two large magnitude historical earthquakes. The Mt. Narryer fault may be an analogy for the types of earthquakes that can occur on the reactivated normal faults along the Cape Range and Exmouth Plateau. In this case, a magnitude 7 or greater earthquake is a possible scenario for faults within 50 to 150 km of major offshore production facilities and onshore processing plants. Depending on the activity rates applied to these potential seismic sources, the incorporation of specific fault sources (in seismic source models for the region) could result in large contributions to the overall ground motion hazard at a site. The presence of active seismic sources also is an important consideration for evaluation of surface fault rupture hazards, selection of time histories used in site-specific site-response analysis, analysis of slope instability, and assessment of liquefaction potential. These seismic sources provide triggering mechanisms for instability and permanent ground deformation and should be considered in the selection and engineering of seabed infrastructure such as manifolds, flowlines, export pipelines, and anchorage systems, as well as onshore LNG plants and port facilities. REFERENCES Anand, R. R. and M. Paine, 2002. Regolith geology of the Yilgarn Craton, Western Australia: implications for exploration, Australian Journal of Earth Sciences 49, 3–162. Audley-Charles. M.G., 1975. The Sumba fracture – A major discontinuity between eastern and western Indonesia, Tectonophysics, 26, 213-228. Bock, Y., Prawirodirdjo, L., Genrich, J.F., Stevens, C.W., McCaffrey, R., Subarya, C., Puntodewo, S.S.O., and E. Calais, 2003. Crustal motion in Indonesia from Global Positioning System measurements. Journal of Geophysical Research 108 (B8), 2367. Clark, D. J., 2003. Reconnaissance of recent fault scarps in the Mt Narryer region, W. Australia, Minerals & Geohazards Div. Earthquake Hazard & Neotectonics Group, Canberra. DeMets, C., Gordon, R.G., Argus, D.F., and S. Stein, 1994. Effect of recent revisions to the geomagnetic reversal time scale on estimates of current plate motions. Geophysical Research Letters 21, 2191–2194. Genrich, J.F., Bock, Y., McCaffrey, R., Calais, E., Stevens, C.W., and C. Subarya, 1996. Accretion of the southern Banda arc to the Australian plate margin determined by

Global Positioning System measurements. Tectonics 15, 288–295. EPRI, 1994. The earthquakes of stable continental regions. Volume1: Assessment of large earthquake potential. Report prepared for Electric Power Research Institute by Johnston, A. C., Coppersmith, K. J., Kanter, L. R. and C. A. Cornell. Geological Survey of Western Australia, 1985. Geology of the Carnarvon Basin, 1:1,000,000, Bulletin 133, Plate 1. Geoscience Australia, 2002, Australian bathymetry and topography grid, CDROM. Karig, D.E., Barber, A.J., Charlton, T.R., Klemperer, S.E., and D.M. Hussong, 1987. Nature and distribution of deformation across the Banda Arc–Australian collision zone at Timor, Geological Society of America Bulletin 93, 18–32. Kendrick, G.W., Wyrwoll, K.-H. and B.J. Szabo, 1991. Pliocene-Pleistocene coastal events and history along the western margin ofAustralia. Quaternary Science Reviews, 10, 419-439. Lambeck, K. and J. Chappell, 2001. Sea level change through the last glacial cycle, Science 292, 679. Landgate, 2009, 30 m digital elevation data, geospatial data CDROM; www.landgate.wa.gov.au. McCaffrey, R, and J. Nabelek, 1984. The geometry of back arc thrusting along the eastern Sunda are, Indonesia: Constraints from earthquake and gravity data, Journal of Geophysical Research, 89, 6171–6179. Myers J.S., 1997. Byro, WA, Sheet SG 50-10 (2nd edition): Western Australia Geological Survey 1: 250,000 Series. Myers, J.S. and R.M. Hocking, 1998. Geological map of Western Australia, 1:2,500,000 (13th ed.), W. Australia Geological Survey. Nugroho, H., Harris, R., Lestariya, A.W., and B. Maruf, 2009. Plate boundary reorganization in the active Banda Arc–continent collision: Insights from new GPS measurements, Tectonophysics, 479, 52–65. Silver, B.A., D.R. Reed, R McCaffrey, and Y. Joyodiwiryo, 1983. Back arc thrusting in the eastern Sunda are, Indonesia: A consequence of arc continent collision, Journal of Geophysical Research, 88, 7429–7448. Simons, W. J. F., Socquet, A., Vigny, C., Ambrosius, B.A.C., Abu, S.H., Promthong, C., Subarya, Sarsito, D.A., Matheussen, S., Morgan, P. and W. Spakman, 2007. A decade of GPS in Southeast Asia: resolving Sundaland motion and boundaries. Journal of Geophysical Research 112, B06420. Van Andel, T.H. and J.J. Veevers, 1965, Submarine morphology of the Sahul Shelf, Northwestern Australia, Geological Society of America Bulletin, v. 76, p. 695–700. Williams, I. R., 1979. Recent fault scarps in the Mount Narryer area, Byro 1:250,000 sheet.: Western Australia. Geological Survey. Annual Report 1978, v. 51–55. Williams, I. R., Walker I. M., Hocking R. M., and S. J. Williams, 1983. Byro, W. Australia. W. Australian Geological Survey, 1:250,000 Geological Series Explanatory notes, p. 25p.



Deepwater Angola part I: Geohazard mitigation A.J. Hill & J.G. Southgate BP Exploration, Sunbury-on-Thames, UK

P.R. Fish Halcrow Group Ltd., Birmingham, UK

S. Thomas Fugro Geoconsulting Ltd., Wallingford, UK

ABSTRACT: BP Angola and its equity partners have embarked upon a rolling programme of subsea developments tied back to multiple FPSOs with future gas export to an LNG plant onshore. The programme concept has meant that BP’s Geohazard Assessment Team has been able to build and refine ground models for each development area. The key geohazards have been evaluated at the appropriate time within the project cycle allowing facilities engineers, drilling and subsurface teams to plan wells and subsea layouts that avoid or mitigate against geohazards where necessary. This paper provides specific examples of geohazards that commonly occur in deep water offshore Angola such as pockmarks, shallow gas and gas hydrates, faults, and seabed and sub-seabed slope instability. It also discusses some atypical features which have been encountered by BP and how they have been investigated.

1

INTRODUCTION

1.1 Background BP’s proposed programme of developments offshore Angola has provided an excellent opportunity for a phased assessment of geohazards and construction of a regional ground model to manage geotechnical risk. The ability to synchronise shallow geophysical and geotechnical data collection with the key stages in a project development (which BP terms APPRAISE, SELECT, DEFINE, EXECUTE and OPERATE) provides the best value to projects (Fig. 1). This is particularly true in relatively remote regions where mobilisation of survey vessels is normally associated with a long lead time. A programme of developments allows, for example, opportunistic data collection at one site whilst carrying out a reconnaissance level investigation at another site nearby. Similarly, a more mature project can benefit from more targeted supplementary information acquired in conjunction with a reconnaissance investigation in an adjacent area. The advantages of a phased approach are well understood by engineers and geoscientists (Campbell 1984). However, opportunities for implementing such a plan are often limited for a variety of commercial, political and logistical reasons. This paper describes how the phased approach adopted by BP Angola has identified key geohazards and provided an evolving ground model at a resolution that matches the project requirements as they develop.


Figure 1. Geotechnical Engineering and Geohazard Mitigation.

BP Angola has formalised the documentation of this process in terms of Geotechnical Engineering and Geohazard Mitigation (GGM) reports which are produced at key milestones; when new data are acquired

or the proposed infrastructure changes significantly. The GGM reports act as baseline information which reflect the project team’s collective state of knowledge of the shallow sub-surface geotechnical conditions and geohazards at a given time. The reports are used to ensure that contractors bidding for work do so on an equal footing and that subsequent designs are consistent. Ultimately the reports provide a reference for seabed-related problems throughout the life of the development through to decommissioning. Figure 1 illustrates the relationship between the project schedule, geo-data collection and the production of GGM reports. Figure 2. Typical pockmark field (hillshaded DEM).

2

GEOHAZARD MANAGEMENT

BP Angola’s overall strategy for management of geohazards is to avoid significant geohazards; if this is not cost effective or practical, the strategy is to implement design mitigations or accept the risks where they are fully understood and tolerable. Ground conditions which are less predictable using a regional model should also be avoided. However, if that is not possible or cost effective, the design must accommodate a wider range of soil conditions. In the case of high-risk structures, the design may need to be supported by additional location-specific data. As well as describing the seabed conditions close to, and at a distance from, locations of geotechnical investigations, a ground model also facilitates the selection of the right scope and tools and techniques to be employed for subsequent surveys and site investigations.

3 3.1

Table 1. Likelihood of activity

Scale of Score activity

Relict Periodically active Recently active Probably active Active

1 2

FEATURE MAPPING AND CHARACTERISATION Deepwater Angola geohazards

This paper provides specific examples of geohazards that commonly occur in deep water offshore Angola such as pockmarks, shallow gas and gas hydrates, faults, and seabed and sub-seabed slope instability. It also discusses some atypical features which have been encountered by BP and how they have been investigated. Examples of these are: a narrow trough feature, which is an extreme example of one of the many linear bedforms; the presence of unusually stiff clays near to the seabed; and the products of hydrocarbon migration which in some cases have reached the seabed. More detailed commentary on the engineering challenges that these geohazards pose is provided in a related paper at this conference (Hill et al. 2010). Some geohazards have had a direct influence on field layout and engineering design and some, so far, have not. There is an inter-relationship between some geohazards, such as the association of pockmarks with faults and salt diapirs, which act as flow-paths for fluid expulsion; the position of gas hydrates and the


Pockmark hazard severity classification.

3 4 5

No seepage Micro-seepage Gentle bubbling Vigorous bubbling Significant outbursts Eruptions of gas and sediment

Hazard Score *Result level 0 1 2

8

High

5

*Result is a multiple of likelihood and scale of activity scores

thermal conductivity of salt; and salt diapirism and the presence of stiff clays near the seabed. The following sections provide a brief overview of some key geohazards offshore Angola and how the risks they pose have been managed in a timely manner throughout the programme of subsea developments. 3.2

Pockmarks

Pockmarks are conical depressions in the seabed formed by fluid expulsion which may be hundreds of metres wide and tens of metres deep (Fig. 2). They are commonly found offshore Angola where potential hazards include expulsion of corrosive fluids and slope instability (Judd and Hovland 2009). In order to rationalise the constraint they impose on facilities, a method has been developed that assesses the hazard severity of individual pockmarks leading to specific guidance for hazard mitigation (Table 1). From the geospecialists that have reviewed pockmarks offshore Angola for BP, the consensus view for placement of seabed facilities nearby is to adopt a nominal 100 m stand-off from the rim (defined by the 5◦ slope angle ‘contour’) of pockmarks of medium and high hazard severity and that low hazard pockmarks

Figure 4. Fault types and relationships to salt.

sediments. This has been proven using CPTs fitted with a temperature element and is one explanation for an elevated base of the GHSZ close to the salt diapir. Figure 3. Relationship between pockmarks, gas and faults.

3.4 are just topographic constraints. Medium and high hazard pockmarks differ in the level additional assessment required to locate structures in stand-off zones. Additional soils data collection is likely to be required for high hazard pockmarks. The pockmark assessment has demonstrated that understanding their relationship with shallow gas and faults is central to determining their impact. High hazard pockmarks not only have an underlying gas pocket, but also a migration pathway along a fault or other plane of weakness such as the edge of a salt diapir (Fig. 3). 3.3 Shallow gas and gas hydrate Shallow gas and gas hydrate occurrences offshore Angola are attributed to a number of mechanisms: gas trapped beneath gas hydrate within thick PlioPleistocene sag basins, gas migrating from reservoir sources via faults or fractures intersecting structural traps, or gas trapped in sandy channel deposits within Miocene sediments. Irrespective of their sources, all shallow gas and hydrates are potential threats to drilling operations and to seabed founded structures, and therefore require systematic mapping. Detailed mapping of bottom simulating reflectors, believed to represent the base of the gas hydrate stability zone (GHSZ), have shown they are found at variable depths below seabed. This variation is thought to relate to local variations in environmental factors such as increases in water salinity, changes to the geothermal gradient around salt diapirs, and chemical composition of the shallow gases forming the hydrate. Recognition of these local factors emphasises the need for detailed mapping, rather than reliance on modelled relationships between base of GHSZ and water depth. One feature of the presence of salt (discussed later) is its relatively high thermal conductivity which elevates the geothermal gradient in the surrounding


Faults and seismic hazard

Faulting evident offshore Angola can be broadly classified into two main types: very shallow polygonal faults, common in thick Plio-Pleistocene clays and caused by dewatering (Cartwright and Lonergan 1996); and faults driven by the extensional regime imposed by salt movement which generally extend no more than 7 km deep. Salt-related faults are classified depending on their relationship to the underlying salt: domino faults sole out on the ductile surface of the salt and are formed by lateral salt movement; flap faults are extensional and occur at the margins between subsiding basins and rising salt diapirs; and keystone faults, which are associated with anticlinal folding from salt uplift (Fig. 4). BP’s concessions are some distance from plate boundaries and consequently no faults are considered to be of tectonic origin (Fugro West, 2008). Measured displacements and calculated displacement rates of less than 1 mm/year (Fig. 5) suggest that there will be no significant movements across these faults during the field life. This is due to very low rates of sedimentation and slow movement of salt. It was concluded in the early SELECT phase that there was very low risk of fault displacement impacting seabed structures, but that there remains a residual risk that human intervention (such as reservoir depletion) may trigger movement along pre-existing faults. The seismic hazard is from background and onshore sources. Seismic hazard analyses have estimated peak ground accelerations of 0.015 g and 0.15 g for 200year and 2,900 year return periods respectively. It is concluded that seismic hazard is not likely to be significant but key structures should be assessed against the full design spectrum according to the period of the structure and foundation. It was concluded that for the development areas in question, no stand-off zone was needed with respect to faults. The reactivation of existing faults is considered unlikely and generation of new faults even more so in the project lifetime.

Figure 6. Mass transport deposits and steeply dipping beds above salt diapir (chirp data).

Figure 5. Fault movement rates.

3.5

Seabed slopes and buried mass transport deposits

Active landslides have not been widely observed on the seabed, except for small failures on the oversteepened flanks of salt diapirs, but seismic data indicate that Mass Transport Deposits (MTDs) are commonly observed in the sedimentary sequence of sag basins between salt diapirs (Fig. 6). This observation suggests that the very gradual uplift of the salt leads to over-steepening, and eventual failure, of slopes. Furthermore, it indicates that slope failure cannot be discounted over the life of field, especially if slopes are loaded by pipelines or other seabed facilities. Consequently site-specific slope stability assessment has been adopted (Hill et al. 2010). MTDs pose a drilling hazard associated with wellbore instability and require assessment on a well-specific basis. Figure 6 also shows steeply dipping beds of older and stiffer clays pushed towards seabed surface by rising salt, which may cause unexpectedly strong materials to be encountered by shallow foundations. 3.6

Narrow trough and other bedforms

Bedforms in a variety of morphologies are very commonly seen in ultra deep water offshore Angola, but they are not generally considered to be hazardous because of their low amplitude and extremely slow rate of migration. One exception is a particularly large bedform (up to 250 m wide, 30 m deep and over 15 km long) and has been separately described as a narrow trough. This feature was identified as a significant geohazard risk driver to the development because it bisects two potential drill centres. Avoidance would have meant a pipeline reroute of over 10 kilometres. There was also

Figure 7. Narrow trough conceptual model.

a concern the feature may result in enhanced potential for erosion or sedimentation and slope instability (Fig. 7). Opportunistic sampling and deployment of seabed current monitoring equipment meant that these concerns could be addressed in the project’s SELECT phase. Sedimentation rates calculated from geochronological tests (14 C and optically-stimulated luminescence) on the flanks and thalweg of the narrow trough were low. Data from both locations indicate no significant active erosion but slightly higher rates of deposition on the flanks of the trough relative to its centre (0.2 mm/yr on flank and 0.08 mm/yr in thalweg). Current meter data show that bottom currents associated with trough are also low (0.04 m/s), with little potential for scour. The more common bedforms are also static features, which are not expected to change in size or migrate over the life of field. They do not require a stand-off


Figure 8. Conformable and cross-cutting HIRs and seabed manifestation as mounds.

zone and are represented as topographic features only which may pose flow assurance constraints. However, the size of the narrow trough is exceptional, meaning it has potential slope instability and flowline spanning issues. Based on these data, it is concluded that the narrow trough is a static feature acting as a topographic constraint to facilities, and its main impact relates to potential slope instability (Hill et al. 2010). 3.7 Hydrocarbon migration Seabed photographs taken during environmental surveys show dark-coloured hydrocarbon extrusions and associated paler-coloured suspected carbonates. The features were also mapped from multibeam echosounder (MBES) data where they were shown to be widespread, and up to 20 m across and 2 m high. Analysis of AUV (chirp) sub-bottom profiler data shows that the subsurface extent of the features is more extensive than their seabed expression, and a High [acoustic] Impedance Reflector (HIR) has been mapped across the development area. There were examples where the HIR is conformable with the bedding and others where it cuts across the beds (Fig. 8). These features were first observed during RoV surveys shortly before they were also identified on side-scan sonar, multibeam echosounder and subbottom profiler data. The reconnaissance geotechnical investigation was then able to target these features using gravity core and seabed CPT systems. Gravity core samples were extremely high strength, and CPT refusal occurred at the intersection of some, but not all, HIRs. More detailed investigation of these features was then scheduled using additional RoV surveys and geotechnical drilling during a supplementary investigation. This was another example of the phased approach to building a ground model where detailed geotechnical investigations were informed by earlier reconnaissance-level investigations.

All of this work was completed prior to DEFINE which prevented it becoming a potentially costly critical-path issue later in the development cycle. Subseabed occurrences of carbonate-rich claystones have been reported by other operators elsewhere offshore Angola, but not as extensively or with such visible seabed manifestations. Following analysis of samples recovered from geotechnical boreholes, it is thought that the HIR is indicative of the stratigraphic position where rising hydrocarbons have interacted with the in situ sediments over geological time resulting in the precipitation of carbonate. In some cases the hydrocarbon reaches the seabed and forms extruded mounds. The majority of samples across the HIR show a predominance of carbonate, mainly in the form of thin, hard carbonate-rich claystone beds or nodules with some intra-bed voids. The hard carbonate-rich claystone beds or nodules are typically only a few centimetres thick and nearly all the carbonate has at least a small component of remnant hydrocarbon associated with it. In some examples there is still a significant proportion of hydrocarbon associated with the hard carbonates. These hydrocarbons have been recovered in the geotechnical samples as either asphalt or as thick oil. Seabed extrusions of asphalt and carbonate mounds are static features which may cause abrasion of flowline coatings, and therefore constrain routing. The corridor width needed for installation of flowlines is typically a few tens of metres and therefore individual mounds have been mapped giving field planners the opportunity to pass through an area of mounds while avoiding individual features. 3.8


Salt diapirism

Salt diapirism is not considered a geohazard in itself. However, salt movement (which is predominantly upwards in the shallower water areas and has a more lateral component in the deeper water) can impact the field layout and foundation designs. Salt uplift and subsequent erosion of the seabed causes stiff to very stiff clays of older strata to be thrust upwards such that clays with undrained shear strengths of greater than 150 kPa have been sampled within 1 m of the seabed. Again, reconnaissance geotechnical investigations provided an early indication of these conditions and supplementary soils data combined with the engineering-quality AUV data enabled the extent and characteristics of these materials to be well constrained. More details on how salt diapirism influences the geotechnical zoning, (known as soil provinces) of the shallow subsurface are provided in Hill et al. (2010). The steep slopes formed by this uplift may be over 20 degrees, which may constrain pipeline routing and can give rise to slope instability, particularly in those areas where recent sedimentation forms a drape over the older materials. The in situ stress regime is also altered by the salt movement, giving rise to tension

cracks and faulting in the most active areas and reorientation of the principal stress directions elsewhere. This can influence the radial stresses providing support to piled foundations. 4

CONCLUSIONS

This paper describes a phased approach to geohazard and geotechnical characterisation used by BP Angola to develop a regional ground model for optimising field layouts and engineering wells and facilities. This strategy has been shown to be highly effective with significant advantages over less systematic methods. It has allowed the project team to identify the potentially high-impact geohazards early in the development cycle; to focus on the geohazards that are most relevant for the planned developments; and to avoid rework and threats to the project schedule by achieving geohazard-tolerant field layouts early. Specifically, this approach has: enabled pockmarks to be ranked according to their severity; to understand the threat posed by shallow gas; discount the impact of faults and seismic hazards; quantify the risk of slope failure; allowed the narrow trough to be classified as a topographic constraint rather than a dynamic feature; provided an understanding of the nature of hydrocarbons and their by-products in the shallow surface; and to identify clearly the areas of atypical soils due to the effects of salt diapirism. Some of the geohazards described in this paper are relatively rare in the developed areas offshore Angola, but others are ubiquitous. As development reaches ever-increasing water depths the authors believe that for similarly geohazardous areas, the systematic approach adopted by BP Angola for evaluating the impact of geohazards is required. There remains however, a need for the approach to managing geohazards to be tailored to project and site specifics.

5

This paper represents the views of the authors and not necessarily those of the companies they represent. The conclusions relate to specific aspects of some of the geohazards present offshore Angola hopefully of interest to the geotechnical community. The authors wish to thank BPAngola, their development partners and Sonangol for permission to publish this material. They are also grateful for the critical reviews, guidance and stimulating conversation provided by Trevor Evans, Mike Sweeney and Mike Fiske.


CLOSING REMARKS AND ACKNOWLEDGMENTS

REFERENCES Campbell, A. J. 1984. Predicting Offshore Soil Conditions. Proceedings Annual Offshore Technology Conference Houston, Texas, 7–9 May, OTC paper Number 4692. Cartwright, J.A. & Lonergan, L. 1996.Volumetric contraction during the compaction of mudrocks. Basin Research 8: 323–331. Dendani, H., Colliat J.L., Puech,A. and Nauroy, JF 2010. Gulf of Guinea deepwater sediments: geotechnical properties, design issues and installation experiences Keynote Paper, Proceedings 2nd International Symposium Frontiers in Offshore Geotechnics Perth, November 2010. Evans, T.G. 2010. A Systematic Approach to Offshore Engineering for Multiple-Project Developments in Geohazardous Areas, Keynote Paper, Proceedings 2nd International Symposium Frontiers in Offshore Geotechnics Perth, November 2010. Fugro West (2008) Seismic Hazard Analysis Block 31, Offshore Angola, Issued to BP Angola December 2008. Doc No. 3193.034 Hill, A. J., Evans, T.G., Mackenzie, B & Thompson, G, 2010. Deepwater Angola Part II: Geotechnical Challenges, Proceedings 2nd International Symposium Frontiers in Offshore Geotechnics Perth, November 2010. Judd, A. & Hovland, M 2009. Seabed Fluid Flow. Cambridge: Cambridge University Press.


Deepwater Angola part II: Geotechnical challenges A.J. Hill & T.G. Evans BP Exploration, Sunbury-on-Thames, UK

B. Mackenzie Fugro Geoconsulting Ltd., Wallingford, UK

G. Thompson Senergy Survey & GeoEngineering, Bath, UK

ABSTRACT: BP Angola and its equity partners have embarked upon a rolling programme of subsea developments tied back to multiple FPSOs with future gas export to an LNG plant onshore. This paper discusses the geotechnical challenges related to geohazards encountered in those development areas and the benefits of a properly constructed ground model in regions that are geologically complex. Examples include: the stability of seabed slopes in and around salt diapirs; and the behaviour of deep foundations in stiff clays up-thrust by salt and in soils modified by hydrocarbon migration.

1

INTRODUCTION

encountered offshore Angola and how they have been managed by BP and its partners.

1.1 Background Hill et al. (2010) and Evans (2010) describe the geohazards identified by BP offshore Angola and the value of constructing a predictive ground model as an effective way of mitigating those hazards. This paper describes how the ground model has been used to provide guidance for geotechnical engineering purposes, with particular reference to suction caissons, driven pile foundations and seabed slope stability. There are many other geotechnical challenges common to deepwater developments in more typical normally consolidated clay environments, but this paper is restricted to those key foundation design issues associated with geohazards. The ground model approach seeks to adequately define a 3D block of soil in a proposed development area. If the ground model is constructed properly, much of the geophysical and geotechnical data acquisition can be done early in a project cycle. It also provides flexibility for changes in field layouts and limits the amount of geotechnical investigation required after the seabed facilities architecture is frozen. Timing the investment in surveys, site investigations and subsequent analysis is critical. The key here is to recognise the complexity of a site (or conversely the simplicity, in some areas) at the right stage in the development cycle. The goal is to produce a ground model that is fit-for-purpose: neither overly complex, nor insufficiently resolute to design reliable foundations, anchors and flowlines. This paper describes the impact of some of the more challenging geotechnical conditions

2 2.1

General

The geotechnical conditions offshore Angola in freefield areas (i.e. those outside the influence of geohazards) are typical of deepwater hemipelagic environments. With the exception of a slightly stronger “crust” within 1m of the seabed, the clays are generally lightly over consolidated to normally consolidated with a shear strength gradient of 1–1.5 kPa/m. The submerged unit weights of the clays are very low (typically only 2–3 kN/m3 in the upper 20 m). As part of the ground modelling process, the shallow subsurface is divided into soil provinces which describe areas or zones within which the geotechnical conditions are interpreted to be broadly similar for engineering purposes (Fig. 1). The criteria for defining the boundaries that separate soil provinces are based on foundation engineering considerations and specifically the inferred geotechnical conditions within 35 m of the seabed. The idealised boundaries between soil provinces are defined by projecting the intersection points of the 35 m depth contour and key seismostratigraphic horizons (Fig. 2). The selected seismostratigraphic horizons were chosen because they represent potentially significant changes in geotechnical conditions which could impact the suitability of suction-installed caissons, currently the most common deep foundation type for BP Angola’s


GEOTECHNICAL CONDITIONS

Figure 1. Soil province map.

Figure 3. Undrained shear strength in each soil province.

occupy the sag basins between salt diapirs and topographic highs created by movement of the underlying salt. Soils in the sag basins are generally uniformly normally to lightly over consolidated with physical characteristics that are reasonably predictable over large areas with a sufficiently advanced ground model. Conversely, soils on the topographic highs that have been uplifted by salt can exhibit significant variability, ranging from very soft to hard clays with possible presence of sand layers. The soils become increasingly less predictable as the influence of salt increases, as shown by the widening undrained shear strength design envelopes in Figure 3.

Figure 2. Definition of soil provinces.

developments. The 35 m depth was chosen because this represents the maximum expected suction caisson penetration with a small margin for tolerances. Each soil province is characterised by a layered sequence of geotechnical units with associated characteristic soil parameters. The soil provinces represent a simplified and idealised model for what are complex conditions. Although the simplified models show the boundaries between soil provinces as sharp breaks, they actually represent transition zones. During the early stages of characterising a development area, the soil provinces are largely based on the interpretation of the 3D exploration geophysical data but as higher resolution (engineering quality) geophysical and geotechnical site investigation data are acquired the soil provinces are refined and engineering guidance developed. 2.2

Impact of salt diapirism on soil provinces

In some parts of BP Angola’s development areas, the soil conditions are heavily influenced by salt diapirism. Salt movement, both uplift and lateral spread, influences the seabed morphology and features, the stress regimes and the engineering properties of the soils. Soil provinces have broadly been defined by areas that

2.3

The results of sub-bottom geophysical (chirp) profiling revealed an anomalous semi-continuous High [acoustic] Impedance Reflector (HIR) in some areas of the ground model. This HIR was within the depth of interest for foundations and anchors and was considered sufficiently important to justify additional investigation. Physical samples, in situ tests of the seabed and ROV surveys, improved the understanding of the geotechnical significance of the HIR. Most recently, geotechnical boreholes have been performed to examine the sub-surface properties of this feature and have developed our understanding even further.


Impact of hydrocarbon migration

Figure 4. Geohazard mitigation process.

Figure 5. Foundation guidance – soil provinces.

When only geophysical data and information from seabed testing were available, advice to subsea engineering teams was necessarily cautious, and mitigation by avoidance was the first option. With the benefit of the new data, it was concluded that the HIR need not necessarily be avoided if driven piles were used rather than suction caissons. However, it would need to be proven that the in-place performance of foundations could be reliably predicted in the materials beneath the HIR. This is discussed further in Section 3.

comprise largely predictable normally consolidated to lightly over consolidated clays, the guidance was that suction-installed and driven piles may be used and that additional structure-specific soils data may not be required. The advice for Soil Province 3 was that suction-installed caissons and driven piles may be used, although structure-specific samples and/or CPTs would be required for design optimisation. In Soil Province 4, suction-installed caissons were not recommended because the older, stiffer and more variable soils may be close to the seabed. The earlier recommendation for driven piles in Soil Province 4 was that they may be used, but that structure-specific soils data would be needed. At the time of writing, a series of geotechnical boreholes has just been completed. The guidance for placement of caissons and piles is being refined but is likely to conclude that driven piles may be used in all soil units and even in some places where the pile may encounter a High Impedance Reflector (see Section 2.3). It is also likely that suction caissons can be used even where there are over consolidated clays near to the seabed, but only where HIRs are absent.

3

ENGINEERING GUIDANCE

3.1 Field layout planning Experience has shown that one of the most effective ways of communicating the rationale behind geohazard mitigation is by use of simple flowcharts such as the one shown in Figure 4. It presents the stepwise guidance for mitigation by avoidance or mitigation by design (with or without additional data collection) with respect to first geohazards (Stage 1) and then geotechnical conditions (Stage 2). The guidance is modified for different seabed structures and foundation types.

3.3 3.2 Suction caissons and piles near salt As our understanding of the effects of salt diapirsm on soil properties develops, engineering guidance for facilities planning and foundation design is evolving to reflect that. The effect of this is that the guidance provided to the facilities engineers becomes less generic, allowing greater flexibility for layouts and the use of different foundation types. Initially the uncertainties in soil properties of some of the older materials led to a conservative strategy of avoidance. This meant that it could not be relied upon that foundations could be installed through certain seismostratigraphic horizons. Now that the ground model is better-calibrated with deeper geotechnical samples, avoidance is no longer the only option. Figure 5 shows the foundation guidance when geotechnical data from only piston cores and seabed CPTs were available. For Soil Provinces 1 and 2, which


Suction caissons and piles in soils modified by hydrocarbons

High Impedance Reflectors (HIRs) were initially thought to pose a number of constraints for facilities and for the suitability of the deep foundation options being considered. The nature of this geohazard was investigated to a limited extent during a reconnaissance site investigation. However the geotechnical seabed tools specified were not designed to penetrate strong materials to sufficient depth to provide data for deep foundations. Geotechnical boreholes integrated with downhole wireline logging were performed at strategic locations (selected using the chirp sub-bottom data) to calibrate the ground model and refine engineering guidance relative to the HIRs. The HIRs have seabed manifestations which are uneven mounds up to 2 m in height, closely spaced in some areas and as discrete features elsewhere. Subsurface HIRs range from one to a series of tabular

which may be problematic for pile driveability, and suction caisson installation. The HIRs are characterized by high CPT cone resistance values or, where sampled, carbonate-rich claystone. Such layers are fully expected to offer increased penetration resistance to driven piles or suction-installed caissons. In the study, HIR hardness and thickness were varied parametrically within credible bounds that were informed by the site investigation and a geological/geochemical review of the material. This led to a maximum modelled thickness and hardness of 2.0 m and 90 MPa respectively. In this context, ‘hardness’ is represented as equivalent CPT cone resistance. Pile driveability results showed a strong dependency on the depth at which the HIR is encountered. For example, assuming the lowest expected background soil strength profile typical of Soil Province 4, an 84-inch pile can be driven through a 0.6 m thick, 90 MPa hard HIR with a Menck MHU 270T hammer, if encountered at 20 m depth. However, if the same pile encounters the same HIR at 70 m depth, it cannot be driven through the layer. This prediction reflects the increasing amount of driving energy taken out of the pile by the soil as it penetrates, such that when driven to 70 m, there is insufficient remaining energy reaching the tip to advance it through the HIR. Predicted driveable combinations of HIR thickness and hardness can be represented graphically (Fig. 6). For suction caisson installation, the study was extended to consider inclination of the HIR, and it was shown that significant problems may be experienced should the caisson encounter an inclined or localised HIR. For example, should a 5.0 m diameter caisson encounter an inclined, moderate strength HIR of 10 MPa at a depth 1.0 m below seafloor, a HIR thickness of as little as 5 cm may cause the caisson to tip during installation. However, if a similarly hard HIR is encountered at 5.0 m depth, then the additional fixity of the caisson at this deeper penetration means it can tolerate a greater HIR thickness before tipping, in this case 20 cm or greater.

horizons conformable with the bedding, to those which cross-cut bedding and are more chaotic in nature. There are not always direct correlations between subsurface and surface features, and therefore seabed mapping cannot be used to determine the presence or absence of the sub-surface HIR. The materials forming the HIR are inferred to be varying degrees of the host soil, gas, oil and asphalt, and carbonate-rich cemented hard clay, and are found as both continuous horizons and discontinuous nodules. Geophysical surveys and geotechnical investigations revealed variable characteristics and properties, ranging from laterally continuous indurated features (causing CPT refusal), to undulating features laterally discontinuous over short distances.The hard layers targeted and encountered during the geotechnical site investigations were between 0.1 m and 0.7 m thick. These hard materials are thought to be authigenic carbonates formed by two primary geological and physicohemical processes: 1. Depositional carbonate formation in the older sediments as alternating layers of relatively carbonaterich and low carbonate sediments, deposited and progressively buried. Diagenesis concentrated the carbonate into layers and cementing the soils into tabular structures to varying degrees depending upon the extent of the diagenesis; 2. Post-depositional carbonate formation associated with the migration and intrusion of hydrocarbonrich fluids into the host soils. This brought about suitable geochemical conditions to preferentially precipitate carbonate within the host sediment and develop the chaotic and laterally discontinuous cemented features observed. These conditions are thought to be driven to some extent by the relatively dynamic hydraulic environment in the vicinity of the salt diapirs. The HIRs pose a number of uncertainties for piled foundations, both for installation and in-place performance. Where these features occur, the soils are either significantly altered through geochemical processes and/or affected by hydrocarbons, modifying their engineering behaviour. Guidance on what types of foundations can be installed where depends on the foundation type and size, and the relative depth and characteristics of the HIR at any given location. The guidance provided prior to the detailed investigation of the HIRs in the boreholes was: all foundation types should avoid seabed manifestations of HIRs; suction-installed caissons should avoid penetrating any subsurface HIR; and driven piles should avoid penetrating any subsurface HIRs unless there is sufficient confidence in the ground model at the proposed pile location. This may require more detailed interrogation of the geophysical data and/or acquisition of additional geotechnical information. An engineering study was designed to enable combinations of thickness, hardness and depth of HIRs

3.4

Among the geotechnical challenges was the possibility of seabed slope instability, and the potential impact this may have on development infrastructure. As discussed above, the seabed morphology in the development area has been strongly influenced by underlying salt movement. This has caused locally steep terrain and topographic highs and has also contributed to the generation of a steep-sided, narrow trough feature (Fig. 7). Further details of this feature are provided in Hill et al. (2010). While these underlying salt processes are considered static within a life-of-field timescale, the legacy of a locally steep terrain together with very soft seabed soils means that zones of the seabed may be susceptible to shallow instability. A study was therefore commissioned in order to identify routing and placement options within the


Slope stability

Figure 7. Seabed renderings showing topographic features. Figure 6. Graphical representation of driveable combinations of HIR thickness and hardness, soil province 4 soil strength profile, Menck MHU 270T hammer, 84” pile.

development area which would help minimise the exposure of the development infrastructure to shallow slope failures. The narrow trough is of particular importance to the project, as pipelines running between two drill centres will be required either to detour around it, or to traverse it. The latter option represents a value opportunity in terms of reduced material and installation cost, but this must be balanced against the possible increased risk of damage from steep trough-side instability. The slope instability study was performed in two phases, and a description of the main study elements within each phase is presented below. In the first phase, a soil model was developed to map the variation of shallow soil strength over the proposed development area, based on a correlation between observed soil strength and interpreted seismic horizon depths. This allowed a spatially-resolute and continuous soil strength model, which avoids the mapped strength discontinuities otherwise associated with a purely soil province-based model. A GIS-based numerical modeling exercise was used to perform a site-wide slope stability analysis. The analysis was premised on infinite slope theory incorporating a shallow slab-type failure mode. Although this was deemed to be the most appropriate geomechanical model, the validity of this assumption was tested by performing comparative analyses based on a deeper rotational failure mode. The GIS-based infinite slope analysis was also informed by seabed slope derived from AUV high resolution bathymetry, as well as soil submerged unit weight and the influence of external loading from the on-bottom weight of various facilities. This first phase yielded a deterministic estimate of instability potential, in the form of estimated Factor


of Safety (FoS), mapped over the study area. In order for the project to understand and manage the instability risk more fully it was decide to advance this deterministic assessment into a probabilistic stability assessment (PSA) to provide estimates of annual probability of slope failure. The first step in the PSA was to quantify the bias and uncertainty in the predictive soil strength model, as well as in the other stability calculation inputs. For soil strength, which is arguably the most influential parameter, this uncertainty was investigated by comparing model-predicted values with actual values measured from geotechnical borehole data or interpreted from seabed in situ tests. A statistical examination of predicted versus measured values at discrete locations gives a probabilistic strength distribution, which is essentially a measure of the ability of the regional soil model to predict the equivalent strength at any location had it been measured using conventional geotechnical methods (Fig. 8). For routine foundation design, the data in Figure 8 would be used to modify the probability density function of the predicted operational undrained shear strength to achieve the same level of design reliability that would be obtained using location-specific measured soils data (Gilbert et al. 1999). However, the data in Figure 8 were used directly in Monte Carlo simulations for initial slope stability screening purposes. The infinite slope equation was again used, but instead of using fixed, deterministic input values, they were chosen at random in each trial from the predefined measured strength (or other input parameter) distributions. The second phase also considered earthquake loading, the effect of which was modelled as an additional, quasi-static downslope load.The PSA allowed annual slope failure probability, under both gravitational-only and earthquake loading to be mapped over the study area (Fig. 9).

4

CONCLUSIONS

This paper has shown that a properly constructed ground model can be used to make informed decisions about the preferred location of foundations and how initially conservative assumptions can be changed with increasing resolution of the ground model. One of the most important applications of this model for offshore Angola was to inform facilities engineers of acceptable locations for installation of foundations and anchors in sediments affected by salt and hydrocarbon intrusion. The same ground model has also been used to decide on the routing of flowlines to lessen the impact of certain geohazards, such as slope stability. 5

Figure 8. Frequency distribution of ratio of measured (su(m)) to predicted (su(c)) shear strength.

CLOSING REMARKS AND ACKNOWLEDGMENTS

This paper represents the views of the authors and not necessarily those of the companies they represent. The authors wish to thank BP Angola, their development partners and Sonangol for permission to publish this material. They are also grateful for the critical reviews, guidance and stimulating conversation provided by Mike Sweeney and other members of the Exploration and Production Technology team at BP. REFERENCES

Figure 9. GIS map showing annual probability of slope failure.

A key outcome of the overall study is that, whereas the deterministic analysis suggested it would be difficult to traverse the narrow trough feature whilst maintaining a FoS magnitude considered as ‘high’ by conventional standards, the PSA suggests that more direct traversing routes exist where the seabed has a very low estimated probability of failure. Also included in the second phase was an assessment of the impact of landslide-related soil movement on installed pipelines and flowlines, and on the residual seabed topography (freespan potential). For a given slope failure probability, the engineering consequence in terms of induced pipeline stress and deformation could then be anticipated.


Dendani, H., Colliat J.L., Puech, A. & Nauroy, JF 2010. Gulf of Guinea deepwater sediments: geotechnical properties, design issues and installation experiences Keynote Paper, Proceedings 2ndInternational Symposium Frontiers in Offshore Geotechnics Perth, November 2010. Evans, T.G. 2010. A Systematic Approach to Offshore Engineering for Multiple-Project Developments in Geohazardous Areas. Keynote paper, 2nd International Symposium Frontiers in Offshore Geotechnics, Perth, November 2010. Hill, A. J., Southgate, J.G, M., Fish, P. & Thomas, S. 2010. Deepwater Angola Part I: Geohazard Mitigation, 2nd International Symposium Frontiers in Offshore Geotechnics, Perth, November 2010. Gilbert, R., Gambino, S. & Dupin R. 1999. Reliability-based approach for foundation design without site-specific soil borings, Offshore Technology Conference, Houston, Texas, OTC 10927.


Shallow gas hazard linked to worldwide delta environments S. Kortekaas, E. Sens & B. Sarata Fugro Engineers B.V., Leidschendam, The Netherlands

ABSTRACT: Shallow gas hazard assessments are normally based on geophysical data, but when these are limited or unavailable, the geological setting can be used to predict whether shallow gas may be present. A GIS hazard map was created using geological settings, because they control local formation of biogenic gas, which is the most common gas (as defined by origin) in shallow sediments. The highest probability of shallow gas occurrence is in shallow marine settings with high terrestrial influx, deltas in particular. Therefore, a delta classification system based on shallow gas hazard was created and was verified using known gas occurrences. Together the hazard map and delta classification system can be an early screening tool for shallow gas hazard to provide input for risk assessments regarding open-hole or riser-less offshore drilling operations, even when no other relevant data are available.

1

INTRODUCTION

Shallow gas, which occurs at depths less than 1000 m below seafloor (Floodgate & Judd 1992), may pose a hazard to offshore open-hole or riser-less drilling operations, such as geotechnical drilling or drilling of the tophole section of oil and gas wells. For this reason, shallow gas hazard assessments are routinely performed prior to geotechnical drilling (Kortekaas & Peuchen 2008). Normally, hazard assessments are based on indications of gas accumulations in geophysical data. When limited or no geophysical data are available, the geological setting can be used to predict whether shallow gas is present. Information on the geological setting may also be instructive in a shallow gas hazard assessment because it can help the user assess the possibility that there is non-pressurised gas or gas in solution. Neither may be visible on seismic data, but both can pose a hazard to geotechnical drilling. Swabbing pressures produced by retrieval of a sample tool or drill string may cause exsolution of gas and induce gas flow or a gas kick (Kortekaas & Peuchen 2008). In this paper, relationships between depositional environment and the probability of shallow gas occurrence are discussed. This information was used to create a map of the probability of shallow gas accumulations based on depositional environment. The map could be used as an early screening tool when planning open-hole or riser-less drilling in these areas. For example, a user might be a project planner whose first act would be to locate the proposed drilling site on the map. If the proposed site lies within an area with a low probability of shallow gas hazard, the planner could relatively confidently proceed with drilling as planned with standard safety equipment and preventative procedures. However, if the site is in a zone of

intermediate or high probability of shallow gas, his/her second act would be to request further information and a detailed risk assessment.A detailed assessment could indicate whether drilling a pilot hole and other mitigation measures are sufficient (Kortekaas & Peuchen 2008), or the drilling site should be relocated.

2

There are two different types of shallow gas as defined by origin. The first is thermogenic gas, which forms at depth under high temperatures and pressures. It may be present in the shallow subsurface when it has migrated up from a deeper reservoir (Floodgate & Judd 1992). The second is biogenic gas, which forms at shallow depths through bacterial activity. Biogenic gas is by far the most common gas in shallow sediments (Lin et al. 2004). Thermogenic gas can migrate upward along natural pathways, through porous strata or along faults, or along leaking wells. Thermogenic gas may pose a hazard to drilling operations, especially when drilling close to existing wells. Nevertheless, it was excluded from this study because the focus was present-day depositional environments, which govern conditions of the formation of only local, biogenic gas. Biogenic gas formation requires a sufficient supply of organic matter and a rapid sedimentation rate to bury organic material before it is oxidised. The gas accumulates only when it can migrate in a free gas phase (Rice 1993). This occurs when the concentration in the pore fluid exceeds gas solubility, or when gas exsolves due to reduction of hydrostatic pressure, which could be caused by erosion of the seabed or a fall in relative sea level.


DEPOSITIONAL ENVIRONMENT AND SHALLOW GAS

The preservation of shallow gas accumulations requires a reservoir, a seal and gas. In the shallow subsurface, reservoirs are most commonly formed by coarser-grained materials such as sand, and seals by fine-grained sediments such as clay. For simplicity, we refer to potential reservoir and seal materials as, respectively, sand and clay in this paper. Given the requirements described above, it can be concluded that shallow biogenic gas is most likely to occur in environments with sand and clay deposits, an influx of organic material and a rapid sedimentation rate. The combination of these factors is most probable on shallow ocean shelves, where a terrestrial influx of sand and organic material can be expected, and, in particular, in delta environments. 3

HAZARD MAP AND DELTA CLASSIFICATION

First, bathymetric data were loaded into a Geographic Information System (GIS). A map created from these data displays the initial, coarse hazard zonation by bathymetry. Most of the ocean was defined as deep sea, where either the water depth exceeds 1500 m, or the water depth is between 1000 m and 1500 m and the seabed slope is less than 1◦ . The continental slope and rise were grouped together for the purposes of the shallow gas hazard map. The slope and rise were defined as the area between 600 m and 1000 m water depth and the area between 150 m and 1500 m water depth where the slope exceeds 1◦ . The continental shelf is the zone between the continental slope and the shoreline. The water depth is generally 0 m to 150 m, but can reach 550 m. In this study, the shelf was defined as the area from the coastline to 150 m water depth and the zone between 150 m and 600 m water depth where the slope is less than 1◦ . Shallow shelf areas where terrestrial influx is high are more likely to contain sand, clay and gas than the deep sea. Given that these conditions are most common in deltas, information about 230 deltas was collected and placed in GIS. The information included: annual river discharge (Hovius 1998, Meybeck & Ragu 1995, Milliman et al. 1995, Orton & Reading 1993, Syvitski et al. 2005) and annual sediment load before damming (Meybeck & Ragu 1995, Milliman & Syvitski 1992); the area of the drainage basin (Beusen et al. 2005, Hovius 1998, Meybeck & Ragu 1995, Milliman et al. 1995, Syvitski et al. 2005); the net slope of the drainage area (calculated from topographic information); the climate in the delta and the climate in the drainage area (both inferred from global climate maps (The Times atlas of the world (1999)); and the presence of known oil and gas fields. As stated above, sand and clay deposits are the most likely combination of materials to provide a reservoir and seal for the accumulation and preservation of shallow gas. However, it is difficult to infer the ratios of sand and clay deposition in deltas from available

datasets. Existing delta classifications are not suitable to predict the presence of sand and clay deposits (e.g. Galloway 1975, Postma 1990, Orton & Reading 1993). Most delta deposits are likely to contain sand and clay, therefore, this analysis has focussed on the factors critical for gas formation: sedimentation rate and the supply of organic material to the delta. To assess the effect of these two factors on the probability of shallow gas presence in a given delta, two datasets were selected for the delta classification. 1. The annual sediment load (before human influence) was assumed to approximate the influx to the delta and to be the most important factor determining the sedimentation rate in the delta. A high sedimentation rate is critical for the formation of shallow gas accumulations because it facilitates burial of organic material before it can be oxidised. 2. The climate in the drainage area affects the amount of organic matter that is transported to the delta. A greater influx of organic matter to deltas of warmer, more humid drainage basins would be expected because these areas would be more biologically productive. Deltas with annual sediment loads greater than 0.1 million tons per year were examined, thus 199 deltas were included. The deltas were categorised by the present-day climate type dominant in their drainage basins (temperate humid, hot humid, cold humid and cold dry). Next, the deltas within each of the four climate types were classified by annual sediment load, also into four groups (Table 1). The deltas examined in this study are represented in Figure 1 by circles. Larger circles denote greater sediment loads. Please note that rather than create a fifth group with only one delta, the Ganges Delta was included in the group with the greatest sediment load though its annual load exceeds a trillion tons per year. To verify the classification, known occurrences of shallow gas from literature (Fleischer 2001, GarciaGarcia et al. 2007) and from Fugro experience were used and are represented by stars in Figure 1. Most published locations of shallow gas occurrences are known from seismic surveys, degassing seabed (observed visually and through acoustic surveys), gas escaping from samples, and from features such as pockmarks on the seabed. The sources (biogenic or thermogenic) Table 1. Number of deltas included in this study, divided by annual sediment load and dominant climate in their drainage basins. *For simplicity, the Ganges Delta is included in this group though the annual sediment load is greater than a trillion tons per year.


Sediment load (106 ton/year)

Hot humid

Temperate humid

Cold humid

Cold dry

0.1 to 1 1 to 10 10 to 100 100 to 1000*

9 13 21 13

21 37 38 8

5 4 7 0

8 8 6 1

Figure 1. Locations of deltas examined in this study (circles) and reported shallow gas (stars). Circle size reflects relative sediment load.

of most of these shallow gas occurrences are known. Only biogenic gas occurrences were included in the verification of the delta classification. In contrast, the sources of most of the Fugro shallow gas occurrences are unknown. Therefore, it is possible that gas reported as biogenic gas in some deltas was, in fact, thermogenic gas that leaked up from deeper hydrocarbon reservoirs. Figure 2 shows the deltas subdivided by climate of their drainage basins and classified according to their annual sediment loads. For each sediment load class, the proportion of deltas in which shallow gas has been encountered is indicated. 4

RESULTS AND DISCUSSION

4.1 Delta classification Comparison of the results of the delta classification with known gas occurrences shows that shallow gas is most likely to occur in deltas with drainage areas in temperate humid and hot humid climates (Table 2). Greater biological production could be expected in such regions, and thus, by extension, more organic material transported by fluvial and overland drainage to these deltas. The probability of gas presence is generally higher in deltas of river systems with high annual sediment loads, which in general can be correlated to the large delta systems. A high annual sediment load correlates to a high sedimentation rate in the delta and thus a greater probability that organic material is buried and available for gas formation. Annual sediment load of a river is directly related to the size of the drainage basin: the larger the drainage area, the higher the sediment load (Milliman & Syvitski 1992). 4.2 GIS and the hazard map A GIS map created from selected data displays the initial, coarse hazard zonation by bathymetry, which


Figure 2. The number of deltas with (black) and without (grey) reported shallow gas by annual sediment load, displayed by climate in the deltas’ drainage areas. Note that vertical scales differ.

essentially correlates with proximity to the coast and expected terrestrial input. A detail from the map is shown in Figure 3. The map includes deltas investigated in this study, annotated by sediment load and the climate in the drainage basin, both of which indicate the likelihood of shallow gas accumulation. Locations of known gas hydrates also appear on the map (Mazurenko & Soloviev 2003). Gas hydrates are not directly hazardous to geotechnical drilling; however, they may indicate the presence of free gas below the gas hydrate stability zone. This map can be used as a first screening tool of the shallow gas hazard in an area where drilling is proposed. 4.3

Although large amounts of organic material are not expected to be transported from cold dry drainage basins, gas was encountered in the large deltas with drainage basins in such climates. These may be older deltas, where in the past, climate conditions were different (i.e. warmer and/or more humid) and more favourable for shallow gas formation. It is important to note that palaeodeltas were not considered individually in this study. Nonetheless, as they are generally located on the shelf, their influence on shallow gas hazard probability is indirectly incorporated in the coarse hazard zonation by bathymetry. The probability of encountering shallow gas accumulations is also affected by wave and current energy. In high energy environments, sediment reworking prevents quick burial of organic material and deposition of clay, which is usually necessary to form seals. This has not yet been investigated in detail.

Data limitations

Known shallow gas occurrences (stars in Figure 1) appear to be concentrated in the northern hemisphere, especially around Europe and North America. The distribution is most likely more even than it appears. The apparent concentration in the northern hemisphere likely reflects the fact that more work has been done in these areas. Table 2.

Studied deltas by climate in drainage area.

Climate

Number of deltas

Deltas with reported shallow gas (%)

Hot humid Temperate humid Cold humid Cold dry All deltas

56 104 16 23 199

23 31.5 12.5 4.5 24.5

5

CONCLUSIONS

To generate and accumulate shallow gas in marine sediments, organic material, sand, clay and rapid deposition rates are usually necessary. All these are most commonly found in shallow seas (on the ocean shelves), and in delta environments, in particular. A classification of deltas by climate in the drainage area and annual sediment influx (river load before human influence) provides information about shallow gas hazards. The highest probability of shallow gas presence is in deltas with drainage basins in temperate humid and hot humid climates and in deltas, which

Figure 3. Part of the global GIS map, showing bathymetry zones, deep sea, continental slope/rise and continental shelf in different shades of blue. Climate of the delta drainage basin is indicated by the colour of the circle and the size corresponds to sediment load. Stars represent locations of reported shallow gas (black) and gas hydrates (yellow).


receive high annual sediment loads. A GIS map of shallow gas hazard was created from the results of this study. This map can be used as a first screening tool for shallow gas hazard when planning geotechnical drilling operations, to provide input for risk assessments and to inform decisions regarding offshore drilling operations, even in areas for which no other relevant data are available. ACKNOWLEDGEMENTS The authors gratefully acknowledge Fugro’s commitment to improving safe and sustainable practice and supporting the work presented in this paper. Thanks are given to Poppe de Boer and Joek Peuchen for their valuable input. We also thank two anonymous reviewers for their comments, which helped to improve this paper. The opinions expressed in this paper are those of the authors and they are not necessarily shared by Fugro. REFERENCES Beusen, A.H.W., Dekkers, A.L.M., Bouwman, A.F., Ludwig, W. & Harrison, J. 2005. Estimation of global river transport of sediments and associated particulate C, N, and P. Global Biogeochemical Cycles 19(4): GB4S05.1GB4S05.17. Fleischer, P. Orsi, T.H., Richardson, M.D. & Anderson, A.L. 2001. Distribution of free gas in marine sediments: a global overview. Geo-Marine Letters 21(2): 103–122. Floodgate, G.D. & Judd, A.G. 1992. The origins of shallow gas. Continental Shelf Research 12(10): 1145–1156. Galloway, W.E. 1975. Process framework for describing the morphologic and stratigraphic evolution of deltaic depositional systems. In Broussard, M.L. (ed.), Deltas: models for exploration: 87–98. Houston: Houston Geological Society. García-García, A., Tesi, T., Orange, D., Lorenson, T., Miserocchi, S., Langone, L., Herbert, I. & Dougherty, J. 2007. Understanding shallow gas occurrences in the Gulf of Lions. Geo-Marine Letters 27(2-4): 143–154. Hovius, N. 1998. Controls on sediment supply by large rivers. In K.W. Shanley & P.J. McCabe (eds.), Relative role

of eustasy, climate, and tectonism in continental rocks: 3–16. SEPM Special publication 59. Tulsa: Society for Sedimentary Geology. Kortekaas, S. & Peuchen, J. 2008. Measured swabbing pressures and implications for shallow gas blow-out. In OTC.08; Proceedings 2008 Offshore Technology Conference, 5–8 May 2008. Houston, Texas, USA. OTC Paper 19280. Houston: Offshore Technology Conference. Lin, C.M., Gu, L.X., Li, G.Y., Zhao, Y.Y. & Jiang, W.S. 2004. Geology and formation mechanism of late Quaternary shallow biogenic gas reservoirs in the Hangzou Bay area, eastern China. AAPG Bulletin 88(5): 613–625. Meybeck, M. & Ragu,A. 1995. River discharge to the oceans: an assessment of suspended solids, major ions and nutrients. Environmental Information and Assessment Report. Nairobi: United Nations Environment Programme. Mazurenko, L.L. & Soloviev, V.A. 2003. Worldwide distribution of deep-water fluid venting and potential occurrences of gas hydrate accumulations. Geo-Marine Letters 23(3–4): 162–176. Milliman, J.D. & Syvitski, J.P.M. 1992. Geomorphic/tectonic control of sediment discharge to the ocean: the importance of small mountainous rivers. Journal of Geology 100(5): 525–544. Milliman, J.D. Rutkowski, C. & Meybeck, M. 1995. River discharge to the sea: a global river index (GLORI). LandOcean Interactions in the Coastal Zone Reports and Studies. Den Burg: LOICZ Core Project Office. Orton, G.J. & Reading, H.G. 1993. Variability of deltaic processes in terms of sediment supply, with particular emphasis on grain size. Sedimentology 40(3): 475–512. Postma, G. 1990. Depositional architecture and facies of river and fan deltas: a synthesis. In: A. Colella & D.B. Prior (eds.), Coarse-grained deltas: 13–27. Special Publications of the International Association of Sedimentologists 10. Oxford: Blackwell. Rice, D.D. 1993. Biogenic gas: controls, habitats, and resource potential. In D.G. Howell (ed.), The Future of Energy Gases: 583–606. U.S. Geological Survey Professional Paper 1570. Washington: United Sates Government Printing Office. Syvitski, J.P.M., Kettner,A. J., Correggiari,A. & Nelson, B.W. 2005. Distributary channels and their impact on sediment dispersal. Marine Geology 222–223: 75–94. The Times atlas of the world. 1999. Tenth Comprehensive Edition. Crown Publishers, New York.



Analysis of submarine flow slides in fine silty sand P.V. Lade The Catholic University of America, Washington, D.C., U.S.A.

J.A. Yamamuro Oregon State University, Corvallis, OR, U.S.A.

ABSTRACT: The mechanism of instability in granular soils is briefly explained, and its requirement as forerunner for liquefaction of level or sloping ground is described. Tests on loose silty sand indicate a ‘reverse’ behavior with respect to confining pressure and as opposed to the behavior of loose, clean sands. Strong correlations between fines content, compressibility and liquefaction potential are found for these soils. A procedure for analysis and evaluation of static liquefaction of slopes of fine sand and silt such as submarine slopes is presented. It involves determination of the region of instability in stress space in which potential liquefaction may be initiated and determination of the state of stress in the slope. The instability line and the region in which potential liquefaction may be initiated can be determined from consolidated-undrained triaxial compression tests, and a method of finding the state of stress is developed to predict the zone of potential liquefaction in simple slopes. 1

INTRODUCTION

Recent experiments involving loosely deposited silty sands dispute the assumption that clean sands always behave similar to silty sands. The tests on loose silty sand indicate a ‘reverse’ behavior with respect to confining pressure and this violates the basic assumption that loose, silty sands behave similar to loose, clean sands. There is a strong correlation between fines content, compressibility and liquefaction potential of these soils. A procedure for analysis and evaluation of static liquefaction of slopes of fine sand and silt such as submarine slopes, mine tailings, and spoil heaps is presented. This procedure involves determination of the region of instability in stress space in which potential liquefaction may be initiated for the soil in question and determination of the state of stress in the slope. It is explained how the instability line and the region in which potential liquefaction may be initiated can be determined from consolidated-undrained triaxial compression tests, and a straightforward method of finding the state of stress is employed to predict the region of potential liquefaction in simple slopes.

2

REVERSE BEHAVIOR OF SILTY SAND AT LOW CONFINING PRESSURES

Figure 1. Drained triaxial tests on 12% relative density silty Nevada sand at confining pressures from 26 to 600 kPa.

Within a range of low confining pressures, drained triaxial tests on very loose silty sand with high compressibility show negligible effect of magnitude of confining pressure on the contractive volume change, as shown in Figure 1 (Yamamuro and Lade 1997). The corresponding undrained tests in the same range of

confining pressures show development of essentially equal pore pressures, as indicated in Figure 2, and consequently the effective confining pressures reach zero faster with decreasing initial consolidation pressures. Thus, the lower the initial consolidation pressure


227

Figure 2. Effective confining pressures during shearing from tests on 12% relative density silty Nevada sand.

Figure 4. Effects of fines content and void ratio on volume compressibility and static liquefaction in experiments on silty Nevada sand.

temporary liquefaction. This second region is characterized by an initial peak stress difference, followed by a decline. As shearing continues the stress path crosses the phase transformation line into the region of dilation and pore pressure decline, resulting in stress differences increasing to much higher magnitudes than the initial peak. In this region the specimens show increasing dilatancy with increasing initial consolidation pressure, contrary to conventional sand behavior. The following two regions of temporary instability and instability are those recognized from conventional sand behavior. Figure 3. Four distinctly different general types of undrained effective stress paths for loose silty sands: static liquefaction, temporary liquefaction, temporary instability, and instability shown in p’-q diagram.

the faster liquefaction conditions are reached in the specimens. This clearly shows that static liquefaction is a low-pressure phenomenon. The effect of increasing the confining pressure is to increase the resistance to liquefaction. This behavior is contrary to observed behavior for conventional undrained tests on clean sands (Seed & Lee 1967). This ‘reverse’behavior observed for very loose silty sand at low confining pressures is accompanied by an inflection in the instability line, as seen (exaggerated) in the schematic diagram in Fig. 3 (Lade and Yamamuro 1997). Four distinctly different types of effective stress paths with corresponding behavior patterns are shown. Static liquefaction occurs at the lowest pressures, and it is characterized by large pore pressure developments that result in zero effective confining pressure and zero stress difference at low axial strains. In this range, the maximum effective friction angle increases with increasing effective confining pressure and it continues to increase through the following region of © 2011 by Taylor & Francis Group, LLC

3

COMPRESSIBILITY AS A MEASURE OF LIQUEFACTION POTENTIAL

This pattern of ‘reverse’ sand behavior is entirely controlled by the very high compressibility of the very loose silty sand. The compressibility is in turn controlled by the amount of fines present in the sand. Figure 4 shows a correlation between the fines content, void ratio, volume compressibility and static liquefaction observed in experiments on Nevada sand (Yamamuro and Lade 1998). The “wall” between the region of stable behavior and the liquefaction regime corresponds to almost constant compressibility. Thus, if the compressibility is higher than this wall, i.e. in the approximate range from (1.4–2.2) · 10−5 (1/kPa), then liquefaction can occur under undrained conditions. Similar results were obtained by Lade et al. (2009) from experiments on Ottawa sand mixed with Loch Raven silt. For this silty sand liquefaction may occur under undrained conditions for compressibilities higher than (1.2–1.6) · 10−5 (1/kPa), which are very similar to the compressibilities for Nevada sand. The development of pore pressures under undrained conditions is directly related to the compressibility of the soil, and loose silty sands exhibit significant volumetric contraction at low pressures. Yamamuro and

228

Lade (1998) proposed to use volumetric compressibility as an alternative indicator of liquefaction potential for silty sands. Volumetric compressibility does not require determination of void ratio and fines content, and it can be measured in many different ways. For example, the last increment in isotropic compression before undrained shearing in a triaxial compression test may be used to obtain the volumetric compressibility. It may also be obtained from a standard oedometer test at the appropriate low stress magnitude, where the liquefaction potential is highest. Furthermore, it may be obtained from an in-situ test on the intact soil by inserting a screw-plate to relatively shallow depth and performing a plate load test to determine the vertical compressibility at the relevant field location. Such an in-situ test also captures the very important effect of the soil fabric or structure (Wood et al. 2008; Yamamuro et al 2008) and avoids the difficult to impossible task of having to recover intact samples of the loose, silty sand for testing in the laboratory. Alternatively, a pressuremeter test may be employed to determine the compressibility in the horizontal direction. Whether the vertical or the horizontal compressibility is more relevant for indication of liquefaction potential remains to be seen. In addition, the actual boundary conditions in the field are important for determining the liquefaction potential in the field. From knowledge of constitutive modeling of soils, it is logic that the volumetric compressibility is one of the significant factors that control the development of pore pressures under undrained conditions. The fact that this property of a loose, silty sand deposit may be determined in-situ by a screw-plate test (vertical compressibility) or by a pressuremeter test (horizontal compressibility) at relatively shallow depths may make determination of liquefaction potential relatively easy. Besides, the fact that all significant factors that influence the volumetric compressibility are already present in the field deposit further increases the importance of such in-situ tests in indicating liquefaction potential.

4 ANALYSIS PROCEDURE FOR STATIC LIQUEFACTION To analyze a slope for its potential for static liquefaction, the states of stress everywhere in the slope are compared with the states of stress in the region of potential liquefaction. According to the recently discovered ‘reverse’ behavior, the critical region reaches all the way down to the stress origin, where true liquefaction occurs. In the following step of the instability method, the state of stress in the slope is superimposed on the stress diagram to find out if any stress states overlap with the region of potential instability. In such overlapping regions, any point is a point of potential instability, and instability will develop under undrained conditions if a suitable trigger mechanism is activated to initiate the © 2011 by Taylor & Francis Group, LLC

Figure 5. Location of region of potential instability and subsequent liquefaction in p’-q diagram.

instability. If the region of potential instability reaches down to the stress origin, then the ground may liquefy.

5

EXPERIMENTS TO DETERMINE THE LIQUEFACTION REGION IN STRESS SPACE

The region of potential instability is located between the instability line and the failure line, and it reaches down to the stress origin for liquefaction to occur, as indicated schematically in Figure 5. The instability line connects the tops of the yield surfaces, and it is a straight line through the stress origin (Lade 1992). The effective stress paths from undrained tests on very loose compressible soil essentially trace the yield surfaces, and determination of the top of one such yield surface is, in principle, sufficient to define the instability line. Thus, the instability line may be characterized by its inclination, similar to a friction angle. However, it should be understood that this inclination cannot be used in a manner similar to a conventional friction angle. For analysis of liquefaction potential, it is necessary to determine the peak shear stress, Sp for which liquefaction can occur, i.e. for which the effective stress path reaches down to the stress origin where true liquefaction occurs, in order to locate the triangular region of potential liquefaction, as indicated in Figure 5. The key difficulty in performance of the undrained tests consists of depositing a silty sand in the laboratory with a structure similar that in the field. Many methods of soil deposition are available for laboratory creation of loose sand deposits, and their effects on the behavior have been investigated in detail (e.g. Ishihara 1993; Vaid et al. 1999, Wood et al 2008; Yamamuro et al. 2008). Each method of deposition produces its own sand fabric and behavior, and these behaviors are very different for different deposition methods. However, it is not known at this time which method is most likely to produce a sand fabric similar to that in the field, simply because the fabric in field deposits of fine silty sand is not known and has not been sufficiently characterized to allow reproduction by an artificial laboratory method. While the difficulty lies in obtaining undisturbed samples of field deposits

229

instability, the addition of the shear stresses may cause the gently inclined slope to have stress states that, on some planes, will reach into the region of potential instability. To determine the required slope inclination for the Mohr circle to become tangent to the instability line, as shown in Figure 6(a), an expression may be developed on the basis of (1) the expressions for the principal stresses from the Mohr circle shown in Figure 6(a), (2) the Mohr-Coulomb failure criterion for sand used with the instability line inclined at φi , (3) the value of K0 = 1-sinφ. The resulting expression for the slope inclination, α, becomes:

Figure 6. (a) Mohr diagram with indication of shear and normal stresses limiting the region of potential instability, and (b) Volume of soil for calculation of stresses along plane parallel with sloping surface.

Thus, for slopes with inclinations smaller than α, the Mohr circle will not reach into the region of potential instability and such slopes will not become unstable under static conditions. For slopes with inclinations greater than α, many planes in the slope will be located in the region of potential instability. For such conditions the zone of potential instability in the submarine slope is also limited by the peak shear stress, Sp , indicated in Figure 5. Thus, for locations with shear stresses less than Sp in the slope, the state of stress is in the region of potential instability. The analysis required to determine this region is simply an infinite slope stability analysis, indicated in Figure 6(b), from which it may be determined that:

to study in the laboratory, such intact samples have been obtained by the freezing method and tested in the laboratory (e.g. Yoshimi et al. 1984, 1989; Vaid et al. 1999). Vaid et al. (1999) recommend using the water pluviation method, because they found that it produces results that are comparable with those on in-situ frozen undisturbed sand specimens from three different sites.

6

ZONE OF POTENTIAL LIQUEFACTION IN SLOPING GROUND

The states of stress in the region of potential liquefaction, shown in Figure 5, can be identified so as to outline a zone in sloping ground, such as in a submarine slope, within which liquefaction may be triggered. Beginning with a slope with moderate inclination, α, the state of stress in the slope may be obtained as shown in Figure 6(a). The state of stress in level ground is obtained from the value of K0 . Thus, σ h ’= K0 ·σ v ’, and the corresponding Mohr’s circle is shown in Figure 6(a). For a gently sloping ground surface, the normal stresses on vertical and horizontal planes may be approximated by those obtained from the K0 stress state. The shear stresses acting on these planes may be obtained from τ = σ v ’·tanα. The corresponding states of stress are represented by the larger Mohr circle in Figure 6(a). This circle represents a reasonable approximation to the real states of stress in the gently inclined slope. While the Mohr circle for the K0 stress state may not reach into the region of potential © 2011 by Taylor & Francis Group, LLC

in which h is the vertical depth below the sloping surface, b is the length considered along the sloping surface, and γ b is the buoyant unit weight of the silty sand. The Mohr diagram in Figure 6(a) shows that the peak shear stress, Sp = (σ 1 −σ 3 )/2, may be expressed in terms of τ in Equation (4):

The vertical depth, hi , to which the zone of potential instability reaches down from the sloping surface may then be determined from the expression in Equation (5):

Thus, the zone in the submarine slope in which instability may be initiated reaches from the surface and vertically down to a depth of hi . Once the

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Note that the resulting zone of potential liquefaction reaches all the way up to the sloping surface and is not bounded by a zone of dilation and stable behavior, as was the zone of instability previously determined (Lade 1992). The reason is the difference between the ‘reverse’ behavior of loose, silty sand and the normal behavior observed for undrained tests on clean sand, as described above. Figure 7. Submarine slope of silty sand with indication of zone of potential instability and subsequent liquefaction.

7

slope inclination increases above α as determined from Equation (1), then the value of hi determines the extent of the zone below the surface of the slope, as indicated in Figure 7. It should be noted that the maximum possible value of the slope inclination is α = ϕ, at which the slope is barely stable against conventional slope stability, which involves comparisons of shear stresses and shear strengths. Therefore, as long as the soil remains drained, it will remain stable. But a small disturbance will cause the silty sand, which has relatively low permeability, to react in an undrained manner and will subsequently cause the slope to become unstable under essentially static conditions. The small disturbance may cause any point within the zone of potential instability, shown in Figure 7, to respond in an undrained manner and trigger the instability and subsequent liquefaction of the slope. A few example calculations are performed for a submarine slope consisting of Nevada sand with 6% fines and a relative density of 12% (Yamamuro and Lade 1997). The friction angle for this silty sand is 33◦ , the instability angle for the isotropically consolidated silty sand is 17◦ , the maximum peak shear stress for a specimen that liquefies is Sp = qmax /2 = 50/2 = 25 kPa, the void ratio after consolidation is approximately e = 0.8, and the buoyant unit weight is therefore γb = (Gs -1)·γ w /(1 + e) = 0.92 g/cm3 . The value of K0 = 1-sinφ = 0.455, which corresponds to a mobilized friction angle of 22◦ . Since this value is greater than the angle of instability ϕi = 17◦ , the sand is potentially unstable under a level ground surface, and all that is required for this soil to become unstable and subsequently liquefy is a trigger to initiate the instability. However, as explained in the Discussion section below, the instability line will be steeper for a K0 consolidated soil than for an isotropically consolidated soil, and in practice it will always be located with greater slope than the K0 -line. Thus, assume for the sake of demonstration that the appropriate instability line is inclined at ϕi = 25◦ . The minimum slope inclination required for potential instability is calculated from Equation (1) to be α = 8.1◦ . At that slope angle the vertical depth of the zone of potential instability may be calculated from Equation (6) to be 17.6 m. As the slope inclination increases above the value of α = 8.1◦ , the instability zone in the slope becomes less deep and at α = ϕ = 33◦ , the vertical depth is hi = 5.4 m. © 2011 by Taylor & Francis Group, LLC

DISCUSSION

The zone of potential instability and subsequent liquefaction comprises a volume of soil parallel with the sloping surface, indicated by the shaded zone in Figure 7, in which any point is on the verge of unstable behavior, and once the instability has been initiated, this is the zone that liquefies first. However, once liquefaction has been initiated, the rapidly increasing pore pressures will cause the pore water to move towards locations with lower pressures, i.e. out of the slope and deeper into the slope. Here the increasing pore pressures will overcome the tendency of the silty sand at greater depth to dilate (according to the ‘reverse’ behavior pattern) and render the deeper dilating soil unstable as well (Lade et al. 1993). Thus, the unstable zone will reach into the slope and tend to create what may look like a concave cavity, as indicated in Figure 7. The difference between conventional slope failure and slope liquefaction is that the conventional slope failure results in a slumping surface in which the soil moves down slope by some small distance, while slope liquefaction produces a scarp with no remaining soil, because the liquefied soil runs far away as a liquid, as observed in the Nerlerk berm failures (Lade 1993).

8

CONCLUSION

A summary of the ‘reverse’ behavior discovered for loose deposits of fine silty sands at low confining pressures is presented. The high compressibility of the loose, silty sands plays an important role in their liquefaction potentials, and it is proposed that this may be identified from in-situ measurements by screw plate or pressuremeter tests on the intact soil in the field. This has the advantage of producing test results for intact soil with the in-situ particle fabric, and it avoids the problems surrounding the recovery of intact samples of soils that are notoriously difficult to sample. Laboratory testing procedures for appropriately deposited specimens to determine the liquefaction region in stress space are reviewed, and an analysis procedure is developed for potential instability and liquefaction of sloping ground resulting in submarine flow slides.

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REFERENCES Ishihara, K. 1993. Liquefaction and flow failure during earthquakes. Geotechnique, 43(3): 351–415.

Lade, P.V. 1992. Static instability and liquefaction of loose fine sandy slopes. Journal of Geotechnical Engineering, ASCE, 118(1): 51–71. Lade, P.V. 1993 “Initiation of static instability in the submarine Nerlerk berm”, Canadian Geotechnical Journal, 30(5): 895–904. Lade, P.V., Bopp, P.A. & Peters, J.F. 1993. Instability of dilating sand. Mechanics of Materials, Elsevier, 16: 249–264. Lade, P.V. & Yamamuro, J.A. 1997. Effects of non-plastic fines on static liquefaction of sands.Canadian Geotechnical Journal, 34(6): 918–928. Lade, P.V.,Yamamuro, J.A. & Liggio, C.D., Jr. 2009. Effects of fines content on void ratio, compressibility, and static liquefaction of silty sand, Geomechanics and Engineering, Techno-Press, 1(1): 1–15. Seed, H.B. & Lee, K.L. 1967. Undrained strength characteristics of Cohesionless soils. Journal of the Soil Mechanics and Foundations Division, ASCE, 93(SM6): 333–360. Vaid, Y.P., Sivathayalan, S. & Stedman, D. 1999. Influence of specimen-reconstituting method on the undrained response of sand. Geotechnical Testing Journal, ASTM, 22(3): 187–195.


Wood, F.M., Yamamuro, J.A. & Lade, P.V. 2008. Effect of depositional method on the undrained response of silty sand. Canadian Geotechnical Journal, 45(11): 1525–1537. Yamamuro, J.A. & Lade, P.V. 1997. Static liquefaction of very loose sands. Canadian Geotechnical Journal, 34(6): 905–917. Yamamuro, J.A. & Lade, P.V. 1998. Steady state concepts and static liquefaction of silty sands. Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 124(9): 868–877. Yamamuro, J.A., Wood, F.M. & Lade, P.V. 2008. Effect of depositional method on the microstructure of silty sand. Canadian Geotechnical Journal, 45(11): 1538–1555. Yoshimi,Y., Tokimatsu, K. Kanoko, O. & Makihara,Y. (1984). Undrained cyclic shear strength of a Niigata sand. Soils and Foundations, 24(4): 131–145. Yoshimi, Y., Tolimatsu, K. & Hosaka, Y. (1989). Evaluation of liquefaction resistance of clean sands based on high-quality undisturbed samples. Soils and Foundations, 29(1): 93–104.

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Hydrate dissociation around oil exploration infrastructure A.K. Sultaniya, J.A. Priest & C.R.I. Clayton University of Southampton, United Kingdom

ABSTRACT: The widespread occurrence of gas hydrates in marine sediments has raised concerns of potential risks to oil production infrastructure, should the hydrate dissociate when pumping hot oil from deeper strata through riser pipes. Hydrate dissociation changes the ice-like structure of hydrate back to its constituent components of gas and water, possibly increasing pore pressure, reducing effective stress, and altering the stiffness and strength of the sediment. To understand sediment behaviour during hydrate dissociation well-controlled laboratory tests on hydrate-bearing sands were conducted at the University of Southampton to mimic temperature in sediment around an oil riser pipe. Measurements were conducted throughout the dissociation stages to determine effective stress and sediment stiffness changes, as well as the associated inherent damping, during undrained and drained conditions. The results will be subsequently used within a FE model to assess the performance and stability of oil exploration infrastructure during hydrate dissociation.

1

INTRODUCTION

Gas hydrates are naturally occurring metastable compounds composed of gaseous molecules encapsulated within a water matrix to form an ice-like structure. The most common gas found within gas hydrates is methane, although gases like CO2 , H2 S, ethane etc are also found in naturally occurring gas hydrates. Gas hydrates exist where there is an ample supply of gas within the sediment, combined with high pressure and/or low temperature conditions. In nature these conditions exist within oceanic sediments on continental margins and deep within sediments in arctic regions below the permafrost. As gas hydrates are metastable they dissociate if temperature or pressure conditions are sufficiently altered. This can change gas hydrate from an ice-like structure back to its constituent parts of gas and water. This will lead to changes in the physical properties of hydrate-bearing sediments. In forming gas hydrates the gas is able to achieve a denser packing (1 m3 of methane gas hydrate contains 164 m3 of methane at Standard Temperature and Pressure (Sloan, 1998)) than it could occupy in its gaseous state, so dissociation may result in an increase in pore pressure in the sediment. Oil and gas exploration activities have begun to extend to significant water depths (greater than 1000 m) where gas hydrates are known to exist. Pumping of hot oil or gas through the hydrate-bearing sediments may dissociate hydrate, resulting in heave or subsidence around oil/gas wells, depending upon the permeability of the adjacent layers. This can lead to casing failure or at the extreme platform subsidence.


This phenomenon has been observed in oil drilling activities by many researchers (Makogon, 1981; Nimblett et al., 2005; Tan et al., 2005). Hydrate dissociation, induced through changes in sea level or rise in ocean bottom temperature, has been associated with large seafloor failure in the geologic past causing devastating tsunamis and widespread flooding (Carpenter, 1981; Kayen & Lee, 1991; Padden et al., 2001). Numerical modeling of sediment behaviour can be used to asses the impact of drilling activities on sediment behaviour. Understanding and mitigating the risk of drilling through hydrate-bearing sediments, using numerical models, can only be effective if detailed knowledge of sediment behaviour during dissociation is available. At present the influence of hydrate dissociation on sediment behaviour is not well understood. Gas hydrate dissociation characteristics can be derived either through field testing, laboratory testing on recovered or testing laboratory prepared gas hydrate samples. However, since gas hydrate exists in deep-water oceanic sediments, or deep permafrost sediments, it is often impractical to perform field testing and difficult to obtain undisturbed in-situ samples for testing (Priest et al., 2008). This paper reports on a series of tests conducted on methane hydratebearing sands to determine the physical properties of these sands during formation and dissociation of gas hydrate. Physical properties included measurements of the small strain stiffness as well as the respective damping ratios under low frequency conditions relevant to seismic geological testing. Factors such as the stress conditions during formation and dissociation were also investigated

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2

LABORATORY APPARATUS

There are a number of different laboratory techniques that have been utilized to determine the physical properties of soils. One such apparatus that is routinely used is the resonant column, which allows estimates of the small strain stiffness of soils and its respective material damping. Generally resonant column testing is performed under low stress conditions (1 MPa) of pressure and at room temperature. But, methane hydrate exists under restricted thermobaric conditions and so pressures (up to 20 MPa) and low temperatures (down to −20◦ C) are required to allow effective formation of hydrate in the laboratory (Stern et al., 1996). To that end the Gas Hydrate Resonant Column (GHRC) apparatus was developed at the University of Southampton (Priest, 2004). The drive mechanism and operating principles are based on a ‘Stokoe’ resonant column, and was modified to apply flexural vibration to the sand (to derive longitudinal wave velocity) in addition to torsional vibration (shear wave) at small strain ( d,

where d = penetrometer diameter, z = penetration depth, su = undrained shear strength, γ = effective unit weight. In some pots slow pseudo-static tests were performed. These gave static resistances within 10% of the estimated values. 4.3

Undrained shear strength of clay

Dynamic penetration resistance

The undrained shear strength is obtained using Equation (1):

The dynamic penetration resistance (qd ) has been obtained by using Equation (3):

where F = load on theT-bar penetrometer; N = bearing factor, assumed to be 10.5 following Stewart & Randolph (1994); A = L × d the projected area of the T-bar

where F = soil resistance = mg – ma – FD ; m = penetrometer mass; g = 9.81 m/s2 ; a = acceleration measured from accelerometer; FD = hydrodynamic drag, ignored in this study; A = penetrometer tip area.


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Figure 6. Influence of undrained shear strength and impact velocity on qd . (d = 20 mm, m = 333 g). Table 2. Association between dynamic penetration resistance and the four independent variables. Figure 5. Dynamic penetration resistance.

Standardised independent variables

It is not possible to separate the friction and end bearing resistances as the penetration resistance is calculated from the acceleration, time record. However, as the tip area is generally much larger than the thin shaft, it is reasonable to assume that the contribution of the shaft resistance is negligible. The dynamic penetration resistance has been computed from impact until at rest. The results of two tests showing the resistance as a function of penetration are presented in Figure 5. Two identical penetrations into the same clay specimen are shown to indicate the repeatability of the data. It can be seen for these tests that the resistance and final penetration are essentially identical. In other repeat tests differences in penetration and resistance were always less than 10%. A feature of the responses in Figure 5 is the practically uniform dynamic resistance from impact until just before the final penetration depth is reached, and this pattern is evident in almost all the tests. As the penetration resistance is uniform with depth, an average value of dynamic penetration resistance has been computed for each drop for use in the further analysis below. The apparent oscillation in qd immediately after impact is believed to occur because the acceleration, from which qd is calculated, is affected by a stress wave, created by the impact, passing up and down the penetrometer shaft. The effect of impact velocity on the dynamic resistance is shown in Figure 6 for three undrained strengths with a fixed tip diameter and mass. A general trend of increasing resistance with velocity is evident. To further investigate the relationship between qd and the four influencing factors (d, m, su and v) multiple linear regression analysis using SPSS has been performed with output as shown in Table 2. Prior to the analysis, all variables were standardised by subtracting the mean and dividing by the standard deviation to provide a common scale for comparison. To confirm the association between qd and the four independent variables, the statistical significance was checked using p-values. At the 95% significance level, it is observed that all four variables produced p-values of less than 0.05, indicating strong association between © 2011 by Taylor & Francis Group, LLC

Undrained shear strength, su Impact velocity, v Diameter, d Mass, m

Beta coefficients

Significance, p

0.703

104 , the residual strain grew faster than proportional to ln(N ) for some of the sands L1 to L7 (see e.g. L2 in Figure 11). The curves εacc (N ) for the more well-graded sands show a bending in the semi-logarithmic scale, which becomes more pronounced with increasing coefficient of uniformity of the tested material (see L11 and L15 in Figure 11). These findings agree well with the results of the earlier study documented by Wichtmann et al. (2009). The parameters CN 1 , CN 2 and CN 3 (columns 7 to 9 of Table 1) were received by fitting the data in Figure 11 with the function fN (solid curve):

For all tested materials the increase of the strain accumulation rate with increasing average stress ratio was confirmed. Figures 9 and 10 compare the accumulation curves εacc (N ) or show the residual strain as a

The loading frequency does not influence the rate of strain accumulation in non-cohesive soils (see the literature review given by Wichtmann et al. (2009)) and is thus not considered in the HCA model.


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Figure 5. Accumulation curves εacc (N ) in tests with different initial relative densities ID0 (all tests: pav = 200 kPa, ηav = 0.75), thick solid curves = recalculation with HCA model.

Figure 6. Accumulated strain εacc /f¯ampl as a function of void ratio e¯ (all tests: pav = 200 kPa, ηav = 0.75).

Figure 7. Accumulation curves εacc (N ) in tests with different average mean pressures pav (all tests: ηav = 0.75, ζ = qampl /pav ), thick solid curves = recalculation with HCA model.

Figure 8. Accumulated strain εacc /(f¯ampl f¯e ) as a function of average mean pressure pav (all tests: ηav = 0.75, ζ = qampl /pav ).

Figure 9. Accumulation curves εacc (N ) in tests with different average stress ratios Y¯ av (all tests: pav = 200 kPa), thick solid curves = recalculation with HCA model.

As an alternative to the “by hand” calibration outlined above, the HCA model parameters were also determined by means of a C++ program. It finds those parameters for which the sum of the squares of the


386

differences between the experimentally obtained εacc data and the data predicted by the HCA model takes its minimum. The method may be seen as some kind of “fine tuning” of the parameters. The parameters

Figure 10. Accumulated strain εacc /(f¯ampl f¯e ) as a function of normalized average stress ratio Y¯ av (all tests: pav = 200 kPa).

Figure 11. Curves εacc (N )/(f¯ampl f¯e fp fY ), fitting of function fN .

summarized in columns 10 to 16 of Table 1 differ from those calibrated “by hand” due to simplifications of the “by hand” method (for example mean values ε¯ ampl and e¯ are used in the diagrams, parameters determined for different N -values are averaged). 4

RE-CALCULATION OF ELEMENT TESTS

The parameters given in columns 10 to 16 of Table 1 were used for recalculations of the element tests with the HCA model. The predicted curves have been added as solid lines in Figures 2, 5, 7 and 9. The parameters determined “by hand” (columns 3 to 9 of Table 1) deliver quite similar curves. In most cases the deviation between the experimental and the calculated data is small, confirming the good prediction of the HCA model. For some sands slightly too low accumulation rates are predicted for small pressures (Figure 7) which is due to deficits of the function fp . This will be inspected in more detail in future. 5

Similarly, the data for CN 1 , CN 2 and CN 3 in Figure 12j-o have been analyzed with Campl , Ce , Cp and CY calculated from Equations (7) to (10). Beside the calibration methods discussed in Section 3, the parameters Campl , Ce , Cp and CY were also estimated from the rate data (see Wichtmann et al., 2010a). CN 1 , CN 2 and CN 3 were determined both, from the data of all curves εacc (N ) and from the curves of the three tests with different amplitudes only. The poor correlation between CN 3 and d50 can possibly be improved by means of data from tests with larger numbers of cycles (N > 105 ).

SIMPLIFIED CALIBRATION PROCEDURE

In Figure 12 the HCA model parameters are plotted versus mean grain size d50 , coefficient of uniformity Cu or minimum void ratio emin , respectively. The data from the tests described by Wichtmann et al. (2009) were re-analyzed with Campl = 2.0 and are included in Figure 12. The correlations defined by Equations (7) to (13) are given in Figure 12 as solid lines and may be used for a simplified estimation of a set of parameters. The parameter Campl does not correlate with d50 or Cu (Figure 12a,b). For Ce both, a correlation with d50 and Cu (Figure 12c,d) and with minimum void ratio emin (Figure 12e) could be established. The values of Cp and CY plotted in Figure 12f–i were obtained calculating Campl and Ce from Equations (7) and (8). © 2011 by Taylor & Francis Group, LLC

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6

SUMMARY, CONCLUSIONS AND OUTLOOK

Based on the data from approx. 350 drained cyclic triaxial tests performed on 22 quartz sands with different grain size distribution curves a simplified procedure for the determination of the parameters of the authors’ high-cycle accumulation (HCA) model has been developed. Correlations of the HCA model

Figure 12. Correlations of the HCA model parameters with d50 , Cu or emin , respectively (SF = Wichtmann et al., 2009).

parameters with mean grain size d50 , coefficient of uniformity Cu or minimum void ratio emin , respectively, have been formulated. In future the simplified calibration procedure will be extended to granular materials with fines content.

ACKNOWLEDGEMENTS The study presented in the paper has been performed in the framework of the project A8 of SFB 398 “Lifetime oriented design concepts” during the former work of the authors at Ruhr-University Bochum, Germany. The authors are grateful to DFG (German Research Council) for the financial support and to M. Skubisch who carefully performed the cyclic triaxial tests.


REFERENCES Niemunis, A., Wichtmann, T. & Triantafyllidis, T. 2005. A high-cycle accumulation model for sand. Computers and Geotechnics, 32(4):245–263. Wichtmann, T., Niemunis, A. & Triantafyllidis, T. 2009. Validation and calibration of a high-cycle accumulation model based on cyclic triaxial tests on eight sands. Soils and Foundations, 49(5): 711–728. Wichtmann, T., Niemunis, A. & Triantafyllidis, T. 2010a. On the determination of a set of material constants for a highcycle accumulation model for non-cohesive soils. Int. J. Numer. Anal. Meth. Geomech., 34(4):409–440. Wichtmann, T., Niemunis, A. & Triantafyllidis, T. 2010b. Towards the FE prediction of permanent deformations of offshore wind power plant foundations using a highcycle accumulation model. In International Symposium: Frontiers in Offshore Geotechnics, Perth, Australia.

388


Understanding cyclic loading behavior of soil for offshore applications J. Yang Department of Civil Engineering, The University of Hong Kong, Hong Kong, China

ABSTRACT: This paper discusses two categories of cyclic loading with special reference to offshore applications: one is the combination of static and cyclic shear stresses and the other is a continuous rotation of the principal stress axes. Particular attention is given to the fist type which is in close relation to the stability of gravity offshore structures. It is shown that the initial static shear stress, due to the installation of an offshore structure, is a key factor for the behavior of the sand beneath the structure. The impact of this factor strongly depends on the initial state of the sand in terms of its relative density and confining stress and on the magnitude of the static shear stress relative to the cyclic shear stress applied. The second type of loading is in relation to the seabed in the free field under ocean waves. It is shown for this loading condition that rotation of principal stress axes without changing the magnitude of deviatoric stress can generate pore water pressures and that the amount and rate of pore pressures is affected by the intermediate principal stress parameter.

1

INTRODUCTION

Instability of the seabed under the action of storms or earthquakes is an important consideration in the design and installation of offshore structures (breakwaters, gravity or pile-supported platforms, pipelines, anchors, etc.), since it may cause severe damage to these structures affecting their operations (Oumeraci 1994; Palmer et al. 2004). Figure 1 illustrates possible failure modes of breakwaters and pipelines in relation to the instability of the seabed. During storms that are of major concern in offshore engineering, cyclic stresses develop in the seabed soil. The stresses can result in a progressive build-up of pore water pressure in the soil and a loss of strength of the soil. Particularly, if the seabed is comprised primarily of sand, the pore water pressure within the sand may build up to the level of the effective vertical pressure resulting in liquefaction and possible instability. The amount and rate of the pore water pressure build-up depends on the properties of the seabed soil including the particle gradation, permeability and relative density; it also depends on the characteristics of the storms such as the height, length, and period of wave components. The installation of an offshore structure into the seabed presents additional considerable difficulty to the problem, because very complex interactions are involved between the structure, the seabed soil and the waves. Note that the characteristics of the structure (e.g., the geometry, weight and stiffness) can play an important role in such interactions. From a geotechnical consideration, the presence of the offshore structure will induce both normal and shear stresses on the soil elements beneath the structure, which are then superimposed by “indirect cyclic loading” due © 2011 by Taylor & Francis Group, LLC

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Figure 1. Schematic illustration of failure modes of offshore structures: (a) breakwaters; (b) pipelines.

to rocking and translation of the structure. This loading condition is considered critical in the design of offshore structures, particularly for the gravity-type structures; it can be characterized by a sustained, static shear stress on the horizontal plane of a soil element throughout the course of cyclic loading. The term “indirect cyclic loading” is used here as opposed to the cyclic loading directly imposed by ocean waves on the seabed in the free field (i.e. without the presence of structures). In this case, the wave loading can be characterized by a continuous rotation

of the principal stress axes in the seabed soil (Madsen 1978; Ishihara & Towhata 1983). Following the pioneering work of Seed & Lee (1966), the cyclic loading behavior of soils has been extensively studied in the past decades with special reference to soil liquefaction during earthquakes. Most of the experimental studies have been focused on the cases where no static shear stresses exit. As reviewed by Yang & Sze (2010), the effect of the static shear stress is not yet well understood due to the lack of physical data and the inconsistency in the existing interpretations. With regard to the stability problems of offshore structures, a proper understanding of the soil behavior under this loading condition is important in developing better technical solutions in terms of safety, economy and reliability. This paper aims to introduce several new findings on the behavior of sand under the combination of static and cyclic shear stresses, established from a comprehensive experimental program (Yang et al. 2009; Yang & Sze 2010), to the offshore engineering community. In addition, recent advances in the study of the effect of rotation of principal stress axes on sand behavior (Yang et al. 2007) are also briefly discussed. 2 2.1

IMPACT OF STATIC SHEAR STRESS ON CYCLIC BEHAVIOR OF SAND

Figure 2. Cyclic triaxial loading conditions: (a) without initial static shear stress; (b) with initial static shear stress.

Simulation of static shear stresses

The cyclic behavior and liquefaction resistance of sand has been commonly studied by undrained cyclic triaxial tests with two-way, symmetrical loading. In such tests a sand specimen is first consolidated under the hydrostatic condition and then subjected to an undrained cyclic deviatoric stress, qcyc , which alters between positive and negative values of the same magnitude (Fig. 2(a)). This symmetrical loading aims to simulate the free-field level ground during earthquakes. To model the static shear stress in the cyclic triaxial test, the soil specimen is first consolidated under the major and minor principal stresses, σ1c and , to produce a static shear stress, τs , on the 45◦ plane σ3c in the specimen (Fig. 2(b)); a cyclic deviatoric stress is then superimposed under undrained conditions to produce a cyclic loading that is non-symmetrical about the hydrostatic stress state. The level of static shear stress can be measured by a parameter α defined below:

where σnc is the effective normal stress on the 45◦ plane. Obviously, with increasing α values the level of static shear increases, and the special case of α = 0 represents the symmetrical loading condition. Using the above method, a comprehensive experimental program consisting of more than 150 cyclic triaxial tests has been conducted on two standard sands (Toyoura sand and Fujian sand) to investigate the


impact of static shear stress (Yang et al. 2009; Yang & Sze 2010). The test program covered a wide range of initial states in terms of the relative density, confining stress and the initial static shear stress. Details on the test equipment, sample preparation and test procedure can be found in Yang & Sze (2010). Three distinct failure patterns have been identified from the test program, namely flow type failure or strain softening, cyclic mobility and plastic strain accumulation. 2.2

Flow failure

Figure 3 shows the undrained cyclic response of a loose Toyoura sand specimen at the relative density of 10% and the confining stress level of 100 kPa. The initial static shear stress was applied at 40 kPa and the subsequent cyclic deviatoric stress was 7 kPa. Given this combination, no stress reversals took place during the entire process of cyclic loading. The excess pore water pressure (PWP) built up progressively until the 28th loading cycle where an abrupt rise of the PWP occurred (Fig. 3(a)). Correspondingly, a sudden, run-away deformation was observed (Fig. 3(b)). Since there was no stress reversal, cyclic loading sat entirely on the compression side and the run-away deformation took place in compression. This failure mode has been observed on all loose specimens (Drc = 10% and 20%), and is characterized by a sudden run-away deformation being similar to that observed in loose sand under monotonic loading conditions (Yang 2002). The occurrence of this failure

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Figure 3. Typical response of loose sand (Drc = 10%, σnc = 100 kPa, τs = 40 kPa, qcyc = 7 kPa).

Figure 4. Typical response of medium dense sand (Drc = 50%, σnc = 100 kPa, τs = 40 kPa, qcyc = 110 kPa).

mode is regardless of stress reversal and is the result of the contractive nature of loose sand. 2.4 2.3 Cyclic mobility Shown in Figure 4 is a typical response referred to as cyclic mobility, which is similar in several aspects to that commonly observed in medium dense to dense sand subjected to cyclic loading without static shear stresses (Castro 1975). In this test, the sand specimen was consolidated at the relative density of 50% and the confining stress of 100 kPa. The static shear stress was at 40 kPa and the cyclic deviatoric stress was applied at 110 kPa, leading to stress reversals in the loading process. Note that the excess pore pressure build-up showed a different pattern with that observed in loose sand (see Fig. 4(a) and Fig. 3(a)) due to the cyclic contraction and dilation of the sand. The pore pressure kept building up but it was unable to reach the level of . At about the 12th cycle, effective confining stress σnc transient softening and then strengthening occurred, which repeated themselves in the subsequent cycles (Fig. 4(b)). This response was more substantial in compression because stress reversals dominated on the compression side. The regain of strength and stiffness was the consequence of dilation in the sample. © 2011 by Taylor & Francis Group, LLC

391

Plastic strain accumulation

Figure 5 shows the typical response called plastic strain accumulation. In this test the Toyoura sand specimen was consolidated at the relative density of 50% and a much higher confining stress, 500 kPa. The static shear stress was as high as 200 kPa and the cyclic deviatoric stress was 440 kPa, leading to almost zero stress reversals in this case. Although the sand specimen was also in a medium dense state, it did not show the features of cyclic mobility. The axial strain accumulated continuously on the compression side (Fig. 5(b)) and, particularly, apart from a pronounced strain in the 1st loading cycle, the strain accumulation in the subsequent loading cycles was at more or less a constant rate. It is worth noting that the excess PWP at the end of the test was only about 30% of the initial effective confining stress σnc . The modes of plastic strain accumulation and cyclic mobility have been observed in all medium dense and dense samples (Drc = 50% and 70%). It has been found that the criterion for the occurrence of cyclic mobility is that the cyclic loading should be with stress reversals. If there is no stress reversal, then plastic strain accumulation will be dominant.

Figure 6. Definitions for failure: (a) flow type failure; (b) cyclic mobility; (c) plastic strain accumulation. Figure 5. Typical response of medium dense sand (Drc = 50%, σnc = 500 kPa, τs = 400 kPa, qcyc = 440 kPa).

2.5

Characterization of cyclic shear strength

To characterize the cyclic shear strength, the failure criterion needs to be rationally defined for each type of response. For the flow type response given in Figure 3, the onset of failure can uniquely be defined as the triggering of run-away deformations, as clearly shown in Figure 6(a). For the cyclic mobility mode and the plastic strain accumulation mode, however, there is no such definite point indicating the initiation of failure or liquefaction. For the former one, the usual way for the symmetrical loading condition is adopted to define the failure as the point where 5% double amplitude (D.A.) axial strains occur (Fig. 6(b)). For the latter, since the strain development only takes place in one direction, the occurrence of 5% peak axial strain (P.S.) in compression (Fig. 6(c)) becomes as reasonable a criterion as the 5% D.A. is for cyclic mobility. Using the failure criteria defined above, the cyclic stress ratio (CRRn ) required to cause the sample to failure at a specified number of cycles (e.g., 10, 15 or 50) can be determined as

As an example, Figure 7 presents the cyclic strength specified at 10 loading cycles for Toyoura sand at © 2011 by Taylor & Francis Group, LLC

Figure 7. Impact of static shear stress on cyclic shear strength of Toyoura sand.

various initial states. The cyclic strength is found to be highly dependent on the three initial state parameters – relative density (Drc ), effective confining stress (σnc ) and initial static shear stress (α). Generally, CRRn always increases with increasing Drc and/or , regardless of the level of initial shear decreasing σnc stress. It means that more dilative sand bears higher cyclic strength. One thing worth noting is the impact of the static shear stress. The impact depends on both relative density and confining stress level. For loose samples tested, the impact is found to be beneficial to cyclic

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shear strength if the level of static shear stress (in terms of α) is low, but it tends to be detrimental when α becomes higher. In this regard,Yang & Sze (2010) have proposed a new concept called threshold α and shown that the threshold α can be estimated by using the noreversal line in the CRRn – α plane (see Fig. 7). This no-reversal line divides between loading with stress reversals (i.e. qcyc /2 > τs ) and without stress reversals (i.e. qcyc /2 < τs ). It is interesting to note that, whenever the CRRn trend line touches the no-reversal line, the CRRn is about to drop with α. In other words, when α is increased to a level such that loading required to bring the sand to failure is without stress reversal, then the cyclic resistance starts to reduce. More importantly, Yang & Sze (2010) have extended the concept to sand at high relative density within the framework of critical state soil mechanics, by introducing a state parameter that collectively accounts for relative density and confining stress (Been & Jefferies 1985). In this framework, the cyclic strength of sand having high relative density may not always increase with increasing α values; rather, it may reduce with α at larger α values when the confining stress is sufficiently high. More details can be found in Yang & Sze (2010). 3

Figure 8. State of stress in seabed due to passage of waves.

IMPACT OF ROTATION OF PRINCIPAL STRESS AXES

During the passage of waves the cyclic loading imposed on the seabed in the free field can be characterized by a continuous rotation of principal stress axes. In this case, the normal and shear stresses induced by the wave loading form a circular path in the plane of τvh and (σv − σh )/2, while the deviatoric stress remains unchanged and the direction of the principal stress (β) continuously rotates from 0◦ to 180◦ (Fig. 8). By assuming that the load induced by the propagation of waves on the surface of the seabed is in a harmonic form, the normal and shear stress components in the seabed can be determined as (Madsen 1978):

where x and z are the horizontal and vertical coordinates, respectively; σ0 is the amplitude of the harmonic wave load, L is the wave length and T is the period of waves. The undrained behavior of sand under this loading condition was investigated by Ishihara & Towhata (1983) using a hollow cylinder torsional shear apparatus. However, the apparatus could not independently control the inner and outer pressures, resulting in the ratio between the principal stresses, measured by b = (σ2 − σ3 )/(σ1 − σ3 ), and the mean total stress, © 2011 by Taylor & Francis Group, LLC

Figure 9. Response of dense sand under continuous rotation of principal stress directions with different b values: (a) pore water pressure; (b) stress path in q-p plane.

measured by p = (σ1 + σ2 + σ3 )/3 to cyclically change during loading. Recently, a continuous rotation of the principal stress axes while maintaining conditions of constant b and p was successfully achieved in an automated hollow cylinder apparatus (Yang et al. 2007). Therefore, this loading condition relating to the passage of waves over the free-field seabed can be better replicated. Shown in Figure 9 are selected results for excess pore water pressure for three Toyoura sand specimens consolidated at the relative density of 70% and mean effective stress of 100 kPa but under different b values. In testing the three samples the deviatoric stress was maintained at 34.65 kPa. The deviatoric stress q is defined as

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where the principal stress σ1 and σ3 are related to the stress components σv , σh , and τvh as

and σ2 is equal to the radial stress (σr ) in the hollow cylindrical configuration. The results shown in Figure 9 indicate that in all the three cases, the rotation of principal axes with constant deviatoric stress was able to generate excess pore pressures in sand specimens. Furthermore, the amount and rate of the pore pressures was strongly dependent on the parameter b. In general, the sand specimen sheared under the condition of b = 0 (i.e. σ2 = σ3 ) showed a much stronger resistance to pore pressure build-up than that under the condition of b = 1 (i.e. σ2 = σ1 ), with the case of b = 0.5 in between. It has also been found that the shear stiffness of the sand specimens degraded during the rotational shear and the degree of degradation was related to the parameter b. A detailed discussion of the tests and results can be found in Yang et al. (2007).

4

CONCLUSIONS

A proper understanding of the cyclic loading behavior of soils is essential in the development of rational technical solutions to tackle various offshore engineering problems. The main points of this paper are summarized as follows: – The cyclic loading can be classified into two major types with reference to offshore applications: (a) a combination of static and cyclic shear stresses, which is in close relation to the stability of gravity offshore structures, and (b) a continuous rotation of principal stress axes with constant deviatoric stresses, which represents the seabed in the free field during the passage of waves. – With regard to the loading type (a), the sand in the vicinity of the offshore structure may fail in three patterns: flow type, cyclic mobility and plastic strain accumulation, and the cyclic shear strength


is closely related to three state parameters: relative ) and density (Drc ), effective confining stress (σnc initial static shear stress ratio (α). – The flow type failure is most catastrophic due to its sudden and run-away nature, whereas in the plastic strain accumulation mode the accumulation of irreversible strains during cyclic loading is more critical than the build-up of pore water pressure. – The loading type (b) can be well replicated using the hollow cylinder torsional shear apparatus that is capable of controlling the axial load, torque, inner cell pressure and outer cell pressure independently. Even dense sand can be weakened by the rotation of principal stress directions without changing the magnitude of deviatoric stress. The degree of pore pressure build-up is significantly affected by the intermediate principal stress parameter b. REFERENCES Been, K. & Jefferies, M. 1985. A state parameter for sands. Géotechnique, 35(2): 99–112. Castro, G. 1975. Liquefaction and cyclic mobility of saturated sands. J. Geotech. Engng. Div., ASCE 101(GT6): 551–569. Ishihara, K. & Towhata, I. 1983. Sand response to cyclic rotation of principal stress directions induced by wave loads. Soils Founds, 23(4): 11–26. Madsen, O.S. 1978. Wave-induced pore pressures and effective stresses in a porous bed. Géotechnique, 28(4): 377–393. Oumeraci, H. 1994. Review and analysis of vertical breakwater failures: Lessons learned. Coastal Eng., 22 (1/2): 3–29. Palmer, A. C., Teh, T. C., Bolton, M. D. & Damgaard, J. S. 2004. Stable pipelines on unstable seabed: Progress towards a rational design method. Proc., Offshore Pipeline Technology Conf., Amsterdam, The Netherlands. Seed, H.B. & Lee, K.L. 1966. Liquefaction of saturated sands during cyclic loading. J. Soil Mech. Founds. Div., ASCE, 92(SM6): 105–134. Yang, J. 2002. Non-uniqueness of flow liquefaction line for loose sand. Géotechnique, 52(10): 757–760. Yang, J. & Sze, H.Y. 2010. Cyclic behaviour and resistance of saturated sand under non-symmetric loading conditions. Géotechnique, to appear. Yang, J., Sze, H.Y. & Heung, M.K. 2009. Effect of initial static shear on cyclic behavior of sand. Proc. 17th Int. Conf. Soil Mech. Geotech Eng., Alexandra, Egypt. Yang, Z.X., Li, X.S. & Yang, J. 2007. Undrained rotational shear and anisotropy of sand. Géotechnique, 57(4): 371–384.

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5 Shallow foundations



Observations of shallow skirted foundations under transient and sustained uplift H.E. Acosta-Martinez AECOM Australia (formerly Centre for Offshore Foundations Systems, University of Western Australia)

S. Gourvenec & M.F. Randolph Centre for Offshore Foundations Systems, University of Western Australia, Perth, Australia

ABSTRACT: This paper summarises results from a four-year experimental programme of centrifuge tests investigating installation resistance and uplift capacity of shallow skirted circular foundations in lightly overconsolidated clay. Centric and eccentric, monotonic, transient and sustained loading were investigated and the effect of gapping along the foundation-soil interface was explored. The results from this study showed that for a skirted foundation with embedment ratio d/D = 0.3, reverse end bearing could be mobilised under transient uplift and loads of up to 50% of the undrained soil resistance could be sustained for months without significant displacement – provided nominal contact along the skirt-soil interface was maintained. The presence of gapping along the skirt-soil interface and eccentric loading reduced undrained capacity by up to 40%, and increased displacement rates under sustained uplift loading by up to a factor five.

1

INTRODUCTION

the timescale over which skirted foundations can withstand sustained uplift.

Skirted foundations, or bucket foundations, provide an economically attractive foundation solution offshore to support or anchor structures for the oil and gas industry. A key benefit of skirted foundations lies in their potential to mobilise uplift resistance due to negative excess pore pressures developed between the soil plug and the base plate. Current industry recommendations (DNV, 1992; API, 2009; ISO, 2003) are based on classical bearing capacity theory (Terzaghi, 1943; Brinch-Hansen, 1970; Vesic, 1975) and provide a conservative estimate of undrained collapse loads for skirted foundation systems, particularly due to ignoring the beneficial effect of suction beneath the foundation during transient uplift or overturning. Although industry guidelines and recommended practices acknowledge that temporary suction caused by dynamic loads may allow greater capacities to be mobilised, since the effect is temporary they advise that suction should not be accounted for in the design unless substantiated by appropriate analysis or experimentation. To date there is no formal guidance regarding the timescale over which negative excess pore pressures may be sustained. The conservative limit loads predicted by classical design methods under general loading constitute the main motivation for the study presented in this paper. A challenging issue to improve design is to investigate the response of skirted shallow foundations using appropriate experimental techniques (i.e. with correct stress scaling) to provide quantitative data regarding the response under transient uplift and about © 2011 by Taylor & Francis Group, LLC

2

OVERVIEW OF METHODOLOGY

This project involved an extensive programme of physical modelling in the UWA geotechnical beam centrifuge (Randolph et al., 1991) to investigate the transient and sustained uplift resistance of skirted foundations. 2.1

Foundation model

Foundation models with embedment depth to foundation diameter ratios d/D = 0.15 and 0.3 were used to represent skirted foundations suitable for gravity based structures, jackets and tension leg platforms. The centrifuge models were machined (in-house) with a diameter of 120 mm and skirt lengths of 18 mm and 36 mm. All centrifuge tests were carried out at an enhanced gravity of 167 g, corresponding to a prototype foundation diameter of D = 20 m and skirt depths of d = 3 m and 6 m.An internal cruciform arrangement of internal skirts (stiffeners) was provided. The four compartments formed by the internal stiffener were connected by a small hole at the cross-over to allow for drainage during installation through a single vent in the base plate. Pore pressure transducers (PPTs) and total pressure transducers (TPTs) were placed on the model for monitoring negative excess pore pressures inside the soil plug, contact and separation of the base plate with

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Figure 2. T-bar penetrometer test results (a) undrained shear strength profile and (b) ratio of extraction to penetration resistance.

Figure 1. Schematic of foundation model (d/D = 0.3) (prototype dimensions).

the soil surface, and radial stresses on the peripheral skirt. All sensors were calibrated in-flight. Previous experience at UWA has confirmed the type of instruments used in the foundation model are reliable in the beam centrifuge (Chen, 2005; Chen & Randolph, 2007). The calibration programme was therefore limited to check the linear response of PPTs andTPTs with applied pressure in water. Calibration factors were validated periodically over the duration of the project. Figure 1 shows a schematic of the foundation model with d/D = 0.3 and location of sensors. 2.2

Soil sample

Lightly over consolidated samples of kaolin clay were used in this study. The kaolin is characterised by plastic and liquid limits of 27% and 61% respectively and a specific gravity of 2.6. The sample was initially consolidated at 1g in a press and then reconsolidated in the centrifuge to give an average (across the 8 samples) su /σv at skirt tip level of 0.45. This soil strength condition was chosen to ensure it could support a free-standing crack along the skirt-soil interface. The consolidation history led to an in situ void ratio e0 = 1.3, wet unit weight γt = 17 kN/m3 and over consolidation ratio at skirt tip level of ∼4. A representative value of the coefficient of consolidation at the vertical stress at skirt tip level (∼40 kPa) was taken as cv(av) = 2.6 m2 /year based on results of one-dimensional consolidation tests, corresponding to a coefficient of vertical permeability of ∼10−9 m/s (Acosta-Martinez & Gourvenec, 2006). The shear strength profile was assessed inflight using a T-bar penetrometer (Stewart & Randolph, 1994). Figure 2 shows typical profiles of undrained shear strength and the ratio of extraction © 2011 by Taylor & Francis Group, LLC

Figure 3. Cyclic T-bar penetrometer test results.

to penetration resistance based on a constant T-bar factor, NT−bar = 10.5 (Stewart & Randolph, 1994). The undrained shear strength was broadly repeatable between samples, although some variations did exist across the 8 samples used during the study; for example, su at tip level for the foundation with d/D = 0.3 varies by up to 20% between samples (which could lead to a 20% difference in measured foundation resistance while indicating the same bearing capacity factor). Figure 3 shows the degradation in undrained shear strength at skirt tip level for the foundation with d/D = 0.3, from cyclic T-bar tests (both carried out in the same sample).The degradation factor gives an indication of the soil sensitivity of the kaolin clay used in this study (Yafrate et al., 2009), and may be considered indicative of the interface friction factor, α, to account for shear on the skirt-soil interface during foundation installation. Figure 3 indicates a value of α between 0.3 and 0.4 is indicated from the cyclic T-bar tests. 2.3

Experimental procedures

The foundation models were attached by a rigid arm to a one-dimensional actuator allowing load or displacement control via a ±3 kN load cell and a 25 mm

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out in virgin sites (i.e. sites were not reused for multiple tests). 3 3.1


Installation resistance

Figure 4 shows a typical measured profile of installation resistance in terms of average stress (load cell measurement divided by the cross-sectional plan area of the foundation). Installation resistance was repeatable for all the tests in the programme. Initially, the skirts and stiffeners penetrated until the base plate touched down on the soil surface (0 < z/D < ∼0.26). Jacking was continued until a distinct increase in the load cell reading was observed, indicating that good contact had been achieved across the base plate. Similarity of readings from diametrically opposed pairs of TPTs confirmed the verticality of the foundation during installation; also the PPT under the lid showed that no excess pore pressures were developed during installation, indicating that the vent hole was sufficient to allow egress of water during installation. The average embedment reached during installation was ∼95% of the total skirt length indicating that the skirt and stiffeners were accommodated equally by inward and outward flow of the soil (i.e. a 50:50 split). A simple theoretical model, considering installation resistance as the sum of plate bearing and skirt friction (Equation 1) was used to back-calculate the interface friction factor, α.

Figure 4. Installation resistance.

stroke linear displacement transducer (LDT). Load and displacement control was achieved via servocontrolled actuators on the centrifuge. Loads and displacements were monitored and recorded by the centrifuge’s data acquisition system at a typical frequency of 10 Hz. The load cell was zeroed once the foundation was suspended above the soil surface and submerged in the free water, such that measured loads are net of the weight of the foundation. The model foundations were installed in-flight, with the drainage valve open to allow water egress from the skirt compartments. The drainage valve was then sealed and sufficient time was allowed for re-consolidation (i.e. dissipation of the excess pore pressure generated during installation) to occur prior to a transient or sustained uplift test. In most cases, re-consolidation was carried out under load-control (at the load achieved at the “end of installation” as marked in Figure 4). Displacement-controlled reconsolidation was required for the eccentric uplift tests to prevent the foundation rotating (under the eccentrically positioned loading arm). A base-line centric load test for comparison of the eccentric load tests, i.e. e/D = 0, was carried out following re-consolidation under displacement-control for appropriate comparison of the results. Transient loading was achieved with displacementcontrol at a rate of 0.1 mm/sec to achieve undrained conditions with respect to the cross-sectional area of the foundation. Taking a representative value of coefficient of consolidation of 2.6 m2 /yr, based on 1-D consolidation tests (Acosta-Martinez & Gourvenec, 2006), the non-dimensional velocity V = vD/cv exceeds 30, the limit for an undrained soil response (Finnie & Randolph, 1994). Local drainage is likely near the skirt tip and internal stiffeners) where V ∼ 1. Sustained load tests were carried out by applying a constant average bearing pressure, q, defined as a fraction of the transient ultimate capacity per unit area under centric uplift, qu . The study described in this paper involved 8 strongbox samples and 32 foundation tests considering transient and sustained uplift, and the effect of gapping along the skirt-soil interface and load eccentricity were also considered. All tests reported were carried

OBSERVATIONS

where Nc = bearing capacity factor for tip resistance; su0 and suav are values of undrained shear strength at foundation tip level and averaged over the embedment depth; Ab = total end bearing area; and As = internal and external surface area of the skirt and stiffeners. Values of interface friction factor 0.3 < α < 0.4 were back calculated from the profiles of installation resistance across the 8 samples and 32 tests. The backcalculated values of α are in good agreement with the degradation factor predicted from cyclic T-bar tests (shown in Figure 3). Detailed interpretation of installation of skirted foundations is presented in Gourvenec et al. (2009) and Acosta-Martinez and Gourvenec (2010). 3.2 Transient (undrained) uplift Figure 5 shows net transient uplift resistance normalised by the initial undrained shear strength at skirt tip level, su0 (i.e. measured by the T-bar), plotted against normalised uplift displacement, w/D. Uplift resistance is taken as load cell measurement divided by the cross-sectional plan area of the foundation. The load cell measurements have been corrected for the increasing difference in overburden pressure outside the foundation and the weight of the soil plug during uplift. Load displacement responses are shown

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Figure 5. Transient uplift response.

for cases of (i) centric loading with nominal contact along the skirt-soil interface for foundation embedment ratios d/D = 0.15 and 0.3; (ii) centric loading with nominal contact and with a gap along the skirt-soil interface for d/D = 0.3; and (iii) centric and eccentric loading with nominal contact along the skirt-soil interface for d/D = 0.3. Considering firstly the effect of foundation embedment ratio (final points on curves labelled “d/D = 0.15 no gap” and “d/D = 0.3 no gap” in Figure 5), a marked difference – nearly 300% – is observed in the measured ultimate uplift capacity between the foundations with d/D = 0.15 and 0.3, while plasticity solutions of bearing capacity for shallow foundations failing in general shear (e.g. Martin & Randolph, 2001) indicate a 20–25% difference over the same change in embedment ratio. It is not strictly appropriate to compare the measured normalised uplift resistance with theoretical bearing capacity factors since an increase in operative shear strength occurs during re-consolidation following installation and prior to uplift. The increase in operative shear strength, compared to the in situ shear strength measured by the T-bar and used to normalise the observed uplift resistance, will imply an artificially high ‘apparent’ bearing capacity factor. Theoretical bearing capacity factors are also only available for linearly increasing shear strength profiles as opposed to the parabolic profile of the lightly overconsolidated soil sample in these tests. Nonetheless, a bearing capacity factor Nc of no less than 6.5 would be expected for a foundation with d/D = 0.15 based on theoretical solutions for reverse end bearing (e.g. Martin & Randolph, 2001). An upper limit to pullout resistance under local shear failure of the foundation, i.e. simple pull-out, can be estimated with Equation 1 using α = 1 (after consolidation). This corresponds to qu /su0 = 2.2, three fold less than the lower limit of reverse end bearing capacity, and indicates that local shear governed failure of the foundation with d/D = 0.15. Bearing capacity factors in the range ∼7 < Nc < ∼12 are predicted by lower and upper bounds for d/D = 0.3 across the full range of skirt-soil


interface roughness, comparing well with the measured apparent bearing capacity factor of 9. The load cell data and the PPT measurements under the base plate, in conjunction with visual observations during testing and inspection of the sites after testing, indicated a local shear mechanism governed failure of the skirted foundation with d/D = 0.15 while general shear reverse end bearing governed failure of the foundation with embedment ratio d/D = 0.3. Current industry guidelines assume local shear governs uplift resistance, which can significantly under-estimate the capacity of a skirted foundation if reverse end bearing is mobilised. For example, for the upper limit of uplift resistance under local shear, the normalised uplift resistance of the foundation with d/D = 0.3 in this study is qu /su0 = 5.4 or 40% less than actually mobilised with reverse end bearing. Since reverse end bearing was not mobilised with the lower embedment ratio, further testing to investigate the effect of gapping and load eccentricity under transient and sustained loading concentrated on the deeper skirted foundation model with embedment ratio d/D = 0.3. The effect of gapping on the transient uplift capacity was investigated by creating a gap (in-flight under horizontal displacement control) along the skirt-soil interface. Two conditions were investigated (i) uplift immediately after gap formation and (ii) uplift after an extended period of re-consolidation following gapping. No difference in uplift resistance, qu /su0 , or failure mechanism was observed under transient uplift immediately after formation of a gap. In contrast, a 40% reduction in capacity was observed if a period of re-consolidation was allowed following crack formation prior to transient uplift. Reduced undrained uplift capacity following gapping and re-consolidation are likely to be related with softening of the surrounding soil and vertical and lateral propagation of the gap, creating a hydraulic connection between the soil plug contained within the skirts and the free water surface, thus preventing development of reverse end bearing. A complete description of the tests involving gapping, including the apparatus development, is presented by Acosta-Martinez et al. (2010a). The effect of load eccentricity was evaluated for normalised eccentricity, e/D = 0.12 and 0.25 compared with a base-line case e/D = 0. The difference in ultimate capacity between the centric load tests with nominal contact for the same embedment ratio (labelled “d/D = 0.3 no gap” and “e/D = 0” in Figure 5) arises due to the load- and displacement-controlled consolidation (discussed in Section 2.3). The consolidation stress continually diminished throughout displacement-controlled consolidation contributing to the lower uplift capacity. As would be expected, load eccentricity has a detrimental effect on the transient uplift capacity, with the effect increasing with the degree of eccentricity. Ultimate undrained uplift capacity was reduced by 30% due to a loading at an eccentricity of one quarter of the distance from the mid-line of the foundation and by

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Figure 6. Displacement time histories during sustained uplift.

Figure 7. Effect of eccentric loading on time–dependent displacements during sustained uplift.

40% due to a loading eccentrically half way between the centre and the edge of the foundation. A kinematic mechanism involving pure rotation followed by simultaneous rotation and vertical displacement was observed during the tests. Gapping along the skirt-soil interface or loss of negative excess pore pressure inside the soil plug were not observed during eccentric uplift study in these tests. Details of the equipment, test procedures and interpretation of the suite of tests investigating eccentric loading are presented by Acosta-Martinez et al. (2010b). Figure 8. Degradation of uplift resistance under sustained load.

3.3 Sustained uplift Figure 6 shows time histories of displacements normalised by the foundation diameter measured during sustained uplift for the cases of nominal skirt-soil contact and for gapping along the skirt-soil interface. Figure 6 shows the time-dependent ‘consolidation’displacements, wc , after deducting the immediate ‘elastic’ displacement, wi . The component of immediate displacement depended on the magnitude of uplift load but typically gave wi /D < 1% (as indicated in Figure 7). Sustained loading tests were stopped at a maximum displacement of wc /D ≥ 2%, or a prototype time tp = 1 year, whichever occurred first. As would be expected, the rate of displacement increased with increasing load. If contact was maintained along the skirt-soil interface, loads of 20% of the uplift capacity were sustained for up to a year with displacements of 1b).

Figure 10. Plastic strain distribution underneath grillage foundation with different spacing under V-M loading.

along with proposed failure envelopes. According to this figure, the normalised resisting moment slightly increases as the spacing increases from 1b to 5b. Comparing the failure envelopes for grillage foundations and solid foundation (Fig. 5 versus Fig. 9), it can be said that grillage foundations resist considerably higher normalised moments. This can be advantageous in cases where the design is mainly governed by moment capacity rather than vertical bearing capacity. Figure 10 depicts the failure mechanisms for 2D grillage systems with different spacing under combined V-M loading. While there are some interactions between the components of the grillage foundation with S = 1b, no interaction exists in the case of S = 4b. In addition, especially for the foundation with higher spacing, the local rotations of grillage components are limited and they resist the moment through their vertical bearing resistance as shown in Figure 10. © 2011 by Taylor & Francis Group, LLC

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4.3

Horizontal load-moment

Figures 11a and b compare the rotational response of 2D grillage foundations with different spacing under positive and negative moments. It is clearly seen in Figure 11a that the rotational capacity of a foundation under negative moment is higher than that of a foundation under positive moment for S = 1b (positive horizontal loads were applied; see Figure 2 for definition of positive direction). Such a visible difference is not indicated when the rotation responses of grillage foundations with S = 3b are simulated under positive and negative moments (Fig. 11b). Insensitivity of the rotational capacity to the direction of the horizontal load for grillage foundations with large spacing can be linked to the rotational resistance mechanism of these foundations being different from that of single foundations (see section 4.2). This also suggests that as the spacing increases, the skewness of failure envelope in the H-M plane would reduce. According to Figure 12a, the failure envelope for S = 1b is slightly skewed to reflect the dependency of

Figure 12. Proposed failure envelope in H-M plane for S = 1b (a) and S.1b (b).

Figure 13. 3D simulation against 2D simulation of grillage foundation under vertical loading (a) S = 1b (B = 0.9 m), (b) S = 4b (B = 2.5 m).

rotational-horizontal bearing capacity on the direction of the applied load. However, the failure envelope for S > 1b is considered to be circular (Fig. 12b). As seen in the Figures 12a and 12b the fixed ratio displacement tests trace the proposed failure surfaces reasonably well.

5 THREE DIMENSIONALITY Figures 13a and 13b present the vertical responses of 3D grillage foundations against those of 2D grillage models. To aid comparison the vertical bearing capacities of square foundations (with the same net areas as the respective grillage foundations) have also been calculated using the classical bearing capacity theory. In the 2D grillage model, an obvious failure point (where response gradient changes significantly) is noted for S = 1b, but no apparent failure point is seen in the 3D grillage foundation response up to the simulated level of displacement (Fig. 13a). Considering the foundation force at the last step as a conservative approximation of the failure (near failure) point, it can be judged that the bearing capacity of the 3D model is close to the bearing capacity of the equivalent square foundation. This is, however, not the case for larger spacings as shown in Figure 13b; the bearing capacities of 3D grillage foundations become significantly lower than equivalent square ones. For a spacing of 4b, the failure loads from both 2D and 3D grillage © 2011 by Taylor & Francis Group, LLC

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Figure 14. Plastic strain distribution under beam crossing points for grillage foundations with different spacings (vertical displacement: 0.01 m).

foundations are remarkably close while the bearing capacity of an equivalent square foundation is significantly higher than the bearing capacities of both 2D and 3D grillage foundations. The distribution of the failure zones on the surface and in depth in 3D simulations are shown in Figure 14 for S = 2b and S = 4b cases for a given displacement of 0.01 m. A difference in the failure mechanisms of

the grillage foundations with S = 2b and S = 4b is distinguished. For S=4b, the failure mechanism is fully developed under the grillage beams both away from and at crossing points (where perpendicular beams cross). For S = 2b, the full failure surface is only about to form under the beam in a section away from crossing points. Full failure surfaces are not seen to develop at crossing points for S = 2b (Fig.14). This implies that the grillage foundation (S = 2b) can accommodate more load due to additional soils resistance under the crossing points. The significant increase in soil resistance under crossing points, as compared to those in mid sections, can be attributed to local increases in the confining pressure in soil due to 3D constraints. 6

CONCLUSIONS

In this paper, the bearing capacity of grillage foundations on sand under horizontal, vertical and rotational combined loading has been investigated numerically. The effects of three dimensions, the soilfoundation interface and foundation spacing have all been addressed. The main findings are summarized below: – The vertical capacity of a grillage foundation on sand is a function of the spacing between the individual beams. – The normalized V-H failure envelope of a relatively rough grillage foundation is very similar to that of a single foundation and is not affected by the spacing. – The normalized resisting moment slightly increases with spacing. Grillage foundations can resist considerably higher normalized moments than solid foundations.


– The rotational capacity of a grillage foundation with higher spacing (S > 1b) is insensitive to the direction of applied horizontal load. – 3D constraints significantly affect the vertical bearing capacity of grillage foundations with low spacing. Based on the results of this study, a 3D failure envelope for these foundations has been proposed through defining the failure surfaces in V-H, V-M and H-M planes. REFERENCES Banimahd M. and Woodward P. K. 2006. Load-displacement and bearing capacity of foundations on granular soils using a multi-surface kinematic constitutive soil model, International Journal for Numerical and Analytical Methods in Geomechanics, 30(9): 865–886. Gottardi G. and Butterfield R. 1995. The displacement of a model rigid surface footing on dense sand under general planar loading, Soils and Foundations, 35 (3):71–82. Javadi A. and Spoor G. 2004. Soil failure patterns and loadsinkage relationships under interacting shallow footing and wheel arrangement, 88(3): 383–393. Kumar J. and Ghosh, P. 2007. Ultimate bearing capacity of two interfering rough strip footings, International Journal of Geomechanics, 7(1):53–62. Stuart, J.C. 1962. Interference between foundations, with special reference to surface footing in sand, Geotechnique, 12 (1): 15–22. Tan K. 1990. Centrifuge and theoretical modeling of footing on sand, PhD thesis, University of Cambridge, UK. Woodward P.K. and Griffiths D.V. 1998. Observation on the computation of the bearing capacity factor Nγ , Geotechnique, 48(1): 137–141.

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The vertical bearing capacity of grillage foundations in sand M.F. Bransby, P. Hudacsek, J.A. Knappett & M.J. Brown The University of Dundee, Dundee, UK

N. Morgan & D.N. Cathie Lloyd’s Register, Aberdeen, UK & Cathie Associates, Brussels, Belgium

R. Egborge Subsea7, Aberdeen, UK

A. Maconochie & G.J. Yun Technip, Aberdeen, UK

N. Brown & A. Ripley Acergy Norway AS, Norway & Acergy, Aberdeen, UK

ABSTRACT: Grillage foundations have been suggested as an alternative to flat plate foundations for seabed infrastructure such as pipeline end manifolds and pipeline end terminations.These foundations have the advantage that they can be installed in an offshore environment more quickly because of their hydrodynamic properties. Seabed infrastructure is typically subjected to a combination of vertical, self-weight loading and horizontal loading from pipelines, snag loads or hydrodynamics and so the combined vertical-horizontal capacities of the foundations are critical. However, little is known about the capacity of grillage foundations either to purely vertical loading or to combinations of load. This paper reports a series of physical model tests designed to address this knowledge gap. The experimental methods are presented first followed by typical results showing drained foundation capacity for pure vertical loading. 1

INTRODUCTION

Shallow foundations may be used to support offshore infrastructure such as pipeline end manifolds (PLEM), pipeline end terminations (PLET) or temporary anchors. Shallow foundations can directly rest on the seabed surface (for example they may consist of a flat steel plate – a ‘mudmat’), or may be skirted if large loadings are expected or soil conditions are poor. Manifold foundation structures are installed by lowering them from a vessel with a crane to the seabed. If the structure is large, good sea conditions are required because of splash zone wave forces. This means that installation may be delayed whilst waiting for appropriate weather conditions with cost consequences. During operation, a manifold is subjected to the vertical dead weight structural loading, W plus any additional loading during operation. These additional loads are likely to be horizontal and caused by (i) snag loads (from trawling gear or anchors), (ii) pipeline expansion or jumper loads, or (iii) hydrodynamic loads (in the case of shallow water). In many cases these will be applied relatively close to the level of the seabed because of the low height of the structures. This means that moment loads will be small and so combinations © 2011 by Taylor & Francis Group, LLC

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of vertical and horizontal load (albeit potentially with some torsional loading) will govern the foundation design. The capacity of foundations under combinations of vertical and horizontal (i.e.V-H) loading has been studied for a range of soil conditions. There are two main design approaches: (i) the Terzaghi bearing capacity method, and (ii) the failure envelope approach. In the bearing capacity method, modification factors for load inclination are applied to the vertical bearing capacity equations as first suggested by Terzaghi (1943). More recently, use of the failure envelope method has become more common. In this method, the load components (V, H and M) applied to the foundation are plotted on an interaction diagram and an envelope of failure states is ascertained (e.g. Zaharescu, 1961; Georgiadis, 1985; Nova & Montrassio, 1991; Gottardi & Butterfield, 1993). Equations describing failure envelopes have been presented by a range of authors for different soil conditions (e.g. sand, clay.) and foundation types (e.g. plan shapes, embedment ratios etc.). It should be noted that the majority of equations for drained conditions (e.g. Nova & Montrassio, 1991; Gottardi & Butterfield, 1993) have been for flat plate foundations resting on the soil surface.

Figure 1. Schematic of grillage geometry (Note: Not to scale. Many more grilles would be used in an offshore foundation).

Recently, grillage foundations have been used in several offshore projects to replace conventional flat plate foundation. Grillages (Figure 1) consist of many shallow, thin vertical grilles connected rigidly together. Typically, each grille is of thickness, t = 5–10 mm, length, d = 50 mm and have a centre-to-centre spacing, s varying from 20 mm to 80 mm. Grillages have the advantage over conventional foundations because water may flow between the grilles during installation through the splash zone. This will allow installation in poorer weather conditions and so have financial advantages to the contractor as they are likely to reduce vessel utilisation time during offshore installation. In addition, there is a possibility that the foundations may require less steel than conventional flat steel plate foundations and have a larger horizontal sliding capacity. Because of their novelty there is currently no generally accepted method to calculate the bearing capacity of grillages under either pure vertical or combined vertical – horizontal loading. In addition, it is not clear how bearing capacity is affected by the spacing, s, of the grilles and their thickness, t (or spacing ratio, s/t) for different soil conditions. For example, what soil condition and spacing ratio combinations are grillage foundations sufficient so that they can be used instead of flat plate foundations? A joint industry project was conducted with the aim of filling this knowledge gap. An extensive series of laboratory physical model tests were conducted by the University of Dundee which investigated first the vertical bearing capacity of grillage foundations and then the combined V-H capacity in drained, sandy soil. This paper considers the capacity of grillage foundations under pure vertical loading for plane strain grillage conditions. It reports first the experimental methodology used in this study followed by presentation of some preliminary results for grillages of different spacing (s/t = 2.5, 4, 6 and 8) in loose siliceous sand. 2

EXPERIMENTAL METHODS

2.1 Introduction In the series of tests reported in this paper, the capacity of grillage foundations under purely vertical loading was investigated. The foundation capacities © 2011 by Taylor & Francis Group, LLC

Figure 2. Direct shear test results (Dr = 8.7%).

were compared to standard design methods for both conventional shallow foundations and piles. 2.2

Foundation properties

Tests were carried out under plane strain conditions in a box of 300 mm width and length 1000 mm with a soil sample of approximately 360 mm depth. Grillage foundations were constructed from steel plates each of which were of thickness, t = 5 mm or 10 mm, width, L = 300 mm and height, d = 150 mm. They were connected together with 6 bars and spacer blocks to ensure that the whole foundation was rigid and each grille was parallel (Fig. 1 and 3). Figure 3b shows a grillage foundation with eight 5 mm grilles with a centre-to-centre spacing, s = 40 mm. The connection system allowed the spacing, s between the grilles and the number of grilles, N to be varied between tests. The individual grilles and their spacing were similar to those used offshore. 2.3 The soil sample A uniform, dry fine silica sand was used in the tests. The sand had d50 = 0.18 mm, d10 = 0.12 mm, ρmax = 1760 kg/m3 and ρmin = 1460 kg/m3 . To prepare samples of known relative density a mesh of similar dimensions to the box was placed at the base of the box. Sand was placed in the box and then the base mesh was slowly extracted from the soil. This sheared the soil to critical state giving a loose condition. The density of the soil was measured and found to be uniform with ρ = 1487 kg/m3 .This gave a relative density, Dr = 8.7%. Direct shear tests conducted on samples prepared with Dr = 8.7% found that φ = 30.8◦ with ψ ≈ 0◦ . These test results are shown in Figure 2. 2.4 Apparatus The apparatus for vertical loading is shown schematically in Figure 3a and a photograph is shown in Figure 3b. The grillage foundation was attached to a hydraulic actuator via an S-shaped tensioncompression load cell to measure the vertical load on the foundation. The actuator had a stroke length of 300 mm and was actuated using a hydraulic pump. A linearly variable differential transformer (LVDT) was positioned to measure the vertical position of the

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Table 1. Test programme.

Figure 3. Vertical grillage testing apparatus.

Test identifier

s (mm)

t (mm)

s/t

N

V16 V5 V18 V11

12.5 20 30 40

5 5 5 5

2.5 4 6 8

5 5 5 5

Figure 4. Load-displacement results for test V5: t = 5 mm; s = 20 mm (s/t = 4); N = 5 (B = 85 mm); L = 300 mm; Loose sand.

foundation. Image analysis was also used to measure foundation and soil movements. This required the front face of the box containing the sand sample to be constructed from Perspex to allow the soil and foundation to be photographed repeatedly with a digital camera during each test. Digital image analysis was later performed with the GeoPIV program (White et al., 2003).

are loose and the grille thickness, t = 5 mm in all tests. The grillage spacing ratio (s/t) is varied from 2.5–8 across the different tests (see Table 1).

2.5 Methodology

3.1

The grillage foundation was initially held several millimetres above the soil surface with the vertical actuator. The hydraulic pump was started and the actuator displaced the foundation vertically downwards at a velocity of approximately 0.4 mm/sec. Digital photographs were captured of the front face of the container throughout the test at the same time as the load cell and LVDT readings were recorded. Each test was stopped after approximately 60 mm of penetration of the foundation, just before the connector bars between the grilles came into contact with the soil. This amount of penetration ensured that data was available for z = 50 mm, the typical maximum length. In many tests, the foundation was then extracted from the soil to estimate the interface friction on the grilles.

The load-displacement results from a test with 5 grilles with thickness, t = 5 mm and spacing, s = 20 mm (so s/t = 4) in loose sand is shown in Figure 4. There is a clear increase in foundation capacity as the penetration, z of the grilles increases and no capacity before the grilles penetrate significantly. Also shown on Figure 4 are the results from separate calculations of foundation capacity assuming that either the foundation behaves as a solid, embedded plate (‘fully plugged solution’) or that each grille penetrates the soil and generates soil resistance as it was a plane strain pile (‘multiple pile solution’). For the solid foundation assumption it is assumed that the foundation is of breadth B (defined by the external dimensions of the grilles; Figure 1) and has embedment depth, z (i.e. the soil between the grilles move as if part of the foundation). The standard Terzaghi equation (1943) is used to calculate the bearing pressure at failure, qf :

2.6 Test programme In this paper four vertical tests are reported to investigate the effect of spacing ratio, s/t and penetration, z on bearing capacity of the grillage. Sand conditions © 2011 by Taylor & Francis Group, LLC

3

411

RESULTS Result for s/t = 4

where qf = Vo /A where the base area, A = BL, γ is the unit weight of the soil, c the apparent cohesion and Nγ , Nc and Nq are the bearing capacity factors which vary with the angle of friction, φ of the soil. Reissner (1924) found the closed form solution for Nq :

Hansen (1970) suggested empirically that:

The solid line with triangular markers (‘fully plugged solution’) on Figure 4 was produced combining Equations 1 to 3 and then also allowing for lateral friction on the outside of the external grilles (using the equations presented for the pile solution later). Parameters used were γ = 14.59 kN/m3 (from the measured ρ = 1487 kg/m3 ), c = 0 kPa and φ = 30.8◦ (from the direct shear test results, Figure 2). Figure 4 shows that the measured bearing capacity of the grillage foundation reaches that of the assumed solid, embedded foundation when z = 50 mm but then seems to slightly exceed this solution. Alternatively, a penetration, z = 23.7 mm is required to obtain the calculated flat plate bearing capacity for no embedment (Vo = 0.27 kN) which is likely to be the design value for a conventional mudmat. Thus, significant penetration of the foundation is required to generate the flat plate capacity, but there is additional capacity should the structure withstand additional foundation settlement (which may occur during installation when better tolerated). The dotted line on Figure 4 is the solution assuming that each grille acts as a single, plane strain pile. Therefore the capacity of each grille is obtained by summing the base resistance and the skin friction as done for axially loaded piles and the capacity of the whole foundation is obtained by multiplying the individual grille capacity by the number of grilles, N . The base capacity, Qb of each grille is given by a slightly modified version of the standard pile endbearing equation to allow for the base area (Ab = tL):

Figure 5. Load-displacement results for test V11: t = 5 mm; s = 40 mm (s/t = 8); N = 5 (B = 165 mm); L = 300 mm; Loose sand.

The dotted line on Figure 4 was calculated using the same parameters as for the shallow foundation assumption, but also using K = 1 (i.e. approximately twice K0 as appropriate for driven piles in sand, Kulhawy, 1984) and δ = φ − 5◦ = 26.4◦ . Figure 4 suggests that the pile equations are appropriate for very small penetrations, but that as the grille penetration increases, the foundation capacity increases more rapidly than given by the multiple pile solution with the parameters used. This increase might be because of enhanced silo pressures between the grilles which raise the normal stress on the interface (or effectively increase K) as in the case of plugged piles (e.g. Randolph et al., 1991). This is discussed in more detail by Bransby et al. (2009) who suggest an analytical solution for the bearing pressure, q before plugging for a large number of grilles:

where a = 2K tanδ/(s − t). The results using Equation 7 are shown on Figure 4 and give a better approximation to the experimental results, but require some improvement. 3.2

For a uniform soil layer, the skin friction on each grille of length, L and penetration depth, z becomes:

where K is the lateral earth pressure coefficient and δ is the angle of interface friction between the soil and the steel grille. Note that each grille has two sides and the sum of these two sides is included in Eq. 5. If it is assumed that each grille is unaffected by the others (i.e. assuming no ‘pile’group effect), the overall capacity of the foundation is:


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Result for s/t = 8

Figure 5 shows the equivalent results for a grillage foundation with a wider grille spacing (s/t = 8), but otherwise identical foundation and soil properties to the previous test. The simple analytical results for the rigid foundation and piled solutions are also shown on Figure 5 for comparison with the test results. For this wider grille spacing the measured load-penetration curve follows the pile solutions more closely. However, the foundation capacity is larger than predicted using the pile solution and this disparity increases with embedment. Again, the reason for the difference between the predicted pile solution and the measured foundation response could be due to grille-to-grille group effects or inter-grille arching/plugging effects

Figure 6. Load-displacement results in loose sand normalised by flat plate capacity (t = 5 mm; N = 5; L = 300 mm).

Figure 7. Penetration required to achieve equivalent flat plate capacity as a function of s/t; Loose sand, N = 5, t = 5 mm.

that increase the normal pressure acting on the grille interfaces and the grille tips. A large spacing ratio, s/t = 8 at first suggests that grille-to-grille interaction is unlikely, although it may be the ratio between the depth of soil between grilles, z, and the space between them (i.e. s − t) which is important for silo effects. At the end of the test, z/(s − t) = 1.7 and so the soil contained between each grille is much deeper than its width. Very significant foundation penetrations are likely to be required to mobilise the calculated equivalent flat plate capacity for zero embedment (i.e. the standard mudmat solution) for the spacing s/t = 8. Figure 5 shows that this displacement was not reached during the test at a penetration of 60 mm, but tentative extrapolation of the loading curve suggests that approximately 120 mm of penetration would be required. This may be a larger displacement than tolerated by the structure supported by the foundation, although working loads should not approach the vertical bearing capacity. 3.3 Result for different grillage spacings Figure 6 shows the load-displacement results from the two vertical tests already reported, along with two additional tests with grille spacings, s/t = 2.5 and s/t = 6. The load in Figure 6 is normalised by the capacity of an equivalent flat plate foundation of the same overall width at zero embedment (the standard ‘mudmat’ design capacity, Vo ). Figure 6 highlights the difference in foundation capacity with spacing ratio. When the plates are very closely spaced (e.g. s/t = 2.5 or 4) the grillage can provide the full flat plate capacity of an equivalent width conventional foundation, though some settlement is necessary to achieve this: 12 mm for s/t = 2.5 and 25 mm for s/t = 4. As the spacing increases still further (s/t = 6) the grillage is not quite able to mobilise the full equivalent flat plate capacity at the maximum penetration achieved in the test (60 mm). For s/t = 8 less than 50% of the equivalent flat plate capacity © 2011 by Taylor & Francis Group, LLC

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has been achieved when the tests were terminated at a penetration = 60 mm. The penetration required to achieve the flat plate capacity has been plotted against s/t in Figure 7. For the case of s/t = 8, which does not achieve this capacity, an estimate of the penetration required has been made by extrapolating the load-penetration curve as discussed earlier. Figure 7 suggests that full flat plate capacity may be achieved by grillages with s/t ≤ 4 in loose sand if only 25 mm of settlement is tolerated. The maximum spacing ratio will be a function of the amount of structural settlement which is permitted. For example, if instead 60 mm is tolerated (dotted line on Fig. 7) a spacing of s/t = 6 may just be acceptable. Note, however, that Figure 7 is based on results for tests with only five grilles (N = 5) and so should not be relied upon for design until the effect of number of grilles has been studied in detail. Soil displacements in the foregoing tests were obtained by comparing the sequential digital images captured during each test.This was done using GeoPIV (White et al., 2003). Figure 8 shows vectors of accumulated soil displacements for the complete installation process (i.e. penetration of the grillage by 60 mm) for the tests discussed previously. The vectors in Figure 8 are all plotted at the same scale factor to facilitate direct comparison across the different tests, and the grillage has been added to scale for each test. Due to a malfunction with camera, there were no images available for test V5 (s/t = 4). There is clearly significantly more soil displacement occurring in Figure 8a (where s/t = 2.5) as a block of soil between the grilles is displaced vertically downwards and the soil displaces with a general shear failure mechanism somewhat similar to that of an embedded foundation. This explains why the bearing capacity approaches that calculated for an embedded foundation when s/t = 2.5; the inter-grille soil has ‘plugged’. In contrast, Figure 8c shows the displacements for the most widely spaced foundation investigated (s/t = 8). Much smaller soil displacements are

foundations under vertical loading only. Such foundations may be used for offshore infrastructure such as pipeline end manifolds, pipeline end terminations or temporary anchors to replace conventional foundations under appropriate load combinations. The preliminary results presented here have demonstrated that the bearing capacity of the grillages is strongly affected by the spacing of the individual grilles in loose sand. Closely spaced grilles will plug after moderately small penetrations and give the capacity of an equivalent flat plate (a conventional ‘mudmat’). In contrast, widely spaced grilles will not plug until excessive displacements which may not be tolerable to the structure. Selection of grille spacing therefore will depend on the levels of tolerable structure settlement as well as the soil characteristics. Results for only five grilles in loose sand were presented here. Consequently, the results of this paper should not yet be used for design without further work. Further investigation is required to quantify performance in different soil types or densities and for more realistic numbers of grilles. In addition, more work is required to investigate their capacity under combined (V-H-M) loading. ACKNOWLEDGEMENTS The work conducted has been funded by Acergy UK, Subsea 7 and Technip. The authors wish to thank their respective companies for permission to publish this paper. The views expressed are those of the authors alone and do not necessarily represent the views of their respective companies. REFERENCES

Figure 8. Vectors of soil displacement at failure for vertical tests in loose sand.

observed for the same grille penetration as the grilles penetrate the soil and just push away the volume of the grilles themselves (as during pile jacking); the grilles have not plugged. In summary, the soil displacement fields confirm the earlier findings. Plugging occurred for the most closely spaced foundation (s/t = 2.5), but not for the more widely spaced ones (s/t = 6, 8). Figure 4 suggests that if images had been available for the test with s/t = 4, the onset of plugging may have been observed in the final images of the sequence. 4

CONCLUSIONS

The paper reports a series of laboratory model tests investigating the bearing capacity of grillage © 2011 by Taylor & Francis Group, LLC

Bransby, M.F., Knappett, J.A., Hudacsek, P. and Brown, M.J. (2009). The vertical capacity of grillage foundations. Submitted to Geotechnique, October 2009. Georgiadis, M. (1985). Load-path dependent stability of shallow footings. Soils and Foundations, 25(1), 84–88. Gottardi, G. and Butterfield, R. (1993). On the bearing capacity of surface footings on sand under general planar loads. Soils and Foundations, 33(3), 68–79. Kulhawy, F.H. (1984). Limiting tip and side resistance, fact or fallacy. Symposium on Analysis and Design of Piled Foundations, ASCE, San Francisco, 80–98. Nova, R. and Montrassio, L. (1991). Settlements of shallow foundations on sand. Géotechnique 41(2), 243–256. Randolph, M.F., Leong, E.C. and Houlsby, G.T. (1991). One-dimensional analysis of soil plugs in pipe piles. Géotechnique 41(2), pp.587–598. Reissner, H. (1924). Zum Erddruckproblem. Proc. 1st Int. Conf. Appl. Mech., Delft, 295–311. Terzaghi, K. (1943). Theoretical soil mechanics. New York: Wiley. White D.J., Take W.A. and Bolton M.D. (2003). Soil deformation measurement using particle image velocimetry and photogrammetry. Géotechnique, 53(7), 619–631. Zaharescu, E. (1961). Sur la stabilité des fondations rigides. Proc. 5th Int Conf. Soil Mech., Paris, 1, 867–871.

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Behaviour of skirted footings on sand overlying clay C.T. Gan, K.L. Teh, C.F. Leung, Y.K. Chow & S. Swee Centre for Offshore Research and Engineering, National University of Singapore

ABSTRACT: On sand overlying soft clay, the bearing resistance developed in the upper soil layer could be substantially larger than that in the lower layer. Installing an offshore foundation on this kind of soil condition could possibly bring about punch-through risk when the imposed load exceeds the upper soil bearing resistance. Intuitively, the punch-through risk could be reduced by minimizing the bearing resistance variance between the upper and lower layers by either reducing the bearing resistance of the upper soil or increasing bearing resistance of the lower soil. The former case is explored in this paper by adopting a footing with extended skirt denoted as skirted footing. This paper reports the findings obtained from a series of centrifuge model tests to investigate the effect of skirt height on the development of bearing resistance of skirted footings in sand overlying clay. The findings suggest that skirted footings produce a lower peak bearing resistance and punch-through depth as compared to spudcan foundation. The footing with skirt height equaling the sand thickness appears to greatly reduce the risk of punch-through failure. 1

INTRODUCTION

The foundation of jack-ups is not custom-designed for a specific site. Therefore, they must be designed to remain stable, regardless of the soil conditions (Poulos, 1988). When a spudcan is installed in strong soil layer overlying weak soil layer such as sand overlying clay where there is a potential of reduction in bearing resistance as a spudcan penetrates, punch-through risk exists. Punch-through happens when the spudcan penetrates uncontrollably through the strong layer into the underlying weak layer, resulting in excessive tilting of the structure. The hull tilting condition can be worsened by additional distributed weight due to shifting of the centre of gravity of the hull towards the punchedthrough leg. A punch-through may subject the leg(s) to very large stresses and cause significant damages to the structure and equipment onboard. This is likely to cause heavy losses in terms of time and cost. A point to note is that punch-through has been occurring at an alarming rate of once per year (Osborne & Paisley 2002). This indicates the needs for improved procedures in identifying and managing such risk. While with detailed site surveys of the area and good quality soil data the punch-through risk can be identified and a safe rig installation requires the risk to be adequately mitigated. Intuitively, this can be achieved by sufficiently reducing the bearing resistance of the upper layer or increasing the bearing resistance of the lower layer such that the difference of bearing resistance between the layers is minimum or reduced to a level where the anticipated spudcan punch-through behaviour is deemed manageable. A limited number of approaches are available to mitigate punch-through risk. Perforation drilling


is a ground preparation method conducted prior to rig installation which involves (partially) removal of strong layer and hence reduces the bearing resistance of the stronger layer. This method has been implemented at some punch-through identified sites with varying degree of success. However, this method is rather time-consuming. Chan et al. (2008) reported a perforation drilling operation conducted in offshore Southeast Asia which took up to 171 hours to drill 106 perforations of 0.91 m diameter and 28 m deep, and only two-thirds of the perforations were successfully drilled. To implement the method, it also requires a sufficient operation time in advanced of the rig installation. There is another form of mitigation which involves no modification to the ground condition but incorporating some cautious procedures during rig installation. According to Rapoport and Young (1987), the procedures include: • Using a small air gap so that the vertical displace-

ment during punch-through failure can be reduced as the leg load is partially countered by the hull buoyancy as it penetrates the water line. • Preload in water so that the leg loads are reduced during penetration. • Preloading one leg at a time. While these procedures are generally found effective in mitigating the punch-through risk, incorporating these procedures inevitably increase the installation time. Also, there are certain conditions where these procedures may not be applicable or suitable. For instance, preloading one leg at a time may lead to intolerance RPDs and the correction procedure will have to perform to ensure that the limit is not exceeded.

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The feasibility of adopting skirted footings, which are flat footings/spudcans equipped with skirts along their circumferences, to mitigate the punch-through effects was conceptualized by Svano & Tjelta (1993). As compared to a spudcan, a skirted footing penetrating in single layer soil would develop a different resistance profile with relatively low resistance during penetration of skirt and followed by sudden a sudden increase in resistance when the flat base is making contact with the ground surface. This trend of resistance profile has been confirmed by test results reported by Byrne & Cassidy (2002) in clay and Teh et al. (2006) in sand. Following the idea of Svano & Tjelta (1993), if the skirt can be designed such that the penetration resistance of the skirt alone is less than the available force during jacking, that is before the hull is lifted free from the sea, the punch-through risk could be greatly reduced. Some preliminary centrifuge test results were reported by Teh et al. (2008) in which the penetration responses of a model spudcan and a skirted footing in medium dense sand overlying soft clay were compared. It should be noted that the skirted footing had a skirt height that was shorter than the upper sand layer thickness. The test results revealed that the skirted footing developed a lower magnitude of peak bearing resistance in the upper layer and the measured punch-through distance, dP-T (as defined in Fig. 2) was shorter, implying a more controllable penetration response as compared to that of the spudcan. Assuming that to a large extent, the skirted footing penetration response in strong soil overlying soft soil is governed by its skirt height or more precisely the ratio of the skirt height to the thickness of the upper layer. Following this, the effectiveness of using skirted footing to mitigate punch-through risk will therefore be affected by its skirt height. This aspect deserves investigation. A series of centrifuge tests was conducted by modeling the penetration of skirted footings with different skirt heights into sand overlying clay. In this paper, the set-up and procedures of the tests are described in the following before the test results are presented. Finally, the practical issues concerning the use of skirted footing particularly the selection of skirt height are discussed. 2

EXPERIMENTAL SET-UP AND MATERIAL PROPERTIES

The centrifuge model tests presented in this paper were conducted on the beam centrifuge at National University of Singapore (NUS).This is a 2 m radius centrifuge that is designed for a payload capacity of 40g-tonnes and can spin-up to a maximum acceleration of 200 g. A stack of 100 tracks silver-graphite slip rings is mounted on top of rotor shaft for power and signal transmission between the centrifuge machine and the control room. Detailed information about NUS centrifuge can be found in Lee et al. (1991) and Lee (1992). © 2011 by Taylor & Francis Group, LLC

Table 1. 2003).

Properties of Malaysia kaolin clay (after Goh

Parameter

Value

Liquid limit Plastic limit Specific gravity Coefficient of consolidation (at 100 kPa) Coefficient of permeability (at 100 kPa)

80% 35% 2.60 40 m2 /yr 2.0 × 10−8 m/s

Modified Cam Clay parameters: Critical state frictional constant Slope for normal consolidation line Slope for swelling line

0.9 0.244 0.053

Table 2.

Properties of Toyoura Sand (after Teh 2007).

Parameter

Value

Specific gravity Uniformity coefficient Average particle size Dmax Dmin D50 D10 Range of density 1 Critical state friction angle, φcv

2.65 1.3 0.2 mm 0.3 mm 0.115 mm 0.2 mm 0.163 mm 1335–1645 kg/m3 32◦

1

Jamiolkowski et al. (2003)

Malaysia kaolin clay was used in this study. The engineering properties of the clay are shown in Table 1. Firstly, the dry kaolin powder was mixed with water at a water content of 1.5 times its liquid limit which is 120% of the clay powder weight. They were mixed thoroughly in a large air-tight mixer with a constant suction applied for 4 hours to form fully saturated clay slurry. Prior to the clay slurry pouring, a 30 mm thick sand layer that serves as drainage layer was placed at the base of a 550 mm diameter model container. A layer of silicon grease was applied on the inner wall of the model container to reduce the friction between the sidewall and the soil. The trapping of air pockets were minimized as the clay slurry was placed under water. To form a relatively firm clay surface for sand placement, the clay slurry was then subjected to a gradually increasing pre-consolidation pressure from 0.5 kPa to 15 kPa at 1 g. The overall pre-consolidation process took 7 days to complete. Next, the surface water was removed before the placement of sand. The type of sand used for the experiment is Toyoura sand. The engineering properties of the sand are shown in Table 2. The sand was prepared following the method adopted by Teh (2007) which is a spot-type sand raining process with proper control of sand drop height. The relationship between the relative density of the sand layer formed and the sand drop height has been established by researchers from the same research team using the same apparatus (Eio 2003 and Teh 2007). In the present tests, the relationship was re-calibrated by conducting a number of trial

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sand raining tests. It is established that raining at a drop height of 600 mm consistently produces a dense sand layer with relative density of 95%. This drop height was adopted in the preparation of the upper sand layer. A 54 mm thick sand layer was formed on top of the earlier prepared clay sample. The sand was then saturated by creating a controlled differential hydraulic head condition. The soil sample was kept fully saturated with a layer of free water layer maintained throughout the consolidation and testing processes. The soil sample was then moved onto the platform of the centrifuge and subjected to self-weight consolidation at 100 g for at least 7 hours. At the end of this consolidation process, the underlying clay sample is believed to have achieved at least 90% degree of consolidation determined based on the surface settlement of similar soil sample monitored previously by Teh (2007). The spudcan/skirted footing was jacked into the model ground at a penetration rate of 0.5 mm/s that results in drained condition in sand and undrained in clay following the velocity group parameter proposed by Finnie (1993). 3

RESULTS AND DISCUSSIONS

Unless otherwise stated, all experiment results presented in this paper are presented in either normalised form or prototype dimension scaled according to the scaling laws published by Garnier et al. (2007). A total of four tests were performed using different model foundations which include a model spudcan and three model skirted footings with different skirt heights, L, of 0.5, 1.0 and 1.5 times the sand layer thickness, Hs , respectively All the model foundations have the same prototype diameter of 8 m. Two tests were performed in a sample. For each model skirted footing, four 1 mm (model scale) diameter holes were drilled at the flat base of the footing to facilitate the water draining. Figure 1 shows the dimensions and load reference points, L.R.P, of the model foundations. The summary of test details is presented in Table 3. The underlying clay layer was normally consolidated under the sand overburden as well as soil self-weight. Hence, the undrained shear strength, su , profile is expected to increase linearly with depth. Although miniature ball penetrometer tests were conducted with an aim to characterize the su profile of the underlying clay, the ball penetration response appears to be questionable with inconsistent development of resistance with depth. Similar anomaly has been observed by Teh (2007) and Lee (2009) which is generally attributed to the unstable state of the soils trapped underneath the ball as the ball penetrate continuously from the upper sand layer into the lower clay. The su profile is therefore approximated using the modified cam clay framework proposed by Roscoe & Burland (1968) and it is expressed as follows:

where z is the depth beneath the sand-clay interface, in meters and su in kPa. Equation 1 generally agrees © 2011 by Taylor & Francis Group, LLC

Figure 1. Schematic diagram showing dimensions and load reference points (L.R.P.) of spudcan and skirted footings (all dimensions are in mm, model scale). Table 3.

Summary of test details.

Test

Skirted height, L m

1

Spudcan Skirted footing 1 Skirted footing 2 Skirted footing 3

0 2.7 5.4 8.1

0 0.5 1.0 1.5

L/Hs

Note: 1 Hs = 5.4 m (prototype scale)

with the strength profiles reported by Teh (2007) who used similar materials in preparing the test samples. 3.1

Spudcan versus skirted footing

The penetration resistance profiles of the model spudcan and skirted footing 1 (L/Hs = 0.5) are compared in Figure 2. The penetration resistance is calculated by dividing the measured load by the widest crosssectional area of the footing and penetration depth, d, is measured from the model ground surface to the L.R.P of the footing. For the spudcan test, the penetration resistance increases with depth and reaches a peak resistance, qpeak , of 338 kPa at d of 0.65 m, followed by a sudden loss in resistance before resistance of similar magnitude as qpeak is developed at 5.6 m. In this paper, the difference in elevation between qpeak and the ‘regained qpeak ’ is defined as punch-through distance, dP-T , as indicated in Figure 2. For profile

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the footing resistance may be still subjected to softening effect from the underlying clay. This explains the slight drop in penetration resistance of skirted footing 1 from 3.0 m to 3.9 m (Fig. 2). While the mechanisms discussed above seem to explain the test results adequately, they are postulated to certain extent. Further investigations on the failure mechanism shall be conducted in the future to verify the postulated mechanisms. 3.2

Figure 2. Penetration behaviour of spudcan and skirted footing 1 in sand overlying clay.

which shows dP-T > 0, which is observed in the spudcan penetration resistance profile, and the qpeak falls between the range of the applied loads i.e. qpeak is greater than the lightship weight but smaller than the preload, punch-through risk should be identified. On the other hand, the penetration resistance of skirted footing 1 increases gradually as the skirt penetrates into the soil. At d = 2.95 m i.e. some distance after the footing base made full contact with the ground, the qpeak is developed but with a smaller magnitude of 250 kPa as compared to that of spudcan. The corresponding dP-T is also shorter. An explanation is offered in the following. By drawing analogy to an open tube pile, the q before the footing base is fully in contact with the sand, is the sum of inner and outer skirt wall frictions, and end bearing on skirt annulus (Houlsby & Byrne 2005a, b). When the flat base of the skirted footing is fully in contact with the ground, the skirt chamber is likely to be filled by sand. After a short distance of penetration which allows the trapped sand to undergo some degree of compaction, the skirted footing and trapped sand may move like a rigid block as the footing advances. At this point onwards, the skirted footing may be analogous to a cylindrical footing with foundation base defined at the level of skirt tip. The q is therefore contributed by the outer skirt wall friction and end bearing mobilised across the footing widest cross-section area at the foundation base level. The thickness of the sand layer where the ‘cylindrical footing’ is founded should be viewed in term of effective value, Hs , which may be approximated as (Hs − L). For instance, for skirted footing 1, the Hs is 2.7 m (see Fig. 1b). When Hs > 0, the end bearing component of © 2011 by Taylor & Francis Group, LLC

Effect of skirt height

Figure 3 shows the penetration resistance profiles of three skirted footing of different skirt heights (refer to Table 3 for test details). Fundamentally, the bearing capacity of a foundation in sand overlying clay is contributed by resistances in both the upper and lower layers in which the correct proportion of these resistance components is partly determined by the relative distance of the foundation base to the underlying clay. For a skirted footing behaving like a ‘cylindrical footing’i.e. composite block of skirted footing and trapped sand, such distance is essentially Hs . For skirted footing 1, Hs is greater than zero. The ‘cylindrical footing’ is viewed as founded on sand overlying clay system and the footing resistance ia expected to be affected by the softening effect due to the underlying clay. Therefore, the penetration profile registers a qpeak and followed by post-peak reduction. The degree of post-peak reduction should be dependent on Hs in which intuitively a higher Hs (i.e. shorter skirt) leads to a higher post-peak reduction. For skirted footing 2 in which the skirt height is equal to the sand thickness (Hs = L), the Hs is essentially zero. During the formation of ‘cylindrical footing’ mechanism, the base of the ‘cylindrical footing’ reaches the sand-clay interface. The footing is likely to be loaded directly on the underlying clay with the upper sand layer serving as overburden. Following this argument, one would not expect the peak and postpeak responses to occur. However, it is worth noting that even when Hs = 0, there could be a possibility of developing peak and post-peak responses during the transition from the ‘open tube pile’to ‘cylindrical footing’ mechanisms if the total resistance of the former mechanism is larger than that of the latter mechanism. The resistance components mobilised by each failure mechanism have been discussed in Section 3.1. When the skirted footing continues to penetrate into the clay layer, there ought to be some progressive reductions in side friction along the outer skirt wall as the shear interface changes from sand-wall to clay-wall. However, this is countered by the increasing overburden pressure and the mobilized shear strength of the underlying clay that yields a higher end bearing capacity with depth. The test results of skirted footing 1 and 2 imply that the latter gain outweighs the former reduction, resulting in a smoothly increasing penetration resistance. Skirted footing 3 represents a case with Hs < 0 in which the skirt height is longer than the sand layer

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two skirted footings, it still performs better than the spudcan.

3.3

Figure 3. Penetration behaviour of skirted footings of different skirt heights in sand overlying clay.

thickness. Interestingly, the measured profile indicates a post-peak reduction. Further, the post-peak reduction takes place before the flat base of the footing is in contact with the ground surface but the skirt tip has penetrated into the clay layer. The mechanism associated to the ‘cylindrical footing’ as described earlier is not yet formed. So the possible explanations for this interesting penetration response should concern only the ‘open tube pile’ mechanism. As mentioned earlier, the total resistance developed under this failure mechanism is the sum of inner and outer skirt wall frictions, and end bearing on skirt annulus. As the skirt advances, there is an increase in side frictions due to a greater skirt wall surface in contact with soils (solely sand before the skirt tip penetrates beyond the sand-clay interface). It is noted that the friction increment is much greater in sand than in clay. However, the shear strength of the sand particularly dense sand is susceptible to large strain reduction, suggesting that the actual increase in friction would not be proportional to the increase in wall surface area. On the other hand, there is a relatively significant reduction in end bearing as the skirt annulus is shifting from in uniform sand to sand-over-clay and later to soft clay. The results shown in Figure 3 imply that the loss in the end bearing outweighs the gain in the side friction. The skirted footing 3 regains the penetration resistance shortly after the flat base fully touches the model ground at around d of 8.8 m (see Fig. 3). It may be worth mentioning that the qpeak and dP-T recorded by skirted footing 3 are smaller than that of spudcan. This suggests that although the performance of skirted footing 3 is less satisfactory as compared to the other © 2011 by Taylor & Francis Group, LLC

Practical issues

The findings obtained from the present study suggest that when a footing equipped with skirt penetrates in sand overlying clay, it mobilises a lower qpeak and experience a smaller dP-T The findings also show that the height of the skirt affects the performance of the skirted footing in which the ‘optimum’ skirt height appears to be equal to the sand layer thickness. Ironically, as pointed out in Section 1 that the foundation of jack-up rigs is not custom-designed for a specific site and the skirt height is normally designed to be less than a quarter of the foundation diameter to ensure smooth rig towing (Purwana 2010, pers. comm.). As the upper sand layer thickness varies with region and area, the optimum performance of skirted footings in mitigating punch-through may not be achieved. Another practical issue is the installation of skirted footings in multi-layered soil particularly a soil condition where a sand layer is sandwiched in between two soft clay layers. The inner space of the skirt may be partially/fully filled up with the upper soft clay before penetrating through the sand layer, and hence altering the soil failure mechanism. In other words, the effectiveness of the skirt may be reduced. This topic is currently being investigated by the NUS research team. Owing to limitations in model fabrication, the skirt thickness modeled in the current study in prototype scale is 0.3 m (note that it is only 3 mm in model scale). Such dimension exceeds the typical dimension adopted in the field. Furthermore, stiffeners that are employed to increase the skirt integrity were not modeled in the present study. These differences may affect the performance of the skirted footing and hence the issues deserve further investigation.

4

CONCLUSIONS

This paper presents a series of centrifuge model tests on spudcan and skirted footings penetration on sand overlying clay. The results suggest that the skirted footings provide a more steady penetration response with reduced peak bearing resistance and punch-through distance when the peak bearing resistance is exceeded. The former enables the footing to penetrate through the strong layer more easily and the latter shortens the leg uncontrollable penetration distance. The post-peak reduction is essentially eliminated in a test featuring a skirted footing with skirt height equal to the sand thickness. However, the optimum use of skirted footing for jack-up rig installation in sand overlying clay depends on various factors such as the sand thickness in relation to the skirt height, the ratio of maximum footing preload to peak bearing resistance, clay shear strength and soil stratification.

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ACKNOWLEDGEMENT The authors wish to acknowledge the officer of the Geotechnical Centrifuge Laboratory, National University of Singapore, Mr. Wong Chew Yuen, for his valuable assistance in conducting the centrifuge model tests.

REFERENCES Byrne, B.W. & Cassidy, M.S. (2002). Investigating the response of offshore foundations in soft clay soils. Proc. 21st Int. Conf. Offshore Mech. And Artic Eng., Oslo, OMAE2002-28057. Chan, N.H.C., Paisley, J.M. & Holloway, G.L. (2008). Characterization of soils affected by rig emplacement and swiss-cheese operations – Natuna Sea, Indonesia, a case study. 2nd Jack-up Asia Conference, Singapore. Eio, T. L. 2003, Sand preparation for geotechnical model test. BEng thesis, National University of Singapore, Singapore. Finnie, I.W.S. (1993). Performance of shallow foundations in calcareous soils. PhD thesis, University of Western Australia. Garnier, J., Gaudin, C., Springman, S.M., Culligan, P.J., Goodings, D., Konig, D., Kutter, B., Phillips, R., Randolph, M.R. & Thorel, L (2007). Catalogue of scaling laws and similitude questions in geotechnical centrifuge modelling. International Journal of Physical Modelling in Geotechnics. 8(3), 1–23. Goh, T.L. (2003). Stabilisation of an excavation by an embedded improved soil layer. PhD thesis, National University of Singapore. Houlsby, G.T. & Byrne, B.W. (2005a). Design procedures for installation of suction caissons in clay and other materials. Proc. of the Institution of Civil Engineers-Geotechnical Engineering, 158, No. 2, 75–82. Houlsby, G.T. & Byrne, B.W. (2005b). Design procedures for installation of suction caissons in sand. Proc. of the Institution of Civil Engineers-Geotechnical Engineering, 158, No. 3, 135–144.


Lee, F.H. (1992). The University of Singapore Geotechnical Centrifuge User Manual. Research Report No. CE001. July 1992. Lee, F.H., Tan, T.S.,Yong, K.Y., Karunaratne, G.P. & Lee, S.L. (1991). Development of geotechnical centrifuge facility at the National University of Singapore. Proc. Int. Conf. Centrifuge 1991, pp. 11–17. Jamiolkowski, M.B., Lo Presti, D.C.F. & Manassero, M. Evaluation of relative density and shear strength of sand from cone penetration test (CPT) and flat dilatometer (DMT). In Soil behaviour and soft ground construction, ed by J.T Germaine, T.C. Sheahan and R.V. Whitman, ASCE, GSP 119, pp. 201–238. Lee, K.K. (2009). Investigation of potential spudcan punchthrough failure on sand overlying clay soils. PhD thesis, University of Western Australia. Osborne, J. & Paisley, J. 2002. South East Asia Jackup Punch-throughs: the way forward? Proceedings of the Conference on Offshore Site Investigation and Geotechnics – Sustainability and Diversity, London. Poulos, H.G. (1988). Marine geotechnics. London: Unwin Hyman. Rapoport, V. & Young, A.G. (1987). Foundation Performance of Jack-Up Drilling Units,Analysis of Case Histories. International Conference on Mobile Offshore Structures. Roscoe, K.H. & Burland, J.B. (1968). On the generalized stress-strain behaviour of wet clay. Engineering Plasticity, Eds. J. Heyman, F.A. Leckie, Cambridge Univ. Press. 535– 609. Svano, G. & Tjelta T. I. (1996). Skirted spud-cans – Extending operational depth and improving performance. Marine Structures 9(1 Spec. Iss.): 129–148. Teh, K.L. (2007). Punch-through of spudcan foundation in sand overlying clay. PhD thesis, National University of Singapore. Teh, K.L., Byrne, B.W. & Houlsby, G.T. (2006). Effects of seabed irregularities on loads developed in legs of jack-up unit. 1st Jack-up Asia Conference, Singapore. Teh, K.L., Gan, C.T., Hu, E.H.J., Leung, C.F. & Chow, Y.K. (2008). A comparison of penetration behaviour of spudcan and skirted footing in sand overlying clay. 2nd Jack-up Asia Conference, Singapore.

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Numerical study of piping limits for suction installation of offshore skirted foundations and anchors in layered sand L.B. Ibsen Aalborg University, Denmark

C.L. Thilsted Dong Energy Power, Denmark

ABSTRACT: Skirted foundations and anchors have proved to be competitive solutions for various types of fixed offshore platforms, subsea systems and an attractive foundation alternative for offshore wind turbines. One main design challenge for skirted structures in sand is to penetrate the skirted deep enough to obtain the required capacity. In order to overcome the high penetration resistance in sand suction assisted penetration is needed. Suction installation may cause the formation of piping channels, which break down the hydraulic seal and prevent further installation. This paper presents a numerical study of failure limits during suction installation in respect to both homogenous and layered soil profile. A numerical flow analysis is performed to determine the hydraulic gradients developing in response to the suction applied, and the results are presented as simple closed form solutions useful for evaluation of suction thresholds against piping. These closed form solutions are compared with large scale tests, performed in a natural seabed at a test site in Frederikshavn, Denmark. These solutions are also valid for penetration studies of other offshore skirted foundations and anchors using suction assisted penetration in homogeneous or layered sand. Due to the complexity of the domain and the governing differential equation, the problem is solved numerically. A numerical solution can be obtained using either finite difference or finite element methods. In this paper, the problem is solved using the commercial finite difference program FLAC3D (Itasca, 2005).

1

INTRODUCTION

More than 485 suction anchors, had been installed for anchoring floaters at more than 50 different sites by the year 2004 (Andersen et al. 2005). Most of these anchors are in clay, but some are also in sand or layered soils. Examples of skirted foundations in sand are the offshore steel platforms at Draupner E and Sleipner T sites in the North Sea (Tjelta 1995). Skirted foundations in sand can also be used to increase the moment fixity and can be an attractive foundation alternative for offshore wind turbine as the bucket foundation installed in Frederikshavn has shown. (Ibsen 2008). In order to overcome the high penetration resistance in sand, suction assisted penetration is needed. The suction creates a pressure differential across the caisson lid, effectively increasing the downward force on the caisson while reducing the skirt tip resistance. This study has been a part of a research project whose aim is to develop a skirted foundation often referred to as the “bucket foundation” as a foundation for offshore wind turbines. At the time of writing, two bucket foundation have been installed, one at Horns Rev II and the other located in Frederikshavn, Denmark, (Ibsen 2008). Figure 1 shows an installation test of a 4 × 4 m bucket at the test site in Frederikshavn. © 2011 by Taylor & Francis Group, LLC

Figure 1. Installation tests on 4 × 4 m buckets in a natural seabed at the test site in Frederikshavn, Denmark.

The installation of bucket foundation for offshore wind turbines differs for several reasons. Compared to oil and gas jackets, the bucket foundation offers less self-weight to assist penetration and the offshore parks are predominantly located at shallow waters, h > 1.2D. In the first case, simulations were conducted to investigate bucket installation in homogeneous soil, the results are shown in Figure 7a. The second case simulates a bucket installed in sand over a impermeable flow boundary, located in the depth L . The results are shown in Figure 7b.

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Figure 7. The results of the FLAC calculation are plotted as normalized seepage length for exit gradient versus relative penetration. a) Installation in homogenous sand. b) Installation in sand over a flow boundary.

4.1

Installation in homogeneous sand

The following empirical expression is given to approximate the numerical data for the installation in homogeneous sand.

Equation (3) is fitted to two boundaries. For a very small h/D ratio, equation (3) approaches 2.86, a theoretical solution for a sheet-pile wall, suggested by Hansen (1978). For an infinitely long bucket, all the hydraulic head loss occurs inside the bucket with evenly spaced horizontal equipotential lines. Therefore, the normalized length tends to unity. For installation in homogenous sand the internal hydraulic gradients have been investigated by several researchers using finite element programs as Plaxis and SEEP. Senders & Randolph (2009) performed calculations with the finite element programme Plaxis and proposed a similar expression for the exit gradient:

Figure 8. Seepage length for exit gradient versus relative pene-tration predicted by equation (4), (5) and (6).

Figure 8 show that these three different formulations predict similar seepage length for penetrations of practical interest 0.1 ≤ h/D ≤ 1. 4.2

For very small h/D ratio equation (4) approaches π, which is a theoretical solution for a sheet-pile wall, suggested by Scott (1963). Feld (2001) performed calculations with the finite element program SEEP and proposed that the seepage length could be estimated as:

Installation in sand over a flow boundary

The following empirical expression is given to approximate the numerical data for the installation in layered sand:

where (s/h)ref is calculated from equation (3). It is seen that equation (6) approaches equation (3) if © 2011 by Taylor & Francis Group, LLC

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the distance to the flow boundary L is large in comparison to the diameter of the bucket D. 5

CRITICAL SUCTION

The formation of local piping channels occurs when the exit hydraulic gradient, next to the caisson wall, exceeds the gravitational force, and thereby reduces the effective stresses to zero. The critical gradient is:

The exit hydraulic gradient i can also be expressed in terms of the applied suction p and the seepage length s as: Figure 9. Normalized critical suction versus relative penetration. The critical suction is calculated with different ratios L /D.

where γw is the unit weight of water and γ’ is the submerged unit weight. The critical suction resulting in formation of local piping channels are therefore

By combining equation (6) with equation (9) the critical suction can be expressed as:

bucket 5 the flow boundary was at a depth of 1.2 m. This increases the suction capacity and the bucket was penetrated with the highest applied suction without any failure occurring. It is shown that these thin silt layers act as flow boundaries and increase the suction thresholds against piping.

7 Figure 9 shows the critical suction calculated by equation (10) with different ratios L /D. If L /D is large then the critical suction approaches the threshold for penetration in homogeneous sand. It is also seen that the presence of a flow boundary will increase the threshold where critical suction will occur. 6

PREDICTION OF FIELD TEST DATA

In Figure 10, the suction needed to install the bucket is plotted against equation (3) and (10). The figure shows that suction close to or higher than critical, predicted by equation (3), can be applied without significant consequences. This is particularly seen in the installation test with bucket 5. It is seen that the suction needed to overcome the resistance during the installation of the bucket 2 never violated the critical suction predicted by equation (10) with the flow boundary at 2.7 m. This was not the case in the installation test with bucket 4. At a depth of 1.56 m the applied suction violated the failure criterion predicted by equation (10) and piping channels were formed and observed during the test. At the test with © 2011 by Taylor & Francis Group, LLC

425

CONCLUSION

By comparing the numerical studies with the installation tests it is shown that it is the exit gradient next to the skirt which controls when piping will occur. For installation in homogeneous sand, the internal hydraulic gradients have been investigated by several researchers using computer programmes such as Plaxis, SEEP and FLAC.These studies have resulted in different formulations, but the empirical expressions predict similar critical suctions for skirt penetrations of practical interest. However, experience from installation of prototype foundations have shown that gradients close to critical, predicted by the expressions for homogenous sand, can be applied without significant consequences. The same was observed in the field test reported in this paper. It is stated that the presence of thin silt layers will act as flow boundaries and increase the suction thresholds against piping. The influence of the flow boundary was studied in this paper. The results are presented as simple closed form solutions and shown to predict thresholds against piping in homogeneous or layered sand. Future studies have to be performed in order to establish the thresholds against piping when the skirt penetrates through a flow boundary.

Figure 10. Installation tests analyzed using equation 10 with the flow boundaries interpret from the CPT tests in Figure 4.

REFERENCES Andersen, K.H., Murff, J.D., Randolph, M.F., Clukey, E., Erbrich, C., Jostad, H.P., Hansen, B., Aubeny, C., Sharma, P., and Supachawarote, C. (2005). “Suction anchors for deepwater applications.” Proc., INT Symp. On Frontiers in offshore Geotechniques (ISFOG). Keynote Lecture, Perth, Western Australlia, p. 3–30. Feld, T (2001). “Suction bucket, a new innovative foundation concept applied to offshore wind turbines.” Aalborg university, Aalborg. Hansen, B. (1978). Geoteknik og fundering del II. Laboratoriet for fundering. DTH. (In Danish). Ibsen, L.B. (2008). Implementation of a new foundations concept for Offshore Wind farms. Proc. Nordisk Geoteknikermøte nr. 15 NGM 2008, 3–6 September 2008 Sandefjord, Norge, 1–15.


Itasca (2005). “FLAC3D – Fast lagrangian analysis of continua: Fluid-Mechanical Interaction”, Itasca Consulting Group Inc., Minneapolis, USA. Scott, R.S. (1963). Principles of soil mechanics. AddisonWesly Publiching Company, Inc. Senders. M. and Randolph M. F. (2009) “CPT-Based Method for the Installation of Suction Caissons in Sand” Jour. of Geotechnical and Geoenvironmental Enginnering. Senepere, D. and Auvergne, G. A. (1982) “Suction anchor piles – a proven alternative to driving or drilling.” Proc., 14th Offshore Technology Conf., Houston, Texas, 483–493. Tjelta T. I. (1995) “Geotechnical experience from the installtion of the Europipe jacket with bucket foundations” Proc., Offshore Technology Conf., Houston, Texas, Paper No. 7795.

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Shallow foundation performance in a calcareous sand B.M. Lehane The University of Western Australia

ABSTRACT: The paper presents results from eight small scale load tests performed on shallow footings founded on an unsaturated calcareous sand at a site located about 100 km North of Perth, Western Australia. These tests highlight inadequacies of traditional bearing capacity approaches and show the potential for more reliable estimates of bearing capacity through direct correlations with CPT data. Both the shear stiffness and creep characteristics of this particular calcareous sand deposit are shown to be consistent with those inferred from footing tests on a siliceous sand.

1

INTRODUCTION

Field research to improve predictions of the performance of shallow foundations in sand has largely concerned siliceous sands because of their predominance in urban areas around the world. However, virtually no field load test data have been reported for shallow foundations in offshore calcareous sands, which are abundant in oil and gas development regions. This shortage of data coupled with the well known high compressibility of calcareous sands lead to relatively conservative shallow foundations designs for offshore facilities. The field testing programme reported in this paper provides some redress for the data shortage and adds insights into a foundation’s stiffness and capacity in a uniformly graded calcareous sand. Figure 1. View of Ledge Point test site.

2 TEST PROGRAMME AND GROUND CONDITIONS AT TEST SITE Eight footings load tests, designated F1 to F8, were conducted at a coastal aeolian calcareous sand site in Ledge Point (see Figure 1), located about 100 km north of Perth, Australia. The tests, details of which are summarised in Table 1, involved loading of three footing types: (i) 300 mm diameter (D), 50 mm thick steel plates, (ii) 300 mm high reinforced concrete cylinders with D = 580 mm and (iii) 600 mm square, 110 mm thick reinforced concrete bases. A CPT truck was employed to provide the reaction for the load tests (see Figure 2). The sand around each footing was removed to a depth E (provided in Table 1); loose sand in its vicinity was then placed to give an overburden height above footing level with the height (H) provided in Table 1. Footings were ‘bedded-in’ and load tested incrementally to the maximum available reaction load of 200 kN or to a maximum footing settlement of at least 20 mm. Results for F3 were only recorded to a © 2011 by Taylor & Francis Group, LLC

maximum settlement of 7 mm due to a data-logger malfunction. The Ledge Point sand is uniformly graded with a mean effective particle size (D50 ) of 0.25 mm, a uniformity coefficient of about 2 and a calcium carbonate content of 90%. An electronic microscope image of the sand is reproduced in Figure 3, where the hollow structure and high aspect ratio of grains is apparent. Maximum and minimum void ratios of 1.21 and 0.90 were recorded by Sharma (2004), who measured a critical state friction angle for the sand of 39◦ . Thirteen, closely spaced, Cone Penetration Tests (CPTs) were conducted to a depth of about 4.5 m at the locations, relative to footing tests, indicated on Figure 4. Each CPT revealed a highly stratified profile comprising 0.5 m–1.5 m thick layers with high CPT end resistances (qc ) of ∼15 MPa alternating with layers of comparable thickness but relatively low qc values of ∼2 MPa. No systematic spatial variation of these

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Table 1.

Footing load test details.

Footing number

D or B (m)

E (m)

H (m)

Zone of influence (m)

qc,avg (MPa)

E at s/D = 1% (MPa)

q at s/D = 20%, qf (kPa)

q at s/D = 10%, q0.1 (kPa)

F1 F2 F3 F4 F5 F6 F7 F8

D = 0.3 B = 0.6 D = 0.3 B = 0.6 D = 0.58 D = 0.58 D = 0.3 D = 0.3

0.5 0.3 0.3 0.3 0.4 0.35 0.2 0.4

0.15 0.11 0.15 0.11 0.2 0.17 0.05 0.15

0.5–0.9 0.3–1.0 0.3–0.7 0.3–1.0 0.4–1.1 0.35–1.0 0.2–0.65 0.4–0.8

6.1 8 7.1 7.2 11 7.2 4.3 8.3

32 15 21 21 15 17 16 41

1100 1100 1150 1420 ∼1500 1100 1250 1350

1000 880 1075 1175 1300 930 930 1120

Figure 2. Load testing a square footing.

Figure 4. Relative locations of footing tests and CPTs.

Figure 3. Ledge Point calcareous sand.

layers was evident, and it would appear that the profile arose due to the aeolian deposition of loose sand within very weakly cemented or much denser, strongly undulating calcareous terrain. Pits excavated for the footings were easily excavated (i.e. cementation levels appeared absent or very low) and their sides remained stable below a depth of 0.3 m due to the presence of © 2011 by Taylor & Francis Group, LLC

some suction. The CPTs indicated that the depth to the water table was in excess of 4.5 m. Figure 5 presents the CPT qc values recorded within the depth of influence of each footing test. CPTs were not performed immediately adjacent to F1 and F2 and, for these cases, the profile plotted on Figure 5 is an average of CPTs conducted close by (e.g. CPT1 and 4 were averaged to produce the F1 profile). Figure 5 indicates that, in the upper horizons (of most relevance to the footings), qc generally increases with depth to an average value of about 10 MPa at 0.8 m depth. The footings’ zone of influence, estimated using the recommendations of Burland & Burbidge (1985), is listed in Table 1 along with the average qc value within this zone (qc,avg ). 3

FOOTING BEARING CAPACITY

The variations of the applied bearing pressure (q) with the settlement (s) normalised by the footing width or

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stress mobilised at settlement ratios of 10% and 20% were estimated by hyperbolic extrapolation (if necessary) of the curves on Figure 6; bearing pressures at a nominal (large) settlement ratio of 20% are referred to here as the footing bearing capacity (qf ). Equivalent footing elastic stiffness values, derived using the equation for a rigid punch, are provided in Table 1 at a typical working stress settlement ratio of 1%. The estimated qf values are about 1300 ± 200 kPa for all footings, irrespective of the footing width, shape or embedment. This trend is unexpected in a granular material but can be explained relatively well using the standard bearing capacity equation (e.g. as provided in API 2007) by a employing a non-zero cementation component (c ) of approximately 6 kPa and a critical state friction angle of 39◦ ; this cementation component contributes 75% of the total bearing capacity. The coefficient of variation for the ratio of predicted to measured capacity (Qp /Qm ) increases from 0.14 when c = 6 kPa to 0.27 if c is assumed zero. For the latter case (c = 0), a mean Qp /Qm ratio of unity is achieved for a friction angle of 46◦ . This angle is far in excess of the angle of tan−1 (0.66 tan φ ) = 28◦ recommended by Finnie & Randolph (1994) for footings on un-cemented calcareous soil. The mis-match with standard bearing capacity theory, if c = 0, may also be due to the size/stress level effect on the Nγ factor identified by De Beer (1965). Footing capacities at Ledge Point did not increase with embedment in the linear manner predicted by the standard bearing capacity equation. This characteristic has also been observed by Lehane (2010) in a review of available footing bearing capacity data and suggests that the mechanism of collapse is more akin to hemispherical expansion than to the classical Prandtl shear mechanism. Given this trend and the difficulties in assessing in-situ levels of cementation (noting that low levels of c can have an enormous influence on capacity), it is suggested here that a more reliable estimate of capacity may be obtained using a simple correlation with the CPT qc value. The ratios of bearing stresses at a settlement to width ratio of 10% (q0.1 ) to the average qc value in the footings’ zone of influence (qc,avg ) are plotted against the qc,avg on Figure 7. Lehane & Randolph (2002) show that a similar trend is indicated by q0.1 values for bored piles and Lehane (2009) established the following relationship from measured q0.1 values at the base of piles in siliceous sand (noting that all terms in this relationship have the same units of stress):

Figure 5. CPT qc values within zone of influence of the test footings.

diameter (s/B or s/D) measured in each of the eight footing tests are presented in Figure 6. Loads were applied in increments with each increment applied for 5 minutes; typically ten increments were employed to achieve the ultimate or maximum load. The bearing © 2011 by Taylor & Francis Group, LLC

with f1 = 2.4 ± 0.7 where σv is the average in-situ vertical effective stress within the zone of influence and pa is atmospheric pressure (=100 kPa). This same relationship is plotted on Figure 7, employing f1 = 2.4 and for a typical range of σv values. It is evident that equation (1) also provides a very good representation of the variation observed in the calcareous sand at Ledge Point.

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Figure 6. Bearing pressure vs. normalised settlement measured in eight footing tests.

Figure 8. Footing settlements during load increments (Footing F6). Figure 7. Dependence of q0.1 on qc and stress level.

4

CREEP SETTLEMENT

Calcareous sands are generally assumed to creep at a higher rate than siliceous sands because of the more compressible nature of their sand grains. The footing experiments showed that, after the settlement that occurred during application of the load increment, © 2011 by Taylor & Francis Group, LLC

creep settlements (sc ) varied approximately with the logarithm of time (t). A typical variation, as indicated by F6, is shown on Figure 8. The creep rate varied approximately with the square of the mobilised strength (q/qf ) or inverse of the factor of safety. Defining a creep coefficient, m, as the slope of relative settlement (s/D) or (s/B) versus natural logarithm of time (s/D/lnt), values for m were determined for each load increment in each footing test. Representative m

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(ii) There is no correlation between E0.1 and qc,avg e.g. the E0.1 /qc,avg ratio varies from 1.4 for F4 to 5.2 for F1. (iii) This range of E0.1 /qc,avg ratios is comparable to that employed in the design of footings on siliceous sand. The higher compressibility of the Ledge Point sand evidently had a low influence on the stiffness operational in these tests. The equivalent stiffnesses of the footings are approximately twice those measured in shallow footing tests on siliceous sand by Lehane et. al. (2009) at the Shenton Park test site in Perth. The average CPT qc value within the footings’ zone of influence at Shenton Park was about 3.5 MPa and hence less than half of the Ledge Point averages (see Table 1). It is clear therefore that the Ledge Point calcareous sand does not exhibit stiffness characteristics which are dissimilar to equivalent siliceous sand deposits.

6

Figure 9. Variation of creep parameter (m) with mobilised strength.

values are plotted on Figure 9 which also plots the following equation:

It is seen that Equation (2) provides an approximate upper-bound to the Ledge Point m data. It is noteworthy that Lehane et al. (2009) showed that the identical equation provided a best-fit to creep data from footing tests on siliceous sands in Perth. It would therefore appear that the Ledge Point sand does not creep more than the siliceous dune sand in Perth. Both sands are above the water table and recent triaxial tests at UWA suggest that the degree of saturation over a certain range can have a marked effect on the shear creep characteristics. The same research suggested that equation (2) provides an upper-bound to the likely level of creep when the sand is fully saturated.

CONCLUSIONS

The footing tests performed at Ledge Point showed that the high calcium carbonate content of this sand (and hence high grain compressibility) did not lead to footing responses that differed significantly from those expected in a siliceous sand. It is shown that footing bearing capacities are best assessed using a correlation with the CPT qc value (or pressuremeter limit pressure) rather than with standard bearing capacity formulae. The operational stiffness and creep characteristics of the near surface (partially saturated) Ledge Point sand are comparable to those of siliceous sands under similar conditions.

ACKNOWLEDGEMENTS The author would like to acknowledge the contribution of four final year students at UWA namely Scott Doncon, Ben Hall, Nicola Reeves and Yuli Yao, all of whom assisted in the execution and interpretation of the tests described here. The technician team at UWA, led by Jim Waters, and the CPT contractor, Probedrill Pty Ltd, who provided free use of their CPT truck, are also gratefully acknowledged. REFERENCES

5

FOOTING STIFFNESS

A detailed analysis of the stiffness of the Ledge Point footings is outside the scope of this paper. Some general trends can, however, be drawn from Figure 6 and Table 1: (i) The equivalent linear stiffness at a settlement to width ratio of 1% (E0.1 ) is approximately 18 MPa for all footings except F1 and F8, which are almost twice as stiff. © 2011 by Taylor & Francis Group, LLC

American Petroleum Institute (API) 2007. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – Working Stress Design, RP2A, Revisions/Edition: 21, Washington. Burland, J.B. and Burbidge, M.C. (1985). “Settlement of foundations on sand and gravel.” Proc. Design and Construction, Institution of Civil Engineers, Vol. 78, December, 1325–1381. De Beer, E. (1965). The scale effect on the phenomenon of progressive rupture in cohesionless soils. Proc. 6th Int. Conf. Soil Mech Found. Eng., 2, Montreal, Balkema, 13–17.

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Doncon S. (2009). Settlement of shallow foundations on calcareous sand. Final Year Honours thesis, University of Western Australia. Finnie, I.M. and Randolph, M.F. (1994). Bearing response of shallow foundations in uncemented calcareous soil. Proc. Centrifuge ’94, 1, Singapore, Balkema, Rotterdam, 535–540. Lehane B.M., Doherty J.P. and Schneider J.A. (2009). Settlement prediction for footings on sand. Keynote Lecture, Proc. 4th International Symposium on deformation characteristics of geomaterials, Atlanta, 1, 133–152, IOS press, The Netherlands. Lehane B.M. and Randolph M.F. (2002). Evaluation of a minimum base resistance for pipe piles in sand.


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J. Geotechnical & Geoenvironmental Engrg. ASCE, 128 (3), 198–205. Lehane B.M. (2009). Relationships between axial capacity and CPT qc for bored piles in sand. Keynote Lecture, Proc. 5th International Symposium on deep foundations on bored and auger piles, Ghent, 1, 61–76, Taylor Francis Group, UK. Lehane B.M. (2010). Assessing the bearing capacity of footings on sand using in-situ test data. (in preparation) Sharma, S.S. (2004). Characterisation of cyclic behaviour of calcite cemented calcareous soils. PhD Thesis, The University of Western Australia.


A numerical study of the vertical bearing capacity of skirted foundations D.S.K. Mana, S. Gourvenec & M.F. Randolph Centre for Offshore Foundation Systems, UWA, Crawley, Australia

ABSTRACT: Finite element analysis is used to investigate the vertical bearing capacity of circular skirted foundations considering the effect of embedment ratio, foundation-soil interface roughness and soil strength heterogeneity. The effect of idealising a skirted foundation as a solid rigid plug and idealising geometry to conditions of plane strain are also addressed through comparison of bearing capacity factors and kinematic mechanisms accompanying failure. A closed-form expression is presented that enables prediction of bearing capacity factors for circular skirted foundations over a practical range of embedment ratio, skirt-soil interface roughness and soil strength heterogeneity, to within ±2.5% of the finite element calculations.

1

INTRODUCTION

Skirted foundations are used to support or moor a variety of offshore structures, such as gravity based structures, tension leg platforms, jacket structures, storage tanks and various sub-sea infrastructure (such as manifolds and pipeline end terminations – PLETs). Investigation of bearing capacity of shallow foundations has typically considered rigid plugs or buried plates (e.g. Martin 2001, Martin & Randolph 2001, Bransby & Randolph 1999, Salgado et al. 2004, Edwards et al. 2005, Gourvenec 2008). The effect of a deformable soil plug (as confined by a skirted foundation) has received less attention, and analysis has been limited to plane strain conditions and a fully rough foundation-soil interface (Yun & Bransby 2007, Gourvenec & Barnett 2010). A particular interest with skirted foundations is whether or not an internal mechanism will develop within the soil plug, hence reducing bearing capacity (unless provision is made for internal skirts or soil improvement within the soil plug). An internal mechanism would not be expected to develop in soils with uniform shear strength with depth, i.e. the soil plug would be expected to displace as a rigid body. However, in soils with increasing shear strength with depth, a failure mechanism may develop within the soil plug reducing bearing capacity. From existing theory, it would be anticipated that an internal failure mechanism would only develop when a Hill-type mechanism governs failure, as opposed to a Prandtl-type mechanism (illustrated in Figure 1). It is well established that a Hill-type mechanism governs failure of smooth-based surface foundations and becomes the critical failure mode for rough-based surface foundations when soil strength heterogeneity is significant (Kusakabe et al. 1986). Martin & Randolph (2001) noted that a Hill-type mechanism is always critical for smooth-based shallowly embedded © 2011 by Taylor & Francis Group, LLC

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Figure 1. Kinematic failure mechanisms of shallow foundations under vertical load (modified from Kusakabe et al. 1986).

foundations and is occasionally critical for roughbased shallowly embedded foundations, notably with increased soil strength heterogeneity – as for surface foundations. Skirted foundations are essentially rough-based, due to the soil-soil interface at foundation level and it would be of use to investigate the embedment ratio, as a function of soil strength heterogeneity, at which an internal mechanism will or will not develop. This paper presents results from an investigation of bearing capacity of circular and strip shallow foundations. Initially, results of skirted foundations, with a deformable soil plug and embedment modelled by a rigid plug are compared for a uniform shear strength and a highly heterogeneous shear strength profile. Subsequently, a parametric study considering the effect on bearing capacity of a range of soil strength heterogeneity and skirt-soil interface roughness is presented for circular skirted foundations. 2

FINITE ELEMENT MODELS

Small strain finite element analyses were carried out using the software Abaqus (Dassault Systèmes 2009). Shallow foundations with embedment provided either by a peripheral skirt or a rigid plug were considered in axi-symmetry and plane strain. Embedment ratios

Figure 3. Bearing capacity factor for rough-sided circular skirted foundations and rigid solid plugs – kD/sum = 0. Figure 2. Finite element mesh, d/D = 0.5.

(d/D or d/B) of 0 (surface), 0.1, 0.2, 0.3 0.5 and 1 and skirt wall thickness (where relevant), t/D = 0.008 were considered, where d is the skirt embedment depth, D is the diameter of the circular foundation and B is the breadth of foundation in plane strain. The soil was modelled as a linearly elastic perfectly plastic Tresca material with a submerged unit weight γ = 6 kN/m3 . Within the elastic regime, the soil had a stiffness ratio E/su = 500, where E is the Young’s modulus of the soil. Poisson’s ratio ν was taken as 0.499. Within the plastic regime, uniform undrained shear strength and linearly increasing strength with depth were considered. Linearly increasing shear strength with depth is described by su = sum + kz, where sum is the undrained shear strength at the mudline and k is the gradient of increase in shear strength with depth z and the degree of heterogeneity is described by the dimensionless group kD/sum . A range of soil strength heterogeneity 0 (uniform with depth) ≤ kD/sum ≤ 20 were considered. The foundation-soil interface was modelled with a contact surface, rough in shear and perfectly bonded preventing slip or separation. For modelling interface roughness, a narrow band of soil elements, the same width as the skirt thickness, was considered adjacent to the outer face of the skirt. This narrow band of soil elements was given shear strength equal to α times the intact shear strength in the rest of the soil mass at that depth. (This approach is necessary as a contact surface with constant interface friction factor cannot be defined in Abaqus.) A typical axi-symmetric finite element mesh is shown in Figure 2. Similar mesh discretisation was maintained for different embedment ratios and for the plane strain meshes. The width of the modelled soil zone was more than three times the diameter from the centre of the skirt and depth was more than six times the diameter of the skirt, to prevent boundary effects on the bearing capacity. Roller supports were provided along the vertical boundaries and along the base of the mesh. Second-order bi-quadratic continuum elements with reduced integration were used to model the soil © 2011 by Taylor & Francis Group, LLC

Figure 4. Bearing capacity factor for rough-sided strip skirted foundations and rigid solid plugs – kD/sum = 0.

(CAX8R and CPE8R).The foundations were modelled as rigid bodies, pre-embedded (i.e. installation was not modelled). Each foundation was brought to failure by a displacement-controlled load path. A reference point (RP) for applying and recovering loads and displacements to the foundation was specified along the centre line of the foundation at skirt tip level. In order to eliminate the effect of foundation weight on the net end bearing capacity of the foundation, the part of the foundations below the mudline were prescribed a unit weight equal to that of soil and the part above was modelled as weightless. 3 3.1

RESULTS AND DISCUSSION Skirted foundations and rigid solid plugs – kD/sum = 0

Figures 3 and 4 show the calculated vertical bearing capacity factors, Nc = qult /su , for rough-sided circular and strip skirted foundations and rigid solid plugs in soil with uniform undrained shear strength, i.e. kD/sum = 0. A vertical bearing capacity factor Nc = 5.95 was calculated for the surface circular foundation, under predicting the exact solution of 6.05 by less than 2% (Cox et al. 1961). This slight under prediction is attributed to the rounding of vertices of the

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Figure 5. Displacement contours for strip (left) and circular (right) rough skirted foundations with kD/sum = 0 (Contour interval δ/D = 0.025).

hexagonal Tresca yield surface in Abaqus (Taiebat & Carter, 2008), possibly exacerbated by the larger deformation of soil elements close to the foundation in axi-symmetry compared with plane strain (Gourvenec et al. 2006). An ultimate vertical bearing capacity Nc = 5.22 was calculated for the surface strip foundation which over predicts the exact solution of 5.14 by less than 2%. The calculated bearing capacity factors for the embedded foundations are validated by comparison with available upper and lower bounds for rigid solid plugs shown in Figures 3 and 4 (Martin 2001, Bransby & Randolph 1999). The bearing capacity factor increases with increasing embedment ratio, as would be expected, and the bearing capacity factors for skirted foundations are coincident with those for a rigid solid plug of equivalent embedment for both the circular and strip foundations (as also noted by Yun & Bransby (2007) for strip foundations). The non-linearity of the relationship between bearing capacity and embedment ratio for the circular foundation geometry (as seen in Figure 3) indicates the transition of failure mechanisms as illustrated in Figure 5; from a traditional Prandtl-type surface failure mechanism (a & b), to an annular flow mechanism where the angle of exit at the soil surface tends to 90◦ (c) to a confined mechanism (d & e). The almost linear relationship between bearing capacity and embedment ratio for the plane strain case (Figure 4) is reflected in the similarity of the failure mechanisms of a Prandtl-type surface failure irrespective of embedment ratio (Figure 5). The more extensive failure mechanisms, both vertically and laterally, under plane strain conditions compared to axi-symmetry are clear from Figure 5. Prandtl-type mechanisms governed failure of both the solid rigid foundations and the skirted foundations i.e. the soil plug displaced as a rigid body, consistent with the coincidence of the bearing capacity factors presented in Figures 3 and 4. 3.2 Skirted foundations and rigid solid plugs – kD/sum = 20 Figures 6 and 7 show the calculated vertical bearing capacity factors, Nc = qult /su0 , where su0 defines the undrained shear strength at foundation level, for © 2011 by Taylor & Francis Group, LLC

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Figure 6. Bearing capacity factor for rough-sided circular skirted foundations and rigid solid plugs – kD/sum = 20.

Figure 7. Bearing capacity factor for rough-sided strip skirted foundations and rigid solid plugs – kD/sum = 20.

skirted foundations and rigid solid plugs in a soil with linearly increasing undrained shear strength with depth, kD/sum = 20 (considered to represent the high end of soil heterogeneity commonly encountered). The calculated bearing capacity factors are validated by comparison with upper and lower bounds for rigid circular plugs (Martin 2001). For the strip foundation, no theoretical solution is available for kD/sum = 20, but an upper bound solution for rigid strips for kD/su0 → ∞ is shown (Bransby & Randolph 1999). The vertical bearing capacity factors for the skirted foundations are coincident with those for rigid solid plugs of equivalent embedment for d/D > 0.23 and 0.28 for circular and strip geometry respectively. For smaller embedment ratios, lower bearing capacity factors are calculated for the skirted foundations; more so for the strip foundation (−8% for d/D = 0.1) compared with the circular foundation (−5% for d/D = 0.1). Examination of the kinematic mechanisms accompanying failure, shown in Figure 8 for the circular case, reveals an internal mechanism inside the skirt for d/D < 0.3, consistent with the diminishing vertical bearing capacity relative to that for the solid plug foundation. It is interesting to consider shape factor, sc = Nc,circle /Nc,strip , as a function of soil strength heterogeneity and embedment ratio. Salençon & Matar (1982) and Houlsby & Wroth (1983) present shape

Figure 8. Displacement contours for rough circular solid plugs (left) and skirted foundations (right) respectively – kD/sum = 20 (Contour interval δ/D = 0.025).

Figure 10. Shape factor as a function of embedment ratio and soil strength heterogeneity.

Figure 9. Shape factor as a function of soil strength heterogeneity for surface foundations.

Figure 11. Bearing capacity factor for circular skirted foundations with varying interface roughness – kD/sum = 0.

factors for surface foundations as a function of heterogeneity parameter kD/sum based on plasticity solutions. Both show diminishing shape factor with increasing shear strength heterogeneity and report shape factors of less than unity for kD/sum > 2. The results from this study indicate sc = 0.83 for d/D = 0 and kD/sum = 20, falling in the line of extrapolation of Houlsby and Wroth’s data (Figure 9). FEA of surface circular foundations with varying kD/sum reported by Gourvenec & Randolph (2003) also agreed well with the lower bound solutions of Houlsby and Wroth. Thus, as reported by Martin & Randolph (2001), the characteristic solutions given by Houlsby & Wroth (1983) are exact, within the accuracy of the numerical implementation, rather than just a lower bound (LB). Figure 10 shows shape factor as a function of embedment ratio for kD/sum = 0 and 20, showing increasing shape factor with increasing embedment ratio and reducing shape factor with increasing soil strength heterogeneity. For kD/sum = 20, the shape factor is less than unity only for d/D < 0.2. Shape factor clearly vary with both embedment ratio and soil strength heterogeneity in a complex manner (resulting from the different kinematic mechanism governing failure), thus highlighting the benefit of considering foundation shape explicitly rather than relying on © 2011 by Taylor & Francis Group, LLC

shape factors to adjust bearing capacity calculations based on plane strain conditions.

3.3

Circular skirted foundations – Effect of foundation-soil interface roughness and shear strength heterogeneity

In the foregoing, a single heterogeneous soil profile was considered and the foundation-soil interface was assumed to be fully rough, with no slip or separation of soil at the interface permitted. The purpose of the analyses was to illustrate different governing modes of failure and the effect of idealising a skirted foundation as a solid rigid plug and idealising geometry to conditions of plane strain. A parametric study was subsequently undertaken to investigate the effect of interface roughness and degree of soil strength heterogeneity on bearing capacity of circular skirted foundations. A range of interface friction αsu , with α between 0 (frictionless) and 1 (fully rough) for shear strength profiles with kD/sum between 0 (uniform) and 20 was considered. Figure 11 shows bearing capacity factors, Nc = qult /su0 , against embedment ratio as a function of interface roughness for the case of uniform shear strength profile (kD/sum = 0). Lower bearing capacity is mobilised with reduced skirt-soil roughness

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Figure 12. Variation of bearing capacity factor with skirt-soil interface roughness as a function of embedment ratio kD/sum = 0.

as would be expected, and the non-linearity between bearing capacity factor and embedment ratio becomes more marked with reducing skirt-soil roughness. Figure 12 shows the data from Figure 11 presented in an alternative form as bearing capacity factor against friction factor as a function of embedment ratio, giving approximately linear relationships, as shown by the dashed lines of best fit. Equation 1 defines the best fit lines giving the bearing capacity factor for a given interface roughness, Nc,α as the sum of the bearing capacity factor for a smooth footing (representing the intercept on the vertical axis in Figure 12) and an additional component due to skirt roughness for varying embedment; clearly the effect of skirt roughness becomes increasingly significant with increasing embedment. Equation 1 predicts bearing capacity factors to within 2.5% of the finite element calculations.

Bearing capacity of shallow foundations has been alternatively presented as the sum of base resistance and shaft resistance (e.g. Byrne & Cassidy 2002,Yun & Bransby 2007), analogous to the calculation of capacity of a deep pile foundation. This approach is only valid if the base and shaft resistance can be decoupled, which is only be the case if a single prevalent mechanism governs failure independent of skirt-soil interface roughness.Yun & Bransby (2007) present upper bound analyses that indicate that this is not the case and that base bearing may be 10% lower for a smoothsided foundation than a rough-sided foundation with d/D = 1. Observations from the finite element analyses in this study also showed the governing failure mechanism is dependent on interface roughness. Bearing capacity factors for circular skirted foundations with varying interface roughness were additionally calculated for conditions of soil strength heterogeneity given by kD/sum = 2, 5, 10, 15 and 20. Although the relationship between bearing capacity factor and embedment ratio differ in form for cases of linearly increasing shear strength with depth compared with the case of uniform shear strength (as shown by © 2011 by Taylor & Francis Group, LLC

Figure 13. Bearing capacity factors for smooth-sided circular skirted foundations, Nc,α=0 , in soils with varying soil strength heterogeneity.

Figure 14. Coefficient C for Equation 2.

comparing Figures 3 and 6), a linear relationship is observed when the data is presented as bearing capacity factor against interface roughness as a function of embedment ratio (as illustrated in Figure 12 for kD/sum = 0). Over the range of kD/sum considered, the relationship between bearing capacity factor and skirt-soil interface roughness can be expressed by a generalised form of Equation 1

where the value of Nc,α=0 and the constant C depend on the value of kD/sum . The bearing capacity factors for smooth-sided foundations, Nc,α=0 , are plotted against embedment ratio for 0 ≤ kD/sum ≤ 20 in Figure 13. (These have also been validated against upper and lower bounds for circular rigid soil plugs (Martin, 2001)). The relationships are not conducive to description by a closed-form expression and must be read from (or interpolated between) the plotted data. The values of the constant C, required for Equation 2, are plotted in Figure 14 as a function of kD/sum and can be described by an exponential expression

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Thus, with Equations 2 and 3 and Figure 13, bearing capacity factors can be predicted for circular skirted foundations over a range of embedment ratios 0 ≤ d/D ≤ 1, with skirt-soil interface roughness 0 ≤ α ≤ 1 in soils with shear strength heterogeneity 0 ≤ kD/sum ≤ 20 to within ±2.5% of the finite element calculations. 4

CONCLUSIONS

Finite element analysis has been used to investigate the vertical bearing capacity of circular skirted foundations as a function of embedment ratio, foundation-soil interface roughness and soil strength heterogeneity and to assess the effect of idealising foundation geometry with a rigid soil plug and to conditions of plane strain. Three-dimensional effects are shown to have a significant influence on bearing capacity of shallow foundations with shape factors varying in a complex manner with foundation embedment ratio and soil strength heterogeneity such that explicit consideration of foundation shape is recommended. The effect of the type of embedment, i.e. solid or skirted, is shown to be less significant than three-dimensional effects with bearing capacity factors for circular skirted foundations a maximum of 5% less than that of an equivalent solid rigid plug. Internal mechanisms are shown to be most prone for lower embedment ratios and high soil strength heterogeneity. In practice, internal mechanisms are likely to be more significant under horizontal loading and overturning than under pure vertical load. A closed-form expression is presented for the calculation of bearing capacity factors for circular skirted foundations over a practical range of embedment ratio, skirt-soil interface roughness and soil strength heterogeneity, to within ±2.5% of the finite element calculations. ACKNOWLEDGEMENT The Centre for Offshore Foundation Systems was established under the Australian Research Council’s Research Centres Programme and is supported by the State Government of Western Australia through the Centres of Excellence in Science and Innovation Program. The work presented in this paper was funded by an ARC grant DP0988904. This support is gratefully acknowledged.

Byrne, B. & Cassidy, M.J. 2002. Investigating the response of offshore foundations in soft clay soils. 21st Int. Conf. Offshore Mech. and Arctic Engng., New York, USA, OMAE2002-28057. Cox,A.D., Eason, G. & Hopkins, H.G. 1961.Axially symmetric plastic deformation in soils. PhilosophicalTransactions of the Royal Society of London (Series A), 254: 1–45. Dassault Systèmes 2009. Abaqus analysis users’ manual, Simula Corp, Providence, RI, USA. Davis, E.H. & Booker, J.R. 1973. The effect of increasing shear strength with depth on the bearing capacity of clays. Géotechnique 23(4): 551–563. Edwards, D.H., Zdravkovic & Potts, D.M. 2005. Depth factors for undrained bearing capacity. Géotechnique 55(10): 755–758. Gourvenec, S. 2008. Effect of embedment on the undrained capacity of shallow foundations under general loading. Géotechnique 58(3): 177–185. Gourvenec, S. & Barnett, S. (2010 accepted, Géotechnique). Undrained failure envelope for skirted foundations under general loading. Gourvenec, S., Randolph, M.F. & Kingsnorth, O. 2006. Undrained bearing capacity of square and rectangular footings. International Journal of Geomechanics 6(3): 147–157. Gourvenec, S. & Randolph, M.F. 2003. Effect of strength non-homogeneity on the shape and failure envelopes for combined loading of strip and circular foundations on clay. Géotechnique, 53(6): 575–586. Houlsby, G.T & Wroth, C.P. 1983. Calculation of stresses on shallow penetrometers and footings. Proc. IUTAM/IUGG Symp. On Seabed Mech., Newcastle upon Tyne, 107–112. Kusakabe, O., Suzuke, H. & Nakase, A. 1986. An upper bound calculation on bearing capacity of a circular footing on a non-homogeneous clay. Soils & Foundations 26(3): 143–148. Martin, C.M. 2001. Vertical bearing capacity of skirted circular foundations on Tresca soil. Proc. Int. Conf. Soils Mech. and Geotech. Engng (ICSMGE), Istanbul Martin, C.M. & Randolph, M.F. 2001. Applications of the lower and upper bound theorems of plasticity to collapse of circular foundations. In (ed.), Proc. 10th Int. Conf. Int. Assoc. of Computer Methods and Advances in Geomech (IACMAG), Tucson, 1417–1428. Salençon, J. & Matar, M. 1982. Capacité portante des foundations superficielles circulaires. Journal de Mécanique théorique et appliquée 1(2): 237–267. Salgado, R., Lyamin, A.V., Sloan, S.W. & Yu, H.S. 2004. Two and three-dimensional bearing capacity of foundations in clay. Géotechnique 54(5): 297–306. Taiebat, H.A. & Carter, J.P. 2008. Flow rule effects in the Tresca model. Computers & Geotechnics 35(3): 500–503. Yun, G. & Bransby, M.F. 2007. The undrained vertical bearing capacity of skirted foundations. Soils & Foundations 47(3): 493–505.

REFERENCES Bransby, M.F. & Randolph, M. F. 1999. The effect of embedment depth on the undrained response of skirted foundations to combined loading. Soils & Foundations 39(4): 19–33.


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The effect of torsion on the sliding resistance of rectangular foundations J.D. Murff & C.P. Aubeny Texas A&M University, Austin, Texas, USA

M. Yang Aker Solutions (formerly Texas A&M University, Houston, Texas, USA)

ABSTRACT: The undrained sliding capacity of a shallow foundation on the sea bed under lateral load is one of the key considerations in its design. Lateral loads may arise from combinations of wind, wave, current, earthquake and/or ice loading and from operational loads. A number of offshore structural systems have recently been developed that are founded on rectangular mats. The existence of torsion will reduce the lateral sliding capacity of the foundation. This paper describes a kinematically admissible model, based on plastic limit analysis, that accounts for the effect of torsion on lateral sliding resistance including the interaction among the passive, active, and side shear components of resistance. The relative effects predicted by the model are supported by finite element analyses. It is demonstrated that the torsion effect can play a very significant role in the foundation design.

1

INTRODUCTION

2

Classical methods for analyzing shallow foundations generally do not include the effects of torsion loading, as such loading is relatively uncommon; however, analyses of a few special cases have been reported. Torsion effects on the sliding resistance of a group of surface footings were addressed by Murff and Miller (1977) using the upper bound method. More recently Finnie and Morgan (2004) used limit equilibrium methods to study torsion-sliding response of rectangular surface footings andYun et al. (2009) have used the finite element method to study the interaction among sliding, torsion and vertical load on surface footings. Recent offshore applications such as LNG facilities in shallow water subjected to asymmetric environmental loads and deepwater pipeline terminals and manifolds subjected to offset pull-in and thermal loads have underscored the importance of torsion effects on rectangular foundations. These foundations normally are equipped with shear skirts so that the foundation is effectively embedded. In some cases a weak horizontal layer below the foundation may exist which is a preferential failure plane. In such cases the lateral resistance includes not only the shear strength of the failure plane but passive and active resistance on the skirts or vertical soil interface and side shear on the skirts or soil interface. To circumvent the requirement for always having to conduct finite element analyses on such complex problems, a relatively simple solution is proposed which employs the upper bound method of plasticity. © 2011 by Taylor & Francis Group, LLC

MECHANISM

The solution approach taken here is an application of the upper bound method of plasticity (Drucker and Prager, 1952). This requires a kinematically admissible mechanism for which the work rate of external forces is equated to the rate of internal dissipation of energy. The unknown force is evaluated and subsequently minimized with respect to parameters defining the mechanism geometry. The failure mechanism employed in this analysis is shown in Figures 1(a), 1(b), and 1(c). The soil defined by DEFG down to a depth d is a rigid block rotating about a vertical axis through point xo , yo . The long wedges shown on each side of the block translate and deform to accommodate the block motion. A brief explanation of the details of the mechanism follows. Figure 1(a) is a plan view of the block DEFG. The zones of the triangular wedges along the sides undergo passive or active failure as indicated by letters “p” and “a”. Figure 1(b) is an elevation view of a cross section HH’ through the rigid block and wedges on sides 1 and 3. Figure 1(c) shows a typical velocity field for a wedge cross section for the velocity component normal to and compatible with the block’s vertical face. The wedge cross section in the example slips along B1-C1 without deforming. The wedge does deform in the direction parallel to the block sides (x direction in this example) to accommodate the variation in the block lateral velocity. The block also has a velocity parallel to its sides but since

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Figure 1. Schematic of sliding-torsion mechanism.

the wedge cannot displace in this direction there must be slip at the interface (A1-B1 in this case). It should be noted that owing to the inclusion of both active and passive zones in the wedge the work rate done by the unit weight of the soil is zero. The sources of dissipation are thus as follows:

where ε1 is the maximum principal tensor strain. The velocities in the wedge are

• Shear deformation in the wedges due to varying

The shear strain rates are thus

velocity along the sides. In the example shown in Figures 1(b) and 1(c) this gives rise to shear strains εxy and εxz while εyz = 0. The dissipation rate for a Tresca yield condition is


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The maximum principal strain rate is then

The dissipation is then integrated over the volume of the wedge in question. It should be noted here that su is taken as a function of z and since the integrand is then a function of z only the dissipation expression can be reduced to a single integral as follows

The subscript RE1 indicates the right end of the wedge on side 1. A similar expression is developed for the other seven wedge ends. • Slip at the base of the rigid block DEFG relative to the rigid soil below the block. The dissipation on the slip surface below the footing is determined by integrating the relative slip velocity over the area below the footing as follows

Note that the integral is independent of whether the wedge is yielding actively or passively. The dissipation in the remaining wedges can be determined in a similar manner. • Slip between the rigid rotation block and the vertical faces of the wedges. The wedges move up relative to the block and the block moves parallel to its edges relative to the wedges. The non-zero slip velocities on the vertical face are then

Taking the resultant and integrating over the vertical interface along the wedge gives

where su is evaluated at depth d. The external work rate of the resultant load F is then

The upper bound estimate of the capacity F is then found by setting the external work rate equal to the ˙ canceling the sum of all dissipation rate terms, D, ˙ and solving for F which gives β’s,

This expression is then minimized with respect to the optimization variables xo , yo , w1 , w2 , w3 , and w4 . Where the depth of the failure plane is obvious, d can be taken as given but it can also be included as an optimization variable. where α is a factor between 0 and 1 indicating the fraction of shear strength mobilized on the vertical interface. The subscript RS1 indicates the resultant velocity on side 1. The dissipation on the other sides is determined in a similar manner. • Slip along wedge ramps (B1–C1 in the example). The resultant velocity up the wedge ramp is then

and the dissipation is thus

The subscript RR1 indicates the resultant velocity on ramp 1. • Slip at the two ends of each wedge relative to the rigid soil outside the wedges. The resultant relative velocity on the ends of the wedge is the same as the slip on the ramps except that x is equal to L/2 or W/2 . The dissipation on the right end of side 1 is then


3

PARAMETRIC STUDY

A brief parametric study was carried out to demonstrate the effect of torsion on sliding capacity for varying aspect ratio (W/L), varying failure surface depth, and varying load angle. Figure 2 summarizes the results of the study for W/L = 0.5 and 1.0 and for d/L = 0.0 and 0.05 for a load angle of 90 degrees. The results are normalized by the sliding and torsion resistance of the respective footings for d = 0, that is by the sliding and torsion results of a simple plate resting on the surface. For the case of d/L = 0.0, the curves for varying W/L have the same limiting values and differ only slightly in shape. For the case where the failure surface is at a depth of 0.05 L the limiting normalized values are, of course, greater than 1.0 due to contributions from the passive and active wedges and the side shear. The shapes of the interaction surfaces however are somewhat similar. For the case of a surface footing the interaction curves are not affected by the load angle. Figure 3 shows the effect of load angle for the embedded footing with a W/L ratio of 1/2. As expected the pure torsion values are unaffected by the load direction. The sliding resistance, however, is significantly affected. The end on loading (0 degrees) gives the smallest resistance due to the smaller end area. The resistance increases with load direction to approximately 45 degrees due to

441

Figure 2. Normalized interaction curves for varying w/l and d/l for a load angle of 90 degrees (broadside).

Figure 4. Comparison of upper bound solution with finite element results.

where

A simpler empirical expression has been determined by curve fitting the analytical solution as follows:

Figure 3. Normalized interaction curves for varying load angle for d/l = 0.5 and W/L = 1/2.

the larger projected areas and then decreases slightly for broadside loading. 4 VERIFICATION One check on the model is to compare model results with available analytical results for limiting cases. The sliding capacity of the surface plate (d = 0.0) is simply Fmax = LWsu which is identical to the model. For the embedded footing a solution can be obtained for purely broadside loading or purely end on loading by adding the plate sliding resistance to that of the active and passive wedges, the wedge ends, and the side shear on faces parallel to the load directions. These too, check exactly. Note that the triangular wedges are only exact solutions for the full plane strain solution with α = 0. For α = 1 the error in the passive/active component for the triangular wedge solution is approximately 10 percent. The overall error is generally much less. It is also possible to obtain a closed form solution for pure torsion, i.e. rotation about the centroid, as follows:


Except for very small errors due to numerical integration, the model results are identical to the closed form solution. Additional checks were made for the maximum torsional capacity with the solutions presented by Yun, et al (2009) and Finnie and Morgan (2004). These were found to check essentially exactly. Figure 4 shows interaction diagrams for the case of a surface footing subjected to combined sliding and torsional loading compared to the solution presented by Yang, et al. (2010). The solution by Yang et al. was obtained for a deeply embedded, infinitely thin plate but the plots normalized by their maximum values are believed to be appropriate for this comparison. As shown in the Figure the differences in the interaction diagrams between the upper bound solutions and the FEM solutions are quite small. It is also noted that the solutions for the square and rectangular footings are in the correct relative order.

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5

SUMMARY AND CONCLUSIONS

In this paper we have presented a solution to the problem of an embedded rectangular footing subjected to sliding and torsion loading. This solution is relevant to recent offshore problems ranging from shallow water LNG facilities to deepwater pipeline terminals and manifolds. The solution presented is capable of considering such complexities as combined torsion and sliding, footing embedment, and non-homogeneous soil strength profiles but is relatively straightforward

to implement on a simple spreadsheet. Example calculations for a range of parameters are presented and various comparisons are made to verify the solution is consistent with other known solutions. REFERENCES Drucker, D. C. and Prager, W. 1952. Soil Mechanics and plastic analysis or limit design. Q. Applied Math., 10: 157–165. Finnie, I.M.S. and Morgan, N. 2004. Torsional Loading of Subsea Structures, Proceedings Fourteenth International Offshore and Polar Engineering Conference, Toulon, France.


Murff, J. D. and Miller, T. W. 1977. Stability of Offshore Gravity Structure Foundations Using the Upper Bound Method, Proceedings, Offshore Technology Conference, Houston, TX., May 1977. Yang, M., Murff, J. D. and Aubeny, C. P. 2010. Undrained Capacity of Plate Anchors Under General Loading, Journal of Geotechnical and Geoenvironmental Engineering, ASCE, Accepted for publication. Yun, G. J., Maconochie, A., Oliphant, J. and Bransby, F. 2009. Undrained Capacity of Surface Footings Subjected to Combined V-H-T Loading, Proceedings Nineteenth International Offshore and Polar Engineering Conference, Osaka, Japan.

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Foundation design challenges of the MCR-A skirted gravity platform L. Tapper & C. Humpheson Arup

B.M. Lehane The University of Western Australia

ABSTRACT: The MCR-A platform is a skirted gravity base platform to be installed in the Caspian Sea offshore Turkmenistan in ground conditions composed of stiff clay. Advanced 3D numerical analysis was required to adequately model the annular geometry and high environmental loading on the substructure. High suction pressures are needed to install the skirts to the depth necessary for foundation stability. The high suction pressures required a complex skirt stiffener arrangement and this posed challenges in quantifying the penetration resistance to be overcome in order to ensure the required skirt penetration could be achieved. In addition, the presence of spudcan footprint ‘craters’ in the direct vicinity of the platform resulting from previous jack-up operations required special design consideration. The approaches carried out to overcome these geotechnical design challenges are described in this paper. 1

INTRODUCTION

The Magtymguly Collector Riser platform (MCR-A) is a skirted gravity base foundation to be installed in the Magtymguly field located in the Caspian Sea, offshore Turkmenistan (Fig. 1). The substructure design for the platform has been undertaken by Arup. This paper provides a high level overview of a selection of the more challenging aspects of the foundation design, including: (i) Assessment of the bearing capacity of the annular foundation geometry under 3D loading. (ii) Quantifying the penetration resistance of skirts with a complex stiffener arrangement. (iii) Evaluation of the impact on foundation installation and in-place performance of spudcan induced ‘craters’ in the vicinity of the proposed platform. 1.1 Platform details and installation approach The MCR-A substructure comprises a square annular shaped, steel plated gravity base foundation, of outer dimension 54 m and inner dimension of 34 m. 5 m deep skirts extend below the base plate to form 12 compartments. Four legs made up of steel plated box girders beneath a four-chord steel tubular lattice are used to support the topsides (Fig. 2). This unusual arrangement allows the platform to be fabricated in Malaysia in components that can be shipped through the Volga-Don canal for final assembly in Turkmenistan. As limited offshore installation kit is available in the Caspian Sea, this configuration also provides the platform with sufficient buoyancy to © 2011 by Taylor & Francis Group, LLC

445

Figure 1. Site location.

float, and to be self-installing using controlled ballasting. The skirts are then penetrated into the seabed using a combination of self weight and under-base suction. Further details of the MCR-A platform are given in Pinna et al. (2009). 1.2

Ground conditions

Geotechnical site investigations were undertaken at the MCR-A location which included piezocone penetrometer testing (PCPT), seismic PCPT and sampling boreholes. The investigation was supported by offshore and onshore laboratory testing involving a suite of characterisation and advanced tests. Ground conditions at the site comprise firm to stiff over-consolidated clay to approximately 11 m below seabed, beyond which interbedded clay and sand layers were present to depth. The design monotonic

(2007). However, accurately assessing the MCR-A foundation capacity using these guidelines was challenging because: a) Whilst shape factors are given to account for rectangular or square shaped foundations, these cannot be readily applied to the annular geometry of the MCR-A foundation. b) Procedures are mostly geared towards in-plane loading rather than the 3D and multi-directional loading that MCR-A is subjected to. c) Solutions are only available for a range of idealised soil profiles.

Figure 2. Sketch of the MCR-A platform.

Figure 3. Three dimensional foundation loading regime.

undrained simple shear strength (su-ss ) profile with depth (z) within the zone of influence of the foundation can be approximated by su-ss = 35 + 4 z (kPa), although a soft layer of roughly 0.5 m thickness was present at 2.5 m. Cyclic strength profiles were also derived to account for the cyclic nature of the foundation loading. 2

FOUNDATION CAPACITY ASSESSMENT

The MCR-A base will be subjected to significant environmental loading (see Fig. 3). A typical design load case due to self weight and a 100 year return period storm event comprises a vertical load of about 240 MN, a horizontal load of up to 40 MN and moments in excess of 1000 MNm. MCR-A is also subjected to high seismic loading, but this aspect is not addressed in this paper.

2.1 Analytical assessment of bearing capacity The ultimate bearing capacity of shallow offshore foundations are commonly assessed using methods based on classical bearing capacity theory provided in design codes such as DNV (1992), ISO (2003) & API © 2011 by Taylor & Francis Group, LLC

As a result of these factors, only a simplistic analytical assessment of the ultimate capacity could be undertaken. This was based on the DNV (1992) method, supplemented with procedures outlined in API (2007) for the determination of effective area for moment loading in two directions. The skirts were included in the assessment, which enabled the weak layer at 2.5 m to be avoided by transferring the foundation loads to more competent deeper soils. The resulting soil plug trapped in the compartments could also be mobilised to assist in overcoming the large overturning moment. The approach adopted was to use the overall foundation area to assess the vertical and lateral stresses applied to separate sections of the base, and check the stability of these in isolation. This approach was considered to be conservative, and doesn’t account for the kinematic constraint provided by the annular shape, which affords additional capacity compared to the individual elements acting in isolation (see for example Fisher & Cathie 2003, Gourvenec & Jensen 2009). After allowing for DNV (1992) partial load and material factors the analytical analyses indicated a minimum reserve capacity of about 5% for the worst load case. 2.2

Finite element analysis

Given the constraints of the analytical analysis, finite element (FE) analysis was used to justify the foundation capacity as it allowed the complex interaction between the annular foundation and non-linear soil to be modeled with greater confidence. FE analysis was undertaken using both 2D and 3D finite element models.Analyses for all design loading cases were performed in 3D using the LS-DYNA FE program (LSTC 2006) whilst the Oasys 2D SAFE FE program (Oasys 2006) was employed for parametric studies (Fig. 4). Details in 3D of the base pontoon, skirts and stiffeners were modeled to provide a realistic structural representation of the foundation and its stiffness (Fig. 4). Vertical loads and moments, depending on their origin, were applied as either a pressure over the foundation base, or on top of the pontoon at the platform leg areas. The total lateral load was distributed evenly over the foundation at seabed level. The 2D model considered a plane strain section through the centre of the foundation in an east-west/ north-south direction. Skirts are modeled as plates

446

Figure 5. SAFE and LS-DYNA results: a) LS-DYNA orthogonal load vertical displacement contours at load factor = 1.6 b) vertical displacement results with increasing load factor.

Figure 4. FE models: a) 3D LS-DYNA b) 2D SAFE.

with equivalent bending stiffness to the stiffened skirts at the centre of a compartment. The base was represented by a continuous beam which was assigned an equivalent stiffness using results of the 3D analyses. The vertical pressure due to the vertical load and overturning moment was applied as a linear distribution across the two pontoons by assuming a rigid annular base. The lateral loads were applied as a horizontal shear stress across the pontoon width assuming the lateral load uniformly distributed over the annular base. In both the 3D and 2D models, (undrained) stability analyses were carried out using a non-linear shear stress-strain model that incorporated the small strain shear modulus (Go ) measured in seismic cone tests and a variation with shear strain inferred from the cyclic analysis for the design storm. The stability runs were undertaken using factored shear stress-strain curves and factored loads.

Figure 6. Indicative details of the MCR-A skirt stiffeners.

in excess of 1.5 are required before excessive displacements occur. As a partial load factor of 1.3 had to be achieved to satisfy code requirements, the FE analysis for this case provides confidence to the finding of the analytical analysis that adequate foundation capacity exists.

2.3 Results of FE modeling An example of the stability results obtained using the 2D and 3D models for a preliminary orthogonal load case is given in Figure 5. The vertical displacements measured at points A and B with the application of increasing environmental load factor are plotted. Similar results are achieved between the 2D and 3D analysis. They show that the foundation on the side of Point B experiences larger vertical displacement and is approaching failure in bearing. Little vertical displacement is occurring at point A on the trailing side of the foundation, which instead experiences more lateral movement such that a sliding type failure is being approached. Figure 5b shows that load factors © 2011 by Taylor & Francis Group, LLC

3

SKIRT INSTALLATION ASSESSMENT

The estimated soil resistance to achieve skirt penetration dictates the maximum suction pressure that needs to be designed for. High installation pressures are expected at the MCR-A site, which proved to be the most onerous design case for the skirt panels inducing high in-plane and out-of-plane loading. This required both horizontal and vertical stiffeners that were arranged to minimise the penetration resistance whilst maintaining structural integrity (Fig. 6).

447

Figure 7. Schematic of penetration resistance with: a) case of full flow-round, b) case of no flow-round, c) soil plug heave.

3.1

Penetration resistance

Due to the complex stiffener arrangement, a key challenge faced was assessing the side friction and end bearing that may be attracted by each of the different skirt panel elements as they penetrate into the clay. Uncertainty was associated with the soil mechanisms that might be occurring during the skirt installation, particularly in relation to the manner in which the soil behaves after coming in contact with the first horizontal stiffener. Above this stiffener, the internal soil may remain self-supporting (Andersen & Jostad 2004), or alternatively, the soil may partially or completely flow around the stiffener, which has also been observed for stiff clay (House & Randolph 2001). These two possible scenarios of full flow-round and non flow-round are illustrated in Figure 7. Two design methods were used to predict the skirt penetration resistance. These were DNV (1992), which uses empirically based coefficients applied to measured cone resistance, and DNV (2005) which utilises the simple shear undrained shear strength. To account for the uncertainties involved, lower-bound, best estimate, and upper bound predictions were made. The resulting maximum suction pressures from these estimates ranged from approximately 50 kPa for a no flowround assumption, to approximately 250 kPa assuming full flow-round. 3.2

Figure 8. Measured vs. predicted penetration resistance calculated for the case of no soil flow-round.

Plug heave

A second challenge associated with the installation was quantifying the soil plug heave within the compartments due to the soil being displaced by the skirt and stiffeners (Fig. 7). Excessive heave could prevent the target penetration depth being reached by the soil surface prematurely coming into contact with the base plate, or causing pumping inlets to clog up. For installation without suction, typically half the skirt wall volume may heave inside the compartment, whilst all the volume displaced by the stiffeners remains inside. For installation using suction, it can be the case that the total structural volume occurs as heave inside the compartment (Andersen & Jostad 2004; Chen & Randolph 2007). The magnitude of heave may © 2011 by Taylor & Francis Group, LLC

be further increased if the soil does not flow around the horizontal stiffeners and instead is pushed upwards, leaving the gaps to be filled with water. For MCR-A, the plug heave assessment was based on a corner compartment as this was most critical having stiffeners on all four walls. The calculation was based on 100% of the displaced soil volume heaving inside the compartment, and allowed for partial water gaps occurring between the horizontal stiffeners. The component of heave resulting from the suction pressure was also estimated using 2D FE analysis. The final heave volume height was predicted to be approximately 300 mm based on a uniform displacement across the compartment. 3.3

Centrifuge testing

Given the challenges posed by the installation, a physical model testing programme was undertaken to examine the effects of the stiffener geometry on the penetration resistance and plug heave in overconsolidated clay. This involved a series of centrifuge installation experiments carried out at the University of Western Australia (UWA) as is described in Westgate et al. (2009). Three separate scale models were fabricated based on a corner compartment of MCR-A. To help investigate the impact of the stiffeners, the Type 1 model had no stiffeners, the Type 2 model had only horizontal stiffeners, whilst the Type 3 model contained the complete arrangement of both horizontal and vertical stiffeners. The results of the penetration resistance measured during the installation of each model type are given in Figure 8. Included in this figure is the predicted resistance calculated based on the assumption that end bearing occurs only on the first (lowermost) horizontal stiffener and does not flow inside the spacing of the stiffeners. It can be seen there is good agreement when this assumption is adopted. This provided evidence that a zero flow-round condition occurs, an assumption which was also supported by visual observations. The plug heave profile was also measured after the installation of each of the model types. As was expected, the magnitude of heave increased with

448

Figure 9. LS-DYNA vertical displacement from spudcan.

increasing model structural volume. For Type 3, representative of the MCR-A, the heave at prototype scale was less than the tolerable value of 500 mm, providing confidence that the degree of plug heave would not pose significant risk to the MCR-A installation.

These findings were in-line with those of Hossain et al. (2004).

4

4.2

Figure 10. Comparison of strengths inferred from T-bar tests.

IMPACT OF SPUDCAN CRATERS

Subsequent to the design of the MCR-A substructure, a jack-up rig was unexpectedly installed for a period in the direct vicinity of the proposed location of the platform (see Fig. 11a). The jack-up spudcans were 18.2 m in diameter and are reported to have penetrated to a depth of approximately 4 m. The disturbed ground conditions remaining after spudcan removal will be to the detriment of the foundation installation and in-place performance. As MCR-A must be linked to adjacent infrastructure, limited scope exists to move the platform location to avoid the disturbance. To support initial planning to resolve this problem, a preliminary investigation was undertaken to quantify the impact of the spudcan craters on the foundation design. In lieu of actual field data after removal of the spudcans, it was necessary to estimate the resulting soil conditions in regard to the extent and nature of the disturbed zone of soil beneath the crater and of the seabed heave. A final assessment is to be made when the results of a future site investigation are available. 4.1 Soil deformation due to the spudcan Soil displacement mechanisms during spudcan penetration in overconsolidated clay were studied experimentally by Hossain et al. (2004). A significant width of soil was observed to move downward with the spudcan whilst soil at the spudcan edge moved laterally and then upward creating surface heave. To enable the actual spudcan shape and soil conditions at the MCR-A site to be considered, 3D FE analysis of spudcan penetration was undertaken using LS-DYNA. The results of the zone of disturbance and heave profiles are shown in Figure 9 for a depth d to diameter D penetration ratio of d/D = 0.2. The zone of soil disturbance was found to have maximum width and depth dimensions of 1.1D (20 m) and 1.2D (22 m) respectively, and a heave profile of maximum width and height of 1.5d (5 m) and 0.45d (1.5 m) respectively. © 2011 by Taylor & Francis Group, LLC

Spudcan impact on soil strength

Experimental studies (e.g. Leung et al. 2007) have shown that spudcan installation and removal can lead to a (short term) reduction by a factor of 2 in the undrained shear strength of normally consolidated clays situated beneath the spud can footprint. To provide an indication of the likely impact of soil strength in higher OCR clays at MCR-A, a centrifuge testing programme was undertaken at UWA. This involved performing 2 T-bar penetration tests (TB1 and TB2) outside the installation area of a square foundation penetrated to a depth d/B = 0.2, and 1 T-bar (TB3) at the centre of the crater after removal of the spudcan. Project constraints meant that the sample clay strength was lower than that of the in-situ material, and that the foundation was not an exact replica of the prototype spudcan. The undrained strengths measured (derived using the T-bar factors given in Lehane et al. (2008) and corrected for shallow penetration effects) are compared in Figure 10 along with estimations using the standard relationship between vertical effective stress (σv ) and OCR. Lower strength is seen below the crater, arising partly because of the lower σv values in the clay and partly because of a permanent reduction in strength due to the soil disturbance. The latter component leads to a strength that is 88% of the undisturbed strength, whilst allowing for the reduced σv results in an overall strength in the clay beneath the craters that is approximately 60% of the undisturbed strength. 4.3

Spudcan impact on foundation design

The preliminary assessment of foundation stability accounting for the presence of spudcans was undertaken using similar soil models and procedures as described in Section 2. The spudcan craters were included in both models as is shown in Figure 11. The soil strengths were adjusted in the estimated zone of disturbance based on the interpretation previously

449

5

CONCLUSIONS

The geotechnical design of the MCR-A skirted gravity based foundation involved a number of challenges, several key ones of which have been presented in this paper. In describing the approaches taken to address them, this paper has highlighted how various methods of assessment, such as finite element analysis and physical modeling, can be used to overcome geotechnical design uncertainties. ACKNOWLEDGEMENTS The authors would like to acknowledge colleagues at Arup who assisted with work that has been discussed. Particular mention is made of Anton Pillai who ran the LS-DYNA analyses, and Zack Westgate (formerly of Arup) who among other things undertook skirt penetration assessment.

Figure 11. FE models with craters a) LS-DYNA b) SAFE.

REFERENCES

Figure 12. SAFE and LS-DYNA results with spudcan craters.

described. The 2D model is conservative since the craters are modeled as infinite parallel trenches. An example of the stability results obtained using the 2D and 3D models for an orthogonal load case is given in Figure 12. The vertical displacement due to increasing load factor is provided for the cases of i) no spudcan present, ii) spudcan crater included, and iii) spudcan crater included with reduced soil strength. It is seen that as these cases are imposed, increasing displacement is experienced for a given load factor. For the load case considered, the results qualitatively indicate that a 30% reduction in capacity could be expected based on the 2D SAFE analysis, or a 10% reduction from 3D LS-DYNA analysis. The estimated installation pressures were also reviewed given crater intersection of the foundation footprint, which may inhibit generation of the required suction in the compromised compartments. The heave mounds surrounding the craters have to be compressed by the base plate to penetrate the skirts, further increasing the penetration resistance. These factors meant that the maximum required suction pressure could increase by as much as 50% and complicate achieving a level foundation installation. A detailed assessment of these problems is to be undertaken following the proposed site investigation. © 2011 by Taylor & Francis Group, LLC

Andersen, K.H. & Jostad, H. P. 2004. Shear strength along the inside of suction anchor skirt wall in clay. Proc. Offshore Technology Conf., Houston, Texas, OTC 16844. API 2007. RP2A Recommended practice for planning, designing and construction of fixed offshore platforms. Chen, W. & Randolph, M. F. 2007. External radial stress changes and axial capacity for suction caissons in soft clay, Geotechnique. 57( 6), pp 499–511. DNV 1992. Classification notes No. 30.4, Foundations. DNV 2005. Recommended practice DNV-RP-E303: Geotechnical design and installation of suction anchors in clay. Fisher, R. & Cathie, D. 2003. Optimisation of gravity based design for subsea applications. Proc. Int. Conf. on Foundations, ICOF, Dundee, Scotland. Gourvenec, S. & Jensen, K. 2009. Effect of embedment and spacing of conjoined foundation systems on undrained limit states under general loading. Int. J. Geomech., 9(6) 267–279. Hossain, M.S., Randolph, M. F. & Hu, Y. 2004. Bearing capacity of spudcan foundation on uniform clay during deep penetration. Proc. 23rd Int. Conf. on Offshore Mechanics and Arctic Engineering, Vancouver, Canada. House, A. R. & Randolph, M. F. 2001. Installation and pullout capacity of stiffened suction caissons in cohesive sediments. Proc. 11th Int. Offshore and Polar Eng. Conf., Stavanger, Norway, pp 574–580. ISO 2000. Offshore structures: Part 4: Geotech., ISO 19900. Lehane, B. M., O’Loughlin, C.D., Gaudin, C. & Randolph, M. F. 2008. Rate effects on penetrometer resistance in kaolin, Geotechnique. 59(1), pp 499–511. LSTC 2006. LS-DYNA Theory Manual, Livermore. Leung, C.F., Gan, C.T. & Chow, Y. K. 2007. Shear strength changes within jack-up spudcan footprint. Proc. 7th Int. Conf. on Offshore & Polar Engineering, Lisbon, Portugal. Oasys. 2006. Safe Users Manual, Ove Arup, London, UK. Pinna, R., McRobbie, I., Altraide, B. & Razak, S. 2009. The first gravity based substructure (GBS) for the Caspian Sea. Proc. 4th Offshore Asia Conference, Bangkok, Thailand. Westgate, Z.J., Tapper, L., Lehane, B.M. & Gaudin, C. 2009. Modeling the installation of stiffened caissons in overconsolidated clay Proc. 28th Int. Conf. on Ocean, Offshore & Arctic Engineering, Honolulu, USA.

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Constructing breakwater with prefabricated caissons on soft clay S. Yan & X. Feng School of Civil Engineering, Tianjin University, Tianjin, China

J. Chu School of Civil and Environmental Engineering, Nanyang Technological University, Singapore

ABSTRACT: A case history of constructing offshore breakwater on soft clay is presented. The breakwater was constructed near the Shanghai Port, China, for deepening of navigation channel along the Yangtze Estuary. The breakwater elements were designed as gravity retaining structures using prefabricated, semi-circular shaped concrete caissons. Some sections of the breakwater were installed on a thick layer of soft soils. During the construction, the caissons in one section failed under a heavy storm. The causes of failure were investigated by running dynamic triaxial tests on undisturbed soil samples taken from the construction site. It was found that the dike failure was induced by the strength weakening of the soft soil layer below the foundation. The design of the guide dike and the soil improvement works are described in this paper. Surcharge preloading and prefabricated vertical drains was adopted to improve the soft soils below the caisson. The soil improvement measure was proven to be effective in maintaining the stability of the breakwater against subsequent heavy storms.

1

INTRODUCTION

piece and thus make the subsequent repair or rehabilitation works relatively easier. (4) The construction process is relatively simple and speedy, as no on-site concrete casting is required.

Guide dike, literally, is the guide for navigation. And it is also constructed to reduce the effects of wave and prevent future sedimentation within the channels. As part of the Yangtze Estuary Development Project, a breakwater was constructed 40 km away from a busy port in Shanghai, China. The guide dike was approximately 50 km. The water depth ranged from 7.0 to 8.5 m. The design wave height was 3.32 to 5.90 m with a return period of 25 years. The breakwater was designed as gravity retaining structures using prefabricated, semi-circular shaped concrete caissons to resist the rough waves (Yan, 2005). The prefabricated, semi-circular shaped concrete caissons for guide dike construction were used in the Channel Deepening Project for the Yangtze Estuary, which has the following advantages: (1) All the wave and water pressures acting on the semi-circular shaped surface pass through the centre of the circle. Thus, the overturning moment becomes very small and the vertical pressure distribution at the base of the caisson becomes more uniform (Jia, 1999). (2) Study shows that the wave force acting on a semicircular shaped guide dike is smaller than that on a vertical guide dike (Xie, 1999). The internal stress induced by the load applied is relatively small for an arch structure and thus the costs for the construction of the breakwater can be reduced. (3) The guide dike is made of prefabricated segments which can be dismantled or replaced piece by © 2011 by Taylor & Francis Group, LLC

However, as the technique is relatively new, there is not sufficient experience on how to treat the foundation soil properly and yet economically when the caissons are to be placed on soft seabed soils. On one hand, these caissons have to be tall and heavy enough to match the water depth and provide stability against the wave forces. On the other hand, when the caissons are too heavy, they cause settlement, bearing capacity, or stability problems (Thiers and Seed, 1969;Yasuhara, 1988). This is particularly the case when the foundation soil is very soft. Treating seabed soft soils offshore in relatively deep water over a large area is difficult and costly. Therefore, it became a challenge on how to construct these large size gravity structures on soft soil in a cost-effective way. Based on the theory analysis and test research, soil improvement work was put forward on the soft soil below the caisson dike to prevent the soft soil from heavily weakening. These measures were proven to be effective in maintaining the stability of the guide dikes against subsequent heavy storms. 2

SOIL CONDITIONS

The typical soil profile of the soil below the guide dike is shown in Figure 1. It consisted of 1.5 m to 3.5 m thick silty sand followed by 2 m to 4 m thick muddy clay and approximately 30 m thick soft clay

451

Figure 2. Cross-section of the prefabricated, semi-circular shaped concrete caisson.

Figure 1. Soil profile beneath seabed soil along the navigation channel. Table 1.

Soil stratum Silty sand Muddy clay Soft clay

Soil properties. Water content %

Unit weight kN/m3

Void ratio –

Liquid limit %

Plastic limit %

29.3

19.0

0.827

–

–

57.5

16.4

1.672

45.2

23.6

51.5

16.8

1.470

45.8

23.8

Figure 3. Failed breakwater after a heavy storm.

underlying the muddy clay. The basic properties of the soils are given in Table 1. The silty sand was very fine with a mean grain size of 0.15 mm. Although the soil below the guide dike in this zone was weak, the initial design was not to treat the seabed soft soil, but to use an adequate strong rubble mount to support the caissons. This approach was based on the following four considerations: 1) It would be very expansive to treat the seabed soil offshore in relatively deep water; 2) The guide dike could tolerate relatively large settlement, as it would be used mainly as a divider; 3) As the guide dike was a strip load, the surcharge load would only be distributed to a certain depth. Under the surcharge load of the rubble mount, the geotechnical properties of the muddy clay near the top few meters would be improved with time, and thus the stability of the guide dike would be improved with time; 4) Similar design had been adopted for other projects with similar site conditions and no failure had occurred before. Such as the breakwater constructed in Miyazaki Harbor, Japan (Sasajima, et al., 1994; Tanimoto and Takahashi, 1994).

The base of each concrete segment was 17 m wide and 19.94 m long. The hollow caisson would be filled with sand after installation through a 600 mm diameter hole on top of the caisson. In order to enhance the stability of caisson and prevent the foundation soil from scouring, a layer of geotextile sheet and sand filled tube composite was placed onto the seabed to enhance the stability of the rubble mount. The sand tubes were 300 mm in diameter and spaced 500 mm apart at the edge and 1000 mm in the centre, as shown in Figure 2. Other detail of the sand tube will be described at a later section. A sand cushion of 700 mm thick was placed on top of the geotextile and sand tube composite layer before another getextile and sand tube composite layer was placed on top of the sand cushion. This was topped by crushed stones of 1∼100 kg in the centre and 200∼400 kg at the edge, which were used to form a rubble mount platform at an elevation of −3.60 m, as shown in Figure 2. The caisson was placed on top of this rubble mount. Berms were also placed on two sides of the caisson to enhance the stability of the guide dike. The berms were 1.7 m thick and made of 400 ∼ 600 kg crushed stones. The hollow caissons were then filled with sand to an elevation of −0.8 m.

4 4.1

3

ORIGINAL DESIGN OF THE GUIDE DIKE

The design of the guide dike is schematically shown in Figure 2. The caisson used was prefabricated, semicircular shaped concrete hollow segment, as shown in Figure 2. The radius of the semicircle was 5.7 m. © 2011 by Taylor & Francis Group, LLC

FAILURE CASE Failure description

The construction for the guide dike was completed in Oct 2002.A heavy storm took place in December 2002. During the storm, 16 segments of the caissons failed. Large settlements occurred suddenly and some caissons also moved laterally. A picture of the guide dike after the storm is shown in Figure 3. The maximum

452

Figure 4. Settlement of #1 caisson and wave height versus time curves.

settlement was 5.78 m. The maximum lateral movement was more than 20 m. The settlement versus time curve measured for caisson #1 is shown in Figure 4. The wave height versus time curve is also plotted in Figure 4. It can be seen that a sudden settlement took place at the time when there was a surge in the wave height on 6 December 2002. The guide dike had failed under the wave action. To understand the causes of failure, dynamic triaxial tests on the muddy and soft clay were carried out. 4.2 Dynamic triaxial tests on the muddy clay and soft clay 4.2.1 General It was understood that the muddy clay below the caissons was in a most critical state as the surcharge loads was just applied for 2 months. Furthermore, it was identified that the following two factors had also contributed to the failure. The first is the additional surcharge load due to the dynamic action of the wave. The second is the softening of soft soil below caissons; that is, a reduction in the undrained shear strength of the clay, under the wave loads. The possibilities of liquefaction of the top silty sand layer and scour at the base of the caissons were ruled out as the causes after the investigation. Based on elasticity theory, the vertical effective stress distributions due to both static and dynamic conditions are calculated as shown in Figure 5. The ratio of surcharge load increment due to the wave action, σv , to the surcharge load under static condition, σv , β = σv /σv , is also calculated and plotted in Figure 5. It can be seen that under the wave action, the vertical overburden stress could increase by 15% due to the wave load. However, this effect becomes much less significant with increasing in depth. 4.2.2 Load path and test equipment After the failure, some undisturbed soil samples were taken from the muddy clay layer and cyclic triaxial tests were conducted (Andersen, et al., 1988; Andersen, and Lauritzsen, 1988). The loading path for the soil samples is shown in Figure 6. Point A is the initial stress of the soil element. Path AB represents the static loading caused by the rubble mount and the caisson © 2011 by Taylor & Francis Group, LLC

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Figure 5. Increase in surcharge load due to wave loads.

Figure 6. Load path before soil improvement.

and path BC denotes the action of the cyclic loading caused by the wave force. Where, τf -shear stress at failure; K0 -consolidation coefficient. τ-shear stress; σc -confining stress.

Table 2.

Properties of soil sample SY3-4.

Soil type

Water content %

Unit weight kN/m3

Sample diameter cm

Sample height cm

Muddy clay

53

17.3

5.0

10.0

Table 3. Test conditions and results of soil sample SY3-4. Cell Static Dynamic Initial Residual Reduction pressure load load strength strength Ratio kPa kPa kPa kPa kPa – 30

20

8

15

10.8

0.72

The tests were performed with the HX-100 dynamic triaxial test equipment. According to the loading path, the sample was first consolidated with the in place initial stresses by self weight. Then the static load was applied in undrained condition. When the displacement of the soil sample was stable, the application of the cyclic load was applied also in undrained condition. The combinations of the applied static load and dynamic load were determined according to the FE analysis results before and after soil improvement (see the later sections). 4.2.3 Test results The triaxial dynamic test was performed on an undisturbed soil sample SY3-4 at depth 3.0–3.5 m. The details of the soil sample, typical test conditions and results are listed in Table 2 and Table 3. The dynamic stress was applied and repeated for 1000 times (with a period of 6 seconds) on the soil sample. The pore pressure (P.P) development is recorded and shown in Figure 7a; the accumulated strain developing with the number of cycles in Figure 7b. The strength of the soil samples were tested before and after the dynamic tests and shown in Figure 8. A total of 25 soil samples were tested in the same way. The strength reduction coefficients are summarized and given in Figure 9. In this figure, both the additional static stress and dynamic are normalized by the confining pressure. Therefore, the strength reduction ratio can be determined with different combinations of σc , σj and σd . 5

SOIL IMPROVEMENT

Figure 8. Shear test results before and after dynamic loading.

Figure 9. The strength reduction ratio.

To ensure the stability of the guide dike against future storms, a soil improvement scheme was adopted to improve the soft soil layer below the caissons, which was 7.0 m to 8.5 m below the sea level. Several soil improvement methods were considered. The method to accelerate the consolidation process of muddy clay using prefabricated vertical drains (PVDs) was considered the most economical option. This method was © 2011 by Taylor & Francis Group, LLC

Figure 7. Typical dynamic triaxial test results.

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also relatively easy to implement. The influence zone of the strip load was estimated to be 7.0 m below the seabed. Therefore, it was only necessary to install the PVDs to a depth of 10 m deep. The weight of the rubble mound was considered sufficient in providing surcharge. The PVDs were installed offshore from a specially designed PVD installation barge. There were

Figure 10. The revised design profile.

Figure 11. The constructed semi-circular concrete breakwater.

12 drain installation rigs on each barge with preset spacing of 1 m in square grid. The position of each drain was located using GPS. Each barge could carry 12 drain installation rigs and could install an average of 1185 drains per day. The procedure for improving the soft soil before the placement of the rubble mount and the caissons was as follows: (1) PVDs were installed from the PVD installation barge at a spacing of 1.0 m in square grid to a depth of 10 m below the seabed; (2) A geotextile and precast concrete block composite was used to cover the seabed around the toe of rubble mount that face the open sea for the prevention of scour. The concrete blocks were 400 mm × 400 mm in square and 160 mm thick. They were attached to the geotextile sheet to form a composite. For the rest, the geotextile and sand tube composite was installed on the seabed. Sand was filled into the 300 mm diameter geotextile tubes from a barge on site before the geotextile and sand tube composite was placed. The tubes were spaced at every 500 mm at the toe and 1000 mm at the centre of the rubble mount; (3) The 700 mm sand cushion was placed before the second layer of geotextile and sand tube composite was laid on top; (4) The crushed stones were laid on to the second geotextile and sand tube composite from a barge which also acted as a surcharge to consolidate the soil below; (5) The caisson segments were only placed after an average degree of consolidation of 80% was achieved, which took about 90 days after the placement of the rubble mound. The adopted soil improvement method has the following advantages: (1) The soft soil can be strengthened to increase the ability of resisting dynamic loads; (2) After consolidation, the weight of the foundation consisting of crushed stones becomes part of the consolidation stress, instead of part of the structure load. This means that after soil improvement, the soil becomes stronger and the static load becomes smaller. As a result, the consolidation stress increases and the vertical subsidiary stress caused by the upper structure decreases. Referring to Figure 9, the strength reduction ratio will increase obviously, so the constructed dike will become much safer. Using the soil improve method, the entire caissons of the navigation guide dike were completed in Dec © 2011 by Taylor & Francis Group, LLC

2003. The guide dike had experienced several strong storms caused by typhoons. Some of them were even stronger than the one that caused the failure. In spite of this, the guide dike was stable and there was no additional settlement caused by the storms. The total settlements were within the expected range. Therefore, the use of PVDs has proven to be a successful method for this project. The guide dike after construction is shown in Figure 11.

6

CONCLUSIONS

A case study of using prefabricated, semi-circular shaped concrete caissons to construct offshore breakwater on soft soil was presented in this paper. A failure case was also investigated by running dynamic triaxial tests on undisturbed soil samples taken from the construction site. It was found that the dike failure was induced by the strength weakening of the soft soil layer below the foundation. It is learned through dynamic triaxial tests that, under a heavy storm, the strength of the soft soil can be seriously reduced. This factor should be taken into consideration in the design. Surcharge preloading plus PVDs was used to improve the soft seabed soil before the installation of the guide dike. With the use of PVDs, the soft soil was consolidated much faster and gained sufficient strength quickly to maintain the stability of the caissons. This scheme has proven to be effective as the dike has experienced several storms since construction without any problems. REFERENCES Andersen, K. H., Kleven, A. & Heien, D. 1988., “Cyclic soil data for design of gravity structures”, Journal of Geotechnical Engineering, Vol: 114(5), pp.517–539. Andersen, K. H. & Lauritzsen, R., 1988, “Bering capacity for foundations with cyclic loads”, Journal of Geotechnical Engineering, Vol:114(5), pp.540–555. Jia, D.H., 1999, “Study on the interaction of water waves with semi-circular shaped guide dike”, China Ocean Engineering, Vol:13(1), pp.73–80.

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Sasajima, H., Koizuka, T., Sasayama, H., Niidome,Y. & Fujimoto, T., 1994, “Field demonstration test on semi-circular shaped guide dike” Proceedings of the International Conference on Hydro-Technical engineering for Port and Harbor Construction, Yokosuka, Japan, (1), pp.593–615. Tanimoto, K., & Takahashi, S, 1994, “Japanese experiences on composite breakwater”. Proceedings of international workshop on wave barriers in Deep Waters, Yokosuka, Japan, pp.1–22. Thiers, G. R. & Seed, H. B., 1969, “Strength and stress-strain characteristics of clays subjected to seismic loading conditions”, Vibration Effects of Earthquakes on Soils and


Foundations, Selig, E.T. ed., ASTM Special publication STP450. Xie, S. L., 1999, “Wave forces on submerged semi-circular shaped guide dike and similar structures”, China Ocean Engineering, Vol:13(1), pp.63–72. Yan, S.W., 2005, “Dynamic cyclic triaxial tests on undisturbed soft clay samples taken along the Yangtze Estuary waterway”, Project Report, Geotechnical Research Institute, Tianjin University, 2005 (in Chinese). Yasuhara, K., 1988, “Cyclic strength and deformation of normally consolidated clay”, Soils and Foundations Vol:22(3), pp.373–381.

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6 Piled foundations



Simplified analysis of laterally loaded pile groups F.M. Abdrabbo & K.E. Gaaver Structural Engineering Department, Faculty of Engineering, Alexandria University, Egypt

ABSTRACT: The response of laterally loaded pile groups is a complicated soil-structure interaction problem. Although fairly reliable methods are developed to predicate the lateral response of single piles, the lateral response of pile groups has attracted less attention due to the required high cost and complication implication. This study presents a simplified method to analyze laterally loaded pile groups. The proposed method implements p-multiplier factors in combination with the horizontal modulus of subgrade reaction. Shadowing effects in closely spaced piles in a group were taken into consideration. It is proven that laterally loaded piles in sand can be analyzed within the working load range assuming a linear relationship between lateral load and lateral displacement. The proposed method estimates the distributions of lateral loads among individual piles in a pile group and predicts the safe design lateral load of a pile group. The benefit of the proposed method is its simplicity for the preliminary design stage.

1

INTRODUCTION

The lateral response of pile foundations is critically important in the design of structures that may be subjected to lateral loads. It is worth noting that, lateral loads are in the order of 10%–15% of the vertical loads in the case of onshore structures, while this value may exceed 30% in case of offshore structures (Rao et al. 1998). The response of a laterally loaded pile is a complicated soil-structure interaction problem, because the pile deflection depends on the soil reaction and the soil reaction in turn depends on the pile deflection. Fairly reliable methods have been developed for predicting the lateral response of single piles, since the pioneer works of Matlock and Reese (1961), and Broms (1964). Frechette et al. (2002) reviewed the design methods for laterally loaded groups of drilled shafts and compared between methods employing a group reduction factor and a p-multiplier. Kumar and Lalvani (2004) analyzed the nonlinear load-deflection behavior of laterally loaded piles using p-y relationships. Full scale and centrifuge model tests on pile groups have been conducted by Brown et al. (1988), McVay et al. (1998), and Rollins et al. (2005). Laterally loaded pile groups may be analyzed using the elastic continuum approach (Poulos and Davis 1980), and the group equivalent pile procedure (Ooi et al. 2004). The p-y relationships, initially developed by Matlock (1970), have been used to model the pilesoil interaction, Reese et al. (1974). As a result of the interaction between piles in a group, the p-y relationship of single pile was modified to be implemented in pile group analysis. The modifications can be carried out by introducing p-multiplier, Ooi et al. (2004), and Rollins et al. (2005). The p-multiplier concept is an effective procedure for implementing in the pile group © 2011 by Taylor & Francis Group, LLC

analysis; nevertheless unique values of p-multiplier for a pile group are not standardized. It is worth noting that the p-y relationship is not a soil property, but rather pile-soil property, Ashour and Norris (2003). In recent years, several simplified approaches for the analysis of laterally loaded single piles or pile groups have been developed that can be used with little computational effort, Liyanapathirana & Poulos (2005), and Castelli & Maugeri (2009). This paper presents a simplified method, for analyzing laterally-loaded pile groups, using p-multipliers in combination with Winkler’s model. 2

GEOTECHNICAL DATA OF THE SITE

Before discussion of the proposed method, obtained geotechnical data, where laterally loading tests on vertical single piles were conducted, are presented. The proposed method was implemented to analyze pile groups constructed at this site. The site is located at the Northeast of Nile River Delta, Demiatta free zone district, Egypt. The soil profile at the site consists of a top layer of medium dense sand. N60% varies from 40 near ground surface to 20 at a depth of 10 m. Fine silty sand layer was encountered at a depth 10 m, and extended to a depth of 15 m below ground surface. This layer is underlain by a thick layer of soft to medium normally consolidated clay, which is extending up to a depth of 36 m. At a depth of 36 m, a very dense sand bed was encountered and explored to a depth of 60 m. The top sand layer has natural unit weight of 19 kN/m3 , angle of internal friction 37◦ , and relative density of 70%. These values were interpreted from the average value of SPT (N60% ) throughout the sand layer. For soft clay, the natural unit weight varies

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between 13.1 and 16.7 kN/m3 , and the undrained shear strength varies from 6 to 20 kN/m2 . The values of unit weight and undrained shear strength increased with depth throughout the clay layer. For medium clay, the natural unit weight varies from 17.4 to 18.7 kN/m3 , and the undrained shear strength varies from 15 to 45 kN/m2 . For sand bed, the peak angle of shearing resistance is 45◦ , while the residual value is 36◦ . The shear strength of clay and sand bed was measured using the triaxial test apparatus. The ground water table is at 1.0 m below ground surface. Bored piles, of 600 mm diameter, were constructed to be seated at a depth 40 m below the existing ground surface. The pile is reinforced by nine bars of 18 mm diameter of steel grade 36/52 and the reinforcement is extending to 16 m below the ground surface. The piles are attached to a pile cap which is resting on ground surface. As the pile is considered a flexible pile, the safe design lateral load of the pile depends on structural capacity of the pile cross section and the allowable lateral deflection at pile head. Based on these design criteria, the safe design lateral load of single pile is 80 kN, dominated by structural capacity of the pile cross section. 3 THE PROPOSED METHOD The aim of the proposed method is to estimate the distributions of lateral load acting on a pile group among the piles in the group. The piles in the group are considered flexible piles. More likely, flexible piles in a group are embedded in a stratified soil, and hence the lateral load may be resisted by soil lateral stresses developed along the top portion of the pile, which is called the effective length (Lef ). One method of assessing the value of (Lef ) is by modeling a single pile as a beam in a soil represents by an elastic uncoupled spring modulus. Lef is assessed as the depth where the lateral deflection of the pile is effectively zero. But the effective depth of single pile differs from the effective depth of a pile in a group, due to pile-soil interaction. The effective length (Lef ) of a pile in a group was calculated by re-analyzing single pile but with softer springs. These spring moduli were obtained by multiplying the spring modulus of the single pile by p-multiplier values. In this analysis, the horizontal subgrade reaction (Kx(s) ) at a depth (Z) below the ground surface within the top sand layer is expressed as; Kx(s) = ηh .Z, where ηh is the modulus of horizontal subgrade reaction at top sand layer. (Kx(s) ) is expressed in units of force per unit area, while (ηh ) is expressed in units of force per unit volume. A constant value of horizontal subgrade reaction along the pile through the clay layer (Kx(c) ) is considered. It is important to note that the modulus of horizontal subgrade reaction is not a unique soil property, but depends on pile characteristics and the lateral displacement of the pile. Using the aforementioned soil profile and soil properties at Demiatta free zone district, the effective depth of single pile and for a pile within a group was founded to be less than the depth of the top sand layer and equal to about 16 times


the pile diameter. In this study, ηh was considered for single pile equal to 16.346 MN/m3 , for medium sand. The relative stiffness factor (T) and the maximum value of depth coefficient (Zmax ) were calculated as; T = (E.I/ηh )0.2 , and Zmax = (Lef /T), where E.I is the flexural rigidity of the pile as un-cracked section. The maximum value of depth coefficient (Zmax ) was found to be 5, which means that the pile is flexible pile. The dimensionless relationships, developed by Reese and Matlock (1956), were used to determine the distribution of pile displacements, bending moments, shearing forces, soil resistances, and slope deflections along the effective length of a single pile due to the safe design lateral load of 80 kN applied at the pile head assuming fixed head piles, without any free length above ground surface. For the single pile, ηh was implemented directly, while for a pile within a group ηh was reduced due to shadowing effects. The shadowing effect depends upon the location of pile row within the group and the location of the pile within the row. McVay et al. (1998) concluded that in the same pile row, the middle pile develops slightly less lateral resistance than the side piles because it is subjected to more substantial shadow effects. However, the authors showed that the difference is not significant and no significant error is developed by assuming that all piles in the same row carry the same lateral load. Consequently, the multiplier factor (p) for all piles within a row was assumed to be the same value. To consider the effects of pile-soil-pile interaction in a group, a pile within a group was analyzed as a single pile and the distribution of pile deflections, bending moments, shearing forces, lateral soil resistances, and slope deflections along the pile were assessed for different values of ηhp where the reduced value of horizontal subgrade reaction (ηhp ) was obtained from; ηhp = p.ηh . The values of p-multiplier factors were obtained from McVay et al. (1998), and Ooi et al. (2004). As a result, a data base containing pile deflections, bending moments, shearing forces, lateral soil resistances, and slope deflections were formed for a single pile embedded in fictitious sand of different (ηhp ) values and subjected to different lateral loads. The data base was formed with the help of computer spreadsheets.

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4 ANALYSIS METHODOLOGY Once the data base was compiled, the analysis of a laterally loaded pile group can be carried out. In the first case study, a pile group configuration containing n-rows of piles and subjected to certain lateral load (PH ) at the ground surface is studied. The piles in the group are attached to rigid pile cap, that is to say the lateral displacements of all piles in the group at their heads are equal. The properties and reinforcement of the piles are mentioned in section (2). The unknowns in this case study are the load distribution among piles in the group and the lateral displacement of the group at ground surface. The piles are considered long flexible piles. The relationship between lateral applied

load (PH ) and lateral displacement at pile head (yG ) is assumed to be linear. The p-multiplier for each row was assessed from documented literature assuming the leading row has the bigger p-multiplier while the trailing row has the smaller p-multiplier. The p-multiplier depends on the number of rows in the pile group, and location of the row in the group. Entering an assumed value of lateral displacement (yG ) and the p-multiplier into the data base, the lateral load acting on a pile (Pi ) in each row corresponding to the assumed lateral displacement of the group (yG ) and the given specified p-multiplier was obtained. If the sum of pile lateral loads (Pi ) is equal to the applied lateral load (PH ) acting on the pile group, the solution is obtained and the process is terminated. But if the sum of pile lateral loads (Pi ) differs from the applied lateral load (PH ), the assumed pile group displacement needs to be altered and the procedure continues in a trial and error process until the equilibrium between the sum of pile loads and the applied total load achieved, that is to say (Pi ) = (PH ). The assumed lateral displacement at equilibrium condition is the lateral displacement of the pile group. Once the equilibrium condition is achieved, the distribution of bending moment, shear force, and soil pressure in each pile in the group can be obtained from the data base. In the second case study, the number of piles in a group is known and the safe design lateral load of the single pile is also known. It is required to determine the safe design lateral load of the pile group. The properties and reinforcement of the piles are mentioned in section (2). This case study represents a practical case in which the safe design lateral load of single pile is evaluated and verified by field loading tests. Usually the pile group configurations are assessed by knowing the vertical applied loads, vertical working load of single pile, and group efficiency. Once the pile groups are arranged, the capability of pile groups to sustain the lateral loads safely becomes essential. To tackle this problem, the value of p-multiplier for each row in the group was assessed. Returning back to the compiled data base, the pile head displacement (y1 ) of the piles in the leading row was obtained corresponding to horizontal subgrade reaction of (p1 .ηhp ) and lateral applied load equal to design lateral load of the single pile. At the same displacement (y1 ) and horizontal subgrade reaction of (p2 .ηhp ), the pile load for the second row is obtained. The procedure is repeated for all rows in the group. Then the design lateral load of a pile group is equal to the sum of individual pile loads within the pile group. The only drawback in the proposed method is that the load of piles in the leading row is assumed equal to the design lateral load of single pile but with a corresponding bigger displacement compared to the displacement of individual pile. The p-multiplier of leading row varies from 0.75 to 1.00, Rollins et al. (2005), and from 0.65 to 1.00, Ooi et al. (2004). Truly at the same lateral displacement of a pile group and single individual pile, the lateral load applied on the leading row in the group is smaller than single individual pile.


Figure 1. Pile arrangement, pile diameter = 0.60 m and length = 40 m.

5

NUMRICAL EXAMPLE

Consider a group of three piles, each of 600 mm diameter installed in one row. This arrangement represents 3 × n-pile groups, where n = 1, 2, 3, etc, figure (1). Single vertical pile was analyzed using curves, developed by Reese and Matlock (1956), under a lateral load of 80 kN. Fixed head pile was assumed, the resulted pile head displacement is 1.95 mm. It is required to determine the lateral load acting on each pile in the group under acting a lateral load of 240 kN. The analysis was started by assuming the p-multipliers as 0.8, 0.4 and 0.3 for leading pile, middle pile, and trailing pile respectively. These values were obtained from McVay et al. (1998). Dodagoudar et al. (2010) reported the values of p-multipliers published by their study, Rollins et al. (1998), Brown et al. (1987), Ilyas et al. (2004), Reese et al. (2006), and Mokwa and Duncan (2005). The reported data are for piles in all soil types. The p-multiplier for leading row varies from 0.60 to 0.93 with an average value of 0.79. The p-multiplier for second row varies from 0.40 to 0.78 with an average value of 0.58. For the third row, the p-multiplier varies from 0.40 to 0.63 with an average value of 0.46. For the fourth row, trailing row, the p-multiplier varies from 0.40 to 0.68 with an average value of 0.52. The reported values by Dodagoudar et al. (2010) excluded values published by McVay et al. (1998), which the present analysis was based. Also it is worth noting that the average value of p-multiplier reported by Dodagoudar et al. (2010) for the fourth row is bigger than the third row. From the compiled data base, the pile head lateral displacements under an applied lateral load of 80 kN and values of ηhp equal to 0.8ηh , 0.4ηh and 0.3ηh are 2.23, 3.40, and 3.98 mm for leading pile, middle pile, and trailing pile respectively. These lateral displacements violate the boundary conditions at the pile heads. Therefore by assuming the lateral deflection of the pile group at ground surface is 3.00 mm and each pile in the group exhibits this deflection, the lateral loads of leading pile, middle pile, and trailing pile shall be 107.62 kN, 70.58 kN, and 60.3 kN respectively. At this stage, the sum of the lateral loads of the piles in the group is 238.5 kN which is less than the applied lateral load of 240 kN. So the pile group displacement should be adjusted to be 3.019 mm to match with the applied load of 240 kN. The process was repeated and the corresponding loads acting on leading, middle, and trailing piles become 108.28 kN, 71.04 kN and 60.68 kN respectively. The sum of the pile loads in the group becomes equal to the applied lateral load. The corresponding lateral displacement

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of the group is 1.548 times the lateral displacement of the single pile. It is worth noting that 9-pile group arranged in square pattern exhibits the same lateral displacement under an acting lateral load of three times of 240 kN. The shadowing approach in which the piles in a row have no interaction effects on piles outside this row is contradicted with the elastic approach. In the elastic approach, the interaction factor between two piles depends upon the angle in plan between the centers of these two piles among other factors, Poulos and Davis (1980). To estimate the safe design load of (3 × 1) pile group, the lateral load of the leading pile was assumed equal to the safe design lateral load of the single pile, which is 80 kN. From the compiled data base and introducing p-multiplier of 0.80, the corresponding lateral displacement at the pile head of the leading pile is 2.23 mm. By enforcing the piles in the group to exhibit the same deflection, thus the lateral loads of the middle pile and the trailing pile become equal to 52.49 kN and 44.83 kN respectively. These values were obtained from the data base using p-multipliers for middle and trailing rows. The maximum bending moments induced in leading, middle, and trailing piles under lateral loads of 80, 52.49, and 44.83 kN respectively and corresponding to p-multipliers of 0.80, 0.40, and 0.30 were obtained from the compiled data base. The structural capacity, expressed as the bending capacity of the pile cross section, was calculated and compared with induced values. It was found that the pile cross section is capable to resist the induced moments safely. If the pile cross section is incapable to resist the induced bending moment, the process is repeated but with a small value of lateral load on the leading pile. In flexible piles, the dominant factor in assessing the safe load of single pile and pile group is the structural capacity of piles. Consequently the safe design lateral load of the pile group is 177.32 kN. Nine-pile group arranged in square pattern carries three times the achieved value of lateral load, at the same value of lateral displacement. The corresponding group reduction factor is 0.739, compared by 0.67 that was reported by Frechette et al. (2002). This analysis indicated that the leading pile carries 45.1% of the applied lateral load acting on the pile group, while the middle and trailing piles carry 29.6% and 25.3% respectively. The corresponding lateral displacement of the group is 1.143 times the lateral displacement of a single pile. McVay et al. (1998) conducted lateral tests on pile groups founded in sand in a centrifuge machine. Their results indicated that, for 3 × 3 pile group the percentage of lateral load carried by lead, second, and trail rows were 43.3%, 31.5%, and 25.2% respectively in case of dense sand. For loose sand, the corresponding values were 46.6%, 29.3%, and 24.1% respectively. A comparison between the results of pile load distribution obtained by simplified method and the measured values revealed that results of the proposed method are in good agreement with the experimental results. The only shortcoming of the proposed method is that the effect of spacing between piles in the group is not considered. However, from


Figure 2. Lateral load versus lateral displacement, test No. 1.

Figure 3. Lateral load versus displacement, tests No. 2 & 3.

a practical point of view, most of designers prefer to arrange the piles at minimum spacing in a group in order to minimize the size of the pile cap. Thus the proposed method is suitable to be implemented for pile groups having practical spacing of 2.5 to 3 times the pile diameter. In this situation it is important to note that the effect of spacing between piles in a group can be considered in the analysis if p-multiplier values for pile groups of different spacing to diameter ratios are developed. 6

JUSTIFICATION OF THE ASSUMPTIONS

The proposed method is based on linear relationship between lateral load and lateral displacement at pile head, which was confirmed by McVay et al. (1998), Yang and Liang (2006), Gaaver (2006), and field test results presented in figures (2) and (3). Figure (2) presents results of a test pile of 600 mm diameter and 40 m length installed at the site located at Northeast side of Nile River delta, having the same succession of soil strata as given before in section (2). Figure (2) illustrates a good agreement between theoretical p-y relationship and the experimental values up to the design lateral load of 80 kN. Two field pile loading tests were conducted on two individual piles at a site nearby Alexandria city, Egypt. The piles are of 500 mm diameter and 13 m depth below the ground surface. The lateral load was applied at the ground surface. Retrieved soil samples from boreholes indicated that the explored site consists of a top layer of sandy silty clay up to 9.5 m depth.The top layer overlies sand stone to 20 m depth. The top layer has a natural unit weight of 18 kN/m3 , undrained shear strength of 23 kN/m2 , and an angle of internal friction of 21◦ , measured using direct shear box apparatus. Figure (3) presents

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Figure 4b. Comparison between measured and computed bending moments of a single pile, Ph = 89 kN, and ηh = 53 MN/m3 .

Figure 4a. Comparison between measured and computed bending moments of a single pile, Ph = 24 kN, and ηh = 53 MN/m3 .

the achieved test results. The safe design lateral load of the pile is 60 kN and the test load 120 kN. Cleary test (2) demonstrates that the relationship is linear up to 100 kN. At the same time, test (3) shows that the relationship is linear, without any appreciable residual displacement. Therefore linear analysis of laterally loaded piles is considered as a good simulation of real behavior of piles under lateral loads within the working load range. It is worth noting that the tested piles were constructed by boring the soil and cast in situ concrete. Therefore, there is a complete contact between the formed pile and the surrounding soil especially near ground surface. The gap that may be formed near ground surface between the pile and the surrounding soil during pile construction as well as the nonlinearity of soil stiffness are the main causes of nonlinearity response of a laterally loaded pile at small values of lateral loads. The proposed method was also based on that the lateral resistance of a pile in a group is a function of row location belonging to that pile within the group, rather than location within a row, contrary to expectation based on the elastic theory. Rollins et al. (2005) and McVay et al. (1998) confirmed the above assumptions. Validation of the proposed method is presented in figures (4a) and (4b). The horizontal subgrade reaction was considered to be increased linearly with depth, from zero at ground surface to (ηh ) at depth 10 m below ground surface. The selected value of horizontal subgrade reaction was used along with the compiled data base to predict the distribution of bending moment along the single individual pile. A comparison between the obtained distribution of bending moment and the measured values by Rollins et al. (2005) indicated that the selected value of horizontal subgrade reaction overestimated the maximum bending moment induced in the single pile by up to 17%, while LPILE (Reese et al. 1997) and SWM (Ashour et al. 2002) methods underestimated the induced values by up to 20%. Cleary the horizontal subgrade reaction can be used for pile


Figure 5a. Bending moments versus p-multiplier.

Figure 5b. Lateral displacement at pile head versus p-multiplier.

group analysis. The induced bending moment in a pile within a group depends upon the location of the pile in the group. According to the proposed method, the distribution of the bending moment can be obtained by analyzing single individual pile using softening modulus of subgrade reaction that can be obtained by multiplying p-multiplier by (ηhp ). The effects of p-multiplier on induced bending moment and the lateral displacement at pile head are shown in figures (5a) and (5b). The applied lateral load at the pile head is 80 kN, while the horizontal subgrade reaction is increasing with depth from zero at ground surface to a value of 163 MN/m3 at depth 10 m below ground surface. The pile diameter is 600 mm. As p-multiplier decreased, the soil gets soft and consequently the pile head deflection and the bending moment increased.

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7

CONCLUSIONS

The effective length of a long flexible laterally loaded pile is equivalent to about 16 times the pile diameter, for pile embedded in sandy soil. Laterally loaded piles in sand can be analyzed within the working load range assuming a linear relationship between lateral load and lateral displacement at pile head. The induced maximum bending moments and lateral displacements at pile head of laterally loaded piles decreased linearly as the values of p-multiplier increased. The paper presents a simplified method for the analysis of pile groups subjected to lateral loads.The proposed method estimates the distributions of lateral loads among piles in a group and predicts the safe design lateral load of a pile group. REFERENCES Ashour, M. & Norris, G. 2003. Lateral loaded pile response in liquefiable soil, J. Geotechnical & Geoenv. Eng., 129 (5): 404–414. Ashour, M., Norris, G., & Pilling, P. 2002. Strain wedge model capability of analyzing behavior of lateral loaded isolated piles, drilled shafts, and pile groups, J. Bridge Eng., 7 (4): 245–254. Brown, A., Reese, C. & O’Neill, W. 1987. Cyclic lateral loading of a large-scale pile group, J. Geotechnical Eng., Vol. 113 (11): 1326–1343. Brown, A., Morrison, C., & Reese, C. 1988. Lateral load behavior of pile groups in sand, J. Geotechnical Eng., vol.114 (11): 1261–1276. Broms, B. 1964. The lateral Resistance of Piles in Cohesionless Soil, J. Soil Mech. Found. Div., Vol. 90 (SM3): 123–156. Dodagoudar, G., Boominathan, A. & Chandrasekaran, S. 2010. Group interaction effects on laterally loaded piles in clay, J. Geotechnical & Geoenv. Eng., Vol. 136 (4): 573–582. Castelli, F., & Maugeri, M. 2009. Simplified approach for the seismic response of a pile foundations, J. Geotechnical & Geoenv. Eng., Vol. 135 (10): 1440–1451. Frechette, D., Walsh, K. & Houston, W. 2002. Review of design methods and parameters for laterally loaded groups of drilled shafts, Deep foundations 2002:1261–1274. Gaaver, K. 2006. Behavior of laterally loaded piles in cohesionless soils, The Tenth East Asia-Pacific Conference on Structural Engineering and Construction.


Ilyas, T., Leung, F., Chow, K., & Budi, S. 2004. Centrifuge model study of laterally loaded pile groups in clay, J. Geotechnical & Geoenv. Eng., Vol. 130 (3): 274–283. Kumar, S. & Lalvani, L. 2004. Lateral load-deflection response of drilled shafts in sand, International Engineering journal, Vol. 84: 282–286. Liyanapathirana, D. S., & Poulos, H. G. 2005. Pseudostatic approach for seismic analysis of piles in liquefying soil, J. Geotechnical & Geoenv. Eng., Vol. 131 (12): 1480–1487. Matlock, H. & Reese, L. 1961. Foundation Analysis of Offshore Pile Supported Structures, 5 Int. Conf. on Soil Mech. and Found. Eng., Paris (2): 91–97. Matlock, H. 1970. Correlation for design of laterally-loaded piles on soft clay, Proc., 2nd Annual offshore Technology Conf., Vol. 1: 557–594. McVay, M., Zhang, L., Molnit, T. & Lai, P. 1998. Centrifuge testing of large laterally loaded pile groups in sand, J. Geotechnical & Geoenv. Eng., Vol. 124 (10): 1016–1026. Mokwa, L. & Duncan, M. 2005. Discussion of ‘Centrifuge model study of laterally loaded pile groups in clay’ by Ilyas, T., Leung, F., Chow, K., & Budi, S., J. Geotechnical & Geoenv. Eng., Vol. 131 (10): 1305–1308. Ooi, K., Chang, F. & Wang, S. 2004. Simplified lateral load analyses of fixed-head piles and pile groups, J. Geotechnical & Geoenv. Eng., Vol. 130 (11): 1440–1151. Rollins, M., Peterson, T. & Weaver J. 1998. Lateral load behavior of full-scale pile group in clay, J. Geotechnical & Geoenv. Eng., 124 (6):468–478. Rollins, M., Lane, D. & Gerber M. 2005. Measured and computed lateral response of a pile group in sand, J. Geotechnical & Geoenv. Eng., 131 (1): 103–111. Poulos, G. & Davis, H. 1980. Pile foundation analysis and design, John Wiley & sons, Inc., New York, N.Y. Reese, C., Cox, R. & Koop. D. 1974. Analysis of Laterally Loaded Piles in Sand, offshore Technology Conference, Houston: 473–483. Reese, C. & Matlock, H. 1956. Non-dimensional solutions for laterally loaded piles with soil modulus assumed proportional to depth, 8th Texas conference on Soil Mech. and Found. Eng., Austin: 1–41. Reese, C., Wang, T., Arrellaga, A. & Hendrix, J. 1997. LPILE plus 3.0 windows, Ensoft, Inc., Austin, Tex., USA. Reese, C., Wang, T. & Vasquez, L. 2006. Computer program GROUP version 7, tech. manual, Ensoft, Inc.,Austin, USA. Rao, S., Ramakrisha, V. & Rao, M. 1998. Influence of rigidity on laterally loaded pile groups in marine clay, J. Geotechnical & Geoenv. Eng., Vol. 124 (6): 542–549. Yang, K., & Liang, R. 2006. Numerical solution of laterally loaded piles in a two-layer soil profile, J. Geotechnical & Geoenv. Eng., Vol. 132 (11): 1436–1443.

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Behavior of piles under combined lateral and axial loading M. Achmus & K. Thieken Institute of Soil Mechanics, Foundation Engineering and Waterpower Engineering (IGBE), Leibniz University of Hannover, Germany

ABSTRACT: In many fields of application piles are used for the transfer of both axial and lateral loads into the subsoil. Combined loading leads to interaction between horizontal and vertical load bearing behavior, which is not usually taken into account in current engineering practice. Therefore numerical investigations were carried out to identify and quantify interaction effects for piles embedded in sand. In this paper the interaction effects are presented in terms of load-displacement curves and in terms of an interaction diagram, which represents the pile system behavior under arbitrary load inclinations. In a small parametric study, different pile diameters and different relative pile stiffnesses are considered to show the dependence of interaction effects on these quantities.

1

INTRODUCTION

Foundation piles are normally and favorably used to transfer axial loads. However, piles with a relatively large diameter can also be and often are exposed to lateral loads acting simultaneously. Due to this combined loading, interaction effects are to be expected, i.e. the horizontal load affects the vertical load bearing behavior and, vice versa, the vertical load affects the horizontal load bearing behavior. In current engineering practice, the interaction effects of combined loaded piles are not taken into account. The deformations in axial and lateral directions are calculated separately regarding only the loads acting in the corresponding direction. Numerous investigations have shown that significant interactions can result from the combined loading of piles in sand. However, it is not yet clear which features of the pile-soil system mainly affect the interaction behavior.

2

from θ = 0◦ (pure tension loading) to θ = 90◦ (pure horizontal loading). Based on their results, Das et al. proposed the following interaction approach (Eq. 1) with regard to the ultimate loads Qu

STATE OF THE ART

Leshukov (1975) carried out tests on 80 cm and 120 cm long piles with a quadratic cross section in silty fine sand. He applied oblique tension forces (combined tension loading) and analyzed the influence of inclination angle on the vertical ultimate load. He found that the ultimate load increases with inclination angles up to 45◦ against the vertical axis and decreases with larger inclination angles. Leshukov proposed taking the effects of combined loading at inclination angles between 10◦ and 40◦ into account. Das et al. (1977) carried out model tests with relatively short, practically rigid piles embedded in loose sand. The direction of the applied load varied © 2011 by Taylor & Francis Group, LLC

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According to this approach, the ultimate horizontal load is relatively more reduced by a vertical load than inversely the ultimate load by a horizontal load component. Ismael (1988) presented the results of field tests with combined tension loading on small bored piles (length L = 1.5 m, diameter D = 0.101 m) in medium dense sand. He concluded that the horizontal component of the ultimate pile load is only slightly affected by a tensile axial force acting simultaneously. For piles under combined tension loading in sand Sharour & Meimon (1991) carried out numerical simulations and concluded that the horizontal load deformation behavior of a pile is hardly or not at all affected by an axial load, whereas a horizontal load leads to a stiffness reduction in the axial direction. Further investigations were carried out by Patra & Pise (2006). They conducted model tests with single aluminum piles D = 19 mm/L = 722 mm in medium dense sand under combined tension load. They found that the vertical ultimate load may significantly increase due to an additional horizontal component. However, it should be noted that the ultimate load in this study was determined at a very large pile head displacement of up to 15 mm (i.e. 75% of the pile diameter). Meyerhof & Sastry (1985) carried out model tests with piles in loose sand under combined compression

Table 1.

Soil parameters for medium dense sand.

Unit weight γ Oedometric stiffness parameter κ Oedometric stiffness parameter λ Poisson’s ratio ν Internal friction angle ϕ Dilation angle ψ Cohesion c

Figure 1. Representation of system and loading parameters.

loads. With regard to the experimental results they proposed an interaction equation (Eq. 2).

Meyerhof and Sastry suggested calculating the vertical ultimate load Quv and the horizontal ultimate load Quh in Eq. (2) dependent on the load inclination angle, with a wall friction angle of δ = 0 for purely horizontal load (θ = 90◦ ) and δ = 0.6 ϕ for purely vertical (compressive) load (θ = 0◦ ). As a result, small vertical loads lead to a larger horizontal ultimate load than purely horizontal loads do. This was also observed in the experiments. Additionally, reference is made to the investigations of Yoshimi (1964), Chari & Meyerhof (1983), Sastry & Meyerhof (1990), Meyerhof (1995), Amde et al. (1997) and Abdel-Rahman & Achmus (2006). Altogether, the existing investigations give no clear view of the effects and their quantitative significance on the interaction under combined pile loading in sand. In most of these investigations, furthermore, only the ultimate loads and not the system stiffnesses are considered.

3

19.0 kN/m3 400 0.60 0.25 35.0◦ 2.5◦ 1.0 kN/m2

The numerical calculation was carried out in three stages. In the first step the initial stress state is generated with vertical stress σz = γ z and horizontal stress σh = γ z k0 for the whole model by using only soil elements. For the coefficient of horizontal earth pressure at rest k0 the usual approach for sand with k0 dependent on the angle of internal friction ϕ was applied: k0 = 1 − sin ϕ . Subsequently, the pile installation process was modeled by replacing the soil elements located at the pile position by pile (concrete) elements (“wished in place”) and activating the contact conditions between pile and surrounding soil. The initial state of loading and deformation is defined after the small pile settlement due to own weight occurred. The unit weight of the concrete pile was thereby determined with γ = 25 kN/m3 . In the final stage, the load was applied to the pile head and increased gradually until the ultimate load was reached. A clear failure state was in general only reached under predominantly tensile axial loads. Regarding compressive axial and horizontal loads, the ultimate loads were defined – as usually done in practice – as the loads inducing deflections of 10% of the pile diameter in the corresponding direction.

3.1

Material and contact modeling

For the simulation of the stress-strain behavior of the soil an elasto-plastic material law including the Mohr-Coulomb failure criterion and stress-dependent stiffness was chosen. In order to account for the nonlinear behavior of the soil a dependency of the stiffness modulus for oedometric compression Es on the mean principal stress σm was implemented as illustrated in Eq. (3). Here pref = 100 kN/m2 is a reference stress. The parameter κ determines the soil stiffness for a current stress state with σm = pref and the parameter λ rules the stress dependency of the oedometric stiffness.

NUMERICAL MODEL

In order to clarify the interaction behavior and to quantify the effects for piles embedded in sand, numerical simulations were carried out. For this, a three-dimensional model of the pile-soil system was established using the finite element program system Abaqus (Version 6.8, Abaqus 2006). The system and loading parameters are depicted in Fig. 1. © 2011 by Taylor & Francis Group, LLC

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For the reference system which is presented in the following, medium dense sand with the parameters given in Table 1 was considered. A small cohesion value was applied in order to enhance numerical stability.

different pile behavior. A negative effect of a horizontal load on the ultimate vertical load, as predicted by the interaction approach Eq. (2), was not confirmed.

4

Figure 2. Contact interaction approach between pile and adjacent soil.

Between pile and soil elements, an elasto-plastic contact behavior was simulated (Fig. 2). The elastic region was defined in a linear increase of the skin friction with relative displacement. This was assumed till reaching a limiting relative displacement uel,slip = 0.005 m, at which the skin friction was set to the maximum skin friction τfric,max resulting from the minimum of the product of horizontal stress σn and coefficient of friction µ and the limiting value of the skin friction τlimit . With regard to limiting values used in design practice (see API 2000), the limiting value was set to τlimit = 80 kN/m2 for medium dense sand. More details concerning configuration and discretization of the model can be found in Achmus et al. (2009). Additionally, it should be noted that geometric nonlinearity was taken into account in order to reflect the influence of relatively large pile deflections due to horizontal loading on the pile behavior.

As a reference system, a reinforced concrete pile (Ep = 30000 MN/m2 , ν = 0.2) with a diameter of D = 1 m and an embedded length of 15 m is considered. The general behavior of a horizontally loaded pile can be assessed by the ratio of the embedded length L to an “elastic length” Le . From the subgrade reaction theory with bedding modulus Ks increasing linearly with depth z (Ks = kr z) the following equation for the elastic length is obtained (e.g. Broms 1964).

With a bedding stiffness of kr = 40 MN/m3 typical for medium dense sand (cf. API 2000) the ratio of the embedded length to the elastic length is L/Le = 7.29 for the reinforced concrete pile considered. Flexible pile behavior with two zero deflection points, i.e. rigid clamping, is normally to be expected when L/Le > 4 to 5. Thus, the reference system reflects a flexible pile. To evaluate all interaction effects of a pile-soil system, numerous inclination angles between α = −90◦ (pure compression loading) over α = 0◦ (pure horizontal loading) up to α = 90◦ (pure tension loading) had to be analyzed. In the course of the study it became clear that the interaction effects depend less on the absolute inclination angle α and more on the normalized inclination angle αnorm (Eq. 5).

3.2 Comparison with experimental results In order to validate the numerical model, backcalculations of the well-reported model tests of Das et al. (1977) for inclined tension loading and of Meyerhof & Sastry (1985) for inclined compression loading were carried out. In these calculations, the soil parameters reported in the aforementioned papers were applied. Due to limited space, the results cannot be presented in detail here. However, the experimental and numerical results agreed well at least qualitatively and in general also quantitatively. Moreover, deficiencies in the experimental investigations were identified by the numerical simulations. Both the interaction approaches regarding ultimate loads proposed by Das et al. (Eq. 1) and by Meyerhof & Sastry (Eq. 2) were found to be unsuitable. In the experiments of Das et al., very large vertical ultimate loads led to a misinterpretation of interaction effects for small horizontal loading portions. The interaction approach according to Meyerhof & Sastry could not be confirmed either although the measured results of the model test agreed very well with the numerical results. However, the numerical analyses showed that the size of the test box used by Meyerhof & Sastry was not sufficient to avoid influences of the boundary on the pile behavior. Assuming a larger test box size in the numerical simulations yielded significantly © 2011 by Taylor & Francis Group, LLC

RESULTS FOR A REFERENCE SYSTEM

Here, for Vult the ultimate compression load (Vult,α=−90 = Vult,c ) or the ultimate tension load (Vult,α=90 = Vult,t ) must be used when combined compression or combined tension loading is considered. For the reference system Hult = 1.54 MN, Vult,c = −4.59 MN and Vult,t = 1.61 MN results. Thus, the absolute value of αnorm is smaller than α for both compression and tension loading. In Fig. 3 the horizontal load-displacement curves obtained for the reference system are presented. With only one exception, namely for the largest load inclination of 61.09◦ for combined tension loading, only minor interaction effects arise. Only the enlarged depiction of the beginning of the load-deflection curves on the right side of Fig. 3 shows differences in the system stiffnesses, which are nevertheless small. For combined compression loading first an increase in the horizontal system stiffness with increasing vertical load is obtained. The reason for this is that the mobilizable passive earth pressure and thus the stiffness in the horizontal direction increases due to vertical

467

Figure 3. Horizontal load-displacement curves (reinforced concrete pile D = 1 m, L = 15 m, medium dense sand).

Figure 4. Vertical load-displacement curves (reinforced concrete pile D = 1 m, L = 15 m, medium dense sand).

downward-directed shear stresses induced by the vertical load transferred via the pile shaft into the ground near the surface. This effect was described by AbdelRahman & Achmus (2006). For higher load levels and thus larger horizontal displacements the importance of this effect decreases. This results from the additional bending moment in the pile due to the moment arm of the vertical load, i.e. from geometrical nonlinearity, which was considered in this investigation. For combined tension loading the horizontal stiffness is decreased, since here the shear stresses acting on the pile shaft are directed upwards and thus reduce the mobilizable passive earth pressure.A distinct effect on the horizontal ultimate load is obtained only for combined tension loading with a large vertical load portion (α = 61.09◦ or αnorm = 60◦ , see Fig. 3). The decrease is induced by large deformations occurring when the ultimate vertical load is approached, which in this case happens due to the large vertical load portion. Besides, the interaction effects on stiffness and ultimate load in the horizontal direction are small, which is in agreement with the general findings of earlier investigations. In Fig. 4 the calculation results for the reference system are shown in terms of vertical load-displacement curves. Here, a larger dependence of the pile-bearing behavior on the load inclination angle than for the horizontal direction is obtained. © 2011 by Taylor & Francis Group, LLC

For combined compression loading a favourable effect of the horizontal load on the pile behavior in the vertical direction is found. The horizontal force increases the horizontal stresses in front of the pile, which leads to larger mobilizable skin friction stresses in the ultimate state. Also with small pile displacements a favourable effect arises, since the upward-directed movement of the passive earth pressure wedge induces a prestressing of the pile (see also Achmus et al. 2009). However, for the reference system this effect is of minor importance. Moreover, for greater horizontal loads the effect is counteracted by the increase in axial deflection due to geometrically non-linear effects. For combined tension loading the opposite effect appears. The upwards-directed movement of the passive earth pressure wedge leads to negative skin friction stresses and thus to a vertical stiffness reduction. Only when large vertical displacements (heave) occur is negative skin friction reduced, and finally even an increase in the ultimate vertical load can arise (α = 61.09◦ , s. Fig. 4). However, with a relatively large horizontal loading part (α = 31.12◦ , see Fig. 4) no increase in the ultimate load occurs, since even with a small pile heave the horizontal ultimate load is reached. For a clear and comprehensive presentation of interaction effects on the pile behavior, the calculation results can be depicted in an interaction diagram. From numerous load-displacement curves calculated with different load inclination angles, the load combinations (H/V) corresponding with certain displacements were derived. From that, lines of equal displacements can be constructed in a H-V interaction diagram. The respective diagram derived for the reference system is shown in Fig. 5. A presentation in a dimensionless form, i.e. the load components are related to the respective ultimate loads without accounting for interaction, was found to be suitable. The displacement values belonging to the different curves shown were taken from the load-deformation curves for purely axial and horizontal loading, respectively. A horizontal course of the curves ux = const or a vertical course of the curves uy = const would mean that no interaction effects occur. Accordingly, the deviation of the curves to the grey-coloured quadratical mesh is a measure of the importance of the H-V interaction effects. Moreover, the load-displacement curves for any load inclination angle can be obtained from an interaction diagram.

5

INFLUENCE OF PILE DIAMETER

The significance of interaction effects is influenced by numerous parameters regarding pile geometry, soil behavior and loading conditions. For practical design it is important to know in what conditions interaction effects become particularly important and should therefore not be neglected. Although a comprehensive parametric study is beyond the scope of this paper, the

468

Figure 6. Comparison of interaction curves for piles with different diameters; left: reference load defined at ux,y = 0.1 D; right: reference load defined at ux,y = 5 cm.

Figure 5. Interaction diagram for the reference system.

Table 2. Investigated systems with varying pile diameter.

Pile

L/D –

Le m

L/Le –

Class –

D = 0.5 m; L = 15 m D = 1.0 m; L = 15 m D = 2.0 m; L = 15 m D = 3.0 m; L = 15 m

30.0 15.0 7.5 3.0

1.18 2.06 3.58 4.95

12.70 7.29 4.19 3.03

Flexible (long) Flexible (long) Nearly rigid Rigid (short)

effect of varying the pile diameter and with that the relative pile stiffness on the magnitude of interaction effects is shown here. Based on the parameters of the reference system, the pile diameter was varied between 0.5 m and 3 m. The parameters of the systems considered are given in Table 2. The piles with D = 0.5 m and D = 1 m can be classified as flexible, whereas the piles with larger diameters are nearly rigid and rigid. Since the quality of interaction is dependent on the load level, two different load levels are considered in the following. On one hand, in order to elucidate interaction effects at a typical service load level, the interaction curves beginning at 50% of the ultimate vertical and horizontal loads are plotted and compared. On the other hand, the interaction curves for the ultimate loads are considered. Two different kinds of interaction diagram are given in Fig. 6. For the diagram in the left part of Fig. 6 the aforementioned definition of failure loads was used, © 2011 by Taylor & Francis Group, LLC

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i.e. at a pile head deflection of 10% of the pile diameter. Since the pile diameter of the compared systems is different, this definition affects the course of the curves, which makes an assessment of the effect of only the pile diameter on interaction quality difficult. To eliminate this, a second comparison of the interaction effects is shown (Fig. 6 right). In this case the reference load is defined as the load which appertains to a deflection of 5 cm in both vertical and horizontal direction, constant for all considered systems. Obviously, the consideration of different horizontal deflections affects the interaction curves. First of all, it has to be stated that there is no uniform effect of the pile diameter and, with that, the pile stiffness on the quality of interaction, since for different loading type (load inclination and load level) different quantitative effects arise. However, with regard to the horizontal load-bearing behavior it can be stated that the effects are by trend more significant, the larger the diameter or the more rigid the pile behavior becomes. Regarding the vertical load-bearing behavior, only slight differences in the interaction effects with respect to pile diameter are obtained if the reference ultimate load is defined at horizontal and vertical pile deflections of ux,y = 5 cm. Larger differences occur when the reference load definition dependent on the pile diameter is used. This is in particular valid for the service load level in combined tension loading (Fig. 6 left). Regarding different pile diameters, the following effects influence the interaction behavior: • An increase in the pile diameter (at constant pile

length) increases the pile stiffness, i.e. leads to a more rigid behavior. With that, the depth of the area in which significant horizontal pile deflections occur is extended, which leads to larger passive earth pressures and so to a more pronounced interaction.

For rigid pile systems therefore greater interaction effects occur than for flexible systems. This was also found in a comparison of systems with constant pile diameter, but variable pile length (results not shown here). • Large horizontal pile deflection leads in all cases to a decrease of system stiffness. The reduction in vertical stiffness results from the maximum value of skin friction τlimit , which limits the skin friction on the passive side of the pile. Simultaneously the skin friction on the active side is decreasing due to horizontal deflection. With regard to horizontal stiffness, the influenced area increases with increasing horizontal deflection. This leads to greater interaction effects. Moreover, the consideration of geometrical nonlinearity leads to a further reduction of the system stiffness. All in all, a tendency to a greater significance of interaction effects for piles with large diameter-tolength ratio was established.

6

CONCLUSIONS

The numerical investigations showed that combined loading of piles in sand induces complex interaction effects. Depending on load inclination, load level and direction of the vertical load component both unfavourable and favourable effects regarding system stiffness and ultimate load of a pile-soil system can occur. For combined compression loading horizontal as well as vertical system stiffnesses are positively influenced, i.e. increased. By contrast, combined tension loading leads to a negative influence on horizontal as well as vertical system stiffness. However, if the vertical load portion dominates, an increase in the vertical ultimate load can happen due to larger normal stresses acting on the pile shaft induced by the horizontal load. A comparison of the interaction effects for piles with different diameters and otherwise identical system parameters showed that by trend interaction effects are the more important, the larger the pile diameter and thus more rigid the pile system is. However, since many different influences affect the interaction behavior, this statement cannot be generalized with regard to the pile behavior under vertical load. Further parametric studies and a more accurate assessment of the significance of pile geometry, soil type and loading conditions are the subject of ongoing investigations. Moreover, since experimental evidence regarding interaction effects under combined loading is scarce, a systematic experimental test programme would be highly desirable.


ACKNOWLEDGEMENT The presented study was carried out as part of a research project funded by the German Research Council (DFG, project no. AC 100/4-1). The authors are grateful for the financial support. REFERENCES Abaqus. 2006. User’s Manual, Version 6.8. Abdel-Rahman, K., Achmus, M. 2006. Numerical modeling of the combined axial and lateral loading of vertical piles. 6th European Conference on Numerical Methods in Geotechnical Engineering., Graz, Austria, September. Achmus, M.,Abdel-Rahman, K & Thieken, K. 2009. Numerical study of the effect of combined loading on the behavior of piles in sand. International Symposium on Computational Geomechanics, Juan-Les-Pins, France, May Amde, A.M., Chini, S.A., Mafi, M. 1997. Model study of H-piles subjected to combined loading. Geotechnical and Geological Engineering (15): 343–355. API. 2000. American Petroleum Institute. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – Working Stress Design. API Recommended Practice 2A-WSD (RP2A-WSD), 21st edition, Dallas. Broms, B. 1964. Lateral Resistance of Piles in Cohesionless Soils. ASCE Journal of the Soil Mechanics and Foundation Division 90 (3): 123–156 Chari, T.R., Meyerhof, G.G. 1983. Ultimate capacity of single rigid piles under inclined loads in sand. Canadian Geotechnical Journal (20): 849–854 Das, B.M., Seeley, G.R., Raghu, D. 1977. Uplift Capacity of Model Piles under Oblique Loads. ASCE Journal of the Geotechnical Engineering Division 102(9): 1009–1013 Ismael, N.F. 1989. Field Tests on Bored Piles Subject to Axial and Oblique Pull. Journal of Geotechnical Engineering, 115 (11): 1588–1598 Leshukov, M.R. 1975. Effect of Oblique Extracting Forces on Single Piles. Togliatti Polytechnic Institute. Translated from Osnovaniya, Fundamenty i Mekhanika Gruntov, No. 5, p. 15, Sept.–Oct. 1975 Meyerhof, G.G. 1995. Behaviour of Pile Foundations under Special Loading Conditions. 1994 R.M. Hardy Keynote Address. Canadian Geotechnical Journal (32): 204–222 Meyerhof, G.G., Sastry, V.V.R.N. 1985. Bearing capacity of rigid piles under eccentric and inclined loads. Canadian Geotechnical Journal (22): 267–276 Patra, N.R., Pise, P.J. 2006. Model pile groups under oblique pullout loads – an investigation. Geotechnical and Geological Engineering (24): 265–282 Sastry, V.V.R.N., Meyerhof, G.G. 1990. Behaviour of flexible piles under inclined loads. Canadian Geotechnical Journal 27(1): 19–28. Shahrour, I., Meimon, Y. 1991. Analysis of behaviour of offshore piles under inclined loads. International Conference on Deep Foundations, pp. 227–284. Yoshimi, Y. 1964. Piles in cohesionless soil subject to oblique pull. ASCE Journal of the Soil Mechanics and Foundation Division, 90 (6): 11–24.

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Investigations on the behavior of large diameter piles under cyclic lateral loading M. Achmus, J. Albiker & K. Abdel-Rahman Institute of Soil Mechanics, Foundation Engineering and Waterpower Engineering, Leibniz University of Hannover, Germany

ABSTRACT: Large diameter monopiles are an established foundation type for offshore wind energy converters (OWEC), although currently practical experience in the design and the behavior of these constructions is rare. Achmus et al. (2008) developed a method for estimating the permanent pile deformation under one-way loading which is to be expected over the lifetime of an OWEC. The procedure, called the stiffness degradation method (SDM), is based on the combination of cyclic drained triaxial test results with numerical simulations. Although it yields plausible results, further validation of the method by comparison with experimental results is very important to ensure its applicability. Recently, LeBlanc et al. (2010) presented the results of a series of 1g-model tests of a monopile in sand under cyclic horizontal loading. These test conditions were modeled with the SDM under drained conditions in order to verify the procedure. The comparison of numerical and test results proves the applicability of the stiffness degradation method. 1

INTRODUCTION

The planned offshore wind farms in the German parts of the North Sea and the Baltic Sea will be constructed in water depths varying from approximately 15 to 40 m. By means of suitable foundation constructions, the large horizontal forces and bending moments resulting from wind and wave loads must be economically and safely transferred to the sea soil. Monopile foundations can be used as one of these foundation types. This foundation method was already implemented for OWECs in the North and the Baltic Sea, but only in water depths of less than about 15 m. Its application is expected to be extendable for water depths up to about 25 to 30 m. However, the diameters of such monopiles will then vary between 5.0 and 7.5 m (Fig. 1). Since wind energy converters are relatively sensitive to deformations, in particular tilting, it is very important to estimate these as exactly as possible. For the mentioned large-diameter piles, there is to date no approved procedure for this. In this paper, the special numerical concept SDM is described, and the results are compared with experimental ones derived from 1g-model tests performed for monopiles under cyclic loading. This method was used in this paper to predict the behavior of monopiles under drained condations.

2

SIMULATION OF THE MONOPILE BEHAVIOR UNDER STATIC LOADING

A three-dimensional (3D) finite element model was established in order to analyze the behavior of © 2011 by Taylor & Francis Group, LLC

Figure 1. System and denominations for a monopile foundation.

monopiles embedded in sand soil. The computations were carried out using the finite element program system ABAQUS (ABAQUS 2008). The most important issue in geotechnical numerical modeling is the simulation of the soil stress-strainbehavior. An elasto-plastic material law with MohrCoulomb failure criterion was used. The soil stiffness is here represented by a stiffness modulus for oedometric compression Es and a Poisson’s ratio ν. To account for the non-linear soil behavior, a stress dependency

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Table 1.

Material parameters used for dense sand.

Unit buoyant weight γ Oedometric stiffness parameter κ Oedometric stiffness parameter λ Poisson’s ratio ν Internal friction angle ϕ Dilation angle ψ Cohesion c

11.0 kN/m3 800 0.55 0.25 37.5 7.50◦ 0.1 kN/m2

of the stiffness modulus was implemented as follows, according to Ohde (1939): Figure 2. Deflection lines of a monopile (D = 2.0 m) calculated using Lpile (2000) and FEM.

Herein σat = 100 kN/m2 is a reference (atmospheric) stress and σm is the current mean principal stress in the considered soil element. The parameter κ determines the soil stiffness at the reference stress state and the parameter λ rules the stress dependency of the soil stiffness. The material parameters used here are typical for dense sand and are given inTable 1. For more details about the numerical modeling see Abdel-Rahman & Achmus (2005). The stress-dependency of the stiffness modulus given by Equation 1 is widely used in soil mechanics. However, no direct experience exists on the magnitude of the two parameters (κ, λ) to be used in the calculation of horizontally loaded piles. In order to calibrate these parameters in connection with the numerical model, firstly monopiles of smaller diameters were investigated. For such diameters the p-y-method is known to give a reasonable estimation of pile deflection, so the numerical results could be compared with the results of the p-y-method for calibration.The calculations with the p-y-method were carried out by means of the Lpile program (Lpile 2000). The calculated deformations of a monopile with a diameter D = 2.0 m with different embedded lengths varying from 20 to 40 m with a wall thickness of 3.0 cm under monotonic loading are shown in Figure 2 and compared with the p-y-method results. The loading consisted of a horizontal force acting at a height h above the soil surface. The magnitude of the load was varied between 0.5 and 4.5 MN. The results are valid for κ = 800 and λ = 0.55. This parameter combination was found to give the best matching results with respect to the p-y-method. This was also verified with similar comparisons for a pile of a diameter D = 1 m and embedded lengths of 20 to 40 m. Further investigations have been made by the authors; see Achmus et al. (2008). 3

SIMULATION OF THE MONOPILE BEHAVIOR UNDER CYCLIC LOADING

Figure 3. Degradation of secant modulus under cyclic loading in a drained triaxial test.

pile-soil system by numerical calculations, taking the behavior of soils under cyclic loading investigated in cyclic triaxial tests into account (Achmus et al. 2009, Kuo 2008). This method is based on the finite element model presented above and accounts for cyclic loading by a special stiffness degradation approach. A sketch of principles of the results of a stresscontrolled cyclic triaxial test under drained conditions is shown in Figure 3. The plastic portion of the axial strain εap increases with the number of load cycles. The increase rate of the plastic strain is mainly dependent on the initial stress state (confining stress) and on the magnitude of the cyclic load portion. The strain increase can be interpreted as a decrease in the secant stiffness modulus. When the elastic strain is negligible, the degradation of the secant modulus EsN can be formulated in the following way dependent on the plastic strain in the first cycle εap,N =1 and in the Nth cycle εap,N :

To investigate the lateral deformation response of a monopile under cyclic loading, a method was developed which yields the permanent displacements of a © 2011 by Taylor & Francis Group, LLC

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The degradation of secant stiffness in a cyclic triaxial test with isotropic initial stress condition can be determined from the plastic strains measured with a regression equation. Such equations were presented, for instance, by Huurman (1996), Gotschol (2002) and Werkmeister (2004). Due to the approach of Huurman used here, the increase in deformation or the decrease in stiffness, respectively, can be described by the following equation:

Here N is the number of cycles, X is a stress-dependent variable (cyclic stress ratio), and b1 , b2 are regression parameters to be determined in triaxial tests. The cyclic stress ratio is defined as

wherein σ1,f is the main principal stress at failure in a monotonic test. Thus, the stress ratio is dependent on the initial stress state and on the cyclic load level. A problem to be dealt with is that the Equations (3) and (4) are valid for triaxial test conditions with isotropic initial stress conditions and a constant confining pressure σ3 during cyclic loading. In the pile-soil system, the initial stress conditions (before application of the horizontal load) are anisotropic and the minor principal stress in the elements as well as the direction of the principal stress axes in general change with the application of the load. To overcome this problem, a characteristic cyclic stress ratio Xc is defined here as

Here the index (1) indicates the cyclic stress ratio at loading phase and the index (0) at unloading phase. At the initial (and unloading) phase, only the vertical load V due to the tower weight is considered, and the lateral load H is applied subsequently in the loading phase. The characteristic cyclic stress ratio is derived from the difference between the stress ratios in the loading and the unloading phase. The accumulation of plastic strain and the degradation of stiffness of the soil element can be obtained from Equation (3) by replacing X by Xc . Figure 4 shows the lateral deflection of monopiles with different embedded lengths (L = 20 m and 40 m) and with a wall thickness tp of 9 cm calculated using the SDM. The long pile shows a better cyclic performance than the shorter one, which could be explained by the different loading levels (H /Hu ). The ultimate horizontal loading (Hu ) was determined here using the hyperbolic method by Manoliu et al. (1984). Figure 5 shows that the accumulation rate for the pile with longer embedded length and, with that, a lower cyclic loading level is smaller than for the shorter © 2011 by Taylor & Francis Group, LLC

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Figure 4. Lateral pile deflection obtained from stiffness degradation model.

Figure 5. Accumulated displacement of a pile under cyclic loading (D = 7.5 m, h = 20 m, H = 15 MN).

pile with a higher cyclic loading ratio. This means that the accumulation rate is a function of the cyclic loading level (H /Hu ). This seems obvious, but with existing approaches an effect of the loading level is not reflected. The new stiffness degradation method is capable of taking soil, geometry and loading conditions into account. For detailed parametric study, refer to Achmus et al. (2008) and Achmus et al. (2009). 4 VALIDATION OF THE STIFFNESS DEGRADATION METHOD For validation of the described method, comprehensive comparison with experimental results is needed. Recently, LeBlanc et al. (2010) (see also LeBlanc 2009) presented the results of a series of 1g-model tests of a monopile in sand under cyclic horizontal loading and developed an approach for estimating the accumulation of plastic deformation with the number of load cycles. The test conditions were modeled with the SDM in order to compare the numerical results with the test results. 4.1 Test program carried out by LeBlanc et al. (2010) In the laboratory model a stiff copper monopile was installed in a basin filled with unsaturated sand. The pile was loaded by a sinusoidal cyclic load, which acted at the top of the pile and was generated by a specially constructed loading rig. The horizontal force creates a moment M = H × h acting at the ground surface,

Table 2. Properties of the small scale copper pile used in the laboratory by LeBlanc et al. (2010). Pile diameter, D (m) Wall thickness, tp (m) Penetration depth, L (m) Load eccentricity, h (m) Pile weight, V (kN)

0.08 0.02 0.36 0.43 0.035

Table 3. Dimensionless parameters introduced by LeBlanc et al. (2010). Moment loading Vertical force Pile rotation (degree) Load eccentricity Slenderness ratio

˜ = M /(L3 Dγ ) M V˜ = V/(L2 Dγ ) θ˜ = θ pa /(Lγ ) e˜ = M /(HL) = h/L η = L/D

where h is the load eccentricity. The secant rotation of the pile was measured by applying two deflectometers at different positions. The dimensions of the eccentrically loaded pile are shown in Table 2. The values of the parameters of the laboratory pile had to be determined accurately in order to be able to compare the test results to the deformations of a typical offshore monopile. For structures in sand the load response is governed by the frictional behavior, which in turn is dependent on the isotropic stress level. The much lower isotropic stress level in the laboratory, compared to a full-scale test, influences the values for the friction angle and the relative density. These issues of scaling are addressed by applying adequate scaling methods, which leads to the development of a complete non-dimensional framework. For a detailed description see LeBlanc et al. (2010). As a result the dimensionless value for the stiffness k˜ is expressed as a function of three further non-dimensional parameters, i.e. the vertical force V˜ , the load eccentricity e˜ , and the pile slenderness ratio η. The non-dimensional moment-rotation relationship in turn is expressed by the following equation:

Thus, a similar behavior of model and prototype is to be expected with regard to both stiffness and strength ˜ against when plotting the non-dimensional moment M the non-dimensional rotation θ˜ , while retaining the three parameters controlling the stiffness constant. Table 3 gives an overview of the definitions of the introduced dimensionless parameters. The parameter pa labels the atmospheric pressure of 100 kN/m2 . The test program carried out by LeBlanc et al. (2010) was developed to investigate the response of the pile and its dependency on the relative density of the sand and the characteristics of the applied loading. In order to choose appropriate values for the relative density in the model, a relation of Schnaid (1990) was used, which relates relative density, friction angle and © 2011 by Taylor & Francis Group, LLC

effective vertical stress. Two test series were conducted at relative densities of Dr = 4% and Dr = 38%, respectively. The peak friction angles used in the laboratory were estimated 35◦ and 43◦ . Under field conditions, these parameter values correspond to relative densities of Dr = 8% and Dr = 75%, respectively, due to the much higher effective vertical stress. The loading amplitude is expressed by the maximum moment of the applied loading in a cycle Mmax , normalized by the static moment capacity MR . Furthermore, a loading cycle is characterized by the values of the minimum and the maximum applied moment, Mmin and Mmax . Hence, for describing the specific loading conditions LeBlanc et al. (2010) introduced the parameters ζb and ζc :

ζb was chosen in a range of between 0.3 and 0.5 to reflect realistic loading conditions. ζc lies in a range of between −1.0 and +1.0, where values of ζc < 0 characterize two way loading. 4.2

Simulation with the SDM

For determining the static moment capacity of the model pile, LeBlanc et al. (2010) initially carried out static load tests in both loose and medium dense sand, ˜ against θ˜ . Here, in order to validate and plotted M the developed numerical model, these static tests were simulated in the original small scale dimensions, and the results were compared to the laboratory test results. Both relative densities of Dr = 4% and Dr = 38% were considered. The values for κ and λ were chosen as 400 and 0.6, respectively, for the loose state (Dr = 4%) and as 800 and 0.5 for the medium dense state (Dr = 38%). The values for the unit weights of the unsaturated sand were γ = 14.7 and 15.8 kN/m3 , respectively. The friction angles in the simulations were varied with ϕ = 32◦ and 35◦ for loose sand and ϕ = 37.5◦ , 40◦ and 43◦ for medium dense sand. The comparison of numerically obtained results and laboratory results is shown in Figure 6. LeBlanc et al. (2010) estimated the friction angles valid for small stresses to 35◦ and 43◦ , respectively. Applying these values in the simulation, the ultimate load is overestimated. Using smaller, but still reasonable values, a fairly good agreement is obtained. However, the agreement with regard to loads less than about 50% of the ultimate load is in all cases very good, which means that at least for service loads the numerical model yields very good results. This is the load range considered here. The validation of the developed numerical model with regard to static loading was therefore successful. For modeling with the SDM, here only the test series with the pile embedded in medium dense sand

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Figure 6. Dimensionless moment-rotation curves obtained by numerical simulation and by LeBlanc et al. (2010). Table 4. Tests chosen for back-calculation with the SDM. e˜

Dr

ζb

ζc

N

1.19 1.19 1.19

38% 38% 38%

0.27 0.40 0.52

0 0 0

8090 7423 17532

Table 5. Pile properties of a typical full-scale offshore pile, scaled by applying the scaling law of LeBlanc et al. (2010). Pile diameter, D (m) Wall thickness, tp (m) Penetration depth, L (m) Load eccentricity, h (m) Pile weight, V (kN)

4.44 0.13 20 23.8 6284.2 Figure 7. Accumulated pile rotation θ(N )/θs against number of load cycles N, results of the laboratory tests and of the numerical simulations with SDM.

with Dr = 38% was regarded. Furthermore only oneway loading with ceasing of the applied moment after each load cycle can be simulated using the SDM, i.e. only simulations where ζc equals 0. Three different tests were chosen for back-calculation with the SDM, which are characterized by the parameters given in the following Table 4. The tests were simulated in the original small scale parameter configurations with the geometrical dimensions given in Table 2 and, furthermore, under application of the mentioned scaling law, in dimensions and parameter configurations that characterize a typical full-scale offshore pile. Table 5 shows the associated pile properties. LeBlanc et al. (2010) presented the results of the laboratory tests in terms of the evolution of the accumulated rotation of the pile by plotting the rotation resulting from cyclic loading θN , diminished by the rotation after the first cycle θ0 , in terms of the rotation θS that would occur in a static test when the applied load equals the maximum cyclic load:


In Figure 7 a) to c) the laboratory results are compared to the results obtained numerically by applying the SDM. For the calculations presented here, the cyclic parameters b1 and b2 were chosen with regard to experience (Kuo 2008), since no suitable cyclic triaxial test results were reported. Values of 0.16 and 0.38, respectively, were used when calculating the system in original dimensions. These values are typical for a medium dense sand. For the system in full scale dimensions values of 0.2 and 5.76 typical for dense sand were used. LeBlanc et al. (2010) found that the pile behavior can be predicted by the following equation:

where Tb and Tc are dimensionless functions, depending on the load characteristics and the relative density. Based on the previous equation it follows that the accumulated rotation increases exponentially with N . For the test results given in Figure 7 (a, b & c) the exponent α was found to be 0.31. Also from the numerical

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modeling the accumulated rotation increases exponentially, though this finding is not absolutely appropriate for low values of N , i.e. here for N = 10. Disregarding the results for this number of load cycles, the exponent describing the increase of the accumulated rotation in the simulations was found to vary between 0.22 and 0.24, slightly ascending with the amplitude of the applied cyclic load. However, it should be noted that the simulation results were obtained with estimated cyclic parameters b1 and b2 , which of course strongly affect the cyclic behavior. For future applications, it could be beneficial to search for a physical way to determine the values of these parameters, in order to make the SDM reflect more realistic results. However, from the presented calculations it is obvious that the numerical results match the model results of LeBlanc et al. (2010) quite well. The laboratory small-scale results show that the accumulated rotation is higher under a higher value of loading level (ζb ). From the numerical modeling using SDM, similar results were also obtained, whereby the dependency between cyclic loading level and accumulated rotation is not as strong as found in the laboratory results. 5

CONCLUSIONS

The stiffness degradation model as a numerical concept enables the estimation of permanent pile displacements and rotations under cyclic lateral loading. The comparison with the model presented by LeBlanc et al. (2010) model results shows good agreement both qualitatively and quantitatively. Despite of SDMsimplifications, this method is suitable to predict the behavior of monopiles under drained conditions in a realistic manner. It can be used to assess the influence of pile geometry and soil type on the performance of piles under cyclic loading. ACKNOWLEDGEMENTS

REFERENCES ABAQUS User’s Manual, Version 6.7. 2008. Simulia, Providence, RI, USA. Abdel-Rahman, K. &Achmus, M. 2005. Finite Element Modelling of Horizontally Loaded Monopile Foundations for Offshore Wind Energy Converters in Germany, International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Perth, Australia. Achmus, M., Abdel-Rahman, K. and Kuo, Y.-S. 2008. Design of Monopile Foundations for Offshore Wind Energy Plants, 11th Baltic Geotechnical Conference – Geotechnics in Maritime Engineering, Gdansk, Poland, Vol.1, pp. 463–470 Achmus, M., Kuo,Y.-S. and Abdel-Rahman, K. 2009. Behavior of monopile foundations under cyclic lateral load, Computers & Geotechnics 36 (2009), pp. 725–735 Gotschol, A. 2002. Veränderlich elastisches und plastisches Verhalten nichtbindiger Böden und Schotter unter zyklisch-dynamischer Beanspruchung, Ph.D. thesis, Universität Kassel, Kassel, Heft 12. Huurman, M. 1996. Development of traffic induced permanent strains in concrete block pavements, Heron, Vol. 41, No. 1. pp. 29–52. Kuo, Y.-S. 2008. On the behavior of large-diameter piles under cyclic lateral load, Ph.D. thesis, Leibniz Universität Hannover, Hannover, Heft 65. LeBlanc, C. 2009. Design of Offshore Wind Turbine Support Structures, Ph.D. thesis, Aalborg University, Denmark, DCE Thesis No. 18. LeBlanc, C., Houlsby, G. T. & Byrne, B. W. 2010. Response of stiff piles in sand to long-term cyclic lateral loading, Geotechnique, Vol. 60, Issue 2, pp. 79–90. Lpile, 2000. User’s manual, Version Lpile plus 4.0. Manoliu, I., Dimitriu, D. V. & Dobrescu, GH. 1985. Loaddeformation characteristics of drilled piers, Proc. 11th International Conference on Soil Mechanics and Foundation Engineering, San Francisco, Vol. 3, pp. 1553–1558. Ohde, J. 1939: Zur Theorie der Druckverteilung im Baugrund. Der Bauingenieur 20, pp. 451–459 (in German). Schnaid, F. 1990. A study of the cone-pressuremeter test in sand, Ph.D. thesis, University of Oxford. Werkmeister, S. 2004. Permanent Deformation Behaviour of Unbound Granular Materials in Pavement Constructions, Ph.D. thesis, Techn. Universität Dresden, Dresden, Heft 12.

The results presented in this paper were obtained as part of the GIGAWIND ALPHA VENTUS research group project funded by the Federal Ministry for the Environment, Natural Conservation and Nuclear Safety, Germany. The support is thankfully acknowledged.


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BP Clair phase 1 – Pile driveability and capacity in extremely hard till T.R. Aldridge & T.M. Carrington Fugro GeoConsulting Limited

R.J. Jardine Imperial College, London

R. Little Fugro GeoConsulting, Inc. (Formerly of Fugro-McClelland)

T.G. Evans EPT, BP Exploration

I. Finnie Advanced Geomechanics (Formerly of Lloyds Register)

ABSTRACT: Evans et al. (2010) describe how the foundation engineering for BP’s Clair Phase 1 Drilling and Production Platform, West of Shetland, UK, had to consider tills with unprecedentedly high undrained shear strengths and unit weights. Boulders were also present. This paper describes the technical approach taken by BP’s foundation assurance team in addressing these challenging conditions, focusing principally on driveability and axial capacity. Advanced field and laboratory investigations were conducted to allow a range of analyses that explicitly considered the effects of cyclic loading, group action, strain softening and possible pilot hole drilling. Instrumented advance driving trials were conducted, while the main jacket installation was also instrumented and back-up drilling options mobilised in case of harder-than-expected driving.

1

INTRODUCTION

A companion paper (Evans et al., 2010) describes the development planning for the Clair field and the strategy adopted by BP and their co-venturers in assuring foundation design under unprecedented geotechnical conditions. The four legged Clair Phase 1 jacket was installed with 3 or 4 2590 mm o.d. × 85–95 mm w.t. tubular sleeve piles driven per corner. This paper describes the technical approach taken in assessing foundation design from basic principles. The work was led by an international assurance team (IAT) assembled by BP to provide industrial and academic experience. The lessons learnt are now being applied in the new Clair Ridge development.

Bulk densities were up to 2.4 Mg/m3 despite ‘standard’ particle specific gravities (2.65–2.75), indicating very low void ratios. Geophysical data and well drilling records both indicated boulders at shallow depths. Boulders or cobbles occurred every 6 m on average in the geotechnical boreholes, but were more frequent close to the geological soil boundaries at 12, 22 and 33 metres below seafloor. The sandy clay tills above the 12 m boundary were the toughest, with some CPT qc pushes up to 120 MPa and some UU su tests maxima >50. High exceeding 2500 kPa, giving su /σvo su /σvo ratios dominated in the tills, but lower strengths < 0.5) were apparent either side of the (with su /σvo 22 m boundary. These softened, possibly weathered, units were also associated with a very dense sand layer encountered between 24 and 26 m.

2 THE CLAIR 1 GEOTECHNICAL PROFILE 3 ADVANCED GEOTECHNICAL TESTING Standard drilling and sampling techniques gave poor results at Clair. A second investigation using the Geobore “S” rotary coring system provided far better samples, and high capacity CPT equipment provided a more accurate geotechnical profile. Fig. 1 summarises the sequence of very hard and dense low plasticity sandy and silty boulder clays, which originate from the Stormy Bank, Otter Bank and Ferder formations. © 2011 by Taylor & Francis Group, LLC

An advanced geotechnical test programme was performed to support ‘first principles’ foundation engineering. Data was required for effective stress design methods and the analysis of cyclic response. For the ‘MTD’ approach (Jardine and Chow, 1996) oedometer tests on remoulded and reconstituted soil were conducted, along with soil-steel ring shear interface

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Figure 1. Soil parameters at Clair.

tests and clay sensitivity measurements. Design YSR data, noting profiles were developed from the su /σvo that the K0 values required for stress path testing might be lower than expected by assuming a stress history of monotonic K0 overconsolidation. Cyclic direct simple shear tests, anisotropic triaxial tests with bender elements, triaxial permeability tests and suction probe tests were also performed. The possibility of cementation in the high strength soils was checked using mineralogy, carbonate content and SEM investigations. Bio-stratigraphic analyses including palynological and micro- and nano-fossil examinations were also used to improve geological understanding of the site. No evidence of cementation was found, and it was concluded that the high densities and strengths might have resulted from compaction under the shearing action of advancing and retreating ice sheets. 4 AXIAL PILE CAPACITY 4.1

Base case design approach by MEI

Mustang Engineering Inc. (MEI), the designer, noted API’s reservations on the use of their method for high su /σv values, due to the lack of pile tests in soils with su /σv values greater than three. Quirós et al. (2000) demonstrated that very high past consolida )nc values substantially tion stresses can lead to (su /σv0 below the 0.25 value implicitly assumed in API. MEI therefore adopted the Randolph and Murphy (1985) approach, adding a constraint that the unit friction be limited to Kp σv tan δ. This provided their ‘Baseline’ static capacity for a single pile. MEI assessed the effects of strain softening, cyclic loading, group action and strain rate, estimating the net effect of these considerations to be neutral, allowing their © 2011 by Taylor & Francis Group, LLC

Figure 2. Skew of pile test trends with YSR and L/D; showing Jardine and Chow (1996) data base points and API trends.

‘Operational’ jacket pile capacities to be equated to the ‘Baseline’ static single pile capacities. 4.2

Re-assessment of main text API by the IAT

Since standard methods of assessing axial group interaction and cyclic loading effects might not be appropriate for Clair, and the potential effects of pilot hole drilling also needed considering (Sullivan and Ehlers, 1973), the IAT decided to investigate these issues, along with potentially positive factors such as loading rate effects and skews in the pile API database with YSR and pile L/D ratio. A multiple parameter analysis involving su /σvo , L/D and plasticity index showed that trends like those in and L/D Fig. 2, when extrapolated to the Clair su /σvo ratios, would result in estimated static capacities up to , and L/D ratios kept 38% higher than API. With su /σvo within the API pile load test data-base, static estimated capacities would be around 16% higher than calculated by the standard API approach.

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4.3

MTD-ICP calculations by the IAT

Calibration checks indicated the MTD-ICP approach (Lehane et al., 1994, Jardine and Chow 1996, Jardine et al. 2005) should cope with the extreme YSRs and low L/Ds at Clair, and be extendable to incorporate cyclic and group effects. Specialist laboratory testing provided the data required to run the MTD-ICP method, and also confirmed the tills to be completely insensitive, with high

interface friction angles, δ , of just below 30◦ . However, cyclic simple shear tests showed the tills to be strongly affected by high level cycling. Initially, an upper limit of 8 was imposed on parameter Kc to keep within the data-base of field-tests on instrumented piles. However, Chow’s (1997) database of industrial piles identified cases where the ICP procedure was consistent with higher Kc values, implying that the Kc restriction could be removed. Capacities 10% higher than API were calculated using a Kc limit of 8, with values 44% higher being calculated without this limit being imposed. 4.4

Cyclic effects

The jacket design indicated that the foundations would have significant cyclic loading. MEI’s non-linear inplace jacket analyses provided data to relate cyclic pile head loads to wave heights. Typical storm wave height distributions provided cyclic load build up and decay during the design storm. A programme of cyclic and post-cyclic static simple shear tests was designed to input into the Jardine (1994) and Jardine et al. (2005) approaches for predicting how the normal (radial) effective stresses σn (and hence shear capacity) on the shaft reduce from initial equilibrium values σnc under cyclic loading. Functions were fitted to the sim to N, the number ple shear data that related σn /σnc of cycles applied, and the cyclic shear loading level, expressed as τ/τmax , where τmax is the static shear capacity and τ is the cyclic load amplitude. Conducting the cyclic tests at appropriate normal stresses and keeping volume constant matched the undrained conditions expected close to the pile shaft. The results defined ratios σr /σrc expected after successive wave packages, with a cumulative ‘equivalent number of cycles’ model quantifying the overall effect. The analysis indicated axial capacity could reduce 12 to 14% during the design storm, although shaft friction recovery was considered probable with time, even after severe cyclic loading. 4.5 Strain-softening effects Local brittleness in shaft friction is implicit in the MTD-ICP approach. Peak and ultimate local δ values from laboratory ring-shear tests are used to estimate the effect on axial capacity in ‘falling-branch’ t-z axial load-deflection analyses. The Clair ring-shear tests showed only a 4% difference between peak and ultimate δ values (from 29.5◦ to 28.5◦ ). These were incorporated into the MTD-ICP capacity calculations. Strain softening effects were also assessed by performing API t-z analyses in which the postpeak t-z reduction was based on triaxial shear strength measurements taken to large strain. These indicated post-peak reductions of 7% in the clays to 15 m below seafloor and 2.5% in the underlying clays. The falling branch t-z analyses gave a peak shaft capacity 4.4% lower than that determined by assuming that all points on the shaft reached their local peaks simultaneously. © 2011 by Taylor & Francis Group, LLC

Figure 3. Axial capacity assessments.

The reduction in overall capacity (i.e. including base resistance) was 2.8%. 4.6

Rate effects

Studies by Bea and Audibert (1979) and Tang (1988) imply that the axial capacities of piles driven into clays may be 1.53 to 1.56 times higher under wave loading at 0.1 Hz than their static capacities. Laboratory loading rate tests were not conducted on Clair samples, but comparable positive effects are thought likely to apply. However, the IAT had to ensure the piles could withstand repeated loading cycles during storms without developing a slow creeping failure, which such positive rate effects might not prevent. 4.7

Group effects

The design spacing of the pile groups left less than one pile diameter between some piles. Equivalent pier analyses indicated that pile groups should not fail as a unit under axial loading. However, recent field tests have shown that shear stress fields emanating from closely-centred piles interact negatively, reducing individual pile capacity (Lehane et al. 2003). The Converse-Labarre equation, which gave a good match with the above field tests, indicated an axial group efficiency of approximately 87% for the Clair groups. Simplified theoretical calculations based on overlapping concentric shear stress fields gave a similar 12% reduction in group capacity. 4.8

Summary for pile capacity

Fig. 3 compares the pile capacities obtained by: 1. The MEI ‘Operational’ capacity; 2. The IAT-modified API method including strain softening, database bias, cyclic and group effects;

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3. The IAT MTD-ICP calculations, including postpeak softening effects; 4. The IAT MTD-ICP capacity including cyclic and group effects, at the 29 m target penetration. At 29 m, approaches 1, 2 and 4 led to operational capacities falling within about 10% of each other, giving confidence in the approaches adopted and also in the direct adoption of the MEI operational capacity.

5 5.1

Well drilling and geophysics data indicated that the probability of a 2438 mm o.d. × 76 mm w.t. pile refusing on a large boulder was ∼10%, with refusals most likely in the upper 25 metres. The probability of one pile per group refusing on a boulder was ∼34%, whilst providing one spare slot per group offered a 92% overall probability of a successful installation. However, it was concluded that having drilling and grouting systems offshore would be a more cost-effective remedy for premature refusals. 5.5

Main considerations

Pile collapse

Consideration was given to the possibility that hard inclusions might cause piles to buckle or collapse inwards, possibly without showing initial refusal. The limiting obstruction size was estimated from the dynamic resistance generated when the pile struck the inclusion, following the approach developed to understand collapses seen in the Goodwyn A and Valhall IP installations (Aldridge et al., 2005). The analysis indicated that the pile tip sections should be at least 76 mm thick 440 MPa yield strength steel.

Soil resistance to driving (SRD)

The upper-bound SRD estimates used a conservative interpretation of the soils data, and because of the unusually high CPT values, upper bound SRD estimates included the cone-based ICP method (Lehane et al., 2000), which exceeded the best estimate by an order of magnitude, indicating possible refusal 10 metres below seafloor – not deep enough to provide adequate foundations. Options considered to reduce the uncertainty in SRD and improve confidence in meeting target penetrations included an internal driving shoe, pre-drilled undersized pilot holes, strengthened pile tips and pile driving trials. Steps to help manage residual installation uncertainties also included spare pile slots for additional piles and a backup drilling/grouting spread to allow for intervention in the event of premature refusals. Noting that shaft friction would be reduced by an unpredictable amount by any internal driving shoes, this option was not taken forward.

5.3

Boulders

PILE INSTALLATION

Factors addressed for pile driveability included whether “standard” approaches for hard clays, such as Toolan and Fox (1977) or Stevens et al. (1982) could apply at Clair, the feasibility of driving through cobble layers, the probability of impacting on large boulders, and the potential for pile collapse when driving through such hard soils.

5.2

5.4

Predrilled pilot hole

The option of a 1524 mm pilot hole pre-drilled through the upper very hard clay layers, to approximately 12 m below seafloor, was considered, to reduce friction during driving. The probable effect on long term pile capacity was estimated by referring to model tests on soft clays by Rojas (1993). For the Clair pile geometry and a pre-drilled 1524 mm pilot hole, Rojas’ tests suggested a most probable long term capacity reduction of 17%, with possible loss of 30%. Carefully designed tests at a heavily over-consolidated site would have been required to make a better assessment. It was concluded that pilot hole drilling should be avoided if possible. © 2011 by Taylor & Francis Group, LLC

6

PILE DRIVING TRIALS

A drilling template was to be installed one year before the jacket, with two 1829 mm o.d. × 75 mm w.t. jacket docking piles being driven at the same time. Instrumenting these piles and driving them beyond the required docking pile penetration to the jacket pile target 29 m penetration provided an opportunity to assess pile driving and set-up. One docking pile included an oversized internal driving shoe (80 mm). The MHU 3000 was used for the docking piles to provide the best information for the jacket sleeve pile driving, expected to use the same hammer. A comparison of observed blowcounts with those predicted using the Stevens et al. (1982) approach for both plugged and unplugged driving is presented for shoed and shoeless docking piles in Fig. 4. Docking pile driving was easier than predicted, blowcounts for MHU 3000 hammer efficiencies of 50 and 70% falling near the prediction for a hammer efficiency of 90%. The trend of actual blowcounts for the shoed DP-1 pile followed the predicted lower bound coring case closely to about 24 m penetration, then increased towards the lower bound plugged case prediction. The blowcounts for the shoeless DP-2 pile fell between lower bound coring case predictions for the shoed and shoeless piles, indicating internal friction for this pile not to be as high as expected, and only slightly higher than for the shoed pile. Below about 20-m penetration, DP-2 blowcounts were well below those predicted for the DP-2 coring case. Monitoring data and the dropping blowcounts predicted for the lower bound plugged case between 21 and 27 m

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7

FINAL PILE DESIGN

The final decision was to install 2.591 m diameter pipe piles of 85 mm wall thickness, except for a 90 mm thick, 1.5 m long, driving head and a 95 mm thickened pile sleeve section. The pile tip was made of high strength (440 MPa) steel. The target pile penetration was 29 m, which satisfied both the standard API guidance, and the IAT’s assessment. The design load was assessed as the highest average load expected within any of the four pile groups under the design load condition. All 14 of the jacket’s piles were driven to the same penetration.

8

Figure 4. Docking Pile Predicted and Observed Blowcounts.

penetration indicate either that DP-2 was driving partially plugged throughout this interval, or that friction fatigue was higher than predicted, or possibly a combination of both factors. A review of the data collected during the docking pile installations revealed damping in the very hard Clair clays to be of the order of 0.23 s/m, i.e. similar to “normal” North Sea boulder clays, not at the higher 0.49 s/m value originally adopted. End bearing during driving was close to the static resistance estimates, instead of 1.67 times higher than the estimated static resistance, as originally adopted. The hindcast analyses, which used these revised damping and end bearing factors, indicated a large improvement in predictions if lower residual friction values and greater friction fatigue were adopted, particularly in the very hard clays in the upper 8 metres. Using these adjusted resistances, good lower bound blowcount matches were subsequently achieved for the jacket skirt piles. The marked friction fatigue effects observed during driving are compatible with the cyclic susceptibility observed in the cyclic simple shear testing and the dependency of static capacity on relative pile tip depth, ‘h/R’, that was implicit in the MTD effective stress capacity assessment. Whilst the pile monitoring instrumentation did not function during the re-strike tests, blow-count and hammer energy data indicated increases in shaft resistance (set-up) of 43% and 46% after pauses of only 14.5 and 22 hours. This is consistent with Fugro’s 30 year data-base of driven steel pipe piles in hard North Sea boulder clays (Bhattacharya et al., 2009). It was estimated from this set-up data that shaft frictions would increase by 50% within 24 hours at Clair. © 2011 by Taylor & Francis Group, LLC

OBSERVATIONS DURING JACKET INSTALLATION

Hindcast analyses of the docking pile driving trials allowed a Foundation Acceptance Plan (FAP) to be developed that specified limits to blowcounts, calibrated to hammer performance. Total counts of 2,350 blows, with the MHU 3000 hammer at 85% efficiency, were permitted. Drilling equipment was provided in case hard driving or boulders threatened pile fatigue or premature refusal. All 14 piles drove successfully, although two piles exceeded the 2,350 blow limit by 5 and 15% respectively. However, the hammer was running at less than 85% efficiency and imposed smaller driving stress cycles than anticipated, allowing the fatigue histories to be deemed acceptable. Noting that blowcounts reflect the frictional resistance along the entire shaft, lower bound blowcount rates were applied as the pile tips approached the design penetrations. The ratio of observed to lower bound rates (adjusted for the MHU 3000 hammer energy) was termed the Acceptability Ratio (AR) and values from 1.16 to 1.92 were recorded over the final metre for all 14 piles. Allowing for a minimum of 50% long-term shaft resistance set-up, all piles were expected to meet or exceed the design capacity.

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9

POST-INSTALLATION VERIFICATION

Post-installation pile verification included comparison, for 10 piles, between hindcast Case Method SRDs from the field force-velocity-time data and the SRDs required to match the required capacity. Using the docking pile data, which showed that long term set up of at least 50% could be assumed, the required long term capacity requirement was reached comfortably in all cases. Signal-matching analyses were conducted for 18 selected hammer blows, including one for each of the ten successfully monitored piles as they approached target depths. A Jc value of 0.2 was used in analysing dynamic resistance. In each case the dynamic resistances exceeded the required static capacities.

Conservative projections that assumed that 60% of the SRD friction was on the external face of the pile, at least 50% long-term set up and a fully-plugged static failure mode led to lower bound static capacity estimates that marginally exceeded the required design capacity. This information, along with SRD increases observed during re-strikes after driving pauses, was important to the overall Foundation Assurance, as described by Evans et al. (2010). 10

CONCLUSIONS

This paper has described how the appropriate design axial capacity values for piled foundations were determined and assured in the unprecedented soil conditions encountered under the Clair-1 platform, West of Shetlands. Specific account was taken of the hard till’s very high shear strengths and potential boulder contents, as well as pile group effects, cyclic loading and the possible effects of pilot hole drilling. Advanced testing, analyses, driving trials and installation observations allowed BP and their partners to be confident in the performance of the foundation piles. ACKNOWLEDGEMENTS The authors are grateful to BP North Sea and its coventurers, ConocoPhillips, Chevron, Shell and Hess for permission to publish this paper and share the unique Clair Phase 1 foundation design and installation experiences with others. REFERENCES Aldridge, T.R., Carrington, T.M. and Kee, N.R. 2005. Propagation of pile tip damage during installation. International Symposium on Frontiers in Offshore Geotechnics, 19–21 September 2005. Bea, R.G. and Audibert, J.M.E., 1979. Performance of dynamically load pile foundations. Proceedings 2nd International Conference on Behaviour of Offshore Structures (Boss 79), Imperial College London, pp. 728–745. Bhattacharya, S, Carrington, T and Aldridge, T. (2009) Observed short term set-up of piles in over-consolidated North Sea clays, Proceedings of the Institution of Civil Engineers: Geotechnical Engineering, 162, (pp. 71–80)


Chow, F. (1997). Investigation into displacement pile behaviour for offshore piled foundations. PhD Thesis, University of London (Imperial College). Evans, T.G., Finnie, I., Little, R., Jardine R.J. and Aldridge, T.R., (2010), ‘BP Clair Phase 1 – Geotechnical assurance of driven piled foundations in extremely hard till, Second International Symposium on Frontiers in Offshore Geotechnics, Perth, Australia Jardine, R.J.(1994) Review of offshore pile design for cyclic loading: North Sea clays. HSE Offshore Technology Report, OTN 94 157.85 Jardine, R.J. and Chow, F.C.(1996) New design methods for offshore piles. MTD Publication 96/103, MTD, London. Jardine, R.J., Chow, F.C., Overy, R.F. and Standing, J.R. (2005) ICP design methods for driven piles in sands and clays”. Thomas Telford Ltd, London p. 105. Lehane, B.M., Jardine, R.J., Bond, A.J. and Chow, F.C. (1994) The development of shaft resistance on displacement piles in clay. Proc. XIII ICSMFE, New Delhi, India, pp. 473– 476. Lehane, B.M., Chow, F.C., McCabe, B.A. and Jardine, R.J. (2000) Relationships between shaft capacity of driven piles and CPT end resistance. Geotechnical Engineering, Vol 143, No 2, pp. 93–102. Lehane, B.M., Jardine, R.J and McCabe, B.A (2003) Pile Group Tension Cyclic Loading: Field test programme at Kinnegar, N. Ireland. HSE Research Report 101; HSE Books, p. 42. Matlock, H., 1970. Correlations for design of laterally loaded piles in soft clay. Proceedings 2nd Offshore Technology Conference, Houston, Paper No OTC 1204. Quirs, G.W., Little, R.L. and Garmon, S. (2000). A Normalized Soil Parameter Procedure for Evaluating In-Situ Undrained Shear Strength. Proceedings Offshore Technology Conference Houston OTC12090. Randolph, M.F. and Murphy, B.S., 1985. Shaft capacity of driven piles in clay. Proceedings Annual Offshore Technology Conference, Houston, pp. 371–378. Rojas, E. (1993) Static behaviour of model friction piles. Ground engineering, May, pp. 26–30. Stevens, R.S., Wiltsie, E.A. and Turton, T.H., 1982. Evaluating pile driveability for hard clay, very dense sand and rock. Proceedings Offshore Technology Conference, Houston, Paper OTC 4205. Sullivan, R.A. and Ehlers, C.J, (1973) Planning for driving of offshore piles ASCE, JCD, Vol 99, CO1, pp 59–79 Tang, W.H. (1988) Offshore axial pile design reliability. Research Report for Phase 1 of the Project PRAC 89-29B sponsored by API. Toolan, F.E. and Fox, D.A., 1977. Geotechnical planning of piled foundations for offshore platforms. Proceedings of the Institution of Civil Engineers, 1, pp. 221–244.

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Photoelastic investigation into plugging of open ended piles J. Dijkstra, E.A. Alderlieste & W. Broere Delft University of Technology, Delft, The Netherlands

ABSTRACT: This paper presents the results of model pile load tests in a transparent photoelastic medium. This medium is composed of broken glass particles in an oil with matching refractive index. The setup allows for quantitative photoelastic measurements in the soil. Reliable interpretation of the stress data in the plug proved to be difficult. In the same test setup, displacements around the pile are measured using digital image correlation (DIC). As opposed to the photoelastic measurement method the DIC method was able to capture the soil deformation in the plug. During monotonic jacking of the transparent pile in loose (n0 = 0.446) and dense (n0 = 0.314) initial conditions only an 8% difference in pile head load was observed. The stress and strain distribution on the other hand show significant differences between the loose and the dense test, both in spatial distribution as in magnitude.

1

INTRODUCTION

Open-ended piles as used for deep offshore foundations can be installed more easily compared to closed ended piles at the penetration depth required for the design tension capacity. During installation of such a pile, given the limited cross section of the pile, only a limited amount of soil is pushed aside, deformed and compacted. However, plugging of soil in the pile can occur as a result of the inflow of soil into the pile. If this occurs, only a limited amount of soil enters the pile during further penetration and the pile will behave more like a closed-ended pile during further penetration. During installation the soil properties and stress state around the pile and in the soil plug are altered. These changes are even more pronounced if plugging occurs. This implies that a prediction of the pile bearing capacity or the driving resistance, based on the undistorted soil properties, should incorporate these installation effects in order to be as accurate as possible. However, before these mechanisms can be predicted reliably first the stress and strain evolution in the soil should be studied more carefully. The stress development in a tubular pipe pile is e.g. studied by De Nicola (1996). He placed strain gauges on several levels on the outside of a model pile in order to monitor the shaft friction distribution and the base load. The shaft friction in a pipe pile can be monitored by adding strain gauges on the inside of the pile (see e.g. Lehane & Gavin 2001 for model pile tests and Paik & Salgado 2003 for model and field tests) or by adding a shear force transducer, see Ogawa et al. (2008). These research efforts do not offer full understanding of the soil behaviour during and after installation. Performing full field stress measurements in natural soils is very difficult. For a decent spatial resolution © 2011 by Taylor & Francis Group, LLC

the size and amount of the required sensors embedded in the simply would negatively influence the characterisation of the stress state. The physical size of the sensors prohibits a realistic failure mechanism. Full field soil stress measurements can only practically be obtained with a photoelastic measurement method. This method uses the birefringent properties of the material for the stress characterisation in the material. In the current paper the photoelastic measurement technique is combined with digital image correlation in order to obtain a stress and strain field of the soil near an open ended pile. The paper is an extension on the work of Dijkstra & Broere (2009), as now displacement fields are measured as well as stress fields. The main aim of the paper is to investigate the plugging mechanism at a more fundamental level rather than directly extrapolating the results to design practice. 2

STRESS AND STRAIN MEASUREMENTS

Whilst photoelasticity has been used extensively to quantify stresses in homogeneous materials (e.g. Coker & Filon 1930, Dally & Riley 1991), the technique is not widely applied as an analogon in granular materials. In the current study the granular material, or soil, is replaced by grains of a photoelastic material. Crushed glass particles are immersed in a liquid with a matching refractive index, in order to prevent light scatter in the sample. The liquid in the current tests is Exxon-Mobil Marcol 82. This technique is similar to Allersma (1982), Drescher (1976) and Wakabayashi (1957). The other properties of this matching liquid, such as viscosity and density, are as close as possible to water. In Allersma (1987) and Dijkstra (2009) the similarity in mechanical behaviour between broken glass and sand is shown using triaxial test results. The

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Table 1. Seven polariscope configurations for the photoelastic measurement. intensity I1 I2 I3 I4 I5 I6 I7

Figure 1. Arrangement of optical elements in a polariscope.

crushed glass grains, however, are more angular than sand and as a result the grain properties are slightly different (as are strength and stiffness of the grains, which renders the material more prone to crushing). On the whole, they are a reasonable substitute material to investigate sand behaviour. The broken glass in the current test setup has a grain size d50 = 2 mm. The remaining contrast in the sample, due a small mismatch in refractive index between the grains and the liquid, allows for the use of digital image correlation to capture the displacement fields next to the stress fields in the same test setup.

β (rad) π 4

− π4 π 4

− π4 − π4 − π4 π 4

γ (rad) 0 0

− π4 − π4 − π4 π 4 π 4

θ (rad) 0 0 − π4 − π4 π 2 π 2 π 2

with respect to the horizontal plane of the sample, the fast axis of the first λ4 plate has an angle β, the second λ plate an angle γ and the analyzer an angle θ. The 4 seven configurations are shown in Table 1. From the seven measured light intensities the isoclinic angle φ and retardation δ in the sample can be calculated by (1) – (3), for a full analysis one is referred to Dijkstra (2009):

2.1 Stress measurements The photoelastic effect is measured using a transmission polariscope. This apparatus projects light with a pre-defined polarization state on the sample. The stressed sample will alter the polarization of the incident light and this alteration of the light polarization is subsequently measured. The polarization state is derived from the measured light intensity as observed from seven preset orientations of the quarter wave plates and polarizers in the polariscope. The polariscope used in this research consists of two linear polarizers and two retardation plates (also called λ4 plate) and is shown in Fig. 1. The emerging light intensity can be calculated using Jones calculus (Theocaris & Gdouto 1979). The Jones vector of the incident light is multiplied with the Jones matrix of the optical element. Depending on the rotational position of the optical elements, the angle between the fast axis compared to the horizontal axis, the general form of the Jones matrix for the polarizer or retarder can be simplified. By the proper choice of the type, order and position of the optical elements the photoelastic properties in the sample can be derived from the emergent light intensities for at least four pre-set configurations of the optical elements in the polariscope. The proposed optical arrangement and configuration of Yoneyama & Kikuta (2006) is used for the measurement of the photoelastic parameters in the sample, i.e. the retardation δ and isoclinic angle φ in Fig. 1. This method allows to compensate for poor wave plate performance. Using this phase stepping technique a full polarization state is measured from at least seven configurations in which the polarization of the incident light and positions of the second λ4 plate and analyzer are varied. The current method incorporates the retardation of the non-ideal retarders before and after the sample in the analysis. The fast axis of the first polarizer is set at an angle π2 (counter clockwise) © 2011 by Taylor & Francis Group, LLC

In these equations is the retardation of the retarders. The sign of sin is always taken positive as the retardation needs to be positive. At positions where cannot be obtained for example when I1 = I2 , a representative average value for the domain is taken. Due to the trigonometric functions in (1) – (3) the phase data is still wrapped in the domains − π4 < φ < π4 and −π < δ < π respectively. Physically, for φ this means that the measurement method cannot differentiate between the first and secondary principal stress direction (Dally & Riley 1991). For full field analysis the data, therefore, needs to be further processed before subsequent analysis is possible. In the post processing of the data the phase wrap is corrected using the Goldstein branchcut algorithm Ghiglia&Pritt (1998), as opposed to the Lp algorithm used in earlier work (Dijkstra & Broere 2009). This algorithm is somewhat more timeefficient, whilst offering similar phase-unwrapping performance. Although phase continuity is recovered, sometimes an absolute phase shift is introduced in the data. This absolute shift (multiples of π) is manually corrected in the analysis. After unwrapping the isoclinic angle and retardation data the mechanical stress differences in the sample and the principal stress direction can be obtained. A necessary assumption in this case is that the stress optic law is applicable and that the isoclinic angle, after unwrapping, is equal to the angle of the main principal stress direction. The complete stress state

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still cannot be derived directly from the measurements, as only the stress difference (σ1 − σ2 ) is derived from the measured photoelastic parameters. The values of the principal stresses need to be separated in order to obtain the complete stress state. Therefore, the individual stress components need to be derived using a full field stress-separation method. In the current research the method proposed by Quiroga & González-Cano (1998) is used. In this procedure the equilibrium equations for plane strain are solved with the measured field data as input (stress difference and principal stress direction). The equilibrium equations can be reformulated into the weighted Poisson equation which in turn is minimized using a multigrid solver. In the stress-separation step a material constant is introduced (4).

Figure 2. Schematized plan view of PE model test setup; laser source, plane strain sample, and photoelastic acquisition.

dimensional case, where Ebiot is a 2×2 matrix. The horizontal strain εxx , the vertical strain εyy , the shear strain εxy and volumetric strain εv are given by

where n is the porosity, the thickness d = 21 mm, the stress-optic constant is taken as 2.7·10−12 Pa−1 (Nissle & Babcock 1973). The wavelength λ of the HeNe laser is 632.8 nm. In both the phase-unwrapping and the stressseparation processing a mask is added in the analysis which masks out unreliable measurement points, e.g. the pile wall. 2.2 Strain measurements For the extraction of the displacement data from the recorded images the minimum quadratic difference method (Gui & Merzkirch 1996) is used for the digital image correlation (DIC).

3

where the subscript denotes to the column i and row j on a rectangular grid with M columns and N rows, and f (1) denotes to an intensity reading in the first image and f (2) is an intensity reading in the second image. The location of the minimum value for Cs,t yields the most probable displacement. The method proves to be more robust in case of illumination and noise differences in the subsequent images. Also the method proves to be more reliable with densely packed particles (soil). In the current research a Gaussian sub-pixel fit is used (Mori & Chang 2003). The strains are derived from the displacement fields after correction for rigid body rotations and translations by polar decomposition of the deformation gradient tensor (obtained from the displacements). This yields the stretch tensor U and subsequently the engineering strains from the biot strain tensor E:

The model test setup is designed to measure the stress field in the soil, composed of broken glass with d50 = 2 mm, near the pile by means of the photoelastic method. The displacement fields in the soil near the pile are measured using DIC, whereas the force on the pile head and the surcharge load are measured with load cells. The setup, as sketched in plan view in Fig. 2, can be divided in three parts. The source of the polarized light (laser source with galvano scanner to project the light on the sample), and the camera for the DIC are located at one side of the sample. The light source is offset from the center line of the sample to reduce reflections and glare effects that interfere with the measurements. The sample is contained in a glass and steel strongbox (H = 400 mm, B = 400 mm and S = 21 mm, and pile width W = 21 mm see Fig. 3) and loaded with two surcharges q1 = q2 = 200 kPa. The square tubular pile (outer dimensions 21 mm with 2 mm wall thickness) is centered in the strongbox , in order to prevent splitting the two side walls, a square tubular pile made out of perspex was used. This material is also photoelasticly sensitive, as a result the photoelastic readings in the pile plug are unreliable as the photoelastic effects in the pile’s front and back wall as well as in the soil plug aremeasured. The compressive stress in the pile walls from the load is influencing

where I is the identity tensor. The Biot strain tensor is composed of the engineering strains, for the two © 2011 by Taylor & Francis Group, LLC

MODEL TESTS

3.1 Test setup

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Figure 4. Pile head load during installation.

Figure 3. Sketch of the strongbox.

the stress reading in the plug. In contrast the glass walls of the strong-box are only loaded in bending (by the soil in the strongbox), the stresses from bending in the front glass plate cancel out the stress in the back wall. On the other side of the sample the remaining components of the photoelastic setup (PE) are situated. As seen in the sketch only about half of the sample is within the field of vision of the cameras. The pile is completely in view, but only half of the soil surrounding the pile. This improves the spatial resolution of the DIC and PE analysis, as less surface area is covered with the fixed resolution of the camera. The light intensity measurements are made with a standard DSLR camera (Canon EOS 400D). In order to improve the reliability of the intensity measurements taken by this camera, only the raw image data was used, the initial CMOS offset was corrected by using black image substraction. In order to improve the intensity readings in each stress measurement the seven polariscope positions are repeated five times. In this way the positioning errors from the optical elements and the random errors of the camera sensor are reduced. 3.2 Test results

4

The strain and stress evolution around the tubular pile during monotic penetration problem is studied in loose conditions (n0 = 0.446) and dense conditions (n0 = 0.314). In these tests the maximum displacement was 36 mm with displacement increments of 1 mm, every 5 mm a photoelastic measurement was taken. The measured stress state, in the form of the shear stress σxy and the sum σxx + σyy , after 21 mm of penetration is shown in Fig. 5 for the loose and dense test. These plots show the full pile width, but only one side © 2011 by Taylor & Francis Group, LLC

of the soil surrounding the pile. The total height and width is 4.4 times the pile width. The scale is normalized on the pile dimension. The evolution of the pile head load is separately measured and shown in Fig. 4. For a similar region the shear and volumetric strain εxy and εv , which are the counterparts of the shear stress and the sum are shown in Fig. 6. Only an 8% difference in pile head load was observed. This difference is much smaller than the stress difference found from the photoelastic measurements. Possibly, the pile-strongbox resistance was much higher in the loose test. Reasonable values for the stress are only found below the pile base. The nonexisting measured shaft resistance is presumably due to the low photoelastic sensitivity of the broken glass. The readings in the plug are unrealistic. What can be noticed in the stress results is that in the loose test only one pile wall is supporting the load, whilst in the dense test two pile walls transfer the load. The loose initial conditions resulted in a somewhat inclined penetration. The loose test shows more loosening of the volumetric strains than the dense test, whereas the shear strains are comparable in distribution, but higher in magnitude for the dense test. The distorted zone in the stress results (with high stress magnitudes) is also found in the strain results. Unfortunately, the analysis of the stress evolution is still an ongoing research effort and could therefore not be presented.

CONCLUSIONS

During monotonic jacking of the transparent pile in loose (n0 = 0.446) and dense (n0 = 0.314) initial conditions only an 8% difference in pile head load was observed. This corresponds well with the observed stress level in the assembly below the pile base as obtained from photoelastic measurements. The stress and strain distribution on the other hand show significant differences between the loose and the dense test, both in spatial distribution as in magnitude.

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Figure 5. From top to bottom: σxy;loose , (σxx + σyy )loose , σxy;dense & (σxx + σyy )dense ; loose: n0 = 0.446 (top), dense: n0 = 0.314, compressive stress is negative; W = pilewidth = 21 mm.


Figure 6. From top to bottom: εxy;loose , εv;loose , εxy;dense & εv;dense ; loose: n0 = 0.446, dense: n0 = 0.314, negative volumetric strain is contraction; W = pilewidth = 21 mm.

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ACKNOWLEDGEMENTS This research was financially supported by a grant from the international press-in organisation (IPA, http://www.press-in.org/), which is greatly acknowledged. REFERENCES Allersma, H. (1982). Determination of the stress distribution in assemblies of photoelastic particles. Experimental Mechanics 22(9), 336–341. Allersma, H. (1987). Optical analysis of stress and strain in photoelastic particle assemblies. Ph. D. thesis, Delft University of Technology. Coker, E. & Filon, L. (1930). A Treatise on Photoelasticity. Cambridge: Cambridge University Press. Dally, J. & Riley, W. (1991). Experimental Stress Analysis Third Edition. Singapore: McGRAW-HILL. De Nicola, A. (1996). The Performance of Pipe Piles in Sand. PhD Thesis, The University ofWestern Australia, Perth, Australia. Dijkstra, J. (2009). On the Modelling of Pile Installation. Ph. D. thesis, Delft University of Technology. Dijkstra, J. & Broere, W. (2009). Experimental investigation into plugging of open ended piles. In Proceedings of the ASME 28th International Conference on Ocean, Offshore and Arctic Engineering, Number OMAE2009-79299. Drescher, A. (1976). An experimental investigation of flow rules for granular materials using optically sensitive glass particles. Géotechnique 26(4), 591–601. Ghiglia, D. & Pritt, M. (1998). Two-Dimensional Phase Unwrapping. New York: John Wiley & Sons, Inc.


Gui, L. & Merzkirch, W. (1996). A method of tracking ensembles of particle images. Experiments in Fluids 21(6), 465–468. Lehane, B. & Gavin, K. (2001). Base Resistance of Jacked Pipe Piles in Sand. Journal of Geotechnical and Geoenvironmental Engineering 127(6), 473–480. Mori, N. & Chang, K.-A. (2003). Introduction to mpiv. http://sauron.civil.eng.osaka-cu.ac.jp/ mori/. Nissle, T. & Babcock, C. (1973). Stress-optical coefficient as related to glass composition. Journal of the American Ceramic Society 56(11), 596–598. Ogawa, N., Ishihara, Y., Yokotobi, T., Kinoshita, S., Nagayama, T., Kitamura, A., & Tagaya, K. (2008, Dec). Soil Plug Behaviour of Open-Ended Tubular Pile During Press-In. In Proceedings of 2nd IPA International Workshop, New Orleans, Lousiana, pp. 15–22. International Press-In Assocation. Paik, K. & Salgado, R. (2003). Determination of Bearing Capacity of Open-Ended Piles in Sand. Journal of Geotechnical and Geoenvironmental Engineering 129(1), 46–57. Quiroga, J. & González-Cano, A. (1998). Stress separation from photoelastic data by a multigrid method. Measurement Science and Technology 9(8), 1204–1210. Theocaris, P. & Gdouto, E. (1979). Matrix Theory of Photoelasticity. Berlin: Springer. Wakabayashi, T. (1957). Photoelastic method for determining of stress in powdered mass. In Proceedings of the seventh Japanese National Conference on Applied Mechanics, pp. 153–158. Yoneyama, S.&Kikuta, H. (2006). Phase-stepping photoelasticity by use of retarders with arbitrary retardation. Experimental Mechanics 46(3), 289–296.

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Soil-pile interaction during extrusion of an initially deformed pile C.T. Erbrich, E. Barbosa-Cruz & R. Barbour Advanced Geomechanics, Perth, Australia

ABSTRACT: During installation of a pile into soil, an initial imperfection from the theoretical pure cylindrical geometry can progressively grow with increasing pile penetration when the stiffness of the surrounding soil exceeds the elastic stiffness of the pile. Eventually, even small initial imperfections can develop to the point where plastic yielding of the pile may occur, ultimately leading to total collapse. This paper addresses this problem by means of a specially developed numerical model (BASIL) which is implemented in a Python script that incorporates all the physics of the pile-soil model and which fires a procession of ABAQUS finite element analyses as required. The structural model of the pile is extruded as it penetrates into the soil, which is represented by a series of fully non-linear ‘p-y’ springs. The implemented numerical algorithm allows analyses of stratified soil profiles and incorporates pile tip forces acting on any tip chamfer type. The BASIL model is described in this paper and an example analysis is presented.

1

INTRODUCTION

At the end of 1992, the Goodwyn A (GWA) platform was installed on the North West Shelf of Australia. During installation of the drilled and grouted foundation piles for this platform it was discovered that 15 of the 20 driven ‘primary’ piles were severely crushed at depths of 80 metres or more below the seabed (Fig. 1; Barbour and Erbrich, 1994). In considering possible mechanisms for the GWA collapse, an ‘extrusion’ type of mechanism was conceived whereby an initial imperfection in the pile would have been forced to grow as it was pushed into a soil of higher stiffness than the pile. At the same time, the first and third authors were engaged in the design of bucket foundations in the North Sea and it became apparent that a similar mechanism might also occur during installation of the thin skirts of these foundations. While bucket foundations are much shorter and ‘squatter’ than the GWA piles, they are many times more slender and subject to high inward radial pressures due to the differential water pressures (‘suction’) induced by pumping water from the skirt compartment to aid penetration. To address this issue a soil ‘extrusion’algorithm was devised and a user element (BASIL – Bucket Adjusted Soil Installation Loading) developed to model this behaviour in the ABAQUS finite element code (Barbour and Erbrich, 1995). The BASIL element was used in conjunction with a shell based structural model of the bucket foundation. The model was carefully validated and then used extensively for verification analyses of the skirt installation process on the Europipe 16/11E buckets and for design of the Sleipner SLT bucket foundations. © 2011 by Taylor & Francis Group, LLC

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Figure 1. Goodwyn A primary pile collapse details.

Attempts to model pile geometries with the original BASIL model were unsuccessful. However, the BASIL model has since been extensively reworked and this objective has now been achieved. As an aside it is noteworthy that Earl (2002) presents a finite element model based on the same principles as used in BASIL for modelling extrusion buckling. A series of parametric analyses were performed and presented along with a superficial investigation into the pile collapse at GWA. However, Earl’s model only considered linear elastic behaviour for both the pile and the soil, which are both inappropriate assumptions for this particular problem. In addition, forces acting at the pile tip and on any tip chamfer were ignored and only very simple soil profiles were considered (i.e. no discrete

layering). Aldridge et al. (2005) also considered this problem in a conceptually similar way, but using only a very simplified analytical implementation. 2. 2 2.1

DESCRIPTION OF BASIL MODEL 3.

Original formulation

The original BASIL model is described in Barbour and Erbrich (1995). The model comprised two parts; a structural finite element model (i.e. shell elements) of the foundation and the BASIL user element, which comprised a ‘brush’ of radial ‘hairs’ that radiated from the centre line through the foundation. Each ‘hair’ defined the location of a p-y soil spring, which acted on the structural model. The origin of each spring was defined by the position of the tip of the skirt or pile as it cut each hair. A permanent (plastic) initial imperfection was imposed on the skirt or pile, which was then pushed into the ‘brush’. Any growth of this imperfection was tracked to establish whether this ‘extrusion’ through the soil would lead to structural collapse. Experience using this model revealed a number of significant difficulties, which principally originate from the fact that the model was implemented into ABAQUS as a user element, written in Fortran: – Definition of soil parameters had to be done on a case-by-case basis, with specific Fortran code written for each new case. This was cumbersome and had a high risk of errors. – A stiffness matrix had to be defined for the BASIL element. This was achieved using numerical differencing, which was an approximate and inefficient process that was found to greatly slow the analysis. – Whilst the ABAQUS solver is very effective in many cases, it doesn’t deal particularly well where instabilities develop in a model. Experience showed that after a certain level of imperfection growth had occurred, it would be impossible to progress the analysis any further. – Using a pile sliding past discrete ‘hairs’ tended to set up oscillation patterns in the mobilised soil pressures, which were caused by the faceted nature of the linear shell elements used to model the pile. These oscillations were considered undesirable and unrealistic.

4.

5.

6.

This approach directly solves all the problems identified with the original algorithm:

For the work presented in this paper the model has therefore been completely re-written albeit the underlying physics remains identical. 2.2

New formulation

The ABAQUS user element is abandoned in the new formulation and is replaced with a Python script that incorporates all the physics of the BASIL element, and also fires a procession of ABAQUS analyses as required. The stages in each analysis are: 1. Define the pile geometry, initial imperfection and boundary fixities as per the ‘original formulation’. © 2011 by Taylor & Francis Group, LLC

However, in the new approach the pile is modelled with a uniform (user defined) element length up the pile. For the defined initial imperfection, use the Python script to determine an initial set of nodal forces to apply to the bottom pile element, assuming that this has penetrated fully into the soil. Trigger anABAQUS analysis to determine the compatible pile displacements for the applied set of nodal soil loads and then output these results to the Python script. Use the Python script to determine a new set of soil forces that are compatible with the updated pile displacements and compare these to the soil forces determined in stage 2. If the difference between the computed nodal forces is less than a defined tolerance then the step has completed. If the nodal force difference exceeds the allowable tolerance then trigger another ABAQUS analysis with this revised set of nodal loads and obtain a new set of pile displacements. Repeat stages 4 and 5 until convergence of nodal forces is obtained After convergence is obtained the analysis proceeds to the next step; this requires advancing the pile another element length into the soil and proceeding from stage 2 to stage 5 above. The only difference compared to the first step is that the deformed geometry obtained at the end of the previous step is used as the starting point for assessing the appropriate soil nodal forces for the next step and nodal forces are also defined for the next element up the pile.

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– Any combination of pile or soil can be configured rapidly in a simple input file. – An iterative solution technique is used and hence no stiffness matrix is generated. The BASIL soil model is solved exactly at all points, rather than through an approximate numerical procedure. The iterative solution technique also means that the ABAQUS solver is not relied on for convergence of the BASIL user element. The ABAQUS solver is only used to determine the solution for the relatively simple problem of a pile shell model subject to defined nodal loads. – Since the solution advances in steps that are one element long, there is no longer any sliding past faceted elements and hence oscillations in soil pressure are effectively eliminated. Despite these advantages, implementation of the new approach proved extremely challenging. The main difficulty centred on the iterative solution technique. Many attempts were made to develop a robust iterative solution technique that would guarantee convergence and eventually after much trial and error we found a procedure that has proven highly successful. Amongst other things, this includes random generation of some of the iterative solution variables in order to prevent the solution getting ‘stuck’ in ‘local minima’.

One major advantage of the new formulation is that the problem can be stopped and restarted at any point, and the solver can be changed between different steps. In the example presented later we invoked a dynamic analysis procedure for a substantial part of the analysis, in which the nodal soil forces were applied to the pile over a time period of 0.04 seconds. This enabled the analysis to proceed into uncharted territory, with almost complete collapse of the pile demonstrated. This would have been impossible with the original BASIL user element since this had no dynamic formulation and the original ABAQUS solver would have failed at a much earlier stage due to the intrinsically unstable nature of the solution at such large deformations. It might have been expected that a dynamic procedure would slow down the propagation of tip deformation somewhat since some component of the applied nodal soil force would be balanced by the inertia forces. However, inspection of the rate of tip deformation suggests this to be a negligible effect; the dynamic analysis simply seems to have added enough ‘damping’ into the system to allow the iterative solver to find the correct, essentially static, solution. 2.3

Figure 2. Ramberg-Osgood model for external BASIL spring.

External BASIL soil spring

As discussed above, the BASIL model includes an external soil spring model (‘p-y’), which represents the response of the soil into which the pile is being penetrated. The original BASIL implementation adopted a simple linear elastic perfectly plastic soil spring model. However, the new formulation uses a fully non-linear Ramberg-Osgood (R-O) form of spring, with the stiffness for ‘unloading’paths defined as the initial (linear) stiffness of the R-O model (Fig. 2). As will be demonstrated in the next section this spring type can give a good match to the type of pressure-displacement response anticipated for the soils in this case. Note that the R-O springs only act in compression, being set to zero if they are displaced inward from their initial position. This implies that any effect from the internal plug has been ignored. The influence of the internal soil plug was explicitly evaluated in some cases, but was not found to be a significant contributor to the overall response. 3 VALIDATION OF BASIL SPRINGS The external BASIL soil springs have been determined assuming that cylindrical cavity expansion conditions apply. A series of small strain cylindrical cavity expansion analyses were performed using ABAQUS to validate this assumption and to calibrate the springs used. The first analysis comprised a validation of the FE model using an elastic perfectly plastic (Tresca type) model for the soil domain. The p-y response from the FE analysis was compared with the theoretical solution proposed by Carter, Booker andYeung (1986) and © 2011 by Taylor & Francis Group, LLC

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Figure 3. Normalised shear strength of su /σv0 = 0.5.

Figure 4. p-y = 0.5). (su /σv0

response

from

cavity

expansion

in Houlsby and Carter (1993) and was found to be essentially identical. For the same normalised undrained shear strength of 0.5, a second FE analysis was performed of su /σvo using the generalised stress-strain properties defined on Fig. 3, which were considered more realistic for the soil under consideration. The resulting p-y response is presented on Fig. 4 along with the equivalent p-y response obtained for the simple linear elastic perfectly plastic Tresca model. Despite the fact that the same ultimate strength is adopted in both cases, it may be seen that the latter model gives a significantly softer response at all stress levels.

Figure 5. FE elliptical cavity into circular cavity analysis.

Figure 6. σr for elliptical cavity into circular cavity analysis.

Since a collapsing pile implies a progressively smaller cavity radius to be expanded, we have also investigated the influence of the initial cavity diameter on the final p-y response. Three small strain FE analyses were performed with initial cavity diameters ranging from 0.75 m to 2.65 m using the elasto-plastic Mohr-Coulomb hardening material shown on Fig. 3. From these analyses it was found that the cavity expansion p-y response is independent of the initial cavity diameter provided the wall displacement from the p-y curves is normalised by the initial cavity diameter. Based on these results, the R-O springs used in the BASIL model were therefore defined in a normalised displacement format. The final parametric study investigated the applicability of the cylindrical cavity expansion analysis to the actual pile buckling problem, where the applied soil springs involve deformation of an initial noncircular (elliptical) cavity back into a circular cavity. Two FE analyses employing an elastic perfectly plastic (Tresca) soil domain were performed for this purpose. The first was a small strain FE analysis of cylindrical cavity expansion starting from a nominal initial diameter (D0 ) of 1.325 m (the dashed line on Fig. 5). The second analysis was a large deformation finite element analysis starting from an initial elliptical section (the dotted line on Fig. 5). Both cases were expanded into the circle shown by the solid line on Fig. 5. In the latter case the perimeter of the evolving elliptical cavity section was also kept constant during the analysis, which is also likely to be the case for a collapsing pile. The radial stress (σr ) averaged across the soil element closest to the minor axis for the initial elliptical cavity geometry has been compared to the radial stress for the equivalent element in the initial circular geometry case. For the cylindrical cavity expansion case σr was found to be pretty uniform within the element, but varied significantly between the different gauss points for the ellipse-to-circle case. However, as shown on Fig. 6, the average value of σr for the four gauss points in this latter case exhibits a similar response, as a function of displacement, to that obtained in the cylindrical cavity expansion analyses. This finding supports the assumption made in deriving the BASIL springs that cylindrical cavity expansion conditions may be (approximately) assumed to apply.

4


4.1

EXAMPLE ANALYSIS Structural model

The structural model of the pile has been developed using the ABAQUS S4R element, which is a 4-node doubly curved shell element with reduced integration, hourglass control and a large strain formulation (including finite membrane strains). These are a general purpose shell element using thick shell theory when the shell thickness is large and thin shell theory when the shell thickness is small. Only a quarter of the pile cross-section is modelled and hence 2-fold symmetry is assumed. The pile model used is 50.0 m long with an outer diameter of 2.65 m. Around the circumference of the pile, each element covers an 8.2◦ arc. Vertically, 0.2 m long elements were used over the full length of the pile. A homogeneous pile wall thickness of 45 mm has been adopted. The pile steel has been modelled as elastic-plastic with aYoungs Modulus of 207 GPa and a yield strength of 420 MPa. A post-yield hardening response was also included, equivalent to a modulus of 5 GPa, which is around 2.5% of the elastic modulus, which is considered reasonable for actual steel. The pile is pushed into the soil through displacement boundary conditions applied to the top and hence any applied vertical tip loads act over the entire length of pile modelled. No frictional resistance along the pile shaft was modelled.

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4.2

External soil springs

For the example case considered here, a suite of FE cavity expansion analyses were performed to determine p-y curves for a number of generalised stressstrain curves, similar to those presented on Fig. 3, . A best-fit R-O curve encompassing a range of su /σvo was then determined for each of the resulting ‘p-y’ curves, as in the example shown on Fig. 2, and the various R-O parameters were generalised such that appropriate values could be selected for any given su /σvo . A strong cemented layer was also included in the analysis for which a separate FE analysis was performed to obtain the required R-O parameters.

Figure 7. Ramberg-Osgood parameters.

The R-O parameters adopted in this analysis are summarised on Fig. 7. Figure 8. Chamfer and tip stresses.

4.3 Tip stresses A chamfer at the pile tip has also been included in the example analysis. In theory a BASIL model could be generated that would automatically account for ‘wedging’ behaviour where a pile tip chamfer is penetrated into the soil. However, in practice this is not readily achieved due to the relatively steep inclination of a tip chamfer, as opposed to the much more gradual inclination expected of a pile wall that is undergoing extrusion buckling. Hence to deal with this problem, we separately estimate any pile tip chamfer forces and impose these, combined with the vertical tip loads, as a set of nodal forces at the tip of the pile. These chamfer/tip forces vary with changes in soil layering as the pile penetrates into the soil. The vertical pile tip forces are generally estimated as being equal to the net cone resistance in uncemented soil layers but rather lower in cemented soils, where the driving process tends to propagate fractures ahead of the pile tip. The same vertical tip pressure is assumed to apply over both flat and chamfered tip sections. However, for the latter, the vertical tip force must be transformed into a normal and frictional stress component acting on the chamfer face itself. The frictional component is determined using an interface friction angle, δ. The profiles of chamfer stresses (q1 ) and vertical tip stress (q2 ) used in the example analysis are presented on Fig. 8. 4.4 Initial imperfection The initial imperfection was defined based on the mode shapes derived from an eigenvalue extraction analysis and was used to adjust the initial pile geometry. The first buckling mode shape assuming a uniform inward radial ring load applied at the pile tip was used for the example analysis. The imperfection magnitude is defined as the maximum change in pile radius relative to the nominal pile radius. For the example analysis a 25 mm initial imperfection was defined (ie. 100 mm ovality). © 2011 by Taylor & Francis Group, LLC

Figure 9. Track of pile tip.

4.5 Analysis results A plot of the of the pile tip radius (‘track of pile tip’) is shown on Fig. 9 for the example analysis. Two lines are presented, one representing the minor axis of the pile and one the major axis. The track of the pile tip defines the origin for each of the BASIL springs at any depth. So if the actual pile radius is larger than defined by the track of the pile tip at any depth, a soil pressure will be imposed on the pile. Conversely, where the actual pile radius is smaller than defined by the track of the pile tip, no soil pressure will be mobilised since this means that the pile is moving away from the soil; as noted earlier no internal soil plug resistance is included. It can be seen that after an initial increase in tip displacement at the start of the analysis, a ‘steady state’ condition appears to arise just prior to the pile entering the well cemented calcarenite layer. On entering this hard layer the tip displacement rapidly increases, attaining a maximum of 48 mm inward displacement per metre of penetration at the base of this layer. After exiting the hard layer the tip displacement continues to increase, but initially at a much reduced rate of only about 9 mm per metre of penetration. However, the pile tip displacement never stabilises and rebounds; instead the pile is progressively crushed as it continues to penetrate into the soil. Down to a penetration of about 35 m, the tip displacement rate is relatively constant

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at around 15 mm per metre, but starts to accelerate once again at greater depths. Initially, this acceleration seems to be initiated by the somewhat stronger layer which commences at a depth of about 35 m. However, we believe that the continued acceleration in tip displacement thereafter, even when this stronger layer has been left well behind, is a function of the softening in the structural response of the pile at the very large deformations that have now been imposed. At the end of the analysis, at a pile penetration of 45.6 m, the pile tip displacement has reached a rate of about 70 mm per metre of penetration, the total inward tip displacement is 761 mm and the final pile radius on the minor axis is 539 mm (ie. only 41% of its initial value). The pile deformed geometry, plotted at true scale, is presented on Fig. 10 at the final penetration depth. To aid visualisation this figure is derived using a feature in ABAQUS that allows the ‘full’ model to be plotted based on the stated axes of symmetry. In addition, the yielded and unyielded parts of the pile are shown by differential colour shading. It can be clearly seen that the pile has ovalised into a ‘peanut’ configuration and extensive plastic yielding has occurred along the pile.

processes could be allowed for by adopting a larger initial imperfection than might otherwise be assumed. Alternatively, the tip end bearing could be enhanced, thereby making some allowance for the inertia effects associated with driving the pile tip. Either way, it is recommended that the initial imperfection assumed in such an analysis should always exceed ‘normal’ fabrication tolerances.

5

REFERENCES

Figure 10. Pile deformed geometry at final penetration depth.

CONCLUSIONS

This paper has presented and demonstrated a sophisticated model that can be used to assess the risk of pile extrusion failure during penetration of an open ended pile into the ground. Whilst not commonly recognised, several major pile failures have arisen due to such behaviour in the past; the aforementioned GWA and the Valhall IP platform piles (Alm, et al. 2004) being the most notable examples. Even when complete collapse is not anticipated major problems might arise with only a small degree of ovalisation, if for example, some other tool had to be subsequently passed down the pile (e.g. if drilling out the pile). It is recommended that this type of analysis should be performed as a matter of course in such cases. When assessing the appropriate magnitude of initial imperfection for any given case it should be appreciated that the analysis does not explicitly account for dynamic pile driving effects; essentially the pile is steadily jacked into the ground. Model tests have indicated that driving rather than jacking is a more damaging process. Earl (2002) proposed that such


Aldridge, TR, Carrington, TM, & Kee, NR. Propagation of Pile Tip Damage during Installation, Proc. International Symposium on Frontiers in Offshore Geotechnics, Perth, September 19–21, 2005. Proceedings. Eds: Gourvenec, S, Cassidy, M. Alm, T., Snell, RO., Hampson, K., and Olaussen, A. (2004). Design and Installation of the Valhall Piggyback Structures, Proc. Offshore Technology Conference, OTC16294, Houston. Barbour, R.J. & Erbrich C. (1994). Analysis of In-situ Reformation of Flattened Large Diameter Foundation Piles Using ABAQUS, UK ABAQUS Users Conference, Oxford, September 1994. Barbour, R.J. & Erbrich C. 1995. Analysis of Soil Skirt Interaction during Installation of Bucket Foundations Using ABAQUS, Proc. ABAQUS Users Conference, Paris, June 1995. Carter, J.P., Booker, J.R. and Yeung, S.K. (1986). Cavity Expansion in Cohesive Frictional Soils, Geotechnique, 36(3): 349–358. Earl R.J. 2002. Growth of Imperfections in Piles During Installation, PhD Thesis, University of Western Australia. Houlsby GT and Carter JP (1993). The Effects of Pressuremeter Geometry on the Results of Tests in Clay, Geotechnique 43(4), 567–576.

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BP Clair phase 1 – Geotechnical assurance of driven piled foundations in extremely hard till T.G. Evans BP Exploration Operating Company

I. Finnie Advanced Geomechanics (Formerly of Lloyds Register)

R. Little Fugro GeoConsulting, Inc. (Formerly of Fugro-McClelland)

R.J. Jardine Imperial College, London

T.R. Aldridge Fugro GeoConsulting Limited

ABSTRACT: BP’s Clair Phase 1 Platform is the first fixed structure on UK’s Atlantic Margin. The conventional steel platform is in 140m of water and is supported on groups of steel pipe piles driven into bouldery glacial clays with undrained shear strengths of up to 2000 kPa. These extreme conditions posed significant challenges for foundation engineering that were managed by a systematic process of design, design assurance and performance monitoring. This paper summarises the project history and foundation risk management process. The foundation engineering is described in more detail by Aldridge et al. (2010).

1

INTRODUCTION

The Clair field is about 75 km off the west coast of the Shetland Islands, UK Sector, North Sea (Fig. 1). Discovered in 1977, it is one of the largest hydrocarbon resources off northwest Europe, with 6 to 7 billion barrels of oil equivalent (bnboe) of reserves. Recoverable oil is limited by a fractured reservoir and low API gravity fluids, so development was not considered feasible until the 1990s, when advances in seismic imaging and engineering technologies offered economic production.

BP (with a 28.6% stake) and its co-venturers, ConocoPhillips (24.0%), Chevron (19.4%), Shell (18.7%) and Hess (9.3%) are developing the field in phases, the first of which extended from 1996 to 2005, as shown on Fig. 2. The second phase, Clair Ridge, is in the conceptual design stage and is expected to come on stream in late 2014. Clair Phase 1 comprised a single fixed steel drilling and production platform in about 140 m of water. Oil is exported to the Sullom Voe terminal in the Shetland Islands through a 22-in, 105 km long pipeline. The Clair Ridge development is about 6 km to the northwest of Clair Phase 1 and will consist of

Figure 1. Location of Clair Phase 1 Platform.

Figure 2. Nine year development of Clair Phase 1.


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Figure 4. Genesis of clair soils.

wall thickness pipe piles that were pre-installed with the template in summer 2003. The docking piles and 14 platform piles were driven to a depth of 29 m below the seabed using a Menck MHU 3000 underwater hydraulic pile driving hammer.

Figure 3. Clair jacket foundation layout.

two bridge-linked steel platforms – one for drilling and production, the other for quarters and facilities. The Clair field soils consist principally of hard clay tills, with gravel to boulder-size rock inclusions. The undrained shear strengths of the clay tills exceed those at all previous offshore sites where steel pipe piles had been driven successfully. This paper summarises the approach taken by BP to design and assure driven pile foundations for the Clair Phase 1 Platform in these exceptional conditions. It focuses on the project history, site characterisation and design and assurance process. The detailed engineering is covered by a companion paper to this Symposium (Aldridge et al., 2010).

2

CLAIR PHASE 1 PLATFORM

The Clair Phase 1 platform is the first fixed structure on the UK Atlantic Margin and is located about 100 km northeast of BP’s FPSO-based developments at Foinaven and Schiehallion (Fig. 1). The structure comprises an 8,800 t four-leg steel jacket with an 11,700 t integrated topsides deck. The jacket is founded on 14 No 2.59 m (102 in) diameter, 85–95 mm wall thickness open-ended steel pipe piles. The platform is designed for 100 year directional storm loads and a 25 year life. The foundations are asymmetrical, with each of the legs on the eastern side of the jacket supported on four skirt piles and the western legs each carried by three skirt piles (Fig. 3). The jacket and topsides were designed by Mustang Engineering Inc. (MEI) and fabricated at Aker Kvaerner’s Verdal yard in Norway. Fugro-McClelland Marine Geoscience (FMMG) was a specialist consultant to MEI for foundation design. The facilities were installed by Saipem UK Ltd. in late 2004 and early 2005 using their S7000 heavy lift vessel. The jacket was docked over a 28 slot drilling template using two 1.829 m diameter 75 mm © 2011 by Taylor & Francis Group, LLC

3 3.1

GEOLOGY Regional

The Clair field lies on the West Shetland Continental Shelf which was glaciated repeatedly during the mid to late Pleistocene Epoch. The platform is in the terminal moraine zones of two coalescing Weichselian-age ice sheets, as indicated on Fig. 1. The geological knowledge of the area pre-1997 was largely based on regional studies by the British Geological Survey (BGS) and exploration-quality seismic data and data from drilling hazard geophysical site surveys. Desk studies in 1997 indicated the depositional environment to be very complex, with the shallow geological conditions varying laterally and vertically. The near-surface soils of interest for platform foundations were inferred as lodgement tills and inter-morainal sediments that have been compressed and sheared by hundreds of metres of ice (Fig. 4). The soils were predicted to be very dense and hard and to contain cobbles and boulders. This prognosis was generally supported by the large number of anchor-dragging incidents and well top-hole re-spuds reported in the Clair field. 3.2

Field-specific

The first engineering-quality geophysical survey and intrusive geotechnical investigations were carried out in April and June 1997, respectively. The work was performed around the preferred platform location at the time, Site A. The main purpose of the geophysical survey was to help develop a regional stratigraphic model and to detect shallow geohazards, including boulders. The techniques used included side scan sonar for imaging the seabed, a Deep Tow Sparker (DTS) for shallow subsurface imaging ( 400/300, due to its excessive deflection (>50 mm or 1 pile diameter). The deflection profiles show rotation, or rotationtranslation of the pile as the Lm /Ls increases. The deflection attains maximum at the surface, and increases with Lm /Ls ratio. Starting at Lm /Ls = 500/200, the pile also begins to translate through the stable layer (Ls ), as the soil movement increases (the pile deflection exhibits an initial rotation, and then rotation-translation at a higher ws ). Figure 6 shows the Mmax against Smax for the FLAC3D models having Lm /Ls of 100/600 to 600/100 (see Table 3). Each model shows a linear correlation between the Mmax and Smax for any magnitude of soil movement, as is noted in model pile tests (Guo & Qin 2010). © 2011 by Taylor & Francis Group, LLC

Figure 5. The pile response at different Lm /Ls ratios.

Figure 6. The relationship between Mmax and Smax (ws = 30∼120 mm)

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The gradient of each line (model), α, is obtained and provided in Table 3. The following are noted: • The α generally reduces with the increase in Lm /Ls

until Lm /Ls = 400/300;

• The α of 0.14 (line 5) and 0.25 (line 6) are correct

for the Lm /Ls of 500/200 and 600/100 respectively,

Table 3.

Gradient α of Mmax against Smax relationship.

Lm /Ls

α

Line

100/600 200/500 300/400 400/300 500/200* 600/100*

0.17 0.19 0.16 0.14 0.14 0.25

1 2 3 4 5 6

layers, independent of soil movement level; and (3)The ratio α for the investigated Lm /Ls ratios is 0.17 ∼ 0.25. ACKNOWLEDGEMENTS The work reported was supported by Australian Research Council (DP0209027) and Griffith University School of Engineering.These financial assistances are gratefully acknowledged.

Note: *Both S max and M max are negative

REFERENCES as negative magnitude of both the M−max and the S−max are noted (see Figure 5(a) for the bending moment profile). • The lines 1 to 4 rotate around the origin in an anticlockwise direction, as Lm /Ls increases. This reflects progressive change in the pile movement mode (discussed previously), as Lm /Ls increases. • The ratio α (= Mmax /Smax ) is 0.14∼0.25 from all the FLAC3D models. This range of values are consistent with 0.13∼0.28 obtained theoretically and experimentally by Guo and Qin (2010), regardless of magnitudes of soil movements. 5

CONCLUSIONS

FLAC3D analysis was conducted regarding the model pile tests subjected to lateral soil movement. The predictions show some difficulty in modeling the magnitude and the profile of the measured pile response. However, the ratio α of maximum bending moment Mmax over shear force Smax induced in each pile is well simulated. The FLAC3D analysis shows that: 1) the bending moment and pile deflection profiles change with the increase in Lm /Ls ratios; 2) The ratio α is unique for each model in the stable and moving soil


American Petroleum Institute. 2000. Recommended practice for planning, designing and construction fixed offshore platforms-working stress design, API RP 2A-WSD. Bhattacharya, S. 2003. Pile stability during earthquake liquefaction. PhD Thesis, University of Cambridge. Chae, K. S., Ugai, K. and Wakai, A. 2004. Lateral resistance of short piles and pile groups located near slopes. International Journal of Geomechanics, Vol. 4, No. 2, pp. 93–103. Ghee, E. H. 2009. The behaviour of axially loaded piles subjected to lateral soil movements. PhD Thesis, Griffith University. Guo, W. D. and Ghee, E. H. 2004. Model tests on single piles in sand subjected to lateral soil movement. Proceedings of 18th Australasian Conference on the Mechanics of Structures and Materials, Perth, Vol. 2, pp.997–1004. Guo, W. D. and Qin, H. Q. 2010. Thrust and bending moment for rigid piles subjected to moving soil. Canadian Geotechnical Journal, Vol. 47, No. 1, pp. 180–196. Guo, W. D., Qin, H. Q. and Ghee, E. H. 2006. Effect of soil movement profiles on vertically loaded single piles. International Conference in Physical Modelling in Geotechnics, Hong Kong, pp. 841–846. Itasca. 2002. FLAC3D version 2.1. Fast Lagrangian analysis of continua in three dimensions manual, Itasca Consulting Group, Inc., Minneapolis. Jaky, J. 1944. The coefficient of earth pressure at rest. Journal of the Society of Hungarian Architects and Engineers, pp. 355–358.

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Cyclic loading of barrettes in soft calcareous rock using Osterberg cells C.M. Haberfield, D.R. Paul, M.C. Ervin & G.A. Chapman Golder Associates, Melbourne, Australia

ABSTRACT: There remains considerable uncertainty with respect to the cyclic performance of drilled and grouted piles founded in soft calcareous rock. Static and cyclic load testing of three test barrettes founded in a weak carbonate siltstone has been carried out using Osterberg cells. The load test results indicate the shaft resistance performance depends on load and displacement history, and that loss of shaft resistance performance under load reversal may not occur unless the ultimate shaft resistance is achieved during load cycling. 1

INTRODUCTION

There remains considerable uncertainty with respect to the cyclic performance of drilled and grouted piles founded in soft calcareous rock. Recent static load testing of 2 No. 65 m and 1 No. 95 m long test barrettes with cross-sectional dimensions of 1.2 m × 2.8 m for the Nakheel Tower project in Dubai provided valuable data on the performance of cast-in-place deep foundations in calcareous sediments under cyclic load. The testing was carried out using two levels of Osterberg cells placed in the bottom 20 m of the barrettes. As part of the testing programme, an 8 m long section of the shaft of the test barrettes (between the two levels of Osterberg cells) was subjected to cyclic loading. This paper describes the ground conditions and properties of the soft calcareous rock, the construction and testing of the barrettes and the measured performance of the barrettes when subjected to static and cyclic loading.

material is generally massive with no significant joints or discontinuities. A conformable sedimentary sequence (Unit D) lies below Unit C and extends to greater than 200 m depth. This material comprises calcareous siltstone or calcisiltite (depending on carbonate content) and is characterised by layers (upto 3.5 m thick) and nodules (cobble size) of gypsum. The thicker gypsum layers could be correlated between boreholes and suggest a general dip within this material of about 8◦ . UCS testing of the calcareous siltstone and calcisiltite generally indicated strengths between about 1 MPa and 5 MPa. Bulk density varies between 1.8 t/m3 and 2.2 t/m3 , void ratio between 0.5 and 0.7 and carbonate content between 50% and 70%. The gypsum is stronger with unconfined compressive strength varying between about 5 MPa and 15 MPa. Unit D material is also massive with no significant joints or discontinuities. 3

2

GEOLOGY AND SUBSURFACE STRATIGRAPHY

The material in which the test barrettes are located is of Quaternary age, comprising shallow marine sediments deposited as the sea level fluctuated during the onset and decline of ice ages. Recent aeolian deposits form a 20 m thick capping over the site. The ground water is highly saline and ground water level is at a depth of about 2.5 m below ground surface level. A unit comprising predominantly variably cemented, calcisiltite (Unit C) of very low to low rock strength underlies the recent aeolian deposits and extends to a depth of about 72.5 m below ground surface level. Unconfined compressive strength (UCS) tests on samples recovered from high quality coring generally range between 0.5 MPa and 4 MPa. Bulk density varies between 1.6 t/m3 and 2.0 t/m3 , void ratio between 0.6 and 0.8, carbonate content between 40% and 80% and hydraulic conductivity is about 10−7 m/s. The Unit C © 2011 by Taylor & Francis Group, LLC

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CONSTITUTIVE BEHAVIOUR

Specialist laboratory testing was undertaken to better understand the constitutive behaviour of the Unit C and D materials. Testing comprised cyclic and monotonic constant normal stiffness direct shear testing on concrete/rock interfaces, resonant column testing, drained triaxial testing, cyclic triaxial testing and high pressure oedometer testing. An extensive programme of insitu testing comprising pressuremeter testing, cross hole seismic testing and water pressure testing was also undertaken. As indicated below, from our observations of the core sample and comparison of the insitu and laboratory test results, it became apparent that, when sampled and brought to the surface, the Unit C and D materials underwent significant stress relief. This resulted in samples tested in the laboratory showing significantly lower strength and deformation properties than measured by insitu testing. Significant emphasis was therefore placed on the results of the insitu testing.

Nevertheless, the laboratory tests provided useful insights into the constitutive behaviour of the Unit C and D materials. Unit C and D materials have a relatively high stiffness below a “bond yield strength” after which the compressibility of the material increases significantly and exhibits properties similar to an uncemented, normally consolidated material at the same void ratio. Prior to reaching the bond yield strength the rock displays approximately linear elastic behaviour with deformations occurring essentially instantaneously. As the bond yield strength is approached, deformations become time dependent and consolidation and creep displacements dominate. Strength testing indicates the behaviour of the rock is dominated by intergranular cementation with little apparent frictional component to strength. The Tresca yield criterion was found to provide a reasonable basis for modeling the behavior of the Unit C and D materials up to bond yield strength. 4

ENGINEERING PROPERTIES

Figure 1 compares the Young’s modulus values estimated from the pressuremeter (initial loading modulus), cross hole seismic and laboratory UCS tests. The pressuremeter test results displayed similar initial loading and unload-reload moduli values which is consistent with the absence of jointing in the rock and the domination of the cementation. The Young’s modulus values obtained from the pressuremeter and cross hole seismic tests show reasonable agreement (see Figure 1) if the small strain modulus values obtained in the cross hole seismic tests are reduced by a factor of five (which is consistent with published data on a range of soil and rock types which compare modulus at different strain levels). Figure 2 compares the shear strengths measured in the UCS tests (taken as UCS/2) and those estimated from the pressuremeter tests assuming a purely cohesive strength criterion (Tresca criterion). We note that the use of the Tresca criteria for rock would be considered unusual as it assumes no frictional component to strength. However, for this material it would appear to be reasonable on the basis of the constitutive behavior observed in the laboratory tests (including drained triaxial tests) where pre-peak strength behavior was dominated by cementation rather than friction). The Tresca criterion also provided a very good fit to the pressuremeter curves up to significant strain levels (8% cavity strain) and was adopted for all subsequent modeling of the test barrettes (see below). Figures 1 and 2 show that stiffness and strength properties measured in the laboratory were significantly less than obtained from insitu tests, and supported a hypothesis that the core samples were undergoing significant stress relief even with the care that was undertaken during the drilling, retrieval, storage, transportation and testing processes. The load testing carried out on the test barrettes (see below) confirmed that the properties obtained from the insitu testing were reasonable and that the © 2011 by Taylor & Francis Group, LLC

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Figure 1. Young’s modulus variation with elevation (surface level RL + 2.5 m DMD).

Figure 2. Shear strength variation with elevation (surface level RL + 2.5 m DM).

laboratory test results significantly under-estimated the properties of the insitu rock. Constant normal direct shear testing was undertaken on concrete/rock interfaces using a range of interface roughnesses (including smooth), normal stiffnesses and initial normal stresses. The results of the tests indicate a residual interface friction angle of about 37◦ .

5 TEST BARRETTES Three test barrettes with cross-sectional dimensions of 1.2 m × 2.8 m were installed to depths of 65 m (TB02 and TB03) and 95 m (TB01) and tested in accordance with specifications provided by Golder Associates Pty Ltd. Test barrette TB02 was installed at the same location as the investigation borehole BH208. Test barrette TB01 was installed about 12 m south east of TB02 and TB03 about 8 m due south of TB02 resulting in a minimum clear distance between test barrettes of about 6 m. The lengths of the barrettes were chosen to provide information on barrette performance in the Unit C and D materials. The test barrettes were installed by a SoletancheBachy/Intrafor Joint Venture using hydrofraise equipment with polymer support. The hydrofraise cutting action results in a relatively smooth excavated surface and hence a concrete rock interface which is essentially devoid of roughness. High slump concrete was placed by tremie. Concrete design characteristic 28 day strength was 60 MPa. Load testing of the barrettes was carried out by Loadtest International Inc under the direction of Golder Associates Pty Ltd. The load tests comprised two levels of Osterberg cells in each test barrette as shown in Figure 3. The Osterberg cells were positioned to measure performance of the lower 20 m or so of the barrettes. The test barrettes were instrumented with displacement tell-tales and strain gauges. In addition, instrumentation was also located in the rock below the toe of the barrettes to directly measure the displacement of the rock at this location. 6

STATIC TEST RESULTS

The measured load versus displacement performance of the two shorter test barrettes (TB02 and TB03) for loading at the lower (LOC) and upper (UOC) levels of Osterberg cells are shown in Figures 4 and 5 respectively. Also shown are predictions of the performance. The predictions were obtained on the basis of the design strength and deformation properties for the site (residual friction angle from CNS tests, strength and deformation properties from pressuremeter tests) and on the as-constructed barrette geometry. The predictions of performance were completed prior to testing of the barrettes. For the Class A prediction, the rock-socket software ROCKET97 (Seidel, 2000) was used to calculate the shaft resistance versus deformation response of the relatively smooth, short test sections of the test barrettes. These analyses assumed a relatively smooth barretterock interface. The shaft resistance versus deformation response calculated in the ROCKET97 analyses were then adopted as the barrette-rock interface behavior in an axisymmetric PLAXIS V8 non-linear finite element model to calculate the load versus displacement responses shown in Figures 4 and 5. The comparison between the measured and predicted response is excellent, which provided further confidence that the design © 2011 by Taylor & Francis Group, LLC

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Figure 3. Shear strength variation with elevation (surface level RL + 2.5 m DMD).

strength and stiffness properties adopted on the basis of the insitu pressuremeter testing were appropriate. Similar comparisons between measured and predicted results were obtained for the deeper test barrette TB01.

7

CYCLIC TEST RESULTS

Following the initial static load testing, both the upper and lower Osterberg cells were sufficiently “open” to allow cyclic testing of the shaft between the upper and lower levels of Osterberg cells. This was undertaken, for example, by pressurising the UOC, while allowing the LOC to bleed off any pressure caused by the shaft between the UOC and LOC moving downwards. Once the nominated downward displacement had been achieved, the process was reversed by depressurising the UOC then pressurising the LOC while allowing the UOC to bleed off any pressure caused by the upwards movement of the shaft between the UOC and LOC. This process was repeated a number of times to investigate the cyclic loading behaviour of the shaft. The initial loading of test barrettes TB01 and TB02 was undertaken by increasing the load in the UOC, whereas the initial loading for test barrette TB03 was undertaken by increasing the load in the LOC. This has implications for the cyclic loading results set out

Figure 6. TB01 – Average shaft resistance versus displacement response for barrette shaft between UOC and LOC. Figure 4. Measured vs predicted performance for loading at upper Osterberg cells.

The average displacement is the average of the measured displacements at the bottom plate of the UOC and top plate of the LOC. Annotations shown on these figures describe various aspects of the tests and shaft resistance behaviour. The cyclic load testing also incorporated significant hold stages during which the load in the Osterberg cells was kept constant. No significant creep displacement was observed during these hold stages. The results for the deepest test barrette TB01 shown in Figure 6 indicate essentially elastic behaviour with a maximum mobilised shaft resistance obtained in the test (for both upwards and downwards loading) of about 1250 kPa. We note the ultimate shaft resistance was not achieved in either upwards or downwards loading and that the response is very stiff with a maximum displacement of less than 3.5 mm at maximum load. The results for TB02 shown in Figure 7 indicate an ultimate average shaft resistance for downward loading of about 550 kPa. At some stages of the test, higher apparent average shaft resistances were measured. However, these were due to contributions from elsewhere along the shaft and should be ignored. Figure 7 also indicates the following behaviour: Figure 5. Measured vs predicated performance for loading at lower Osterberg cells.

below. The average shaft resistance versus displacement performance of the barrette shaft between the UOC and LOC are shown in Figures 6, 7 and 8 for TB01, TB02 and TB03 respectively. The average shaft resistance was calculated by dividing the load difference between UOC and LOC by the perimeter area of the shaft between UOC and LOC. © 2011 by Taylor & Francis Group, LLC

1. A significant reduction (from about 550 kPa to about 250 kPa) in ultimate shaft resistance on complete shear reversal from downwards loading to upwards loading. We note that this was not observed with TB01 and may suggest that the reduction in shaft resistance under tension loading may only occur if the ultimate shaft resistance is exceeded. 2. No reduction in shaft resistance performance during cyclic loading for downwards loading provided the displacement on reversal (i.e. in upwards

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Figure 7. TB02 – Average shaft resistance versus displacement response for barrette shaft between UOC and LOC.

Figure 8. TB03 – Average shaft resistance versus displacement response for barrette shaft between UOC and LOC.

loading) is less than required to achieve peak shaft resistance (for upwards loading). 3. No significant reduction in ultimate shaft resistance with displacement under downwards loading. 4. Shaft resistance versus displacement behavior during cycling appears to depend on the magnitude of post-peak displacement experienced during upwards loading. 5. Following cycling, full shaft resistance is achieved with further downwards displacement. © 2011 by Taylor & Francis Group, LLC

6. The stiffness of the unloading response (whether in upwards or downwards loading) is relatively constant across all loading cycles.

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The results for TB03 shown in Figure 8 are reasonably consistent with those for TB02 and indicate an ultimate average shaft resistance of 600 kPa for downward loading and about 300 kPa to about 100 kPa for upward loading. Note that during some cycles higher (and increasing) negative shaft resistances were

measured and these were due to errors in the testing procedure which allowed resistance contributions from other parts of the barrette. Comparison of Figures 7 (TB02) and 8 (TB03) indicates that the shaft resistance in TB03 is mobilised at significantly lower displacement than for TB02. The ultimate shaft resistance measured in TB03 is also slightly higher than for TB02. This is consistent with other data obtained from the tests (e.g. base resistance) and indicates the ground at TB03 may be stronger and stiffer than at TB02. The results for TB02 and TB03 appear to indicate a significant loss in shaft resistance on reversal of loading. However, it is interesting to note that loading of the test section for pile TB01 and TB02 initially occurred by pushing the test section downwards, whereas TB03 the test section was pushed upwards. Ultimate shaft resistance was not achieved during testing of TB01 and there was no loss in shaft resistance on reversal of load. For TB02, ultimate shaft resistance was achieved during initial downwards loading and on load reversal a significant reduction in ultimate shaft resistance was observed for upwards loading. For TB03, ultimate shaft resistance was not achieved in initial upwards loading, but on load reversal, ultimate shaft resistance (about 600 kPa) was achieved in downwards loading. On load reversal again to upwards loading, a significant reduction in ultimate shaft resistance was observed. It would appear that the reduction in ultimate shaft resistance may only occur if ultimate shaft resistance is achieved prior to load reversal. The measured values of ultimate shaft resistance (downwards loading) for the test barrettes TB02 (550 kPa) and TB03 (600 kPa) are reasonably consistent with those estimated using the estimated pressure of the fluid concrete and the residual friction angle of the concrete – rock interface (580 kPa). This is probably to be expected due to the relatively smooth barrette-rock interfaces formed using the hydrofraise equipment. The measured ultimate shaft resistance for TB01 (1250 kPa) is higher than estimated using this


simplistic approach (1000 kPa) and this may be potentially explained by greater roughness of the interface within the gypsum layer and the higher strength of the gypsum compared to the calcisiltite.

8

CONCLUSIONS

The results of load tests on three test barrettes in calcareous siltstone indicates that for the barrettes tested, it would appear that: 1. a reasonable estimate of ultimate shaft resistance for smooth barrette-rock interfaces can be made from the pressure applied by the fluid concrete and the residual friction angle of the concrete-rock interface. 2. the ultimate shaft resistance measured on load reversal (following loading to ultimate shaft resistance) is significantly less than measured under static loading. 3. this reduction in ultimate shaft resistance may only occur if ultimate shaft resistance is achieved prior to load reversal. 4. the stiffness of the unloading response (whether in upwards or downwards loading) is relatively constant across all loading cycles.

ACKNOWLEDGEMENTS The authors gratefully acknowledge the assistance of Nakheel, Soletanche-Bachy/Intrafor Joint Venture and LoadTest International; and Foundation QA for the use of Rocket.

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REFERENCE Seidel J.P. (2000) ROCKET97 Help Manual. Department of Civil Engineering, Monash University


Shaft capacity of drilled and grouted piles in calcareous sandstone B.M. Lehane University of Western Australia & Consultant to Arup Australasia

ABSTRACT: The paper addresses the significant shortage of full scale load test data for drilled and grouted (D&G) piles in weak calcareous rocks by presenting results from an instrumented pile test programme that involved tension testing of 240 mm, 340 mm and 450 mm diameter D&G piles at a calcareous sandstone site north of Perth, Australia. It is shown that the capacities predicted using a variety of approaches employed by local practitioners are often significantly larger than measured static capacities, and that these approaches themselves yield a wide range of capacities. The testing programme highlighted the considerable uncertainties associated with the prediction of both static and cyclic pile response in variable soft rock deposits such as the Tamala Limestone. 1

INTRODUCTION

Drilled and grouted (D&G) piles are presently the preferred pile type to resist axial loads in the coastal limestones and calcareous soils present along the west coast of Western Australia. There is, however, a great shortage of reported field load tests on such piles and consequently a diverse range of design approaches are in current use. Motivated by the need for actual field data to test the validity of various approaches, Arup International, supported by Belpile Pty. Ltd., commissioned three static tension load tests (followed by cyclic testing) on instrumented D&G piles installed in a calcareous sandstone. This paper presents the results of these tests and compares the observed capacities with Class A predictions made by local practitioners. 2

SITE LOCATION

The site selection process was constrained by the need to be within easy reach of Perth city centre (thereby reducing costs associated with piling). The selected site was a limestone quarry in Pinjar, which is about 25 km north of Perth. The rock in this quarry was described by its owner as a “medium grade limestone”, which forms part of the Tamala Limestone formation. Piles were installed in the centre of the quarry, which had been excavated some years ago to a depth of about 10 m below the surrounding ground level. 3

GROUND CONDITIONS

Cone Penetration Tests (CPTs) were conducted in advance of the piling and included tests within 1m of each of the pile test locations. The CPT end resistance (qc ) profiles at these locations are shown on Figure 1 and indicate qc values typically varying from about © 2011 by Taylor & Francis Group, LLC

Figure 1. CPT qc profiles in the vicinity of the test piles.

15 MPa to 50 MPa; some horizons with low qc values ( 1000 for the SLT (Hölscher et al., 2008). Although the SLT is the most reliable method, it is often too expensive and time consuming to apply routinely. The RLT is increasingly used because it is better in terms of execution, elaboration and quality assurance than the DLT (Middendorp et al., 1992) and is more suitable for use in offshore foundation engineering than the SLT. Open-end piles generally behave as though fully plugged during static loading but they can behave in a partially plugged way during rapid or dynamic loading, especially when loading rates are high (Bruno & Randolph, 1999). The degree of plugging depends not only on the loading rate but also on the type of soil. Different degrees of plugging are expected to result in different levels of soil resistance. An understanding of plugging during an RLT is important for the application of RLTs to open-end piles: if a pile plugs during an SLT but does not plug during an RLT, the RLT will be unreliable and may underestimate pile capacity. Scale modelling pile load tests offers a good possibility for this research. It avoids the high costs of field testing and offers additional possibilities compared © 2011 by Taylor & Francis Group, LLC

with field testing. Centrifuge modelling is considered to be a reliable method due to the accurate representation of the stress state, especially the self-weight stress gradient, around and inside the model pile at a reduced scale. An experimental study of RLTs and SLTs with open-end piles was performed with different soil types to examine plugging behaviour in silt and sand, especially during RLTs, and to compare soil resistance in rapid and static conditions. This paper presents the results from four test series comprising several RLTs and SLTs.

2 2.1

DESCRIPTION OF RESEARCH Centrifuge modelling

Given the requirement of stress similarity between the model (with the centrifuge length Lmodel and the centrifuge acceleration of amodel ) and the prototype (with the length Lprototype and the earth’s gravity aprototype ), the scale factor is defined as:

Table 1 shows the scale factors of some parameters on the basis of dimensional analysis (Taylor 2005): The experimental study was carried out in the GeoCentrifuge at Deltares (The Netherlands). Figure 1 shows the facility. It was described in detail by Huy et al. (2008).

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2.2

Model piles

The model pile was made from steel with a length of 300 mm, a diameter of 11.3 mm (D), wall thickness of 0.5 mm and mass of 875 gram (M ); this mass includes the pile mass and the mounting gear on the pile head. Table 1.

Scale factors in centrifuge test.

A load cell was mounted on the pile head to measure the applied force. 2.3

Model materials

Baskarp sand (d50 = 130 µm) and silt (d50 = 58 µm) were chosen for the tests. Table 2 lists the basic parameters for the soils (the quoted values for friction angle

Parameters

Model

Prototype

Table 2.

Length/Displacement Acceleration Time (dynamics) Mass Velocity Force Stress Strain

1 N 1 1 1 1 1 1

N 1 N N3 1 N2 1 1

Parameters

Dimension

Sand

Silt

Grain vol. mass d50 Min. porosity Max. porosity Friction angle Permeability

kg/m3 µm % % degree m/s

2647 130 34 46.9 40◦ 12 × 10−5

2650 58 42.2 53.9 38◦ 1.5 × 10−5

Figure 1. Centrifuge test setup (Huy, 2008).


538

Properties of soils.

and permeability are at 65% relative density) and Figure 2 shows the grain size distribution curves. To minimise the scale effects, the ratio of pile wall thickness to the mean grain size d50 needs to be larger than 10 and the ratio of the inner diameter of pipe pile to d50 must be larger than 200 (Nicola & Randolph, 1997). The silt almost satisfies this condition (8.6 and 178). In the sand, the ratios are 3.9 and 79. In prototype terms, the test with silt corresponds to the normal use of open-end piles in sea-bed sand, while the test with sand is an extreme case in a fine gravel layer which is sometimes to be found in reality. The soil sample was prepared by drizzling sand into water, followed by densification using impact loading (Rietdijk et al., 2010). This method made it possible to achieve a reasonably homogeneous and reproducible sample of 65% relative density (for these types of soils). Water was selected as the model pore fluid for all tests. It is therefore reasonable to assume drained behaviour. Based on the results of Huy (2008), the response of the pile under rapid loading will be drained, with water as the pore fluid in both cases (Baskarp sand and silt). The effects of excess pore pressure can be ignored. Furthermore, it is very difficult to saturate the silt with viscous fluid and the silt and sand can be used again easily after the tests if water is used.

in sand (Huy et al., 2008) are also shown here for the purposes of comparison. 3

RESULTS OF THE CENTRIFUGE TEST

Figure 3 shows two typical results for measured pile head force and applied pile displacement. The pile head forces have been corrected for the self-weight of the pile. During an RLT, the pile can be seen as a rigid body. In that case, the force on the pile head (Fmeasured ) is equal to the sum of the soil resistance (Fsoil ) and the inertia force (Finertia ) of the pile (Middendorp et al., 1992). The soil resistance can therefore be calculated from:

where M is the pile mass and a is the pile acceleration. The acceleration is calculated numerically as the second derivative of the measured pile displacement at all time steps.

2.4 Test programme Three tests were performed at the gravity level N = 40 with the same loading programme: two tests in silt (one with an open-end pile (OEP) and one with a closedend pile (CEP)) and one test with an OEP in sand. During the tests, the pile was first pushed from the pre-embedded depth of 10D to a depth of 20D using the large hydraulic actuator. Two RLTs were then performed with displacements of 1% D (Rapid u = 1%) and 10% D (Rapid u = 10%) respectively (duration 10 ms) and, finally, an SLT with a displacement of 10% D (Static) was performed. The results from one test conducted previously (also at Deltares) with a CEP

Figure 2. Grain size distribution curves.


Figure 3. Measured Load-Displacement curves, OEP.

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Figure 4 shows an example of the measured pile head force, inertia force and resulting soil resistance and prescribed pile displacement from the RLT with silt. This soil resistance still includes velocity effects due to rapid loading. The applied force can be considered rapid, even though, compared with a field test (e.g.Matsumoto & Nishimura, 1996), the generated force has very steep flanges and a long duration of maximum force. 4

and Figure 7 shows the results of the tests in sand. Part a) shows the results for the OEP and part b) the results for the CEP. The test for the CEP in sand can be found in Huy et al. (2008). Generally, the soil resistance-displacement curves of RLTs have quite similar patterns: the force first rises quickly to its maximum value, then stays high at about the maximum value before finally falling rapidly.

DESCRIPTIONS AND DISCUSSION OF THE MODEL PILE TEST RESULTS

This section describes the comparison of SLTs and RLTs in silt and sand in detail. It should be pointed out that, from this point on, the soil resistance force during the RLT will be the calculated pile head force after eliminating the inertia force of the pile, and that all the numbers and quantities are in terms of model scale (N = 40 g). 4.1 Pile installation As described above, the model piles were pushed into the soil medium with the large hydraulic actuator from the initial depth of 10D to the final depth of 20D with a driving velocity of 10 mm/min.At this very low driving speed, the installation process can be considered as static jacking. Figure 5 shows the pushing records from the installation phase. It is clear that the installation of the model pile in sand requires about 30% more force than in silt. A possible explanation is the grain size of sand, which is quite large compared to the thickness of the pile wall. 4.2

Figure 5. Load-Displacement curve for installation phase.

Soil resistance

Figures 6 and 7 show the soil resistance-displacement curves for different maximum displacement values. Since the duration of the loading was the same in all tests, the loading speed also varies between these tests. Figure 6 shows the results of the tests in silt

Figure 4. Example of measured and calculated signals.


Figure 6. Load-Displacement curve for pile in silt.

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This loading pattern deviates from the loading pattern observed in field tests with a shallower increase to the maximum load and a shallower decrease to zero. This is a limitation of the hydraulic loading system, as seen in Figure 4. With the sand sample, the maximum soil resistance during the RLT is comparable with the maximum soil resistance during the SLT (RM = 1); with the silt sample, the maximum soil resistance during the RLT is 20% higher than the maximum soil resistance during the SLT (RM = 1.2). These differences apply to both the closed-end and open-end piles.

The maximum soil resistance of the closed-end pile is higher than the maximum soil resistance of the openend pile in both the RLT and SLT: about 30% for the sand sample and 10% for the silt sample. The ratios of soil resistance at the unloading point during the RLTs to maximum soil resistance during the SLTs (RUP in Table 3) were quite different in all tests. For the CEP in sand, the ratio was 0.9 while, in the other tests, the ratio was significantly less than 1. Figures 6 and 7 suggest that the unloading point method is not altogether appropriate for the RLTs in this study. This may be due to the steep loading pattern or the high inertia forces during these RLTs. The soil resistance observed during the SLTs in sand was higher than in silt: soil resistance with the OEP was 1.5 times higher; a factor 2 was found for the CEP. These differences could possibly be explained by the properties of the soil materials. Firstly, the friction angle of Baskarp sand is 1–2◦ higher than the friction angle of silt (at a relative density of 65%). Secondly, the d50 of the sand is 2.5 times larger than the d50 of the silt. The d50 governs the thickness of the shear band along the pile shaft, at the outer surface for the CEP pile and at the outer and inner surface for the OEP pile (Wolf et al., 2003; Wood, 2002), and at the pile tip. 4.3

Stiffness

Figure 7 also shows clearly that the stiffness of rapid loading is higher than that of static loading. This concurs with the numerical results for dynamically loaded piles in saturated soil of Hölscher and Barends (1992). 4.4

Figure 7. Load-Displacement curve for pile in sand. Table 3.

Plugging

After installation and all loading phases, the pile was dug out. The final plugging length of the soil inside the model piles was 55 mm (5D) with silt and 22 mm (2D) with sand. The total displacement of each pile was 122 mm (10.8D), with the total embedded length of each pile being 241 mm (20.8D). Plugging length as a percentage of the total embedded length of pile was about 23% for silt and 9% for sand. These are relatively extreme values for plugging length when compared to those generally observed in reality (10–20% of the embedded length of the pile) (Randolph et al., 1991). Since the measured plugging length is highly dependent on the material, it is important to use a correctly scaled material; in this case of N = 40, silt must be used.

Soil resistance in RLT and SLT at displacement of 10% D. Rapid Load

Test

Max Load [kN]

UP Load [kN]

Static Load [kN]

RM =

Closed-end Sand Open-end Sand Closed-end Silt Open-end Silt

1.33 1.00 0.81 0.73

1.21 0.38 0.19 0.26

1.35 0.95 0.66 0.60

0.99 1.05 1.22 1.21


541

FMax Load FStatic

RUP = 0.90 0.40 0.28 0.43

FUP Load FStatic

A close inspection of Figure 6 shows that the SLTs for the OEP and the CEP are almost identical. The RLTs for all piles show that the force declines after reaching the maximum. With the OEP, the force decreases slightly more than for the CEP and is slightly more perturbed. The soil column inside the pile may have slipped during the RLTs. However, the differences are small and the soil resistance of the open-end pile was quite comparable to the soil resistance of the closed-end pile. This suggests that the piles plug during both SLTs and RLTs. The motion of the plug would have to be measured directly to obtain more accurate information. 5

CONCLUSIONS

This paper described experimental work investigating soil plugs in open-end pipe piles in a geotechnical centrifuge. Both static and rapid load tests were studied in two types of soil: fine-grained sand and silt. The results of the model tests show that centrifuge testing is a feasible and efficient approach to studying the behaviour of open-end piles. The conclusions can summarised as follows: 1. The plugging length in sand is about 2.5 times less than the plugging length in silt. 2. The soil resistance of a closed-end pile is about 30% higher in sand and 10% higher in silt than the soil resistance of an open-end pile in both RLTs and SLTs. 3. The ratio of the maximum soil resistance in the RLTs to that in the SLTs depends on the soil type: 1.0 for sand and 1.2 for silt. 4. In the test in silt, the soil resistance at the unloading point in the RLTs seems to be unrelated to the soil resistance in the SLTs. The reason for this is not fully understood but it may be due to the steep loading pattern, which deviates from the smooth pattern assumed in the unloading point method, or due to the high inertia forces. 5. Rapid stiffness is significantly higher than static stiffness. 6. The proper scaling of an open-end pile requires proper scaling of the grain size. Silt must be used for a 1:40 scale. The research is still ongoing. To improve out understanding of plugging behaviour and the impact of plugging on open-end pile capacity during RLTs, the preliminary tests can be improved by:

2. Varying the loading rate to investigate its impact on both plugging and pile capacity. REFERENCES Bruno, D. & Randolph, M.F. 1999. Dynamic and static load testing of model piles driven into dense sand. Journal of Geotechnical and Geoenvironmental Engineering 125 (11) p. 988–998 De Nicola, A. and Randolph, M.F. 1997. The plugging behavior of driven and jacked piles in sand. Geotechnique 47 (4) p. 841–856 De Nicola, A. and Randolph, M.F. 1999. Centrifuge modeling of pipe piles in sand under axial loads. Geotecnique 49 (3) p. 295–318 Hölscher, P. & Barends, F.B.J. 1992. The relation between soil-parameters and one-dimensional toe-model. Proc. 4th Int. Conf. Application of Stress Wave Theory to Piles p. 413–419 Huy, N.Q., van Tol, A.F. and Holscher, P. 2008. Rapid model pile load tests in the geotechnical centrifuge. Rapid Load Testing on Piles. p. 103–127 Huy, N.Q. 2008. Rapid load testing of pile in sand. PhD thesis, Delft University of Technology Matsumoto, T. and Nishimura, S. 1996. Wave propagation phenomena in statnamic test of a steel pipe pile. Proc. 5th Int. Conf. Application of Stress Wave Theory to Piles p. 1015–1030 Middendorp, P., Bermingham, P. & Kuiper, B. 1992. Statnamic load testing of foundation piles. Proc. 4th Int. Conf. Application of Stress Wave Theory to Piles p. 581–588 Paik, K., Lee, J., Salgado, R. and Kim, B. 2003. Behavior of open- and close-ended piles driven into sand. Journal of Geotechnical and Geoenvironmental Engineering 129 (4) p. 296–306 Randolph, M.F., Leong, E.C. and Houlsby, G.T. 1991. One-dimensional analysis of soil plugs in pipe piles. Geotechnique 41 (4) p. 587–598 Randolph, M.F., May, M., Leong, E.C., Hyden, A.M. and Murff, J.D. 1992. Soil plug response in open-ended pipe piles. Journal of Geotechnical Engineering 118 (5) p. 743–759 Rietdijk, J., Schenkeveld, F.M., Scahminée, P.E.L. and Bezuijen, A. 2010. The drizzle method for sand sample preparation. Accepted in Proceedings of the International Conference on Physical Modelling in Geotechnics. Taylor, R.N. 2005. Centrifuges in modeling: principles and scale effects. Geotechnical Centrifuge Technology, Blackie Academic & Professional p. 20–34 Wolf, H., Konig, D. and Triantafyllidis, T. 2003. Experimental investigation of shear band patterns in granular material. Journal of Structural Geology 25 p. 1229–1240 Wood, D.M. 2002. Some observations of volumetric instabilities in soils. International Journal of Solids and Structures 39 p. 3429–3449

1. Measuring the plugging length during installation and all successive static and rapid loading steps;


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Field measurements on monopile Dolphins A. Sadeghi-Hokmabadi & A. Fakher School of Civil Engineering, University of Tehran, Iran

ABSTRACT: A monopile dolphin includes a single pile used for berthing and anchoring of large vessels in offshore or onshore terminals. Several methods have been developed to analyze piles under lateral loading. One of the most effective methods is the Strain Wedge Model (SWM) which has a number of advantages in comparison with traditional p-y curves. In the Pars Special Economic Energy Zone (Asalouyeh) in the south of Iran, a number of single piles as dolphins were constructed and some full-scale lateral loading tests were conducted on them under the supervision of the second author. In the present paper, a program called Lateral Analysis of Piles (LAP), which has been developed by the authors, is used to examine the Strain Wedge Model for pile analysis using the results of these full-scale loading tests. The research shows that the SWM calculates a greater pile head displacement than the test data, and illustrates the need for local calibration.

1

INTRODUCTION

In general, there are two types of berth structure; quay and jetty. A quay (or wharf) is a landing place parallel to a navigable waterway that provides access to ships and boats (Figure 1.a). Because of its high lateral resistance, the fenders must be well-designed to absorb the berthing energy of a ship. A jetty (or pier) extends out into the water from the shore. It is in the perpendicular direction to the shoreline serving as a landing place and where loading equipment allows the

use of a lighter structure. Ships can berth directly at the structure, but usually require separate structures, such as dolphins, to absorb the high energy of the ship (Figure 1.b). In some cases, dolphins consist of a number of piles. This type has low lateral deformation and, therefore, a reduced ability to absorb energy. A monopile comprises a single large-diameter pile which is embedded in the soil and behaves as a console. The ability of monopiles to absorb a high amount of energy, their low cost, and simple construction method has made

Figure 1. a) Schematic picture of a quay, b) Schematic picture of a Jetty with two berthing dolphins in middle and four mooring dolphins in sides, c) construction of Jetty with its Monopiles in Asalouyeh.


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them common alternatives for offshore structures such as wind turbines and mooring or berthing dolphins (Quinn 1972). In the analysis of monopiles, lateral behavior is important and the interaction between the pile and soil should be modeled accurately. A number of researchers have investigated laterally loaded pile behavior, providing a number of different approaches. These methods can be classified in to the following categories: (a) Continuum-base approaches; (b) Load-transfer (or subgrade reaction) approaches. In the first category, the soil has been modeled as a continuum media, requiring several soil properties inputs for analysis (Fleming et al. 1992). The complexity and unavailability of soil properties of this first approach make it less attractive. The load-transfer approach is more commonly used and was selected for this study. The load-transfer method models the pile as an elastic member and the soil as series of nonlinear springs (p-y curves). The nonlinear soil springs describe the local variation of lateral soil–pile interacting resistance with lateral displacement. Traditional p-y models were initially developed by Matlock (1970) and Reese et al. (1974). Later, a number of p-y curves were developed by different researchers (like Murchinson & O’Neill (1984) & Scott (1980)). Traditional p-y curves do not consider pile properties such as pile bending stiffness, pile cross-sectional shape, pile head restraint, and pile installation method (Ashour et al. 2004). SWM is an advanced method in comparison with traditional p-y curves. It can consider three-dimensional behavior of soil, the effect of piles dimension and shape, and the piles head conditions. However, SWM, as like as traditional p-y curves, is a semi-empirical method. In the other words, the main drawback to these approaches is that they are based on empirical parameters (i.e. the modulus of subgrade reaction) which can only be back figured from the results of pile load tests (Basile 2003). The aim of this study is to assess the accuracy of SWM by using the results of some full-scale tests in the Pars special economic energy zone area (Asalouyeh) in Iran. In the present paper, at first the characteristics of SWM are briefly discussed. Details and the results of the undertaken full-scale tests are shown and specs of developed computer program (LAP) has been describes. Later on, the tested monopiles are analyzed with LAP and the results are compared with the tests’ data and general conclusions are made.

2

STRAIN WEDGE MODEL

The Strain Wedge Model (SWM) is an approach that has been developed to predict the response of a flexible pile under lateral loading (Norris 1986). In the Strain Wedge Model (SWM), the soil resistance against the lateral loading is determined by the three-dimensional © 2011 by Taylor & Francis Group, LLC

Figure 2. Basic Strain Wedge in Uniform Soil (Ashour et al. 1998).

Figure 3. Soil-Pile interaction in Multisublayer Strain Wedge Model (Ashour et al. 1998).

passive wedge of soil that develops in front of the pile (Figure 2). As shown in Figure 2, this passive wedge is characterized by base angles, θm and βm , the current passive wedge depth h, and the spread of the fan angle, ϕm (the mobilized friction angle). The horizontal stress changes at the passive wedge face, σh , and the side shear τ, act. Indeed, SWM allows the assessment of the nonlinear p-y curve response of a laterally loaded pile based on the envisioned relationship between the three-dimensional responses of a flexible pile in the soil to its one-dimensional beam on elastic foundation parameters (Ashour et al. 1998) as in Figure 3. The main objective behind the development of the SWM is to solve the beam on elastic foundation (BEF) problem of a laterally loaded pile based on the envisioned soil-pile interaction and its dependence on both soil and pile properties. Compared to other approaches, the SWM depends on well known on accepted principles of soil mechanics (the stressstrain-strength relationship) and an effective stress soil analysis. For more information about SWM refer to Ashour et al. (1998 & 2004). This method is used in the present research to analyze the full-scale tested monopiles.

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3

UNDERTAKEN FULL-SCALE TESTS ON MONOPILES

Table 1.

Experimental researches conducted on the behavior of laterally loaded piles could be divided into two basic types, namely full-scale and small-scale or model testing. Full-scale tests are generally believed to provide the most accurate results, but they are rare because of the large costs required and difficulties involved. Therefore, the results of full scale tests are valuable. In the presented research, a number of full scale tests were performed on monopile dolphins.

Section

Monopile No. 2 1 ST52 2 ST52 3 ST60 4 ST70 5 ST70 6 ST70 Monopile No. 3 1 ST52 2 ST60 3 ST70 4 ST70 5 ST70 6 ST60 Monopile No. 4 1 ST52 2 ST60 3 ST70 4 ST70 5 ST70 6 ST70 7 ST70

Asalouyeh is located in southern Iran on the Persian Gulf. It is 300 km east of the city of Bushehr on the coast of Iran. Pars Petrochemical Port in Asalouyeh has 15 berths. At piers 5 and 15, monopiles are used as berthing and anchoring dolphins (Fig. 1.c). Four monopiles were tested. Monopiles No. 1 and 2 are the inner and outer piles of Berth 15 at a water depth of 14 m. Monopiles No. 3 and 4 are the inner and outer piles of Berth 5 at a water depth of 26 m. The final elevation of the monopile heads after installation was 5 m above mean sea level. These monopiles have a cylindrical shape and were made from three types of steel. The thickness and types of steel used are variable in depth and are shown in Table 1. Details of the monopiles are shown in Figure 4. The soil parameters in the field were obtained for each layer using borings. Because of the high soil stiffness, it was not possible to perform in-situ tests such as the standard penetration test. The geotechnical properties of the soil are shown in Table 2. These parameters were obtained by describing the disturbed samples and laboratory tests. For instance the internal friction angle is determined from laboratory shear box. Table 2 presents the drained density (γd ), wet density (γt ), estimated value of standard penetration test (Nspt ), effective cohesion (C ), internal friction angle in degrees (ϕ ) and undrained cohesion (C u ).

Outer diameter (m)

Thickness (mm)

Yielding stress (kN/m2 )

1.778 1.778 1.778 1.778 1.778

25.40 25.40 28.58 31.75 34.93

360000 420000 490000 490000 490000

1.905 1.905 1.905 1.905 1.905 1.905

25.40 28.58 28.58 34.93 41.28 44.45

360000 360000 420000 490000 490000 490000

1.778 1.778 1.778 1.778 1.778 1.778

25.40 25.40 28.58 31.75 34.93 34.93

360000 420000 490000 490000 490000 420000

1.905 1.905 1.905 1.905 1.905 1.905 1.905

25.40 25.40 28.58 31.75 34.93 44.45 41.28

360000 420000 490000 490000 490000 490000 490000

applied load and may be disregarded. Also, since the spacing between the piles (21.5 m) is more than eight times the diameter of the piles, there is no pile group effect (Fleming et al. 1992). Four monopiles were tested under lateral static loading. Monopiles No. 1 and 2 were loaded in five steps. Monopiles No. 3 and 4 were loaded in three steps to accommodate the displacement limitation of the jacking system. At each step, the displacement of each pile was measured using Total Station. The loading steps increased and, for each step, the load was applied for 15 min for small loads and 30 min for large loads. Figure 5 shows the results. As it mentioned, the loads are applied at the head of the monopile dolphins and the displacement is measured in their head as well.

3.2 Tests method and results


Type of steel

Monopile No. 1 1 ST52 2 ST60 3 ST70 4 ST70 5 ST70

3.1 Tests location

A heavy duty tension system was designed and constructed that uses a hydraulic jack to provide force and a cable to transfer tension force from one monopile to another. The testing followed ASTM D3966-81, item 24 (ASTM 1995). The tension system sat on one monopile and pulled the other one. Bolts placed in the head of the monopiles for a quick release system were used for the temporary installation of the tension system on one monopile and a pulley on the other. Cables were installed between the tension system on one monopile and support on the other with a 56 in diameter pipe between them to support the weight of the cables and avoid any initial force from them. This pipe is allowed to have axial displacement. Analysis shows that the maximum friction between the cable and pipe was less than 3% of the

Details of monopile sections.

4

LATERAL ANALYSIS OF PILES (LAP) PROGRAM

A program was developed to analyze the monopiles. The Lateral Analysis of Piles (LAP) program was written in FORTRAN programming language to solve the governing equation for a beam on an elastic foundation (Equation 1) by Hetenyi (1946),

545

Figure 4. Details of monopiles in Asalouyeh. Table 2.

Geotechnical properties of the soil in the field.

Layer description

Sand

Sand and gravel

Sand stone

Depth (m) classification γd (ton/m3 ) γt (ton/m3 ) NSPT C (ton/m2 ) ϕ (◦ ) Cu (ton/m2 )

0.0–8.0 SP 1.7 2 >50 0 38 0

8.0–21.0 GP 1.95 2.1 >50 0 40 0

21.0–30.0 — 1.8 2.1 — — — —

where EI = bending stiffness of the pile; Px = axial load on the pile; y = lateral deflection of the pile at point x along the length of the pile; and Es = soil subgrade reaction (spring stiffness). LAP uses the finite difference method proposed by Matlock and Reese (1961) to solve Equation 1. It considers four sets of boundary conditions at the top of the pile, such as free-head or fixed-head pile. Also, LAP can use different types of spring stiffness (Sadeghi-Hokmabadi et al. 2009) like linear springs, Non-linear p-y curves, and SWM. In addition, LAP can assess pile group behavior under lateral and dynamic lateral loading such as earthquake loads (Sadeghi-Hokmabadi 2009). In the © 2011 by Taylor & Francis Group, LLC

546

Figure 5. Results of lateral loading tests on monopile dolphins.

present research, it was used as a means to analyze the dolphins at Asalouyeh and examine the accuracy of SWM. 5

COMPARISON AND DISCUSION

In the present research, LAP is used to analyze the tested monopiles using SWM. In addition, the mentioned monopiles are analyzed using COM624

(Resse & Sullivan 1980) as well. COM624 is a program for analyzing single piles under lateral loads and uses the p-y curves suggested by Reese et al. (1974). Figure 6 presents the results of analyses for the four monopiles in the term of head displacement versus lateral load at the head of monopiles. As Figure 6 shows, both the SWM and COM624 calculate a greater pile head deflection than the measured data. In comparison, the SWM gives closer answers with measurements undertaken in the presented case study than COM624. The SWM receives force at the pile head as an input and gives the pile head displacement as output (Ashour 1998). This method calculates p-y curves during the computation. In other words, the SWM does not use pre-defined p-y curves like the traditional p-y method (Fakher et al. 2009), and it is not possible to define a certain modification factor for this method like the p-y method. The average of ratio between the calculated pile head displacement and the observed one for monopiles number 1 to 4 is calculated as 0.82, 0.85, 0.87, and 0.98 respectively. Also, the total average of this ratio for these 4 set of monopiles is 0.88. It means that the data of performed tests are 12 percent less than predicted pile head displacement using the SWM. The real behaviour of pile head displacement is nonlinear. The p-y curves and SWM have difference with real situation in the flexure of pile head-displacement curves. Indeed, the total lateral stiffness in the real situation declines sooner, but in these methods it decline later and have approximately linear behaviour in the tests loads. The difference between proposed p-y curves and SWM with the real situation is occurred because of the development of plastic region near the soil surface. In fact, in the real situation under the testing loads the near surface soil has a plastic manner and yields, but p-y curves and SWM do not show this behaviour under the tests loads level. It should be noted that the total behaviour of pile-spring system is very sensitive for the near surface soils, and these soils should be modeled carefully (Fakher et al. 2009).

6

CONCLUSION

The results of full-scale tests on large diameter piles showed that the monopile dolphins behave like long piles. The LAP program was developed to analyze piles under lateral loading. This program has the ability to consider different boundary conditions and types of spring stiffness like p-y curves and Strain Wedge Model. According to the results of full-scale in-situ test, the accuracy of Strain Wedge Model has been investigated. In granular marine soils, the traditional p-y curves and SWM calculate the pile head displacement as being greater than the test data from the present study. This means that real piles withstood large amounts © 2011 by Taylor & Francis Group, LLC

547

Figure 6. Pile head horizontal displacement versus lateral load according to test data and LAP analysis for monopiles No.1, 2, 3, and 4.

of force for the specified displacements. Thus, using these curves without calibration leads to overestimating the piles displacements and demonstrates the need for local calibration.

In comparison, the SWM gives the closest answers to the measurements undertaken in the presented case study than COM624. The shape of the pile head displacement under real conditions declines sooner than in the calculated results because the analytical models do not show the soil plasticity near ground depth. To modify, the ultimate resistance of the non-linear springs should be decreased and the primary stiffness should be increased. REFERENCES Ashour, M., Norris, G. & Pilling, P. 1998. Lateral loading of a pile in layered soil using the strain wedge model. Journal of Geotechnical and Geoenviromental Engineering, Vol. 124, No.4. Ashour, M., Pilling, P., & Norris, G. 2004. Lateral Behavior of Pile Groups in Layered Soils. Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 130 (6), pp. 580–592. Basile, F. 2003. Analysis and design of pile groups, In Numerical Analysis and Modeling in Geomechanics (eds J. W. Bull). Spon Press, London, Chapter 10, 278–315. Fakher A., Sadeghi-Hokmabadi A., & Saeedi-Azizkandi A. 2009. Assessment of lateral load-transfer methods of piles by full scale in-situ tests. Proceeding of the 2nd International Conference on New Developments in Soil Mechanics and Geotechnical Engineering, Nicosia, Cyprus; pp. 230–238. Fleming, W.G.K., Weltman, A.J., Randolph, M.F., & Elson, W.K. 1992. Piling Engineering. 2nd Edition, Blackie Academic & Professional, Glasgow, UK.


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Hetenyi, M. 1946. Beams on elastic foundation. The University of Michigan Press, Ann Arbor. Murchison, J. M. & O’Neill, M. W. 1984. Evaluation of p-y relationships in cohesion less soil. Analysis and design of pile foundations, ASCE, New York, 174–191. Matlock, H. 1970. Correlations for design of laterally loaded piles in soft clay. Proc., 2nd Annual Offshore Technol. Conference, Houston, Texas. Norris, G.M. 1986. Theoretically based BEF Laterally Loaded Pile Analysis. Proceedings, Third International Conference on Numerical Methods in Offshore Piling, Nantes, France, pp. 361–386. Quinn, A.D. 1972. Design and construction of ports and marine structures. McGraw-Hill Inc., USA. Reese, L.C. and Cox, W.R., & Koop, F.D. 1974. Analysis of laterally loaded piles in sand. 6th Annual Offshore Technology Conference, Austin Texas, 2(OTC2080): 473–485. Reese, L.C., & Sullivan, W.R. 1980. Documentation of computer program COM624 parts 1 and 2: analysis of stresses and deflections for laterally loaded piles including generation of p-y curves. Geotech. Eng. Ctr., Bureau of Eng. Res., Uni. of Texas, Austin. Texas. Sadeghi-Hokmabadi, A. 2009. Development of a computer program for the analysis of single piles and pile groups under lateral loads. Post-graduate research thesis, University of Tehran, Iran. Sadeghi-Hokmabadi, A., Seyfi, H. & Fakher, A. 2009. Analysis of single piles under lateral loading using the Strain Wedge Model. 8th International Congress of Civil Engineering (8ICCE), Shiraz University, Shiraz, Iran (In Farsi Language). Scott, R.F. 1980. Analysis of centrifuge pile tests: Simulation of pile driving. Research Rep. OSAPR Project 13, American Petroleum Institute, Washington, DC.


Behaviour of driven tubular steel piles in calcarenite for a marine jetty in Fujairah, United Arab Emirates J. Thomas WorleyParsons Services Pty Ltd, Perth, Western Australia

M. van den Berg Delta Marine Consultants, The Netherlands

F. Chow WorleyParsons Services Pty Ltd, Perth, Western Australia

N. Maas Vopak Horizon Fujairah Ltd, Fujairah, United Arab Emirates

ABSTRACT: Vopak Horizon Fujairah Limited is the biggest oil storage and bunkering company in the Emirate of Fujairah with a sizeable terminal that has recently undergone its fifth expansion. This includes an extension to the existing marine jetty with the provision of two additional berths in 15m water depth for vessels with a capacity up to 110,000 DWT. The jetty structures are founded on tubular steel piles driven into shallow variably cemented very weak to weak sedimentary rocks with varying calcium carbonate content. A static uplift load test to failure followed by a static compression load and numerous dynamic End of Initial Drives (EoID) and restrike tests with signal matching were carried out for a better understanding of the pile behaviour. This paper describes the pile tests conducted, results obtained and their interpretations. The suitability of the API RP2A method for clays in estimating capacity of piles driven into the calcarenite at this site is investigated. The pile response from the static load test due to reversals in loading directions is also discussed.

1

INTRODUCTION

2

The jetty structures at Vopak Horizon Fujairah’s terminal are founded on tubular steel piles driven into variably cemented very weak to weak sedimentary rocks with varying calcium carbonate content. To verify pile design, a static uplift load test to failure followed by a static compression load test was carried out. During production piling, numerous dynamic tests with signal matching were carried out to verify the axial pile capacities. The static load test was carried out on a sacrificial pile. The pile was driven to a penetration of 18.41 m from the seabed level. An average shaft friction of 81 kPa was estimated from the uplift load test. During production piling, dynamic EoID and restrike test were carried out. VHFL asked WorleyParsons to review the dynamic test results and confirm that the tests were carried out in accordance to generally accepted quality standards. As part of the review WorleyParsons also assessed the pile capacity based on the API RP2A WSD (2007) clay recommendations, utilising borehole specific design UCS profiles. © 2011 by Taylor & Francis Group, LLC

PROJECT DESCRIPTION

The jetty expansion is part of the Phase V expansion project of VHFL in the Emirate of Fujairah in the United Arab Emirates. For this project DMC was designer for the Civil Marine works. The new expansion involves two new berths in the shape of a finger pier. The new berths are required to increase the ship handling capacity of the terminal due to the increase of the storage capacity on land. The berths are designed for vessels with a capacity from 5,000 to 110,000 DWT and are situated in 15 m water depth in an area sensitive to seismic activities (Zone 2b according to the Uniform Building Code). Construction started in September 2007 and was completed in December 2009. The finger pier consists of: • An access trestle with a length of approximately 300 m. • Two loading platforms • Eight breasting dolphins • Six mooring dolphins • Interconnecting walkways in between the dolphins.

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Figure 1. Overview of new berths. Table 1.

Pile working loads (kN).

Loading

Min

Max

Tension Compression

0 400

1500 3000 Figure 2. Carbonate content.

In total 218 piles have been driven. Rock coring was performed using PQ double tube core barrel producing a nominal diameter hole of 121 mm and a nominal core diameter of 84 mm. Standard Penetration Tests (SPTs) in accordance with BS 1377 (1990) were conducted in uncemented to slightly cemented soils. Laboratory tests including particle size distribution, Uniaxial Compressive Strength (UCS) and carbonate content were carried out on recovered SPT and core samples.

• 200 OD 914 mm 19 mm WT • 18 OD 1016 mm 22 mm and 25 mm WT

The structural design is carried out in accordance to British Standards. The design of the foundation is executed based on working loads using an overall factor of safety on the pile bearing capacity. All the piles driven have been subjected to dynamic testing. For this reason the factors of safety for the final assessment after driving have been taken as 2.0 for tension and 1.75 for compression. The range of working loads for the piles is given in Table 1 above. 3

SITE GEOLOGY

The jetty site is located in Fujairah, situated along the east coast of the United Arab Emirates along the Gulf of Oman. The near-surface geology of the coastal Fujairah region is dominated by alluvial “wadi” gravel plains. The alluvial deposits are generally derived from predominantly ophiolitic, igneous rock forming the nearby Hajjar Mountain Range to the west.The alluvial gravel deposits typically extend to the coastline, where intercalation and reworking with carbonate marine sediments is commonly evident. 4

4.1

The soil and rock were classified in accordance to the Clark and Walker Carbonate Sediment Classification System. The total carbonate content was estimated in accordance with the rapid titration method of BS 1377 assuming all of the carbonate present was calcium carbonate. Figure 2 shows the carbonate content profile at the site. The rock materials encountered were generally logged as siliceous calcarenite to a depth of −27 m CD and underlain by calcareous sandstone to the remaining investigated depth. 4.2


Uniaxial compressive strength

UCS tests were carried out on selected rock core samples in accordance with the methods outlined in ASTM D2938 (2002). The measured UCS values are provided in Figure 3.

SITE INVESTIGATION

4.3

Thirteen boreholes at the project site were sunk generally to a depth of 20 m below existing seabed. It is understood that an Edeco T30 rotary drive drill rig was used mounted on a jack-up drilling platform. The boreholes were advanced through variably cemented sand using rotary wash boring equipment and techniques.

Carbonate content

Design UCS profile

A UCS strength profile is required for pile design and was assessed from a combination of laboratory testing, field strength estimation and core photographs. Reliable UCS testing of calcarenite tends to be difficult as pre-existing but visually unidentifiable defects and/or anisotropic cementation results in premature

550

5

Figure 3. Design UCS profile.

failure of the UCS specimens and gives results that are generally lower than the true intact rock strength. For drilled and grouted piles, a strength profile chosen from inaccurate laboratory test results might only cause a conservative pile design, whereas a conservatively chosen design UCS profile based on inaccurate test results may result premature refusal of driven piles. It is therefore important to assess an optimum UCS design profile both for the sake of capacity assessment and pile driveability. The design UCS profile was selected based on the strength description in the borehole log, core photographs, measured UCS values from the recovered samples and experience with previous pile driving operations in similar geological formations. If the strength is highly variable with depth (generally observed for cemented sedimentary rocks), a certain averaging of the UCS strength profile relying on experience and judgment is required. At this site, Point Load Index tests (Is50 ) were not carried out. If Is50 values were available, then these values would have been added to the strength data base by converting to inferred UCS values. The inferred UCS values may be obtained by linearly correlating the Is50 values to the adjacent UCS results, after filtering out unreliable tests. Field/laboratory Point Load Index tests are generally numerous and likely to provide better representation of the inherent vertical strength variability in calcarenite than UCS tests. The pile capacity assessment should be based on reasonable lower bound strength parameters, while pile driveability should be based on reasonable upper bound strength parameters. The Is50 values are more likely to provide an upper band of strength data and are therefore valuable for a better prediction of driveability. © 2011 by Taylor & Francis Group, LLC

PILE DESIGN APPROACH

The required pile penetrations to carry the loading platforms access trestle and dolphin loads were verified by a static load test carried out on a test pile (TS38) driven to a penetration of 18.41 m from the seabed level. Based on the uplift load test, an average shaft friction of 81 kPa for the test pile was estimated (excluding the pile and the plug weight). The estimated pile penetration based on this value was found to be too optimistic for other piles, indicated by dynamic tests during production piling. In this case it appears that the longer test pile had a higher average shaft capacity than the shorter working piles. This contrasts with the normal expectation of skin friction degradation, i.e. a lower average shaft capacity for longer driven piles in sands and clays, e.g. Jardine et al. (2005). Rock strength variability may have contributed to this observation. The effects of variations in rock strength can be reduced by using a design method based on sound engineering principles that takes account of strength variability. Beaumont and Thomas (2007) used the clay design method provided in API RP2A WSD (2007) to predict pile capacity of steel tubular piles driven into similar geological formations in the Pilbara coast of the northwest Australia. In their approach, the very weak to weak rock was considered as a cohesive material and the undrained shear strength (su ) of the material was taken as 0.5 × UCS for the purpose of estimating pile capacity. The suitability of this approach for the Fujairah site was assessed and is discussed below. The test pile was installed 10 m from BH8 and 30 m from BH9. The predicted pile capacity based on the API RP2A clay method is provided in Figures 4 and 5 for boreholes BH8 and BH9 respectively. The design UCS profiles provided in Figure 3 were used for estimating pile capacity. The skin friction in tension was assumed to be 0.75 times the skin friction in compression. The end bearing capacity was estimated by applying unit end bearing resistance of 4.5 × UCS over the gross area of the pile, multiplied by a factor of 0.77 to account for the compressibility of the plug, as suggested by Bruno and Randolph (1999). At strata boundaries the full end bearing resistance of the stronger material was assumed to mobilize after a pile penetration of one pile diameter into the stronger material.

6

STATIC LOAD TESTS

A 914 mm OD × 19 mm WT steel tubular test pile without a driving shoe was installed on 25 January 2008 using a Delmag D100-13 diesel hammer. The test pile TS38 is part of the access trestle. The seabed level at the pile location is −14.5 m CD and the pile penetrated to about 18.41 m (pile toe level at about −32.91 m CD).

551

Table 2.

Pile stiffness (kN/mm).

Test direction

Tension

Compression

Unload-reload loop Initial/reload Unload

178 300 250

218 135 253

Figure 4. Ultimate pile capacity – BH8.

Figure 6. Static load test – Pile TS38 (tension negative).

Figure 5. Ultimate pile capacity – BH9.

The test pile was surrounded by four permanent piles in a square grid with a spacing of 5.5 m, which served as reaction piles during the load test. The reaction piles were located 4.2 pile diameters away from the test pile. A static uplift test was carried out 46 days after pile installation (42 days after the dynamic restrike test) followed by a static compression test 3 days after the completion of the static uplift test. The uplift pile capacity was about 4610 kN and appears to have fully mobilized during testing. The compression test was stopped at a pile load of about 5000 kN before full © 2011 by Taylor & Francis Group, LLC

mobilization of pile capacity to avoid overloading of the reaction piles. The load test data is shown in Figure 6. The pile toe movement was assessed to be about 49.4 mm at the maximum uplift load (pile head movement of 66.5 mm at maximum uplift test load with elastic pile elongation of 17 mm). One unload-reload load cycling in both tension and compression loading was carried out during testing. The pile stiffnesses are provided in Table 2. During the uplift load test, a reversal of loading direction was experienced by the pile. The pile was subjected to another reversal of loading direction during the subsequent compression loading. The residual stresses present in the pile after uplift unloading, elastic pile compression and the reversal in principal stress direction are probably responsible for the reduced stiffness apparent during the first compression cycle at a compression load of about 1500 kN. The increased stiffness at the initial stage of the reloading phase of the uplift test following the unload-reload loop is also likely to be due to presence of residual stresses. The compression load cycle indicated a shaft friction reduction to about 1500 kN. This is probably due to the reversal in principal stress directions and Poisson loading effects (reduction or increase in pile diameter as the pile is elongated in tension or compressed in compression). Similar but smaller (maximum 26%) reductions in shaft capacity were recorded by the relatively stiff Imperial College instrumented pile in medium dense silica sand due to a change in loading

552

direction (Chow, 1997). Restrike tests on some piles at Fujairah also showed apparent reductions in shaft friction as the piles were driven deeper. In some cases the magnitude of the reduction was greater than the increase in shaft friction observed during initial set-up (the latter being based on the comparison of the EOID blowcount and the first few blows during restrike). It is generally expected that driven or drilled and grouted piles in siliceous carbonate or carbonate material show post peak friction degradation behaviour. Thomas (1998) among others observed strain softening response in piles jacked in/driven into overconsolidated stiff clays upon a reversal of loading direction. However, in this particular case a reversal of loading direction did not apparently cause any overall strain softening behaviour. It is possible that localized strain softening was masked by progressive shaft failure and/or the gradual mobilization of base capacity with pile toe displacement.

7

Figure 7. Results of dynamic pile tests.

CAPWAP RESULTS

Dynamic testing was carried out on all of the piles in the pile group where TS38 test pile was present. The EoID soil resistance of these piles is provided in Figure 7. The borehole logs and the core photographs indicate a high degree of strength variability in the vertical and lateral directions. During pile driving operations, this variability was evident as the observed blowcounts are quite different for similar adjacent piles driven with same hammer. Dynamic tests carried out on other piles in the same group as TS38 clearly shows this variability. A variation of about 100% in EoID soil resistance to driving was observed between the least resistant and most resistant piles. A larger variability was observed for the end bearing component of the soil resistance. This indicates the presence of bands of stronger rock where a pile can refuse and may result in inadequate uplift capacity if the chosen hammer is not adequately sized.

8

Figure 8. Results of driveability analysis.

DRIVEABILITY ANALYSIS

A driveability analysis was carried out using the GRLWEAP (2005) computer program. The soil resistance to driving was assumed equivalent to the ultimate pile capacity estimated using the API RP2A clay method for a coring pile.This assumption may be warranted for driveability due to the high level of strength variability in calcarenite. The default parameters provided in the GRLWEAP program for the Delmag D100-13 diesel hammer with a hammer efficiency of 65% was used in the analysis. In this analysis, a skin quake of 2 mm, a toe quake of 1 mm, a skin damping of 0.5 sec/m and a toe damping of 0.35 sec/m were used. Driveability analyses were carried out for borehole profiles BH8 and BH9. The predicted blowcounts for these two borehole profiles are provided in Figure 8. © 2011 by Taylor & Francis Group, LLC

Also included in the figure are the observed blowcounts of test pile TS38. The predicted and observed blowcount profiles show excellent comparison and this is another indication of the applicability of the API RP2A clay method for capacity prediction for the siliceous calcarenite present in this site. The observed blowcounts of another pile in the group is also provided as an indication of the variability that was observed. Variability in observed blowcounts during production piling were found to be quite high, some of the piles in a particular group were driven to target penetration effortlessly whilst some other piles in the same group met refusal prior to reaching target penetration. There was no apparent link between driving blowcount and the order that the piles were driven in.

553

• Reversals in pile test direction caused apparent

reductions in shaft capacity due to changes in principal stress direction and Poisson loading effects. • Significant pile set-up may be expected for piles driven into siliceous calcarenite and/or calcareous sandstone. • Pile uplift capacity is about 75% of the compression shaft friction obtained from a reliably conducted dynamic test.

REFERENCES

Figure 9. Measured setup for shaft friction and end bearing.

9

PILE SET-UP

A restrike test was carried out on TS38, 4 days after pile installation. Comparing the results of EoID and restrike tests indicates that the shaft friction increased by about 18% and the end bearing increased by about 16%. Restrike tests were carried out on 14 additional piles of individual pile groups during production piling to assess the setup factor for the pile group. The shaft friction set-up for the other piles was generally observed to be higher than the test pile as shown on Figure 9. The reduction in end bearing resistance was due to the under-mobilisation of end bearing, probably due to inadequate delivery of hammer energy and/or insufficient hammer size. 10

CONCLUSIONS

The following conclusions are drawn from the static and dynamic tests carried out on a 914 mm diameter driven steel tubular pile:

American Petroleum Institute Recommended Practice 2AWSD (2007), Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – Working Stress Design. ASTM D2938 (2002), Standard Test Method for Unconfined Compressive Strength of Intact Rock Core Specimens. Beaumont, D. and Thomas, J. (2007), Driving Tubular Steel Piles into Weak Rock – Western Australian Experience, Proceedings, 10th Australia New Zealand Conference on Geomechanics, pp. 430–435. British Standards Institute (1990). BS 1377 Methods of test for Soils for civil engineering purposes. Bruno, D. and Randolph, M.F. (1999), Dynamic and Static Load Testing of Model Piles Driven into Dense Sand, Journal of Geotechnical Engineering Division, ASCE. Chow, F.C. (1997). Investigations into displacement pile behaviour for offshore foundations, Ph.D Thesis, University of London (Imperial College). Clark, A.R. and Walker, B.F. (1977), A proposed scheme for the classification and nomenclature for use in the engineering description of Middle Eastern sedimentary rocks, Geotechnique, Vol. 27, pp. 93–99. Jardine, R.J., Chow, F.C., Overy, R.F. & Standing, J.R. (2005). ICP design methods for driven piles in sands and clays. Thomas Telford (publ). Thomas, J. (1998), Performance of piles and pile groups in clay, PhD Thesis, The University of Western Australia. Uniform Building Code (1997), Structural Engineering Design Provisions, Volume 2, Table 16 I.

• The capacity of a steel tubular pile driven into

rock at this site may be estimated using the API RP2A clay method. However, this approach may not be applicable at all calcarenite sites. Its use should be undertaken with care until proven through driveability trials and pile load tests.


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CPT-Based design method for axial capacity of offshore piles in clays B.F.J. Van Dijk & H.J. Kolk Fugro, Leidschendam, The Netherlands

ABSTRACT: A statistically reliable design method is proposed for large pipe piles driven into clays. The method relies on Cone Penetration Test (CPT) input parameters and a database containing high quality data on axial pile load tests in clays. Factors such as soil plasticity, overconsolidation ratio, pile length, pile slenderness and time between pile driving and loading were considered.

1

INTRODUCTION

Most offshore design methods use peak undrained shear strength as the primary input parameter for determining axial capacity of pipe piles driven into clays, e.g. API (2000) Main Text and Kolk & Van der Velde (1996). However, undrained shear strength is not a unique parameter. A particular “reference” laboratory undrained shear strength must be considered, usually defined by some combination of soil sampling method and laboratory test method. For example, API (2000) is largely based on tube push sampling and unconsolidated undrained triaxial compression and miniature vane testing. In practice, “reference” values are commonly inferred from correlations with non-reference values, such as Cone Penetration Test (CPT) inferred values, supplemented by engineering judgement. This practice introduces significant uncertainties in axial pile design (Van der Wal et al. 2010). The (Piezo) Cone Penetration Test (CPTu) is a more robust method to assess in-situ soil strength. Thus, it is no surprise that many CPT-based design methods are available for determining axial capacity of piles driven into clays. These semi-empirical methods have typically been calibrated for pile geometries and pile types common in onshore practice. They may not be appropriate for relatively long and large diameter pipe piles typically used offshore. This paper proposes a statistically reliable CPT based design method for large pipe piles driven into clays, based on a database containing results of high quality pile load tests in clays.

2

DATABASE

2.1 Selection of pile load tests Thirty-three high quality pile load tests on driven piles at fifteen different locations were selected from public literature (included in the reference list) for this study. High quality refers to both the pile load test and © 2011 by Taylor & Francis Group, LLC

the soil data. Typical reasons for excluding other pile load tests include: test pile pushed rather than driven, other non-representative installation methods, implausible differences in repeat pile load tests, uncertainty on representative CPT data, and significant amounts of sand inclusions within clay layers. Starting point for the database was an earlier study by Kolk & Van der Velde (1996). The Kolk & Van der Velde database (KV-database) was extended with results from eighteen other high quality pile load tests. The database contains results of both axial compression and tension pile load tests. The piles selected for the database are predominately friction piles in clay for which sufficient high quality soil and load test data were readily available. Details of the pile load tests are presented in Table 1.

2.2

Soil parameters

For the database, CPT cone resistance qc , CPT pore pressure u2 and equilibrium in-situ pore pressure u0 estimates were determined at 0.1 m depth spacing. Additionally, average values of soil unit weight, plasticity index PI, liquid limit LL and undrained shear strength cu were determined. The net cone resistance qn was derived using:

where σvo = total in-situ overburden pressure; and a = the cone dependent net area ratio of the crosssectional steel area at the gap between cone and friction sleeve to the cone base area. No values for the net cone resistance qn and/or the pore pressure (u2 ) were available for 12 pile load tests. For these instances, qn was derived from qc by estimating pore pressure u2 according to correlations such as presented by Mayne (1990) and Robertson (1990). For 15 pile load tests the cone dependent net area ratio a was not available, in which case a value of 0.75 was selected.

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Table 1.

Database of selected pile load tests.

Location

Load Length Diameter Pile tip Set-up test** m m – days –

1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20 21 22 23 24 25 26 27 28 29 30 31 32 33

22.6 4.5 3.7 3.6 9.1 9.1 9.0 9.0 48.0 48.0 11.6 11.6 11.6 13.1 13.1 10.8 18.6 21.9 21.9 9.4 50.0 10.0 10.0 10.0 40.0 7.5 9.6 7.6 14.5 14.5 30.0 71.4 18.3

Aquatic Park* Canons Park Canons Park Canons Park Cowden Cowden Cowden Cowden Garigliano Garigliano* Hamilton Hamilton* Hamilton* Houston* Houston* Lulu Island Kontich* Kontich* Kontich Merville Mortaiolo* Onsoy Onsoy Onsoy Pentre* Pentre* Tilbrook* Tilbrook* Tilbrook Tilbrook Tilbrook* West Delta West Sole*

0.762 0.168 0.102 0.102 0.457 0.457 0.305 0.203 0.381 0.381 0.114 0.114 0.114 0.273 0.273 0.324 0.61 0.61 0.61 0.508 0.477 0.219 0.219 0.812 0.762 0.219 0.219 0.219 0.219 0.273 0.762 0.762 0.762

Open Closed Closed Open Open Closed Open Open Closed Closed Closed Open Open Closed Closed Closed Open Open Open Open Closed Closed Closed Open Open Closed Closed Closed Closed Open Open Open Open

60 60 2 2 30 30 – – – – 18 18 18 105 105 82 12 21 26 41 25 40 40 40 44 31 60 60 60 60 134 133 0.25

T C T T C C T T C C C C C C C C C C T C C T T T C T T T T T C T C

Figure 1. Predicted versus measured capacity.

Bustamante et al. (1994) reported that the tip of pile 1 at Garigliano (no. 9) was installed in a sand layer with a high end-bearing capacity. The end bearing was measured during the test. The measured ultimate pile capacity for this pile was corrected for this measured end bearing value. Thus, only the shaft friction of the pile was used in the study.

* Derived from Kolk & Van der Velde (1996) database. ** T = Tension loading; C = Compressive loading test.

3

The normalised cone resistance Qt was used as a measure for the overconsolidation ratio OCR:

where σvo = effective in-situ overburden pressure. This links with Chen & Mayne (1996) who found OCR = 0.32 Qt for 205 clay sites around the world.

2.3

Pile test results

The ultimate resistance Q of a pile was defined as the maximum resistance measured during the test, corrected for the effective weight of the pile (total weight of the pile minus the weight of the displaced soil), giving consideration to whether testing was in compression or tension. Unit shaft friction f in occasional sand layers was calculated using API (2000) Main Text recommen dations for dense sand: f = min (βσvo , fmax ) where β = shaft friction factor; and fmax = limiting unit shaft friction value. © 2011 by Taylor & Francis Group, LLC

EXISTING DESIGN METHODS

Three existing CPT-based design methods to determine axial capacity of piles in clays were considered (Fig. 1): Bustamante & Gianeselli (1982), Almeida et al. (1996) and CUR (2001). Table 2 shows statistics for the ratio of predicted to measured pile capacity, R. The CUR (2001) method provides the better predictions with R = 0.98. However, 39% of the tests are either under or over predicting by more than 20%. Only the Almeida et al. (1996) method considers the overconsolidation ratio. Although their study did not explicitly reveal any length effects, they recommend to reduce shaft friction as suggested by Semple & Rigden (1984) or Randolph & Murphy (1985) in case the ratio embedded pile length L to diameter D exceeds 60. The Almeida et al. (1996) method also considers a more conservative relation in case PI < 20%. 4 4.1

DEVELOPMENT OF NEW DESIGN METHOD General equation

A new design method was developed, considering possible dependency of shaft friction on plasticity and

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overconsolidation of the soil, and pile characteristics such as length, diameter and wall thickness WT. The general equation for pile capacity in compression was used, assuming a fully plugged pile tip condition:

where q is unit end bearing and z is depth below seafloor. In case of tension, this study considered reversed end bearing where applicable. However, reversed end bearing may not always occur (e.g. when thin sand layers are present). For design purposes, following standard engineering practice for tension piles, the authors suggest not to rely on pile tip resistance for tension piles (i.e. no suction) and thus to omit the first term in Equation (3). It is noted that offshore piles are usually long and slender steel pipe piles with a relatively large diameter. Usually the pile tip condition of these piles is plugged under static loading.

Figure 2. R versus Plasticity Index PI.

where z = depth below seafloor; h = distance between pile tip level and z. A basic equation for computing unit shaft friction using a CPT method can be written as:

The following mathematical expression allows exploring dependency of ks on a range of parameters:

4.2 End bearing The following equation was adopted for calculating unit end bearing q in clays:

where qn,av = average qn value. This equation corresponds with the commonly used relation q = 9 cu , with a cone factor Nk = qn /cu of 13 (NGI 2006). For qn,av the authors suggest using the average qn value between 1.5D above and 1.5D below pile tip level. This is according to principles for CPT-based methods as given by API (2000) for sandy soils. Equation (4) could not be verified because of lack of reliable data. However the end bearing of long offshore piles in clay is usually small compared to the shaft friction. Any prediction errors thus have limited effect on Q. 4.3 Outer shaft friction The Kolk & Van der Velde (1996) method was considered as starting point for prediction of the outer unit shaft friction f . This method accounts for pile length effects and soil overconsolidation. The method is an α-method. It uses the peak intact undrained shear strength cu (based on unconsolidated undrained triaxial compression tests) as the primary soil parameter, according to:

and


where Ar = 1-((D-2WT )/D)2 = pile tip displacement ratio; j, k, l, m, n and o are dimensionless parameters; and uL = unit length to render the expression dimensionless = 1.0 m = 3.3 feet. Parameters j, k, l, m, n and o were determined using regression analysis, exploring lowest coefficient of variation on the ratio of predicted to measured pile capacity, R. Using the most suitable values considering only the pile load tests of the KV-database, Equation (8) becomes:

A value of 0.08 was selected for kmax . This value corresponds with a unit shaft friction equal to cu when Nk is 13 (NGI 2006). For the pile load tests of KV-database the new method (combining Equations (3), (4), (7) and (9)) results in an average of 0.99 and a coefficient of variation of 0.13 for R. 4.3.1 Soil plasticity Figure 2 confirms that the pile friction as described by Equation (9) is not significantly affected by the plasticity of clay. This conclusion was also drawn by Kolk & Van der Velde (1996). 4.3.2 Pile length Previous studies (Randolph & Murphy 1985; Kolk & Van der Velde 1996; Jardine et al. 2005) revealed that pile unit shaft fiction reduces in relation to pile length. This may be attributed to remoulding of clay and large soil strains during pile installation and subsequent pile loading.

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Figure 3. R versus pile length L.

Figure 6. R versus pile displacement ratio Ar .

Figure 4. R versus pile length L/D.

Figure 7. R versus time factor T .

Figure 5. R versus normalised cone resistance Qt .

4.3.5 Time between pile driving and loading Enough time should be allowed between pile driving and pile loading, to allow for dissipation of excess pore pressures and set-up of the soil. According to Randolph & Wroth (1979) 90% consolidation around solid driven piles occurs at times t between 5 D2 /c and 15 D2 /c, where c = coefficient of consolidation. Figure 7 shows a time factor T = ct/D2 > 5 to 15 for most tests, based on an empirical relation between c and LL according to c = 10−LL/40 , derived from NAVFAC (1981). No clear change of R with time is observed suggesting that both the predicted and the measured capacities in the database are representative for long-term capacity. It should be noted that this assessment is crude and therefore no design specific recommendations can be given with respect to set-up behaviour of piles in clay.

It was expected that pile slenderness L/D would affect pile capacity. Surprisingly, Figures 3 and 4 confirmed that pile frictional capacity as described by empirical Equation (9) was affected by pile length rather than by pile slenderness. 4.3.3 Overconsolidation Both API (2000) and Kolk & Van der Velde (1996) methods indicate that the pile friction to soil strength ratio, α, reduces as the OCR increases. Figure 5 confirms that the pile friction to cone resistance ratio, ks , decreases with increasing Qt as described by Equation (9). Here, Qt serves as measure of overconsolidation. 4.3.4 Soil displacement during installation It was explored if pile frictional capacity would be related to the amount of soil displaced by the penetrating pile. Pile tip displacement ratio Ar was taken as a measure for soil displacement. No direct relation between the pile displacement ratio and the measured shaft friction was found as can be observed in Figure 6. The new design method is applicable for both open ended and closed ended piles. © 2011 by Taylor & Francis Group, LLC

5 VALIDATION OF NEW DESIGN METHOD The entire database was used to validate the new method. Table 2 shows statistics for the ratio of predicted to measured pile capacity, R. Figure 8 presents results for the new design method. Table 2 and Figure 8 show improved predictions by the new CPT-based design compared to the selected existing design methods. Only 12% of the load tests is either under or over predicted by more than 20%.

6

LIMITATIONS

The proposed method is subject to limitations, most of which are common also to other design methods. Comments are as follows.

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The proposed design method considers unit shaft friction affected by overconsolidation and pile length, but not by load direction, soil plasticity, pile diameter and pile tip displacement ratio.

Table 2. The average and coefficient of variation of R for existing design methods and the new design method.

Average of R

Coefficient of Variation of R

Database:

Database:

Method

full

KV

full

KV

ACKNOWLEDGEMENTS

Bustamante Almeida CUR New method

0.55 0.87 0.98 1.00

0.52 0.86 1.03 0.99

0.19 0.19 0.22 0.14

0.17 0.22 0.20 0.13

The authors gratefully acknowledge Fugro’s commitment and support to improving geotechnical practice. REFERENCES

Figure 8. Measured versus predicted capacity for the new design method.

Database limitations include interpretational issues on pile failure load and representative soil profiles (Jeanjean et al. 2010). Measurement uncertainties also apply. Influence of soil sensitivity was not studied. The bearing capacity may be adversely affected for sensitive soils. Screening of CPT data will be necessary for intermediate soils where cone penetration may be partially drained, giving high qn values compared to undrained conditions. Pile loading may be undrained for intermediate soils.

7

CONCLUSIONS

A simple and reliable CPT-based design method is proposed for prediction of the axial capacity of long steel pipe friction piles installed in clay, typically used offshore. The method is based on regression analysis, exploring lowest coefficient of variation. It was initially developed from an existing pile load test database and subsequently verified using an augmented database. © 2011 by Taylor & Francis Group, LLC

APIAmerican Petroleum Institute 2000. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – Working Stress Design, API Recommended Practice 2A-WSD, 21st Edition. (With Errata and Supplement 1, December 2002, Errata and Supplement 2, October 2005, and Errata and Supplement 3, March 2008). Almeida, M.S.S., Danziger, F.A.B. & Lunne, T. 1996. Use of the piezocone test to predict the axial capacity of driven and jacked piles in clay, Canadian Geotechnical Journal 33 (1), 23–41. Bogard, D. & Matlock, H. (1998). Static and Cyclic Load Testing of a 30-inch-diameter Pile over a 2.5-year Period, Proceedings 30th Annual Offshore Technology Conference, 4–7 May, 1998, Houston, Texas, U.S.A, Vol. 1: 455–468 Bond, A.J. & Jardine, R.J. 1990. Research on the behaviour of displacement piles in overconsolidated clay.Technology Report OTH 89296, The imperial College London. BRE 1985. Comparison of piezocones in overconsolidated clays. Building Research Establishment and Norwegian Geotechnical Institute. Report No. 84223-1. Bustamante, M. & Gianeselli, L. 1982. Pile Bearing Capacity Predictions by Means of Static Penetrometer CPT, Proceedings of the Second European Symposium on Penetration Testing, ESOPT-II, Amsterdam, The Netherlands, Vol. 2: 493–500. Bustamante, M., Gianeselli, L., Mandolini, A. & Viggiani, C., 1994. Loading Tests on Slender Driven Piles in Clay, Proceedings Thirteenth International Conference on Soil Mechanics and Foundation Engineering, New Delhi, Vol. 2: pp. 685–688. Chen, B.S.Y. & Mayne, P.W. 1996. Statistical Relationships between Piezocone Measurements and Stress History of Clays, Canadian Geotechnical Journal, Vol. 33, No. 3: 488–498. Clarke, J., Rigden, W.J. & Senner, D.W.F. 1985 Reinterpretation of the West Sole Platform ‘WC’ Pile Load Tests, Géotechnique, Vol. 35, No. 4, pp. 393–412. CUR 2001. Bearing Capacity of Steel Pipe Piles,CUR-report 2001-8, July 2001, CUR, Gouda, Netherlands. Gibbs, C.E., McAuley, J., Mirza, U.A. & Cox, W.R. 1992. Reduction of Field Data and Interpretation of Results for Axial Load Tests of Two 762 mm Diameter Pipe Piles in Clays, Proceedings of the Conference ‘Recent Large-scale Fully Instrumented Pile Tests in Clay’: 285–345. Heerema, E.P. 1979. Pile Driving and Static Load Tests on Piles in Stiff Clay, in EleventhAnnual OffshoreTechnology Conference, Vol. 2: 1135–1145. ISSMGE 1999. International Reference Test Procedure for the Cone Penetration Test (CPT) and the Cone Penetration Test with Pore Pressure (CPTU): Proceedings of the Twelfth European Conference on Soil Mechanics and Geotechnical Engineering, Amsterdam, Netherlands,

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7–10 June 1999, Vol. 3, A.A. Balkema, Rotterdam: 2195–2222. Jardine, R., Chow, F.C., Overy, R.F. & Standing, J.R., 2005. ICP Design Methods for Driven Piles in Sands and Clays, Thomas Telford Ltd., London. Jeanjean, P. Watson, P.G., Kolk, H.J. and Lacasse, S., 2010. RP 2GEO: The New API Recommended Practice for Geotechnical Engineering, Proceedings Offshore Technology Conference, 3–6 May 1996, Houston, Texas, U.S.A Karlsrud, K., Hansen, S.B., Dyvik, R. & Kalsnes, B., 1993. NGI’s Pile Tests at Tilbrook and Pentre – Review of Testing Procedures and Results, Proceedings of the Conference ‘Recent Large-scale Fully Instrumented Pile Tests in Clay’: 405-429 Karlsrud, K., Kalsnes, B. & Nowacki, F., 1993. Response of Piles in Soft Clay and Silt Deposits to Static and Cyclic Axial Loading Based on Recent Instrumented Pile Load Tests, Advances in Underwater Technology, Ocean Science and Offshore Engineering, Vol. 28, pp. 549–583. Kirby, R.C. & Roussel, G. 1979. ESACC Project Field Model Pile Load Test, Hamilton Air Force Base Test Site Norato, California, Amoco Production Company. Kolk, H.J. and Van der Velde, E. 1996. A Reliable Method to Determine Friction Capacity of Driven Piles in Clay, Proceedings 28th Annual Offshore Technology Conference, 6–9 May 1996, Houston, Texas, U.S.A., Vol. 1: 337–346. Lambson, M.D., Clare, D.G. & Semple, R.M. 1992. Investigation and Interpretation of Pentre and Tilbrook Grange Soil Conditions, Proceedings of the Conference ‘Recent Large-scale Fully Instrumented Pile Tests in Clay’: 134– 196. Mayne, P. W. Kulhawy, F. H. and Kay, J.N. 1990. Observations on the development of pore-water stresses during piezocone penetration in clays, Canadian Geotechnical Journal, Vol. 27: 418–428. McClelland Engineers, inc. 1982. Geotechnical Investigation Borings 4, 5 & 6, Block 58, West Delta Area, Gulf of Mexico, Report No. 0181–0217. NAVFAC DM-71 1981. Soil Mechanics, Design Manual 7.1, Department of the Navy, Naval Facilities Engineering Command, 7.1–144. Norwegian Geotechnical Institute, 2006. Shear Strength Parameters Determined by In situ Tests for Deep Water Soft Soils, NGI Report No. 20041618-6. Olson, R.E. 1984. Analysis of Pile Response under Axial Loads, Final Report on Project 83–42B to API


O’Neill, M. W., Hawkins, R. A., & Mahar, L. J., 1981. Field Study of Pile Group Action. Final Report, Report No. FHWA/RD-81/002 and Appendix C, FHWA/RD-81/005. Pelletier, J.H. and Doyle, E.H. 1982. Tension Capacity in Silty Clays - Beta Pile Test, in Proceedings 2nd International Conference on Numerical Methods in Offshore Piling, April 29–30, 1982, Austin, Texas: 163–181. Price, G. & Wardle, I.F. 1982. A comparison between cone penetration test results and the performance of small diameter instrumented piles in stiff clay. Proceedings of the Second European Symposium on Penetration Testing, Amsterdam. Randolph, M.F. and Murphy, B.S. 1985. Shaft Capacity of Driven Piles in Clay, Proceedings 17th Annual Offshore Technology Conference, Vol. 1, 371–378. Randolph, M.F. & Wroth, C.P. 1979. An Analytical Solution for the Consolidation around a Driven Pile, International Journal for Numerical and Analytical Methods in Geomechanics, Vol. 3, No. 3: 217–229. Rigden, W.J., Pettit, J.J., St. John, H.D. & Poskitt, T.J. 1979. Developments in Piling for Offshore Structures, Proceedings of the Second International Conference on Behaviour of Offshore Structures, London: 1177–1182. Robertson, P.K. 1990. Soil Classification Using the Cone Penetration Test, Canadian Geotechnical Journal, Vol. 27, No. 1, 151–158. Robertson, P.K., Campanella R.G., Davies M.P. & Sy, A, 1988. Axial Capacity of Driven Piles in Deltaic Soils using CPT, Penetration Testing, 1988, ISOPT-1: 919–928 Semple, R.M. & Rigden, W.J. 1984. Shaft Capacity of Driven Pipe Piles in Clay, Proceedings, Symposium on Analysis and Design of Pile Foundations at ASCE National Convention, San Francisco, California, American Society of Civil Engineers, New York, 59–79. Totani, G., Marchetti, S., Calabrese, M. & Monaco, P. 1994. Field Studies of an Instrumented Full-scale Pile Driven in Clay, in Proceedings Thirteenth International Conference on Soil Mechanics and Foundation Engineering, New Delhi, 5–10 January 1994, Vol. 2, Oxford & IBH Publishing, New Delhi: 695–698. Togliani, G. 2008. Pile capacity predictions from in situ tests. Proceedings of the 3rd International Conference site Characterization, Taipei, Taiwan: 1187–1192. Wal, T. van der, Goedemoed, S., & Peuchen, J. 2010. Bias reduction on CPT-based correlations, Proceedings CPT10, Huntington Beach, California, May.

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7 Foundations for renewable energy



Evaluation of pile capacity approaches with respect to piles for wind energy foundations in the North Sea M. Achmus & M. Müller Institute of Soil Mechanics, Foundation Engineering and Waterpower Engineering, Leibniz University of Hannover, Germany

ABSTRACT: The design of axially loaded piles in non-cohesive soils is usually based on the β-approach. Recently, alternative CPT-based approaches have been developed by different research groups. However, in general the different approaches give different results, and it is not clear which approach is the best suited, i.e. most realistic, for special conditions. For the conditions of offshore wind energy foundations in the North Sea and with regard to tensile axial loading, a parametric study was carried out to ascertain the differences in pile capacity between the different approaches. It was found that in most cases the β-approach yields smaller pile capacities than the new CPT-based approaches. However, a calibration of the CPT-based methods for the considered conditions of piles with large diameter in dense to very dense sand has not yet been carried out. The new methods must therefore be applied with due caution until verified by experiments or sufficient experience. 1

INTRODUCTION

Offshore wind energy converters (OWECs) offer huge potential for the expansion of renewable energy in the North Sea in Germany. For water depths between 25 and 50 m tripod or jacket structures are suitable. These structures are supported by open steel pipe piles with diameters between 1.5 and 3 m. Under extreme loads induced by wind and wave loading, large axial tension forces have to be transferred to the ground, which in most cases is design-driving with regard to the required pile length. Since the erection of thousands of wind energy converters is planned for the North Sea, the optimization of the design and thus an accurate prediction of tensile pile capacities is very important. In the API regulations (API 2000, API 2007) usually applied in offshore engineering, the β-approach for calculating the axial pile resistance of offshore piles in non-cohesive soils is given. This method has been used successfully in offshore engineering for a long time. However, it is now known to possibly either underestimate or in some cases overestimate the actual pile capacity significantly. To overcome the disadvantages of the current design method, CPT-based approaches have been developed by different research groups. These empirical approaches are already included in current regulations for offshore structures. However, in general the different approaches give different results, and it is yet not clear which approach is the best suited, i.e. most realistic, for special boundary conditions. The soil conditions in the North Sea are mainly characterized by sandy soils, whose relative densities range from at least medium dense to dense and often very © 2011 by Taylor & Francis Group, LLC

dense. For these conditions a parametric study was carried out to ascertain the differences in tensile pile capacity between the different CPT-based approaches and the β-approach. The experimental data base used for the approaches was evaluated in order to identify the approach best suited to the particular geometric, loading and subsoil conditions considered here.

2

DESIGN METHODS

In water depths of more than 20 m, tripod or jacket foundation structures can be favourably used for the support of the wind tower. Usually open-ended steel pipe piles with a constant outer diameter between 1.5 m and 3.0 m are used to transfer the loads into the ground. Since the required depth of the driven piles in most cases depends on the tension loading case, only tensile pile capacities are considered in this paper. In general the tensile bearing capacity of piles consists of the pile’s weight and outer and inner skin friction. In the case of plugging, the latter is limited to the total weight of the soil plug inside the pile.

where fto = outer unit skin friction for tension; fti = inner unit skin friction for tension; Ao = outer pile shaft area; Ai = inner pile shaft area; Gs = effective steel weight of pile; and Gp = effective weight of soil plug inside the pile. The design values for skin friction f may be established either on the basis of test results or according to empirical approaches. In offshore technology pile

563

Toolan et al. (1990) published the results of pile tests which showed an overestimation of pile capacity in the case of long piles in loose to medium dense sand. Lehane (2005) reported that the β-method is conservative for relatively short offshore piles ( 0.3b, were carried out as well. For the 64 neighbouring Nysted turbine foundations the sliding criterion turned out to govern the design due to the possible existence of an upper, weak layer of remoulded soil on top of the intact soil. However, careful planning between Contractor and Designer of pit excavation, cleaning equipment and processes proved it possible to avoid adverse influence from a remoulded layer in the design criteria for the Rødsand 2 foundations. The foundation level with cu ≥ 250 kPa was achieved by excavation using a backhoe with a tolerance of ±0.2 m. After excavation the foundation pit was cleaned for soft material using an air lift mounted on a screeding frame and guided by diving assistance. This method results in an uneven surface with only tens of millimetres of softer material in pockets on top of the strong, intact soil. This condition was verified by multi-beam survey, CPTs as well as diver’s knife tests and video sweeping of the pit immediately before placing the gravel bed material. A survey demonstrating the unevenness of the excavated bottom is shown in Figure 5. Figure 6 illustrates schematically the potential small amounts of settling suspended sediments and local pockets of remoulded/disturbed clay till which will be

penetrated by the gravel layer. The gravel has a mean grain size of about 20 mm and a maximum grain size of 90 mm. In principle, sliding surfaces could pass through the intact clay till, the gravel bed, the interface between concrete and gravel bed or through a combined surface of gravel bed and intact clay till. The characteristic intact clay till strength at excavation level is in excess of cu = 250 kPa, corresponding to a sliding capacity of Hd = cu,d · A where cu,d is the undrained design shear strength and A is the effective sliding area. For the gravel and the concrete/gravel interface the sliding capacity is Hd = Vd · tan δd , where δd is the design interface friction angle, and Vd is the vertical load. All other things considered equal the undrained shear strength increases as a result of the vertical loading by the wind turbine. Due to the initial high value of cu , sliding surfaces passing through clay till or clay till/gravel will have a significantly higher capacity compared with surfaces in the gravel and concrete/gravel interface. The non-existence of a soft soil layer was verified by the execution of five shallow CPTs immediately before placing 0.5 m thick layer of gravel bed material. The interpretation of these required due consideration of the near-surface measurement issues like cracking of the surface soil surrounding the penetrating cone, influence of geometrical conditions of the probe as well as the tip resistance depending on the vicinity to the surface. In this way strict success criteria


589

for the measurements in the upper decimetres were defined to ensure the anticipated soil strength based on a conservative cone factor of 15. This value is higher than the commonly adopted value of 10 for Danish clay till. However, the higher value was chosen in agreement with the Owner and DNV to avoid lengthy testing to accommodate effects of cyclic degradation and because all general experience has demonstrated, that correlation between CPT and triaxial strengths in Danish clay till is difficult. Potential pockets of soft, remoulded material relative to competent clay till were easily identified by a tip resistance < 0.1 MPa, and removed. 3.2

Soil stiffness

Since the turbine loads and the natural frequencies of the turbine and foundation are mutually dependent, the design loads as well as the foundation geometry and properties are determined iteratively between the turbine supplier and foundation designer. Hence, in order to determine the natural frequencies of the turbine and foundation structures, the dynamic foundation stiffness, i.e. the dynamic structural stiffness and the dynamic soil stiffness combined, was determined. The dynamic soil stiffness is directly linked to the dynamic shear modulus of the soil, which in turn is highly dependent on the shear strains induced in the soil. The stiffness at very small strain levels is typically many times higher than the stiffness at large strains. The clay till at the wind farm site is at places so hard that it is not possible to confidently indicate a maximum strength and stiffness of the soil. Hence, only a lower bound value of the soil stiffness was determined using the structural stiffness as the upper limit of the foundation/structure stiffness range. For dynamic shear strains in the range of γ ∼ 0.1%, a lower bound value of the dynamic shear moduγ=0.1% lus was determined as Gdyn = 165cu in accordance with instructions in the Owner’s Design Basis. For the majority of the locations where the soil profile consists of clay till only, the linear-elastic method of George Gazetas, as tabulated in DNV (2007), was used to determine the soil stiffness based on a constant value of the dynamic soil modulus. To demonstrate the magnitude of the induced shear strains, finite element models were employed, i.e. Plaxis 2D v9 with soil model parameters calibrated towards results of the triaxial testing of the soil as well as previous plate loading tests on similar gravel bed material. The stiffness for first time loading as well as for reloading of the gravel bed material were determined from plate loading tests on uncompacted and compacted material, respectively. 3.3 Settlement and tilt The maximum tilt of the wind turbine shall be less than 0.5◦ in accordance with Owner’s Design Basis. This includes an installation tolerance of 0.25◦ . The © 2011 by Taylor & Francis Group, LLC

differential settlement and tilt due to the soil behaviour originate from inevitable variations in the soil below the foundation base plate as well as influence from the prevailing wind direction. The differential settlement originating from variations in the soil was determined dependent on the vertical settlement. It was based on observations of the relationship between the foundation width, vertical settlement and tilt of offshore foundations. Because of the high overconsolidation of the material, the initial and consolidation contributions of the total vertical settlement were determined in accordance with a classic linear elastic calculation based on tangent values of the constrained modulus derived from present oedometer tests. The results of these were found to be in line with general Danish clay till experience. The creep settlements were estimated in accordance with experiences retrieved from extended investigations of Danish clay till conducted by Kristensen et al. (1995), where creep settlements for piers were observed up to a value of 20% of the sum of the initial and consolidation settlements per log cycle of time. For a consolidation time of 1 month, the creep amounted to some 40–50% of the vertical settlement during the lifetime of 25 years. The differential settlement due to a prevailing wind direction was calculated for two cases: A permanent overturning moment corresponding to 0.25Mmax , where Mmax is the maximum unfactored overturning moment in the ultimate limit state, as well as for a case with a temporary, short time, fixed high value of 27 MNm in accordance with Owner’s Design Basis. Settlements and differential settlements proved not to be a design issue.

4

DESIGN AT LOCATIONS WITH LAYERED SOIL

At around 10% of the foundation positions, the soil profile proved to be non-homogeneous, i.e. layered soil consisting of both young and old meltwater deposits as well as Tertiary clay and chalk interbedded in clay till. Furthermore, the soil strength was found to be significantly decreasing with depth at some locations. The common bearing capacity formula equations were not suitable for the layered soil conditions, since introduction of very conservative assumptions would be necessary. Hence, both two- and three-dimensional numerical modelling was applied in the geotechnical design in these situations. These analyses are exemplified in the following by the design process of the foundation at position M14, cf. Figure 1. The soil profile at this location consists of clay till underlain by varying low strength meltwater deposits and low strength clay till, as listed in Table 1. The effective strength parameters turned out to be governing for the design. Where non-horizontal layering was found, the design was carried out assuming worst-case conditions.

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Table 1.

Design soil profile at foundation location M14. Top level [m] ϕ [◦ ] c [kPa] cu [kPa]

Clay till Clay, rather fat Clay till Sand Clay till

−12.5 −15.6 −17.5 −19.0 −22.0

33 17 30 31.5 32

25 4.5 7.5 0 21

250 45 75 – 210

Figure 8. Abaqus calculation model. Left: Entire domain with soil layering. Right: Meshed foundation base plate.

would be strained unacceptably in case of a redesign and construction of a revised foundation geometry, a full three-dimensional modelling was established as non-3D calculation methods were assumed to underestimate the bearing capacity.

4.2 Three-dimensional modelling

Figure 7. Meshed 2D calculation domain showing deformation at failure. The eccentrically loaded foundation is equalised with a centrally loaded effective area.

4.1 Plane strain modelling The finite element software Plaxis 2D v9 was employed based on a plane strain assumption. The intension with the analyses was to reflect an extension of the code-based bearing capacity formula calculations to include soil layering. Thus, the analyses inherently use the same simplifications as the bearing capacity formula equations, e.g.: 1. The moment loading on the foundation is equalized with a centrally loaded effective area of a strip foundation. 2. The torsional moment is applied as an equivalent horizontal force in accordance with DNV (2007). The Mohr-Coulomb material model with characteristic (unfactored) soil parameters was employed for the failure analysis and the safety against failure was calculated by means of a ϕ/c reduction calculation step. Calibration tests were carried out on models with non-layered soil conditions, and an accurate match with results of the bearing capacity formula was found. Thereafter, an extension of the finite element analyses to layered soil conditions was trivial. Figure 7 illustrates the deformed calculation domain for location M14. For nearly all wind turbine locations with layered soil profiles, sufficient safety against failure was verified by the plane strain calculations. Thus, the level of safety was similar to the locations on uniform soil conditions designed in accordance with the bearing capacity formula. However, for the above location M14, the material factor of safety was calculated as Msf = 1.05, i.e. less than the code requirement of γm = 1.15. Since the time schedule and economy © 2011 by Taylor & Francis Group, LLC

A proper modelling of the soil-structure interaction was evaluated to be of highest importance in this analysis. The correct 3D octagonal foundation base plate geometry was modelled with the general-purpose finite element software Abaqus v.6.9-1 as illustrated in Figure 8. The circular hole in the foundation base plate centre models the circular recess in the gravel bed. The foundation was modelled as a rigid body, and the soil was modelled with the Mohr-Coulomb material model and second-order hexahedral solid elements. The torsional moment was applied directly. The analysis was carried out by establishing the initial stress state, applying the foundation loads, and thereafter reducing the soil strength parameters by increasing the material factors until failure was fully developed. The analysis showed an increase in the material factor of safety to Msf = 1.40 compared to Msf = 1.05 as determined by the plane strain analysis, thus verifying sufficient bearing capacity. The increased bearing capacity is attributed to proper modelling of the contact area under the foundation, direct modelling of the torsional loading, and development of a three-dimensional failure figure. In the plane strain analysis, the plastic stress distribution under the effective foundation area is assumed to be uniform, with the failure condition fulfilled everywhere in the area. However, in the 3D total area analysis, it is not necessary to adopt this assumption. Furthermore, the contact area between soil and structure is allowed to be more widespread than in the effective area approach. In the present analysis of location M14, the contact area has been found to be approximately 50% larger than the corresponding effective area. The failure figure and the contact area are depicted in Figure 9, where the failure figure is found to be limited at depths by the interbedded sand layer. A significant amount of the drastic increase in the material factor of safety is attributed to the specific soil conditions at the actual site, with a stronger layer overlying weaker deposits. Thus, the failure will have

591

Figure 9. Left: Magnitude of plastic strain indicated under failure at location M14. Right: Contact area under ULS loading.

Figure 10. A view from the construction site.

a character of punch-through, and the difference from a plane strain analysis with an effective area approach to a three-dimensional model will be increased. In this specific case, the increase in material factor of safety is approximately 35%. Comparative calculations have been carried out with the actual foundation placed on a uniform soil profile. The corresponding increase in material factor of safety for this case was calculated to around 20% depending on failure state (drained/undrained) and the actual material strength properties. 5

of load combinations, and specific advanced numerical modelling. Cast foundations are depicted in Figure 10. The geotechnical design has benefitted fruitfully from the careful planning and knowledge sharing between project parties, resulting in an optimized design already before the early commencement of activities on site. Designer’s incorporation of the contractor’s experience with the offshore equipment and working processes made it possible to close previously unsolved issues with regards to the sliding capacity. Thus, an upper remoulded layer was demonstrated not to exist with the planned excavation and cleaning processes. Furthermore, specific purpose-planned supplementary investigations with triaxial testing made it possible to re-evaluate and increase the original design soil properties with considerable ballast saving as a result. At locations with layered soil conditions, twodimensional plane strain analyses with Plaxis provided a straightforward extension of the normal, well winnowed design scheme, whereas the establishment of an Abaqus model allowed a three-dimensional modelling of the foundation geometry, loading and stress states.The three-dimensional model was found to yield minor to major increases of the bearing capacity compared with the plane strain approach, depending on the actual conditions in question. In the present case with location M14, the establishment of an advanced model allowed the generic geometry of the foundation to remain unchanged, with significant savings both for the design and construction processes as outcome. ACKNOWLEDGEMENT The authors gratefully acknowledge the permission by Owner E.ON Vind Sverige and Contractor Per Aarsleff – Bilfinger Berger Joint Venture to publish the paper.

CONCLUDING REMARKS

The Rødsand 2 project was considered an important show case in connection with the UN COP15 meeting on climate change in Copenhagen, December 2009. Thus, the foundation design process and the foundation production had to meet a very strict deadline. This was achieved with a practical and economical design due to a combination of simple, robust bearing capacity formula design schemes to handle the multitude


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REFERENCES Kristensen, P.S., Regtop, J. & Balstrup, T. 1995. Predicted and observed settlements and tilts of offshore bridge piers. DGF-Bulletin11. Proceedings XI ECSMFE, Copenhagen. Det Norske Veritas. 2007. Offshore standard DNV-OS-J101. Design of offshore wind turbine structures. DNV.


Geotechnics for developing offshore renewable energy in the US M. Landon Maynard University of Maine, Orono, Maine, USA

J.A. Schneider University of Wisconsin-Madison, Madison, Wisconsin, USA

ABSTRACT: With climate change and energy security issues being of concern globally, interest in renewable energy generation has greatly increased in the US, particularly with the 2005 Energy Policy Act. A major focus has been on development of offshore ocean and wind force energy conversion technologies. However, a disconnect exists between technology designers and developers and the operational environment. Little attention has been given to seabed-foundation interaction, with the exception of offshore wind, and site investigation, soil properties, and foundations are largely ignored until final development stages. Designs that incorporate mooring and foundation system responses with metocean conditions will likely lead to increased efficiency of power production and cost. This paper identifies challenges and recommendations for offshore geotechnics of developing renewable energy facilities, including costs, site investigations and foundations. 1

INTRODUCTION

The development of commercial scale renewable energy sources is growing in interest within the United States (US). Provisions within the 2005 Energy Policy Act increased funding for research and development of clean, renewable technologies such as wind, tidal, and wave energy. Since this time, government agencies such as the National Renewable Energy Laboratory (NREL) and Minerals Management Service (MMS), among others, began publishing wind and ocean energy resource documents, which have served to target technologies and locations for renewable energy development (wind: Musial et al. 2006; current: MMS 2006; wave: Bedard et al. 2005; tidal: Bedard et al. 2006; Musial 2008). Table 1 summarizes the extractable US offshore energy potential for each of the above mentioned energy resources. The estimates for wind include a 60% area of exclusion and 40% efficiency. While much attention has been given to offshore wind in the US in the last decade, particularly to Cape Table 1. Resource comparison of extractable US offshore renewable energy (after Musial 2008).

Energy Source

Extractible Capacity (GW)

Deep water offshore wind (>30 m) Shallow water offshore wind ( 1), even in uniform soil, are not currently available. Details of geotechnical design considerations are given in a companion paper (Palix et al. 2010).

2

FINITE ELEMENT ANALYSIS

2.1 HARMONY Proprietary program HARMONY computes the response of axisymmetric bodies subjected to nonaxisymmetric (horizontal, moment and twist) load. The soil material model is linear elastic – perfectly plastic (Mohr-Coulomb). By using Fourier expansions in the circumferential (θ) direction, the full 3D problem is solved with a quasi 2D analysis. This leads to less cost and time than for a comparable 3D analysis. Since its initial development (Griffiths 1985), HARMONY analyses have been made of various soilstructure interaction problems (e.g. Kolk et al. 2001). 2.2 Generic mesh design Eighteen cases (6 caisson embedment ratios L/D × 3 soil undrained shear strength su profiles) were analysed. The caisson diameter D was always 5 m, with embedded lengths L of 7.5, 10, 15, 20, 25 and 30 m. The lowest L/D value (1.5) is no longer a shallow foundation (L/D up to 1) and avoids possible problems with internal scoop failure. L/D ≈ 6 represents a practical © 2011 by Taylor & Francis Group, LLC

Figure 1. Caisson geometry, HM load sign convention and soil su profiles – Constant, Normally Consolidated and Stepped.

upper limit for suction installation in normally consolidated clay. To cover offshore conditions, “constant”, “normally consolidated”, and “stepped” su profiles were taken, with ez,su /L values of 1/2, 2/3 and 3/4 (Fig. 1). For consistency, each case used the same mesh, and mesh geometry was related to L and D. A 16 × 16 axisymmetric rectangular mesh of 8noded quadrilateral elements using “reduced” (2 × 2) integration was used. An inner (15 × 15) mesh of finite elements was surrounded by mapped infinite elements (Marques & Owen 1984). In the r-z plane, 8 rows × 5 columns of elements modelled the plug inside the caisson. The caisson had a top plate. In the radial direction, the caisson wall was surrounded on the outside by a “thin” steelsoil interface zone D/500 (10 mm) thick. The first column of soil outside the caisson wall was always

741

D/20 (250 mm) thick. Vertically, element rows immediately above and below the caisson tip were always D/20 (250 mm) and D/50 (100 mm) thick. The finite mesh extent was always 2L deep and 2L wide radially. Infinite mesh boundaries were at 6L depth and 6L radius. Material data included: (i) infinite tensile strength, (ii) constant soil su per element row, (iii) soil Eu /su = 500, νu = 0.49, (iv) E = 210e9 kPa for both plug and caisson to provide a rigid response, (iv) to give the correct α value, the outer interface zone had Eu and su values equal to 0.65 that of the adjacent soil, (v) soil submerged unit weight was 6.7 kN/m3 and Ko,total was unity. All analyses applied prescribed displacements to the caisson head. This facilitated “failure” definition (after 40 equal steps, the maximum displacement somewhere on the caisson was D/10) and allowed direct assessment of ez (=H load vertical eccentricity) from seafloor M and H reactions. Three harmonics were used (0, 1 and 3). The number of freedoms was 2304. Stresses were checked for yield at 30◦ intervals in the circumferential (θ) direction. 2.3 HARMONY verification To verify the performance of the “thin” interface elements, a laterally loaded disk model was analysed by extracting a single row of elements from the generic mesh: 8 finite elements modelled the soil. HARMONY Np,fixed errors were less than 0.5% of the analytical (Randolph & Houlsby 1984) solution when α = 0.65. To verify VHM loading, two laterally loaded caisson 3D FE benchmarks reported by Andersen et al. (2005) were simulated using the generic mesh. HARMONY agrees well with ABAQUS and BIFURC (Fig. 2). Extreme checks for axial capacity Vmax included a first run with “zero su ” soil below the caisson, and recovering the correct outer friction resistance Fo = α su,av,L π D L. Then, to check superposition, a second run with α = 0 gave the caisson end-bearing resistance component Qtip . 3

MH ELLIPSES AT V = 0

For each caisson L/D and soil su profile, on average 35 HARMONY analyses were made. Each applied a different displacement/rotation probe δx , θxz . “Fixed head” and “free head” conditions were included to establish four key data points Hmax , Ho , Mo and – Mmax (Fig. 3a). For caissons with L/D > 1, appropriate non-dimensional loads are M* = M/(D L2 su,av,L ) and H* = H/(D L su,av,L ). Typical M*H* load paths are given in Figure 3b. Pre-displacing the caisson to 0.9 |Hmax | minimised numerical problems associated with locating data points around -Mmax . At final displacement D/10, all “load-settlement” curves had flattened out. Even though the finite element method is a slight upper bound, it was assumed that final points lay on the resistance envelope. Mirroring was used (i.e. final point M∗ , H∗ gave –M∗ , –H∗ ). © 2011 by Taylor & Francis Group, LLC

Figure 2. HARMONY verification – laterally loaded anchor piles in NC soil su = 1.25z (Andersen et al. 2005) (a) caisson L/D = 1.5, Wsub = 300 kN (b) L/D = 5, Wsub = 1100 kN.

Figure 4 shows that M∗ H∗ final points lie on ellipses inclined at angles almost equal to ez,su /L. Rotated ellipses have the parametric form:

To find MH (rotation angle) and aMH , bMH (major and minor semi-axes’ lengths), least squares minimisation was used. Trials with differing numbers of data points and constraints showed that the best overall fits (highest correlation coefficient (Pearson’s r) R2 and lowest coefficient of variation) for all 35+ data points were obtained by simply using (M∗ , H∗ ) co-ordinates of 3 key data points Hmax , Ho and Mo , and a vertical gradient at Hmax . This fit procedure was applied to all 18 cases. Values of MH aMH , and bMH for the above 18 fits can be derived from su , Ho and Hmax using:

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Figure 4. MH ellipses at V = 0: summary final data points, caisson D = 5 m, L/D = 1.5, 2, 3, 4, 5 and 6. Soil su profiles: (a),(b) Constant, ez,su /L = 1/2, (c),(d) NC, ez,su /L = 2/3 and (e),(f) Stepped, ez,su /L = 3/4.

4 V-HMAX ELLIPSOIDS Figure 3. MH ellipses at V = 0: (a) 4 key points Hmax , H0 , M0 and -Mmax (b) load paths and final data points, case L/D = 5, NC soil.

Correction parameter MH is due to the ellipse semi-major axis lying slightly below point Hmax . Figure 5 shows data points and fitted ellipses for 6 of the 18 cases. Overall fit quality is good. As can be seen on the insets, there is a slight deterioration for L/D = 5 near Hmax : data points are not equidistant from the ellipse major semi-axis, have a flat top surface, and a general shape resembling a “World War I tank” originally noted by Poulos & Davis (1980). Figure 6 is a key figure, plotting lateral bearing capacity factors Np,fixed and Np,free versus L/D. Trends of Np for both uniform and NC clay agree well with those presented by Randolph et al. (1998). At L/D ≈ 6, all Np values flatten out, and Np,fixed values are close to the Randolph & Houlsby (1984) value of 11.2 for L/D = ∞ and α = 0.65. Differences are due to caisson base shear/moment contributions and “shallow wedge” failure mechanisms close to seafloor. Before checking that MH resistance envelopes do not change shape when V = 0, it is first necessary to establish their shape in the V-Hmax plane. © 2011 by Taylor & Francis Group, LLC

For caissons with L/D values of 1.5, 3 and 5, Supachawarote et al. (2004) showed that V-Hmax resistance envelopes for NC clay were ellipsoidal and could be fitted with the equation

The above work was repeated and extended using HARMONY. For all 18 cases, twelve δx , δz “fixed head” (θxz = 0) displacement probes were made. Vmax and Hmax were obtained when δx = 0 and δz = 0 respectively. Envelopes remained ellipsoidal. Figure 7 shows resulting best fit aVH and bVH values. For the NC profile, good agreement was obtained between HARMONY and Supachawarote et al. (2004). Values of aVH and bVH are similar for both “NC” and “stepped” su profiles. In addition, for “constant” su profiles, the corresponding recommended simplified equations are:

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Figure 7. V-Hmax plane: ellipsoid best fit parameters aVH and bVH versus L/D.

Figure 5. MH ellipses at V = 0: typical fits; caisson L/D = 1.5 and 5, Constant, NC and Stepped soil su profiles.

Figure 6. MH ellipse at V = 0: Np versus L/D (α = 0.65). H reference point location: caisson head.

Increased aVH and decreased bVH values are due to more competent soil near seafloor reducing interaction between V and H. 5

MH ELLIPSES AT V = 0

Applying V load decreases caisson capacity. To check that MH ellipses had similar shapes, six cases (caisson L/D = 1.5 and 5, each with 3 soil su profiles) were analysed for three additional load levels (V/Vmax = 0.6, 0.8 and 0.9). Vmax values had been established in Section 4 above. Figure 8a compares typical MH ellipses. Rotation angle MH is again essentially constant.

Figure 8. MH ellipses at V = 0: Typical V/Vmax results; L/D = 1.5, NC soil (a) “as-is” (b) same major semi-axis lengths aMH .


Figure 8b re-plots the ellipses all with the same aMH value. It is seen that the corresponding bMH decreases are modest. Therefore, it is reasonable to assume that, for non-zero V load, ellipse shape ratio aMH /bMH remain essentially unaltered: M and H reduce equally.

Figure 10. Example: effect of anchor load depth on caisson capacity in NC soil su = 1.25z, V/H = tan(30◦ ), L/D = 1.5 and 5.

8

DISCUSSION/ASSUMPTIONS

Assumptions in the above work included:

Figure 9. ellipse/ellipsoidal VHM envelope. VHM reference point location: caisson head.

Hence MH ellipses at V = 0 can be derived from MH ellipses at V = 0, plus load V and resistances Vmax and Hmax . 6 VHM ENVELOPE AND DESIGN EQUATIONS The complete VHM resistance envelope is given by rotated ellipses in the MH plane (Eq. 9) plus ellipsoids in the V – Hmax plane (Eq. 10). Factor Hmax,V /Hmax accounts for non-zero V load.

Figure 9 gives the resulting “tongue”-shaped VMH envelope. 7

DESIGN EXAMPLE

Anchor pile capacity is sensitive to lug level depth and chain/wire load angle. Supachawarote et al. (2004) and VHM equation results are compared on Figure 10. Good agreement for optimum depth (ez /L = 0.7) was obtained. For depths above optimum, the VHM equations slightly underestimate capacity decrease at L/D = 5. This is because the FE final data points are unsymmetrical (e.g. Fig. 5d). However, for depths below optimum, agreement is good, even for L/D = 5. This zone, where restoring rotation occurs, is of interest to suction pile anchor design.


i ii iii iv v vi vii viii

zero caisson twist/tilt suction possible/no cracks unique soil shear strength su rigid caisson D = 5 m and 1.5 ≤ L/D ≤ 6 caisson top at seafloor clay-steel α = 0.65 MH ellipse symmetrical about semi-major axis Vmax and Hmax derived from FE analyses (not simplified analytical equations).

Ad (i) through (iii) these imply a symmetrical VHM envelope (e.g. identical V in tension and compression, no reduced reverse end-bearing). Ad (iv) use with caution for shallow foundations, but with confidence L/D > 6. Minor differences can occur for non-rigid caissons and/or other D values. Ad (v) provided that su,av,L is evaluated over the caisson embedded length, equations can also be used to assess torpedo pile capacity – pile head embedment has a small effect on Np,fixed at high L/D. Ad (vi) Randolph & Houlsby (1984) give guidance on Np,fixed variation with α for infinitely long caissons. All geotechnical foundation models are inaccurate to a certain degree. The resistance model described above is no exception, even though the basic building blocks (Randolph et al. 1998, Supachawarote et al. 2004) have been improved. Should assumptions be violated (for example reduced end-bearing or tensile gapping), it is also reasonable to consider this ellipse/ellipsoidal failure envelope for this modified situation without a significant decrease in model accuracy. 9

CONCLUSIONS

HARMONY is reliable and fast: a quasi 3D approach permitted over 1500 analyses in a rigorous and consistent manner. Based on the results in this paper, a VHM resistance envelope for caissons in clay may be reasonably approximated by three equations defining its ellipse/ellipsoidal shape. Envelope parameter values are functions of caisson geometry and soil shear strength profile. The equations obviate the need for

non-linear 3D FE analyses (except for assessing responses or soil reactions), and facilitate probabilistic and optimisation analyses.

10

NOTATION

su,z = soil undrained shear strength at any depth z su,av,L = average su,z between caisson head (seafloor) and caisson tip (L) L = 0 su,z dz/L t = ellipse parameter (0 < t < 2π) V = vertical load at caisson head (seafloor) Vmax = caisson vertical resistance (H & M = 0) Wsub = caisson submerged weight z = depth below seafloor

= clay-steel outer adhesion factor = caisson head lateral displacement = caisson head vertical displacement = caisson rotation REFERENCES = load factor = ellipse rotation angle Andersen, K.H., Murff, J.D., Randolph, M.F, Clukey, E.C., = soil undrained Poisson’s ratio Erbrich, C.T., Jostad, H.P., Jansen, B., Aubeny, C., = ellipse major semi-axis length Sharma, P. & Supachawarote, C. 2005. Suction Anchors = ellipse minor semi-axis length for Deepwater Applications, in Frontiers in Offshore = ellipsoid parameter Geotechnics ISFOG 2005: Proceedings of the First International Symposium on Frontiers in Offshore Geotech= ellipsoid parameter nics, Perth, Australia, Taylor & Francis, London, pp. 3–30. = caisson outer diameter Kolk, H.J., Kay, S., Kirstein, A. & Troestler, H. 2001. North = soil undrained Young’s Modulus Nemba Flare Bucket Foundations, Offshore Technology = H load vertical eccentricity with respect to Conference, Houston, Texas, U.S.A., OTC Paper 13057. caisson head (seafloor) Griffiths, D.V. 1985. HARMONY – A program for pre= M/H dicting the elasto-plastic response of axisymmetric bodez,su = analytical ez based on su,z ies subjected to non-axisymmetric loading, Report to L L Fugro-McClelland Engineers B.V., Simon Engineering = 0 su,z zdz/ 0 su,z dz Laboratories, University of Manchester, U.K. ez,Hmax = ez based on M/H at Hmax Marques, J.M.M.C. & Owen, D.R.J. 1984. Infinite EleFo = caisson outer skin friction resistance ments in Quasi-static Materially Nonlinear Problems, H = horizontal load at caisson head (seafloor) Computers and Structures, Vol. 18, No. 4, pp. 739–751. = non-dimensional H value H∗ Randolph, M.F. & Houlsby, G.T. 1984. The Limiting Pres= H/(D L su,av,L ) sure on a Circular Pile Loaded Laterally in Cohesive Soil, Hmax = caisson maximum “fixed head” horizontal Géotechnique, Vol. 34, No. 4, pp. 613–623. resistance (V = 0) Palix, E., Kay, S. & Willems, T. 2010. Caisson Capacity = Np,fixed L D su,av,L in Clay: VHM Resistance Envelope – Part1: 3D FEM Numerical Study, in Frontiers in Offshore Geotechnics Hmax,V = as Hmax , but with V = 0 ISFOG 2010: Proceedings of the Second International Ho = “free head” horizontal resistance (V&M = 0) Symposium on Frontiers in Offshore Geotechnics, 8-10 = Np,free L D su,av,L November, 2010, Perth, Australia. L = caisson embedded length Poulos, H.G. & Davis, E.H. 1980. Pile Foundation Analysis M = moment load at caisson head (seafloor) and Design, John Wiley and Sons, New York, Series in = non-dimensional M value M∗ Geotechnical Engineering, Chapter 7. 2 = M/(D L su,av,L ) Randolph, M.F., O’Neill, M.P., Stewart, D.P. & Erbrich, C. Mmax = caisson maximum moment resistance 1998. Performance of Suction Anchors in Fine-grained Calcareous Soils, Offshore Technology Conference, Hous(V = 0) ton, Texas, U.S.A., OTC Paper 8831. = caisson moment resistance (V & H = 0) Mo Supachawarote, C., Randolph, M.F. & Gourvenec, S. 2004. Np,fixed = “fixed head” lateral bearing capacity Inclined Pull-out Capacity of Suction Caissons, Proc. factor Fourteenth Int. Offshore and Polar Engineering Con= Hmax /(D L su,av,L ) ference ISOPE, Toulon, France, International Society of Np,free = “free head” lateral bearing capacity Offshore and Polar Engineers (ISOPE), Cupertino, pp. factor 500–506. = Ho /(D L su,av,L ) α δx δz θxz λL MH νu aMH bMH aVH bVH D Eu ez



Installation and in-place assessment of drag anchors in carbonate soil M.P. O’Neill, S.R. Neubecker & C.T. Erbrich Advanced Geomechanics, Perth, Western Australia

ABSTRACT: The design of drag anchors as fixed moorings for offshore oil and gas facilities requires a detailed level of understanding of anchor behaviour in different soil types. An adequate level of safety in the system needs to be ensured, but also the anchor point must be prevented from moving during the design storm loads. In carbonate soils, factors such as sensitivity, cyclic loading and consolidation will heavily influence the level of preload required to ensure no anchor movement in service. These considerations are discussed in detail in this paper. 1

INTRODUCTION

Drag anchor installation involves preloading the anchor and chain system to a specified level in order to ensure sufficient capacity (or no movement) under the design loads. However, it is acknowledged that the in-place capacity of a drag anchor in cohesive sediments will differ from the preload applied to it. This is due to the combined effects of consolidation and cyclic loading. With fine grained carbonate soils, the difference between installation and in-place anchor capacities is amplified by the very high sensitivity they can exhibit and their generally strong susceptibility to degradation under cyclic loading. Following on from Neubecker et al. (2005), this paper describes simple anchor relationships to take account of the effect of soil sensitivity on the in-place capacity. The effect of consolidation and cyclic loading are also examined.

2

Figure 1. Anchor force system for N-R model.

where f is the anchor form factor, Ap is the total projected anchor area in the direction of travel, Nc is a bearing capacity factor (taken as 9), and su−m is the local monotonic undrained shear strength. The anchor resistance normal to the direction of travel, Tn , is defined as:

N-R ANCHOR EMBEDMENT MODEL

The drag anchor model described by Neubecker & Randolph (1996) (the ‘N-R model’) is a relatively simple method for predicting drag anchor behaviour in cohesive materials. The basis of the method is the assumption that two fundamental anchor parameters, namely the anchor form factor, f, and the resultant angle, θw , are unique to each anchor type and describe the behaviour of any sized anchor of that type in any cohesive soil strength profile. The method also assumes that a drag anchor travels along a plane parallel to its flukes inclined at an angle, β, to the horizontal (see Figure 1). The anchor resistance acting in the direction of travel, Tp , can be calculated at any depth as:

where θw is the inclination relative to the direction of travel of the total soil resistance force, Tw , which is the resultant of Tp and Tn (see Figure 1). The buoyant anchor weight, Wa , is then combined with Tw to form the resultant, Ta , which is taken as the anchor holding capacity orientated at the angle, θa , of the chain to the horizontal. The magnitude of θa may be determined using closed form solutions also developed by Neubecker & Randolph (1996) for embedded anchor chain behaviour in cohesive soils. The fundamental anchor parameters, f and θw , may be determined via calibration against field or model centrifuge test data, calibration against existing drag anchor simulation software or through analysis using the results of detailed finite element modelling. For this current discussion, the Vryhof Stevpris (Vryhof,


The parameter Nc−DNV is the bearing capacity factor recommended by DNV (2000), which for this discussion has been taken as 12.5. The bearing capacity factors calculated using this simple equation are virtually identical to those derived using the equations presented in Zhou & Randolph (2009) for a T-bar, assuming no rate effect and with their strain softening parameter, ξ95 = 15, which is a reasonable value for many carbonate soils. Hence, during anchor penetration, Equation 3a and Equation 4a (describing the bearing and frictional resistances on the fluke respectively) can be re-written as: Figure 2. Anchor force system for component model.

2010) with a 50◦ fluke shank setting (typical for use in cohesive sediments) has been adopted as the ‘base case’ anchor type. The results of Stevpris model drag anchor centrifuge tests conducted in Speswhite Kaolin soil and presented in O’Neill (2000) indicate best-fit f and θw values of 1.4 and 35◦ respectively, which are adopted as the reference parameters herein. 3 ANCHOR FORCE COMPONENT MODEL An anchor force component model has been developed following the DNV (2000) general methodology. The soil resistance on the anchor projected in the direction of travel can be divided into four components, namely friction on the fluke, RFF , bearing resistance on the fluke, RFB , friction on the shank, RSF , and bearing resistance on the shank, RSB (see Figure 2). The bearing resistance components may be calculated as:

where AFB and ASB are the projected areas of the fluke and shank bearing components respectively in the direction of travel. The frictional resistance components may be calculated as:

where AFF and ASF are the friction areas of the fluke and shank respectively, α is an adhesion factor taken as 1/St and St is the soil sensitivity. Recent advances suggest that the bearing capacity factor, Nc , is strongly dependent on St . As a simple model, it can be considered that the bearing resistance on an anchor element can be divided into two components, namely one acting in virgin (i.e. intact) ground, and one acting in disturbed ground (i.e. in the ‘back-flow’ region) as the anchor penetrates through the soil. Hence, for the anchor force component model the value of Nc may be calculated as follows:


Equation 3b and Equation 4b (describing the bearing and frictional resistances on the shank respectively) can be re-written in the same way. With regards to the Vryhof Stevpris model anchor tests reported in O’Neill (2000), it has been assumed that the Speswhite Kaolin soil used for the tests has a sensitivity of 2. It then follows that the N-R model fundamental anchor parameters calibrated for the Vryhof Stevpris anchor in Speswhite Kaolin (i.e. f = 1.4 and θw = 35◦ for St = 2) can then be used as a base case to match the anchor capacity calculated using the anchor force component model. This is achieved by varying the projected area of the shank in the direction of travel, ASB , on the basis that due to the complexity of the shank, its apparent area is not simply calculated. The other anchor area parameters AFB , ASF , and AFF are taken as fixed anchor characteristics based on the actual anchor geometry. Although the obtained shank bearing area may not strictly relate to the actual anchor geometry, it does permit the formation of an alternative anchor model which is comprised of separate anchor force components and which is calibrated against reliable model anchor centrifuge test data. 4

INCLUSION OF SOIL SENSITIVITY IN N-R MODEL

Unlike the anchor force component model described in the previous section, the ‘standard’ N-R anchor embedment model does not directly account for the sensitivity of the soil. However, it is possible to match the anchor capacity calculated by the N-R model at a given embedment depth with that calculated by the force component model for different values of St by altering the fundamental anchor parameters f and θw . Considering that θw should vary as a function of St such that the anchor bearing forces vary consistently with the bearing factor described by Equations 3a, 3b and 5, and the shear forces given by Equations 4a and 4b, an iterative approach can be employed to obtain

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Figure 3. Anchor form factor and resultant angle versus sensitivity for 10 tonne Vryhof Stevpris anchor.

corresponding f and θw values for a given St , such that the anchor capacities predicted by the N-R model match those from the force component model. An example of the calculated f and θw values obtained using the process described above for sensitivities ranging between 1.2 and 20 and corresponding to a 10 tonne Stevpris anchor is presented on Figure 3. For increasing St , it can be seen that θw increases while f decreases. This is a simple reflection of the fact that as the sensitivity increases, both the normal and shear forces reduce, though the shear forces reduce more rapidly. Additional results from the same example are presented on Figure 4 in terms of the anchor capacity at the padeye, Ta , and the padeye embedment depth, da , during installation versus the horizontal padeye drag length, xa , normalised by the anchor fluke length, Lf . Profiles are included corresponding to the f-θw -St values contained on Figure 3. Generally, higher St implies an increased level of soil remoulding during installation, leading to lower anchor capacity and reduced anchor embedment. This trend between St , Ta and da is similar but less marked than presented by Aubeny & Chi (2010), who performed an assessment of drag anchor embedment in soft soils using a new numerical procedure. However, the Aubeny-Chi method ignored the effect of sensitivity on the bearing resistance components, and it is thought that this may explain the greater influence of sensitivity on the anchor trajectory obtained in their work. 5

Figure 4. Effect of sensitivity on installation capacity and embedment depth for 10 tonne Vryhof Stevpris anchor.

Equation 7 can be rewritten in terms of the consolidation factor, Ucons , and the cyclic loading factor, Ucy :

Tentative recommendations are made in DNV (2000) on values of Ucons and Ucy to be used for drag anchor design. However, these are not applicable to the high sensitivity carbonate soils encountered offshore Australia, since the recommended values for these factors only account for sensitivities up to 2.5, whereas carbonate soils generally exhibit much higher sensitivities. Instead, the consolidation and cyclic effects represented by Ucons and Ucy may be calculated for each individual term of the anchor force component model (i.e. RFB , RSB , RFF RSF ) and at each increment of the corresponding N-R anchor embedment analysis.

CONSOLIDATION & CYCLIC EFFECTS

As outlined in DNV (2000) and discussed in Neubecker et al. (2005), the ultimate holding capacity (or ‘characteristic anchor resistance’), Rc , of a drag anchor comprises the sum of the installation anchor resistance, Ri , and the predicted post-installation effects of consolidation (Rcons ) and cyclic loading ( Rcy ):


5.1

Soil consolidation effects

Following installation, the effects of consolidation of the soil around the anchor will generally act to increase the anchor holding capacity. As detailed in Neubecker et al. (2005), it has been observed that in the case of vertical surfaces (such as pile walls or skirts) the loss of normal effective stress at the surface after the penetration process is only partially recovered after consolidation, due to the high soil

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compressibility and arching mechanisms that generally occur in carbonate soils. However in the case of a horizontal surface, Neubecker et al. (2005) considered that the normal effective stress would ultimately revert to the in-situ conditions. An additional consideration is that, where the normally consolidated strength, su−NC , is greater than the remoulded strength, su−m /St the reconsolidated soil strength following full remoulding is assumed to not exceed the normally consolidated strength, su−NC , which for typical carbonate materials may be calculated as:

and where σvo is the in-situ vertical effective stress. Hence, for such cases the consolidated soil strength, su−con , acting on a frictional surface having an inclination to the horizontal, θ, may be calculated as:

The above scenario is typical of carbonate materials which generally have a relatively high sensitivity. As a consequence of the ‘dual-component’ nature of the bearing capacity factor, Nc , described by Equation 5, the fraction of the bearing resistance mobilised in disturbed ground (i.e. in the ‘backflow’ region) will be subject to consolidation effects. This may be accounted for by enhancing the bearing resistance of this component with due consideration of the appropriate re-consolidated undrained shear strength. Incorporating the soil consolidation effects described above into the component model expressions describing the bearing and frictional resistances on the fluke (RFB and RFF respectively) results in the following:

the remoulded zone. The excess pore pressures dissipate over time, and this consolidation process leads to a recovery of the effective stresses. In a simple consolidation model, it is assumed that as the excess pore pressures dissipate the effective stresses eventually recover to their original in-situ values, which as discussed previously, is also the assumption adopted by Neubecker et al. (2005) for horizontal surfaces (as incorporated in Equation 10). For the case of carbonate soils with high sensitivity, the high compressibility of the remoulded soil compared to the surrounding in-situ soil leads to the potential for arching mechanisms to develop. For these soils it cannot necessarily be assumed that complete dissipation of excess pore pressures will result in full recovery of effective stresses. In order to capture the complexity of such a consolidation process, a numerical analysis was performed using the finite difference program FLAC (Itasca, 2005) for a typical fine grained carbonate soil with a sensitivity of 20. The analysis represented a plane stain slice through an embedded anchor and surrounding soil perpendicular to the direction of anchor drag, for a particular geometry. The aim was to assess the degree of consolidation within the soil around the anchor over a 30 day period following installation. Although the anchor and soil properties adopted in the analysis were representative of a specific offshore location, the analysis results provide some useful insight into the post-installation consolidation process. The key result of the analysis is presented on Figure 5, which shows the average degree of consolidation, U, plotted against the elapsed time following installation of the anchor. The average degree of consolidation is defined as the average vertical effective stress at a given time divided by the in-situ vertical effective stress. It can be seen that after an elapsed time of 30 days, U had reached approximately 74%. However, it is also clear that the consolidation process is substantially complete after 30 days, and therefore the vertical effective stresses may only ever recover to about 75% of the in-situ stresses, unless creep processes increase this over time.

Similar expressions may be obtained for the bearing and frictional resistances on the shank (RSB and RSF respectively). The parameter U is defined as the average degree of consolidation and is discussed below.

5.2 Assessment of average degree of consolidation During installation, the process of anchor penetration results in the remoulding of soil around the anchor within the ‘back-flow’ region. This in turn results in the generation of excess pore pressures and a corresponding reduction in the effective stresses within


Figure 5. Example assessment of consolidation progress within remoulded soil zone after anchor installation.

5.3 Cyclic loading effects After a period of soil consolidation, the anchor may be subjected to a period of one-way cyclic loading arising from a passing storm. The effects of this cyclic loading will generally act to decrease the anchor holding capacity. These cyclic loading effects on the in-place anchor capacity may be accounted for by calculating the cyclic undrained shear strength, su−cyc , acting on each anchor component. The magnitude of su−cyc may be calculated as:

where A, B and C are empirical coefficients determined via examination of laboratory cyclic soil are strength test data. The parameters su−x and σv−x set to su−m and σvo respectively for those force components acting in virgin (i.e. intact) soil, and to either su−NC or su−con and σv (the effective vertical stress after remoulding and consolidation) respectively for those force components acting in disturbed soil (i.e. in the back-flow region). An example of the relationship between su−cyc /su−m and su−m /σvo as described by Equation 12 for a typical carbonate sediment under one-way cyclic loading is presented on Figure 6. It can be seen that as the ‘apparent overconsolidation’ of the soil (represented ) increases, the relative undrained strength by su−m /σvo under cyclic loading (represented by su−cyc /su−m ) decreases. This aspect of carbonate soils is further discussed in Erbrich (2005).

5.4

An example set of results obtained from such an assessment is presented on Figure 7 for a 10 tonne Vryhof Stevpris with a 50◦ fluke-shank setting and a fluke length, Lf , of 3.41 m. The monotonic shear strength, su−m , was assumed to equal 5 + 1.5z kPa (where z is depth below mudline), while the effective unit weight of the soil, γ , was taken as 5.5 kN/m3 . The average degree of consolidation, U, was assumed equal to 0.7, while the one-way cyclic undrained shear strength, su−cyc , was determined according to the relationship presented on Figure 6. For an assumed St = 5, the data contained on Figure 3 imply f = 1.01 and θw = 37.8◦ . Furthermore, it was assumed that an 80 mm bar diameter chain was connected to the anchor padeye with a friction coefficient, µ, of 0.4 and a zero uplift angle (i.e. horizontal) at the mudline. The results on Figure 7 show the anchor holding capacity at the padeye, Ta , during installation, after a period of consolidation and then after a period of cyclic loading, plotted against xa /Lf . Also included on Figure 7 are the anchor padeye and fluke tip embedment depths, da and dt respectively, plotted against xa /Lf . It can be seen that during installation and after a drag length of 40Lf , the anchor has embedded to a padeye depth of 14.6 m and has a holding capacity at the padeye, Ta (≈ Ri , see Equation 8) of 2.5 MN. If the anchor was then left at this depth for a consolidation

In-place anchor capacity – example

The N-R and force component drag anchor models described above may be used to calculate the embedment trajectory and resistance of an anchor during installation, as well as the subsequent in-place anchor holding capacity after any given amount of consolidation and after application of cyclic loading.

Figure 6. Typical one-way cyclic strength data for carbonate sediments.


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Figure 7. Example assessment of in-place capacity and embedment depth for 10 tonne Vryhof Stevpris anchor.

period sufficient to achieve U = 0.7, the analysis indicates that the holding capacity at the padeye would increase to 3.5 MN. Then, if the anchor were subjected to a period of one-way cyclic loading resulting from a passing storm, the holding capacity at the padeye would decrease to 3.0 MN. In terms of Equation 8, these capacities imply ‘overall’ values of Ucons = 1.40 and Ucy = 0.86. Although DNV (2000) does not provide any specific values of Ucy , it does suggest a Ucons range of 1.35–1.55 for St = 2.5.

the Bass Strait (Erbrich, 2005), where failure occurred at a cyclic preload considerably less than the static preload that had previously been applied. Drag anchors are generally not permitted to move under the design loads, but after cyclic failure there is the potential for ongoing and permanent anchor displacements (i.e. ‘ratcheting’) at loads less than the limiting cyclic failure load. Hence, avoidance of cyclic failure should be a key design objective. REFERENCES

6

SUMMARY & CONCLUSIONS

This paper has considered a simple force component model acting on different elements of a drag anchor. A method for determining variations in each of these independent force components (as a function of whether it is a shear or bearing force, element orientation, soil consolidation, cyclic load effects) has then been presented to determine the time dependent anchor capacity. Such considerations are particularly important in highly sensitive carbonate soil conditions where the anchor capacity can be shown to increase substantially after consolidation of the soil. Cyclic load effects will also influence the anchor capacity, by reducing the soil strength and force required to initiate movement of the anchor. It is only through detailed consideration of these combined effects that a final in-place capacity of a drag anchor can be established. Although it would be beneficial to obtain anchor field test data to validate the model predictions, there is a high level of confidence in the model given its sound theoretical basis. An analog to the problem of anchor behaviour in cyclically degradable soils was the spudcan preloading operations at the Trefoil field in


Aubeny, C.P. & Chi, C. (2010), Mechanics of Drag Embedment Anchors in a Soft Seabed, Journal of Geotechnical & Geoenvironmental Engineering, ASCE, Vol.136, No. 1, pp. 57–68. DNV (2000), Design & Installation of Fluke Anchors in Clay, Recommended Practice RP-E301. Erbrich, C.T. (2005), Australian Frontiers – Spudcans on the Edge, Proc. International Symposium on Frontiers in Offshore Geotechnics – ISFOG 2005, Perth, Australia, Balkema: Rotterdam. Itasca (2005), FLAC: Fast Lagrangian Analysis of Continua, User Manual, Itasca Consulting Group. Neubecker, S.R. & Randolph, M.F. (1996), The Performance of Drag Anchor & Chains Systems in Cohesive Soil, Marine Georesources & Geotechnology, 14, pp. 77–96. Neubecker, S.R., O’Neill, M.P. & Erbrich, C.T. (2005), Preloading of Drag Anchors in Carbonate Sediments, Proc. International Symposium on Frontiers in Offshore Geotechnics – ISFOG 2005, Perth, Australia, Balkema: Rotterdam. O’Neill, M.P. (2000), The Behaviour of Drag Anchors in Layered Soils, PhD Thesis, Department of Civil & Resource Engineering, The University of Western Australia. Vryhof (2010), Anchor Manual 2010, Vryhof Anchors BV. Zhou, H. & Randolph, M.F. (2009), Resistance of Full Flow Penetrometers in Rate-Dependent and Strain-Softening Clay, Geotechnique, Vol. 59, No.2.

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Caisson capacity in clay: VHM resistance envelope – Part 1: 3D FEM numerical study E. Palix & T. Willems Fugro Offshore Geotechnics, Paris, France

S. Kay Fugro Offshore Geotechnics, Leidschendam, The Netherlands

ABSTRACT: This paper is about the development of HM resistance envelopes using PLAXIS 3D Finite Element Analyses for rigid circular caissons with embedment ratios L/D of 0.5, 1, 2 and 4. Clay with constant and normally consolidated undrained shear strength profiles was considered. The resistance envelopes were compared to results given by two Fugro software packages: CANCAP 2 based on limit equilibrium solutions and HARMONY, a quasi 3D finite element program. Satisfactory comparisons were obtained which improves the confidence in their use for the determination of foundation caisson dimensions. A companion paper deals with the development of VHM resistance envelope equations, applicable to offshore caisson foundation design and valid for L/D ratios between 1.5 and 6 and a wider range of undrained shear strength profiles. 1

INTRODUCTION

This section, which reviews caisson capacity assessment in clay using numerical modeling, shows that resistance envelopes for caissons under general VHM loading are not currently available, even in uniform soil. Offshore caissons are rigid circular open-ended steel cans (embedment ratio L/D > 1), usually vertical and installed by self-weight and suction. They are a technically efficient and economically effective solution to anchor floating vessels (such as FPSOs) in deep water or for jacket foundations (see Fig. 1a). Today, in the 2010’s, caisson foundations are also frequently used to support seafloor structures such as deepwater manifolds, PLEMs, PLETs, pumps, etc. (Fig. 1b). 1.1 Suction anchors Caissons were first used in the 1980’s as anchors to moor floating vessels such as FPSOs and FPUs. Design methods to assess anchor caisson installation resistance and subsequent holding capacity under inclined tension loading are now well established and documented (e.g. Andersen et al., 2005). In deep water normally consolidated (NC) clays, suction anchor caisson diameters D are typically in the range 4 m to 6 m with maximum achievable embedment ratios L/D around 6. A frequent assumption, based on limit lateral pressure distribution theory, is to take an optimum attachment point slightly deeper than 2/3L for NC clays to ensure a slight backward rotation (almost pure anchor translation and no crack behind the anchor). Uncertainty on the optimum point increases for soil profiles with a highly non-linear su profile and/or low

caisson L/D. Supachawarote et al. (2004) established ellipsoidal equations for the V-Hmax plane resistance envelope (i.e. no anchor rotation) for caissons with L/D values of 1.5, 3 and 5 in NC clay and a range of caisson-soil adhesion coefficients. 1.2

Unlike anchors, such caisson types are generally “stubbier”, with lower L/D values around 2–4. In service, they are subjected to numerous combinations of overturning H and M loads (generated by seismic activity, pressure or thermal expansion of the connected pipelines/flowlines). Vertical load components (dead weight) are generally small and compressive, and HM loads are usually overturning. Due to the high M load, associated failure mechanisms are usually rotational. Due to the number of VHM load cases, many analyses are required to optimize foundation design. When M/H is significant and L/D > 1, classical VHM bearing capacity theory, (e.g. ISO 19904-1, ISO (2003)), is inappropriate and 3D FE analysis is impractical. Offshore geotechnical engineers generally use a variety of tools: either based on upper-bound theory (Murff & Hamilton 1993) or limit equilibrium analyses (Fig. 5 and Fugro 2009). 1.3 VHM resistance envelopes Ideally, design of all the above foundation types should use VHM resistance envelopes. Poulos & Davies (1980) gave the first MH resistance envelope equations for rigid piles subjected to lateral earth pressures (pu ) either constant or varying linearly with depth.


Small seafloor structure supports

Figure 2. Plaxis 3D mesh for L/D = 3 (52461 elements).

and various failure conditions. Soils other than NC clay are encountered in frontier areas. This was the reason for including a “stepped” profile in Kay & Palix (2010). 2

Figure 1. Suction caisson: recent developments are fixed platforms (Kolk et al. 2001) and manifolds (a) and (b). Sign convention (c). VHM reference point location: caisson head.

The resulting envelope is “laurel-leaf ” shaped (i.e. smooth-edged and lens shaped). Besides V = 0, the model does not account for non-uniform pu distribution (e.g. due to shallow wedge/deep flow failure mechanisms) and base shear/moment. Both effects are important for small L/D values, where rotational failure also occurs, and alter ez,su for anchors. Thanks to reliable 3D FE software running on cheap high performance digital computers, envelope methodology is re-emerging (e.g. Bransby & Randolph 1999 for rough surface foundations). More recently, papers on “shallow” strip foundations (L/D up to 1) have appeared by Bransby & Yun (2009) and Gourvenec (2008). The former assume HM loading of skirted or solid foundations and soil of constant or linearly increasing su , whereas Gourvenec researched VHM loading on solid foundations and constant su soil. Their work showed that various failure mechanisms are possible and limit resistances are related to embedment ratio. In addition, a suitable “load reference point” has to be defined: this affects the envelope shape, even though foundation capacity is unaltered. No elegant simple solutions exist for deeper caisson foundations in more complex soil conditions. In general, there is a requirement for a reliable method of assessing caisson capacity for all L/D under general (both restoring and overturning) VHM loading


PLAXIS FINITE ELEMENT MODEL

All the finite element analyses used PLAXIS 3D version V2.2. Caissons of 5 m diameter were modeled as full rigid body avoiding any internal scoop mechanism. Only half of the problem needed to be represented because of symmetry in the lateral loading direction (M and H in the same direction) and the absence of torque load. 2.1

Mesh

PLAXIS 3D uses 15-nodes wedge elements. This type of element is a good compromise between the accuracy required for soil calculations, and memory and time consumption imposed by 3D calculations. Various mesh densities were investigated to use the maximum elements available for the longest caisson (L/D = 4) and to ensure a good accuracy. Meshes in the horizontal plane and at the caisson tip were the same for the four geometries studied. A typical three-dimensional FE mesh used for the L/D = 3 caisson is shown in Figure 2. The mesh extends 5D from the centre of the foundation and 4D beneath the foundation. Zero-displacement boundary conditions prevent out-of-plane displacements of vertical boundaries, and the base of the mesh is fixed in all three coordinate directions. 2.2

Material properties

Both constant and normally consolidated undrained shear strength profiles were considered (su = 10 kPa

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and su = 2 + 1.5 z). The soil was modeled as an undrained, cohesive linear elastic-perfectly plastic (Tresca) material by using the PLAXIS MohrCoulomb strength model. A Poisson’s ratio ν = 0.45 was used to simulate undrained (i.e. no volume change) conditions, stiffness:strength ratio Eu /su = 500, submerged soil unit weight γ = 5 kN/m3 and initial stress ratio K0,total = 1. The caisson was modelled as a linear elastic material with very high stiffness and with the same unit weight than the surrounding soil (i.e. weightless). Isoparametric (curved) interface elements were considered with an α value (i.e. reduced interface Rint in PLAXIS) of 0.65. This value is commonly used for suction anchor design (Andersen et al. 2005) and is considered more representative of reality than a rigid interface. 2.3 Loading conditions The shape of the resistance envelope depends on the VHM load reference point, and there is a lack of consensus between researchers about the location of the reference point. For embedded shallow foundations, reference points have been defined at caisson tip or beneath the top cap (e.g. Bransby & Randolph 1999, Yun & Bransby 2007). In the present study, the reference point is at the top of the caisson (see Fig. 1c). This is consistent with offshore practice: for the vast majority of seabed support structures, external loads are provided by structural engineers at the point of load transfer, i.e. seafloor. Moment was applied by two diametrically opposite vertical loads acting on top of the caisson. FE simulations were performed considering no vertical load (V = 0). This was because Gourvenec (2007) showed that for circular shallow foundations the effect of vertical load on the resistance envelope was negligible for V < 0.3 Vmax. In addition, suction anchor pile resistance (in the V-Hmax plane) was found to be essentially independent of vertical load for V < 0.4 Vmax (Supachawarote et al. 2004) and was confirmed by Kay & Palix (2010). 2.4 Verification with suction anchor benchmark As the previous version 2.1 of PLAXIS was recognized to overestimate caisson capacity (Andersen et al. 2008), a prerequisite was to check the capability of the newly released version in which isoparametric interface elements were introduced. This was done by performing a benchmark and comparing PLAXIS 3D data with ABAQUS and BIFURC 3D data obtained by OTRC, COFS and NGI on one of the reference cases treated during the API/Deepstar industry sponsored project (Andersen et al. 2005). The anchor has been modelled with PLAXIS 3D as a rigid body with a submerged unit weight γ = 9 kN/m3 equivalent to (Wsub + Wplug )/AL. The V-H envelopes, obtained for pure caisson translation, agree well with the ones obtained by ABAQUS and BIFURC (Fig 3.). © 2011 by Taylor & Francis Group, LLC

Figure 3. Comparison of PLAXIS 3D Foundation and benchmark suction anchor computations after Andersen et al. 2005 (Case C2: D = 5 m, L/D = 1.5, su = 1.25 z, α = 0.65 and no crack at active side).

The above verification, and work done by Edgers et al. (2009), suggests that the isoparametric interface elements used in PLAXIS 3D version 2.2 accurately model interaction between curved soil-structure surfaces. 3

PLAXIS RESULTS

Resistance envelopes are presented in HM load space for V = 0. Both horizontal and moment loads are proportionally incrementally increased (i.e. M/H remain constant) until soil failure. The final loads set is defined as the (M,H) values obtained when the maximum displacement of the top of the caisson was about D/10. Due to caisson rigidity, the plastic plateau is generally obtained before this displacement. In the following, it was assumed that final points lay on the resistance envelope. Only the first quadrant (H and M positive) and fourth quadrant (H positive and M negative) are presented. The remainder of the envelope can be obtained by symmetry. The first quadrant corresponds to a load case where the moment is induced by the horizontal external load (overturning). This is interest for the design of seabed support structures where moments are induced by horizontal forces applied above the top of the foundation. The fourth quadrant is mainly of interest for design of jacket foundations and suction anchors when the moment is restoring (M counteracts with the horizontal load H) or when the load application point is below the seabed (lug level selected to minimize anchor rotation). Resistance envelopes under combined moment and horizontal loading are found to be close to rotated ellipses. For L/D > 1 the envelope is almost linear in the first quadrant. Figures 4 and 8 show that, close to Hmax , data points on the resistance envelope are not equidistant from the

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Figure 4. non-dimensional MH envelope curves for uniform clay.

ellipse major axis. The shape of the envelope tends to flatten for slight backward caisson rotation. This phenomenon is accentuated when the L/D ratio increases. For a L/D ratio of 4, the shape of the HM envelope is close to those proposed by Poulos & Davis (1980) for unrestrained rigid piles under HM loading applied at top of the pile. On Figures 4 and 6, resistance envelopes are drawn in non-dimensional space using M* = M/(DL2 su,av,L ) and H∗ = H/(DLsu,av,L ) for L/D ≥ 1 and M∗∗ = M/(D2 Lsu,av,L ) and H∗∗ = H/(D2 su,av,L ) for L/D < 1. The variation of lateral bearing capacity factors Np,fixed and Np,free with L/D for uniform clay profile can be estimated by comparing respectively Hmax and Ho values given on Figure 4. The trends of Np,fixed and Np,free factors are given in a companion paper (Kay & Palix 2010) for 1.5 < L/D < 6 and three different soil profiles, and these confirm and extend data obtained by Randolph et al. (1998) with AGSPANC analytical software. It can be observed that for L/D ratios of 2 and above, the envelopes are similar. The grouping of the curves is mainly due to a common global shape and that Np,fixed and Np,free are almost constant for L/D ≥ 2.

4

COMPARISON WITH LIMIT EQUILIBRIUM MODEL

For preliminary design, the use of 3D FE software is not appropriate: engineers need less sophisticated and time consuming tools for capacity analyses. Hence, limit equilibrium analyses are often used at preliminary stage to identify critical load cases and optimize caisson geometry. For caissons (cans) with embedment ratios (L/D) up to around 2, CANCAP2 (Fugro 2009) can be used to compute Factors of Safety against rotational failure under combined VHM loading in layered clay soils. Three potential rotational failure modes are considered (Fig. 5): 1 rotational failure of the whole can in the direction of the overturning moment (shallow rotational failure, SR), © 2011 by Taylor & Francis Group, LLC

Figure 5. CANCAP2 rotational failure modes: (a) Shallow rotational failure, SR, (b) intermediate rotational failure IR (c) deep rotational failure DR. Typical CANCAP2 results (L/D ≈ 0.5, NC clay) DR mode (d) failure inside can, minimum FOS = 1.1, (e) failure outside the can, minimum FOS = 1.9.

2 rotational failure of the whole can, combined with active and passive failure wedges at part of the back and front of the can wall (intermediate rotational failure, IR) 3 bearing capacity failure – either tangential to can toe level or penetrating soil plug within can – combined with active and passive failure wedges at the back and front of the can wall (deep rotational failure, DR). Circular cans are usually modeled as a square can with the same cross sectional plan area. Failure is assumed to occur in a vertical “slice” of soil enclosing the square can. Resisting forces (including frictional forces on the two sides of the failing soil slice) in the plane of the applied VHM loads are computed using limit equilibrium solutions for various soil segments. These resisting forces use the maximum soil shear stress values divided by an unknown material factor (γm ) which is found from moment equilibrium of the resisting forces and the external (VHM) loads. Analyses are done for various centres of rotation (on and off the centre-line) until the minimum γm value is found. In the present paper, deep rotational failure was forced to be tangential to the can toe level (Fig. 5e) in order to simulate a solid foundation. For V = 0, HM envelopes determined using PLAXIS 3D and CANCAP2 are comparable (Fig. 6). Insets show that PLAXIS soil displacements around the caisson emulate those in CANCAP2. An exception is for H values close to Hmax . This corresponds to a shallow wedge / deep flow failure type model for

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Figure 8. Comparison between PLAXIS 3D and HARMONY HM envelopes (V = 0 and su = 10 kPa).

5

Figure 6. Comparison between PLAXIS 3D and CANCAP2 HM envelopes (V = 0): (a) for Su = 10 kPa & (b) for normally consolidated clay and L/D = 1.

An alternative time saving approach for caisson modelling can be to use a quasi 3D finite element program. HARMONY (Fugro 1994) computes the response of axisymmetric bodies subjected to non-axisymmetric load. Details of HARMONY model, verification and results are given in a companion paper (Kay & Palix 2010). Basically, HARMONY was extensively used to analyze six caisson L/D ratios between 1.5 and 6 and three su profiles. Based on these data, ellipse/ellipsoidal equations were developed for defining the complete VHM envelope. Figure 8 compares results for L/D values of 2, 3 and 4. HARMONY results are seen to superimpose on every part of the corresponding PLAXIS 3D envelope. 6

Figure 7. (a) Soil displacement around the caisson for M/H =− 11, (b) incremental displacement at 15 m depth (L/D = 4 and uniform soil profile).

(almost) pure lateral translation of the caisson. Nonrotational mechanisms are (cautiously) not considered in CANCAP2. On all other parts of the HM envelopes, good matches were found between the lines of failure and centers of rotation obtained by both methods (Fig. 6b). The use of CANCAP2 is limited to caisson with L/D up to 2. For higher L/D values, CANCAP2 failure envelopes become unconservative, particularly in the fourth quadrant (H positive and M negative). Flow mechanism starts to occur at caisson base (Fig. 7). This type of mechanism is not considered by CANCAP2. © 2011 by Taylor & Francis Group, LLC

COMPARISON WITH HARMONY

CONCLUSIONS

The aim of the work presented here was to investigate numerically the shape of HM resistance envelope for caisson foundations in clay with embedment ratios between shallow foundations (L/D < 1) and short rigid piles. Based on the above results, HM resistance envelopes (at V = 0) have been found to be dependent on the embedment ratio L/D and the soil shear strength profile. The global envelope shape is almost elliptical. A companion paper (Kay & Palix 2010) gives VHM ellipse/ellipsoidal equations for the design of offshore caisson foundations with L/D values of 1.5 and above. Simpler tools than 3D FE software can be used for preliminary design: limit equilibrium analysis using CANCAP2 has proved to be reliable and efficient for caissons with L/D ≤ 2, with or without internal crossplates, and under significant V load. For final design, 3D FE analyses will still be needed to assess foundation responses and soil reactions. 7

NOTATION

A = caisson area = πD2 /4 α = clay-steel outer adhesion factor

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D ez ez,su γm H H∗ H∗∗ Hmax Ho

L M M∗ M∗∗ M Mmax Mo Np,fixed Np,free pu su,z su,av,L

V Vmax Wsub Wplug z

= caisson outer diameter = H load eccentricity with respect to caisson head (seafloor) = analytical ez based on su L L = 0 su,z z dz / 0 su,z dz = material factor = horizontal load at caisson head (seafloor) = non-dimensional H value [−] = H/(D L su,av,L ) = H/(D2 su,av,L ) = caisson maximum (“fixed head”) horizontal resistance (with V = 0) = Np,fixed L D su,av,L = caisson “free head” horizontal resistance (with V = M = 0) = Np,free L D su,av,L = caisson embedded length = moment load at caisson head (seafloor) = non-dimensional M value = M/(D L2 su,av,L ) = M/(D2 L su,av,L ) = transformed moment = maximum moment resistance (V = 0) = caisson moment resistance (with V = H = 0) = “fixed head” lateral bearing capacity factor = Hmax /(D L su,av,L ) = “free head” lateral bearing capacity factor = Ho /(D L su,av,L ) = ultimate lateral resistance = soil undrained shear strength at any depth z = average su,z between caisson head (seafloor) and caisson tip (L) L = 0 Su,z z dz / L = caisson vertical load at caisson head (seafloor) = caisson vertical resistance (with H = M = 0) = caisson submerged weight = soil plug submerged weight = depth below seafloor

Deepwater Applications”, in Gourvenec, S. and Cassidy, M. (Eds.), Frontiers in Offshore Geotechnics ISFOG, Perth, 2005, Taylor & Francis, London pp. 3–30. Andresen, L., Edgers, L. and Jostad, H. P. (2008), “Capacity analysis of suction anchors in clay by PLAXIS 3D Foundation”. PLAXIS Bulletin, issue 24, October, 5–9. Bransby, M.F. & Randolph, M.F. (1999) “The effect of embedment depth on the undrained response of skirted foundations to combined loading”, Soil Found. 39, N◦ . 4, 19–33. Bransby, M.F. & Yun, G.J. 2009. “The undrained capacity of skirted strip foundations under combined loading” Géotechnique 59, N◦ . 2, 115-125. Edgers, L., Andersen, L. & Jostad, H.P. 2009. “Evaluation of loading-carrying capacity of suction anchors in clay by 3D finite element analysis”, 1st Int. Symp. on Computational Geomechanics (ComGeo I), Juan-les-Pins, France. Fugro. 1994. HARMONY user’s manual HARMONY 00.09. Fugro. 2009. CANCAP2 user’s manual CANCAP2 00.17. Gourvenec, S. 2007. “Failure envelopes for offshore shallow foundations under general loading”, Géotechnique 57, N◦ . 9, 715–728. Gourvenec, S. 2008. “Effect of embedment on the undrained capacity of shallow foundations under general loading”, Géotechnique 58, N◦ . 3, 177–185. ISO International Organization for Standardization. 2003. “Petroleum and Natural Gas Industries – Specific Requirements for Offshore Structures – Part 4: Geotechnical and Foundation Design Considerations”, International Standard ISO 19901-4:2003. Kay,S. & Palix, E. 2010. “Caisson Capacity in Clay: VHM Resistance Envelope – Part 2: Envelope Equation and Design Procedures”, in Frontiers in Offshore Geotechnics ISFOG 2010: 8-10 November, 2010, Perth, Australia. Kolk, H.J., Kay, S., Kirstein, A. & Troestler, H. 2001. “North Nemba Flare Bucket Foundations”, Offshore Technology Conference, 30 April-3 May 2001, Houston, Texas, U.S.A., OTC Paper 13057. Plaxis BV. 2008. PLAXIS 3D Foundation Version 2.2. Poulos, H.G. & Davis, E.H. 1980. “Pile Foundation Analysis and Design”, John Wiley and Sons, New York, Series in Geotechnical Engineering, Chapter 7. Randolph, M.F., O’Neill, M.P., Stewart, D.P. & Erbrich, C. 1998. “Performance of Suction Anchors in Fine-grained Calcareous Soils”, in 30th Annual Offshore Technology Conference, 4–7 May 1998, Houston, Texas, U.S.A.: Proceedings, Vol. 1, OTC Paper 8831, pp. 521–529. Supachawarote, C., Randolph, M.F. & Gourvenec, S. 2004. “Inclined Pull-out Capacity of Suction Caissons”, ISOPE, Toulon, France, May 23–28, 2004, International Society of Offshore and Polar Engineers (ISOPE), Cupertino, pp. 500–506.

REFERENCES Andersen, K.H., Murff, J.D., Randolph, M.F, Clukey, E.C., Erbrich, C.T., Jostad, H.P., Jansen, B.,Aubeny, C., Sharma, P. & Supachawarote, C., (2005) “Suction Anchors for


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Numerical investigation of the behaviour of suction caissons in structured clays S. Panayides & M. Rouainia Newcastle University, Newcastle Upon Tyne, UK

A. Osman School of Engineering, Durham University, Science Laboratory, Durham, UK

ABSTRACT: This paper investigates the ultimate capacity of suction anchor foundations, using an advanced constitutive model. The paper explores the effect of changing the aspect ratio of the caisson on the undrained load-carrying capacity of the bucket foundations in clay. Illustrative numerical results for an inorganic clay of low sensitivity from the Norrköping region in southern Sweden demonstrate the potential of the constitutive model.

1

INTRODUCTION

The relative inefficiency of piles in resisting lateral forces has led the offshore industry to consider alternative anchorage systems such as suction caissons. Suction caissons can be installed very quickly and precisely at the desired location with less heavy installation equipment and at lower cost. Therefore they are considered as a viable anchorage system in a wide variety of soils ranging from soft clay to dense sands and overconsolidated clays and for a wide variety of structures ranging from floating exploration platforms to permanent production facilities. The development of suction caissons in recent years has seen them used around the world in more than 36 fields in the last decade alone (Andersen et al., 2002). Suction caissons are large cylindrical shells, with an open bottom and a closed top fitted with valves. The aspect ratio of these piles, defined as the length to diameter ratio, is relatively small when compared with the aspect ratio of conventional piles, typically six or less (Andersen et al., 2005). Internal stiffeners are usually added, to resist buckling during the installation process, since the caisson walls are relatively thin. They are installed partly by self weight and partly by differential pressure between the surrounding environment and the inside of the skirted foundation. In some cases, dead weights can be applied on the top of the cap to ensure that compressive loads are acting on the suction anchors (Zdravkovic et al., 2001). Once full penetration has been achieved, the valve is closed. Any vertical movement during service will result in the generation of suction pressure inside the anchor which will mobilize the reverse end-bearing mechanism, as it is described by Byrne and Finn (1972). Foundations for offshore structures, however, experience significant environmental loads from waves, currents and wind giving


rise to lateral loads. The direct consequence of that is that resultant loads can be inclined to the vertical. It is well documented that suction anchors are capable of resisting both lateral and axial loads as well as inclined loads. The ultimate capacity of suction caissons has been the focus of many investigations in recent years. Following the work of Hogervorst (1980), Keaveny et al. (1994) showed that lowering the load attachment point at mid depth increased the capacity significantly. Suction caisson capacity studies based on upper bound limit analyses from Randolph et al. (1998) and a finite element study from Sukumaran et al. (1999) indicate that the anchor capacity can be maximised when the load is located at a point which forces the anchor to fail in a translational mode of failure rather than rotational. Murff and Hamilton (1993) presented a three dimensional quasi upper bound formulation for predicting the ultimate capacity of laterally loaded piles. The three dimensional mechanism which they proposed comprised of a conical wedge near the free surface and a flow around zone below the wedge (Randolph and Houlsby, 1984). This paper presents a study of the shot-term pullout capacity of suction caissons in soft clay using an advanced constitutive model. The failure envelopes were produced for two reference caissons with length to diameter ratios of 1.5 and 3 respectively. The soil was modelled using the Kinematic Hardening Soil Model (KHSM) as proposed by Rouainia & Muir Wood (2000). 2

GEOMETRY

The geometry for the first reference suction anchor foundation adopted for this study is provided in

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Figure 2. Finite element model for the reference suction caisson with an aspect ratio of L/D = 1.5. Figure 1. Schematic representation of the suction caisson model (Supachawarote, 2004).

estimated using the strength reduction factor and the soil properties as follows:

Figure 1. It comprises of a cylindrical suction anchor with closed top with a diameter D of the cylinder of 5 m, while the skirt length L is varied between 7.5 m and 15 m, giving rise to an aspect ratio of L/D = 1.5 and 3 respectively. A wall thickness of 50 mm (or D/100) was used in all cases. The caisson is embedded with the top cap flush with the surrounding ground level and the load attachment point, or padeye location, at a depth zp along the caisson shaft. Loads are applied at an angle θ from the horizontal, and the depth to the point of intersection of the line of action of the load with the centre-line of the caisson is denoted by zcl . The caisson is considered to be very stiff compared to the soil. The pullout loads are applied on different points on the side of the caisson with at an inclination θ to the horizontal to produce the failure envelopes. The geometry is extended 3 times the length L around to avoid influence of the geometry boundaries.

3

where ϕi and ci are the interface effective friction angle the interface effective cohesion, respectively and Eoed is the constrained modulus of the soil. ϕ and c are the friction angle and effective cohesion of the soil. Since pore water pressures or the installation of the suction anchor, is not considered the phreatic level was placed at the bottom of the geometry. In this study, the Kinematic Hardening Soil Model, developed by Rouainia & Muir Wood (2000) is use. It should be noted that this constitutive model admits the possibility, with certain combinations of soil parameters, of creating rapidly strain-softening materials. For this reason, displacement controlled analyses were carried out in Plaxis 2D software, in order to capture the strain softening behaviour of the clay as it can be seen in Figure (4).

FINITE ELEMENT ANALYSES

The cylindrical suction anchor was modelled using the Plaxis 3D with 15-noded wedge elements. The anchor was modelled by linear elastic wall elements with a high stiffness making them virtually rigid. Since the governing failure mechanisms do not involve the soil plug inside the anchor, this soil was modelled as a stiff, elastic material. The mesh used for the short caisson can be seen in Figure 2. Approximately ∼13000 elements and ∼26000 nodes was found to be sufficiently refined in order to minimize the discetization error for the first case, whereas for the L/D = 3 a mesh of ∼16500 elements and ∼45000 nodes was used. For all the FE-models in this study, interface elements along the outside skirt walls have been used. with the strength reduction factor of the interface (Rinter) set to 0.65. The interface properties are


4

MATERIAL MODEL

The model used in this study was formulated for natural clays within the framework of kinematic hardening with some elements of bounding surface plasticity. It is a rate independent model and it takes into account the effects of damage to structure caused by irrecoverable plastic strains, resulting from sampling or geotechnical loading. KHSM is an extension of the well known Cam-Clay model. The steady fall of stiffness with strain is controlled by an interpolation function which ensures a smooth advancement of the elastic domain (which is enclosed in a small bubble) towards the bounding surface during loading. A scalar variable r, which is a monotonically decreasing function of the plastic strain, represents the progressive degradation of the material. Accordingly,

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Table 1.

Soil parameters for Norrköping Clay.

Material constants

Value

Slope of swelling line κ∗ Slope of normal compression line λ∗ Poisson’s ratio v Critical state stress ratio M Ratio of size of bubble and reference surface R Stiffness interpolation parameter B Stiffness interpolation parameter ψ Destructuration parameter k Destructuration strain parameter A Initial degree of structure r0 Anisotropy of initial structure η0

0.0297 0.252 0.22 1.35 0.145 1.98 1.547 4.16 0.494 1.75 0.5

the following exponential destructuration law, is proposed

where ro denotes the initial structure and kis a parameter which describes the rate of destructuration process with strain. The rate of the destructuration strain εd will be assumed to have the following form

where A is a non-dimensional scaling parameter and p p εq and ε˙ v are the plastic shear strain and the plastic volumetric strain, respectively. The governing constitutive relations of the KHSM are summarized in the Appendix. 5

CLAY PROPERTIES

The parameters required for the analysis correspond to inorganic clay of low sensitivity from the Norrköping region in southern Sweden. The soil has undrained shear strength of 10 kPa down to 3 m, increasing linearly with depth below that level at a rate of 2 kPa/m. An effective unit weight γ = 10 kN/m3 was used. The model parameters for the soil and the interface were taken from Westerberg (1995). An over consolidation ratio (OCR) of 1 was adopted for the analyses. The coefficient of lateral earth pressure (K0NC = 1-sin ϕ ) was taken as 0.5 which corresponds to an effective friction angle φ’ of 300 . For the soil-structure interface, an oedometric Young’s modulus (Eoed ) of 1800 kPa and cohesion c of 2.1 kPa were used. These optimized parameters are described as reference parameters and correspond to the KHSM model in all the analyses below. 6 6.1

RESULTS AND DISCUSSION Failure mechanisms

Figures 3(a) and 3(b) show the 3D and 2D deformed meshes from the analyses of horizontally loaded

Figure 3. Deformed mesh from analyses of pure horizontal loading, for caisson L/D = 1.5 (a) 3D model and (b) 2D model.

suction caissons with an aspect ratio of L/D = 1.5. It can be seen that clear failure surfaces form on the active side of the caisson. On the passive side, the well defined failure surfaces form very close to the caisson, while the caisson translates horizontally. The failure mechanisms become more distinctive with the plot of the displacement vectors as it will be discussed in the following section (Figure 6). As it is evident in Figure 4, the KHSM model can replicate the strain softening behaviour of the clay. The KHSM predicts a peak load of 583 KN, which is ∼9% higher than the ultimate load predicted by the bubble model for which the structure behaviour is switched off. This behaviour would correspond to reconstituted soils.The increased ultimate capacity can prove critical when designing caissons for offshore structures, as it will enable engineers to utilize more economic designs in the future. As a direct consequence, the cost of offshore founding systems may be reduced significantly. It should be noted that this difference in the behaviour is also observed when the caissons are analysed using the three-dimensional finite element model. Figure 5 depicts the displacement shadings in 3D for the case where purely vertical load was applied on the suction caisson with an aspect ratio of L/D = 1.5. It can be seen that the model predicts the resistance to uplift which corresponds to full mobilisation of a reverse bearing mechanism. Figures 6 depict the displacement vectors and failure mechanism obtained from the analysis corresponding to pure horizontal loading where the load was attached at the optimal point. It can be seen that a


Figure 6. Displacement vector for horizontal ly loaded suction caissons L/D = 3.

Figure 4. Load-displacement curve in 2D for horizontally loaded suction caissons L/D = 1.5.

Figure 5. Failure kinematics for vertically loaded 3D suction caissons L/D = 1.5.

well defined failure surface develops on both sides of the caisson. This is in very good agreement with the failure kinematics that have been proposed in the literature for lateral loading of caissons. It can also be seen that the failure extends to the bottom of the caisson, with no flow around zone occurring for the caisson. It should be noted however that the gradient of the wedge varies with depth. As the wedge approaches the tip of the caissons it tends to curve passing tangentially at the bottom of the caisson. 6.2

Failure envelope comparison

Figure 7(a) shows the comparison of the failure envelope for non-horizontal loadings for the structured model (KHSM), the Bubble model and two analytical methods suggested by Supachawarote (2005) and Senders & Kay (2002). It is evident that the structured model predicts the ultimate load for all loading angles very well. In contrast, the bubble model, consistently, underestimates the ultimate load by an average of ∼12%. This behaviour is as expected, since the bubble model was formulated to represent the behaviour of reconstituted material and cannot account for added strength the natural clay deposits exhibits.

Figure 7. Failure envelopes for suctions caissons of different aspect ratios: (a) L/D = 1.5 and (b) L/D = 3.

The analytical equation proposed by Supachawarote et al (2004) for the failure envelope has the following form:


where: a = 0.5 + L/D; b = 4.5 − L/3D

into account, this results in an increase in ultimate capacity of suction caissons, which in turn can lead to a more economical design.

The expression is similar to the one proposed by Senders and Kay (2002), who suggested that both coefficients a and b are equal to 3 for all aspect ratios. Figure 7(b) provides an insight on the effects an increasing of the embedment length of a caisson, has on the capacity. It is evident from the two failure envelopes that the capacity increases significantly with increasing aspect ratio. Moreover, the longer caisson exhibit maximum capacity for loads that have higher lateral components, which is in contrast with the shorter caisson, where the maximum capacity is observed under loads with high vertical components. The capacity for pure horizontal loading is four times larger when the aspect ratio is increased from 1.5 to 3, whereas the maximum load corresponding to pure vertical loading, increases by a factor of ∼2.5. As it can be seen the finite element results agree very well with the curve fitting equation proposed by Supachawarote (2004). The equation recommended by Senders and Kay (2002), predicts lower capacities for inclined loading. For load angles that are close to the pure vertical or pure horizontal loadings, the results from the analytical solution are very similar to the finite element capacities. As the ellipsoidal equation from Supachawarote is made to depend on the caisson geometry, the predicted results are deemed to be more reliable for all caisson geometries.

7

8 APPENDIX I. Non-linear elastic constitutive law

where κ∗ is the swelling line in a volumetric strainlogarithmic mean compression plane and K and G are the bulk and shear modulus, respectively. The stress and strain tensors can be expressed in terms of the volumetric and the deviatoric components and the basic elasto-plastic assumption is the additive decomposition of the strain rate, ε, into an elastic and a plastic part. II. Yield functions


This paper presented numerical analyses of the undrained ultimate capacity of suction caissons installed in a soft structured clay deposit. Two finite element models were simulated for caissons with aspect ratios L/D set to 1.5 and 3 in order to compute the ultimate capacity, for a variation of load inclinations in order to provide the full failure envelopes. The clay deposit was modelled with an advanced constitutive model for natural clays, namely the Kinematic Hardening Structure Model (KHSM), which can simulate the destructuration process. The 2D study included displacement controlled analysis in order to capture the strain softening nature of the soil. Plaxis 3D was used to produce the full vertical and horizontal interaction diagram for the two cases. The two failure envelopes were compared with two analytical equation as proposed by Senders and Kay (2002) and Supachawarote et al (2004). The shape of the yield envelopes obtained with 3D calculations is in good agreement with the analytical solutions and may be used in practice to obtain the optimal capacities on suction caissons. Furthermore, the kinematics obtained from the FEM in both 2D and 3D simulations for various loading inclinations agree very well with the failure mechanisms proposed in analytical methods. The above conclusions imply that the KHSM model can predict the undrained ultimate capacity of suction caissons accurately. When the soil structure is taken


where M is the critical state stress ratio and pc is a scalar variable which defines the size of the outer surface. a¯ denotes the location of the centre of the bubble and R represents the ratio of the sizes of the elastic bubble and the outer surface. η0 is a dimensionless deviatoric tensor controlling the structure and r is the ratio of the sizes of the structure surface and the reference surface. Note that the position of the centre of the structure surface is given by aˆ = {rpc , (r − 1)η0 pc }. III. Isotropic hardening law

λ∗ is the slope of the normal compression line expressed in the same plane as κ∗ . IV. Kinematic hardening law

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where µ ˙ is a positive scalar of proportionality and σ c is there the conjugate stress tensor. V. Plastic modulus at current stress

The additional material parameters ψ and B control the rate of decay of stiffness with strain and the magnitude of the contribution of the interpolation term, respectively. For more information regarding the formulation of the constitutive model, the reader is referred to Rouainia &Muir Wood (2000).

REFERENCES Andersen, K.H. and Jostad, H.P, (2002). “Shear strength along outside wall of suction anchors in clay after installation,” Proc. 12th International Offshore and Polar Engineering Conference, pp 785–794. Andersen,K.H., Murff J.D., Randolph M.F.,Clukey E.C., Erbrich C., Jostad H.P., Hansen B., Aubeny C., Sharma P., and Supachawarote C. 2005. Suction anchors for deepwater applications. Int. Symp. on Frontiers in Offshore Geotechnics, ISFOG. Sept. 2005. Perth, Western Australia. Proc. A.A. Balkema Publishers. Andersen L.,Edgers L.and Jostal H.P.(2008) Capacity Analysis of Suction Anchors in Clay by Plaxis 3D Foundation. Plaxis Bulletin, Issue 24. Oct. 2008. Aubeny, C., S. Moon, and J. Murff (2001). Lateral undrained resistance of suction caison anchors. International Journal of Offshore and Polar Engineering 11(2), 95–103. Aubeny, C., J. Murff, and J. Roesset (2001).Geotechnical issues in deep and ultra deep waters. International Journal of Geomechanics 1(2), 225–247. Aubeny, C.P., Moon, S.K., and Murff, J.D. (2001b). “Lateral undrained resistance of suction caisson anchors.” International Journal of Offshore and Polar Engineering., 11(2), 95–103. Aubeny, C.P., Han, S.W., and Murff, J.D. (2003a). “Inclined load capacity of suction caissons.” International Journal for Numerical and Analytical Methods in Geomechanics., 27(14), 1235–1254. Aubeny, C.P., Han, S.W., and Murff, J.D. (2003b). “Refined model for inclined load capacity of suction caissons.” 22nd International Conference on Offshore Mechanics and Arctic Engineering, Cancun. Bransby, M.F., and Randolph, M.F. (1998). “Combined loading of skirted foundations.” Geotechnique., 48(5), 637–655.


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Hogervorst, J.R. (1980). Field trials with large diameter suction piles. Offshore Technology Conference,Houston, Texas. Paper 3817. Keaveny, J.M., Hansen, S.B., Madshus, C. and Dyvik, R. (1994). Horizontal capacity of large-scale model anchors. Proceedings of the 8th International Conference on Soil Mechanics and Foundation Engineering, New Delhi, India. Murff, J.D., and Hamilton, J.M. (1993). “P-Ultimate for undrained analysis of laterally loaded piles.” ASCE, Journal of Geotechnical Engineering., 119(1), 91–10. Randolph, M.F., and Houlsby, G.T. (1984). “The limiting pressure on a circular pile loaded laterally in cohesive soil.” Geotechnique., 34(4) 613–623. Randolph, M.F., O’Neill, M.P., Stewart, D.P., and Erbrich, E. (1998). “Performance of suction anchors in fine-grained calcareous soils.” Proc. Offshore Tech. Conf., Houston, OTC 14236. Randolph, M.F., & House A.R. (2001). “Analysis of Suction Caisson Capacity in Clay” Proc. Offshore Tech. Conf., Paper No.8831, 521–52. Rouainia, M. and D. M.Wood (2000). A kinematic hardening constitutive model for natural clays with loss of structure. Geotechnique 50(2), 153–164. Senders, M. and Kay, S. (2002) “Geotechnical Suction Pile Anchor Design in Deep Water Soft Clays”, Conference Deepwater Risers, Moorings and Anchorings, London, UK. Sukumaran, B., and McCarron, W.O. (1999). “Total and effective stress analysis of suction caisson for Gulf of Mexico conditions.” OTRC 99 International Conference on Analysis, Design, Construction, and Testing of Deep Foundations, Austin, TX, 247–260. Supachawarote, C., Randoplh, M.F and Gourvenec, S. (2004). “Inclined Pull-Out Capacity of Suction Caissons” Proceedings of the 14th International Society of Offshore and Polar Engineering Conference, 500–506. Zdravkovic, L., Potts, D.M. and Jardine, R.J. (1998). “PullOut Capacity of Bucket Foundations in Soft Clay”. Proceedings of the International Conference on Offshore Site Investigation and Foundation Behaviour, 301–324. Zdravkovic, L., Potts, D.M. and Jardine, R.J. (2001). “Parametric Study of the Pull-Out Capacity of Bucket Foundations in Soft Clay”, Geotechnique, 51 (1) 55–67.


Cyclic moment loading of suction caissons in sand B. Zhu Department of Civil Engineering, Zhejiang University, China

B.W. Byrne & G.T. Houlsby Department of Engineering Science, University of Oxford, UK

ABSTRACT: Suction caissons are being considered as a foundation option for offshore wind turbines. One configuration involves use of the caisson as a mono-foundation, in which case the moment-rotation response of the caisson must be well understood by the designers. The monotonic response has been the focus of recent research work and there is some guidance now available for engineers to use. However, less focus has been applied to the cyclic response and particularly to the long term response under cyclic loading. Over the lifetime of the turbine the foundation is likely to be exposed to many millions of cycles, of varying amplitude and period. One of the key issues that must be addressed by a designer is whether the accumulated rotation, over the life of the turbine, is acceptable. This paper presents experimental data from long term cyclic loading tests on caisson foundations. These tests can be used to develop a framework of response for long term cyclic loading. This paper describes the small scale cyclic loading rig used at Oxford and presents some typical results that give insight into the caisson response to a range of cyclic moment loading patterns. 1

INTRODUCTION

Offshore wind turbines can be founded on either a multi-footing structure, usually when the water depth is greater than 20 m, or a single footing structure, typically for water depths less than 20 m. Only a few multi-footing structures have been installed (for example the two Beatrice wind turbines north of Scotland) so most installed wind turbines are located on monostructures. The monopile is the usual design, although gravity base foundations have also been adopted by some developers. A minor variation on the gravity base concept is the use of a skirted foundation such as a suction installed caisson, as shown schematically in Figure 1(a). Figure 1(b) shows a suction installed foundation at Frederikshaven in Denmark. This was installed in 2002 and supports a V90 wind turbine, though the site is onshore (in a shallow lagoon) rather than truly offshore. More recently a suction caisson foundation was installed offshore at Horns Rev 2 Offshore Wind Farm (also in Denmark) as a foundation for a meteorology mast (LeBlanc, 2009). The suction installed foundation might be a preferred option to a monopile, not because they may use a lower amount of steel, but because the concept allows the installation contractor to use lighter duty installation processes. In fact it is possible, as appears to be the case for the meteorology mast foundation, to develop a floating, self-installing structure that does not require a heavy lift vessel. This option therefore has potential not only around the UK coast but also at other locations such as China, where there is an increasing demand for electricity from renewable energy sources.


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Figure 1. Suction caisson foundation for wind turbine.

For the monopod configuration the designer of the suction caisson must understand accurately the moment-rotation response. The monotonic response is important for developing ultimate capacity calculations and for developing appropriate stiffness parameters for structural dynamics calculations. The cyclic loading response is important for understanding accumulated deformations and changes in stiffness with

time. For design the accumulated angular rotation must be limited in the lifetime of the wind turbine. For the gravity base foundation at Thornton Bank the design rotation under cyclic loading was limited to 0.25◦ (Peire et al. 2009). In China the design codes require the cyclic accumulated angular rotation of the foundation to be less than 0.17◦ for onshore wind turbines, and there is no corresponding criterion for offshore turbines yet. Any changes in stiffness arising from the cyclic loading must also be well understood as there could be significant effects on the dynamic response of the turbine. Wind turbines are normally designed to have natural frequencies in the range between the frequency bands of the rotor rotation and the blade passing frequency, usually denoted by 1P and 3P. This avoids resonances, but changes in stiffness might result in interference between the first natural frequency and the excitation frequencies, 1P or 3P of the wind turbine. Research on suction caisson foundations has recently been carried out at the University of Oxford, Aalborg University and the University of Western Australia (Senders, 2008). This has concentrated on the monotonic response (Feld 2001, Houlsby et al. 2005, Villalobos 2006) and with a small amount of work on the cyclic response (though for only small numbers of cycles). There has been very little work looking at the long term response of caissons to cyclic loading. This paper addresses this gap by presenting results from cyclic moment loading experiments on suction caisson foundations. A novel cyclic loading rig, located at the University of Oxford, is described and some typical results from the tests are given. These results can be used to develop a framework to predict the longterm cyclic accumulated angular rotation of the suction caisson foundation.

2 TEST EQUIPMENT AND TEST PROGRAM 2.1

Cyclic loading rig and testing procedure

The loading rig is shown in Figure 2. A simple, motor driven, loading system is used to apply cyclic loading to the caisson. The rig consists of a soil container (0.55 m by 0.6 m by 0.6 m), a steel frame with pulleys, three weight-hangers and a lever with a driving motor. The lever is attached to the steel frame through a pivot and carries a motor, which rotates a mass m1 to provide cyclic loading on the foundation. The weights can be adjusted to impose different loading regimes. The motor is a geared single-phase AC motor rotating at a frequency of 0.106 Hz. The rig is very stable and can accurately provide a sinusoidal loading for a large number of cycles. The rig was originally developed by Rovere (2005) but more recently described and used by LeBlanc (2009) and LeBlanc et al. (2010) to carry out a series of cyclic loading tests on stiff piles (representing monopiles). The maximum number of cycles applied during LeBlanc’s work was 65,370.


Figure 2. Long-term cyclic loading rig. Table 1.

Redhill 110 properties (Villalobos, 2006).

Parameters

Values

D10 , D30 , D50 , D60 (mm)

0.08, 0.10, 0.12, 0.13 1.63, 0.96

Coefficients of uniformity, Cu and curvature Cc Specific gravity, Gs Minimum dry density, γmin (kN/m3 ) Maximum dry density, γmax (kN/m3 ) Critical state friction angle, φcs

2.65 12.76 16.80 36◦

For the tests described in this paper a load cell was used to monitor the overturning load on the caisson. Three LVDTs were also installed on the caisson to measure the rotation and deflection of the caisson. The diameter (D), the length of skirt (L) and the thickness (t) of the skirt wall of the caisson used was 0.2 m, 0.1 m and 0.001 m, respectively. The dimensions of the caisson were chosen to represent 1:75 model of a proposed suction caisson mono-foundation for an offshore wind turbine. A fine, silty sand was used for the experiments, as this is one of the more usual soil types at the sites of potential offshore wind farms in China. The sand was Redhill 110, a commercially produced sand, with properties given in Table 1. The sand bed and equipment were set up in a repeatable way to ensure similar conditions for each test. The sample-box was carefully filled with sand by pouring the dry sand from a very low drop height to achieve a loose state. The relative density of the fine sand was controlled to approximately 11%. The caisson was installed by a pushed installation rather than by suction, as the soil used was dry and not saturated. Previous work by Villalobos (2006) has provided guidance on the effects of suction installation, as compared to pushed installation, on the caisson response under loading. To ensure a vertical penetration of the caisson, in the centre of the soil container, a vertical rod within a guide was used. The load for the pushing installation was applied by

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Figure 3. Determination of ultimate moment capacity.

Figure 4. Cyclic loading ranges in relation to monotonic moment capacity of caisson.

adding weights to a loading hanger located centrally on the caisson by a point contact. The maximum vertical load applied to the caisson was 113 N which corresponds to a non-dimensional vertical load of V˜ pre = Vpre /(γ D3 ) = 1.07. Once installed the loading wires were connected to the vertical rod on the caisson. The vertical installation rig was carefully dismantled and the appropriate weight, m1 , was added to the hanger connected to the motor in order to apply the right amount of cyclic loading. The load cell and LVDTs enabled load and displacement data to be electronically recorded. The instruments were powered by a RDP M600 system which also allowed the resulting signals to be amplified. The electronic data was collected by a National Instruments USB-6009 digital card. The data collecting frequency was set at 2 Hz and the data were collected continuously during each cyclic test. Processing of the signals, by means of a digital filter, was also carried out to remove unwanted electrical noise. 2.2 Test program The test program consisted of both cyclic and monotonic testing of the caisson. The monotonic tests were carried out at different ratios of M /(HD) to determine the ultimate moment capacity of the caisson, MR . A typical moment-rotation curve is shown in Figure 3. To define the point of yield (or capacity) the method described by Villalobos (2006) was used. Straight lines are fitted to the initial elastic section and the later plastic section and the intersection is used to define the yield load. This exercise is repeated for different values of M /(HD) to define a yield envelope. Figure 4 shows the results of five tests as well as the line of best fit plotted through the experimental data points to represent the yield envelope, which is approximately straight in this region. Also shown on Figure 4 is a line of constant M /(HD) extending from the origin. This line is chosen to reflect the field case where typically the diameter may range from 15 m to 25 m, and the height of the resultant lateral load (both wind and wave) might be about 30 m (Byrne and Houlsby, 2003) which gives


Figure 5. Definition of cyclic parameters (from LeBlanc et al., 2010).

an eccentricity e = M /(HD) between 1.2 and 2.0. Along this line are plotted four important design loads for the wind turbine. These are: (i) the ultimate loadcarrying capacity related to the limit state ULS; (ii) the worst expected transient load related to ULS/1.35; (iii) the serviceability state SLS which occurs 102 times during the lifetime of the wind turbine; and (iv) the fatigue limit state FLS which occurs 107 times during the lifetime of the wind turbine (DNV, 2007). Devising the cyclic test program requires considerable care to limit the number of tests carried out. There is also a need for a consistent approach to defining the applied loading. LeBlanc et al. (2010) defined two parameters to characterize the applied cyclic load:

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Table 2. tests.

Normalized parameters of caisson in 1 g model

Parameters

Normalized expressions

Moment M Vertical load V Angular rotation θ Unloading stiffness in each cycle k Vertical displacement w

˜ = M /(γ D4 ) M ˜ V = V /(γ D3 ) θ˜ = θ(pa /γ D)0.5 k˜ = k/[(γ pa )0.5 D3.5 ] w˜ = (w/D)(pa /γ D)0.5

Figure 7. Angular rotation calculated by displacements.

Figure 6. Displacements measured by LVDTs.

Here Mmax and Mmin are the maximum and minimum moment in the load cycle as shown in Figure 5, and MR refers to the static capacity at the selected ratio of M /(HD). As illustrated in Figure 5 the dimensionless parameter ζc denotes the type of loading ranging from 0 for one-way loading to −1 for two-way loading. A test program was therefore devised that considered a range of these parameters to explore the accumulated deformation behavior of the caisson. By carrying out a dimensional analysis it is possible to develop a number of normalized groups with which to present the results and these are shown in Table 2. 3 3.1

CYCLIC MOMENT LOADING TEST RESULTS Cyclic accumulated deformation

Some typical test results are shown in Figures 6 through to 9. Figure 6 shows the raw displacement data as collected with time. By accounting for the geometry of the set-up it is possible to deduce the angular rotation and the settlement of the caisson and this is shown in Figures 7 and 8 respectively. Together with the data of continuous cyclic loads, the plot of the normalized angular rotation in response to the applied moment is given in Figure 9. It can be observed that the first loading cycle generates a larger displacement than the following ones, and the accumulated cyclic angular


Figure 8. Angular rotation calculated by displacements.

rotation increases with the number of cycles but that the angular rotation specific to one cycle decreases with the number of cycles. In a similar way to a piled foundation (Lin & Liao 1990, LeBlanc 2009) the cyclic accumulated angular rotation of the caisson can be evaluated in terms of the dimensionless ratio (θ N − θ0 )/θ s . Here θ N and θ 0 are illustrated in Figure 5 and θ s is the angular rotation that would be experienced in a monotonic test if the maximum load under cycling were applied. The relationship between (θ N − θ 0 )/θ s and the number of cycles for different characteristics of the cyclic load is shown in Figure 10. It would appear that the dimensional ratio (θ N − θ 0 )/θ s is almost linear with the number of cycles in logarithmic coordinates within the 10,000 cycles applied. The normalized accumulated settlements of the caisson for different ζ b are shown in Figure 11. The values increase with the increasing cycles and amplitude of the cyclic loads. Settlement of the caisson foundation is of course also an important consideration for this application.

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Figure 11. Cyclic accumulated settlement against cycle number.

Figure 9. Relationship between moment and angular rotation.

Figure 12. Instantaneous centre for monotonic and cyclic tests.

position of the instantaneous centre of rotation. For a common eccentricity e = M /(HD) = 1.875, the movements of the instantaneous centre for monotonic and cyclic tests are shown in Figure 12. For the monotonic test, the centre moves from beneath the skirt to the horizontal plane at the skirt tip at the end of the test. For the cyclic loading test the instantaneous centre starts at a much greater depth below the caisson and does not change much during the test.

3.3

Figure 10. Normalized accumulated angular rotation against cycle number.

3.2 Rotation of the caisson During both monotonic and cyclic moment tests it is possible, given the displacements x1 , y1 and y2 measured by three LVDTs (see Figure 1), to deduce the © 2011 by Taylor & Francis Group, LLC

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Unloading stiffness

If the foundation stiffness changes during the lifetime of the wind turbine then there might be a concern that dynamic characteristics of the structure will be altered. In particular the 1st natural frequency of the structure might start to move towards one of the excitation frequencies. Figure 13 presents curves of normalized unloading stiffness for three typical cyclic tests. It can be observed that the unloading stiffness of the caisson is constant irrespective of the accumulated angular rotation, for the number of cycles explored here. For these tests the stiffness appears to decrease with increasing maximum cyclic load.

ACKNOWLEDGEMENT This work was supported by National High-Tech R&D Program of China (863 Program) (research grant: 2007AA05Z427) and National Natural Science Foundation of China (research grant: 50979097). The first Author acknowledges the support from the Department of Engineering Science at Oxford. REFERENCES

Figure 13. Unloading stiffness in relation to angular rotation.

4

CONCLUSIONS

Suction caisson foundations might be used as monopod foundations for offshore wind turbines. One of the key issues will be the accumulated rotation over the lifetime of the turbine and whether this can be predicted. This paper describes a series of tests aimed at understanding the long term cyclic loading response of suction caissons. A unique loading rig for applying cyclic loading to foundations was described. A set of experimental results were presented and the test results were normalized. A framework for understanding the cyclic accumulated angular rotation and unloading stiffness of the caisson was introduced. As would be expected the accumulated angular rotation of the caisson is significantly affected by the characteristics of the applied cyclic load. The dimensional ratio of accumulated rotation, (θ N − θ0 )/θ s , was found to be almost linear with the logarithm of the number of cycles within the limits of 10,000 cycles applied. Further work is required to explore whether this relationship can be used to predict the cyclic accumulated angular rotation of the caisson during an ocean storm or the lifetime of the wind turbine. This will require tests to larger numbers of cycles. In addition, centrifuge and full scale tests as well as tests on a higher relative density sand would be required to provide added confidence in the scalability of these results.

Byrne, B. W. & Houlsby, G. T. 2003. Foundations for offshore wind turbines. Phil. Trans. R. Soc. Lond. A, 361: 2909– 2930. LeBlanc, C. 2009. Design of Offshore Wind Turbine Support Structures. PhD Thesis, Aalborg University, Denmark. LeBlanc, C., Houlsby, G.T. and Byrne, B.W. 2010. Response of stiff piles in sand to long term cyclic loading. Geotechnique, 60(2): 79–90. DNV. 2007. Offshore Standard (DNV-OS-J101): Design of Offshore Wind Turbine Structures. Det Norske Veritas, Hovek, Norway. Feld, T. 2001. Suction Buckets, a New Innovative Foundation Concept, applied to Offshore Wind Turbines. PhD Thesis, Aalborg University, Denmark. Houlsby, G. T. Ibsen L. B. & Byrne B. W. 2005. Suction caissons for wind turbines. In Gourvenec & Cassidy (eds.), Frontiers in Offshore Geotechnics: ISFOG, Perth, Australia, 19–21 September 2005. London: Taylor & Francis. Lin, S. S. & Liao, J. C. 1990. Permanent strains of piles in sand due to cyclic lateral loads. Journal of Geotechnical and Geoenvironmental Engineering, ASCE. 125(9): 798–802. Kelly, R. B., Houlsby, G. T. & Byrne, B. W. 2006. A comparison of field and laboratory tests of caisson foundations in sand and clay. Geotechnique 56(9): 617–626. Peire, K. Nonneman, H. & Bosschem, E. 2009. Gravity based foundations for the Thornton Bank Offshore Wind Farm. Terra et Aqua, 115: 19–29. Rovere, M. 2005. Cyclic loading test machine for suction caisson foundations. Project Report. Politecnico di Milano. Senders, M. 2008. Suction caissons in sand as tripod foundations for offshore wind turbines. PhD Thesis, The University of Western Australia, Australia. Villalobos, F. A. 2006. Model Testing of Foundations for Offshore Wind Turbines. DPhil Thesis, University of Oxford, UK.


10 Pipelines and risers



Multidirectional analysis of pipeline-soil interaction in clay R.G. Borges PETROBRAS Research & Development Centre, Rio de Janeiro, Brazil

J.R.M.S. Oliveira Military Institute of Engineering, Rio de Janeiro, Brazil

ABSTRACT: This paper discusses the interaction, in all directions, between buried pipes and the surrounding soil. In numerical terms, this interaction is normally modelled via separate elasto-plastic springs positioned in the main directions (vertical and horizontal). Many experimental results are available in the literature dealing with soil-pipe interaction in the main directions on a plain submarine slope. Equations are accessible to calculate the maximum soil resistance for design purposes. However, the direction of movement of the pipeline is not taken into account and, depending on the soil characteristics, it may lead to a sub-estimation of the confinement forces. A comprehensive set of numerical analyses was carried out to study the soil response in all directions. The software SIGMA/AEEPECD was used in these Finite Element analyses. A reference curve for soil response in any directions was achieved. The numerical results were compared with a standard for pipe-soil interaction, showing good agreement. 1

INTRODUCTION

seabed, which can induce overloads that can cause the pipeline to rupture.

Offshore pipelines in shallow waters are generally buried to provide mechanical protection and constraint. A safe buried pipeline design must take into account a reliable evaluation of the pipe-soil interaction forces and the associated displacements, in order to assure structural integrity during operation. Simplified usual design practice (ALA, 2005) based on Hansen (1961), assumes the pipe as a beam and the soil restraint as axial and transversal Winkler springs placed along its external body. The properties of these springs vary with direction and position along the pipeline, showing no relationship with each other. Many authors have proposed theoretical, numerical and experimental approaches to investigate the soil-pipe interaction. Among those who have chosen the numerical simulation, Fernando & Carter (1998), Popescu & Konut (2001), Popescu et al. (2002), Guo & Popescu (2002), Oliveira et al. (2005) and Borges (2008) focused on horizontal or vertical pipe-soil interaction. Furthermore, Zhang et al. (2002) and Calvetti et al. (2004) studied failure envelopes for pipelines subjected to combined force and displacement loadings both experimentally and numerically. In that way, the main purpose of this work is to investigate the soil response envelope for any direction the pipe may take between fully downwards and upwards. Such loading happens when pipelines are installed in harsh environments, such as along an unstable slope, crossing an active fault plane, or buried in soil layers likely to suffer liquefaction upon the occurrence of an earthquake. When one of these conditions occurs, the pipeline can be affected by relative movements of the

2

Numerical analyses using Finite Element Method (FEM) were used in this research to investigate the soil-structure interaction behaviour of an offshore steel pipe buried in marine clay, considering several burial depths and a comprehensive set of displacement directions, all in monotonic conditions. 2.1 The SIGMA/AEEPECD software For the numerical simulation of the non-linear physical behaviour of the soil, a continuous medium was considered, which demanded the application of interactive incremental integration algorithms. The software SIGMA v. 5.32 (Amaral et al., 1997, LIRA, 1998), was used for the bi-dimensional Finite Element discretization and model post-processing. This system is based on the integration of numerical analysis modules developed by the Scientific Methods Team of Petrobras Research and Development Centre, with graphical tools conceived by an agreement signed between Petrobras and Tecgraf/PUC-Rio. These tools consist of programs used for geometric modelling, specification of attributes, finite element mesh generation and results visualization provided by the numerical simulators. Such tools were used as pre- and postprocessors of the models, assisting in the modelling of the Engineering problem. The version 3.02 of the numerical simulator AEEPECD (Cardoso, 2005), was used to solve the


NUMERICAL MODELLING

Figure 1. Espadarte Field. Offshore Brazil. Fugro (2005).

equilibrium of continuum differential equations. This software allows linear and non-linear analyses of plane strain, plane stress and axi-symmetric models. The program uses constitutive laws for plastic and elastoplastic rheologies. The adopted plastic criteria are Von Mises and Mohr-Coulomb, incorporating geometrical and physical non-linearities. 2.2

Soil material properties

The soil strength profile was based on the undrained shear strength (su ) results from in situ and laboratory tests in a Southeastern Brazilian coast marine site. Piezocone penetration tests (PCPT) were undertaken in Espadarte marine field (Figure 1), using the Deepwater Seacalf System as well as laboratory tests in Jumbo Piston Core (JPC) samples (Costa et al., 2002). The JPC samples were obtained using a 20 m long piston launched in free fall 2.5 m above the seafloor. Afterwards, the soil undisturbed samples were conditioned in special cases and taken to the laboratory. Data from the PCPT tests were interpreted using the methodology suggested by Amaral & Costa (1998). Based on the results obtained from the set of tests in Espadarte oil field, the geotechnical borehole GT212 was chosen as a minimum strength profile, with a water depth of 963.39 m, as shown in Figure 2: Equations (1) and (2) show total and effective stress regression lines, respectively:

Figure 2. Geotechnical strength profile for borehole GT-212.

The total strength regression line was chosen to simulate the soil resistance with depth. The mean value of the submerged unit weight was adopted from the test JPC-212 (γ = 5.45 kN/m3 ). The earth pressure coefficient k0 was assumed as 1, the Poisson ratio ν = 0.49 and the soil Young modulus (E) was obtained from Equation (3), proposed by Amaral et al. (2002):

The mobilized shear strength in the soil-pipe interface (τ u ), was calculated using Equation (4):

Where r is an interface parameter which varies from 0 (perfectly smooth surface) to 1 (perfectly rough surface). In this work, a mean value of r = 0.5 was used to simulate an intermediate situation. The non-linear behaviour of the foundation soil was represented by the Mohr-Coulomb plastic yielding model, considering undrained conditions, which means that the internal friction angle is zero. This implies that the Mohr-Coulomb plastic yielding model turns into the Tresca criterion.


Table 1.

Geometrical and mechanical properties of the pipe.

Property

Value

External Diameter (D) Thickness (e) Length (L) Young Modulus (E) Poisson’s Ratio (ν) Yielding Stress (σy ) Ultimate Stress (σu )

0.60 m 0.0254 m Infinite 2.06 108 kPa 0.30 3.58 105 kPa 4.55 105 kPa Figure 4. FE spatial discretization for H/D = 50%.

Figure 3. Definition of angle α and burial depth H .

2.3 Pipe material properties

Figure 5. Displacement vector distribution for H /D = 100% and α = 45◦ .

Table 1 resumes the main geometrical and mechanical properties adopted for the weightless empty pipe:

depth. A typical mesh for this problem, along with the applied displacement boundary conditions, is show in Figure 4 (Borges, 2009): The total lateral and vertical dimensions of the models were adopted as 8 m and 4 m, respectively. These values were considered large enough to avoid any boundary influence.

2.4 Description of the simulation models Small strain analyses were performed on the preembedded pipe, considering six different burial depths (H /D = 25, 50, 75, 100, 150 and 200%), in order to obtain the soil response for a pipe subjected to constant incremental displacements in various directions, which are defined by the angle α with the vertical (Figure 3). The chosen angles were α = 0, 15, 30, 45, 60, 75, 90, 105, 120, 135, 150, 165 and 180◦ , leading to a total number of 78 processed analyses. The finite element mesh was composed by isoparametric quadrilateral quadratic elements with 8 nodes, and the pipe-soil contact elements were composed by special interface elements with 6 nodes and quadratic variation of the relative displacements. The soil-pipe contact was represented by special interface elements, assuming a non-linear constitutive law for the shear stress-strain relationship, and a Mohr-Coulomb criterion for the contact material. In these interface elements, it was considered the nonlinear behaviour of the soil in the tangential direction, assuming a maximum shear stress equal to the soil undrained shear strength. In the normal direction, it was considered a linear behaviour. The actual distribution and concentration of elements varied as a function of the pipe embedment


3 3.1

SIMULATION RESULTS Displacement vectors

For better visualization of the deformation mechanisms occurring within the soil mass, Figures 5 and 6 show the displacement vectors for H /D = 100% and direction angles α = 45◦ and 135◦ . It was noticed the formation of a gap behind the pipe (because separation is permitted), but the analyses performed didn’t consider large displacements/large deformations. It could only be captured the initial pipe-soil interaction behaviour when the maximum resultant force was mobilized. The associated maximum displacements were small (4.5% of the pipe diameter).

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3.2

Resultant force versus displacement curves

Figures 7 and 8 show the resultant force FR (FR = (F2H + F2V )1/2) versus displacement δ relationship for

Figure 9. Resultant forces with α angle and H /D ratio.

Figure 6. Displacement vector distribution for H /D = 100% and α = 135◦ .

Figure 10. Comparison between SIGMA/AEEPECD lateral forces and ALA (2005).

Figure 7. Resultant force versus displacement for α = 45◦ .

For a better evaluation of the resultant force variation with α and H /D ratio, the Figure 9 is presented below. It can be seen that the resultant forces increase with the α angle, and also with the pipe embedment ratio H /D. The smallest resistance offered by the soil against the pipe motion occurred for purely vertical upwards pipe movement, because of the lesser weight of soil to be displaced by the pipe, and the highest one was the vertical downwards. 3.3 ◦

Figure 8. Resultant force versus displacement for α = 135 .

six distinct H /D ratios and direction angles α = 45◦ and 135◦ , respectively: The results show that for α = 45◦ the full mobilization of soil resistance happens in earlier displacements than for α = 135◦ . This fact is probably associated with the relevant increase in soil mass involved in failure process from α = 45◦ to 135◦ , as can be seen in Figures 5 and 6.

ALA (2005) presents a series of equations to represent the soil springs in axial, lateral and vertical (upwards and downwards) directions. However, those equations only consider α = 0, 90 and 180◦ . Figure 10 shows a comparison between the lateral force values obtained in this work with SIGMA/ AEEPECD software and ALA (2005): The results show that ALA (2005) data for lateral response (α = 90◦ ) agree well with SIGMA/ AEEPECD. However, according to Borges (2009), the comparisons for vertical directions (α = 0 and 180◦ ), ALA (2005) values are lower than those obtained


Comparison with ALA (2005)

Figure 11. Failure envelopes for a pipeline buried in clay.

Figure 12. Failure envelopes normalized by Equation (5).

with SIGMA/AEEPECD, suggesting that the first methodology is more conservative. 3.4 Failure envelopes The resultant loading paths from the numerical analyses of combined displacements in the H-V plane are shown in Figure 11. They follow the same trend of those previously reported for shallow foundations (Bransby & Randolph, 1998), and for “wished-inplace” pipelines (Merifield et al., 2008). In addition, arrows in this chart indicate the directions of pipe motion, indicating the resultant vector of incremental displacements. By plotting the peak values for each loading component in the H-V plane, it is possible to recognize the influence of the relative depth H /D and the direction of movement of the pipeline. Besides that, through Figure 11 it is easy to identify a remarkable coupling effect between the horizontal and vertical forces, and one can define the soil-pipe interaction coupled domain through these failure envelopes. These results show how the soil-pipe interaction in different monotonic loading directions can be conveniently approximated by numerical models using the Finite Element Method, considering the undrained behavior and an elasto-plastic constitutive model for the clayey soil. Horizontal and vertical components of the resultant forces in buried pipes in clay are usually normalized by the soil undrained strength at the pipe invert and pipe diameter, in terms of the Equation (5):

Figure 12 shows the normalized failure envelopes for H /D from 25% to 200%, where a clear increase in horizontal and vertical forces can be observed: It was observed that the failure envelopes are not symmetrical around the horizontal plane, because the undrained shear strength of the soil increases linearly with depth in the models. The maximum horizontal © 2011 by Taylor & Francis Group, LLC

Figure 13. Failure envelopes normalized by the maximum values.

force occurs at α = 90◦ , and the maximum vertical force at α = 180◦ , as expected. Figure 13 presents the same horizontal and vertical force values presented in Figures 11 and 12 but, this time, they were normalized by their respective maximum values. The maximum horizontal force assumed for each H /D ratio was the value associated with α = 90◦ . For the maximum vertical force, the values associated with α = 0◦ and α = 180◦ were adopted for the upward and downward movements, respectively. The result is a set of superimposed curves that can be considered roughly as a single curve. Based on these data, Equation (6) allows to calculate the horizontal and vertical force components acting on a buried pipe moving towards any direction:

4

CONCLUSIONS

This work studied the soil-structure interaction for the case of pipelines subjected to combined vertical and horizontal loadings when buried in a clayey soil in

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the submerged condition. In other words, results of a numerical investigation on the resistance of the soil against the oblique motion of a pipe in marine clay were presented. The failure envelopes, drawn with the purpose to establish a basis for estimating the maximum resistance offered by the soil, tried to reproduce the force-displacement response for a steel buried pipe in marine clay, obtained from finite element analyses of a pipe with intermediate roughness considering several embedment/diameter ratios. The software SIGMA/AEEPECD, developed by PETROBRAS, was used in these finite element twodimensional analyses to achieve a set of failure envelopes that is able to describe the mobilized soil resistance behaviour during pipe movement. It was found that there is a relationship between the vertical and horizontal loading directions. The curves obtained in this work are similar to those achieved by Bransby and Randolph (1998) for shallow foundations, but capture the additional effects due to the curved shape of the pipe surface. Also, the results were comparable to those obtained by Merifield et al. (2008) for buried pipes. Based on Equation (6), it was possible to determine the horizontal and vertical force components acting on a buried pipe moving towards any direction. The numerical simulations proved to be useful as a pseudo-experimental tests in order to gauge the shape and size of the soil-pipe interaction domain, when displacement directions between the vertical and horizontal are imposed to the pipe. Even though these relationships must be used with criteria, they proved to be useful in design practice, allowing quick access to complex simulations. ACKNOWLEDGEMENTS The authors would like to gratefully acknowledge to TRANSPETRO and PETROBRAS/CENPES for the sponsorship and support, respectively, as well as all people involved in this research program. REFERENCES Amaral, C.S., Costa, A.M., Carvalho, M.T.M. et al. 1997. Description of SIGMA System – Geotechnical Integrated System for Multiple Analyses (in Portuguese), RT SEDEM n◦ 007/1997, PETROBRAS/SUPEN/DIPREX/SEDEM, Rio de Janeiro, Brazil. ALA. 2005. Guidelines for the Design of Buried Steel Pipe. American Lifelines Alliance. http://www.americanlifelinesalliance.org/pdf/ Update061305.pdf. Amaral, C.S. & Costa, A.M. 1998. Methodology for Interpretation of Wilson Piezocone Tests using the Seacalf Penetration System (in Portuguese). Partial Report 600.234 PETROBRAS/CENPES/PDP/MC, Rio de Janeiro, Brazil.


Amaral, C.S., Costa, A.M., Cardoso, C.O. 2002. Soil Reaction due to Riser Displacement in Roncador Field, Partial Report 600.728 PETROBRAS/CENPES/PDP/MC, Rio de Janeiro, Brazil. Borges, R.G. 2008. Guidelines for Soil-Pipe Interaction Design–Phase 1 (in Portuguese), PETROBRAS/CENPES/PDP/MC, RT MC n◦ 062/2008, Rio de Janeiro, Brazil. Borges, R.G. 2009. Numerical Modelling of the Pipe-Soil Relative Movement in the Oblique Direction in Clay (in Portuguese), PETROBRAS/CENPES/PDP/MC, RT MC n◦ 086/2009, Rio de Janeiro, Brazil. Bransby, M.F. & Randolph, M.F. 1998. “Combined Loading of Skirted Foundations”, Géotechnique, v. 48, n. 5, pp. 637–655. Calvetti, F., Di Prisco, C., Nova, R. 2004. “Experimental and Numerical Analysis of Soil-Pipe Interaction”, Journal of Geotechnical and Geoenvironmental Engineering, ASCE, v. 130, n. 12, pp. 1292–1299. Cardoso, C. O. 2005. Methodology for the Analyses and Design of Submarine Pipelines Subjected to High Pressures and Temperatures via Finite Element Method (in Portuguese), D.Sc. Thesis, COPPE/UFRJ, Rio de Janeiro, Brazil. Fernando, N.S.M. & Carter, J.P. 1998. “Elastic Analysis of Buried Pipes under Surface Patch Loadings”. Journal of Geotechnical and Geoenvironmental Engineering, v. 124, n. 8, pp. 720–728. Fugro. 2005. Geotechnical Investigation – Espadarte and Espadarte Sul Fields, Campos, Offshore Brazil. FugroMcClelland Marine Geosciences, Houston, Texas, USA. Guo, P. & Popescu, R. 2002. “Trench Effects on Pipe/Soil Interaction”, In: Proc. 2nd Canadian Specialty Conference on Computer Applications in Geotechnique, Winnipeg, Canada, pp. 261–269. Hansen, J.B. 1961. The Ultimate Resistance of Rigid Piles Against Transversal Forces, The Danish Geotechnical Institute, Bulletin n. 12, pp. 5–9, Copenhagen, Denmark. Lira, W.W.M. 1998. An Integrated Configurable System for Simulations in Computational Mechanics (in Portuguese). M.Sc. Thesis, PUC-Rio, Rio de Janeiro, Brazil. Merifield, R., White, D.J., Randolph, M.F. 2008. “The Ultimate Undrained Resistance of Partially Embedded Pipelines”, Géotechnique, v. 58, n. 6, pp. 461–470. Oliveira, J.R.M.S., Almeida, M.S.S., Almeida, M.C.F. et al. 2005. “Physical and Numerical Modelling of Lateral Buckling of a Pipeline in Very Soft Clay”, In: Proc. International Symposium on Frontiers in Offshore Geotechnics – ISFOG 2005, Perth,Australia, pp. 607–613. Popescu, R. & Konuk, I. 2001. “3D Finite Element Analysis of Rigid Pipe Interaction with Clay”, In: Proc. 10th International Conference on Computer Mechanics Advances in Geomechanics, Tucson, Arizona, v. 2, pp. 1203–1208. Popescu, R., Phillips, R., Konuk, I. et al. 2002. ”Pipe-Soil Interaction: Large-Scale Tests and Numerical Modeling”, In: Proc. International Conference on Physical Modelling in Geotechnics – ICPMG’02, St. John’s, Newfoundland, Canada, pp. 917–922. Zhang, J., Douglas, P.S. & Randolph, M.F. 2002. “Modelling of Shallowly Embedded Offshore Pipelines in Calcareous Sand”, Journal of Geotechnical and Geoenvironmental Engineering, ASCE, v. 128, n. 5, pp. 363–371.

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Geotechnical challenges for deepwater pipeline design – SAFEBUCK JIP D.A.S. Bruton, M. Carr & F. Sinclair Atkins Boreas, UK

ABSTRACT: When the SAFEBUCK JIP started research into the interaction between pipelines and marine clays in 2002, no one could have predicted how that research would impact deepwater pipeline design, or how far the research test methods would develop and improve. The complexity that lies behind the lateral and axial pipe-soil response in soft clays has led to a radical overhaul of established geotechnical practice, based on a significant research effort to evaluate, quantify and understand these pipe-soil interaction mechanisms. This work has led to a number of key insights into the fundamental pipe-soil behaviour and the development of models to simulate this behaviour. This paper demonstrates why pipe-soil interaction is so important to pipeline design, and why it is the largest uncertainty that pipeline designers face in addressing the challenges associated with lateral buckling and pipeline walking; with serious implications for field layouts, pipeline configurations and mitigation solutions.

1

INTRODUCTION

When the SAFEBUCK JIP started research into the large displacement cyclic response of pipelines on very soft deep-water marine clays in 2002, it would not have been possible to predict how that research could now impact pipeline design, or how far the research test methods would develop and improve. From the beginning, this JIP pioneered the idea of small-scale lateral pipe-soil tests in a centrifuge, which is now established as a reliable evaluation method for new projects. Neither could the JIP have envisaged the development of a large-scale in-situ test rig called SMARTPIPE® (Hill & Jacob 2008), using the lessons learned from large-scale laboratory tests pioneered by SAFEBUCK. The complexity that lies behind the lateral and axial pipe-soil response in soft clays has led to a radical over-haul of established geotechnical practice in pipesoil interaction. The JIP and many recent projects have invested significant research effort to evaluate, quantify and understand these pipe-soil interaction mechanisms. This work has led to a number of key insights into the fundamental pipe-soil behaviour and the development of models to simulate this behaviour. There have also been some surprises along the way, for example, the observed link between pipe-soil friction response and the flow regime inside the pipeline. This paper demonstrates why pipe-soil interaction is so important to design, and why it is the largest uncertainty that deepwater pipeline designers face. The pipe-soil response affects the design limit states associated with lateral buckling, pipeline walking, route-curve pullout and flowline anchoring. It can also influence the choice of flowline configuration, for example by choosing pipe-in-pipe, increasing the wall thickness or modifying the coatings. These design


issues can also affect the layout and architecture of new developments, requiring careful definition of pipeline routes, bathymetric profiles and the location of production facilities, as well as the design of pipeline crossings and tie-ins to risers and subsea structures. Lateral buckling and pipe walking behaviour is extremely sensitive to pipe-soil interaction forces and there is significant uncertainty associated with the characterisation of these forces in design. A rigorous theoretical understanding of the basic phenomena involved has only recently emerged, principally through research activity associated with the SAFEBUCK JIP.

2

DESIGN CHALLENGES

Any pipeline which is subjected to above ambient temperatures and pressures has a tendency to relieve the resulting high axial stress in the pipe wall by expanding longitudinally. This expansion is resisted by the axial soil resistance between the pipe and the seabed. This restraint causes an axial compressive force to develop in the pipeline, which can cause buckling. Where pipelines are laid on the seabed without trenching, there is no lateral or uplift restraint acting on the pipeline to prevent buckling, apart from the lateral resistance between the pipe and the soil. A key challenge for the design of such pipelines is the control of lateral buckling, pipe walking, or route curve instability. If left uncontrolled, these behaviours can have serious consequences for the integrity of a pipeline. Uncontrolled lateral buckling can lead to high strains and cyclic loads at the buckle crown, which can compromise the integrity of a pipeline, due to

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levels of 1 to 10 kPa, associated with pipeline submerged weights; (ii) the near-surface soils to a depth of one or perhaps two metres of the seabed; (iii) more significant disturbance of the seabed, from the pipe embedment process and large-amplitude lateral and axial movements. It is important to bound the soil properties as tightly as possible (Bruton et al., 2007), to minimise the range of lateral and axial friction response.

3.1

Figure 1. Engineered buckle initiators.

local buckling of the pipe section, weld fracture or fatigue failure. Careful consideration is also required to prevent unplanned ‘rogue’ buckles forming at inline structures or pipeline crossings. Lateral buckling is therefore controlled using engineered buckle initiators (Figure 1) placed at regular intervals along each flowline, thereby triggering regular lateral buckles and minimizing the lateral buckle loads (Sinclair et al., 2009). Pipe-walking can occur when a pipeline is subjected to thermal cycling while tension is applied at one end by a riser, or the pipeline is laid down a slope or is subjected to steep thermal transients during shutdown and restart cycles (Carr et al., 2006 & Bruton et al., 2010). Over a number of cycles, this movement can lead to very large global axial displacements, with associated overload of the connections. Where the predicted pipe walking displacements threaten system integrity, walking will be controlled by the use of pipeline anchors, typically installed at the end of the pipeline from which it is walking. However, pipeline anchors result in very high levels of tension at shutdown, which can lead to route-curves becoming unstable. This is usually overcome by increasing the routecurves radius, or removing route curves altogether by changing the field layout.

3

GEOTECHNICAL UNDERSTANDING

Conventional methods of geotechnical investigation and analysis that have evolved for foundation design are not suited to the analysis of pipe-soil behaviour. Compared with conventional foundation design, pipesoil interaction is concerned with (i) far lower stress


Geotechnical survey data – from the field

To narrow the range of soil properties, it is essential that a detailed soil investigation provides undrained shear strength profiles, sensitivity, and unit weight in as much detail as possible. Tests carried out in-situ on softer soils (or box core samples of those soils) should include T-bar tests, including cyclic loading to quantify the remoulding behaviour. It is also critical to provide a good datum reference to the original surface of the seabed. Often data is truncated to remove very weak surface soils (or the weak slurry is removed from the box-core before testing) and yet this information is fundamental to the assessment of pipeline embedment, which affects every aspect of the subsequent pipe-soil response.

3.2

Project specific testing and interpretation

It is common now to supplement geotechnical survey investigations with project-specific model test programmes, to assess the axial and lateral pipe-soil response. Project-specific testing requires the collection of bulk samples of near-surface soils for subsequent laboratory tests and pipe-soil modelling studies (Bruton et al., 2009). The SAFEBUCK JIP reviewed historical data and conducted a number of pipe-soil interaction test programmes to provide generic guidance for future projects and improve current understanding (Bruton et al., 2006). Although some of the historical test data had limited information on some of the test parameters, the proposed formulations from this work remain sound for relatively light pipes. However, pipes that are relatively heavy have a significant influence on the cyclic response (Bruton et al., 2008). ‘Heavy’ and ‘light’ pipes are distinguished by the ratio of the pipeline weight to the seabed strength. In simple terms, values of V/Su .D 2.5 give a ‘heavy-pipe’response, characterised by the pipe diving with displacement. Data from the early tests has since been augmented by a large quantity of high quality project-specific tests donated by JIP participants and tests carried out by SAFEBUCK, to create a database that now spans a very wide range of pipe diameters, weights and soil conditions. A detailed review of this lateral friction response database was recently issued to participants (White & Cheuk, 2010).

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It was also clear from this review that the quality of test data has improved significantly over recent years, with improved test equipment and testing methods, based on lessons learned from earlier tests. For example, SAFEBUCK pioneered the idea of small-scale lateral pipe-soil tests in a centrifuge and this type of test is now established as a reliable evaluation method for new projects, requiring smaller soil samples and much reduced time-scales for testing. The lessons learned have also been captured in large-scale laboratory tests and, more recently with the development of a largescale in-situ test rig called SMARTPIPE® (Hill & Jacob, 2008). 3.3

Remaining uncertainties

Nevertheless, much uncertainty remains in key areas of pipe-soil interaction response: 1. Complex lateral frictional response of pipe that is heavy in relation to the soil shear strength 2. Predicting axial friction in soft clays. 3. The influence of internal flow regimes on lateral buckling, pipe-walking and route-curve pull-out These issues are discussed in the following sections. 4

‘HEAVY’ PIPE RESPONSE

Increasing the flowline wall thickness increases its bending stiffness and capacity but this also increases the load in the pipeline due to the lateral resistance increasing with pipe weight. The restrained thermal expansion force also increases with wall thickness but the overall influence of increasing wall thickness at a lateral buckle is not obvious and depends upon the lateral friction response. If the lateral resistance is proportional to the pipe download, then the buckling becomes more severe as the wall thickness increases. However, if the lateral resistance does not change with the pipe download, then the buckling becomes less severe as the wall thickness increases. The use of pipe-in-pipe (PIP) systems has significant advantages for design over more traditional externally-insulated systems. Although PIP systems are usually more expensive per metre than external insulation systems and are much more complex to assemble offshore, they offer advantages in terms of improved thermal insulation and enhanced lateral buckling and pipe walking performance. The good thermal performance of PIP systems is well known. The ability to tailor the bending stiffness and moderate the axial load by increasing the outer pipe wall thickness (which usually experiences little thermal loading) is also known to greatly assist lateral buckling design. Finally the increased submerged weight of such systems (in comparison with externally insulated pipe) can greatly reduce or eliminate the propensity for pipe walking, due to the high weight and increased axial soil resistance. Thus they may actually offer a cost advantage. However, the use of heavy pipe-in-pipe systems introduces an extremely challenging lateral © 2011 by Taylor & Francis Group, LLC

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Figure 2. Cyclic embedment at a lateral buckle.

cyclic response into the design process (Bruton et al., 2009). Under conditions of constant vertical load, a laterally sweeping ‘heavy’ pipeline will rapidly embed. However, in the areas where there is little cyclic movement, such as at the stationary inflection point in a lateral buckle, there is little embedment. This causes the contact pressure to vary, so that the rate of embedment and the lateral soil resistance varies along the length of the buckle. Therefore, in regions of lateral sweeping the contact force and lateral soil resistance gradually reduce as the trench gets deeper and soil berms are established at the extremes of displacement. Consequently, the friction curves formulated using the nominal pipe weight can significantly overestimate the lateral resistance at the crown of the buckle. The structural response of heavy pipe therefore required a totally new approach, to capture the reducing contact pressure along the buckle as the lobes of the buckle displace laterally and penetrate vertically. Meanwhile, the regions between the lobes experienced little lateral movement and sustained increasing contact pressure as the load is transferred to these regions. Modelling of the changing seabed elevation as the pipe sweeps and displaces soil to form a trench is extremely challenging. Figure 2 shows a section through a lateral buckle model, where the excavation of the seabed is captured at the end of each sweep of the pipe, based on predictions from experimental data that account for variation in contact pressure and the sweep amplitude. This complex modelling approach can be used to assess the cyclic response for a selected pipe weight and pipe bending stiffness. As the lateral buckle displaces cyclically in operation, the analysis defines the shape of the lateral buckle and the variation of contact pressure along it. To do this, sufficient knowledge of the rate of vertical displacement and lateral resistance for the given pipe weight and cyclic amplitude is required from project specific tests. This approach has been used to verify a simpler method of analysis, using re-formulated friction curves to suit more traditional lateral buckling models that employ an elastic seabed surface that is not deformed by cyclic loading. The re-formulated friction curves exhibit a heavy pipe response in the early cycles which gradually migrates towards a light pipe response, as the contact pressure reduces with

increasing embedment. Work is ongoing to simplify this necessarily complex approach but the ultimate solution is a better way of modelling lateral pipe soil interaction in finite element analysis (FEA), which is a key aim of the SAFEBUCK GEO JIP, described further in Section Section 7. A major challenge to designers, associated with changing operating conditions and cyclic loading, is the predicted increase in operating temperature late in field life. Often this is combined with increasing density of the production fluids as the level of produced water increases. This combination of increasing load and lateral resistance may cause a pipeline to break through an established soil berm at the apex of a lateral buckle, causing strain localisation in the pipe. Assessing this potential failure condition requires a very good understanding of the gradual growth in berm strength and the ability to model cyclic response of the lateral buckle over many operating cycles.

pipelines, is the influence of internal flow mechanisms on the axial friction response. On a relatively light pipeline with external insulation, the weight of the pipe can vary significantly as alternate slugs of gas and liquid pass along the pipeline. This frequent change in contact pressure is likely to affect the excess pore pressure in the soil around the pipeline, causing changes in the axial friction response. Observations of operating pipelines has certainly demonstrated very low levels of axial friction in operation when regular slugging occurs, compared with much higher levels of axial friction on shutdown when slugging has ceased. This is the only plausible explanation for some observed end expansion behaviour and is thought to be responsible for high levels of tension at shutdown that has resulted in route-curve instability.

6

PREDICTING PIPELINE WALKING AND END DISPLACEMENTS

5 AXIAL FRICTION IN SOFT CLAYS 6.1 Walking response of long pipelines 5.1

Drained and undrained axial response

Drained axial friction is relatively straightforward to establish using devices such as a tilt table (Najjar et al. 2003), or direct shear or ring-shear devices adapted for low stress levels. Undrained axial friction is much more complex to predict, although fast direct shear or ring shear tests can be performed. Undrained axial resistance can lead to a very low lower-bound friction value that dominates the pipewalking response. Recent large-scale axial tests (Bruton et al., 2009) showed that undrained (fast) and drained (slow) axial movements lead to dramatic differences in the axial resistance, due to the generation of excess pore pressure. Although an effective stress friction approach provides a consistent interpretation of both drained and undrained responses, it is necessary to assess and predict the development of excess pore pressure in order to calculate the axial resistance in undrained conditions. The excess pore pressure is thought to be a function of: 1. Pipe velocity – where the range of pipe velocities can span the full undrained to drained response; 2. Pipe weight – where heavier pipes are associated with higher levels of excess pore pressure; 3. Cumulative displacement – where the generation of excess pore pressure can reduce with displacement. The important transition from undrained to drained conditions is not well understood. Improving understanding is a key aim of current research, including the SAFEBUCK GEO JIP. 5.2

Influence of internal flow regime on axial resistance

Another unexpected interaction mechanism that has recently emerged, from observations of operating


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Lateral buckles along a pipeline can significantly influence both pipeline walking and end expansion predictions. When a pipeline undergoes lateral buckling it effectively splits the pipeline into a number of shorter lines between buckles. The pipeline expansion feeds into and out of each lateral buckle (as well as the pipeline ends) with each shutdown and restart cycle (Carr et al., 2006). As a result of differential expansion the size and shape of the buckles change. This modifies the effective axial force profiles and hence the walking response. As pipe feeds into and out of each buckle, the resistance to lateral displacement of the pipe gradually increases as soil berms build up to each side of the buckle, which can cause gradual axial displacement over long lengths of the pipeline. Meanwhile, the virtual anchors that form between each buckle shift between load and unload conditions causing additional localised walking of the pipeline that can significantly influence cumulative pipeline end expansion. Understanding of this complex cyclic interaction is improving and methods for predicting the cyclic response analytically or using FEA techniques are available (Bruton et al., 2010). However, the walking response is extremely sensitive to the axial pipe-soil resistance, Figure 3. The figure shows the typical end expansion response of a long pipeline with multiple buckles. In this example, the walking behaviour is driven by slope (making the displacement more positive), thermal transients (making the displacement more negative) and a changing buckle force profile. The figure shows the load–unload end expansion over a number of shutdown/start-up cycles; the position on load is highlighted by the thick lines. For low axial resistance (µA = 0.4), the initial pipeline expansion is approximately 0.65 m, which falls to 0.2 m on unload. Over the first 20 shutdown

gas accumulating at the top of the slope. This density variation can increase the rate of pipeline walking and significantly modify the shutdown response of lateral buckles (Bruton et al., 2010).

7

SAFEBUCK GEO AND PHASE III

The issues raised in this paper are the subject of ongoing research both for projects and for the SAFEBUCK Joint Industry Project. The current phase of SAFEBUCK GEO has two main aims:

Figure 3. Impact of axial friction on walking.

cycles, the end expansion increases rapidly. Thereafter the rate of increase slows, but there is a steady increase in end expansion on each cycle; the end expansion reaches 1.9 m after 100 cycles. Depending upon the number of cycles anticipated during the field life, anchors may be required. If the axial friction is increased (to µA = 0.8), there is a fundamental change in behaviour. The initial expansion is reduced modestly, to 0.5 m, which falls to 0.15 m on unload.The end expansion increases slightly over the next 40 shutdown cycles, to reach a maximum expansion of 1.1 m, and then begins to reduce as the thermal transients begin to dominate the response. After 100 cycles the end expansion on load is reduced to 1 m and falling. 6.2

Restraining walking – operational feedback

For long pipelines, predicting this cyclic response is extremely challenging, making the design requirements to mitigate walking extremely uncertain. This is why some project have installed anchors on some pipelines and only made provision for anchoring on others (Jayson et al., 2008). Several recent deepwater projects with and without pipeline anchors are now in operation and feedback on the walking response has gradually emerged, though Integrity Monitoring activities that are essential for pipelines that experience cyclic lateral and axial displacements (Baker et al., 2006). This includes the observation of an unexpectedly rapid pipeline walking response, which led to the identification of a multiphase flow mechanism as the main cause. During shutdowns in multiphase pipelines, the flow stream can quickly separate into gas and liquid; which can result in significant variations in submerged weight along the pipeline. This is a major influence on both lateral and axial friction response. In pipelines on steep slopes, liquids settle at the bottom of the slope with © 2011 by Taylor & Francis Group, LLC

1. To fundamentally improve the way that lateral pipesoil response is addressed in design by developing a new ‘force-resultant plasticity model’ to run inside standard software packages, which will capture experience from modelling and testing of lateral pipe-soil interaction. 2. To improve our understanding and quantify key uncertainties in predicting the axial friction response, by a review of all recent project-specific tests, supplemented by additional JIP tests. The current phase of SAFEBUCK Phase III has a number of aims, one of which is to collect and share data and lessons learned from operating pipelines. This will include the pipe-soil response observed on operating pipelines, including embedment levels, axial displacements and the detailed responses at lateral buckles. The other aim is to formalize the SAFEBUCK Design Guidance (Carr, Bruton & Cosham, 2008), including the approach to pipe-soil response, within a recognized DNV Industry Recommended Practice.

8

NOMENCLATURE

V = vertical pipe-soil force Su = soil undrained shear strength D = outside diameter of pipe (including coatings) µA = axial friction factor

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REFERENCES Baker, J.H.A., Bruton, D.A.S. & Matheson, I.C. 2006. Monitoring and Effective Integrity Management of Laterally Buckled Flowlines in Deep Water. Offshore Technology Conference. OTC-17932. Bruton, D., White, D., Bolton, M., Cheuk C. & Carr, M. 2006. Pipe/Soil Interaction Behavior during Lateral Buckling. SPE Projects, Facilities & Construction SPEPFC106847. Bruton, D.A.S., Carr, M. & White, D. 2007. The Influence of Pipe-Soil Interaction on Lateral Buckling and Walking of Pipelines – The SAFEBUCK JIP. 6th International Offshore Site Investigation and Geotechnics Conference. Bruton, D.A.S., White, D., Carr, M. & Cheuk, C. 2008. PipeSoil Interaction During Lateral Buckling and Pipeline Walking – The SAFEBUCK JIP. Offshore Technology Conference. OTC-19589. Bruton, D.A.S., White, D., Langford, T. & Hill, A. J. 2009. Techniques for the assessment of pipe-soil interaction

forces for future deepwater developments. Offshore Technology Conference. OTC-20096. Bruton, D.A.S., Sinclair, F. & Carr, M. 2010. Lessons Learned From Observing Walking of Pipelines with Lateral Buckles, Including New Driving Mechanisms and Updated Analysis Models. Offshore Technology Conference OTC20750. Carr, M., Sinclair, F. & Bruton, D. 2006. Pipeline Walking – Understanding the Field Layout Challenges, and Analytical Solutions developed for the SAFEBUCK JIP. Offshore Technology Conference. OTC-17945. Carr, M., Bruton, D. & Cosham, A. 2008. Design Guideline. SAFEBUCK JIP. (Confidential to JIP Participants) Jayson, D., Delaporte, P., Albert, JP., Provost, M.E., Bruton, D. & Sinclair, F. 2008. Greater Plutonio Project – Subsea Flowline Design and Performance. Offshore Pipeline Technology Conference.


Hill, A.J. & Jacob, H. 2008 In-Situ Measurement of PipeSoil Interaction in Deep Water. Proc. Offshore Technology Conference, Houston, USA. Paper OTC-19528. Najjar, S.N., Gilbert, R.B., Liedtke, E.A. & McCarron, W. 2003. Tilt Table Test for Interface Shear Resistance between Flowlines and Soils. International Conference on Ocean, Offshore and Arctic Engineering. OMAE-37499. Sinclair, F., Carr, M. Bruton, D. & Farrant, T. 2009. Design Challenges and Experience With Controlled Lateral Buckle Initiation Methods. International Conference on Ocean, Offshore and Arctic Engineering. OMAE-79434. White, D.J. & Cheuk, C.Y. 2010. SAFEBUCK Joint Industry Project – Pipe-Soil Interaction Models for Lateral Buckling Design – Phase IIa Data Review. Univ. of W. Australia Report GEO: 09497v2. (Confidential to JIP Participants)

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Large deformation finite element analysis of vertical penetration of pipelines in seabed S. Chatterjee, M.F. Randolph, D.J. White & D. Wang Centre for Offshore Foundations System, The University of Western Australia

ABSTRACT: A large deformation finite element analysis method has been developed based on frequent remeshing and interpolation, linked to small strain analyses performed using the finite element package, ABAQUS. Initially, in order to validate the finite element approach, a set of simulations were undertaken based on the parameters obtained from a set of centrifuge model tests. Numerical simulation results matched well with the centrifuge results, provided both the effects of strain rate and softening were taken into account. The results of the subsequent parametric study showed that the vertical capacity depends strongly on the shear strength profile of the seabed, but is best correlated to the nominal shear strength at the depth of the pipeline invert. A simple approach to allowing for the rate of loading (in respect of strain rate dependency of the shear strength) and partial softening of the soil is described. It is concluded that predicting vertical embedment of pipelines on the seabed can be erroneous if secondary effects are not taken into consideration, including strain rate and softening, local soil heave and buoyancy forces. 1

INTRODUCTION

As offshore energy extraction facilities are gradually shifting to deeper water, greater lengths of pipeline are required, both within the fields and for transporting the hydrocarbon products to land. For the most part, flowlines and export pipelines are just laid on the seabed and become partially embedded due to their self weight and the lay process. Pipes carrying oil at high temperature and pressure tend to deform axially. This is resisted by frictional forces from the seabed soil, resulting in high compressive forces in pipe, which in turn tend to buckle the pipe. Buckling is helpful in reducing excessive compressive force in the pipe but leads to high bending stresses in the pipe wall. For this reason, controlled buckling is a common design practice (Bruton et al., 2006). The lateral buckling response depends on the pipeline embedment and also on the lateral resistance provided by the seabed soil. To achieve controlled buckling, it is essential to estimate the initial vertical pipe embedment and pipe soil interaction forces. Vertical penetration of pipelines in the seabed has been widely researched. Solutions based on classical plasticity theory (Randolph & Houlsby, 1984 Murff et al., 1989, Randolph & White, 2008b) are available in the literature. Many studies (Aubeny et al., 2005; Merifield et al., 2008; Merifield et al., 2009) have been conducted with the help of small strain finite element analysis. Experimental data from centrifuge model tests are also available (Dingle et al., 2008). When a pipe is embedded into the seabed, heave occurs at the sides of the pipe. This phenomenon has been neglected in most of the theoretical studies

mentioned above. Also, dynamic pipe lay effects result in significant amount of softening in the surrounding soil (Randolph & White, 2008a). Effects of strain rate and strain softening on the vertical resistance of pipeline have mostly been ignored in previous studies. For this study, a large deformation finite element approach based on the commercial finite element software ABAQUS (Dassault Systèmes, 2007) was developed, following a similar approach to that described by Wang et al. (2010). The ‘remeshing and interpolation technique with small strain’ (RITSS: Hu & Randolph 1998) method was used, with the original shear strength of the soil modified to account for any increase due to high strain rates and any decrease due to softening (through the accumulation of plastic strain). After the results had been validated by comparing with existing centrifuge test data, a parametric study was undertaken over a wide range of soil properties.

2 2.1

Methodology

A large deformation finite element (LDFE) method based on the RITSS approach was developed. The basis of LDFE analysis using this approach is to divide the large displacements into a series of small strain analyses. Python, the in-built scripting language of ABAQUS, was used for writing pre-processing and post-processing codes. First of all, a python script was written to generate the input file of the first step. This was submitted to ABAQUS for small strain analysis


LARGE DEFORMATION FINITE ELEMENT METHOD

2.3

A total stress approach, assuming an undrained soil response, was adopted for the analyses. A MohrCoulomb plasticity soil model was used with zero friction angle, which is equivalent to the Tresca failure criterion. The pre-failure soil response is assumed to be elastic, with Young’s Modulus E taken equal to 500su , where su is the undrained shear strength of the soil. Poisson’s ratio ν was chosen as 0.499 to minimise volumetric strain. This simple elasto-plastic soil model was modified to account for effects of strain rate and strain softening. The limiting shear strength in the pipe-soil interface, τmax = α sum , with α = 0.5 was assumed, where sum is the shear strength of soil at the mudline.

Figure 1. Typical initial mesh and boundary conditions.

and the output database was post-processed with the help of another python script. Nodal displacements, stresses and other parameters on integration points were recorded.The recorded nodal displacements were used to establish the new displaced boundary of the system. The whole model was then remeshed with the help of a Python script. A Fortran code was written to recover stresses and strains from the old integration points to the old nodes. The superconvergent patch recovery (SPR: Zienkiewicz & Zhu 1993) method was used for stress or strain recovery. These parameters were then interpolated from the old nodes to the new integration points using a subroutine written in Fortran. Interpolated total and plastic strains were used to modify the original shear strength to account for the effects of strain rate and strain softening on shear strength. The detailed methodology regarding this is described later in the paper. Stresses at new integration points are given as initial conditions of the remeshed structure, using the in-built ABAQUS subroutine SIGINI. With these initial conditions, the remeshed model is then submitted for another small strain analysis. The cycle is repeated until the required total displacement is achieved. The whole process is controlled by a main Fortran program, which calls other subroutines or submits Python scripts as required, with no need for user intervention during the analysis.

2.2

Model and boundary conditions

A 2-D plane strain model was used with the pipe considered as rigid and the soil as deformable. Vertical boundaries of the soil were free to move vertically but restrained horizontally. The bottom horizontal boundary was fully fixed, while the top surface was free to move. Plane strain element CPE6 with 3 vertex nodes and 3 mid-side nodes was used for soil elements. Fine meshing, with a minimum size of D/20 (where D is the pipe diameter), was used near the pipe (see Figure 1). At each step, a small displacement (1% of the diameter of the pipe) was applied to the centre of the pipe.

2.4

Effects of strain-rate and softening

Effects of strain rate and strain softening on the shear strength of soil are well accepted. The shear strength of soil is increased by increasing strain rate and decreased as it is remoulded under accumulating shear strains, depending on the soil sensitivity and ductility. The combined effects of strain-rate and softening can be incorporated by extending the simple Tresca soil model, multiplying the original shear strength by two factors (Einav & Randolph 2005, Zhou & Randolph 2007). Here, the undrained shear strength at integration points was modified prior to any step in the analysis according to

Where the first part of the equation adjusts the strength according to the strain rate effect and the second part allows for strain softening. ε1 and ε3 are the major and minor principal strains, respectively, resulting from a displacement increment, δ/D, where D is the pipe diameter. Vp is the pipe velocity. µ is the rate of strength increase per decade of strain rate, which is usually taken in the range of 0.05–0.2, γ˙ ref is the reference strain rate. Softening is taken as an exponential function of the cumulative absolute plastic shear strain ξ. Here δrem denotes the ratio of fully remoulded strength to the initial strength, hence is the inverse of sensitivity, St . The parameter ξ95 reflects the relative ductility of soil and is the value of ξ at which the soil has undergone 95% remoulding. 2.5

Comparison with centrifuge result

Initially, a set of parameters (shear strength profile of soil, unit weight of soil, pipe diameter) were chosen to match conditions in a centrifuge model test (Dingle et al., 2008). The soil used in the study was lightly over consolidated with strength at mudline, sum = 2.3 kPa, linear shear strength gradient k = 3.6 kPa/m and with


Constitutive model

Figure 2. Comparison with centrifuge data. Table 1.

Case I II III IV V VI VII

Figure 3. Variation of penetration resistance with embedment for different cases.

Shear strength parameters during parametric study. Pipe Dia. m

k kPa/m

sum kPa

κ

0.8

1.00 1.00 4.00 6.25 8.00 10.0 10.0

20.0 10.0 8.00 5.00 1.60 1.00 0.50

0.04 0.08 0.40 1.00 4.00 8.00 16.0

Three values of submerged unit weight of soil (3 kN/m3 , 5 kN/m3 and 7 kN/m3 ) and three values of sensitivity of soil (1, 3 and 6) were chosen and repeated with all the cases presented in Table 1. Strain rate parameters described in first part of equation 1 were kept constant during all the simulations with: µ = 0.1; Vp / Dγ˙ ref = 5000. 4

submerged unit weight, γ = 6.5 kN/m3 . With these parameters the pipe was vertically embedded to a depth of 0.45D. The resulting vertical reaction force V was then normalised by Dsu0 , where D is the diameter of the pipe and su0 is the original shear strength at te invert of the pipe. The normalised vertical capacity V/Dsu0 is plotted against embedment ratio w/D in Figure 2. It can be seen from Figure 2 that if effects of strain rate and softening are not considered the response does not match the centrifuge result well. If both the effects of strain rate and softening are considered, results are matching quite well with the centrifuge test data, where the pipe was penetrated at a normalised rate Vp /D = 0.015 s−1 .

4.1

PARAMETRIC STUDY

3.1 Parameters chosen Shear strength parameters, submerged unit weight and the sensitivity of the soil were varied in a parametric study. The shear strength su at any depth z is expressed by su = sum + kz, where sum is the shear strength at mudline and k is the shear strength gradient. Values of sum and k were varied so that a non-dimensionalised shear strength gradient κ (= kD/sum ) captured a wide range of values. Table 1 shows all the shear strength parameters chosen for the study.

Effects of unit weight variation

To investigate the effects of varying the submerged unit weight, the vertical response is plotted in Figures 4 and 5 for the two extreme values of κ, and for the three submerged unit weight values (γ = 3 kN/m3 , 5 kN/m3 and 7 kN/m3 ). It is clear from these that the effect of the soil weight is much greater in the case of soils with low shear strength at the mudline, and hence high values of γ D/sum , as would typically occur where the value of κ is high. For low γ D/sum , the effect of unit weight is negligible. Note that for the analyses in Figure 5, the value of γ D/su0 (normalising using the soil shear strength at pipe invert) ranges from 0.53 to 1.1, so still much greater than for the analyses in Figure 4.


Effects of shear strength profile variation

The non-dimensionalised vertical reaction force is plotted against embedment ratio for different values of κ in Figure 3. For a particular submerged unit weight and sensitivity of soil, the responses are quite similar. This confirms the general approach of normalising the vertical reaction force by the shear strength at the pipe invert. It can, however, be seen that although the vertical response is more or less similar overall, the normalised penetration resistance can vary by as much as 40% at lower embedment ratios.

4.2 3

RESULTS

Figure 4. Effect of submerged unit weight on penetration resistance (κ = 0.04).

Figure 6. Best fit single power law curve for soil with St = 3 (geotechnical resistance only). Table 2. Comparison of power law coefficients for bearing capacity factor, Nc , from different studies (for 0 < w/D < 0.5).

Aubeny et al. (2005) Merifield et al. (2008) Merifield et al. (2009) Present Study

Figure 5. Effect of submerged unit weight on penetration resistance (κ = 16).

The penetration resistance during embedment has two components: the geotechnical resistance and buoyancy due to of the displaced soil. It can be expressed as

The geotechnical resistance, given by the Nc term, can be fitted by a simple power law expressions with coefficients ‘a’ and ‘b’. The buoyancy is expressed in terms of As , the (nominal) submerged cross-sectional area of the pipe and an adjustment factor, fb , to account for local heave. If Archimedes’ principle were to apply (i.e. for a completely level soil surface), the value of fb should be 1. However, as shown by Merifield et al. (2009), the value of fb is around 1.5 when heave is considered. To evaluate a best-fit power law curve to the geotechnical resistance, it is first necessary to subtract the effect of buoyancy from the total capacity. This © 2011 by Taylor & Francis Group, LLC

a

b

Comments

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0.22

Wished-in-place, no strain rate effects Wished-in-place, no strain rate effects Pushed-in-place, no strain rate effects Pushed-in-place, strain rate effects

brings together the responses in Figures 4 and 5. For soils with low values of κ, if fb = 1.5 is assumed, the response merge into a single trend. For soils with high κ, a more appropriate value of fb = 1.9 was required to bring the curves together. Once the geotechnical resistance was isolated, values at a particular embedment were averaged for different κ. This gave an average vertical response curve for soil with particular strain rate and softening parameters. This curve was fitted to the power law expression as shown in Figure 6. The coefficients ‘a’ and ‘b’ are compared in Table 2 with other published values (all for fully rough pipe-soil interfaces). In this way of representing the penetration resistance, the effect of κ is neglected, other than by taking the invert soil strength, su0 . Though this approach is reasonable for embedment ratio w/D ≥ 0.1, there can be significant discrepancy for w/D < 0.1 (as is also evident from Figure 3). The power law fit can be improved for the portion w/D ≥ 0.1, although becomes less good for w/D < 0.1. For this reason, separate power laws have been fitted for the portions less than or greater than w/D = 0.1 for different κ. For w/D < 0.1, the expression

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Table 3. Nc power law coefficients for w/D ≥ 0.1 for different κ.

κ = 0.04 κ = 0.08 κ = 0.40 κ = 1.00 κ = 4.00 κ = 8.00 κ = 16.0

a

b

6.536 6.517 6.431 6.242 6.015 5.809 5.751

0.170 0.167 0.154 0.139 0.117 0.107 0.114

Figure 8. Effect on penetration resistance of varying softening parameter ξ95 .

Figure 7. Effect on penetration resistance of varying strain rate parameter µ.

was used to get a good match with the FE data. The value of m depends on κ. For low κ, m is 0.55–0.5, whereas it is 0.35–0.25 for κ > 1. The power law fit coefficients for w/D ≥ 0.1 for different κ is presented in Table 3.

Figure 9. Effect on penetration resistance of varying sensitivity.

5

4.3 Effects of strain rate and softening Strain rate and strain softening parameters also have a marked effect on the penetration resistance. The effects of variations in strain rate parameter µ, softening parameters ξ95 and St , as described in Equation 1, were explored. Figure 7 shows the effect of varying µ whilst keeping other parameters constant. It can be seen that the penetration resistance can vary up to as much as 56% due to varying µ over the plausible range (for the adopted dimensionless pipe penetration velocity, Vp /Dγ˙ ref = 5000.). Softening parameters ξ95 and St have smaller effects on penetration resistance. As shown in Figures 8 and 9, the penetration resistance increases as ξ95 is increased, and decreases with increasing sensitivity. The penetration resistance can vary by up to 6% due to varying ξ95 , while it can vary by up to 17% for different values of sensitivity in the range 1–6.



The as-laid embedment of a pipeline is extremely important for many aspects of pipeline design, in particular for the assessment of lateral buckling of deep water pipelines. In this paper, a large deformation finite element method has been developed which addresses some of the short-comings of the previous studies in this field. The LDFE method implemented is a robust approach which captures the changes in geometry, in particular the adjacent soil heave, as the pipe is pushed continuously into the soil. The method can also account for the effects of strain rate and strain softening on the shear strength of soil. The present method was validated by comparing results with data from centrifuge model tests.An excellent match between the result of this study and the centrifuge test data was obtained.

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A detailed parametric study was performed to explore the effects of varying shear strength, buoyancy, strain rate and softening parameters on the pipeline penetration resistance. Improved curve fits were obtained for the penetration resistance for base case strain rate and strain softening parameters. These secondary parameters were also found to have an effect on the predicted vertical response. ACKNOWLEDGEMENTS This work forms part of the activities of the Centre for Offshore Foundation Systems (COFS), established under the ARC Research Centres Program and now supported through Centre of Excellence funding from the State Government of Western Australia. The first writer is supported by an International Postgraduate Research Scholarship and a University Postgraduate Research Award from the University of Western Australia. REFERENCES Aubeny, C. P., Shi, H., & Murff, J. D. 2005. Collapse load for cylinder embedded in trench in cohesive soil. International Journal of Geomechanics, 5(4), 320–325. Bruton D. A. S., White D. J., Cheuk C. Y., Bolton M. D. and Carr M. C. 2006. Pipe-soil interaction behaviour during lateral buckling, including large amplitude cyclic displacement tests by the Safebuck JIP. Proc. Offshore Technology Conf., Houston, Paper OTC 17944. Dingle, H. R. C., White, D. J. & Gaudin, C. 2008. Mechanisms of pipe embedment and lateral breakout on soft clay. Canadian Geotechnical Journal, 45, 636–652. Einav, I., & Randolph, M. F. 2005. Combining upper bound and strain path methods for evaluating penetration resistance. Int. J. Numer. Meth. Engng, 63, 1991–2016.

Dassault Systèmes 2007. Abaqus analysis users’ manual, Simula Corp, Providence, RI, USA. Hu, Y., & Randolph, M. F. 1998. A practical numerical approach for large deformation problems in soil. Int. J. for Num. & Analytical Meth. in Geomechanics, 22, 327–350. Martin, C. M. & Randolph, M. F. 2006. Upper-bound analysis of lateral pile capacity in cohesive soil. Geotechnique, 56(2), 141–145. Merifield, R. S., White, D. J. & Randolph, M. F. 2008. The ultimate undrained resistance of partially embedded pipelines. Géotechnique, 58(6), 461–470. Merifield R. S., White, D. J. & Randolph, M. F. 2009. Effect of surface heave on response of partially embedded pipelines on clay. ASCE Journal of Geotechnical and Geoenvironmental Engineering, 135(6), 819–829. Murff, J.D., Wagner, D.A., & Randolph, M.F. 1989. Pipe penetration in cohesive soil. Géotechnique, 39(2), 213–229. Randolph, M. F. & Houlsby, G. T. 1984. The limiting pressure on a circular pile loaded laterally in cohesive soil. Geotechnique, 34(4), 613–623. Randolph, M. F. & White, D. J. 2008a. Pipeline embedment in deep water: process and quantitative assessment. Proc. Offshore Technology Conference, OTC19128. Randolph, M. F. & White, D. J. 2008b. Upper-bound yield envelopes for pipelines at shallow embedment in clay. Géotechnique, 58(4), 297–301. Wang, D, White, D.J. & Randolph, M.F. 2010. Large deformation finite element analysis of pipe penetration and largeamplitude lateral displacement. Canadian Geotechnical Journal (accepted April 2009, in press for publication). Zhou, H. & Randolph, M. F. 2007. Computational techniques and shear band development for cylindrical and spherical penetrometers in strain-softening clay. International Journal of Geomechanics, 7(4), 287–295. Zienkiewicz, O. C. & Zhu, J. Z. 1993. The superconvergent patch recovery and a posterior error estimates. Part 1: The recovery technique. International Journal for Numerical Methods in Engineering, 33, 1331–1364.



Implementation of geotechnical techniques in the analysis of pipeline response G. Cumming & N. Brown J P Kenny Pty, Perth, Australia

ABSTRACT: Several design approaches can be used to analyse the interface between offshore pipelines and the seabed. This paper summarises some key areas of pipeline design that require the implementation of geotechnical techniques and the state of the art research and testing of the behaviour of that pipeline/foundation interface. The integration of the response of the pipeline to the behaviour of the foundation is a critical interface between pipeline engineers and geotechnical engineers, which requires open and productive dialogue to solve often complex interaction scenarios.

1

INTRODUCTION

The design of submarine pipeline systems to DnV OS-F101 requires that the seabed characteristics, such as uneven, unstable, subsidence, seismic activity and soil properties are factors that shall be taken into consideration for the selection of pipeline location. The seabed properties necessary for evaluating the effects of relevant loading conditions shall be determined from testing of the seabed deposits. The modelling of pipe-soil interaction in pipeline design is discussed in more detail in the recommended practices DnV RP F105, DnV RP-F109, and DnV RP-F110, for pipeline spanning, stability and global buckling respectively, where the seabed material is classified as clay or sand/gravel/rock. Pipeline engineering practice uses conservative upper or lower bounding values to characterize the pipeline response in deterministic analysis; low interface friction to conservatively predict maximum expansion or pipelay tensioner clamping requirements; high interface friction to maximize axial or bending stresses. It is not always clearly defined whether an upper or lower bound presents the most onerous case for pipeline design. This is particularly the case in global buckling design, where high axial friction increases the effective axial force in a pipeline, increasing the potential for buckling to occur, but decreases the feedin (or expansion) into buckle locations. Conversely, low axial friction decreases the effective axial force in a pipeline, decreasing the potential for buckling to occur, but increases the feed-in (or expansion) into buckle locations, increasing the consequences of buckling. The design combination of feed-in due to low friction, and initiation due to high friction, may provide a design solution that is neither economic nor practicable.

Equally the most conservative design assumptions for pipeline stabilization or pipeline embedment may similarly lead to design solutions that are neither economic nor practicable. This leads to an increasing requirement for pipeline engineers to understand more about the behaviour of the pipe-soil interface, to assess the risk inherent in a pipeline design solution. The requirement for pipeline engineers to more fully understand the pipe-soil interface is equally required where the seabed material classification does not fall into the clay or sand/gravel/rock categories.

2 2.1

General

In the design of pipeline supports, be they engineered or seabed, the initial assumption in pipeline design is that the pipeline moves, often a significant distance, and that the loads imparted by the pipeline on the support may significantly exceed the resistance that the seabed may provide. This is contrasted by traditional foundation design, where the design criteria consider that the foundation resists significant movement of the structure. This can lead to misunderstanding between pipeline and geotechnical engineers, where they are essentially talking different languages, but using the same words. Consider the example of a concrete mattress overlaid at the end of an offshore pipeline for protection. The design of the pipeline-mattress-seabed interface will see loads much greater than the pipeline-mattressseabed can resist. However, the mattress-seabed interface will have the functional requirement to be stable, under the action of the pipeline and hydrodynamic loads.


PIPELINE RESPONSE

The ‘length’ of a pipe support, particularly an engineered support, may be short, even with regard to the pipeline diameter. For a pipeline supported on a seabed, the length of pipeline that is supported may vary from continuous to discrete short supports. The response of the pipeline may equally vary between long system effects or short local effects. Typically axial expansion (feed-in) / restraint is associated with a long length of pipeline. The support of pipelines at span shoulders and at the crowns of global buckles is typically a local restraint. In stability analyses, restraints such as secondary stabilization measures are generally a local response, where primary stabilization away from any local restraints is pipeline system behaviour. It is often required to consider the interaction of system and local behaviour, in particular for global buckling design.

The comparison of foundation and pipeline design for the example of lateral buckling is further discussed in White and Gaudin (2008), DNV RP-F110 (2007) and Bruton et al. (2008). For the purpose of this paper the implementation of geotechnical techniques into the consideration of pipeline response is split into three approaches: – a traditional friction factor approach, – an empirical approach, – the fundamental approach. In many ways these approaches all originate from an experimental/testing basis whether that is from in-situ testing, laboratory tests, centrifuge testing or full scale experiments. The difference is in how the idealised soil response experimental data is interfaced in to the (often idealised) pipeline response models. In the traditional friction factor approach, the results of the geotechnical testing and/or analysis are interpreted into a Coulomb, bi-linear (or multi-linear) friction factor response. In the empirical approach the geotechnical results may be interpreted to account for some of the more non-linear aspects of pipe-soil behaviour which may include passive resistance, cyclic, and load history effects. In the fundamental approach the geotechnical analysis interprets the test results to address the underlying mechanics of the pipe-soil behaviour. 2.2

Pipeline response – system or local

Offshore pipelines are long with regard to their diameter, but also with regard to the wavelength of ‘global’ response modes.


Implementation of traditional friction factor approach

In many ways a section on the traditional friction factor approach in the proceedings of the symposium on frontiers in offshore geotechnics could seem misplaced. Many authors have highlighted the influence of loading history, loading rate and hydrodynamic interactions, on the pipe-soil interaction behaviour, and that the pipe-soil resistance is often not directly proportional to the vertical pipeline reaction. There are many reasons for the traditional friction factor approach remaining a core tool in the pipeline design ‘toolbox’:

Pipeline surveys

While pipeline route surveys are normally carried out along the total length of a planned pipeline route, it is worth noting that a pipeline route survey is not necessarily a geotechnical pipeline route survey. Guidance on geotechnical investigations for marine pipelines is further discussed in OSIF (2004), http://sig.sut.org.uk/sutosig.htm. Part of the functional requirements for these route surveys should include the assessment of seabed features including geophysical, geotechnical and geological information. The assessment of geotechnical properties typically includes information from geological databases, insitu testing, laboratory testing, and classification testing. Characteristic soil parameters – ‘design’ values – are subsequently defined. Where the design of the pipeline requires the use of model or full scale testing for the calibration or definition of input parameters for specific pipe-soil models, the collection and preservation of a substantial quantity of soil may be required. Where the design of the pipeline system requires characteristic soil parameters for soils that are not classified ‘clay’ or ‘sand’, there is limited guidance in the present range of published offshore pipeline rules and guidelines. 2.3

2.4

– The friction factor approach remains an appropriate assumption for many pipeline responses. – At conceptual/front end phases of pipeline design, it is often required to make major assumptions in lieu of little or no seabed survey information. – It is common practice in design and research to consider idealised cases, where simplifying assumptions are utilized to focus on a particular aspect of the response. – Before an experimental approach may be specified it is often necessary to bound the expected input ranges using idealised cases. – The friction factor approach can be used to quickly implement geotechnical experimental results into standard pipeline engineering tools, leading to an experimentally calibrated friction factor approach. It is often assumed that friction coefficients for pipelines and soils are historically available. However there are frequently differences in pipeline coatings, even within the same ‘type’ of coating. The typical concrete coating application method in SE Asia differs from the application method applied in European coating plants, leading to significant differences in concrete coating roughness, which may have an effect on the pipe-coating/soil interface friction.

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Figure 2. UWAPIPE vertical displacement versus reaction. Figure 1. Typical Verley model force-displacement curve.

Where the interface friction has a significant impact on the pipeline design, laboratory shear tests are recommended for the pipe-coating/interfacing materials. 2.5 Implementation of geotechnical techniques – empirical sand model The Verley & Sotberg (1992) pipe-soil interaction model for sandy soils predicts the development of pipe penetration into the soil and the associated soil resistance that may be mobilized against horizontal pipe displacements (Figure 1). Zeitoun et al. (2009) describe the implementation of the Verley & Sotberg (1992) soil response model into the J P Kenny Simulator package that uses ABAQUS as the finite element engine. The pipe-soil model has been implemented as a FORTRAN subroutine, and could be equally extended to include the Verley and Lund (1995) soil response model for clay soils. An empirical model for partially drained conditions, which often occur in silts or calcareous sands, would require further testing to develop and implement. 2.6 Implementation of geotechnical techniques fundamental approach – plasticity model / UWAPIPE The plasticity models set out in Zhang et al. (2001, 2002a, and 2002b) provide a fundamental approach to simulate the mechanics of pipe-soil behaviour. Zhang and Erbrich (2005) present applications of the plasticity models demonstrating how the ‘traditional friction factor approach’does not capture the effect of the loading path on the minimum friction coefficient. Tian & Cassidy (2008) present the application of the plasticity models in terminology consistent with structural analysis frameworks and finite element models, referring to this model as the UWAPIPE model. The UWAPIPE model, developed by, and licensed from the University of Western Australia, has been implemented into the J P Kenny Simulator package. Examples responses of UWAPIPE, including the associated failure envelopes in vertical-horizontal (V-H) load space are shown in Figure 2 and Figure 3. © 2011 by Taylor & Francis Group, LLC

Figure 3. Pipe positions and failure envelopes.

2.7

Comparison of pipeline dynamic lateral stability response to Coulomb, Verley and Sotberg (1992) and UWAPIPE models

The key issues that are faced when designing for pipeline stability include; the acceptance criteria, the hydrodynamic loads acting on the pipeline, the pipe structural response and the pipe-soil response including resistance, liquefaction and scour as discussed in Zeitoun et al (2008). For these key issues, idealised cases are often utilized to define loads and resistances. The applicability of these idealised cases has previously been a subject of study and debate (e.g. Palmer (1996) and Teh et al. (2006)). The recently published DnV RP-F109 considers that soil resistance consists of a Coulomb component and a passive penetration contribution and that hydrodynamic load reduction may be attributed to pipe embedment and for a permeable seabed. DnV PR-F109 acknowledges the issues with calcareous soils, liquefaction and scour, and that at present a recognized model for incorporating scour and liquefaction into a generalised pipeline stability design methodology remains a subject for ongoing research.

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increased embedment, and therefore lateral resistance, compared to the original Zhang / UWAPIPE model, which is for drained conditions. The development of further fundamental understanding of pipeline-soil-fluid behaviour is ongoing and includes research such as the STABLEpipe JIP, as discussed in Griffiths et al. (2010), and partially drained plasticity models. In addition to the resistance of the soil, these developments consider the effect of the fluid/pipe on the soil properties and the effect of the pipe on the soil resistance. Figure 4. Lateral movement comparison.

2.8

Figure 5. Vertical displacement comparison.

From a pipeline lateral stability design consideration, how do the present methods of modelling pipe-soil resistance compare when considering the response of the pipeline system to a particular hydrodynamic load history and these different resistance models? For the purpose of comparing the pipeline dynamic lateral stability response to Coulomb, Verley and Sotberg (1992) and UWAPIPE models, no load reduction, scour or liquefaction effects are considered. The input parameters for the Verley & Sotberg (1992) and UWAPIPE models are as detailed in Verley & Sotberg (1992) and Tian & Cassidy (2008), and these are compared to case using a Coulomb friction coefficient of 0.8. In Figure 4, the lateral movement for the Coulomb model and the UWAPIPE model are similar, with a significantly lower lateral displacement for the Verley and Sotberg (1992) model. This implies that the input parameters utilized in the UWAPIPE model are comparable to a Coulomb friction of around 0.8, for the pipe size and SG under consideration. The embedment obtained in the UWAPIPE and Verley and Sotberg (1992) models are significant, with a lower embedment predicted from the UWAPIPE model (Figure 5). It is noted that Cathie et al. (2005) suggest that the centrifuge test data used to develop the Zhang / UWAPIPE model gives a lower rate of increase in equivalent friction factor with increasing embedment than the test data used to develop the Verley & Sotberg model. Recent centrifuge studies to assess a partially drained response of a pipeline in calcareous soil have led to a proposed partially drained model which offers


Implementation of geotechnical techniques – modelling of pipelay induced embedment and its effect on lateral resistance

Observations of pipelay-induced embedment have been documented by Lund (2000) and Westgate et al. (2009). The observations show that the embedment can be variable based on the activity of the pipelay vessel, from planned activities such as pipe end laydown, and unplanned events that reduce the lay rate. Bruton et al. (2008) and DnV OS-F110 discuss pipesoil interaction responses that influence the design of pipelines susceptible to lateral buckling, which include pipe embedment due to installation effects, its effect on lateral resistance, and the effect of axial resistance on feed-in. Cheuk and White (2008), and White and Gaudin (2008) present how centrifuge modelling techniques can be utilised to study the behaviour of the soil under dynamic pipelay load histories. This is further developed by Wang et al. (2009) who present numerical simulations of dynamic pipelay embedment, andYu and Konuk (2007) who utilize a continuum approach to simulate lateral buckling. In an operational pipeline, the as-found pipeline embedment may be considered when determining the most likely range of lateral resistances, taking into account any potential for seabed mobility. In the design phase this is less straightforward. A lateral buckle is a local feature, and the feed-in to the buckle is a global pipeline effect. Pipeline design should consider an approach that addresses both of these global and local effects. The combination of lower, upper bound and best estimate frictions for a range of potential pipelay embedments can be overly conservative. A probabilistic approach may be used to describe the potential range of pipeline responses such as the method presented in Rathbone et al. (2008). Soil responses from no pipelay effect to a full pipelay effect may therefore be addressed, capturing the range of potential lateral resistances. Engineering judgment is still required in the definition of the axial and lateral resistance probability distribution functions. For the lateral buckling case these should consider that the axial resistance is a global feature and that the lateral resistance is a local feature and the statistical distributions may be limited by the no embedment case and a maximum predicted embedment case. Engineering judgment is also required to interpret the results of the probabilistic assessment, or to communicate the risk of

Figure 6. Typical pipelay lateral friction distributions.

Figure 8. Superspan soil testing in a geotechnical centrifuge.

Figure 7. Jansz Superspan illustration Equid (2008).

buckling from the design to the operational / integrity management phases of the project. There are many challenges that are the subject of ongoing research in lateral buckling design. These include the consideration of pipe-soil model complexity (force resultant, empirical, plasticity, continuum), the requirements for, and variability of, soil testing, and how to describe appropriate idealised pipeline design cases. Even when engineered interfaces such as rock or vertical/lateral trigger systems are utilized, the small risk of an unplanned lateral buckle on the seabed may govern the pipeline design.

geotechnical and pipeline design team worked closely together to determine the interaction of pipeline loads with the potential geotechnical response at the span shoulders. In order to further understand the behaviour of the soil at the shoulder of the Jansz scarp span, detailed assessments including centrifuge testing and modelling program were undertaken (Figure 8), to determine and validate the pipe-soil interaction model to be used in the analysis of the pipeline response. The pipe-soil interaction model considered the axial and lateral soil behaviour, and in particular the effect of both the pipelay loads and operational loads on the potential embedment and vertical stiffness. The pipe-soil interaction model was interpreted into input parameters for the pipeline structural model, considering the potential for various pipelay and operational loads, and the potential range in soil conditions determined by the in-situ geotechnical tests. The resulting structural model was extensively utilized to assess the sensitivity of the base case design solution to perturbation of the constructed trench/pipe profile/location, pipelay loads and back tensions. 3

2.9 Implementation of geotechnical techniques – modelling of Jansz scarp crossing span support The challenges of the Jansz development and the crossing of the scarp are discussed in Equid (2008). Figure 7 shows an impression of the scarp and resulting pipeline span. DnV OS F105 requires that the modelling of pipe-soil interaction is considered in the detailed evaluation of free spanning pipelines, preferably by means of geotechnical tests on undisturbed soil samples, in particular where the in-situ soils differ from the typical geotechnical values given in DnV RP-F105. Borehole samples of the seabed were collected and utilized for laboratory testing. The over consolidated calcareous soil found at the Jansz scarp has very different characteristics to typical deep water soils. The



Projects have significant benefits to gain through sound management of the risks associated pipeline geotechnical design. Mitigation strategies need to address both technical and design execution risks. Key factors that need to be addressed include: – the availability of geotechnical data, – the availability of pipe-soil models that are applicable for the soil classification, – the availability of pipeline design tools that can utilise the pipe-soil model, – the maturity of the pipeline design. A key design execution risk is management of the interaction between pipeline designers and geotechnical specialists. It must be recognized that the technical

languages used by respective parties are different and poor communication can lead to incomplete and inconsistent overall design outcomes. Recognition and mitigation of this risk is an essential element of the pipe-soil interaction design problem. A well managed design effort including clear recognition and management of technical and communication risks can lead to significant benefits from both design integrity and project execution perspectives. REFERENCES Bruton, D. A. S., White, D., Carr M. and Cheuk, J. C. Y. (2008) Pipe-Soil Interaction During Lateral Buckling and Pipeline Walking – The SAFEBUCK JIP, Proc. Offshore Technology Conference, OTC 19589, Houston, USA. Cathie, D.N. Jaeck, C., Ballard, J.C. and Wintgens J.F.:(2005) Pipeline Geotechnics: State of the Art, Proc. Int. Symp. on Frontiers in Offshore Geotech. (ISFOG) Perth, Australia. Cheuk, C. Y. and White, D. J. (2008). Centrifuge modelling of pipe penetration due to dynamic lay effects. Proc. Int. Conf. on Offshore Mech. & Arctic Engng., OMAE200857923. Estoril, Portugal. Det Norske Veritas. Offshore Standard DNV OS-F101 Submarine Pipeline Systems, October 2007. Det Norske Veritas, Recommended Practice DNV RP-F105 Free Spanning Pipelines, February 2006. Det Norske Veritas, Recommended Prac. DNV RP-F109 On-bottom Stability Design of Submarine Pipelines April 2009. Det Norske Veritas, Recommended Practice DNV RP-F110 Global Buckling of Submarine Pipelines Structural design due to high temperature/pressure., October 2007. Equid, D. (2008) Challenges of the Jansz deepwater tieback. Proc. Deep Off. Tech. Conf. (Asia-Pacific), Perth, Australia. Griffiths, T. J., White, D. J. and Cheng, L., (2010) Progress in investigating pipe-soil-fluid interaction: The STABLEpipe JIP. Proc. 20th Int. Offshore and Polar Engng. Conf., ISOPE 2010-TPC-790, Beijing, China. Lund, K. M. (2000) “Effect of increase in Pipeline Soil Penetration From Installation” Proc. Int. Conf. on Offshore Mech. & Arctic Engng. OMAE2000/PIPE-5047. Offshore Soil Investigation Forum (OSIF) 2004. Guidance notes on geotechnical investigation of marine pipelines. Rev 03. http://sig.sut.org.uk/sutosig.htm. Palmer, A.C., (1996) A Flaw in the Conventional Approach to Stability Design of Marine Pipelines. Proc. Conf. on Offshore Pipeline Technology, Amsterdam. Rathbone, A. Hakim, M. A. Cumming, G. and Tørnes, K., (2008) Reliability of lateral buckling formation from planned and unplanned buckle sites. Proc. Int. Conf. on Offshore Mech. & Arctic Engng. OMAE2008-57300, Estoril, Portugal.


Teh, T.C., Palmer, A.C., Bolton, M.D. and Damgaard, J.S., (2006) Stability of submarine pipelines on liquefied seabed. ASCE J. of Waterway, Port and Coastal Engineering. Tian,Y. and Cassidy, M. (2008). Explicit and implicit integration algorithms for an elastoplastic pipe-soil interaction macroelement model. Proc. Int. Conf. Offshore Mech. & Arctic Engng. OMAE2008-57237, Estoril, Portugal. Verley, R.L.P. and Sotberg, T. (1992) A soil resistance model for pipelines placed on sandy soil. Proc. Int. Conf. Offshore Mech. & Arctic Engng. Alberta, Canada. Verley, R.L.P. and Lund, K. M. (1995) A soil resistance model for pipelines placed on clay soils. Proc. Int. Conf. Offshore Mech. & Arctic Engng Copenhagen, Denmark. Wang, D., White, D.J. and Randolph, M.F. (2009), Numerical simulations of dynamic embedment during pipe laying on soft clay. Proc. 28th Int. Conf. on Offshore Mechanics and Arctic Engineering, OMAE2009-79199, Honolulu, Hawaii. Westgate, Z. J., White, D. J. and Randolph, M. F. (2009), Video observations of dynamic embedment during pipelaying in soft clay”, Proc. 28th Int. Conf. on Offshore Mechanics and Arctic Eng., OMAE2009–79814, Honolulu, Hawaii. White D.J. and Gaudin, C. (2008) Simulation of seabed pipesoil interaction using geotechnical centrifuge modelling. Proc. Deep Offsh. Tech. Conf. (Asia-Pacific) Perth, Aust. Yu, S. and Konuk, I. (2007) Continuum FE modelling of lateral buckling Proc. Offshore Tech. Conf., Houston USA. Paper OTC18934. Zhang, J., Stewart, D. P. and Randolph, M. F. (2001). Centrifuge modelling of drained behaviour for pipelines shallowly embedded in calcareous sand. Int. J. Physical Modelling in Geotechnics 1, 25–39. Zhang, J., Stewart, D. P. and Randolph, M. F. (2002a). Kinematic hardening model for pipeline-soil interaction under various loading conditions Int. J. Geomech., 2(4), 419–446. Zhang, J., Stewart, D. P. and Randolph, M. F. (2002b). Modelling of shallowly embedded offshore pipelines in calcareous sand. ASCE J. Geotech. & Geoenv. Eng., 128(5), 363–371. Zhang J., and Erbrich C.T.: (2005) Stability design of untrenched pipelines – geotechnical aspects, Proc. Int. Symp. on Frontiers in Offsh. Geotech. (ISFOG) Perth, Aust. Zeitoun, H.O., Tørnes, K., Cumming, G., and Brankoviæ, M. (2008) Pipeline stability – State of the Art. Proc. 27th Int. Conf. on Offshore Mech. & Arctic Eng., OMAE2008– 57284, Estoril, Portugal. Zeitoun, H.O., Tørnes, K., Li, J., Wong, S., Brevet, R., and Willcocks, J. (2009). Advanced dynamic stability analysis, Proc. 28th Int. Conf. on Offshore Mechanics and Arctic Eng., OMAE2009–79778, Honolulu, Hawaii.

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Lateral soil resistance to an untrenched pipeline under the action of ocean currents F.P. Gao Key Laboratory for Hydrodynamics and Ocean Engineering, Institute of Mechanics, Chinese Academy of Sciences, Beijing, China

S.M. Yan Key Laboratory for Hydrodynamics and Ocean Engineering, Institute of Mechanics, Chinese Academy of Sciences, Beijing, China China Petroleum Pipeline Engineering Corporation, Langfang, China

E.Y. Zhang CNOOC Research Center, Beijing, China

Y.X. Wu Key Laboratory for Hydrodynamics and Ocean Engineering, Institute of Mechanics, Chinese Academy of Sciences, Beijing, China

X. Jia CNOOC Research Center, Beijing, China

ABSTRACT: A plane-strain finite element model is proposed to investigate the pipe-soil interaction mechanisms for the partially embedded pipe with two kinds of constraint conditions, i.e. freely-laid pipe and anti-rolling pipe. The numerical model is verified with updated mechanical-actuator experiments. The magnitude of lateralsoil-resistance coefficient for the examined anti-rolling pipes is much lager than that for the freely-laid pipes, indicating the end-constraint condition affects significantly the lateral stability of the untrenched pipeline in ocean currents. Parametric study indicates that the variation of lateral-soil-resistance coefficient with the dimensionless submerged weight of pipe is affected greatly by the internal friction angle of soil, pipe-soil friction coefficient, etc. 1

INTRODUCTION

To avoid the occurrence of pipeline on-bottom (lateral) instability, i.e. the breakout of the pipe from its original site, the seabed must provide enough soil resistance to balance the hydrodynamic loads upon the untrenched pipeline. For pipeline geotechnical engineers, one of the main concerns for pipeline on-bottom stability design is to properly predict the ultimate soil resistance in the severe ocean environments, and to further determine the thickness of coating layers based on nominal pipe weight (Det Norske Veritas 2007). In the past few decades, the pipe-soil interactions have attracted much interest from pipeline researchers and designers. Numerous experimental studies on wave-induced pipe instability have been carried out with 1g mechanical actuators (e.g., Wagner et al. 1987; Palmer et al. 1988), with centrifugal pipe-soil interaction tests on calcareous sand (e.g. Zhang et al. 2002), and with flume hydrodynamic simulations (e.g. Gao et al. 2003; Teh et al. 2003). Several empirical


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“pipe-soil” or “wave-pipe-soil” interaction models were developed to improve the conventional Coulomb friction theory. Some reviews on pipeline geotechnics and pipe-soil interactions have been made recently by Cathie et al. (2005), White and Randolph (2007), etc. Note that the aforementioned studies mainly focused on the pipeline on-bottom stability subjected to ocean waves. As the oil and gas exploitation moving into deeper waters, ocean current becomes the prevailing hydrodynamic load for on-bottom stability of submarine pipelines.Although the pipe on-bottom stability in currents seems less complicated than in waves, till now, the underlying physical mechanism has not been well revealed (Gao et al. 2007). To further explore the mechanism of pipeline onbottom stability in ocean currents, a plane-strain finite element model is proposed and verified with the mechanical-actuator tests. The ultimate lateral soil resistance to the untrenched pipes with two kinds of constraint conditions, i.e. freely-laid pipes and anti-rolling pipes, is investigated numerically.

Figure 1. Typical plane-strain finite element mesh (not in scale) and boundary conditions for pipe lateral stability analyses.

2

DEVELOPMENT OF A PLANE-STRAIN PIPE-SOIL INTERACTION MODEL

2.1 The finite element model 2.1.1 Finite element mesh and boundary conditions As the length of a submarine pipeline is much larger than its diameter, the pipeline lateral stability can be treated as a plane-strain problem. A plane-strain finite element model is proposed for simulating the breakout of the pipeline from its original site. The typical finite element mesh is illustrated in Figure 1. The boundary conditions are set as follows: (1) at the left and right boundary, no displacement in the x direction takes place; (2) the bottom boundary is fixed, i.e. the displacement and rotation are not permitted; (3) at the pipe-soil interface, the contact-pair algorithm provided in the ABAQUS software (Hibbitt, Karlsson and Sorensen Inc, 2006) is adopted to simulate the moving pipe along the deformable soil. The non-contact soil surface is treated as a free boundary. In the numerical modelling of the pipeline losing onbottom stability, it is crucial to properly describe the contact conditions between the pipe and the neighbouring soil. The pipe-soil friction is defined by the Penalty Function with the advantage that it guarantees the positive definiteness of sparse matrix in the calculation. In order to avoid large distortion of finite elements causing the calculation misconvergence, the self-adaptive mesh technology is employed. To obtain high calculation efficiency, the finite element mesh gets more refined at closer proximity to the pipe. Based on the results of a series of trial calculations, the width of the numerical model is set as 17.5D and the depth as 5D, and the pipeline is located at x = 7.5D (Dis the pipeline diameter), see Figure 1. 2.1.2 The end-constraint and the simulation of ocean current loading on the pipeline For a long-distance laid pipeline, the on-bottom stability of the pipeline at its separate sections is different. Due to the constraints from risers and the pipeline own anti-torsion rigidity, the pipeline movement is neither purely parallel nor purely rotational. As such, the following two end-constraint conditions are taken account in the present study: Case I: Anti-rolling pipe. © 2011 by Taylor & Francis Group, LLC

Pipeline’s rolling is restricted, but pipeline can move freely in horizontal and vertical directions; Case II: Freely-laid pipe. The pipe may rotate around its axis without any end constraint. When a pipeline is laid on the seabed under the action of ocean currents, there exists a dynamic balance between the submerged weight of the pipe, the hydrodynamics forces (including the horizontal drag force FD and the vertical lift force FL ), and the soil resistances. When the ultimate lateral soil resistance is not able to balance the horizontal drag force, the pipe would breakout from its original site, i.e. the lateral instability occurs. To efficiently simulate the ocean currents induced hydrodynamic loads upon a submarine pipe-line is crucial for evaluating pipeline lateral on-bottom stability. According to Morison’s equation, the horizontal and lift (vertical) components of the steady flow induced horizontal drag force and vertical lift force are expressed as FD = 0.5CD ρw DU 2 , FL = 0.5CL ρw DU 2 , respectively. Herein, CD is the drag coefficient, CL is the lift coefficient, is the mass density of water, U is the effective water particle velocity. The variations of the drag and lift coefficients, CD and CL , with the Reynolds number (Re) for various values of pipe surface roughness have been obtained by Jones (1978). The resultant hydrodynamics force upon the pipe is obliquely upwards with the inclination angle:

Referring to the experimental results by Jones (1978), the inclination angle (θ) is approximately between 530 − 570 . It is therefore reasonable to apply an inclined force in the θ direction to simulate the hydrodynamic loads on the pipe in steady ocean currents. Constitutive model for soils and the material properties The sandy soil under drained conditions can be essentially assumed to behave as an elastic c-φ material (e.g. Mohr-Coulomb or D-P material). The seabed soil is simulated with the well-known Drucker-Prager (D-P) elastoplasticity constitutive model. In the simulations, the parameters of soil are chosen as follows: Young’s modulus E = 0.18 MPa, Poisson’s ratio ν = 0.32, the cohesion c = 0, the buoyant unit weight of the soil γ = 9.3 × 103 N/m3 , the values of soil internal friction angle φ are various for the parametric study in Section 3.1. As aforementioned, the pipe is treated as a rigid cylinder with outer diameter D = 0.15 m (same as the test pipes). The submerge weight of the pipe per meter (WS ) and the pipe-soil friction coefficient (µ) are various for parametric studies in Section 3. Due to that the stiffness of the steel pipeline with concrete cover is normally larger than that of the soil, the wall of the pipeline is regarded as a rigid cylinder in this finite element analysis.

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2.1.3

Figure 2. Experimental setup for pipe lateral stability.

2.2 Verification of numerical model To verify the proposed numerical model, an updated experimental facility by employing the mechanicalactuator simulation method has recently been designed and constructed, as depicted in Fig. 2. The facility mainly consists of a sand box with glass wall, a mechanical-actuator, and the measurement system, etc.. In the sand box (2 m long, 0.5 m wide and 0.6 m deep), a saturated sand-bed with certain relative density can be prepared by employing the sand-raining technique. In the mechanical actuator system, a stepper motor was capable of generating inclined force onto the test pipe via a cable passing through a fixed pulley, for simulating steady currents induced drag force and lift force on the pipeline. Meanwhile, a lifter was used to adjust the inclination angle, which was maintained in the range of 530 ∼ 570 according to the above analyses. Figure 3(a) illustrates typical development of lateral soil resistance and the corresponding vertical pipesoil contact force for an anti-rolling pipe when losing lateral stability. With the increase of horizontal displacement (Sx ) during the pipe losing lateral stability, the horizontal lateral soil resistance (FH ) increases gradually to its maximum value (Fu = 0.10 kN/m) when the additional settlement is nearly fully developed according to the experimental observation. Meanwhile, the corresponding vertical pipe-soil contact force (WS − FH tanθ) decreases gradually to its minimum value (0.085 kN/m). The FEM numerical results match well with the test results. Figure 3(b) shows the numerical results of plastic deformation beneath the anti-rolling pipe while losing lateral stability. It is indicated that the shear band is distributed underneath the deformed soil layer; meanwhile, the soil just in front of the moving pipe upheaves obviously (see Figure 3b). The variation of ultimate lateral soil resistance (Fu ) with the vertical pipe-soil contact force (WS − Fu tanθ) is given in Figure 4, indicating the numerical and the experimental results are quite comparable. The ultimate lateral soil resistance increases linearly with the vertical pipe-soil contact force. The proposed FEM model is capable of predicting the lateral resistance for the untrenched pipeline on-bottom instability. © 2011 by Taylor & Francis Group, LLC

Figure 3. (a) Development of lateral soil resistance and the corresponding vertical pipe-soil contact force for an anti-rolling pipe when losing lateral stability: Comparison between numerical and experimental results; (b) Plastic deformation beneath the anti-rolling pipe while losing lateral stability (D = 0.15 m, µ = 0.7, WS = 0.225 kN/m, φ = 26.7◦ ).

Figure 4. Variation of ultimate lateral soil resistance with vertical pipe-soil contact force: Comparison between the numerical and the experimental results (D = 0.15 m, µ = 0.7, φ = 26.7◦ ).

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3

NUMERICAL RESULTS AND ANALYSES

In the process of a pipeline losing lateral stability under the action of ocean currents, the soil plastic deformation beneath the untrenched pipeline may be

Figure 5. Plastic deformation beneath the pipe while losing lateral stability: (a) Anti-rolling pipe; (b) Freely-laid pipe (D = 0.15 m, Ws = 0.439 kN/m, µ = 0.7, φ = 20◦ ).

created due to the intensive pipe-soil interaction. Figure 5(a) and (b) illustrate the plastic strain in the proximity of an anti-rolling pipeline and that of a freely-laid pipeline, respectively. It is indicated that, the end-constraint condition has much influence on the distribution of the plastic strain zone in the soil. For the anti-rolling pipeline, an obvious shear strain band may be formed in the underlying soil layer, and soil upheave occurs in front of the moving pipeline (see Figure 5(a)). Nevertheless, for the freely-laid pipeline, the smaller plastic-strain zone is created just underneath the rolling pipeline (see Figure 5(b)). As discussed above, many factors influencing the pipe-soil interaction could be incorporated in the proposed finite element model. In the following sections, the effects of soil internal friction angle and the pipesoil friction coefficient on the on-bottom stability of the pipelines with two kinds of end-constraint will be further investigated numerically.

3.1

Figure 6. Lateral stability of anti-rolling pipes for various values of internal friction angle: (a) em /D vs. G; (b) η vs. G(D = 0.15m, µ = 0.7).

horizontal lateral soil resistance to the corresponding vertical pipe-soil contact force, i.e.

The commonly-used dimensionless submerged weight (G) of the pipe is

Effects of soil internal friction angle

For better understanding the pipe-soil interaction mechanism for on-bottom stability, a lateral-soilresistance coefficient (η) is proposed, whose physical meaning is the ratio of the ultimate value of the © 2011 by Taylor & Francis Group, LLC

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where γ is the buoyant unit weight of the saturated sand. Both the experimental and numerical results show that, in addition to the initial embedment due to selfweight of the pipe in the process of losing lateral stability, some additional settlement may be developed while the horizontal lateral soil resistance increases gradually to its maximum value. Figure 6(a) and (b) give the variation of maximum pipe settlement (em /D. with the dimensionless pipe submerged weight (G) and that of the corresponding lateral soil resistance coefficient (η) with

Figure 7. Lateral stability of freely-laid pipes for various values of internal friction angle: (a) em /D vs. G; (b) η vs. G (D = 0.15 m, µ = 0.7).

G for various values of soil internal frictional angles for the Case of anti-rolling pipeline with a given diameter (D = 0.15 m). The maximum pipeline settlements (em /D in the process of pipeline losing stability increase approximately linearly with the increase of G. For the same value of G, em /D increases with the decrease of soil internal friction angle, especially for the larger pipeline submerged weights (see Figure 6(a)). The lateral-soil-resistance coefficient (η) decreases gradually to a constant value with the increase of G. The effect of soil internal friction angle on η gets more significant with increasing submerged weight of the pipeline. Similarly, the variation of em /D with G and that of η with G for the case of freely-laid pipelines are given in Figure 7(a) and (b). Compared with the case of anti-rolling pipelines (see Figure 6), the relationships between em /D and G for the freely-laid pipes follow similar trends, but the maximum settlements are somewhat less in magnitude. Unlike the case of antirolling pipe, the effect of φ on the variation of η with G for the freely-laid pipeline is different, i.e. η decreases with the increase of φ for a fixed value of G (e.g.

Figure 8. Effects of pipe-soil friction coefficients on the lateral stability of anti-rolling pipes: (a) em /D vs. G; (b) η vs. G (D = 0.15 m, φ = 26.70 ).

G > 1.0, see Figure 7(b)). This may attribute to that the pipe settles shallower into the soil with bigger internal friction angle, and that the freely-laid pipe tends to roll away from its original site. Note that the range of η for the examined anti-rolling pipes is between 1.0 ∼ 2.0 (see Figure 6(b)), but that for the freely-laid pipes only between 0.2∼0.3 (see Figure 7(b)). Therefore, the end-constraints have significant influence on the lateral stability of the untrenched pipeline in ocean currents. 3.2

The submarine pipeline is usually constructed with concrete cover. As imagined, the pipe-soil friction coefficient may affect the lateral stability of the pipeline. Figure 8 (a) and (b) give the variation of em /D with G and that of η with G for various values of pipe-soil friction coefficient (µ), respectively. For the case of anti-rolling pipes, the increase of µ brings an increase of maximum settlement in the process of pipe losing lateral stability (see Figure 8(a)). The effect of


Effects of pipe-soil friction coefficient

pipe-soil friction coefficient is more obvious for the smaller value of G. Its effect on the variation of η with G gets less with the decrease of µ (see Figure 8(b)). 4

CONCLUDING REMARKS

As the offshore oil & gas exploitation goes into deeper waters, ocean current becomes the prevailing hydrodynamic load for on-bottom stability of submarine pipelines. In this paper, a finite element model is proposed and verified with the updated mechanicalactuator experiments. Parametric study is made to investigate the pipe-soil interaction mechanism for the current-induced pipeline lateral instability. The following conclusions can be drawn: • The finite element model can effectively simulate

the behaviour of pipeline losing lateral stability in ocean currents under two end-constraint conditions, i.e. the anti-rolling pipe and the freely-laid pipe. The ultimate lateral soil resistance can be obtained from the load vs. displacement curve. • A lateral-soil-resistance coefficient (η) is presented for better understanding pipe-soil interaction mechanism. The value of η decreases gradually to a constant with the increase of G. The magnitude of η for the examined anti-rolling pipes is much lager than that for the freely-laid pipes, indicating the end-constraint condition affects significantly the lateral stability of the untrenched pipeline in ocean currents. • For a certain case of end constraint (anti-rolling or freely-laid), the variation of η with G is affected by various parameters, including soil internal friction angle, pipe-soil friction coefficient, etc. The effect of pipe-soil friction coefficient is more obvious for the smaller value of G. • When evaluating the capacity of lateral resistance, it would be beneficial to further examine and get correlation with the maximum pipeline penetration (including initial and additional settlement) and the development of the plastic-strain zone beneath the pipeline.

REFERENCES Cathie, D.N., Jaeck, C., Ballard J-C. & Wintgens, J-F., 2005. Pipeline geotechnics – state of the art. In: Frontiers in Offshore Geotechnics: ISFOG 2005. Eds: Gourvenec S. and Cassidy M., New York: Taylor & Francis, 95–114. Det Norske Veritas, 2007. On-Bottom Stability Design of Submarine Pipelines. Recommended Practice, DNV-RPF109. Gao, F.P., Gu, X.Y. and Jeng, D.S., 2003. Physical Modeling of Untrenched Submarine Pipeline Instability. Ocean Engineering, 30 (10): 1283–1304. (SCI, EI) Gao, F.P., Yan, S.M., Yang, B. and Wu, Y. X., 2007. Ocean currents induced pipeline lateral stability on sandy seabed. Journal of Engineering Mechanics,ASCE, 133(10): 10861092. Hibbitt, Karlsson and Sorensen Inc, 2006. ABAQUS Theory Manual, Version 6.5–1. Jones, W.T., 1978. On-bottom pipeline stability in steady water currents.Journal of Petroleum Technology, Vol. 30, 475–484. Palmer, A. C., Steenfelt, J. S. and Jacobsen, V., 1988. Lateral resistance of marine pipelines on sand. Proceedings of 20th Annual Offshore Technology Conference, OTC 5853, 399–408. Teh, T.C. and Palmer, A.C. and Damgaard, J.S., 2003. Experimental study of marine pipelines on unstable and liquefied seabed. Coastal Engineering, 50, 1–2: 1–17. Wagner, D. A., Murff, J. D., Brennodden, H., and Sveggen, O., 1987. Pipe-soil interaction model. Proceedings of Nineteenth Annual Offshore Technology Conference, OTC 5504, 181–190. White, D.J. and Randolph, M.F., 2007. Seabed Characterisation and Models for Pipeline-Soil Interaction. Proceedings of the Seventeenth International Offshore and Polar Engineering Conference, Lisbon, 758–769. Zhang, J., Stewart, D. P., Randolph, M. F., 2002. Modeling of shallowly embedded offshore pipelines in calcareous sand. Journal of Geotechnical and Geoenvironmental Engineering, ASCE, Vol. 128, 363–371.

ACKNOWLEDGEMENTS The work reported herein is financially supported by Knowledge Innovation Program of the Chinese Academy of Sciences (No. KJCX2-YW-L02) and China National S&T Major Project (No. 2008ZX05026-005).


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Vertical cyclic testing of model steel catenary riser at large scale T.E. Langford & V.M. Meyer Norwegian Geotechnical Institute, Norway

ABSTRACT: Deepwater offshore developments often use steel catenary risers to carry oil and gas between subsea facilities and floating platforms or vessels. Multidisciplinary research on steel catenary risers has investigated the interaction between these long cylindrical structural elements and the often soft and compressible seabed. The seabed is typically included within a structural model as an equivalent vertical stiffness, or hyperbolic curve, which may be based upon model riser tests or numerical simulations. Although these models may be sufficiently complex to capture the non-linearity of a single load phase, the vertical loading itself is variable throughout the lifetime of a riser. A large scale model has been subjected to vertical cyclic loading to demonstrate how set-up of a high plasticity soil leads to large regains in strength and stiffness, likely to be the result of thixotropy. Details of the test performance and results are presented, together with a discussion of the differences between the tests and other recent work.

1

INTRODUCTION

2.2

Offshore oil and gas developments in deepwater locations often use Steel Catenary Risers (SCRs) to link seabed facilities with floating platforms. The interaction of SCRs with typically soft seabed soils is an important research area in offshore geotechnics. Recent work (e.g. Hodder et al., 2009) investigated the effects of reconsolidation between cyclic phases of vertical loading. Such work relates to locations where seasonal storms are prevalent, with relatively benign loading conditions in between, and have been modeled in the centrifuge using kaolin. The tests described herein permit comparison with large-scale 1g model tests in high plasticity West African clay.

Soil properties

The soil for the current test program was high plasticity West African clay with natural water content around 150%. The liquid and plastic limits of the clay were measured to be 160% and 60% respectively, giving a plasticity index IP = 100%. The clay was consolidated using a vacuum applied within a rubber membrane at a stress of 15 kPa. The typical clay depth during testing was around 1.25 D. Mini T-bar (25 mm × 120 mm) penetration tests were performed immediately prior to the riser model tests in order to evaluate the shear strength profile. The T-bar tests were performed on the submerged soil at a penetration rate of 2 cm/s. The undrained shear strength, su , has been interpreted from the T-bar resistance using a factor NT-bar = 10.5. Figure 1 shows the

2 TEST PROGRAM 2.1 Test apparatus The pipe/riser model test apparatus at NGI uses a biaxial rig with hydraulic actuators and 50 kN maximum load capacity. The tests are executed using an MTS FlexTest SE control system and MOOG servo valves. For the tests presented herein, the model was subjected to vertical loading only using the vertical actuator. Further details of the apparatus are described by Langford et al. (2007). The test bin has plan dimensions of 3.6 × 1.75 m and has space for 6 footprints, 3 of which were used for the testing presented herein. The test riser section has a length of 1300 mm and diameter of 174 mm. It is finished with a roughened polypropylene coating. The base of the pipe is fitted with filter-capped holes which are in turn linked to pore-pressure transducers.


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Figure 1. Undrained shear strength from T-bar tests.

Table 1.

Summary of tests.

Penetration (z/D) Time between cyclic phases Test Max Min (days) 2-1 0.5 2-2 0.5 2-3 0.5 ∗

0.0 0.0 0.0

1 (36∗ ) 0.1 16–18

Additional cyclic phase after 36 days. Figure 2. Initial penetration to 0.5 D.

interpreted shear strength profiles, together with socalled ‘corrected’ profiles which are based upon the average of the measurements. This correction is based upon a non-linear increase in T-bar until it reaches a steady-state value (White & Randolph, 2007). Different interface coefficients are used ranging from α = 0.0 (smooth) to α = 1.0 (rough) to define the penetration factor Np for smaller penetrations according to Aubeny et al. (2005). In summary, su is inferred to be approximately 5 kPa at the surface, increasing to around 7 kPa at the base of the test bin. 2.3 Test parameters Each test footprint was subjected to 3 cyclic episodes, each comprising 20 cycles of vertical loading. Penetration to 0.25 D was performed using a constant displacement rate of 0.5 mm/s. Subsequent vertical cyclic tests were performed to consistent maximum and minimum embedments of 0.5 D and 0.0 D respectively. Cyclic penetration was performed according to a displacement controlled sine function with a cyclic period of 350 seconds (0.0029 Hz) giving an average displacement rate of 0.5 mm/s. The test program is summarized in Table 1. The delay between cyclic phases was 0.1 to 18 days for which the pipe was fixed at 0.25 D (i.e. above the surface of the trench). These delays allowed for set-up of the soil before retesting. Following completion of the third cyclic phase for Test 2-3, the rig was shifted back to the position of Test 2-1 and an additional cyclic phase was performed after a longer delay of 36 days. Successive cyclic test phases are given suffices A, B and C (and D for Test 2-1) within the paper. Pore pressures were measured at 3 positions along the pipe invert: U1 and U3 were at either end of the model whilst U2 was at the midpoint. 3 3.1

RESULTS AND INTERPRETATION Initial penetration phase

The initial penetration of the model riser into the soil is shown in Figure 2 and comprises monotonic loading to 0.25 D and further penetration to 0.5 D as part of the first cyclic phase. There was a small delay between these two phases which caused a drop in load at constant penetration, followed by an increase in load due to the short set-up period.

Figure 3. Vertical cyclic penetration for 1st cyclic phase.

A reduction in test rate towards maximum penetration occurs with the sine-time displacement loading function. This gives a corresponding reduction in soil resistance which can be explained by the change in soil strength with rate of strain. Test results show the contribution from soil resistance, qs , where the buoyancy component, qb , has been subtracted from the total resistance, qt , as suggested by Hodder et al. (2009). 3.2

Plots of soil resistance versus normalised penetration are shown on Figure 3 for the first cyclic phase of each test footprint (2-1-A, 2-2-A and 2-3-A). The results show the same response in each case whereby the soil resistance for penetration and extraction reduces for each cycle. It can also be seen that the soil appears to detach from the pipe and re-penetrates at a normalized penetration of around 0.35 to 0.45 D, increasing during the course of each test phase. The soil resistance for the 2nd cycle is around 65% of that for the 1st cycle, and reduces to as low as 10% of the original value after 20 cycles. The second and third cyclic phases for Test 2-1 (2-2-B and 2-2-C) are shown on Figure 4. The trench depth remains around 0.40 to 0.45 D for these phases, and the range of measured resistance during each phase is much smaller than for the first phase where the clay was originally intact. Similar results were obtained for Tests 2-1 and 2-3 respectively.


Cyclic penetration

Table 2.

Summary of soil resistance values.

Test / phase

Time (days)

2-1-A 2-1-B 2-1-C 2-1-D 2-2-A 2-2-B 2-2-C 2-3-A 2-3-B 2-3-C

–

Soil resistance, qs (kPa)

Figure 4. Vertical cyclic penetration for 3 cyclic phases of Test 2-2; scale on y-axis for left hand plot is 4 times larger than for middle and right hand plots.

1 1 36 – 0.1 0.1 – 18 16

Cycle N

Cycle N + 19

qs,N+1 / q∗s,N

17.8 2.5 2.2 3.9 17.8 4.6 3.4 17.7 5.2 3.7

1.9 1.5 1.4 1.2 4.4 3.2 2.2 3.8 2.6 1.8

– 1.32 1.45 2.79 – 1.06 1.05 – 1.38 1.44

∗

Where ‘N’ is last cycle of previous phase and ‘N + 1’ is first cycle of current phase.

Figure 5. Change in pore pressure for 3 cyclic phases of Test 2-2; scale on y-axis for left hand plot is 5 times larger than for middle and right hand plots. Figure 7. Definition of normalised secant stiffness, Ksec .

response’ after 10 cycles. In contrast, the results from the current tests show that the soil resistance does not stabilize even after 20 cycles. The reason for this difference is unclear. From earlier T-bar testing the sensitivity (St ) of the soil is judged to be between 2 and 3 (Langford & Aubeny, 2008b). However, the observed pipe model test behavior suggests that water may be becoming entrained within the soil body during tests, thereby causing further reduction of the resistance. Table 2 summarises values of soil resistance for the first and last cycle of each test phase. The peak soil resistance for the first cycle is around 18 kPa for all three footprints, which suggests a homogeneous clay bed. However, the value after 20 cycles varies between 1.9 and 4.4 kPa, where the lowest ‘remoulded’ strength is given for the footprint closest to the tank wall where boundary effects may have some influence.

Figure 6. Change in peak soil resistance per cycle.

Figure 5 shows the measured change in pore pressure from Test 2-2. These data are not currently used in the interpretation, however general agreement between the changes in soil resistance and pore pressure are evident from the tests. Figure 6 shows the change in soil resistance per cycle for the three cyclic phases in each test. The Figure has been arranged so that the test with the shortest set-up period (2-2, 0.1 days) is on the left and that with longest set-up period (2-3, 16–18 days) is on the right. Hodder et al. (2009) presented results in kaolin where the soil resistance reached a ‘steady remoulded

3.3

The method used to derive the normalised cyclic stiffness, Ksec , is shown in Figure 7, and may be defined by the following equation:


Normalized secant stiffness

Figure 8. Normalised secant stiffness, Ksec , versus normalised extraction, z/D, for Test 2-1; scale on y-axis for left hand plot is 10 times larger than for middle and right hand plots.

Figure 10. Normalised secant stiffness, Ksec , at z/D = 0.025 versus cycle number.

Figure 9. Normalised secant stiffness, Ksec , versus normalised extraction, z/D, for the last cycle of each test (N = 20).

where Vs is the change in unit pipe-soil load (kN/m), z is the change in displacement (m) and qs,initial is the initial pipe bearing capacity (kN/m2 ). The normalised secant stiffness has been evaluated for each cycle at different normalized penetrations. The resulting values for the different phases of Test 2-1 are shown on Figure 8. The results show that the normalised secant stiffness is highly non-linear when compared to the normalised penetration. Furthermore, Ksec drops rapidly during the first cyclic phase before reaching a range which is more typical of the values for the 2 subsequent cyclic phases. Figure 9 shows contours of Ksec from the different tests, as derived from the last cycle of each phase. There is a general trend of reduction in Ksec during the 3 phases for each test; however the curves plot remarkably close together. The range in cyclic secant stiffness between the different tests is similar to that for the soil resistance shown previously. Figure 10 shows the change in Ksec per cycle for the three cyclic phases in each test, based on a normalized extraction of z/D = 0.025. There is an increase in Ksec after each set-up period followed by a trend of general decrease throughout each cyclic episode. This trend is the opposite to that presented by Hodder et al.


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Figure 11. Comparison of results for Test 2-3 with data given by Hodder et al. (2009).

(2009) for a test series in kaolin where there was also no contact stress between the riser model and soil during the set-up periods. Figure 11 shows an example of the results from Hodder et al. (2009) where Ksec at the end of each phase increased from one phase to the next. Inspection of the graphs for soil resistance versus normalised displacement reveals the same trend, where the measured resistance at the end of one phase is greater than that preceding it. Since the remoulded strength is linked to the water content, it is therefore reasonable to assume that the water content has decreased during the test series. This could be due to the properties of the kaolin or stress regime within the centrifuge. It should be noted that the set-up factor for kaolin differs from natural clays. As reported by Andersen & Jostad (2004), kaolin is not thixotropic, but it may give higher setup than natural clays because the effective stress after redistribution of pore pressures is higher and because of faster global pore pressure dissipation. For the current tests, once the soil has been remoulded, and possibly experienced an increase in water content, the stresses near the seabed are much lower than those used for the initial consolidation, and the general trend of decrease soil resistance and normalised secant stiffness is not unexpected. If many more phases were performed, or if the number of cycles

Table 3.

Summary of normalised secant stiffness values. Normalised secant stiffness, Ksec (−)

Test / phase

Time (days)

2-1-A 2-1-B 2-1-C 2-1-D 2-2-A 2-2-B 2-2-C 2-3-A 2-3-B 2-3-C

– 1 1 36 – 0.1 0.1 – 18 16

Cycle N

Cycle N + 19

Ksec,N+1 / K∗sec,N

57.7 8.1 7.2 11.6 59.1 12.1 9.5 54.1 13.9 10.6

5.7 4.8 4.7 3.7 11.5 9.2 5.9 9.9 6.9 5.5

– 1.42 1.51 2.50 – 1.05 1.03 – 1.40 1.53

Figure 13. Effect of time delay between cyclic phases on qs and Ksec ; ratio of values from the 1st cycle of phase p to values from 20th cycle of phase p.

∗

Where ‘N’ is last cycle of previous phase and ‘N + 1’ is first cycle of current phase.

Figure 12. Effect of time delay between cyclic phases on qs and Ksec ; ratio of values from the 1st cycle of phase p+1 to values from 20th cycle of phase p.

was increased, it is possible that values may stabilize. Different results may be expected where there is a contact stress between the riser and seabed during the set-up periods. Table 3 is similar in format to Table 2, but rather presents values of Ksec at the start and end of each cyclic phase, giving the increase due to set-up in each case.

The figure shows a clear trend of increasing soil resistance and normalised secant stiffness with time, which is as expected.There are some issues that require further attention. For instance, the values for 1 day setup are very similar to those for 16-18 days set-up. The additional phase for Test 2-1 (2-1-D) included a longer 36 day set-up period. The associated increase in soil resistance and normalised secant stiffness is greater than would be expected from the trends for the other test phases. If the location for Test 2-1 is affected by the tank boundaries, it is possible that the horizontal and mean stresses have been increased by the deformed soil pushing on the tank walls, thereby giving a higher degree of set-up. This effect may have been less relevant for the 1 day set-up periods for phases 2-1-B and 2-1-C, since these would have been dominated by thixotropy. Nonetheless, this phenomenon will be explored during future tests. Figure 13 presents the results in a different way, where the relative decrease in soil resistance or normalised secant stiffness is plotted versus the corresponding time delay for set-up. This ratio can be considered analogous with the development of sensitivity in the soil. For reference, the ratios based on the soil resistance for the first cyclic phase in each case vary from around 4 to 9, using the values in Table 2. The corresponding ratios for Ksec range from 5 to 10. Once again, the results for the additional test phase at 2-1 (2-1-D) for 36 days set-up seem to lie above the trend for the other tests. This anomaly warrants further investigation at a later stage. 3.5

3.4 Time delay effects The time delay between cyclic phases allows set-up of the soil. For the current tests, since the set-up periods are relatively short, the set-up is likely to be dominated by thixotropy, which can be significant for high plasticity soils. Set-up due to changes in effective stress may be more relevant for longer set-up periods. Figure 12 presents the results from Tables 2 and 3 in terms of set-up. Also shown are typical thixotropy test results for high plasticity clays (Ip = 75 to 90%) taken from Andersen & Jostad (2004).


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Comparison with earlier work

Langford & Aubeny (2008a) present results from vertical displacement-controlled tests with set-up periods in between cyclic phases. There were also some important differences between the tests: • The displacement range was much smaller (0.01 D

instead of 0.25 D) • The set-up periods were limited between 1 and

4 hours • The riser model was kept in contact with the soil

during the set-up periods

• The soil consolidation stress was lower, giving a

secant stiffness. This increase seems to be related to the length of set-up period in a manner analogous to thixotropy in high plasticity clays.

lower shear strength profile Nonetheless, similar trends were obtained as for the current testing. The soil resistance dropped significantly throughout the test series. The magnitude of this decrease was far greater than that shown for the current tests, although this could be due to the very small range of displacement. However, there were also significant relative increases in measured resistance after the short set-up periods. Furthermore, the values of normalised secant stiffness at the end of the cyclic phases were very similar in spite of the decrease in soil resistance. Load-controlled tests from the same test bin also indicated significant set-up of the soil between cyclic test phases, in this case resulting in a reduction of permanent displacement per cycle. 4

CONCLUSIONS

Large scale vertical cyclic riser model tests were performed to investigate the performance of a high plasticity West African clay. The tests included set-up periods of different durations to allow the remoulded soil to regain strength between cyclic test phases. The tests showed the following trends • There is good agreement between the load-

displacement curves for the initial penetration phase of each test, which suggests a homogeneous soil profile. • The cyclic test phases reveal incomplete stabilization of the soil resistance, which differs to the trend shown by recent centrifuge tests with kaolin where stabilization occurred after 10 cycles. Incomplete stabilization suggests that water may have become entrained within the soil body during the pipe model tests. • There is a general trend for decrease in soil resistance and normalised secant stiffness from the first cyclic episode to the subsequent episodes. • The set-up periods of between 0.1 and 36 days allow a partial increase of soil resistance and normalised


ACKNOWLEDGEMENTS The testing presented herein was partially funded by the Norwegian Research Council. The authors are grateful for help from colleagues in the laboratory and workshop at NGI, Oslo. Chuck Aubeny from Texas A&M provided helpful advice during planning of the test program and subsequent interpretation of results. REFERENCES Andersen, K.H. & Jostad, H.P. 2004. Shear strength along inside of suction anchor skirt wall in clay. OTC16844, Offshore Technology Conference, 3–6 May 2004, Houston, Texas, USA. Richardson: OTC. Hodder, M., White, D.L. & Cassidy, M. 2009. Effect of remoulding and reconsolidation on the touchdown stiffness of a steel catenary riser. OTC19871, Offshore Technology Conference, 4–7 May 2009, Houston, Texas, USA. Richardson: OTC. Aubeny, C.P., Shi, H. & Murff, J.D. 2005. Collapse loads for a steel cylinder embedded in trench in cohesive soil. Int. J. Geomechanics, ASCE 5(4): 320–325. Langford, T.E., Dyvik, R. & Cleave, R. 2007. Offshore pipeline and riser geotechnical model testing: practice and interpretation. Proc. Conf. on Offshore Mechanics and Arctic Engineering (OMAE), San Diego, California, USA. New York: ASME International. Langford, T.E. & Aubeny, C.P., 2008a. Model tests for steel catenary riser in marine clay. OTC19495, Offshore Technology Conference, 5–8 May 2008, Houston, Texas, USA. Richardson: OTC. Langford, T.E. & Aubeny, C.P., 2008b. Large scale soilriser model testing on high plasticity clay. Proc. 18th Int. Offshore & Polar Engineering Conference (ISOPE), Vancouver, Canada, Vol. 2, pp. 80–86. Cupertino: ISOPE. White D.J. & Randolph M.F., 2007. Seabed characterisation and models for pipeline-soil interaction, Proc. 17th Int. Offshore & Polar Engineering Conference (ISOPE), Lisbon, Portugal. Cupertino: ISOPE.

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Kupe gas project pipeline – optimisation of discrete rock berm design shore approach B.L. Larsson Technip Oceania Pty. Ltd., Perth, WA, Australia

ABSTRACT: The Kupe Gas Project, located 30 km offshore the NW Taranaki Peninsula of the New Zealand North island consists of a Wellhead Platform, 30 km of subsea 12 pipeline and a 6 service umbilical connected to the onshore processing plant and export pipeline. This paper describes the optimisation of the near shore approach rock dumping engineering design and operations executed to provide a long term solution for the rigid pipeline and unarmoured umbilical crossing of an extensive and culturally sensitive boulder field in very shallow water conditions. The use of discrete rock berms constructed at regular predetermined intervals provides a robust solution for crossing the boulder zone. By controlling the elevation of the ocean floor, the risks to product arising from point loading, excessive free-spans and long term erosion and stability problems have been managed. 1

INTRODUCTION

1.2

1.1 Kupe Gas Project The Kupe Gas Project consists of an unmanned Wellhead Platform, 30 km of subsea 12.75 pipeline and a 6 service umbilical with an onshore processing plant, located in the Taranaki geological basin on the West side of the North Island of New Zealand. Origin Energy Resources (Kupe) Limited is the nominated operator for the Joint Venture partners Origin Energy, Genesis, New Zealand Oil & Gas and Mitsui. The project, commissioned in March this year, will provide 15% of the domestic gas consumption of New Zealand. The Engineering, Procurement, Installation and Commissioning contract was executed by the Technip-Origin Energy Alliance, see Figure 1.

Scope of paper

The emphasis of this paper is the near shore crossing discrete rock berm design and construction.The design provided the Project with a technically and economically feasible solution for the near shore approach and crossing of a 2 km wide boulder field. 2 2.1

SUBSEA TRANSPORTATION SYSTEM Rigid export pipeline

The majority of the 12.75 (323.9 mm) line pipe is DNV 450 MPa with 22.2 mm Wall Thickness (WT) and 2.2 mm 3LPP coating. For the HDD crossing the WT increased to 25.4 mm and for the boulder field section, the coating increased to 6 mm. 2.2

Umbilical

The unarmoured umbilical is designed and supplied by Duco – a Technip umbilical manufacturing company. The service umbilical includes MEG and chemical injection lines, 3-phase 33 kV power and fibre optics sheathed in HDPE plastic. 3

Figure 1. Kupe field. © 2011 by Taylor & Francis Group, LLC

NEAR SHORE APPROACH

The subsea HDD shore crossing exits 2 km from the shoreline in sand at a water depth of 14 m (LAT). The near shore approach was categorised as the route alignment from the HDD subsea exit to the end of an extensive boulder field. Here, the pipeline and umbilical traverses a rock boulder zone in a water depth between 14 m and 22 m over a length of approximately 2 km.

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Figure 3. Shoreline boulders.

Figure 2. The Taranaki coastline.

Table 1.

Table 2.

Metocean data.

Description

Water Depth U∗s U∗cwd Hs Hmax

Soil As-built data. Location

Unit –

1YR KP0

1YR KP27

100YR KP0

100YR KP27

m m/s m/s m m

34.0 1.09 0.44 6.38 11.86

14.0 1.67 28.0 4.71 8.75

34.0 1.88 0.76 9.17 17.05

14.0 2.25 0.42 7.6 13.14∗∗

Description

From-KP To-KP Soil description

Grey sands

0

14.4

Sand/Peat

14.4

17.6

Boulder field 25.6

28.0

∗

( ) Note 1: Significant velocity (Us ) and Current velocity (Ucwd ) taken at 1.0 m Above Sea bed (ASB). (∗∗ ) Note 2: The back-analysed maximum wave height derived from the statistical data is capped at 12.2 m due to seabed bathymetry capping Hmax in more shallow waters.

4 WEATHER Situated on the South side of the Taranaki peninsula the field is exposed to the weather arriving from the Southern Ocean, and subject to severe (winter) storms, see Figure 2 and Table 1. 5

GEOLOGY

The shoreline is formed by 40 m high cliffs, and subject to an average yearly erosion rate of 1 m. The seabed is generally covered by boulders resting on top of an underlying siltstone base over the 4 kms closest to the shore, with only isolated areas of sand with less rock boulders. The majority of the boulders are 200–400 mm, but boulders up to 2.5 m in diameter are present. Boulders are generally made up of andesite,formed from lahar lava flows and erosion of the coastline, see Figure 3. 5.1

Dark grey fine to course sand. Generally very loose to dense to very dense within the trenching depth. Grey sand as per above. Lenses of peat and silt. High undrained shear strength >180 kPa generally of limited thickness. Heavy black (Iron) sands Underlying siltstone Substrata. Isolated patches of superimposed sand. The majority of the boulders are 200–400 mm ranging up to approx. 2.5 m in diameter.

HDD exit (2 km offshore) to 34 m water depth at the WHP (30 km offshore), see Table 2. Cone penetration tests performed along the route alignment indicated soil strength of 2 m boulders

Geotechnical data

Offshore the boulder field section, the seabed is predominantly sand and silty-sand, with isolated patches of dense sand and peat. The seabed is relatively flat with a gradual decline from 14 m water depth at the © 2011 by Taylor & Francis Group, LLC

6

LOCAL CONSENT

The Alliance had a clear objective to keep an open dialogue and work together with the local population in

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the Taranaki region, accounting for regional employment, financial and environmental considerations. Attention was given to how the ancestral owners of the land, the Iwi people commonly known as Maori of New Zealand and their ethnic, beliefs and religion could be respected without excessive intrusion by the Project. As the rock boulders are considered to be the ancestors for the local Iwi people, the extended HDD shore crossing in combination with the rock dumping solution allowed the Alliance to install the pipeline production system with a minimum of disturbance. 7

DESIGN CRITERIA

7.1 Route alignment The final route alignment accounts for optimised alignment vs. dominant design weather direction, depth of boulder field and ensured that the HDD subsea exited in an area of sand.

In short all the excavation options were subject to: – A constraint on availability of suitable equipment; – A risk of excavation works not being completed within one summer causing delay to the offshore campaign; – A potential for extensive scour creating free-spans and a risk of boulders being displaced by storms in an uncontrolled way upon the pipe or umbilical product. 8.2 Infill methods To avoid excavation, the alternative was to use infill methods, dominated either by pre-lay mattresses or pre-pipeline installation rock dumping options. Due to the extent of the boulder field, the prelay mattress option was considered not to be technically, nor financially viable for the project. 9

ROCK BERM DESIGN CONCEPT

9.1 Design approach 7.2 Boulder field requirements The near shore boulder field represents a significant risk to the pipeline and umbilical during installation and throughout field life, e.g. point loading, mechanical damage, deformation and over stress of product around rock outcrop, increased risk of free-spans development etc. 7.3 Environmental loading The offshore bathymetry of the Kupe field is quite complex, with a dominant S-SW long period swell semi-aligned with the pipeline with potential for a more wind-driven wave loading approaching as a quartering sea. The tidal current is orientated almost perpendicular to the pipeline due to strong currents through the Cook Sound in W-E direction. 8

NEAR SHORE APPROACH SOLUTIONS

A number of options were evaluated before the Project decided in favour of the discrete rock berms. The various options can be grouped into either clearance or infill methodology. 8.1 Clearance methods Initially a number of excavation options were assessed, and included barge and surface excavator spreads, suction-hopper dredger, excavator modified for subsea use or blasting. The blasting alternative was dismissed quickly, as it would have had the largest environmental impact and would be inconsistent with the objectives of the Alliance of mutual respect for the Iwi people’s beliefs and religion. © 2011 by Taylor & Francis Group, LLC

Typically, in the past, rock dumping has been performed as a continuous rock berm. Either as a prepipelay operation to provide a more level seabed or for stabilisation purposes post-pipelay to provide protection against scour and related free-spanning, stability in storm conditions and protection against fishing and trawling gear. Based on the technical, economical and sociological/ethnical assessments and interfaces, the Project decided to proceed with a near shore approach based on rock dumping. The construction of a series of discrete rock berms effectively addresses a number of constraints including: – Less disturbance of the seabed and the Iwi ancestral rocks; – The elevated and globally level seafloor, i.e. well defined allowable difference in height between each adjacent rock berm, reduce the required pre lay and post lay rectification of free-span; – Height control of rock dumping is less critical with discrete berms, whereas only the shorter alignment of a rock berm providing the actual support for the pipeline is in contact with the product; – Known dynamic free-span allows the structural integrity including fatigue issues to be assessed and addressed during the detailed engineering phase ahead of pipeline installation; – Product seabed stability improves since a spanning pipeline generates minimal lift and hence less mattress weight requirements; – Less pipeline/umbilical length subject to abrasion; – Lower overall cost. 9.2 Detailed design Rather than implementing an extensive hydro dynamic physical modelling, the Project has worked closely with the design consultant Coastal Engineering Solutions based in Melbourne, Australia, with the

811

Offshore Engineering Division from Technip’s Specialist Centre in Aberdeen, UK, reviewing the results. The winter between pipelay installation and rock berm construction allowed the rock berm materials to consolidate prior to pipelay. Rather than the actual rock berm design itself, the allowable mattress spacing and pipeline free-span defines the individual distance between two adjacent rock berms. The detailed design included: – Rock stability assessment; – Identifying the rock sizing and grading curves (D15 / D85); – Assessment of discrete rock berm vs. a continuous rock berm including stabilisation methodology; – Assessment of implications for pipeline upheaval and/or lateral buckling; – Hydrocarbon flow assurance; – Pipeline design for dynamically loaded pipeline; – Umbilical design incorporating dynamic loading; – Fatigue analysis for pipeline & umbilical; – Fishing and trawling gear interaction; – Rock quality specification; – Rock quantity assessment.

rock particles were subject to an abrasion test and a requirement for cubical, compact unit-rock shape. Not more than 10% of the rock particles were allowed to have an elongated, brick like shape. During construction, samples were taken from every 10th truck leaving the quarry by employing the Queen Mary and Westfield (QMW) abrasion test, the elongation check was performed by visual monitoring and recording at discrete intervals.

10.4

With the rock berms being fully immersed in seawater there were limits set on absorption of water for both individual stones and a sample average. Even though not subject to freeze and thaw cycles, the chemical composition in combination with the absorption was controlled to reduce the risk of excessive erosive deterioration potential due to chemical reactions over the long term, generally related to (smectite) clay minerals.

11 10

ROCK SPECIFICATION

To fulfil the design criteria, there were also specific requirements on the rock aggregate itself. These criteria arise from International and local New Zealand/ Australian Standards, industry research findings, design specific criteria and Technip Group experience. The combined criteria reflect the requirements of the detailed design and define the quality of the rock aggregate in terms of geometric shape, chemical composition, density and gradation etc.

DISCRETE ROCK BERM IMPLICATIONS

The decision to construct discrete rock berms also imposed additional engineering requirements for the design of the production system. Arising from the use of discrete rock berms, as opposed to continuous support, additional engineering should include assessment of:

10.1 Rock aggregate density The minimum density of the rock aggregate particles accounts for the 100YR RP wave load condition. The unit-rock particle size vs. specific gravity needs to be assessed carefully, as it quickly impacts the construction methods and rock aggregate size. 10.2 Edginess To improve stability and unit-rock particle interlocking, the rock aggregate is an angular quarried and crushed rock with a certain edginess, rather than a smooth rounded shape of each rock itself. By implementing a series of selected mesh sizes and percentages from four (4) defined more narrow spanning ranges, the overall rock gradation ensured a rather steep gradation with a certain minimum of larger rocks to ensure the overall stability if subject to the design 100YR RP wave condition. 10.3 Shape To maintain the long term integrity of the rock berms, in addition to the basic shape criteria and density, the © 2011 by Taylor & Francis Group, LLC

Chemical composition

812

– Risk for excessive point loading at contact points exerted either during the lay, or long term field life; – Predefinition of criteria for pipelay installation including layback and contact loads.Attention to the mattress design including exit/entry of the product at either edge of a stabilisation mattress, and in turn mattress installation landing velocity limitation; – Use of static design criteria is required such that if subject to hydro-dynamic loading or other lateral loading the system is not moved beyond the elevated rock berm support; – Pipeline and umbilical lay tolerances to account for lay tolerances and temporary stabilisation. Width of rock berm to accommodate mattresses; – Pipeline potential to move or oscillate, could lead to both local coating damage, with increased risk of local corrosion or fatigue damage, with the potential to cause a pipeline / umbilical failure; – Shape and construction of the stabilisation mattresses vs. rock aggregate shape support taking into account the geometry of the product. Consider coating thickness increase; – Structural integrity criteria, including fatigue considerations, for pipeline and umbilical when used for a dynamic application; – Piggy-back of light-weight product umbilical to pipeline, allows increased predefined free-spans. Design to account for strap/saddle spacing vs. dynamic loading including fatigue and local point

– The created predefined free-spans results in pipeline and umbilical product being exposed to dynamic loading and fatigue issues; – By securing the pipeline with concrete stability mattresses on top of the discrete rock berms and using the pipeline as carrier pipe for the umbilical, the long term structural integrity is maintained for the whole production system; – Rock unit shape and static stability measures accounts for pipe cross section to prevent local buckling or coating damage during installation and permanent design alignment; – Design of the stabilisation mattresses and other stabilisation measures shall reflect the geometry and spacing of the discrete rock berms; – Reduced pre-pipelay span rectification with rock berms constructed within predefined height tolerance; – Less pipeline/umbilical areas subject to abrasion; – Pipeline stability is improved since a spanning pipeline generates minimal lift and hence mattress weight requirements are reduced;

Figure 4. Rock berm route alignment.

Figure 5. Rock berm cross section.

Particular to the discrete rock berm design, it is concluded that:

loading. Here, the rock berm distance was determined by pipeline allowable free-span accounting for VIV, fatigue and ultimate structural integrity.

12

– The use of discrete rock berms provides a robust solution for crossing boulder zones and provides system stability. By controlling the elevation of the ocean floor, the risk of point loading, excessive freespans and long term erosion and stability problems for the pipe can be managed; – Discrete rock berms allow increased control of project cost and construction duration; – The solution provides a predefined and controlled free-span situation with reduced post pipelay rectification; – Careful design of rock-unit size and gradation vs. inherent specific gravity for rock material that can be sourced close to site is essential for viability of solution; – Toughness, abrasion resistance and durability requirements of the rock aggregate to reflect site environment and design life; – Height control of rock dumping is less critical with discrete individual berms, than for a continuous rock berm.

ROCK BERM CONSTRUCTION

The rock is installed as discrete rock berms at intervals coinciding with the permissible free-spans. The berms are oriented perpendicular to the alignment route with a 6 m × 2.5 m stability mattress at each rock berm. The rock berm flat top surface is sized for lay tolerances and mattress size, see Figure 4 and Figure 5. A total of 86 rock berms were constructed using a side dump vessel with survey capability allowing monitoring, for practical purposes, down to ±10 cm of rock berm level and geometry. During the construction of the rock berms, particular attention was paid to: – Limit sea state for rock dumping activities; – Transversally offset rock berm widened to maintain lay corridor at top of rock berm; – Control of rock berm top level to reduce need of post-lay free-span rectification.

ACKNOWLEDGEMENTS 13

CONCLUSION

A careful review of installation criteria and associated weather forecasting services available was carried out. For the installation duration, temporary stabilisation measures were assessed. Particular aspects for the discrete rock berm design and interaction with pipeline & umbilical systems that can be highlighted are:

Technip and the Alliance would like to direct a special thank you to Peter Riedel of Coastal Engineering Solutions who provided the detailed design, and Technip OED Aberdeen, in particular Alasdair Maconochie and John Oliphant for the geotechnical advice during the project engineering phase.

– The pipeline expansion is relatively unaffected by the rock berm support at the cold-end of the production system; © 2011 by Taylor & Francis Group, LLC

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PREFERENCES, SYMBOLS AND UNITS – 3LPP: 3 Layer Poly-Propylene – EHU: Electro Hydraulic Umbilical

– – – – – – – – –

HDD: Horizontal Directional Drill WHP: Wellhead Platform KP: Kilometre Point Lahar: Volcanic type of mudflow or landslide composed of pyroclastic material and water LAT: Lowest Astronomical Tide OD: Outer Diameter RP: Return Period WT: Wall Thickness YR: Year

REFERENCES ABS. 2006. Guide for Building and Classing Subsea Pipeline Systems. Houston: American Bureau of Shipping API 1104. 2005. Welding of Pipelines and Related Facilities, 20th Edition. Washington D.C.: American Petroleum Institute API-RP-1111. 1999. Design, Construction, Operation and Maintenance of Offshore Hydrocarbon Pipelines, 3rd Edition.Washington D.C.: American Petroleum Institute API Specification 5L. 2004. Specification for Line Pipe, 43rd Edition. Washington D.C.: American Petroleum Institute ASME Section VIIl Div 2. 2004. Rules for Construction of a Pressure Vessel. New York: American Society of Mechanical Engineers ASME B16.5. 2003. Pipe Flanges and Flange Fittings. New York: American Society of Mechanical Engineers Brown, G. et al. 2004. Reliability Based Assessment of Minimum Reelable Wall Thickness for Reeling, Proceedings of International Pipeline Conference, IPC04-0733. Calgary: American Society of Mechanical Engineers BS 7910. 2005. Guide to Methods for Assessing the Acceptability of Flaws in Metallic Structures. London: British Standard Institution DNV OS-F101. 2000. Submarine Pipeline Systems. Oslo: Det Norske Veritas DNV CN 30.5. 2000. Environmental Conditions. Oslo: Det Norske Veritas


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DNV RP-E305. 1988. Recommend Practice for On-Bottom Stability of Pipelines. Oslo: Det Norske Veritas DNV RP-F111. 2006. Interference between Trawlgear and Pipelines. Oslo: Det Norske Veritas DNV RP-F105. 2006. Free Spanning Pipelines. Oslo: Det Norske Veritas DNV RP-F107. 2002. Protection of Pipelines. Oslo: Det Norske Veritas Kupe Project. 2005. Final Report for the Geophysical and Geotechnical Pipeline Route Survey, South Taranaki Bight, New Zealand, Document Number P0071 Rev 0. New Plymouth: Kupe Project HR Wallingford. 2005. Kupe Metocean Design Criteria, Kupe Field, New Zealand, Assessment of Pipeline Exposure Risk, Technical Note EBR4048/01, Document Number 5510-02. Wallingford: HR Wallingford Metocean Solutions. 2006. Wave and current Design values for the Kupe Pipeline Doc. No. DESIGN-DATA-01 Rev. B. New Plymouth: Metocean Solutions Metocean Solutions. 2006. Summer season waves and currents for the Kupe pipeline route Doc. No. DESIGN-DATA02 Rev. B. New Plymouth: Metocean Solutions CIRIA. 1991. Manual on the use of rock in coastal engineering and shoreline engineering. London: Construction Industry Research and Information Association CIRIA. 1995. Special Publication 83, CUR Report 154 – Manual on the Use of Rock in Coastal and Shoreline Engineering. London: Construction Industry Research and Information Association Rance, P.J et al. 1970. The threshold of movement of loose material in oscillatory flow, Proc. 12th International Conf. on Coastal Engineering. Washington: International Conference on Coastal Engineering Soulsby, R. 1997. Dynamics of Marine Sands – A manual for practical applications. London: Thomas Telford Publications U.S.Army Coastal Engineering Research Centre. 1984. Shore Protection Manual. Washington D.C: Department of the Army, Corps of Engineers


Model test studies on soil restraint to pipelines buried in sand R. Liu & S.W. Yan School of Civil Engineering, Tianjin University, Tianjin, China

J. Chu School of Civil and Environmental Engineering, Nanyang Technological University, Singapore

ABSTRACT: The buckling of submarine pipelines may occur due to the action of axial soil frictional force caused by relative movement of soil and pipeline, which is induced by the thermal and internal pressure. The likelihood of occurrence of this buckling phenomenon is largely determined by the capability of the soil to resist pipeline movements. A series of large-scale model tests were carried out to facilitate the establishment of substantial data base for a variety of burial pipeline topologies. Results show that the soil resistance depends on the pipe diameters and the depth of cover. According to the uplift test results, the force-displacement topologies with smaller depth of cover are greatly different from those with larger depth of cover. The results of the lateral sliding and axial pull out tests show that the soil resistance initially increases before a peak value is reached and then keeps the same level. For the same covered depth, the lateral soil resistance is more than twice that for uplift.

1

INTRODUCTIONS

Vertical buckling of buried submarine pipeline caused by high temperature and pressure is an important issue endangering the safety and stability of pipeline (Nielsen et al. 1990, Guijt 1990, Liu et al. 2005). In-service hydrocarbons must be transported at high temperature and pressure to ease the flow and prevent solidification of the wax fraction. Thermal stress together with Poisson effect will cause the steel pipe to expand longitudinally. If such expansion is resisted, for example by frictional affects of the foundation soil over a kilometer or so of pipeline; compressive axial stress will be set up in the pipe-wall. When the value exceeds the constraint of foundation soil on the pipeline, sudden deformation will occur to release internal stress, which is similar to the sudden deformation of strut due to stability problems. The upheaval buckling may cause the pipeline destroyed suddenly, as a result, the effective prediction method and protection measure against this phenomenon are important parts in the design of buried submarine pipelines. From the analysis above, it can be known that under a given temperature and pressure design condition, the occurrence of buckling is largely determined on the soil resistance acting on the pipeline. However, the soil resistance depends on a good many factors, such as direction of pipeline movement, amplitude of pipeline buckling, burial depth of pipeline and soil properties. Unfortunately, there is no theoretical method which can determine the soil resistance accurately up to now. As a result, model tests and numerical analysis are both the primary means to research the interaction between the soil and pipelines. Many researchers have done © 2011 by Taylor & Francis Group, LLC

some researches in this area since the early eighties of the last century. The first published work in the field of pipe-soil interaction surfaced in 1981. To design the pipeline for lateral stability and determine the winch capacity for pulling the pipeline, Anand and Agarwal (1981) carried out small-scale model and large-scale prototype experimental studies to determine the frictional resistance between the concrete-coated pipes and the soil in the lateral as well as the longitudinal directions. Taylor et al. (1985; 1989) chose sand as the supporting medium in view of North Sea conditions and carried out the pull-out tests and axial friction tests. In paper Boer et al. (1986) described the results of full scale pull-out tests for a 2 m long test section of a 12.75 inch O.D. concrete coated pipe covered with gravel. Horizontal and vertical pull out tests for pipeline buried in sand and soft clay were performed by Friedmann in 1986. Schaminee et al. presented the results of a fullscale laboratory test program on the uplift and axial resistance of a 4 inch pipe embedded in cohesive or cohesionless soil in 1990. Finch (1999) confirmed that soft clay backfill can be effectively modeled as a frictional material when considering uplift resistance, and derivation of applicable axial friction factors for coated pipelines in sand. Schupp et al. (2006) described a plane strain pipe unburied tests in loose dry sand and initial small scale three-dimensional buckling tests. White et al. (2008) presented a limit equilibrium solution for the uplift resistance of pipes and plate anchors buried in sand based on the model tests. Some centrifuge model tests were conducted by Dickin (1994), Moradi and Craig (1998), Bransby et al. (2001, 2002), Palmer et al. (2003) and Cheuka et al. (2007) to assess

815

Figure 1.

Model test tank scheme and Pull-out Topology.

the uplift capacity pipelines. Based on the tests, the force-displacement relationships of the pipe for different embedment depths had been produced and a semi-empirical design formula had been proposed. However, there are great differences among the results calculated by these various formulas. The differences are owing to the following reasons. One is that the understandings of the soil failure mechanism are at variance, the other is that the soil properties are quite different in different region. This paper chooses fine sand as the soil medium in view of Bohai Gulf geotechnical conditions and carries out a series of large-scale model tests to facilitate the establishment of substantial data base for a variety of burial pipeline topologies. Pipe segments with diameter of 30 mm, 50 mm and 80 mm are used respectively. The pipe segments are buried in different depth-todiameter ratios between 1 and 9. The uplift, lateral and axial resistances are recorded during the tests.

2 2.1

MODEL TESTS Design of experiment model

The purpose of the tests was to obtain the forcedisplacement relationship under different test conditions. To do this, the pipe segments were pulled out in vertical, lateral and axial directions, buried at different depths. The test data could be applied to determine the soil resistance in upheaval analysis of submarine pipelines. The model test tank dimensions were 1 m × 1 m × 3 m, which was assembled by PVC plates. In order to pull out the pipeline in three different directions and eliminate end effects (Tran 1994), two “L” shape notches were slotted symmetrically on each side of the tank. Sliding retaining device was used to prevent the sand from flowing out of the hole with pipeline moving. An organic-glass-plate was installed on one side of the tank to facilitate the observation of soil deformation during the pipe moving. Photograph of the device are shown in Figure 1. Stainless steels pipe segments with diameter of 30 mm, 50 mm, 80 mm and length of 1200 mm were selected to simulate the real pipelines in the test. All the test pipe segments were sealed on ends and extend out of the tank with 100 mm for each side to eliminate the disturbance of pipe ends to soil during the pipe moving. The pipe segments were buried in different depth-to-diameter ratios between 1 and 9. To record © 2011 by Taylor & Francis Group, LLC

Figure 2. Silty sand grading curve.

the peak values of soil resistance during pipe moving, it required that the movement speed of the pipe should be smooth and slow enough and the force acting on the pipe should be steady and continuous. Therefore, the electromotor with a secondary reducer was adopted to control the force application. The rotate speed of the electromotor was controlled at 0.06 mm/s during the test. The force was transferred to the pipe by the steel rope with a group of dynamic and static pulleys. Figure 1 is the sketch of the force acting as pipe moves in different directions.

2.2

Soil sample preparation

Sand was chosen as the soil medium in view of Bohai Gulf conditions and a sieve analysis identified the requisite fine sand (ref. to Fig. 2). Dry condition testing was employed for convenience. According to sieve analysis results, the average particle size of sand used in the test was 0.248 mm, and its non-uniform coefficient was 3.16, the curvature coefficient was 0.95, which belongs to uniform fine sand according to the soil classification code. Laboratory data showed that the maximum and minimum dry density of the sand is 1.76 g/cm3 and 1.57 g/cm3 separately. The fine sand was compacted to a relative density (Dr) equal to 0.5, and the corresponding dry density was 1.66 g/cm3 . The natural repose angle of the sand was 32◦ and internal friction angle was 35◦ .

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2.3

Data recording

Pipeline displacement and soil resistance were the main measurement data in the test. The displacement transducers were installed on both ends of the pipe in accordance with pipe movement direction. Electric pressure sensors were installed on the steel rope which transfers the force in the test. A KYOWA-PCD300A type dynamic data processor had been used. During the test, the displacement transducers, pressure sensor and dynamic data processor were all connected to a computer, which could record the displacement and force data simultaneously by special data processing software.

2.4 Tests procedure The weight of the pipes was measured before the test. Sand was filled into the test tank layer by layer. For each filling layer, sand was compacted to a target bulk density of 16.6 kN/m3 . As the filling depth reaches a required level, a horizontal trench was then excavated to 100 mm depth to the tank bottom where the pipe emplaced. During vertical pull-out test, firstly measured the pull-out force without covered sand and took it as initial value. Secondly, a trench was dug to the design burial depth with 30◦ sidewall. The pipe was laid and loose sand was filled without compaction. Finally, a vertical force was applied to the pipe and the pipe move upwards slowly and smoothly. The time path curve of vertical pull vs. displacement was recorded until the pull out force remains the same level. During lateral pull-out test, firstly, placed the lateral pull out setup. Secondly, a trench was dug to the design burial depth with 30◦ sidewall. The pipe was laid and loose sand was filled without compaction. Finally, applied a lateral pull-out force to the pipe and keep it moving.The lateral pull and the displacement of the pipe were recorded simultaneously until the force approached to a constant value. During axial pull out test, firstly, the axial pull out setup was placed on both ends of the pipe to realize changing the movement direction during the test. Secondly, a trench was dug to the design burial depth with 30◦ sidewall. The pipe was laid and loose sand was filled without compaction. Finally, the axial pull out force and the pipe displacement had been recorded in the process of the pipeline moving back and forth for two cycles. Diameters of the pipe were 30 mm, 50 mm, 80 mm and the depth-to-diameter ratios are between 1 and 9 in the test. The weights of the pipe were 2.8 kg, 6.3 kg and 13.45 kg, separately. For all three movement directions, 81 group tests had been carried out.

Figure 3. Vertical Pull-out Tests Results for H /D =2.

Figure 4. Vertical Pull-out Tests Results for H /D = 8.

Figure 5. Lateral Pull-out Tests Results for H /D = 2.

3 TEST RESULTS ANALYSIS 3.1 Influence of diameter to the soil resistance Averaged vertical pull-out characteristics are illustrated in Figures 3–4 for cases of H /D = 2, 8 and D = 30 mm, 50 mm and 80 mm respectively. The vertical force is denoted by Fv and pipe vertical displacement by Sv. Figures 3–4 indicate that the soil resistance follows a similar procedure for pipes of different diameters. The loci show that only small deformations are onset up to the maximum pullout state, deflection then increasing rapidly down the post maximum falling branch. The maximum pull-out values are indicative of the mechanical effect of pipeline diameter and burial depth. The larger diameter, the greater soil resistance can be reached and relatively the greater displacement is needed. In the case of H /D = 2, the pipe displacement corresponding with maximum soil resistance are 0.3 mm, 1.5 mm, 3.2 mm © 2011 by Taylor & Francis Group, LLC

for D = 30 mm, 50 mm and 80 mm respectively. In the case of H /D = 8, the pipe displacement corresponding with maximum soil resistance increase to 0.375 mm, 3.6 mm, 8.0 mm for D = 30 mm, 50 mm and 80 mm respectively. Averaged lateral pull-out characteristics are illustrated in Figure 5 and Figure 6 for cases of H /D = 2, 9 and D = 30 mm, 50 mm and 80 mm respectively. The lateral force is denoted by FL and pipe lateral displacement by SL . Figures 5–6 indicate that the soil resistance loci of lateral pull out tests are quite different from that of vertical pull out tests. Lateral test results show that the soil resistance initially increases before a peak value is reached and then keeps the same level. The larger the pipe diameter is, the greater the soil resistance reached. And such changes decrease with the pipe depth of cover increasing. For smaller depth of cover, the pipe displacement corresponding to the

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Figure 6. Lateral Pull-out Tests Results for H /D = 9.

Figure 9. Vertical Pull-out Tests Results for D = 80 mm.

Figure 7. Axial friction force for D = 30 mm and H /D = 8. Figure 10. Lateral Pull-out Tests Results for D = 80 mm.

soil resistance initially increases before a peak value is reached and then keeps the same level. Comparing Figure 7 to Figure 8, it can be seen that the pipe displacement corresponding to the maximum soil resistance increases with the pipe diameter. Due to soil disturbances, the soil resistance in axial direction is reduced after reversed pull out tests. Under cyclic reversal, the soil resistance for the pipe with large diameter decreases faster than the pipe with small diameter. For example, the soil resistance after two times reciprocation decreases to 81 percent of that after the first pull-out test for the case of D = 30 mm. And the decrease ratio falls to 69 percent for the case of D = 80 mm.

Figure 8. Axial friction force for D = 80 mm and H /D = 8.

maximum soil resistance is approximately 0.1D for the pipe of different diameters, which is much larger then the one corresponding to the vertical maximum soil resistance. For smaller depth of cover, the lateral soil resistance is more than twice that for uplift. Since the depth of cover will enlarge this change of soil resistance, under thermal and internal pressure actions the vertical buckling is particularly of interest with respect to entrenched submarine pipelines. Pipelines in operation usually experience cyclic temperature changes which lead to pipeline expansive and shrink along the axial direction. In order to study the influence of pipeline extending and shrinking on the soil resistance, the pipe is pulled back and forth twice during the axial pull out test. The axial resistances before and after loading in such way are recorded as shown in Figures 7–8. The axial force is denoted by FA and pipe axial displacement by SA . Figures 7–8 indicate that the © 2011 by Taylor & Francis Group, LLC

3.2

Influence of burial depth to soil resistance

Averaged vertical and lateral pull-out characteristics are illustrated in Figure 9 and Figure 10 respectively. Figure 9 shows that the vertical soil resistance increases obviously with the pipe burial depth. However, there is no linear proportional relationship between the soil maximum resistance and depth-todiameter ratios. When the depth-to-diameter ratio is less than 5, the soil resistance initially increases rapidly before a peak value is reached and decreases to a residual level. When depth-to-diameter ratio is greater than 5, the soil resistance has no significant trend of decrease after its peak. Figure 10 illustrates that the lateral soil resistance initially increases before a peak value is reached and then keeps the same level for all depth-to-diameter ratio. The soil resistance increases

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Figure 13. Soil failure modes for H /D = 2 and H /D = 8. Figure 11. Axial friction force for D = 50 mm and H /D = 2.

Figure 12. Axial friction force for D = 50 mm and H /D = 8.

significantly with pipe burial depth. For the same depth of cover, the lateral soil resistance is more than twice that for uplift. The axial pull-out characteristic is illustrated in Figures 11–12 for cases ofH /D = 2, 8 and D = 50 mm respectively. The results of the axial pull out tests show that the soil resistance initially increases before a peak value is reached and then keeps the same level with the pipe displacement developed. Comparing Figure 11 to Figure 12, it can be seen that the pipe displacement corresponding to the maximum soil resistance increases with the depth of cover. Due to soil disturbances, the soil resistance in axial direction was reduced after reversed pull out tests and the soil resistance decrease faster with larger depth of cover than the one with smaller depth of cover. For example, soil resistance reduces to 74 percent of its initial value for the case of H /D = 2 and reduces to 70% for the case of H /D = 8. 3.3 Soil failure mode analysis In order to observe soil deformation as pipe moving, dyed sands are laid between different backfill sand layers and black straight lines can be seen before test (ref. to Fig. 13). According to the uplift test results, the soil failure modes with smaller depth of cover are greatly different from those with larger covered depth. The smaller © 2011 by Taylor & Francis Group, LLC

cases show that the pipe moving up mobilizes the soil wedge above it extending to the surface and upheave the soil surface within a certain ranger over the pipe. This type of soil failure mechanism is called shallow failure. Figure 13a is a photo of soil deformation corresponding to the maximum soil resistance for the case of D = 50 mm and H /D = 2, which shows that black sign lines between the pipe and soil surface all curve with the pipe moving up. The influence region looks like a reverse trapezoid. The soil failure mechanism above the pipe belongs to shallow failure. With the burial depth increasing, the upheaval phenomenon of soil surface fades away. For the case of H /D > 5, the displacement on the soil surface is hardly observed and a very localized failure mechanism appears in close proximity to the moving pipe, which can be called larger failure. Figure 13b is a photo of soil deformation corresponding to the maximum soil resistance for the case of D = 30 mm and H /D = 8, which shows that movement of the pipeline just affect the region about five times of the pipeline diameter, and the shape of the affected region is like rectangular. Therefore, a larger failure occurred in soil over the pipeline. Based on the test data, H /D = 5 can be taken as transition from smaller to larger failure mechanisms for burial pipelines.

4

CONCLUSIONS

Model tests have been employed to facilitate the establishment of a substantial data base for a variety of pipeline-burial relationships. Pipe segments with diameter of 30 mm, 50 mm and 80 mm are used respectively. The pipe segments are buried in different depth-to-diameter ratios between 1 and 9. The uplift, lateral and axial resistances are recorded during the tests. Based on 81 tests, it can be concluded that: The vertical pull-out tests show that the soil resistance depends on the pipe diameter and depth of cover. The force-displacement relationships with small pipe diameter and shallow depth of cover are greatly different from those with large pipe diameter and deep depth of cover. In both cases, the soil resistance initially increases rapidly before a peak value is reached. In the former case, the soil resistance decreases to a residual level, whereas in the latter case the soil resistance remains the same level.

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The results of the lateral sliding tests show that the soil resistance initially increases before a peak value is reached and then keeps the same level. The displacement corresponding to peak resistance increases with pipe diameter. For the same depth of cover, the lateral soil resistance is more than twice that for uplift and increase with pipe diameter, which indicates that in practice, buried pipeline usually occurs vertical buckling rather than lateral buckling under thermal and internal pressure action. The results of the axial pull-out tests also show that the soil resistance initially increases before a peak value is reached and then keeps the same level. The displacement corresponding to peak resistance increases with pipe diameter. The soil resistance decreases with cyclic reversal. For the conservation, the residual axial soil resistance should be used in pipeline upheaval analyzing. According to the uplift test results, the forcedisplacement relationships with smaller depth of cover are different from those with larger depth of cover. The soil deformation also depicted the different between the pipe with smaller covered depth and larger covered depth. Based on the test data, H /D = 5 can be taken as the limit to divide the soil failure mode into shallow failure and deep failure. ACKNOWLEDGMENTS The work described in this paper was funded by China National Natural Science Foundation (No. 40776055). REFERENCES Anand S. and Agarwal, S.L., 1981, Field and laboratory studies for evaluating submarine pipeline frictional resistance,Transactions ofASME, Journal of Energy Resources Technology, 103: 50–254. Boer S., Hulsbergen C.H., Richards, D.M. et al, 1986, Buckling considerations in the design of the gravel cover for a high temperature oil line, Proc. 18th OTC, Houston,Texas, May, 5294: 1–8. Bransby M.F., Brunning P., and Newson T.A., 2001, Numerical and centrifuge modelling of the upheaval resistance of buried pipelines Proceedings of the International Conference on Offshore Mechanics and Arctic Engineering, 4(6): 265–273. Bransby M.F., Newson T.A., and Brunning P., 2002, The upheaval capacity of pipelines in jetted clay backfill, International Journal of Offshore and Polar Engineering, 12(4): 280–287.


Cheuk C.Y., Take W.A., Bolton M.D., Oliveira J.R.M.S., 2007, Soil restraint on buckling oil and gas pipelines buried in lumpy clay fill, Engineering Structures, 29: 973–982. Dickin, E. A., 1994, Uplift resistance of buried pipelines in sand,Soils and Found, 34(2): 41–48. Finch, M., 1999, Upheaval buckling and floatation of rigid pipelines: the influence of recent geotechnical research on the current state of the art, Proceedings of the Annual Offshore Technology Conference, 1(5): 27–43. Guijt, J., 1990, Upheaval buckling of offshore pipeline: overview and introduction, In proceedings of the 22nd Annual OTC, Houston, Texas, 4: 573–578. Liu R., Yan S.W. and Sun G.M., 2005, Improvement of the Method for Marine Pipeline UpheavalAnalysis under Thermal Stress£¬Journal of Tianjin University, 38(2): 124–128. Moradi, M., and Craig, W. H., 1998, Observation of upheaval buckling of buried pipelines, Centrifuge 98, Kimura, Kusakabe and Takemura(eds): 693–698. Nielsen, N.J.R., Lyngberg, B. and Pedersen, P.T., 1990, Upheaval buckling failures of insulated burial pipelinesa case story, In proceedings of the 22nd Annual OTC, Houston, Texas, 4: 581–592. Palmer A. C., White D. J., Baumgard A. J., et al. 2003, Uplift resistance of buried submarine pipelines: comparison between centrifuge modeling and full-scale tests, Geotechnique, 53(10); 877–883. Peng L.C., 1978,Stress Analysis Methods for Underground Pipe Lines: Part 2, Pipeline Industry: 65–75. Schaminee, P.E.L., Zorn, N.F. and Schotman, G.J.M., 1990, Soil response for pipeline upheaval buckling analyses: Full-scale laboratory tests and modeling. OTC 6486, 22nd annual otc, Houston, texas, may 7–10: 563–572. Schupp, J., Eacott, N., Byrne, B.W. et al, 2006, Pipeline unburial behaviour in loose sand, Proceedings of the International Conference on Offshore Mechanics and Arctic Engineering, 2006(6). Taylor, N., Richardson, D., and Gan, A.B., 1985, On submarine pipeline frictional characteristics in the presence of buckling, Proc. 4th International Symposium on Offshore Mechanics and Arctic Engineering, ASME, Dallas, Texas, February: 508–515. Taylor, N., Tran, V.C., and Richardson, D., 1989, Interface modeling for upheaval subsea pipeline buckling, In proceedings of 4th International Conference on Computational Methods and Experimental Measurements, Capri, Italy, Springer-Verlag, May: 269–282. Tran, V., 1994, Imperfect upheaval buckling of subsea pipelines,PhD Thesis, Sheffield Hallam University. White, D.J., Cheuk, C.Y. and Bolton, M.D., 2008, The uplift resistance of pipes and plate anchors buried in sand, Geotechnique, 58(12): 771–779.

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Pipe-soil interaction on clay with a variable shear strength profile D.R. Morrow & M.F. Bransby Civil Engineering Department, University of Dundee, Dundee, Angus, UK

ABSTRACT: Pipe-soil interaction is an important consideration in pipeline design for sub-sea oil and gas developments, particularly when pipeline burial is not undertaken as is commonly the case in a deep water setting. This paper presents the results of a suite of small strain finite difference analysis, undertaken to investigate the influence of shear strength gradient and a shear strength crust on pipeline penetration under vertical loading. Both the shear strength gradient and the presence of shear strength crusts were found to influence pipeline penetration resistance significantly. In particular, some geometries of shear strength crust resulted in a punchthrough mechanism significantly increasing pipeline penetration for a given vertical load, relative to an equivalent linear increasing shear strength gradient.

1

INTRODUCTION

1.1 Background Subsea pipelines fulfill a range of important functions in the development of offshore oil and gas fields. For example, infield pipelines are used to link wells and longer distance export pipelines are used to transport products ashore or to a central offshore facility. Ancillary pipelines may also be present, providing water or gas injection to the reservoir or transporting additives. Within a deepwater setting it is typical for these subsea pipelines to remain on the seabed through the course of their design life, increasing the importance of understanding pipe-soil interaction (Bruton et al., 2006; Perinet and Fraser, 2006). Pipelines in shallower water settings may also remain on the seabed if consideration of protection and other requirements indicates no need to undertake burial. Perhaps the earliest method of considering pipe-soil interaction, and the most commonly used in design, was the development of empirical methods based on the results of model testing (e.g. Brennodden et al. 1986; Verley and Lund, 1995). More recently, further work has been undertaken using this approach specifically for deep water developments, as described by Bruton et al. (2006), and utilizing centrifuge model testing (Hodder et al., 2008). Hill and Jacob (2008) also describe the migration of “model testing” to the field, or even to site specific testing. In addition to model testing based approaches, there has been progress in understanding pipe-soil interaction using analytical techniques, such as finite element analysis (Aubeny et al., 2005; Merifield et al., 2008) and upper bound plasticity calculations (Randolph and White, 2008). Useful information on failure mechanism has also been gained using image processing techniques in conjunction with centrifuge © 2011 by Taylor & Francis Group, LLC

model testing (Dingle et al., 2008). With literature largely confined to a flat seabed case some preliminary investigations into the effects of seabed slopes were undertaken by Morrow and Bransby (2009). Clay soils present in subsea development areas are likely to have increasing shear strength with depth, as a result of normal consolidation from self weight. This is discussed by Puech et al. (2005) for Gulf of Guinea soils and Yun et al (2006) for Gulf of Mexico soils. Increase in shear strength with depth may also be associated with various over-consolidation events, although these will not be specifically considered within the scope of this paper. An example of a somewhat usual shear strength profile that can be present in deepwater development areas is the presence of a near surface increased shear strength “crust” as described by Ehlers et al. (2005), Puech et al. (2005), and by Kuo and Bolton (2009). This type of shear strength profile may be significant for pipe-soil interaction and initial investigations will be presented here. Aubeny et al. (2005) undertook some investigations into the influence of shear strength gradient, comparing pipeline penetration for a uniform shear strength and a shear strength gradient with a zero strength intercept at mudline. However, it does not appear that the shear strength variation associated with crusts has been investigated previously with regards to pipeline penetration resistance. A similar design problem, albeit with some differences in geometry, is the problem of a strip footing on an undrained soil. For this problem a linearly increasing shear strength gradient has been investigated for surface foundations (Davis and Booker, 1973) and embedded foundations (Bransby and Yun, 2009). Again it appears that a shear strength crust has not been considered specifically.

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Figure 2. Representation of a shear strength crust.

Figure 1. Problem definition.

1.2

Scope of study and problem definition

The analysis and results presented in this paper are part of a wider study into pipe soil interaction on clay seabeds. The aim of this element of the study is to investigate the influence of a range of variable undrained shear strength profiles on the vertical penetration resistance of a pipeline. This paper will initially considered shear strength profiles with a linear variation in strength with depth before being extended to consider the influence of shear strength crusts. As is implicit with the use of an undrained shear term, this problem is considered at a rate of loading that provokes an undrained soil response. A pipeline of a given diameter, D, embedded in a clay seabed to depth zp is considered (Figure 1a). The seabed can be uniform strength (Figure 1b), have a linear increasing shear strength profile with a zero strength intercept at mudline (Figure 1c), or have a linearly increasing shear strength but a non-zero shear strength at mudline (Figure 1d). To increase generality of the results from the analyses it is useful to express resistance to penetration in terms of the dimensionless group R/su .D, where R is the vertical bearing capacity, su is the undrained shear strength and D is the diameter of the pipeline. When shear strength varies with depth, the shear strength at the base of the pipeline will be used, suzp at depth zp . This approach is consistent with Aubeny (2005). To quantify the different linear shear strength gradients (Figure 1b-d), the shear strength at mudline, su0 was divided by suzp . For example su0 /suzp = 0.5 represents the case where the shear strength intercept at mudline is 50% of the shear strength at the pipeline embedment depth. For this problem this provides an easier visualization of the non-uniformity of the soil strength than the more conventional used dimensionless term kD/suzp , where k is the gradient of shear strength change with depth. Note that the nonuniformity will vary with pipeline penetration and suzp © 2011 by Taylor & Francis Group, LLC

increases for all cases except that with uniform soil strength (Figure 1b). A representation of a shear strength crust for use in analysis is shown in Figure 2. This treats the crust as a departure from a given linear increasing shear strength gradient. The geometry of the crust is captured by a steep positive linear shear strength gradient extending to a peak of suct , at a depth zcp . The shear strength then returns to the underlying shear strength gradient using a negative shear strength gradient, intercepting at a depth zc . Although only a limited number of examples of crusts are considered in this paper, this approach should be relatively robust in representing a wide range of shear strength crusts. A total of 50 analyses were undertaken in this study to investigate the influence of shear strength gradient. These analyze investigated two interface conditions, perfectly smooth and perfectly rough, five “wished in place” embedment depths from 0.1D to 0.5D, and five shear strength gradients. The shear strength gradients investigated were the extremes of a constant shear strength and a gradient with an intercept at mudline of suo = 0. Intermediate shear strength gradients of su0 /suzp = 0.25, su0 /suzp = 0.5 and su0 /suzp = 0.75 were also analyzed. For the case of a shear strength crust. there is very little generality with a large range of possible cases, at least from a geometrical standpoint. In addition there is little discussion in the literature on common geometries, with only a limited number of examples reported in the public domain (e.g. Ehlers et al., 2005). This paper presents analysis for a limited number of cases with the aim of highlighting some of the issues related to shear strength crusts and pipe-soil interaction. Further work is ongoing. A total of sixteen analyses where undertaken to consider shear strength crusts. Both completely rough and smooth interface conditions were considered for one embedment depth, 0.3D. The crust was taken as being symmetrical, i.e. zcp = zc /2 (see Figure 2), and could therefore be defined in terms of just two parameters, zcp and suct . The depth to crust peak, zcp was expressed in dimensionless form, with respect to pipeline diameter, and suct was described as a multiple of the

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underlying shear strength at zcp . Two strength ratios were considered: five and ten. Four crust depths were analyzed, zcp /D = 1, 0.433, 0.3 and 0.2 was selected for the smooth interface case and zcp /D = 1, 0.5, 0.3 and 0.2 for the rough interface. 2

METHODOLOGY

The analysis presented in this paper was undertaken using the Finite Difference code FLAC (Fast Lagrangian Analysis of Continua) Version 6.0 (Itasca, 2008). Analysis was undertaken under plane strain conditions in a small strain calculation mode. Computational time was reduced by considering half a pipeline around a central axis of symmetry. Grid density was selected from grid convergence studies. Grid density requirements varied with embedment depth and the highest grid densities were required for zp = 0.1D and 0.5D. The overall boundary dimensions were selected to ensure they did not affect the calculated failure load. The width of the grid also needed to be widened significantly to accommodate a punch-through mechanism for the shear strength crust analysis. For example a 0.65 m by 0.65 m grid was required for a 0.3D embedment with D = 0.3 m. However for later analysis this was extended laterally to 1.20 m due to the larger failure mechanism associated with punch-through. The pipeline shape was formed from a section of grid above the soil surface, before being wished in place to the selected depth of embedment. The interface between the pipeline and the soil was controlled by interface elements (considering perfectly rough and smooth conditions) and the pipeline shape was given rigid behavior by application of a uniform fixed velocity (or displacement) boundary. A series of displacement controlled analysis were undertaken. The associated load-displacement behavior was calculated by summing the vertical nodal reaction forces on the pipe-soil interface. Each analysis was run until a constant load was reached representing the ultimate capacity. The seabed soil was modeled as a linear elastic perfectly plastic material with a Tresca yield criterion. Initially, a constant shear strength was assigned, but subsequent analyses were undertaken with a variation in shear strength with depth as previously described. Bulk Modulus (K) and Shear Modulus (G) were assigned based on a Poisson’s Ratio of ν = 0.49 and a Young’s Modulus of E = 200.su . On this basis, elastic stiffness parameters also varied with depth in line with changing shear strength. 3


3.1 Linear increasing shear strength Results were expressed in terms of the dimensionless group R/su .D (see Figure 3 and 4). There was reasonable agreement between the results for the constant shear strength case and previous work by Aubeny et al. © 2011 by Taylor & Francis Group, LLC

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Figure 3. R/su .D against z/D for a smooth interface condition.

Figure 4. R/su .D against z/D for a rough interface condition.

(2005) and Merifield et al. (2008). Analysis results were also normalized by the capacity for the constant shear strength case, Rconstant , which simplified interpretation. Figure 5 presents results for the case of a smooth interface condition and Figure 6 shows results for the rough interface case. The relationship between shear strength gradient and peak resistance to vertical penetration is relatively complex, with the shear

Figure 5. Change in vertical capacity for different shear strength gradients and embedments for a pipeline with a smooth interface.

Figure 6. Change in vertical capacity for different shear strength gradients and embedments for a pipeline with a rough interface.

strength gradient having a clear influence. However, this influence varies with both depth of embedment and interface condition. For the smooth interface condition, larger variations in shear strength generally result in reduced resistance to pipeline penetration for larger embedment. However, at shallower depth, resistance to penetration is generally increased. In contrast, for the rough interface condition, a larger shear strength gradient increases the resistance for all but the deepest embedments and steepest strength gradients. Additional insight into the reasons for these results can be gained by considering the failure mechanisms associated with variation of the shear strength gradient. Figures 7a to 7e show changes in failure mechanism associated with variation in shear strength gradient © 2011 by Taylor & Francis Group, LLC

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Figure 7. Displacement vector plots at failure for a smooth pipeline with zp /D = 0.3.

Figure 8. Displacement vector plots at failure for zp /D = 0.1 and smooth interface. Uniform shear strength (left) and suo /suzp = 0 (right).

Figure 9. Displacement vector plots at failure for zp /D = 0.3 and rough interface. Uniform shear strength (left) and suo /suzp = 0 (right).

for a pipeline embedded at zp = 0.3D with a smooth interface. With a smaller value of suo /suzp there is an increase in the shear strength below the base of the pipeline and a decrease above. Figure 7 shows that non-uniformity produces a change in the failure mechanism as it becomes preferential for the mechanism to migrate upwards into the lower strength material. With a smooth interface condition this effect will reduce the resistance at larger embedments (Figure 3 and 5) At shallower depths the shear strength below the base of the pipeline increase rapidly, for given gradient, in conjunction with a relatively small zone of lower shear strength material above the pipeline base. Figure 8 provides an example of the deformation mechanisms for a shallowly embedded pipeline. Relatively high shear strength material beneath the pipeline must be sheared, consistent with a greater resistance then for the constant shear strength case. For the rough interface condition, increased variation in shear strength gradient increases the resistance for all but the deepest embedments and steepest gradients (Figure 4 and 6). A rough interface mechanism (Figure 9) is larger and mobilizes soil at a greater depth then the equivalent smooth interface behavior, which is consistent with the increased failure loads (Figure 4 and 6). 3.2 Shear strength crusts The influence of shear strength crusts was investigated and comparison made with the results previously described for linear increasing shear strength gradients (Figure 10). This shows the normalized pipeline capacity with the peak strength of the crust at different depths for four conditions: (i) for rough and smooth interface conditions, and; (ii) for a crust where the shear strength © 2011 by Taylor & Francis Group, LLC

Figure 10. Normalized penetration resistance against crust depth for a pipeline with zp = 0.3D.

Figure 11. Failure mechanism showing punch-through: smooth interface; zp /D = 0.3; zcp /D = 0.43; x10 crust.

is 5 or 10 times the strength of the underlying shear strength gradient. Note that the penetration resistance is normalized by the strength at the embedment depth, zp , which varies with crust depth. When the crust peak, zcp , is relatively deep, for example 1.0D in the data shown in Figure 10, the vertical bearing capacity is identical to that in soil with a linear increasing shear strength profile extending to infinite depth (the solid lines on Figure 10). This is because the bearing capacity is dominated by the strength and gradient in the upper part of the crust. In addition, the failure mechanism was identical to those previously seen (Figure 7 and 9). The results for a crust peak depth of 0.43D and 0.5D shows a reduction in resistance. This can be attributed to a punch-through mechanism, which can be seen in Figure 11. The failure mechanism extends into the lower shear strength material below the crust resulting in a very large deformation mechanism, but a smaller bearing capacity. The analysis for shallower crust depths (e.g. zcp /D = 0.3, 0.2 in Figure 10) also give reduced resistance relative to the case of constant linear increasing gradient, with the biggest reduction occurring when

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the pipeline embedment depth coincides with the depth of the crust peak. This case in particular highlights that the failure mechanism can encompasses a large proportion of lower shear strength material than is characterized by the shear strength at the base of the pipeline. The case with zcp /D = 0.2, is perhaps the most complex. Normalized bearing capacity is slightly larger than the case with zcp /D = 0.3, albeit significantly below the resistance for zcp /D = 1. This can be attributed to a mechanism that encompasses lower strength material below zp and higher strength material above zp . The mechanism was shallower as it intercepted the increasing shear strength gradient below the crust, and this also contributes to the relative increase in resistance. As well as the crust depth being important, the strength of the crust peak compared to the underlying shear strength gradient was also critical, especially close to the onset of punch-through. When there was a larger strength difference, punch through was more marked. This is likely to relate to an increased size mechanism, including proportional weaker soil encompassed by the mechanism, associated with a stronger crust. 4

FURTHER CONSIDERATIONS

The analysis presented in this paper incorporates a number of simplifications as described in section 2. These simplifications may need to be addressed or otherwise quantified for specific pipelines or soil conditions. In addition a number of uncertainties remain with regards to the origin and properties of shear strength crusts incuding time dependent variation. A limited number of examples have been presented in this paper and work is currently ongoing to consider a wider range of crusts and produce generalized design guidance. 5

CONCLUSIONS

This paper reports the results of a study investigating the influence of clays with variable shear strength profiles on pipe-soil interaction. Various linear increasing profiles were considered, both as an area of interest in itself and to provide background material for subsequent consideration of shear strength crusts. In addition, results have been shown for analyses with shear strength crusts of different strength relative to the underlying strength gradient, variation of the depth to crust peak, and variation in the interface conditions. The presence of a shear strength gradient was seen to influence vertical bearing capacity, particularly for the rough interface conditions. However, the presence of a shear strength crust had the most dramatic effect. When the crust was deep the behavior was dominated by the upper shear strength gradient, but as the crust peak got closer to the base of the pipeline a punchthrough mechanism was observed. © 2011 by Taylor & Francis Group, LLC

A number of examples have been presented in this paper, where the bearing capacity cannot be calculated accurately using a single shear strength without consideration of either shear strength gradient and/or the geometry of the shear strength crust. REFERENCES Aubeny, C., Shi.,H & Murff, J. 2005. Collapse load for cylinder embedded in trench in cohesive soil. International Journal of Geomechanics 5(4): 320–325. Bransby, M.F. & Yun, G.J. 2009. The undrained capacity of skirted strip foundations under combined loading. Geotechnique 59(2): 115–125. Brennodden, H., Sveggen, O., Wagner, D.A. & Murff, J.D. 1986. Full scale pipe-soil interaction tests. Proceedings of OTC 1986. OTC Paper No. 5338. Bruton, D., White, D., Check,C., Bolton, M. & Carr, M. 2006. Pipe soil interaction behavior during lateral buckling including large amplitude cyclic displacement tests by the SAFEBUCK JIP. Proceedings of OTC 2006. OTC Paper No. 17944. Davis, E.H. & Booker, J.R. 1973. The effect of increasing shear strength with depth on the bearing capacity of clays. Geotechnique 23(4): 551–563. Dingle, H.R.C., White, D.J. & Gaudin, C. 2008. Mechanisms of pipe embedment and lateral breakout on soft clay. Canadian Geotechnical Journal, 45(5): 636–652. Ehlers, C.J., Chen, J., Roberts, H.H. & Lee, Y.C. 2005. The origin of near-seafloor “crust zones” in deepwater. Proceedings of ISFOG 2005: 927–933. Hill, A.J. & Jacob, H. 2008. In-situ measurement of pipe-soil interaction in deep water. Proceedings of OTC 2008. OTC Paper No. 19528. Hodder, M.S., Cassidy, M.J. & Barret, D. 2008. Undrained response of shallow pipelines subjected to combined loading. Proceedings of ICOF 2008: 897–908. Itasca Consulting Group. 2008. FLAC – Fast Lagrangian Analysis of Continua – Users Guide. Kuo, M.Y.H. & Bolton, M.D. 2009. Soil characterization of deep sea west African clays: is biology a source of mechanical strength. Proceedings of ISOPE 2009: 488–494. Merifield, R., White, D.J. & Randolph, M.F. 2008. The ultimate undrained resistance of partially embedded pipelines. Geotechnique 58(6): 461–470 Morrow, D.R. & Bransby, M.F. 2009. The influence of slope on the stability of pipelines subjected to horizontal and vertical loading on clay seabeds. Proceedings of OMAE 2009. Perinet, D. & Fraser, I. 2006. Mitigation methods for deepwater pipeline instability induced by pressure and temperature variation. Proceedings of OTC 2006. OTC Paper No. 17815 Puech, A, Colliat, J.L., Nauroy,J-F. & Menier,J. 2005. Some geotechnical specificities of Gulf of Guinea deepwater sediments. Proceedings of ISFOG 2005: 1047–1053. Randolph, M.F. & White, D.J. 2008. Upper bound yield envelopes for pipelines at shallow embedment depths in clay. Geotechnique 58(4): 213–229. Verley, R.L.P. & Lund, K.M. 1995 A soils resistance model for pipeline place on clay soils. Proceedings of OMAE 95, 225–232. Yun, T.S., Narsilio, G.A. & Santamarina, J.C. 2006. Physical characterization of core samples recovered from Gulf of Mexico. Marine and Petroleum Geology 23: 893–900.

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Sweeping behaviour of shallowly-embedded pipeline during cyclic lateral movement T. Takatani Department of Civil Engineering, Maizuru Nat’l College of Technology, Maizuru, Kyoto, Japan

ABSTRACT: A non-linear finite element analysis based on an effective stress theory for pipeline-seabed interaction problem was carried out in order to simulate the sweeping behaviour of a shallowly-embedded pipeline on carbonate sandy soil subjected to cyclic lateral movement under a constant vertical load. Pipeline sweeping behaviour during cyclic lateral movement was numerically investigated in terms of the initial depth of pipeline, the carbonate soil conditions, the amplitude and frequency of cyclic lateral movement under constant vertical loading. The large-amplitude and cyclic pipe-soil interaction is discussed. Pipeline sweeping behaviour greatly depends on the amplitude and frequency of cyclic lateral movement.

1

INTRODUCTION

Shallowly-embedded offshore pipelines are directly exposed to the vertical and horizontal forces induced by the hydrodynamic environment. The cyclic movement of pipeline due to both drag and lift forces caused by waves and currents will lead to a large deformation of pipeline. On the other hand, the seabed soil berms created at the extremities of the cyclic lateral sweeping range and the remoulding of the seabed soil generated by the cyclic lateral movement of a pipeline have recently investigated in the offshore pipeline engineering field. Recently, Dingle et al. (2008) observed the deformation mechanism during cyclic lateral movement of pipeline through some centrifuge tests, and the in-flight images which indicate pipeline breakout and large amplitude sweeping were evaluated using Particle Image Velocimetry (PIV) analysis. White and Cheuk (2008) proposed a simplified modelling of cyclic lateral pipe-soil interaction, based on the accumulation and deposition of berm materials. Hodder et al. (2008) conducted a series of centrifuge model tests to evaluate a riser-soil interaction within the touchdown zone of a steel catenary riser during pipeline laying process. Pipeline sweeping behaviour due to cyclic lateral movement strongly depends on a pipe-soil interaction. It is, therefore, very important to investigate the largeamplitude and cyclic pipe-soil interaction behaviour using experimental tests and numerical analyses in order to accurately evaluate the pipeline sweeping behaviour during cyclic lateral movement. The purpose of this paper is to investigate the pipeline sweeping behaviour during cyclic lateral movement, focusing on the large-amplitude and cyclic pipe-soil interaction behaviour during severe storm © 2011 by Taylor & Francis Group, LLC

condition. In this paper, a two-dimensional non-linear finite element analysis based on an effective stress theory is employed for a pipe-soil interaction problem. 2

CYCLIC LATERAL PIPELINE-SEABED INTERACTION ANALYSIS

An advanced numerical analysis in this paper is two-dimensional dynamic non-linear finite element method (Takatani et al., 2005, 2008) based on the effective stress theory in order to simulate a sweeping behaviour of shallowly-embedded pipeline on carbonate sandy soil under undrained condition. In this finite element analysis, a non-linear relationship between shear stress and shear strain of soil element is accurately expressed by a multi shear spring model (Towhata and Ishihara, 1985) and the Masing rule for loading and unloading curves is employed so as to adjust the amplitude of hysteresis damping for the multi shear spring model. Also the cyclic mobility model (Iai et al., 1990), which is of a generalized plasticity-multiple mechanism type, is adapted to simulate excess pore water pressure. Pore fluid is assumed to be imcompressible, and also the viscous boundary technique (Lysmer et al., 1969) is used to create the infinite of seabed soil in this analysis. There are three governing equations of a kinematic equation between soil and pipeline, pore water input/output balance equation in each pore fluid element, and dynamic water pressure wave propagating equation for pore fluid. Pore water pressure can be expressed by an increment of volumetric strain of soil skeleton because of undrained condition, and also dynamic water pressure wave propagating equation for pore fluid can be represented by a technique that the effect of pore fluid

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Figure 1. Fem mesh for pipeline-seabed interaction analysis (z0 = 0.1 m.)

existence can be taken into consideration by applying an additional mass of each pore fluid element to the soil-structure kinematic equation. In this paper, the sweeping behaviour of a shallowlyembedded pipeline resting on carbonate sandy soil is investigated from a view point of carbonate sandy soil characteristics. Pipe is assumed to be 1.0 m diameter and both 0.1 m and 0.25 m for its initial depth, and the seabed is assumed to be a carbonate soil. Pipeline sweeping behaviour during cyclic lateral movement is numerically investigated in terms of the stiffness of joint element between pipe and seabed soil, the initial depth of pipeline, the carbonate soil conditions, the amplitude and frequency of cyclic lateral movement under a constant vertical loading. Figure 1 shows a finite element mesh for an unburied offshore pipeline-seabed interaction analysis considering a liquefaction phenomenon in the seabed around the pipeline. The joint element is used at contact area between pipeline and sand layer to represent a slip phenomenon at contact area between pipeline and seabed surface (Takatani et al., 2005). The initial depth of pipeline, z0 , is 0.1 m as shown in Figure 1.The model of initial depth of pipeline, z0 = 0.25 m, is described in Figure 1, too. The analytical domain is 7 m × 10 m and is assumed to be a carbonate sand layer. Both 6node triangle and 8-node square elements are used in this mesh, and also the Selective Reduced Integration method (Hughes, 1980) by which each soil element integration can be separately evaluated for both the volumetric and deviation components is employed in order to make an accurate evaluation for each soil element integration. At every incremental time step, the coordinate of each nodal point is renewed according to the soil deformation, and the stress loading of each element is re-evaluated by a self-weighted analysis result at every time step. This finite element analysis with a coarse mesh in the vicinity of the pipeline shown in


Table 1.

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Material properties of carbonate sand (Dr = 60%).

Initial shear modulus, Gma (kPa) Elastic tangent bulk modulus of soil skeleton, Kma (kPa) Friction angle, φf (degree) Phase transformation line angle, φp (degree) Material parameters for dilatancy S1 w1 c1 p1 p2

48,395 126,207 38.16 28 0.005 4.634 1.548 0.500 1.037

Figure 1 can be carried out with sufficient accuracy (Ozutsumi, 2003). The mechanical properties for carbonate soils with the relative densities Dr = 60% and 80% are evaluated from the liquefaction resistance curves, that is, the effective stress ratio vs. the number of cycles. The material properties of carbonate soil for the relative density Dr = 60% is shown in Table 1, and also are obtained from the liquefaction resistance curve for carbonate sand indicated in Figure 2(a) (Aramaki, 1997). Figure 2(b) shows the liquefaction resistance curve for carbonate sand with the relative density Dr = 80%. As a reference, the liquefaction resistance curves for Toyoura sand are indicated, too. Five parameters shown in Table 1 which specify the dilatancy are determined by the back-fitting technique to the liquefaction resistance curves of the carbonate sand obtained from the laboratory test. In general, the liquefaction resistance curve is determined by combining the laboratory test data and the bearing capacity test data at the sites for taking the in-situ conditions of soils into account (Morita et al., 1997). Before the cyclic lateral movement of pipeline, the self-weighted analysis for pipeline-seabed interaction

Figure 2. Liquefaction resistance curve for carbonate sand (Aramaki, 1997).

problem is carried out under the completely drained condition to obtain the initial effective stress of each soil element. In this numerical pipeline-seabed interaction analysis, a strain space plasticity approach is assumed to be used for cyclic mobility in order to represent the realistic hysteretic damping factor under cyclic loading. In this approach, actual cyclic shear mechanism is decomposed into a set of one dimensional virtual simple shear mechanism. Material properties of dilatancy S1 , w1 , c1 , p1 and p1 shown in Table 1 are five parameters to define the cumulative volumetric strain of plastic nature for representing cyclic mobility. These parameters define the correlation between the liquefaction front parameter (Iai et al., 1990) and the normalized shear work. The liquefaction front parameter is given by a function of shear work, and Towhata and Ishihara (1985) obtained the correlation between the shear work and the excess pore pressure, and also concluded that the correlation is independent of the shear stress paths with or without the rotation of principal stress axes. 3

NUMERICAL RESULTS

In the sweeping behaviour analysis of a shallowlyembedded pipeline subjected to cyclic lateral movement, the vertical load, V , is assumed to be maintained a constant value during cyclic lateral movement, H .


Before cyclic lateral movement, the static analysis subjected to a constant vertical load, V , is carried out, and then both a constant vertical load, V , and cyclic lateral movement, H , operate at the centre of pipeline as shown in Figure 1. Figure 3 shows seabed deformation around a pipeline after 1,000 cyclic lateral movements for two initial depths of pipeline z0 = 0.1 m and 0.25 m in the relative density Dr = 80%, the frequency f = 0.5 Hz of cyclic lateral movement and the constant vertical load V = 4 kN/m. The unit tangential stiffness for normal and shear directions, Kn and Ks , f or a joint element are used 1.0 × 106 (kN/m) and 1.0 × 105 (kN/m), respectively. The friction angle of joint element is assumed to be 25 degree in this analysis. It can be observed from these figures that the settlement of pipeline increases with the number of cycles, and that greatly depends on the amplitude of cyclic lateral movement, H . The larger the amplitude of cyclic lateral movement becomes, the more widely and deeply the seabed surface around a pipeline will be excavated by the cyclic lateral movement of pipeline. Pipeline settlement after 1,000 cyclic lateral movements in the initial depth of pipeline, z0 = 0.1 m, is much larger than that in z0 = 0.25 m for each amplitude of lateral movement, H . This is because that the soil resistance force to pipeline increases with the initial depth of pipeline, z0 . Although the seabed deformation after 1,000 cyclic lateral movements in the relative density Dr = 60% for each amplitude of cyclic lateral movement, H , is not illustrated in this paper on account of space consideration, pipeline settlement slightly increases with the decrease of the relative density, Dr. Figure 4 shows pipeline sweeping behaviour during cyclic lateral movement in Dr = 80%, H = ±0.05 m and V = 4 kN/m. Pipeline settlement behaviour in the frequency f = 2.0 Hz of cyclic lateral movement is indicated in Figure 4(a). It can be seen from this figure that the large settlement in pipeline sweeping behaviour occurs within several cyclic lateral movements at initial cyclic stage and then gradually approaches a certain value with the number of cycles. Figure 4(b) illustrates pipeline settlement behaviour during 1,000 cyclic lateral movements in the frequencies f = 0.5, 1.0 and 2.0 Hz. Pipeline settlement increases with the frequency, f , of cyclic lateral movement, and also each settlement behaviour seems to increase with the number of cycles, N , after 1,000 cycles. It should be noted that pipeline settlement greatly depends on the frequency, f , of cyclic lateral movement, H . Table 2 indicates pipeline settlement after 1,000 cyclic lateral movements for each amplitude of cyclic lateral movement, H , in the relative density Dr = 60%, the initial depths of pipeline, z0 = 0.1 m and 0.25 m, the cyclic frequencies f = 0.5, 1.0 and 2.0 Hz. It can be seen from this table that pipeline settlement has a tendency to increase with the amplitude, H , and the cyclic frequency, f , of cyclic lateral movement. On the other hand, Table 3 illustrates pipeline settlement results after 1,000 cyclic lateral movements

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Figure 3. Seabed deformation around a pipeline behaviour after1,000 cyclic lateral movements (Dr = 80%, f = 0.5 Hz, V = 4 kN/m).


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Figure 4. Pipeline behaviour during cyclic lateral movement (Dr = 80%, H = ±0.05m, V = 4 kN/m). Table 2. Pipeline settlement after 1,000 cyclic lateral movements (Dr = 60%, V = 4 kN/m)

Table 4. Pipeline settlement after 1,000 cyclic lateral movements (f = 2.0 Hz, H =±0.1 m)

Settlement (m) after 1,000 cycles f = 1.0 Hz

f = 2.0 Hz

(a) z0 = 0.1 m ±0.05 −1.292 ±0.1 −1.859 ±0.2 −2.784 ±0.5 −4.761

−1.346 −1.974 −2.928 −5.548

−0.997 −1.756 −3.238 −5.839

(b) z0 = 0.25 m ±0.05 −0.697 ±0.1 −0.875 ±0.2 −1.247 ±0.5 −2.571

−0.787 −1.114 −1.882 −3.641

−0.892 −1.518 −2.419 −5.415

H (m)

f = 0.5 Hz

Table 3. Pipeline settlement after 1,000 cyclic lateral movements (Dr = 80%, V = 4 kN/m) Settlement (m) after 1,000 cycles f = 1.0 Hz

f = 2.0 Hz

(a) z0 = 0.1 m ±0.05 −1.054 ±0.1 −1.244 ±0.2 −1.954 ±0.5 −3.620

−0.988 −1.382 −2.221 −4.137

−1.032 −1.549 −2.488 −4.137

(b) z0 = 0.25 m ±0.05 −0.426 ±0.1 −0.615 ±0.2 −1.018 ±0.5 −1.916

−0.514 −0.830 −1.408 −2.872

−0.678 −1.148 −2.076 −4.364

H (m)

f = 0.5 Hz

for each amplitude of lateral movement, H , in the relative density Dr = 80%. It can be observed from Tables 2 and 3 that pipeline settlement increases with the decrease of the relative density, Dr. This is because that the soil resistance force to pipeline increases with the relative density, Dr. © 2011 by Taylor & Francis Group, LLC

V (kN/m)

Dr = 60%

Dr = 80%

(a) z0 = 0.1 m 0.5 1.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0

−0.844 −1.071 −1.223 −1.386 −1.756 −1.787 −1.800 −2.088 −2.273

−0.996 −1.161 −1.174 −1.309 −1.549 −1.817 −1.843 −1.891 −2.138

(b) z0 = 0.25 m 0.5 1.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0

−0.659 −0.804 −0.988 −1.172 −1.518 −1.846 −1.657 −1.800 −2.000

−0.539 −0.623 −0.993 −0.979 −1.148 −1.319 −1.351 −1.521 −1.564

Table 4 shows the effect of constant vertical load, V , on the pipeline settlement in the initial pipeline depths, z0 = 0.1 m and 0.25 m. It can be seen from this table that pipeline settlement increases with the constant vertical load, V . As the soil resistance force to pipeline increases with the initial depth of pipeline, pipeline settlement may trend to decrease if the initial pipeline depth, z0 , is large and the constant vertical load, V , is small. Table 5 indicates the effect of joint element stiffness on the pipeline settlement in the initial pipeline depths, z0 = 0.1 m and 0.25 m, the relative densities, Dr = 60% and 80%. It can be clearly found from this table that the settlement of pipeline increases with the stiffness of joint element. This is because the pipeline movement is directly transmitted to the seabed soil around pipeline with the increase of joint element stiffness. Although this effect of joint element stiffness is not

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Table 5. Pipeline settlement after 1,000 cyclic lateral movements (H = ±0.1 m, f = 2.0 Hz, V = 4 kN/m). Dr = 60%

Dr = 80%

Kn , Ks (kN/m)

z0 = 0.1 m z0 = 0.25 m z0 = 0.1 m z0 = 0.25 m

108 ,107 106 ,105 104 ,103 103 ,102 102 ,101

−1.756 −1.751 −1.513 −1.273 −0.547

−1.583 −1.518 −1.209 −1.134 −0.604

−1.537 −1.249 −1.053 −0.851 −0.392

−1.202 −1.148 −0.858 −0.670 −0.383

REFERENCES

mentioned in detail in this paper on account of space consideration, Takatani (2005) discussed the effect of joint element stiffness on the pipeline behaviour during cyclic loading. 4

CONCLUSIONS

The advanced finite element analysis based on an effective stress theory was conducted to simulate the sweeping behaviour of a shallowly-embedded pipeline on carbonate sandy soil during cyclic lateral movement. In summary, the following conclusions can be made based on the results presented in this paper. 1. Pipeline settlement greatly depends on the amplitude, H , and frequency, f , of cyclic lateral movement, the initial pipeline depth, z0 , the vertical load, V , and the stiffness, Kn and Ks , of joint element between pipeline and seabed surface. 2. Because the soil resistance force to pipeline increases with the initial depth of pipeline, the settlement of pipeline after 1,000 cyclic lateral movements in the initial depth of pipeline,z0 = 0.1 m, is much larger than that in z0 = 0.25 m for each amplitude of cyclic lateral movement. 3. Pipeline settlement after 1,000 cyclic lateral movements in the relative density Dr = 80% is smaller than that in Dr = 60%, because the soil resistance force to pipeline increases with the relative density. In this paper, the sweeping behaviour of a shallowlyembedded pipeline on carbonate sandy soil was investigated under undrained condition. In future, there seems to be a need for further consideration on this analytical condition in the pipeline-seabed interaction problem. Although the pore pressure accumulation in


the seabed around a pipeline is not presented in this paper due to the limited space, it is necessary for an intensive study on the effect of pore pressure accumulation on the pipeline sweeping behaviour because it is the most important role in the design of offshore pipeline. In addition, further investigation may be needed to simulate these phenomena mentioned above and make some concrete conclusions.

Aramaki, N. 1997. Undrained cyclic and monotonic triaxial behaviour of crushable carbonate soil, Dr. Eng. Thesis, Yamaguchi University. Cheuk, C.Y., White, D.J. and Bolton, M.D. 2008. Uplift mechanism of pipes buried in sand, J Geotech & Geoenvironmental Eng, Vol.134, No.2, pp.154–163. Dingle, H.R.C., White, D.J. and Gaudin, C. 2008. Mechanism of pipe embedment and lateral breakout on soft clay, Canadian Geotech J, Vol.45, No.5, pp.636–652. Hodder, M.S., White, D.J. and Cassidy, M.J. 2008. Centrifuge modeling of riser-soil stiffness degradation in the touchdown zone of a steel catenary riser, Proc 27th Int Conf Offshore Mech Arctic Eng (OMAE), ASME, Estoril, Portugal, CD, OMAE2008-57302. Hughes, T.J.R. 1980. Generalization of selective integration procedures to anisotropic and nonlinear media, Int J Num Meth Eng, Vol.15, pp.1413–1418. Iai, S., Matsunaga, Y. and Kameoka, T. 1990. Strain space plasticity model for cyclic mobility, Report of Port and Harbour Research Institute, Vol.29, No.4, pp.27–56. Lysmer, J. and Kuhlemeyer, R.L. 1969. Finite dynamic model for infinite media,J Eng Mech Div, ASCE, No.EM4, pp.859–877. Morita,T, Iai, S, Liu, H, Ichii, K, and Sato,Y. 1997. Simplified Method to Determine Parameter of FLIP, Material of Port and Harbour Research Institute, No 869. Ozutsumi, O. 2003. Numerical analysis on seismic damage estimation for soil-structure system on liquefied area, Dr. Eng. Thesis, Kyoto University. Takatani, T. 2005. Pipeline-seabed interaction analysis subjected to horizontal cyclic loading, Proc Int Symp Frontiers Offshore Geomech, Perth, pp.629–635. Takatani, T. and Kaya, T. 2008. Unburied offshore pipeline stability analysis based on non-linear relationship between pipeline and carbonate soil, Proc 18th Int Offshore Polar Eng Conf, Vancouver, Canada, Vol.2, pp.176–185. Towhata, I. and Ishihara, K. 1985. Modelling soil behaviour under principal stress axes rotation, Proc 5th Int Conf Num Method Geomech., Nagoya, pp.523–530. White, D.J. and Cheuk, C.Y. 2008. Modelling the soil resistance on seabed pipelines during large cycles of lateral movement,Marine Structures, Vol.21, No.1, pp.59–79.

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Advanced nonlinear hysteretic seabed model for dynamic fatigue analysis of steel catenary risers I.H.Y. Ting, M. Kimiaei & M.F. Randolph Centre for Offshore Foundation Systems, The University of Western Australia, Perth

ABSTRACT: Fatigue of steel catenary risers (SCRs) in the touchdown zone (TDZ) remains one of the greatest challenges in designing SCRs. This is because of the nonlinear pipe-soil interaction in the TDZ and the random cyclic motions to which the SCRs are subjected. The soil parameters used to model the interaction can also significantly affect the TDZ fatigue response of SCRs. Most traditional fatigue design approaches for SCRs are based on an assumed linear stiffness for the seabed and tend to provide very conservative results, particularly in deep water. The response of SCRs as a result of hysteretic nonlinear pipe-soil interaction is difficult to predict and improved understanding of this interaction is therefore vital for more accurate fatigue design. In this paper, the results of a parametric study are presented, using a nonlinear pipe-soil interaction model to investigate how the main geotechnical parameters influence the fatigue life of SCRs. The main parameters considered are: maximum normalised stiffness, soil suction ratio and shear strength gradient. The paper summarises the relative effects of these parameters on the fatigue life, concluding that the shear strength gradient and soil suction ratio and are the most critical parameters affecting the SCR fatigue life. 1

INTRODUCTION

As the offshore industry continues to progress developments in deepwater fields, steel catenary risers (SCRs) are often the preferred riser option for subsea tieback to floating platforms. This is due to their conceptual simplicity, ease of construction and installation and simple interface with the flowlines. Modelling of pipesoil interaction in the touchdown zone (TDZ), where the riser meets the seabed, is one of the critical challenges for SCRs, since it has a significant effect on the fatigue life in that zone. The interaction exhibits complex behaviour and is highly nonlinear in response to the random cyclic motions to which the SCR is subjected (Grealish et al. 2007). The pipe-soil interaction in the TDZ is critical for accurate estimation of the fatigue life. Traditionally fatigue assessment of SCRs is carried out using linear soil springs, with or without damping, despite the awareness of the nonlinear interaction in the TDZ (Clukey et al. 2004). Linear pipe-soil interaction models ignore or simplify much of the fundamental response of soil, such as variation of the secant stiffness depending on the amplitude of cyclic displacement, suction during uplift and softening of soil under cycling motions. In spite of this, linear springs are adopted partly because of previous software limitations and also partly because linear solutions greatly simplify fatigue study. The limited understanding of the actual nonlinear interaction has also restricted development of appropriate pipe-soil interaction models, although a number of models have been proposed recently © 2011 by Taylor & Francis Group, LLC

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(Aubeny & Biscontin 2009, Randolph & Quiggin 2009). It is important, however, not to restrict the modelling of such effects to linear approximation, in particular for ultra deepwater developments, as it can lead to a conservative design approach (Clukey et al. 2007, Grealish et al. 2007). It is therefore important to model the interaction accurately to minimise conservatism in fatigue response predictions for SCRs. Significant research programmes, laboratory testing and field-scale experiments such as the STRIDE and CARISIMA JIPs (Bridge et al. 2004) have helped to understand the interaction and provide a basis for nonlinear soil models. The soil parameters used for SCR analysis can have a significant effect on riser response especially on the predicted fatigue life (Bridge et al. 2004). It is therefore important to study the sensitivity of fatigue results to soil input parameters. The soil model presented by Randolph & Quiggin (2009) has been used in this study. This soil model is based on a hyperbolic secant stiffness formulation such as proposed by Bridge et al. (2004) with asymptotic limiting penetration and uplift resistance. It is able to capture the variation of stiffness with the amplitude of cyclic displacements. The results of a parametric study of fatigue damage based on this nonlinear pipe-soil interaction model are presented here. A series of dynamic riser response analyses were carried out to investigate how the main geotechnical parameters influence the fatigue life of SCRs in the TDZ. The main parameters considered are: maximum normalized stiffness, soil suction ratio and shear strength gradient. The paper summarises how these

input parameters influence the fatigue results of an example deepwater SCR.

The key soil input parameters, and the manner in which they affect the response, are described below.

2

2.1

PIPE-SOIL INTERACTION MODEL

The hysteretic nonlinear pipe-soil interaction model presented by Randolph and Quiggin (2009) is used in this study. As shown in Fig. 1, in general there are four different penetration modes in this model: not in contact, initial penetration, uplift and repenetration. The primary input data into the model are the pipe diameter, seabed soil shear strength profile (assuming a linear strength profile with mudline intercept sum and gradient ρ) and soil density. Additional parameters include: normalised maximum stiffness of the pipe-soil response, suction resistance ratio (relative to penetration resistance at the given embedment), normalised suction decay distance and normalised repenetration offset (which controls the incremental additional embedment with each cycle). These various input parameters are used to define the non-linear hyperbolic functions that model the seabed resistance force as a function of the penetration history as detailed by Randolph and Quiggin (2009). Function parameters are updated each time a penetration reversal occurs allowing the model to capture the hysteretic behaviour of the seabed response and the increasing penetration of the pipe under vertical cyclic loading. No attempt is made to model softening of soil due to remoulding directly, although incremental embedment occurs during load controlled cycles. The model is drawn from a number of research programmes such as reported by Bridge et al. (2004) and has been calibrated against model tests reported by Aubeny et al. (2008). It has also been calibrated against field-scale experiments carried out at Watchet harbour, UK.

Soil shear strength gradient, ρ

The pipe-soil response is closely linked to the shear strength profile, which is assumed to vary linearly with depth in the present implementation of the model. For deepwater applications, it is reasonable to assume that the strength intercept at the mudline, sum , is zero, since any small positive value would rapidly be eliminated due to remoulding. The strength profile is therefore controlled entirely by the rate of change of shear strength with depth (ρ). For foundation and anchor design, strength gradients are typically in the range 1.5 to 2 kPa/m over depths in the 1 to 50 m range (and even deeper). However, many deepwater sediments show shear strengths of 5 to 15 kPa at a depth of about 0.5 m, and hence high strength gradients over that depth range, before the strength gradient reverts to the more typical 1.5 to 2 kPa/m. As such, shear strength gradients for SCR design may range typically from 1.5 to 30 kPa/m. 2.2

Normalised maximum stiffness, Kmax

The normalised maximum stiffness, Kmax , controls the stiffness of the pipe-soil response during initial penetration mode, uplift mode and repenetration mode. It also controls how fast the resistance asymptotically approaches its limiting value. A higher value leads to greater stiffness and translates to increase in fatigue damage. The soil stiffness is given by:

where Pu is the ultimate penetration resistance (force per unit length of riser), z is the penetration of the riser invert below the seabed and D is the riser diameter. The ultimate resistance (net of the buoyancy correction) is expressed as

where Nc is a bearing capacity factor (expressed as a power law function of normalised embedment as Nc = a(z/D)b and su,invert is the shear strength at the current depth of the riser invert. The hyperbolic response linking the initial stiff gradient to the ultimate limiting resistance is expressed as

where Figure 1. Different penetration modes in nonlinear pipe-soil interaction model by Randolph & Quiggin (2009). © 2011 by Taylor & Francis Group, LLC

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It may be seen that substitution of Eq. 4 into Eq. 3 leads to Eq. 1 as z reduces towards zero, and Eq. 2 as z becomes large. Similar approaches, using changes in penetration, are used for reversals of displacement from penetration to uplift and vice versa. Typical values of normalised maximum stiffness for soft sediments have been shown to lie in range 150 to 250 (Bridge et al. 2004, Clukey et al. 2005).

2.3 Suction ratio, fsuc The suction ratio controls the ultimate suction (or uplift) resistance of the model, which is given by:

It should be noted that in reality the suction resistance will depend on a variety of factors, in particular the rate at which the pipe is lifted up, the length of time over which upward motion is sustained and the recent history of cyclic motion (Bridge et al. 2004); use of a constant fsuc in Eq. 5 is therefore a simplification, reflecting the very limited experimental data currently available. For single uplift motions of a pipe that has been undisturbed for a period, values of fsuc between 0.5 to 1 may be appropriate. However, for fatigue studies or other applications with many cycles of loading, values in the range 0 to 0.3 are recommended (Randolph & Quiggin 2009).

3

Figure 2. General view of the case study in this paper. Table 1.

CASE STUDY

The example SCR configuration used for the parametric studies in this paper is the same configuration adopted by Kimiaei et al. (2010). A general view of the configuration is shown in Figure 2. It comprises a 9 inch (0.228 m) diameter SCR, with 1 inch (25.4 mm) wall thickness, submerged weight of 1.01 kN/m and bending stiffness, EI, of 17.7 MNm2 . The riser is 1600 m long and is hung in 1000 m water depth from the pontoon of a semi-submersible platform. The mean departure angle of the riser is 10◦ as measured from the downward vertical. A flexjoint with stiffness of 10 kNm/deg is incorporated at the connection of the riser to the platform. The riser touchdown point (TDP) is approximately 1170 m arc length measured from the top connection while the TDZ during dynamic motions varies between 1130 m and 1200 m arc length. The soil properties adopted for this study are characteristic of typical soft clay offshore sediments such as in the Gulf of Mexico. The default set of soil model parameters, referred to as ‘base case’ here, are presented in Table 1. In order to investigate the sensitivity of the results to key input parameters, nine different load cases were considered in addition to the base case; in each case only one of the soil parameters (Kmax , fsuc or ρ) were changed, relative to the values for the base case, as presented in Table 2. © 2011 by Taylor & Francis Group, LLC

Soil parameters for the base case.

Parameter

Symbol

Value

Pipe diameter Mudline shear strength Shear strength gradient Saturated soil density Power law parameter Power law parameter Soil buoyancy factor Normalized maximum stiffness Suction ratio Suction decay parameter Repenetration parameter

D sum ρ ρsoil a b fb Kmax fsuc λsuc λrep

0.228 m 0 kPa 1.5 kPa/m 1.5 te/m3 6 0.25 1.5 200 0.2 1 0.3

Table 2.

Load cases used for the parametric studies.

Load case

Symbol

Value

1 2 3 4 5 6 7 8 9

Kmax Kmax fsuc fsuc fsuc ρ ρ ρ ρ

100 300 0 0.4 0.6 3 kPa/m 5 kPa/m 10 kPa/m 20 kPa/m

The analysis software OrcaFlex (Orcina 2009) was used for dynamic time domain response analysis of the SCR. The nonlinear pipe-soil interaction model described previously has been incorporated in OrcaFlex (Randolph & Quiggin 2009). For fatigue damage calculations, deterministic regular fatigue

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analysis, together with a Palmgren-Miner approach, was used, since this is one of the most widely accepted methods for fatigue design of offshore facilities. In this method, environmental loads on the system were represented by a wave scatter table comprising a number of regular wave packets (giving wave heights, wave periods and total number of wave occurrences). The cyclic stress ranges in the SCR (as relevant for the fatigue analysis in the TDZ) were obtained from the numerical simulations and then, using the standard stress-cycle (S-N) approach, the corresponding fatigue damage for each wave packet was determined. The overall fatigue damage of the system is then obtained from the summation of fatigue damage due to each individual wave packets. The example wave scatter table used in this study, broadly representative of wave data in deepwater Gulf of Mexico, is presented inTable 3. In this study, loading time histories for the riser dynamic response analyses comprised a series of 15 consecutive wave packets selected from Table 3, but with the order varies as detailed later. Note that the wave packets tabulated represent an original set of 30 storms with irregular wave scatter table (covering total exposure time of 20 years). In order to reduce the computational efforts only a 15 wave packet regular wave scatter table was adopted. The main objective here is to provide a relative assessment of how the fatigue life is affected by different orders of waves and variations in key soil properties, rather than obtain absolute estimates of fatigue life. Results of riser fatigue response analyses, which are influenced by nonlinear pipe-soil interaction behaviour, are sensitive to the order of the wave packets in each loading time history (Kimiaei et al. 2010). In this paper, results of only two following sample loading sequences (LS) have been studied:

Table 3. Wave scatter table. Wave packet no.

Wave height (m)

Wave Period (s)

Number of annual wave occurrences

1 2 3 4 6 7 8 10 11 12 14 15 16 18 19

1 1 1 1 3 3 3 8 8 8 13 13 13 18 18

3 8 13 18 8 13 18 8 13 18 8 13 18 13 18

24603799 4584144 770892 288213 1170825 88515 15607 28572 5042 572 868 440 44 63 4

Figure 3. Fatigue damage profiles for different normalised maximum stiffness.

LS1: Loading time history comprising wave packets 4, 8, 12, 16, 19, 3, 7, 11, 15, 18, 2, 6, 10, 14 and 1, sequentially. LS2: Loading time history for each wave packet (1, 2, 3, 4, 6, 7, 8, 10, 11, 12, 14, 15, 16, 18 and 19) separately. Measure fatigue damage for each wave packet in these separate analyses. In each case, the fatigue damage for each wave packet in the loading time history was determined and the total fatigue damage for the system then determined by summation. Note that only sufficient wave cycles were analysed for each wave packet to reach a steady state response (generally 20 cycles) and then the total fatigue damage per year obtained using the stabilised cyclic stress range obtained in the final cycles. 4 4.1

NUMERICAL RESULTS Fatigue results for LS1

Results of 20-year life fatigue damage for LS1 due to different normalized maximum stiffness (Kmax ), suction ratio (fsuc ) and soil shear strength gradient (ρ) over © 2011 by Taylor & Francis Group, LLC

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Figure 4. Fatigue damage profiles for different suction ratios.

the touch down zone, are shown in Figures 3, 4 and 5, respectively. The figures show consistency of the results in respect of the influence of the three parameters on the SCR fatigue damage. In Figure 3 it is seen that the maximum fatigue damage increases slightly with increase of the maximum normalised stiffness and the location of the maximum fatigue damage is largely

Figure 5. Fatigue damage profiles for different shear strength gradients.

Figure 7. Suction ratio effect on predicted fatigue life.

Figure 8. Soil shear strength gradient effect on predicted fatigue life.

Figure 6. Normalized maximum stiffness effect on predicted fatigue life.

unaffected (remaining close to the initial TDP, which is around 1170 m arc length). Figure 4 shows that the maximum fatigue damage increases as the suction ratio increases, while the location of maximum damage shifts towards the floating platform. For a realistic maximum suction ratio of 0.4, the maximum damage is increased by 50% Figure 5 shows that the maximum fatigue damage increases markedly as the shear strength gradient is increased. It is also observed that as the shear strength gradient increases the location of maximum damage shifts away from the floating platform. Figures 6, 7 and 8 show how the estimated fatigue life of the system is affected by each parameter. Note that the fatigue life is dictated by the most critical section of the riser in the touchdown zone, equivalent to 20 years divided by the maximum 20-year life fatigue damage. From Figure 6 it is clear that the normalised maximum stiffness has no significant effect on the fatigue life of the system. Increasing the normalised maximum stiffness from 100 to 300 only leads to a reduction of approximately 10 years in the predicted fatigue life of the SCR. This translates to only 5% drop in predicted fatigue life. Figure 7 shows a more significant effect of the suction ratio on the SCR fatigue life. Increasing the suction ratio from 0 to 0.6 shows a reduction in fatigue life from around 220 years to 120 years, i.e. a 45% © 2011 by Taylor & Francis Group, LLC

reduction. A similar reduction, from 180 years to 100 years, occurs as the soil shear strength gradient is increased from 1.5 to 10 kPa/m, as shown in Figure 8. The flattening of the curve for shear strength gradients higher than 10 kPa/m suggests that further increase in the shear strength gradient has a more limited effect on the fatigue life. Note that a more realistic upper limit of ∼0.3 on the suction ratio would limit the reduction in fatigue life to about 25% (see Figure 7). 4.2

Effect of loading sequence on fatigue life

The effect of the loading sequence on fatigue life was explored by (a) running LS1 in reverse order and (b) running LS2 (i.e. each wave packet separately). These analyses showed that the original LS1 gave the highest damage, with the fatigue life increased by 10% (reverse LS1) and 15% (LS2). 4.3

Equivalent linear stiffness

The fatigue analyses using the non-linear model gave a fatigue life between 120 and 220 years. It is of interest to note what linear seabed stiffness, k (kN/m/m or kPa) would give a similar fatigue life. This is shown in Figure 9, from which it may be seen that a linear stiffness of 10 to 50 kPa would give the similar range of fatigue life. This is considerably lower than is commonly adopted in fatigue design studies.

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It is recommended that nonlinear pipe-soil interaction should be taken into account in order to allow reliable, but economic, fatigue design of SCRs.

ACKNOWLEDGEMENTS

Figure 9. Variation of fatigue life with linear seabed stiffness.

5

CONCLUSIONS

The nonlinear soil model recently incorporated in the OrcaFlex software for dynamic analysis of risers was used to conduct a parametric study to investigate the influence of three key soil parameters on the fatigue response of a typical deepwater SCR. The key soil parameters explored were: the normalised maximum stiffness, the suction ratio that determines the maximum uplift resistance, and the soil shear strength gradient. Results of the sensitivity analysis show that: • The normalised maximum stiffness has no signifi-

cant effect on the maximum fatigue damage or its location along the riser. • Increasing the suction ratio or shear strength gradient lead to a decrease in fatigue life. • Fatigue life is more sensitive to the soil shear strength gradient than to the suction ratio. • Increasing the suction ratio shifts the point of maximum fatigue damage towards the floating platform whereas increasing the shear strength gradient shifts it away from the floating platform. It was also found that, while the ordering of the wave packets had some effect on the fatigue life, this was of secondary importance (less than 15%). Analyses using linear seabed stiffness showed that a low stiffness, in the range 10 to 50 kPa, was required to obtain a similar range of fatigue life to that predicted using the nonlinear pipe-soil model. The results from this fatigue parametric study confirm that the soil parameters in the nonlinear soil model can have a significant effect on the SCR fatigue damage in the touchdown zone. However, the fatigue life obtained for typical soil parameters is rather greater than would be obtained using typical linear seabed stiffness values as currently adopted in design.


This work forms part of the activities of the Centre for Offshore Foundation Systems (COFS), established under the Australian Research Council’s Research Centres Program and now supported by the State Government of Western Australia as a Centre of Excellence. The authors would also like to thank Orcina Ltd for their technical support for this study.

REFERENCES Aubeny, C.P. & Biscontin, G. 2009. Seafloor-Riser Interaction Model. Int. J. of Geomechanics, ASCE, 9(3), 133–141. Aubeny, C.P., Gaudin, C. & Randolph, M.F. 2008. Cyclic tests of model pipe in kaolin. Proc. Offshore Technology Conference, Houston, Paper OTC19494. Bridge, C., Laver, K., Clukey, E. & Evans, T. 2004. Steel catenary riser touchdown point vertical interaction models. Proc. Offshore Technology Conference. Houston, Texas, Paper OTC16628. Bridge, C. & Willis, N. 2001. Steel catenary risers – results and conclusions from large scale simulations of seabed interaction, 2H Offshore Engineering Ltd, Woking, Surrey, UK Clukey, E., Haustermans, L. & Dyvik, R. 2005. Model tests to simulate riser-soil interaction effects in touchdown point region. Proc. Int. Symp. on Frontiers in Offshore Geotechnics, ISFOG, Perth, Australia, 651–658. Clukey, E., Ghosh, R, Mokarala, P. & Dixon, M. 2007. Steel catenary riser (SCR) design issues at touch down area, Proc. 17th Int. Conf. on Offshore and Polar Engineering, ISOPE, Lisbon, 814–819. Grealish, F., Kavanagh, K., Connaire, A. & Batty, P. 2007. Advanced nonlinear analysis methodologies for SCRs. Proc. Offshore Technology Conference 2007. Houston, Texas, Paper OTC18922. Kimiaei, M., Randolph, M.F. & Ting, I. 2010. A parametric study on effects of environmental loadings on fatigue life of steel catenary risers. Proc. 29th Int. Conf. on Ocean, Offshore and Arctic Eng., Shanghai, Paper OMAE201021153. Orcina 2009. OrcaFlex User Manual, www.orcina.com, UK. Randolph, M.F. & Quiggin, P. 2009. Non-linear hysteretic seabed model for catenary pipeline contact. Proc. 28th Int. Conf. on Offshore Mech. and Arctic Eng., Honolulu, USA, Paper OMAE2009-79259.

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Mobilization distance in uplift resistance modeling of pipelines J. Wang & S.K. Haigh University of Cambridge, UK

N.I. Thusyanthan & S. Mesmar KW Ltd.

ABSTRACT: Upheaval buckling (UHB) of pipelines is a phenomenon by which buried pipelines buckle due to the increased temperature and pressure of operational conditions. UHB is resisted by the resistance of the soil cover. This paper presents a series of experiments designed to investigate the mobilization required for soil cover to provide its peak uplift resistance. It is shown that the DNV-RP-F110 code recommended mobilization is unconservative, leading to overly stiff force-displacement response, especially for higher H/D ratios. As buckling is a stiffness governed behavior, underestimation of peak mobilisation and hence overly stiff force-displacement response will lead to unconservative designs. Based on test results from this research and data available from literature, a new equation in terms of H/D is proposed for predicting peak mobilization.

1

INTRODUCTION

Pipeline networks are instrumental for transporting hot crude oil from offshore platforms to onshore refineries. At shallow water sites (water depth ≤ 15 metres), the trench-and-burial method is typically adopted for pipeline laying projects and the excavated material during trenching is used as primary backfill. One of the main purposes of burial is to preventthe pipeline from heaving upwards, the result of a phenomenon known as upheaval buckling (UHB). UHB is a thermally induced structural effect. The operating conditions of high temperature and large internal pressure, which are significantly above the ambient seabed conditions at first laying, lead to thermal extension along the pipeline. These axial movements are restricted by the frictional forces at the soil-pipeline interface. Large compressive forces are then developed, which can cause the pipeline to buckle globally if lateral restricting forces are inadequate. At locations where the pipeline profile features an over-bend, the most likely buckling mode is for the pipeline to heave upwards through the backfill soil, hence the name upheaval buckling. Once UHB has initiated, additional upward movement of the pipeline would lead to reduced axial compressive force. At the same time, the imperfection curvature would increase, making it easier for the pipeline to buckle. Hence, the stability of the pipeline would depend on the interaction of these two effects. The Current understanding ofuplift resistanceis based on analyses and experimental work (Vesic, 1971; Rowe and Davis, 1982; Hobbs, 1984; Randolph and © 2011 by Taylor & Francis Group, LLC

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Houlsby, 1984; Trautmann et al., 1985; Palmer et al., 1990; Schaminée et al., 1990; Dickin, 1994; Croll, 1997; Moradiand Craig, 1998; Baumgard, 2000; White et al., 2001; Bransby et al., 2001; and Cheuk, 2005) by researchers on both structural and geotechnical fronts. Most research effort has been directed at identifying the maximum available uplift resistance in granular soils. Little has been concluded, however, on the full uplift force-displacement regime. Small-scale centrifuge tests are widely used as a modeling technique for uplift resistance. Comparison between model and full-scale experiments shows good agreement on the maximum available uplift resistance, but there are significant discrepancies in the dimensionless mobilization displacement (Palmer et al., 2003). Current design guidelines and practice for pipelines are more empirical than analytical. Improved design efficiency would arise from a better understanding of the deformation mechanism during the uplift event and hence a more robust theoretical basis for the prediction of the mobilization of available uplift resistance with pipe upward displacement. A series of both full scale (1 g) and centrifuge (30 g) experiments have been conducted at the Schofield Centre, University of Cambridge, to model the uplift response of pipelines buried in saturated sand. One of the key objectives of this research is to investigate the mobilization distance to peak uplift resistance. The Particle Image Velocimetry (PIV) technique (White et al. 2003) has been employed to reveal the true deformation mechanism involved in the uplift process. This research paper will elaborate on these findings.

Table 1.

Parameters for the DNV tri-linear design curve.

Soil type

Parameter

Range

Loose Sand (H/D range 3.5 to 7.5)

fp δf ∗ α β

∈ ∈ ∈ =

[0.1, 0.3] [0.5%, 0.8%] H [0.75, 0.85] 0.2

Medium/Dense Sand (pre-peak) (H/D range 2 to 8)

fp δf ∗ α β

∈ ∈ ∈ =

[0.4, 0.6] [0.5%, 0.8%] H [0.65, 0.75] 0.2

Rock (H/D range 2 to 8)

fp δf ∗ α β

∈ ∈ ∈ =

[0.5, 0.8] [20 mm, 30 mm] 0.35 D 0.2

Figure 2. Trautmann-Pedersen vertical slip surface model.

where σ configuration is the standard deviation for the survey accuracy of the pipeline configuration and has a minimum value of 0.025 m. However, this minimum value would lead to γUR = 0.925 for cohesionless backfill, i.e. less conservative than without applying this safety factor, which contradicts its original intention. The Trautmann-Pedersen (Pedersen & Jensen, 1988) Vertical Slip Surface Model (Figure 2) and the associated design equation is adopted by DNV to provide an estimate for Rmax :

*: Recommended values from DNV

where: D is the external diameter of the pipe; H is distance between the soil surface and the pipe crown; γ is the submerged unit weight of the soil; and fp is the dimensionless Pedersen Uplift Resistance Factor. The recommended range of fp values for different types of cohesionless soils is also shown in Table 1. The limited range of H/D ratio for which Equation 2 applies should be noted (Table 1). For design scenarios with H/D ratios below 1.0, the typical current practice is to force fp = 0, and limiting Rmax arbitrarily to the weight of the soil cover alone:

Figure 1. Tri-linear uplift force-displacement model with global safety factor applied.

2

REVIEW ON CURRENT DESIGN PRACTICE

The current industry design practice for uplift resistance modeling is prescribed in Appendix B of the offshore design code, DNV-RP-F110. Description of soil models for both cohesive (clay) and cohesionless (sand and rock) soils are given. This research has concentrated on the latter soil type only. The DNV design code states that the characteristic response of cohesionless soils to UHB can be approximated by a normalized tri-linear uplift forcedisplacement curve (Figure 1), with Rmax and δf being the maximum available uplift resistance per meter along the pipeline and the corresponding mobilization distance respectively. The geometry of this tri-linear characteristic curve can be accurately defined by three parameters: α, β, and δf .The prescribed ranges of these parameters for different types of cohesionless backfill are summarized in Table 1. To obtain the more conservative design curve, a global safety factor, γ UR , must be applied as shown in Figure 1. For cohesionless soils, this global safety factor, γ UR , is given by:

The tri-linear design curve can then be applied. This extra-conservatism at low H/D ratios can lead to large quantities of rock dump material being required as secondary backfill, which can cost millions of dollars. The major reason behind this conservatism is the absence of available data at these low H/D ratios. Recent research (Wang et al., 2010; Thusyanthan et al., 2010) suggests that Equation 2 still provides a good estimate for Rmax at H/D ratios below 1.0 in both sand and rock. 3

RESEARCH OBJECTIVES

A series of full-scale pipeline uplift resistance tests have been devised and conducted at the Schofield © 2011 by Taylor & Francis Group, LLC

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Table 2.

Figure 3. Full-scale plane-strain test tank in (a) schematic plot, and (b) photograph of complete set-up.

Centre, University of Cambridge. The test series can be further divided into two categories: 1. Plane-strain testing in saturated sand 2. Full-scale field testing in moist sand

Test No.

Backfill

H/D

Pipe Diameter (mm)

1 2 3 4 5 6 7 8 9a 9b

Sand Sand Sand Sand Sand Sand Sand Sand Sand Sand

0.1 0.4 0.5 1 2 3.5 0.5 1.0 6 8

100 100 100 100 100 100 258 258 200 200

(W) × 850 mm (H). More details of the test tank can be found in Wang et al., (2010). Two model pipes with external diameters of 100 mm and 258 mm with PTFE front and back faceswere manufactured. The PTFE material has a very low friction angle, which minimizes the end effects in 2D plain strain modeling. Either pipe will be connected to the actuator via a 12 mm diameter aluminum rod. The cross sectional area of the rod represents 1.03% of the projected area of the 100 mm diameter model pipe and 0.40% of the 258 mm diameter model pipe. Hence its effect on the measured uplift resistance is negligible. Fraction E sand of relative density ID = 35% (loose), median diameter D50 of 0.15 mm and saturated unit weight γ sat of 18.5 kN/m3 was used as backfill. The Particle Image Velocimetry (PIV) technique was employed using a single Canon G10 digital camera to capture the displacement field of the backfill around the pipe throughout the pull-out process. Tests No. 9a and 9b were conducted at full-scale in the field via vertical hydraulic jacking. Fine sand with properties very similar to that of Fraction E Sand was used. The measured moisture content of the soil was approximately 4.6% and the bulk unit weight 15 kN/m3 . The model pipe used had an external diameter of 0.2 m and a length of 1 m.

The principal objective of these tests is to understand how mobilization distance varies with H/D ratios in cohesionless soils. Another parallel objective is to understand the reliability of the shear component of the “true” uplift resistance in cohesionless soils at H/D ratios below 2, with a particular focus on H/D ratios below 1. Detailed results and conclusions on this aspect are explained in Wang et al., (2010).

4 APPARATUS AND TEST PROGRAM The test program for the entire test series is summarized in Table 2. Figure 3 illustrates the full-scale plane-strain test tank used for Tests No. 1 to 8. The container has internal dimensions of 1000 mm (L) × 76 mm © 2011 by Taylor & Francis Group, LLC

Full-scale test program for uplift resistance in sand.

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5


Representative uplift force-displacement raw data are illustrated in Figure 4. This data can be normalized using the DNV approach so as to be compared directly with the recommended tri-linear design curve. This is illustrated in Figure 5. Figure 5 illustrates that, if normalized with measured values for mobilization distance δf , the DNV upper and lower bound normalized plots for loose sand are reasonably representative of the experimental data. However, the measured δf values are very different from the DNV recommendations of the mobilization distances being between 0.5% H and 0.8% H. Comparison between the measured and the DNV recommended values for δf is summarized in Table 3.

Table 3. values.

Summary of actual and DNV recommended δf DNV Recommendation (mm)

Test No.

H (mm)

Measured δf (mm)

0.5 % H

0.8% H

1 2 3 4 5 6 7 8 9a 9b

10 40 50 100 200 350 129 258 1200 1600

2.5 3.0 2.5 3.3 4.6 10 10 8 110 215

0.05 0.2 0.25 0.5 1.0 1.75 0.65 1.29 6 8

0.08 0.32 0.4 0.8 1.6 2.8 1.0 2.1 9.6 13

Figure 4. Representative uplift force-displacement data.

Figure 6. Summary of normalized mobilization distance.

Figure 5. DNV characteristic response compared with representative:(a) plane-strain test data; and (b)field test data, normalised by measured δf instead of values suggested by DNV.


It is apparent that the DNV code vastly underestimates the mobilization distance required to reach Rmax . In addition, DNV suggests that δf is only a function of H and independent of the H/D ratio, hence the diameter of the pipe should not affect δf if H remains constant. According to Table 3, this does not seem to hold: The cover heights in Tests No. 4 (D = 100 mm) and 7 (D = 258 mm) only differ by 29 mm, but δf values differ by 200%. Thusyanthan et al. (2010) suggests that a good correlation can be obtained by normalizing the mobilization distance with pipe external diameter and plotting this dimensionless mobilization, δf /D, against the H/D ratio on a log-linear scale. This correlation is shown in Figure 6, using experimental results from Tests No. 1 to 9 as well as other available uplift resistance data in loose sand from literature. A linear trend line is evident from Figure 6 which can be described by Equation (4). The latest full-scale uplift tests in rock exhibit similar linear trends with gradient apparently dependent upon grain size.

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strain originate from both edges of the pipe crown, and start to propagate almost vertically towards the soil surface. Newborn compression fronts start to converge on and merge into the two macroscopic shear bands. Rmax is usually reached when this macroscopic shear band just reaches the soil surface. 3. At post-peak displacements (δ > 4.6 mm in this case), the existing mechanism reinforces itself, and the two macroscopic shear bands start to move sideways and widen in a very gradual manner. The widths of both the smaller compression fronts and the bigger macroscopic shear bands seem independent of the H/D ratio. At peak uplift resistance, the centre lines of the two shear bands coincide very well with the two shear planes specified in the Vertical Slip Surface Model. However, their finite but significant width suggests that this model is at most an approximation to the true uplift deformation mechanism in loose sand at these H/D ratios. PIV strain analysis also provides useful insight into the width of the influence zone above the soil surface during the uplift event. In loose sand, the width of this influence zone seems to be approximately 2.5 times the pipeline external diameter. Hence if additional downward force is to be provided by rock dump, the width of dumping should be of similar dimensions to ensure maximum efficiency. Figure 7. Evolution of total shear strain for Test No. 5.

7 6

CONCLUSIONS

DEFORMATION MECHANISM

For Tests No. 1 to 8, the soil displacement field during the uplift event was accurately measured at 5-second intervals, which corresponds to 0.025 mm of upward displacement by the model pipe. This was achieved using the non-contact digital image correlation technique of particle image velocimetry (PIV), described in detail by White et al. (2003). As an example, the evolution of soil shear strain with pipe upward displacement for Test No. 6 (H/D = 2) is illustrated in Figure 7. It is clearly visible that the uplift mechanism in loose sand can be divided into three phases: 1. At very small displacements (δ < 1 mm in this case), thin strands of compression fronts (interpreted as shear bands in total shear strain plots) originate from one side of the pipe crown, fanning out gradually to the other side and swiftly propagating through the backfill soil medium. Localized dilation shear zones start to appear underneath the pipe, which ultimately creates a wedge-shaped void. 2. At small pre-peak displacements (1 mm < δ < 4.6 mm in this case), propagation becomes slower and slower and almost comes to a standstill when these compression fronts have rotated by over 90◦ . Subsequent compression fronts start to superimpose on their predecessors. Two cumulated macroscopic shear bands of more than 5% total shear © 2011 by Taylor & Francis Group, LLC

This paper presented the results of full-scale tests on the upheaval buckling resistance of pipelines in loose saturated sand in an attempt to clarify the mechanics of the pipe-soil interaction. The results were compared with behaviour suggested by the DNV-RP-F110 design code which was shown to be non-conservative in estimating mobilization distance. On the other hand, Wang et al., (2010) shows that the DNV code is conservative in estimating the maximum available uplift resistance. Buckling is a stiffness dominated process, so the most important parameter in determining whether or not a strut or pipeline will buckle is the stiffness of the restraining “spring”. The mobilization distance predicted by the DNV code underestimates measured values by factors between 5 and 50, providing a non-conservative design approach. Equation 4 gives an expression for mobilization distance that is a function of both cover H and pipe diameter D.

ACKNOWLEDGEMENT The authors would like to thank all staff at the Schofield Centre, University of Cambridge for their help and advice throughout the testing program. The first author would also like to thank Trinity College, University of Cambridge, and KW Ltd. for

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their generous financial support towards this research effort. REFERENCES Baumgard, A.J. (2000). Monotonic and cyclic soil response to upheaval buckling in offshore buried pipelines. PhD Thesis.University of Cambridge. Bransby, M.F., Newson, T.A., Brunning, P., and Davies, M.C.R. (2001). Numerical and centrifuge modelling of upheaval resistance ofburied pipelines. Proc. OMAE, Rio de Janeiro, June 2001. Cheuk, C.Y. (2005). Soil pipeline interaction at the seabed.PhD thesis.University of Cambridge. Croll, J.G.A. (1997). A simplified model of upheaval thermal buckling of subsea pipelines.Thin-Walled Structures 29(1-4): 59–78. Dickin, E.A. (1994). Uplift resistance of buried pipelines in sand. Soils and Foundations 34(2): 41–48. DNV-RP-F110, Global buckling of submarine pipelines – structural design due to high temperature/high pressure. Det Norske Veritas, Norway, 2007. Hobbs, R. (1984). In service buckling of heated pipelines.ASCE Journal of Transportation Engineering 110(2): 175–189. Moradi, M. and Craig, W.H. (1998). Observations of upheaval buckling of buried pipelines. Centrifuge 98, Kimura, Kusakaba&Takemura(eds), ISBN 90 5410 986 6 Palmer, A.C., Ellinas C.P., Richards, D.M., and Guijt, J. (1990).Design of submarine pipelines against upheaval buckling.Proc. Offshore Technology Conf., Houston, OTC 6335: 551–560. Palmer, A.C.White, D.J., Baumgard, A.J., Bolton, M.D., Barefoot, A.J.Finch, M., Powell, T., Faranski, A.S., Baldry, J.A.S. (2003). Uplift resistance of buried submarine pipelines: comparison between centrifuge modeling and full-scale tests. Géotechnique 53(10): 877–883.


Pedersen, P.T. & Jensen, J.J. (1988).Upheaval creep of buried pipelines with initial imperfections. Marine Structures 1:11–22, 1988. Randolph, M. F., &Houlsby, G. T. (1984). The limiting pressure on a circular pile loaded laterally in cohesive soil. Géotechnique, 34(4): 613–623. Rowe, R.K., and Davis, E.A. (1982).The behaviour of anchor plates in sand.Géotechnique 32 (1): 25–41. Schaminée, P.E.L., Zorn, N.F., and Schotman, G.J.M. (1990). Soil response for pipeline upheaval buckling analysis: Full-scale laboratory tests and modelling. Offshore Technology Conference, Houston, OTC 6486 Thusyanthan, N.I., Mesmar, S., Wang J., and Haigh, S.K. (2010). Uplift resistance of buried pipelines and DNV-RPF110 guideline. Proc. Offshore Pipeline and Technology Conference.Feb 24–25, Amsterdam, Netherlands. Trautmann, C.H., O’Rourke, T.D., and Kulhawy, F.H. (1985). Uplift force-displacement response of buried pipe. ASCE Journal of Geotechnical Eng. Division 111(9): 1061–1075. Vesic, A.S. (1971). Breakout resistance of objects embedded in ocean bottom. ASCE Journal of the Soil Mechanics and Foundation Division.97 (9): 1183–1205. Wang, J., Ahmed, R., Haigh, S.K., Thusyanthan, N.I., and Mesmar, S. (2010). Uplift resistance of buried pipelines at low cover-diameter ratios. Proc. Offshore Technology Conference.May, 2010, Houston, USA. White, D.J., Barefoot, A.J., Bolton, M.D. (2001). Centrifuge modelling of upheaval buckling in sand. International Journal of Physical Modelling in Geotechnics, 2(1): 19–28. White, D. J., Take, W. A. & Bolton, M. D. (2003). Soil deformation measurement using particle image velocimetry (PIV) and photogrammetry.Géotechnique,53(7): 619–631.

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Theoretical, numerical and field studies of offshore pipeline sleeper crossings Z.J. Westgate, M.F. Randolph & D.J. White Centre for Offshore Foundation Systems, University of Western Australia, Perth

P. Brunning Acergy, Singapore

ABSTRACT: Offshore pipelines experience axial stresses due to internal pressure and thermal cycles during start-up and shut-down, leading to the formation of lateral buckles. Pipeline lay routes often include strategicallyspaced transverse sleepers, creating a vertical imperfection from which a controlled lateral buckle can be initiated. The as-laid pipeline embedment affects the pipe-soil interaction forces and therefore the buckle initiation response. The potential for increased embedment in the touchdown zones to either side of the sleeper exists, which can cause higher than expected lateral breakout forces. The magnitude of this embedment is difficult to quantify due to dynamic lay effects and changes in catenary forces as the pipe is laid over the sleeper. Theoretical solutions for the touchdown force at a sleeper crossing are presented for the case of an elastic seabed, and are compared to as-laid survey data in soft clay. Static and dynamic numerical analyses are also presented to illustrate the changes in pipe-soil contact force as a pipe is laid across a sleeper. This provides a rationale for asymmetry observed in as-laid embedment profiles and its influence from dynamic lay effects. General guidance is provided for pipeline designers to assist the assessment of pipe-soil interaction forces in the vicinity of sleepers.

1

INTRODUCTION

As offshore hydrocarbon developments have progressed into deeper waters, new pipeline design issues have arisen, such as lateral buckling, which occurs due to axial pipe stresses during thermal cycles of start-up and shut-down. This has led to development of lateral buckle mitigation techniques, such as sleepers. Sleepers allow controlled buckles to occur in predetermined locations along the route by creating a vertical imperfection on which the pipeline slides laterally (Sinclair et al. 2009). In deep water developments where soft fine-grained soils are prevalent, the lay process induces significant pipeline embedment due to dynamic lay effects (Randolph & White 2008).As-laid field surveys (Lund 2000, Westgate et al. 2010) show that pipeline embedment during normal lay conditions in the field can be up to an order of magnitude greater than the predicted static pipeline embedment based on the intact soil strength. Using the remoulded strength in this calculation has been shown to predict embedment closely matching field observations (Westgate et al. 2010). Pipeline embedment influences the lateral breakout resistance provided by the seabed soil. Together with dynamic lay effects, the variation in pipe-soil contact force across a sleeper influences the magnitude of pipeline embedment. This can lead to differences in the lateral breakout resistance along the touchdown © 2011 by Taylor & Francis Group, LLC

zones to either side of the sleeper, affecting the pipeline stresses along the buckle. This study illustrates the variation in pipe-soil contact force in the vicinity of sleepers, and the dependence of the resulting pipeline embedment on the asymmetric and dynamic nature of the lay process. The paper presents results of theoretical analyses used to calculate the pipe-soil contact force and embedment based on the standard catenary solution, and static and dynamic numerical analyses of the lay process. These are compared to as-laid field data for a pipeline installed at a soft clay site in deep water. 2 THEORETICAL ANALYSES Theoretical methods for calculating pipe-soil contact forces during pipe laying are well-established (Lenci & Callegari 2005, Palmer 2008). The governing equation for the response of a pipe being laid on an elastic seabed can be written as:

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where E = elastic modulus, I = second moment of area, T0 = horizontal pipe tension, p = submerged pipe weight, k = seabed soil stiffness, w = pipe embedment, and x = the distance along the pipeline.

equation 1 to be solved. It is worth noting that for the extreme (though unrealistic) case of T0 = 0, the character of the solution to equation 1 changes from a catenary to a beam solution. Several variations of pipe lay over a sleeper were analysed, increasing in realism and computational effort (Table 1). Table 2 lists the pipeline properties and lay conditions for these analyses. These correspond to a pipeline for which the lay conditions and as-laid embedment is known (Case 8). 2.2

Figure 1. Idealised sleeper crossing (vertical scale exaggerated for clarity).

The horizontal pipe tension is calculated based on the standard catenary solution as:

where φ = the lay angle, defined as the inclination to the horizontal at the lay ramp departure point, and zw = the water depth. The horizontal component of tension is constant along the suspended pipeline. The pipe catenary creates a vertical force concentration at the touchdown zone expressed as the pipe-soil contact force V , normalised by the submerged pipe weight p. The ratio V /p is related to the characteristic length λ = (EI/T0 )0.5 , after Pesce et al. (1998). 2.1 The sleeper crossing problem The as-laid sleeper crossing problem can be idealised as a beam on an elastic seabed due to the small vertical deformation of the pipeline in the vicinity of the sleeper (Figure 1). For the case of a pipeline laid from a vessel, the maximum force concentration Vmax /p in the touchdown zone may be approximated as (Randolph and White 2008):

For the case of a pipeline lowered from a horizontal plane above a flat seabed under zero tension, i.e. ‘placed’, Vmax /p = 1 everywhere along the pipeline due to the absence of the catenary. The presence of a sleeper of height h above the seabed complicates this condition due to the additional boundary conditions imposed (Figure 1). Visual inspection shows that deflection, slope, shear and moment are continuous at x = 0, with deflection w = 0. Similarly, at x = L, w = − h and slope = 0, which allow © 2011 by Taylor & Francis Group, LLC

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Static pipe lay and placement on linear seabed

Figure 2 shows the variation in V /p for the zero tension, pipe placement condition (Case 1) and the as-laid catenary tension of the real case (Case 2). For this geometry and tension there is little difference between the as-laid catenary case with tension and the zero tension placement case, with a maximum V /p equal to 1.38 and 1.5 respectively. The adopted stiffness of k = 20 kPa for the elastic solutions is consistent with the theoretical pipe penetration prediction of Merifield et al. (2009) for an intact shear strength gradient of su = 20 kPa/m (negligible mudline strength intercept) as applicable to the field case discussed later. This stiffness gave a nominal embedment due to the pipe weight p of w/D = 0.09 (i.e. w = 0.09 pipe diameters) and a maximum embedment due to the catenary contact force V of w/D = 0.13. Reducing the seabed stiffness in the theoretical solution represents a case where the seabed has softened due to dynamic lay effects. A reduced stiffness of 8 kPa represents a fourfold drop to remoulded conditions, based on a remoulded shear strength gradient of 5 kPa/m as applicable to the field case. This decreases the maximum force concentration by ∼10% so the nominal and maximum embedment values increase by less than fourfold to w/D = 0.22 and w/D = 0.27 respectively (Figure 3). These solutions represent symmetric sleeper crossing cases, with the same maximum contact force and embedment on each side of the sleeper. However, the lay process is asymmetric, as illustrated conceptually in Figure 4. As the pipe is laid from the lay vessel, it approaches the sleeper with a catenary configuration independent of the sleeper’s presence (dotted line). On the trailing side of the sleeper, the pipe lifts off of the seabed as it contacts and rotates over the sleeper (due to the pipe stiffness), causing the touchdown point to move away from the sleeper through the ‘uplift zone’ (dashed line to left of sleeper). On the leading (i.e. vessel) side of the sleeper, the additional weight of the suspended pipe causes the pipe to embed further compared to the nominal pipe catenary force (dashed line to right of sleeper). This load-unload mechanism results in a similar uplift zone on the leading side of the sleeper. The end result is an asymmetric embedment profile on a real (plastic) seabed (heavy solid line) which exceeds that of an idealised (elastic) seabed (light solid line).

Table 1.

List of analyses.

1 2 3 4 5 6 7 8

Theoretical Theoretical Theoretical Numerical Numerical Numerical Numerical Empirical

Table 2.

Pipeline properties and lay conditions.

Placement Static lay Static lay Static lay Static lay Static lay Dynamic lay Real pipe lay

Linear/intact Linear/intact Linear/remoulded Linear/intact Non-linear/intact Non-linear/remoulded Non-linear/intact Real soil

Parameter

Idealised

Field case

Outside diameter, D (m) Steel thickness (mm) Bending rigidity, EI (MNm2 ) Coating thickness (mm) Submerged pipe weight, p (kN/m) Water depth, zw (m) Lay angle, φ (deg) Horizontal pipe tension, T0 )kN) Sleeper height, h (m) Significant wave height, Hs (m)

0.32 19 44 0 0.57 1300 83 103 0.9 2

0.32 19 44 2.6 0.58 1240–1310 82.6 106–112 0.9–1.0 0.7–2.4

Zero As-laid As-laid As-laid As-laid As-laid As-laid As-laid

Increasing realism and computational effort

Figure 3. Theoretical analysis for remoulded seabed (Case 3).

For a horizontal tension of 103 kN, the catenaryinduced contact force is 1.73 and the final sleeper-induced contact force is 1.38, a reduction of 20%. For the final pipe-sleeper configuration, the maximum V /p reduces towards a value of 1.5 at zero tension. At this point, the maximum V /p for the catenary is infinite, but rapidly reduces for more realistic values of T0 /λp. At high normalised tension values, both conditions converge to V /p = 1. Figure 2. Theoretical analyses for intact seabed (Case 1 and 2).

An analysis of the lay process on a non-linear (plastic) seabed will show this asymmetry since the overloading history will force irreversible embedment. However, the elegant elastic cases allow trends related to the lay tension and seabed stiffness to be explored – in particular the variation in force concentration in the touchdown zones.

2.4

Figure 6 shows V /p as a function of the normalised seabed stiffness K = λ2 k/T0 . As the seabed stiffness increases, both the absolute V /p values and the ratio of V /p between the two cases increase. As the stiffness reduces, V /p converges to unity. Normalised stiffness values for soft clay seabeds are typically in the range of K = 100–1000, but for very weak remoulded soft clays K can be lower, and for stiffer clays K can exceed 10,000.

2.3 Effects of pipe tension

3

Figure 5 shows V /p as a function of the normalised pipe tension T0 /λp, for a seabed stiffness of k = 20 kPa. The temporary pipe-soil contact force from the pipe catenary is always greater than the final contact force after the pipeline has crossed the sleeper, with the difference increasing with reducing pipe tension. © 2011 by Taylor & Francis Group, LLC

Effects of seabed stiffness

FIELD STUDIES

Variations in pipeline embedment across sleepers were obtained from as-laid field survey data from a development in deep water with a soft clay seabed. The intact undrained shear strength gradient in the upper 0.5 m of the seabed is about 20 kPa/m, with a remoulded

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Figure 7. As-laid field survey embedment profiles compared to as-laid theoretical solutions (Cases 2 and 3).

Figure 4. Sleeper crossing lay process (vertical scale exaggerated for clarity).

rotational motion, thus minimising effects of the vessel motions on the pipe-soil contact forces. Profiles of the as-laid normalised pipeline embedment w/D across the sleepers (Case 8) are shown in Figure 7, compared to the as-laid theoretical solution for the intact and remoulded strength gradients (Cases 2 and 3). The mean embedment for this pipeline away from the sleepers was 0.31D. At the sleeper crossings, the embedment ranged from about 0.2D to 0.5D. The leading side of the sleeper exhibited slightly deeper embedment consistent with the lay process discussed in Section 2. The as-laid embedment data showed that the height of the sleeper crown h above the seabed was between 0.9 and 1.0 m, which affects the length of the hanging spans L on each side of the sleeper (Figure 1). The theoretical solution for the intact seabed (Case 2) exhibits the correct shape of the embedment response, but significantly under predicts the magnitude of the embedment and over predicts the length of the hanging spans. If the remoulded strength is used in the theoretical solution (Case 3), the embedment and span length are closer to the field values, but the asymmetry is absent.

Figure 5. Influence of horizontal pipe tension on maximum pipe-soil contact forces.

4

Figure 6. Influence of seabed stiffness on maximum pipe-soil contact forces.

The dynamic riser analysis software, OrcaFlex (Orcina 2009), was used to explore the trends shown in the field data. The first numerical analysis (Case 4) comprised static pipe lay on a linear elastic seabed. Further analyses (Cases 5, 6, 7) show the influence of non-linear seabed stiffness, a reduced strength from remoulding and vessel motions respectively. The idealised pipe properties used in the numerical model are those used in the theoretical analyses (Table 2).

strength gradient of 5 kPa/m. The properties and lay conditions for the surveyed pipeline are summarised in Table 2. The pipeline was laid from a J-lay vessel with a lay ramp that permitted 3 degree-of-freedom © 2011 by Taylor & Francis Group, LLC

NUMERICAL ANALYSES

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4.1

Static pipe lay on linear seabed

Figure 8 shows the variation in V /p and w/D for the linear seabed (Case 4). The solid line shows the final profile, which matches the theoretical solution

Figure 8. Variation in pipe-soil contact force and embedment for static pipe lay on linear intact seabed (Case 4).

from Case 2. The dashed line shows the maximum (temporary) profile exhibited during the lay process (Figure 4). As the pipe catenary approaches the sleeper, the force concentration factor is constant at V /p = 2.1, with a corresponding embedment w/D = 0.16. The final embedment reduces (elastically) to w/D = 0.09, corresponding to the submerged pipe weight. After making contact with the sleeper, V /p reduces as the pipe weight is redistributed to the sleeper and eventually to the seabed on the leading side of the sleeper. The increase in contact force on the leading side (due to the heavier catenary) is 20% higher than that on the trailing side, with a maximum V /p = 2.7. The pipe embedment for this force is w/D = 0.2. Once the sleeper crossing is completed, the maximum force concentration factor returns to the nominal value of V /p = 2.1, for which w/D = 0.16. For the elastic seabed, the increased V /p has no effect on the final embedment. Although an asymmetric final profile cannot be predicted by the elastic seabed model, the progressive laying simulation of Case 4 shows a maximum V /p profile that is consistent with the asymmetry in the field data of as-laid embedment (Figure 7).

Figure 9. Variation in pipe-soil contact force and embedment for static pipe lay on non-linear seabed (Cases 5 and 6).

uplift (transient tensile) zone is evident on both the trailing and leading sides of the sleeper, as discussed in Section 2. The final pipe embedment is close to the maximum embedment profile due to the plastic seabed deformation, and captures the asymmetry well. A second static non-linear case was carried out for the remoulded shear strength gradient of 5 kPa/m (Case 6). The final embedment profile for this case is also shown in Figure 9. The weaker penetration resistance due to the lower strength gradient results in deeper pipe embedment. This final ‘remoulded’ embedment is closer to the mean embedment profile observed in the field and captures the asymmetry.

4.2 Static pipe lay on non-linear seabed A non-linear seabed with the intact shear strength gradient of 20 kPa/m (mudline strength intercept of zero) was analysed for static pipe lay. The non-linear soil model in OrcaFlex has a stiff unloading response following penetration, thus capturing the effect of a previous overloading event (Randolph & Quiggin 2009). The model also captures increasing pipeline penetration with cycles of vertical loading, simulating the effects of soil softening. Default soil model parameters were used for the static analysis, which were based on intact strength profiles. Figure 9 shows the range in V /p (upper plot) and w/D (lower plot) for static pipe lay across the sleeper on a non-linear intact seabed (Case 5). Also shown is the final force concentration and embedment. An © 2011 by Taylor & Francis Group, LLC

4.3

Dynamic pipe lay on non-linear seabed

Vessel-induced pipe motions result in higher transient contact forces, which (for the non-linear plastic seabed) lead to greater penetration as well as incremental pipe embedment with each cycle. A dynamic analysis (Case 7) was carried out by adding a regular (Dean stream) significant wave height of Hs = 2 m with a wave period of 13 seconds (to match the field conditions) to the static non-linear Case 5.The vessel and pipe payout advanced at 0.1 m/s, i.e. a lay rate of 360 m/hr. This lay rate corresponds to the time period between welding operations, i.e. the minimum number of cyclic pipe motions in the touchdown zone. The intact soil strength gradient of 20 kPa/m was adopted as the non-linear soil model accounts for

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zone, dictated by the lay rate, sea state, and vessel and pipeline dynamics. These analysis techniques can be used in pipeline design to determine the range of embedment likely to occur near sleepers, aiding assessment of the pipe-soil response during buckling. ACKNOWLEDGEMENTS

Figure 10. Variation in embedment for dynamic pipe lay on non-linear seabed (Case 7) compared to field survey data.

incremental penetration under cyclic motions. Inspection of individual nodes along the pipeline indicated that the force – penetration response reached near steady-state conditions within the 15 to 20 cycles of pipe motions to which it was subjected. The analysis is therefore not sensitive to the range of lay rate recorded during normal laying operations. This represents a touchdown zone length of about 23 m, which is realistic for this seabed, pipe and lay geometry. The non-linear soil model parameters were then optimised to match the shape of the field data profiles by reducing the non-dimensional maximum suction resistance factor and increasing the non-dimensional repenetration resistance depth factor (Randolph & Quiggin 2009). Figure 10 shows the final embedment profile from this analysis, which provides the best match to the field data. This example illustrates how soil non-linearity and the dynamic aspects of pipeline embedment can be simulated. 5

CONCLUSIONS

This study has presented a series of theoretical and numerical analyses showing the variation in pipe-soil contact force and the resulting pipeline embedment in the vicinity of sleeper crossings along offshore pipelines. The results have been compared to field survey embedment data for a pipeline installed at a soft clay site in deep water. The asymmetry within the pipe-soil contact force profiles across sleepers can only be captured realistically using non-linear seabed models that account for the loading history of the pipe-soil contact force during the sleeper crossing. Static pipe lay analysis using nonlinear seabed models and a remoulded shear strength to account for dynamic lay effects provided an embedment profile that generally matched the trends in the field data. A marginally closer match to field observations was obtained using an optimised dynamic pipe lay analysis that simulated incremental penetration due to cycles of vertical pipe movement in the touchdown


This work forms part of the activities of the Centre for Offshore Foundation Systems, established under the Australian Research Council’s Research Centres Program and now supported by the State Government of Western Australia through the Centre of Excellence in Science and Innovation program. The work was carried out while the primary author was an Endeavour International Postgraduate Research Scholar. The second author is supported by an ARC Federation Fellowship (grant FF0561473). The third author is supported by an ARC Future Fellowship (grant FT0991816). This study is part of a COFS-Acergy research collaboration. Orcina Ltd, UK, assisted in developing a model to simulate pipe lay in OrcaFlex. REFERENCES Lenci, S. & Callegari, M. 2005. Simple analytical models for the J-lay problem, Acta Mechanica, Vol. 178: 23–39. Lund, K.M. 2000. Effect of Increase in Pipeline Soil Penetration from Installation, Proc. ETCE/OMAE2000 Joint Conf., Paper OMAE2000-PIPE5047. Merifield, R., White, D.J. & Randolph, M.F. 2009. The effect of soil heave on the response of partially-embedded pipelines in clay, ASCE J. of Geotechnical and Geoenvironmental Eng., 135(6): 819–829. Orcina. 2009. OrcaFlex User Manual, Version 9.2e, www.orcina.com.uk. Palmer, A. 2008. Touchdown indentation of the seabed, Applied Ocean Research, 30(3): 235–238. Pesce, C.P., Aranha, J.A.P. & Martins, C.A. 1998. The soil rigidity effect in the touchdown boundary layer of a catenary riser: Static problem, Proc. 8th Int. Offshore and Polar Eng. Conf., 207–213. Randolph, M.F. & Quiggin, P. 2009. Non-linear hysteretic seabed model for catenary pipeline contact, Proc. 28thInt. Conf. on Offshore Mech. and Arctic Eng., Paper OMAE2009-79259. Randolph, M.F. & White, D.J. 2008. Pipeline Embedment in Deep Water: Quantification and Processes, Proc. Offshore Tech. Conf., Paper OTC19128. Sinclair, F., Carr, M., Bruton, D. & Farrant, T. 2009. Design challenges and experience with controlled lateral buckle initiation methods, Proc. 28th Int. Conf. on Offshore Mech. and Arctic Eng., Paper OMAE2009-79434. Westgate, Z.J., White, D.J. & Randolph, M.F. 2010. Pipeline laying and embedment in soft fine-grained soils: field observations and numerical simulations, Proc. Offshore Tech. Conf., Paper OTC20407.

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Observations of pipe-soil response from the first deep water deployment of the SMARTPIPE® D.J. White Advanced Geomechanics and Centre for Offshore Foundation Systems (COFS), Univ. of Western Australia

A.J. Hill BP Exploration, Sunbury-on-Thames, UK

Z.J. Westgate Advanced Geomechanics and Centre for Offshore Foundation Systems (COFS), Univ. of Western Australia

J-C. Ballard Fugro Engineers SA, Brussels, Belgium

ABSTRACT: The Fugro SMARTPIPE® is a new site investigation tool to measure pipe-soil interaction parameters in situ, at the seabed. It comprises a seabed frame with an instrumented model pipe that can be driven in the vertical, axial and lateral directions. This paper presents results from the first SMARTPIPE® campaign in deep water, focusing on axial pipe-soil interaction on soft clay. Pore pressure measurements on the surface of the test pipe provide data that allows the cyclic axial response to be interpreted in an effective stress framework. An effective stress failure criterion shows better agreement with the response across several axial sweeps than a total stress interpretation. In the example test presented here, transient positive excess pore pressure is generated when the pipe changes direction. This means that in this case, surprisingly, the full axial resistance is mobilised over a particular time rather than a particular distance.

1

INTRODUCTION

The Fugro SMARTPIPE® is a new site investigation tool that is designed to measure pipe-soil interaction parameters in situ, at the seabed. It comprises a seabed frame with an instrumented model pipe that can be driven in the vertical, axial and lateral directions whilst the corresponding loads are recorded. Descriptions of the original design of the SMARTPIPE® are given by Hill & Jacob (2008). A key motivation for the development of the SMARTPIPE® is the need to better quantify the interaction forces between on-bottom pipelines and the seabed in order to provide geotechnical input for the assessment of pipeline buckling and walking. The first offshore SMARTPIPE® field testing campaign was carried out at a deep water location during 2008. This paper describes some of the results from that campaign, focussing on axial pipe-soil interaction. Pipe-soil resistance forces depend on properties of both the soil (including the drainage conditions, and the relevant undrained or drained strength) and the pipe (including the embedment, diameter and surface coating). The SMARTPIPE® should be viewed as a model test. Through interpretation, this can yield information about the soil properties, and the response of a given pipe resting on that soil. © 2011 by Taylor & Francis Group, LLC

Just as the SMARTPIPE® is an evolving tool (the equipment has been enhanced since the campaign described here), so are the techniques to analyse, interpret and predict pipe-soil interaction. This paper provides insights into the best framework in which to interpret axial pipe-soil interaction. 2

DESCRIPTION OF THE SMARTPIPE®

The SMARTPIPE® can be deployed via a single lifting/communication cable, from vessels equipped with the equivalent of a stern-mounted 20 tonne capacity A-frame at least 5 m wide and 7 m tall. The maximum operational water depth is currently 2,500 m. Figure 1 shows a diagram of the SMARTPIPE®, as used in this deployment. The key dimensions of the model pipe are summarised in Table 1. The central measurement section of the SMARTPIPE® model pipe is attached to the main frame of the device by a pair of triaxial load cells, which indicate the vertical, axial and lateral loads applied to the measurement section (Figure 2). The measurement section is also equipped with a set of pore pressure transducers (PPTs) which record the pore water pressure at the surface of the pipe relative to a hydrostatic reference. There are five PPTs located along the invert of

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Table 2.

Instrumentation and nomenclature.

Instrument

Measurement and symbol

Triaxial pipe load cells

Vertical pipe-soil load (/unit length) Lateral pipe-soil load (/unit length) Axial pipe-soil load Vertical pipe movement 1 Lateral pipe movement Axial pipe movement Excess pore pressure on pipe surface (at 5 locations on invert, and 4 locations at 30◦ from invert)

Displacement transducers Pore pressure transducer

V H F w uh ua u

1

Figure 1. Schematic SMARTPIPE® frame. Table 1.

view

from

This value is corrected for any frame settlement based on a displacement transducer linked to a settlement plate resting on the seabed.

beneath

Key dimensions and properties of SMARTPIPE®.

Dimension

Value

Model pipe diameter Overall length of model pipe Length of central measurement section Surface coating

225 mm 1100 mm 776 mm Polypropylene

Five types of test were carried out in the SMARTPIPE® campaign. These were (i) cyclic T-bar penetration tests, (ii) vertical pipe penetration tests, (iii) pore pressure dissipation tests, (iv) axial pipe tests, and (v) lateral pipe tests. This paper focuses on the results from one particular axial pipe test, which serves to illustrate several important phenomena that control the resistance that can be mobilised during axial pipe movement. 3

Figure 2. Cutaway view of instrumented pipe showing load cells and pore pressure transducers.

Figure 3. Pore pressure transducers on pipe surface.

the model pipe, and at each end of the measurement section there are PPTs located 30 degrees around the circumference on each side (Figure 3). The instrumentation and the resulting measurements are summarised in Table 2. A hydraulic system provides the actuation of the model pipe. The vertical axis can operate in displacement-rate control or in load-controlled modes, although the load-controlled mode currently involves some manual intervention. © 2011 by Taylor & Francis Group, LLC

SITE CHARACTERISATION

The seabed conditions at the site comprise very soft high plasticity marine clay, with a shear strength profile that increased monotonically with depth. The liquid and plastic limits were 183% and 77% respectively, giving a plastic index of 106%. The SMARTPIPE® features a miniature T-bar penetrometer, which is 12 mm in diameter (although the bar can be interchanged with other sizes, if required). This device is smaller than the conventional T-bar penetrometer, which has a diameter of 40 mm, to provide better strength resolution close to mudline. At this site, the T-bar results showed that a layer of slurry, 50– 100 mm in thickness was present, but had negligible strength. Below the slurry, the intact soil strength rose from approximately zero at the soil-slurry interface with a gradient of ∼ 10 kPa/m over the first 0.3–0.4 m depth (which is the range relevant to the pipe tests) before rising at a lower, more usual rate, at greater depth. The remoulded strength, derived from cyclic T-bar tests, rose at 3 kPa/m from zero at the soil-slurry interface. 4

EXCESS PORE PRESSURE DISSIPATION

The first stage of each SMARTPIPE® test involved vertical penetration of the pipe into the seabed, followed by a period of excess pore pressure dissipation. The vertical penetration response is not presented here. However, the resistance recorded during the faster penetrations was consistent with the calculated resistance based on the strength profiles

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Table 3.

Cyclic axial test parameters.

Sweep

Length (mm)

Speed (mm/s)

1 2 3 4 5 6

+150 −200 +200 −200 +250 −250

0.04 mm/s

0.15 mm/s

Figure 4. Excess pore pressure dissipation after vertical penetration: comparison of SMARTPIPE® data with FE results.

derived from the T-bar tests, and the solutions presented by Randolph & White (2008), based on the analysis of Merifield et al. (2009). Slower rates of embedment led to reduced resistance, which can be attributed either to concurrent consolidation settlements or the effect of strain rate on soil strength. After penetration to a specified depth, the vertical load was held constant (or near-constant), typically for 4 hours, and the dissipation of excess pore pressure at the various PPT locations was monitored. The results from one particular test, which involved the highest and most steady vertical load level and therefore the largest and most consistent pore pressure response, have been compared with the numerical solutions presented by Krost et al. (2010) and Gourvenec & White (2010). These solutions are based on finite element (FE) analyses adopting an elastic soil model with coupled consolidation. Analyses with and without soil heave around the pipe were performed, showing minimal effect (Gourvenec & White 2010). The average excess pore pressure measured at the pipe invert shows excellent agreement with these solutions, where the numerical results are scaled according to a coefficient of consolidation of cv = 75 m2 /year (Figure 4). The good agreement throughout the dissipation period gives confidence in this back-analysis, which indicates a value of cv that is 200 times higher than values inferred from reconstituted samples of the same soil tested in the laboratory. This observation has implications for assessments of the drainage condition during pipe movements and other consolidation events at this site. One key benefit of the SMARTPIPE® tool is that the consolidation characteristics of the surficial soil can be measured in situ – in this case leading to a significant reassessment of this design soil property.

5

CYCLIC AXIAL PIPE TEST

5.1 Overall forces and displacements The cyclic axial pipe test report here was conducted at an embedment of 0.65D, over a period of 7 hours. The vertical load on the pipe was maintained in the range 1–1.2 kN/m throughout most of the test, which consisted of six axial sweeps with the lengths and rates of movement given in Table 3. © 2011 by Taylor & Francis Group, LLC

Figure 5. Time history during cyclic axial pipe test.

The time histories of imposed vertical load, imposed axial displacement and measured axial resistance are shown in Figure 5. The vertical load gradually decayed over the test, and a small cyclic component of vertical load (∼5% of the steady value) was also present, due to a slight misalignment in the drive spindle – which is evident in the load-displacement response (Figure 6). The axial response included a modest peak in resistance during the first sweep, and a ductile response during all subsequent sweeps (Figure 6). 5.2

Interpretation of effective stresses

The nine PPTs allow an effective stress interpretation of the pipe-soil response to be performed, although some assumptions must be made, in order to link the applied loads (which act around the entire pipe-soil contact surface) to the measured pore pressures (which are predominantly at the invert). We have adopted averaged quantities, using the inadvertent ‘wobble’ in the SMARTPIPE® spindle to calibrate between the invert and average pore pressures. The total load on the pipe-soil surface, N, is the applied vertical load, V, multiplied by a ‘wedging factor, taken here as ζ = 1.27 (White & Randolph 2007). The wedging effect arises because the sum of the normal forces on the curved pipe-soil interface exceeds the vertical force. The mean total normal stress is then found as σn = N/LP where P is the contact perimeter. A value of πD/2 is adopted for P: although the embedment is slightly greater than 0.5D, the soil was observed, via a camera mounted on the seabed

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Figure 6. Load-displacement response in cyclic axial pipe test.

Figure 8. Effective stress interpretation: response during change of direction.

5.3 Total and effective stress failure criteria Figure 7. Mean measured pore pressure, total stress and effective stress response during cyclic axial pipe test.

frame, to stand clear of the pipe above the centreline. The mean total normal stress varies between approximately 4.5 and 3.5 kPa, showing a general decrease throughout the test. The mean measured excess pore pressure, uav,meas , based on 7 PPTs, of which 4 were located at the pipe invert, also decreased over the test period (2 PPTs were inoperative). Superimposed on this slow trend are sharp variations associated with changes of pipe direction (Figure 7). Since the mean measured pore pressure is based predominantly on data from the pipe invert, this value is likely to be an over-estimate of the mean pore pressure around the full pipe-soil contact width – which extends to the soil surface, where the excess pore pressure must be zero. The small cyclic load component proved to be useful in allowing an adjustment factor to be derived to account for this effect. It was found that if the mean measured pore pressure, uav,meas , is scaled by 80%, then the cyclic component is minimized within the inferred mean effective stress, σn = σn – 0.8 uav,meas – as shown in the short period of data from 1–1.4 hours, close to a change in direction (Figure 8). This adjustment is based on an assumption that the small cyclic load is undrained and causes no changes in effective stress. A reversal of the axial movement results in a sharp rise in the mean pore pressure, and a corresponding reduction in the effective pipe-soil stress (Figure 8). In this case σn falls from ∼3 kPa to ∼1 kPa, for about 0.15 hours (10 minutes) before steady conditions are re-established. This effect is linked to the mobilisation of axial resistance in Section 5.4. © 2011 by Taylor & Francis Group, LLC

Having inferred the variation mean total and effective stresses at the pipe soil surface, the data from all six sweeps can be compared via both total stress and effective stress failure criteria. The mean shear stress on the pipe surface is calculated as τav = H/(πD/2). In pipeline design, it is common to consider a friction factor defined based on the total forces: FF = H/V. The equivalent effective stress quantity is FF = τav /σn,av . The FF response is smoother than the FF response. This is partly from the influence of the PPT data, which showed some small variations between sensors, indicating that the excess pore pressure was non-uniformly distributed (Figure 9). However, when these responses are plotted in the form of total and effective stress failure criteria, the effective stress interpretation is the most consistent. A total stress failure criterion for shearing on a prescribed interface takes the form of a limiting shear stress, independent of the imposed total normal stress. The data from the six sweeps shows that the steady limiting shear stress spanned the range 1–1.5 kPa, with cycle 2 showing a higher strength than cycles 1 and 3 (Figure 10). In an effective stress interpretation – σn vs. τav – all three sweeps in each direction overlie each other (Figure 11). The first sweep shows a peak in resistance, to a stress ratio of τav /σn = 0.81, suggesting an initial peak in the mobilized friction angle. The remaining data lie close to the effective stress failure criterion of White & Randolph (2007):

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with fitting parameters of A = −0.4 and B = 0.6. This failure envelope and the axial test data are compared to results from a set of model pipe tests conducted at the Norwegian Geotechnical Institute

Figure 12. Comparison of effective stress failure response with data from other sources.

5.4

Figure 9. Friction factor response during cyclic axial pipe test.

Figure 10. Total stress interpretation of axial response.

Figure 11. Effective stress interpretation of axial response.

using the same clay (Figure 12). These tests were back-analysed in the same manner, using measured pore pressures and a wedging factor to derive effective stress quantities. Good agreement is evident, with the SMARTPIPE® data falling amongst the NGI data, towards the lower bound of the envelope. When the NGI results are interpreted in a total stress manner, the friction factor varies by a factor of 4 (Bruton et al. 2009). The effective stress approach gives closer bounds, and shows that a lower total friction factor is associated with high excess pore pressures around the pipe surface. Also shown on Figure 12 are values of friction angle derived from tilt table tests using the same clay. These also agree closely with the model test data. © 2011 by Taylor & Francis Group, LLC

Mobilisation: A function of distance or time?

The stress and pore pressure responses during the initial 60 mm of each axial sweep are shown in Figure 13. This figure also illustrates the extent of the ‘wobble’ of the spindle mentioned earlier, showing that it is evident as cyclic “noise” and does not mask the overall trend of the data. At every reversal there is an increase in pore pressure during the initial movement, resulting in a corresponding reduction in the effective stress. The pore pressure then recovers close to the initial value, as does the effective stress. The exception to this behaviour is the first sweep, in which the total stress reduces over the first 30 mm of movement, causing a corresponding reduction in the effective stress. In these tests the generation and recovery of excess pore pressure is strongly influenced by the rate of pipe movement. During the first two cycles, the pore pressure transient takes place over ∼20 mm of movement, whereas during the final cycle (which took place at a higher pipe velocity) the transient is spread over ∼60 mm. The same behaviour is evident in the shear response. Figure 14a shows the mobilisation of axial shear stress during the initial 60 mm of each sweep with the negative values of resistance retained for clarity. The general shape of each response is consistent, although in some cases there is an initial residual shear stress from the previous sweep. The exceptional case is the first sweep, which shows a brittle peak. The faster sweeps show a more compliant response, and a higher mobilisation distance. The shape of the mobilisation response in the different sweeps is better compared by considering the normalised shear resistance, S, defined as:

where τinit is the initial residual shear stress and τ60 is the fully-mobilised resistance at 60 mm displacement. When plotted in this manner, the responses at each speed overlie each other closely, with the exception of the brittle first sweep response. The effect of velocity is captured by relating the normalised shear stress to the time since the reversal point, rather than the distance. When plotted in this

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transient observed over the same period of time after each reversal. This mechanism adds complexity to the selection of a mobilisation distance for design, if pipe movements occur at an undrained or partially-drained rate, creating excess pore pressure. 6

CONCLUSIONS

Results from the first field deployment of the SMARTPIPE® have been presented. The resulting data provides new insights into the mechanisms of pipe-soil interaction, with the benefit of having been gathered in situ, in deep water, on intact soil. A cyclic axial test has been back-analysed using an effective stress frictional failure criterion, which provides a more consistent interpretation than a total stress approach. However, the generation of excess pore pressure means that the pipeline friction factor for a given soil varies if movements occur at an undrained or partially-drained rate (or when lay-induced pore pressures remain). The importance of pore pressures is highlighted by evidence that, in this case, the full pipe-soil resistance is only mobilised when steady pore pressures are reached, which is after a fixed time, rather than a fixed displacement.

Figure 13. Detailed stress-pore pressure response during the initial 60 mm of each sweep.

ACKNOWLEDGEMENTS The authors wish to thank BP Exploration for permission to publish these data. The SMARTPIPE® was developed by Fugro with support from BP. REFERENCES

Figure 14. Shear stress mobilization during the initial 60 mm of each sweep.

Figure 15. Shear stress mobilization with elapsed time.

way, all of the mobilisation curves overlie each other, except for the first breakout (Figure 15). This remarkable agreement suggests that in this case the mobilisation of axial resistance is a time-related process, rather than being linked to the distance of shearing. This is consistent with the pore pressure © 2011 by Taylor & Francis Group, LLC

Bruton D., White D.J., Langford, T.L. & Hill A. 2009. Techniques for the assessment of pipe-soil interaction forces for future deepwater developments Proc. Offshore Technology Conference, Houston, USA. Paper OTC20096, 12pp. Gourvenec S.M. & White D.J. 2010. Elastic solutions for consolidation around seabed pipelines. Proc. Offshore Technology Conference, Houston. Paper 20554, 16pp. Hill, A.J. & Jacob, H. 2008. In-Situ Measurement of PipeSoil Interaction in Deep Water. Proc. Offshore Technology Conference, Houston, USA. Paper OTC 19528, 18pp. Krost K., Gourvenec S.M. & White D.J. 2009. Consolidation around partially-embedded submarine pipelines. Géotechnique, Accepted 19 January 2010, in press. Merifield, R., White, D.J. & Randolph, M.F. 2009. Effect of surface heave on response of partially embedded pipelines on clay. ASCE J Geotech..& Geoenv. Eng. 135(6):819– 829. Randolph, M.F. & White, D.J., 2008. Pipeline Embedment in Deep Water: Processes and Quantitative Assessment, Proc. Offshore Technology Conference, OTC 19128, 16 pp. White D.J. & Randolph M.F. 2007. Seabed characterisation and models for pipeline-soil interaction, Int. Journal of Offshore & Polar Engineering. 17(3), pp.193–204.

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11 Trenching, ploughing, excavation and burial



Influence of object geometry on penetration into the seabed A. Ivanovi´c, R.D. Neilson & G. Giuliani University of Aberdeen, Aberdeen, UK

M.F. Bransby University of Dundee, Dundee, UK

ABSTRACT: The seabed is disturbed by predominently horizontal cutting mechanisms during the installation of pipelines (e.g. by ploughs), by fishing gear components and by natural processes (e.g. iceberg scour). In each case the zone of soil disturbed by the device and/or the soil resistance force to the movement of the object is of interest to the geotechnical engineer. Consequently, this paper reports an experimental investigation of how different shaped objects penetrate the seabed. The results show that the penetration is influenced by both the front face angle and the weight of the object while the drag force is mostly influenced by the weight.

1

INTRODUCTION

Cutting processes that commonly disturb the seabed and have direct implications for the offshore industry are iceberg scour, anchor dragging, trawling and the cutting and ploughing processes associated with the installation of pipelines and cables. The first three processes are out of the control of an engineer and if contact of any of these is made with a pipeline this can cause damage. Pipeline burial to beneath the affected zone is the primary form of mitigation. Ploughing is a more controlled process, which is used during pipeline installation. Both iceberg scour and ploughing processes disturb the mechanical state of the soil as a rigid body penetrates horizontally through the seabed near the surface. These processes are examined in this paper by means of physical modelling. 1.1 Ice gouging The challenge that offshore pipeline engineers have been faced with in recent years is the estimation of the most economical burial depth for a pipeline, at which it remains safe and in a good condition. During the movement of an iceberg it has been reported by various studies (e.g. Palmer et al., 1990; Woodworth-Lynas et al., 1996) that the soil below a scouring ice keel will displace laterally, in the direction of ice movement, transverse to the ice movement and vertically. The scouring process will produce soil deformations, which can produce strains and stresses that can be transferred to any nearby buried pipeline, which in turn could be damaged. The seabed affected by an ice scour process is described generally through three zones (Palmer et al., 1990). Zone 1 is the top layer soil that is gouged by ice – there is direct contact with the ice and consequently, there is a high probability © 2011 by Taylor & Francis Group, LLC

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Figure 1. The geometry of the penetrating object.

that any pipeline within this zone would be subjected to damage. Zone 2 is situated below the level of the base of the ice keel and is subjected to large plastic soil deformations. The magnitude of these soil deformations and therefore the depth of zone 2 are difficult to estimate because they are dependent on many factors such as the type of soil and the shape of the ice feature causing the ice scour. Finally, Zone 3 is where the soil deformation is limited to elastic (or at least small displacement) behaviour. Consequently, the safe burial depth is usually estimated to be within Zone 3. Since this may be very deep for large icebergs it may not be necessarily the most economical solution. The optimal burial depth therefore tends to be located in Zone 2 where the pipeline can perform as designed but without being damaged (Fig. 1). The pipeline does not have the same capability as the seabed to resist the ice contact forces. The challenge for the pipeline designer is to determine the safe burial depth of pipeline in order to avoid potential damage to the pipeline if the ice scour path coincides with the position of the pipeline. It is very difficult to define exactly ice scour scenarios because they depend on environmental forces, such as wind (in shallow water) and the current, as well as the seabed conditions. This

Figure 2. The geometry of the penetrating object.

explains why ice gouging poses a threat to submarine pipelines in arctic areas. The influence of iceberg scour on pipelines has been considered by Barrette & Timco (2008) where small scale laboratory experiments were undertaken in a 6 m long sand channel using real ice models. The research confirmed the presence of the three zones identified previously. 1.2

Figure 3. Sand channel used to contain the soil sample at the University of Aberdeen.

Ploughing

Pipelines are often protected by burial. One common method of burial is by ploughing, where a large (e.g. 20 m long) plough is towed along the seabed to form a trench (e.g. Palmer et al., 1979). The pipeline is placed in the open trench and the trench is then backfilled. One important aspect of the operation is the tow force from the support vessel required to progress the plough and so semi-empirical relationships have been developed to predict how this force varies with the weight of the plough, the depth of trench formed and the velocity (e.g. Reece & Grinsted, 1986; Cathie & Wintgens, 2001). Like an iceberg, the plough share (which performs the main ‘cut’) penetrates the seabed and moves horizontally during steady-state ploughing. However, unlike an iceberg, the face of the share is inclined backwards (i.e. with β < 90◦ in Fig. 2). 1.3 Aim of study This study of ice scouring and ploughing will focus on how the mechanical properties of the soil and the shape of the penetrating object affect the trenching depth, drag force and the subsurface deformations.The geometry considered is shown in Figure 2. The object moves horizontally through the seabed and cuts a trench of depth, D, measured from the original surface, requiring a force, F to move. The key characteristic of the geometry is the angle between the base and the front face of the object, β (Fig. 2). In terms of the soil resistance, F, to movement, the most applicable theory is that of retaining walls where the passive earth pressure may be considered to act in front of the object (and depend on D and the width of the object, L) with a base shear resistance (depending on the base roughness and self weight, W) also acting against the direction of movement. © 2011 by Taylor & Francis Group, LLC

2

EXPERIMENTAL METHOD

2.1 Test samples Laboratory experiments were conducted in which the penetration behaviour of objects of different shapes were examined. The objects were tested in a 4.8 m long and 50 cm wide channel in which the penetration of the object into the sediment, and the towing force were measured (Fig. 3). This channel has rails that support a trolley, to which the component to be tested is attached. The trolley allows the component to move freely in the vertical direction under a fixed vertically applied load, W, and includes a load cell which measures the drag force, F, and a LinearVariable DifferentialTransformer (LVDT) that measures the penetration of the component into the sediment, D, as the object was towed along the channel at constant speed by a winch mechanism. Potential rotational motions were restrained but reaction moments were not measured. Uniform grading, dry silica sand (rounded particles; D50 = 144 µm; φcrit = 32◦ etc.) was used in all tests to form the seabed sample. This was prepared by raking and leveling the sand between the tests to produce a sample with a unit weight of 15.6 kN/m3 (Dr = 25.6%). All samples were prepared dry to ensure fully drained conditions were generated during the cutting process for the preliminary experiments reported here. The test pieces were constructed using steel boxes with base dimensions, B = 150 mm and L = 150 mm where the angle of the front face, β, was varied from 60◦ to 150◦ as shown in Figure 4. An additional box with the front face at 90◦ and with a curved edge of

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Figure 5. Drag force vs. penetration into the seabed for the 90◦ box towed at 0.1 m/s and 0.5 m/s (W = 50 N).

Figure 4. The steel boxes used as the penetrating objects.

15 mm radius was constructed to investigate how the curved edge affected its performance during seabed penetration. Each component was towed at a constant rate along the sediment placed in the tank and the drag force and the penetration measured. The tests show how the drag forces and soil displacements are affected by the shape and penetration depth. The test pieces were dragged at two different speeds, 0.1 and 0.5 m/s, and different masses were added to the original weight of the boxes themselves to give total weights of 50, 100 or 150 N during the tests.

3

Figure 6. Drag force vs. penetration t into the seabed for the 90◦ box towed at 0.1 m/s and 0.5 m/s (W = 100 N).

RESULTS

3.1 Effect of velocity on drag force The influence of the speed at which the boxes were towed (0.1 and 0.5 m/s) was examined first. Figures 5–7 show the force-penetration relationship for different weights for the box with the front face at 90◦ . In general it can be seen that the drag force increases with an increase in penetration which would be expected from retaining wall theory. For the three different weights taken together, there is no consistent difference in the F-D relationships for different velocities. For dry sand the drag force would not be expected to change with velocity for reasons of pore fluid movement, although differences might be expected at much higher rates because of dynamic inertial effects (i.e. local accelerations of soil around the fast-moving object as new soil is pushed away by the object). The variable forces shown for the lighter boxes towed at higher speeds are due to build up of a heap of sand in front of the test piece. The heap then spills to the side, reducing in height, and then the process repeats. This oscillation becomes less prominent as steady state conditions are achieved. © 2011 by Taylor & Francis Group, LLC

Figure 7. Drag force vs. penetration for box with β = 90◦ , towed at 0.1 and 0.5 m/s (W = 150 N).

3.2

Effect of geometry and weight on penetration and drag force

A series of tests were conducted in which each box shape (defined by angle β) was tested separately with total weights of 50, 100 or 150 N. Near steady-state conditions were generally achieved after towing the object between 2 and 3 m along the channel (see Figure 8). The results presented in Figures 9 and 10 show the penetrations and the drag forces for different weight and speed scenarios at steady-state conditions.

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Figure 8. Vertical penetration against horizontal displacement along the seabed for the boxes with different front face angles (v = 0.1 m/s, W ≈ 150 N).

Figure 10. Steady-state drag force against front face angle for different object weights.

Figure 9. Steady-state penetration against front face angle for different object weights.

Figure 8 shows that for an approximately fixed W, the highest penetration is obtained when β = 60◦ which represents a shape applicable for ploughing applications. In contrast, the smallest penetration was noted for β = 150◦ , applicable to the ice scouring problem. The latter shape reaches the steady state penetration after a horizontal displacement of 0.4 m while for β = 60◦ , steady-state behaviour is obtained after a displacement of between 1.5 and 2 m. There is a clear trend of reduction in penetration with an increase in front face angle, β as shown in Figures 9a and 9b. Drag force however does not change as significantly with front face angle (Fig. 10), although there is a small reduction in drag force with increasing β. Given that penetration is reducing significantly, this suggests that the drag force for similar penetrations will be larger as β increases, and this will be investigated later. © 2011 by Taylor & Francis Group, LLC

Figure 11. Drag force vs. penetration relatiosnhip for 105◦ box towed along the seabed at 0.1 m/s with different weights.

There is a general trend of increasing penetration depth (Fig. 9) and drag force (Fig. 10) with increasing weight, W, for all β which would be expected. It is however noted that the results for β = 105◦ do not quite follow the general trend of the penetration data. Figures 11 and 12 show how the drag forces vary with object penetration during tests with β = 105◦ and 150◦ respectively. Each graph shows results for the three weight conditions. There is clearly a substantial drag component at zero penetration, which would be expected to be due to base shear. The interface friction coefficient, H/V = µ = tan δ was estimated by plotting the drag versus normal force at D = 0 and taking the gradient of the resulting plot. The drag force values for D = 0 used in this process were found by curve fitting the graph of drag force

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Table 1. Passive coefficient, Kp values for different angles, β.

Figure 12. Drag force vs. penetration relatiosnhip for 150◦ box towed along the seabed at 0.1 m/s with different weights.

Figure 13. Drag force vs. weight relationship for estimation of interface friction coefficient.

against penetration and identifying the intercept at D = 0. This analysis revealed that the resistance at zero penetration is described by F ≈ 0.31 W as shown in Figure 13. This implies a friction angle, δ = 17.2◦ for the smooth, rolled steel-sand interface. For larger penetrations (Fig. 11 and 12), the drag force, F, increases with depth of penetration. This drag term is likely to be due to the frontal resistance and side shear resistance, which are depth dependent. Similar results were reported by Phillips et al. (2005) for iceberg scouring and by Reece & Grinsted (1986) for ploughing. The drag force – penetration relationship might be expected to vary as:

where Kp = the passive earth pressure coefficient, D = penetration depth below the original ground surface and L = frontal width. Curve fits using Eq. 1 with δ = 17.2◦ , γ = 16 kN/m3 and Kp = 20 for β = 105◦ and Kp = 40 for β = 150◦ are shown in Figures 11 and 12 alongside the original test results. In order to achieve the best fit the value of Kp was adjusted for each front face angle β while keeping the measured interface friction angle and the soil unit weight constant. The selected values of Kp are shown in Table 1. The data show that Kp increases with front © 2011 by Taylor & Francis Group, LLC

Front face angle, β (degrees)

Kp

60 90 105 120 150

10 12 20 30 40

face angle β reflecting the increasing lateral and side pressures for blunter front face angles (larger β). The difference in Kp values to fit the data as front face angle changes, reflect the quite different flow patterns around the 60◦ and 150◦ shapes, which give the large difference. When β = 150◦ the object will push forward and compress vertically the sand in front causing a partial bearing capacity failure (e.g. see Palmer et al. 1990 for clay conditions). This will cause a large vertical reaction force which will prevent significant penetration into the seabed without large W and increase local effective stresses and normal contact forces and thereby reaction forces. In contrast the object with β = 60◦ has a face inclined with an upward normal component. The sand will flow over the front face (causing a small additional vertical downward load onto the base) and then slide to the sides with little vertical restraint. This may account for the reduced drag, despite the greater penetration and the variation of Kp with β. Finally, it should be noted that the Kp values backcalculated (Table 1) did not take into account the build up of sand in front of the object. Consideration of the spoil heaps would reduce the effective Kp values used in calculations, but require additional knowledge of the changing spoil heap size to calculate penetration resistance as penetration and translation varied. This will be studied in future experiments. An additional experiment was undertaken to examine soil deformation mechanisms by pulling a test piece with β = 150◦ adjacent to a Perspex box side with the same loading conditions as used in the main test series. Multiple digital images (e.g. Fig. 14) were captured of the process and analysed using geoPIV (White et al., 2003) to measure soil displacements during penetration. Figure 14 clearly shows the build up of sand in front of the object during steady-state (i.e. horizontal) penetration. Figure 15 shows the calculated instantaneous soil movements which occurred during penetration of the 150◦ test piece. The soil beneath shows downward components of movement, which generate the upward reaction force, which reduces penetration and increases drag force. 3.3

Influence of the curvature of front edge of the object

The data for 90c compared to 90 is interesting as there is a large difference in penetration (e.g. 55%

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Figure 14. Digital image of object penetration at steady-state in apparatus at the University of Dundee (β = 150◦ ; W /BL = 3.6 kPa).

The results show that the front face angle of the penetrating object and the vertical load applied to it significantly affect both the steady-state drag force and, to an even greater extent, the penetration depth through the soil. The more acute the angle between the base and the front face, the greater is the penetration. The practical implication of this finding is that if the geometry of a typical iceberg is known then it may be possible to estimate the required burial depth of pipelines. Furthermore, the results of this study indicate that by introducing a curved front edge, the penetration reduces but the drag force does not drop proportionally. This suggests that an object with a curved edge is likely to cause less damage to a buried pipeline than one with a sharp edge. ACKNOWLEDGEMENTS The authors would like to acknowledge the assistance of Steven Boertien (University of Dundee) who conducted the test shown in Figure 14. REFERENCES

Figure 15. Incremental soil displacements for object with β = 150◦ before steady-state penetration (D/B = 0.28; Scale factor = 3).

reduction for the condition shown in Fig. 8), which is not reflected in the drag force for higher speeds. The difference in penetration depth is due to the curved frontal edge which allows the sand to flow under the object rather than the edge acting as a cutting tool, reducing penetration depth for the object with the curved edge. A similar effect has been found while undertaking experiments on trawl otter doors. Whether such differences occur for the cases for β > 90◦ (for example for icebergs with abraded, rounded edges) requires further study. 4

CONCLUSIONS

The work presented in this paper focuses on the influence of iceberg and ploughing objects on the seabed. An experimental rig has been set up and boxes of different geometries tested.


Barrette, P. & Timco G. (2008) Ice scouring in a large flume: Test set-up and preliminary observations ICETECH08133-RF. Cathie, D.N. & Wintgens, J-F. (2001). Pipeline trenching using plows: performance and geotechnical hazards. Proc. Offshore Technology Conference OTC 13145, Houston, May 2001. Palmer, A. C., Kenny, J. P., Perera, M. R. & Reece, A. R. (1979). “Design and operation of an underwater pipeline trenching plough” Géotechnique 29 (3): 305–322. Palmer, A.C., Konuk, I., Comforg, G. & Been, K. (1990). Ice gouging and the safety of marine pipelines. Proc. Offshore Technology Conference OTC 6371, Houston, May 1990. Phillips, R., Clark, J.I. & Kenny, S. (2005) PRISE studies on gouge forces and subgouge deformations. Procedeeings of the 18th International Conference on Port and Ocean Engineering under Artic Conditions (POAC), Potsdam, USA, Vol. 1 pp 75–84. Reece, A. R. & Grinsted, T. W. (1986). Soil Mechanics of Submarine Ploughs. Proc. Eighteenth Annual Offshore Technology Conf., Houston (5341) May 1986, 453–461. White, D. J.,Take, W.A. & Bolton, M.D. (2003). Soil deformation measurement using particle image velocimetry (PIV) and photogrammetry. Géotechnique 53(7): 619–631. Woodworth-Lynas, C.M.L., Nixon, J.D., Phillips, R. & Palmer, A.C. (1996) Subgouge deformations and the security of Arctic marine pipelines. Paper OTC8222 in Proceedings, Twenty-eighth Annual Offshore Technology Conference, Houston, 4 657.

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Investigation into the effect of forecutters on plough performance K.D. Lauder, M.J. Brown & M.F. Bransby University of Dundee, Dundee, UK

J. Pyrah CTC Marine Projects Ltd, Darlington, UK

ABSTRACT: Pipeline ploughing is a common method used to bury offshore pipelines for protection. During ploughing, rate effects which occur in fine grained granular soils increase the tow force required by the support vessel and may reduce achievable plough velocities, thereby increasing the duration of a contract. The share of the plough ‘cuts’ the trench with some ploughs featuring a second, smaller cutting tool known as a forecutter which sits in front of the share. The effectiveness of the forecutter in reducing tow forces during ploughing has been investigated by reduced-scale model testing and is outlined herein. The forecutter is shown to be beneficial in reducing the rate effect but has a negative impact on the ‘static’ component of tow force. Conditions where the forecutter should be of overall benefit to reducing the tow force during ploughing are described.

1

INTRODUCTION

1.1 Pipeline burial Offshore pipelines are commonly buried around 1.5 metres below the sea bed surface. Burial is used as a means of protection from fishing activities and hydrodynamic loading and if backfilled can prevent movements due to thermal expansion on commissioning of the pipeline. A pipeline plough, towed along the sea bed by a support vessel is one of the most common means by which pipelines are buried. Ploughs use a large share (see Fig. 1) to ‘cut’ a trench and often the pipeline runs through the plough and is laid directly into the trench as it is formed. It is important that a minimum cover to the pipeline is maintained, along with a relatively constant trench profile to maintain support, which can be particularly tricky when ploughing through geohazards such as sandwaves, (see Bransby et al. 2010a, b). For commercial reasons all these aims need to be achieved in the shortest possible time period.

Figure 1. Components of a typical pipeline plough.

The rate effect is therefore very important as it may control the velocity at which the plough can be towed. In addition to the share, a second smaller cutting blade, known as a forecutter (see Fig. 1) may be attached in front of the share and used as a means to reduce the rate effect. Although literature does exist on the merits of multiblade ploughing (e.g Hata, 1979), there is limited accessible general information available to make an informed decision as to whether a forecutter actually reduces rate effects during ploughing. The work outlined in this paper is a step towards a better appreciation of the forecutter’s effect on plough performance.

1.2 Rate effects Although ploughs are used widely by industry to bury pipelines, the soil mechanics occurring during ploughing are not wholly understood. When ploughing in sands and silts, the rate of ploughing is such that it causes partial drainage of the soil as it shears and in turn causes a rate effect whereby the tow force increases with plough velocity. The support vessels which provide the forward thrust to the plough have a limited capacity and so the velocity achieved may be limited by the tow force. © 2011 by Taylor & Francis Group, LLC

2 2.1

EXPERIMENTAL METHODS Introduction

Scale model testing was used to investigate the significance of a forecutter on plough performance. This approach allowed for sand conditions and the surface of each test bed to be prepared in a highly repeatable manner. A 50th scale pipeline plough was manufactured with a detachable forecutter. This allowed for

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Table 1. depth.

Figure 2. Schematic of experimental apparatus.

Numbered skid settings and the resulting plough

Skid setting

Arm angle (deg)

Mean plough depth (mm)

1 2 3 4 5

67 50 44 30 17

19 26 30 36 41

a normal effective stress range of 3–26 kPa. Standard laboratory tests gave Gs = 2.63, ρmax = 19.48 kN/m3 and ρmin = 14.33 kN/m3 . 2.4

Sample preparation

The dry sand test beds were prepared by a two stage process of stirring the sand in the test tank to create a uniform density followed by removal of the uneven surface by scraping with a flat edge. The saturated test beds were prepared by first saturating the sand and then stirring and scraping the surface flat.The resulting densities were a loose state (Dr = 32%) in dry sand and medium dense state (Dr = 53%) in saturated sand. Figure 3. Comparison of the generation of steady state values.

3 a comparison of plough performance during ploughing tests both with and without a forecutter. Because all lengths scale by a factor of 50, the normal effective stresses during model testing are 50 times smaller than at prototype scale, with associated implications on dilation as described by Bolton (1986). 2.2 Apparatus and procedure The apparatus used to perform the ploughing tests is shown in Figure 2. In preparation for tests where the plough was to be used without a forecutter it was ballasted to keep the plough mass and its centre of gravity the same as when the forecutter was used. Before each test the plough was placed onto the sand at the end of the tank (left hand side in Fig. 2). The trolley was then moved into position above the plough and a load cell of 200 N capacity and a 250 mm stroke LVDT were attached to the plough. The tank is 0.5 m wide by 0.5 m deep and is 2 m long, which allowed the 250 mm long plough enough travel to find its steady state depth in the sand according to the skid settings and provide data at the desired trenching depth. For a description of how a plough maintains its depth by moment equilibrium, see Palmer (1979) and Lauder et al., (2008). 2.3 Material properties All of the tests described in this paper were conducted in poorly graded fine Congleton sand with D60 = 0.15 mm and D10 = 0.10 mm. Shear box tests revealed ϕcrit = 31◦ and for Dr = 53%, ϕmax = 37◦ over © 2011 by Taylor & Francis Group, LLC

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3.1

RESULTS Establishing steady-state values

Figure 3 shows the plough sits on the surface of the sand (depth ≈ 0 mm) at the start of the test and by around 600 mm displacement it has achieved a constant depth known here as steady state. Steady state values are the value of tow force, plough depth and pitch when the plough is in dynamic equilibrium, i.e. it has reached its optimum depth according to its skid settings and from this point onwards the tow force, depth and pitch (defined by θ in Fig. 1) all show constant values with increasing plough displacement. Figure 3 compares results from two tests in loose dry sand: one with a forecutter and one without. Although the skid settings are the same for both tests the plough trenched slightly deeper without a forecutter than with, which was due the forecutter causing the plough to pitch forward slightly more than without. Comparing the tow force generated during the two tests it is clear that the forecutter causes a slight increase of tow force in dry sand even though it caused the plough to trench slightly shallower. 3.2 The forecutter’s influence on ploughing depth 50th scale ploughing tests were conducted in dry sand at a range of model depths from 19 mm–41 mm, which when scaled up by 50 times correspond to prototype scale depths of 0.95 m–2.05 m. Table 1 shows the skid settings used during the following tests to dictate the plough’s depth. The depth

Figure 4. Depiction of arm angle, defined as the angle of the arm relative to the heel of the share.

Figure 6. Relationship between plough depth and tow force in dry loose sand.

3.3 The forecutter’s influence on tow force in dry sand Figure 5. The change in plough pitch and depth due to the forecutter.

Cathie & Wintgens (2001) advanced earlier work by Reece and Grinsted (1986) and found that the ‘static’ tow force generated could be broken into two components, (see Eq. 1).

associated with each skid setting is the average plough depth expected for each skid setting and varies slightly from test to test. Arm angle is defined in Figure 4 as the angle between the heel of the share and the arm which supports the skids. Figure 4 shows two different arm angles of 17◦ and 44◦ respectively and highlights how arm angle influences the vertical distance between the skids and share. Figure 5 shows the effect of the forecutter on pitch and depth where aft pitching is described as being positive and forward pitching being negative. Each data point represents the average steady state value (as described in the previous section) of depth and pitch for a particular test. The data points that have been grouped together (circumscribed) are from tests carried out with the same skid settings, the exact setting (see Table 1 for reference) is indicated by the number which the oval also surrounds. The data in Figure 5 shows that the plough’s pitch will change significantly with depth and ranges from 0.15◦ at 19 mm depth to −1.2◦ at 41 mm depth. The difference between the tests with a forecutter and those without is only marginal but the forecutter does tend to make the plough pitch forward slightly more than it would without. This results in a small reduction in the plough depth for any particular skid setting. Although the difference between the depth of the plough when a forecutter is used and when not is only small, (around 2%) it should still be considered when examining the forecutter’s effect on the tow force generated during ploughing.

where Cw is a constant depending on interface friction between the plough and sand, W is the weight of the plough, z is the cutting depth, Cs is a constant dependent on relative density and γ is the unit weight of the sand. The first term in Eq. 1 is a base friction term which is the product of the plough’s self weight and the coefficient of friction between the plough and the surrounding sand. The weight of the 50th scale model plough is 16 N which is equivalent to 1/503 the weight of prototype weight. The interface angle between the plough and fine Congleton sand is 26◦ and the interface friction force is the product of the plough’s weight and the tangent of the interface angle and is 8 N. Figure 6 shows the tow force generated during the shallowest test conducted, plough depth, D = 19 mm model scale (or 50 × 19 mm = 0.95 m prototype scale) is only 9 N and therefore almost entirely due to interface friction. The deepest test (D = 41 mm) by comparison achieved a tow force of 20 N and therefore the base friction accounts for less than half of the total in this case. The second component of tow force during ploughing in dry sand is a function of the cube of the plough depth and is influenced by soil properties such as unit weight and angle of friction. Figure 6 shows the steady state tow force increasing with depth for two series of tests, one with and one without a forecutter. The force prediction lines in Figure 6 were generated by Eq. 1, where Cw = 0.48 and was the measured interface coefficient of friction. Cs is an arbitrary multiple of unit weight which best fits the data for each series and is


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Figure 8. Comparison between ploughing tests with and without a forecutter in saturated fine sand. Figure 7. Variation of tow force during saturated ploughing tests without a forecutter.

taken as 10.5 for tests without the forecutter and 12 for tests with the forecutter. Note that the value of Cw used to best fit the data is similar to Cathie & Wintgens (2001) published value of 0.4. The values of Cs used here are greater than the Cathie & Wintgens (2001) published value for loose sand of 5 and fit between their values of medium sand (Cs = 10) and dense sand (Cs = 15). The reason for the difference in Cs values could be due to the inaccuracies in deriving empirical parameters from field tests where soil conditions may not be known over large distances between discrete investigation points.Additionally the scale model results may produce higher than expected tow forces due to the low effective stresses resulting in increased dilation and causing the sand to act as if it were in a denser state. To give an idea of the engagement of the forecutter at each plough depth, the forecutter in the model was approximately 12 mm shallower than the share during the tests shown in Figure 6, however this value was sensitive to pitch. The two tests show that the forecutter causes an increase in tow force in dry sand of around 7%, which appears constant over the range of depths tested compared to the tests without the forecutter. It has not been investigated why this is the case but it is thought that the forecutter may increase the tow force by causing some sand to be sheared twice, firstly by the forecutter and secondly by the share. As rate effects are not present during ploughing in dry sand, the dry sand test results allow the static component of tow force to be examined in isolation. Figure 7 shows the force and displacement data recorded during two saturated ploughing tests, each performed without a forecutter and at different velocities of 43 m/h and 159 m/h respectively. The 43 m/h test shows the increase of tow force as the plough translates through transition and into steady state which is very similar to that of the transition and steady state data shown in Figure 3. © 2011 by Taylor & Francis Group, LLC

3.4 Tow force-displacement profiles of saturated ploughing tests The data recorded during the 159 m/h test is more irregular and is typical of saturated ploughing test data at velocities of around 120 m/h and upwards. The reason for this may be due to the unsteady nature of the resistance of shear planes. Strain dependant resistance is magnified by the increase in normal effective stresses due to reductions in pore water pressure during dilation, which then dissipate post peak. Comparison of the irregular data was made easier as the post transition data was averaged from 550 mm horizontal displacement to the end of the test, for all tests.

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3.5 The effect of a forecutter on rate effects Saturated tests were carried out, both with and without a forecutter at depths of 37 mm and 38 mm respectively over a range of velocities from 28 m/h to 193 m/h in order to investigate how the tow force varied with plough velocity. Figure 8 shows how the steady state tow force varies with velocity for 19 saturated ploughing tests and allows comparison between tests where a forecutter was present and those where it was absent. On initial inspection, the results appear to contain less experimental error during low velocity tests than it does for tests where the velocity is greater than 120 m/h, as suggested by Figure 7. Figure 8 shows the tow force increases in an approximately linear manner with velocity for both test series and therefore linear trend lines were fitted to the data in using least squares regression. The rate effect is reduced for the test series where the forecutter was present and is highlighted by its shallower sloping trend line but increases the tow force at zero velocity. It has already been shown (see Figure 6) that the forecutter increases the tow force in dry sand and therefore it is reasonable to assume that when ploughing at speeds where the rate effect is relatively small the forecutter will increase the force required to tow the

plough. Consequently at velocities less than 65 m/hr the tow force is greatest for tests where a forecutter is present and for tests at velocities above 65m/h the tow force is greater when the forecutter is absent. The forecutter in effect creates a two stage ploughing process by cutting a shallow trench ahead of the share and in doing so reduces the distance between the share tip and the sands’ surface when the share makes its cut. This decreases the length of drainage path from the share and in turn reduces the rate effect by allowing a greater degree of dissipation during shearing. Although the rate effect is reduced in tests where the forecutter is present, the average plough depth was slightly less during tests where the forecutter was present and will have slightly skewed the results, making the forecutter appear additionally beneficial. It is felt however that this small difference in plough depth (3% less with a forecutter) will have little bearing on the results and the overall findings.

4

Figure 9. 50th scale plough data scaled to prototype scale.

IMPLICATIONS FOR PLOUGHING IN SANDS AND SILTS

test depth and Figure 9 therefore represents a plough trenching at 1.9 m in medium dense fine sand. Note that the actual tow forces may be increased slightly because of scale effects (higher φ and ψ because of low stress levels), but the type of effects observed are very likely to occur in full-scale ploughs. The equations of the best fit straight lines in Figure 9 show the tow force (F) as the sum of a constant (which is the result of the ‘static’ resistance to ploughing) and the rate effect, which is the product of velocity (v) and a multiplier. For the test series where the forecutter was used, the multiplier on the rate effect is only 60% of the tests without the forecutter and confirms that a forecutter is of use in reducing rate effects during ploughing. Consequently, it appears that its use is beneficial in soils where significant rate effects (and so high tow forces) are anticipated.

4.1 Limitations The results displayed in the previous section show clearly the effect of the forecutter on the performance of a 50th scale model pipeline plough. However for the results to be used directly by industry there are two main limitations that need to be addressed. Firstly all tests were performed in medium dense, fine Congleton sand, which was selected on the basis that fine sands and silty sands appear to cause the greatest problems to industry. This was noted by Reece and Grinsted, (1986) and is due to the fact that they are of relatively low permeability and readily dilate upon shearing. It would therefore be desirable to find out how sensitive the results are to permeability and also relative density. For example, ploughing in a slightly coarser sand may produce a force-velocity graph which shows ploughing with a forecutter will increase tow force at any velocity due to the rate dependant part of the overall tow force being small in comparison to the static component. The second limitation of this study is the reduced scale of testing and it is necessary to investigate how the rate effect might scale with plough size.

4.2 Use of a forecutter during ploughing Figure 9 shows the 50th scale plough data with the tow force scaled up to prototype scale. The mass of the model plough is 1/503 of prototype, the volume of sand affected by the model is assumed to be 1/503 of the prototype. Shear forces are assumed to be 1/503 the prototype assuming a Coulomb failure envelope, as depths and therefore effective stresses are 1/50 times those of the prototype and act on planes 1/502 the area of the prototype. This allows the assumption that tow force can simply be increased by a factor of 503 . The scaled depth of the tests is simply 50 times the model © 2011 by Taylor & Francis Group, LLC

5

CONCLUSIONS

A series of model scale pipeline plough tests have been conducted to investigate the effect of the forecutter on plough performance. The main findings were that: 1. The forecutter has a fairly minor impact on the pitch and overall stability of the plough. 2. Although the forecutter is relatively small in comparison to the share, its presence does cause a measureable (7%) increase in the ‘static’ component of tow force (see Fig. 5 and Fig. 7). 3. The forecutter is effective in reducing the rate effect during ploughing. 4. As the reduction in tow force achieved by its use is offset by an increase in the static component of tow force, it will only be beneficial for greater plough speeds. Further work is required to investigate how these speeds change with trench depth, plough scale and soil conditions.

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ACKNOWLEDGEMENTS The first author is supported by an EPSRC DTA award with additional funding and technical assistance for this project provided by CTC Marine Projects Ltd, Darlington, England. REFERENCES Bolton, M. D. (1986). The strength and dilatancy of sands. Geotechnique 36, No. 1, 65–78. Bransby, M.F., Brown, M.J., Hatherley, A. & Lauder, K. (2010) Pipeline plough performance in sand waves. Part 1: Model testing. Canadian Geotechnical Journal. Vol. 47, No. 1. pp. 49–64. DOI: 10.1139/T09-077. ISSN 0008-3674. Bransby, M.F., Brown, M.J., Lauder, K. & Hatherley, A. (2010) Pipeline plough performance in sand waves. Part 2: Kinematic calculation method. Canadian Geotechnical Journal. Vol. 47, No. 1. pp. 65–77. DOI: 10.1139/T09091. ISSN 0008-3674.


Cathie, D. N. & Wintgens, J. F. (2001) Pipeline Trenching Using Plows: Performance and Geotechnical Hazards. Proc. Thirty-third Annual Offshore Technology Conf., Houston (13145) April/May 2001, 1–14. Grinsted, T.W. (1985): Earthmoving in Submerged Sands, Unpublished Ph.D. thesis, University of Newcastle upon Tyne, UK. Hata, S. (1979). Submarine cable: multi-blade plough Geotechnique 29, No.1, 73–90. Lauder, K.D., Bransby, M.F., Brown, M.J., Pyrah, J., Steward, J. & Morgan, N. (2008) Experimental testing of the performance of pipeline ploughs, Proc. Eighteenth (2008) Int. Offshore and Polar Engineering Conf. Vancouver, Canada, July 6–11 2008. Palmer, A. C., Kenny, J. P., Perera, M. R. & Reece A. R. (1979). Design and operation of an underwater pipeline trenching ploughGeotechnique 29, No. 3, 305–322. Reece, A. R. and Grinsted, T. W. (1986). Soil Mechanics of Submarine Ploughs Proc. Eighteenth Annual Offshore Technology Conf., Houston (5341) May 1986, 453–461. Wood, D. M. (2004). Geotechnical Modeling. Oxfordshire: Spon Press. 246–258.

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State-of-the-art jet trenching analysis in stiff clays J.B. Machin & P.A. Allan Geomarine Limited, Newcastle, UK

ABSTRACT: Apparently simple upon first inspection, following careful analysis the water jet excavation of stiff clay soils is found to be a complex process. Due to the early lack of technical data, engineers working in the field of soil jetting first used empirical approaches and rules-of-thumb to quantify and model the process. Over the last ten years or so more data has become available, both from research work and actual field operations. As a result, a more scientific approach to the problem is now available. A design methodology is described in this paper. It has recently been applied by the authors to the design of a number of jet trenching and excavation machines and in the assessment of a number of jet trenching and excavation projects in stiff clay soils.

1

INTRODUCTION

Water jetting is widely used in applications where underwater cutting and removal of a material is required. Typical applications include cutting of steel and concrete, rock cutting and excavation, surface cleaning, and soils dredging. Underwater jetting is rather less energy efficient than jetting in air because of the large amount of mainly useless viscous shear currents that are created by a turbulent jet prior to impact. However, in many cases this draw-back is more than countered by two major advantages: The first being the perception of relative safety and reduced likelihood of accidental damage associated with use of water jets. The second is the assumption that jetting is a fairly simple operation requiring equipment that uses only a small number of moving parts. Underwater soils jetting is becoming an increasingly widespread construction technique. Engineers involved, among them the authors, have found that a careful jet nozzle design that optimizes jet stream pressures and velocities, while minimizing energy requirements, dramatically improves system efficiency. The result is improved productivity as well as permitting new soil types and strength classifications, for example stiff to hard clays, to be excavated that have not previously been possible.

2

CHARACTERISTICS OF SUBMERGED WATER JETS

Figure 1 shows the idealized penetration of a submerged water jet impinging on a clay stratum. Standard references on fluid mechanics e.g. Blevins (1984) provide solutions for the propagation of such continuous non-cavitating submerged water jets. In an initial region, close to the jet nozzle, the velocity and © 2011 by Taylor & Francis Group, LLC

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Figure 1. Idealized soil penetration of submerged water jet.

pressure in a “potential core” are approximately constant across the cross-section. This is followed by a transition region which is then followed by a region of fully developed flow. Here the jet entrains surrounding fluid via a viscous shear layer which leads to loss of jet energy. The jet diameter increases with distance from nozzle, and the jet velocity and pressure decay with distance from nozzle. It should be noted that these jets under analysis are generally high speed turbulent jets which possess a Reynolds Number greater than about 3,000. Figure 2 taken from Blevins (1984) shows experimental data of the decay in centre-line axial velocity along a submerged water jet versus is axial distance. Based on this and similar data which confirms the relationship, the authors have adopted a model of velocity ratio V/V◦ versus normalized distance d/D◦ from nozzle, where V is the jet velocity at distance d

where P is the pressure on the area of the small element, ρw is the mass density of the jet fluid, V is the axial velocity of the water jet, and f() is a function that expresses the velocity distribution of a submerged water jet. On the question of the shape of velocity distribution, as part of their research into the efficiency of jet mining of China Clay deposits in England, Davies and Jackson (1981) measured the impact pressure distribution underneath the jet impact zone and found it to resemble a “bell” shape. If the impact stagnation pressure is equated to the bearing capacity of the soil excavation face by q = Nc Su for a cohesive soil, then the threshold velocity required to cause bearing capacity failure follows: Figure 2. Decay in centre-line velocity, Blevins (1984).

where V is in m/s, Su is in kPa, and the approximation has been obtained by putting Nc = 6 and ρw = 1Te/m3 . An effect of a bearing capacity failure will be the development of some form of depression, hole or cavity in the soil. This will alter the flow characteristics and pressures. As the water turns and flows laterally over the soil, erosion due to viscous shear will occur. Equation (3) is of significance in that it determines the minimum required operating point or threshold for an underwater jetting system to begin to cut into a cohesive soil of known undrained shear strength. Equation (3) can be combined with Equations (1) and (2) to give us a value for the total depth of cut as a function of pressure:

Figure 3. Initial jet impact behaviour.

from a nozzle of diameter D◦ , and V◦ is the jet exit velocity at the nozzle. V is assumed constant up to six diameters from the nozzle, and then decreases with distance according to:

where d > 6 D◦ . Figure 1 shows a divergent fully developed jet angle forming a conical shaped cavity. The authors have found from experience that this angle is typically about 14 degrees. However the angle does seem to vary somewhat according to different authors although experience suggests that this typical value works well with the range of nozzle types normally used. 2.1

Where d is measured in metres and nozzle pressure P is measured in bar and a nozzle coefficient of discharge of 1.0 is assumed. Equation (4) is used to check that jetting systems have at least a minimum necessary ability to excavate clay of a given shear strength Su. For example, the equation suggests that a well designed nozzle of pressure 10 bar and diameter 20 mm can cut up to 15 cm into clay soil of shear strength 100 kPa. The equation can also be correlated with in-situ Cone Penetration Test (CPT) measurements of shear strength.

Initial jet impact behavior

When the high speed jet impacts on a perpendicular soil face the water velocity in the direction of travel is brought to an abrupt halt and even partially reflected backwards. A stagnation pressure must therefore arise at the point of impact. To again escape from the excavation face water will then also try to deflect sideways. Figure 3 illustrates this process, again after Blevins (1984). From classical fluid mechanics theory, the equation for the stagnation pressure function exerted on a small element of a plane perpendicular to the jet axis is:


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2.2

Correlation with test observations

Data was gathered and back-analyzed from four major jet impact testing research programs by Machin et al. (2001). These programs were carried out under controlled conditions in clay soils with strengths in the range 2.5 kPa to 120 kPa. Several hundred individual tests have been performed and back-analyzed. All of these tests to date have confirmed that upon impact the resulting soil cavity depth develops approximately as represented on Figures 4 and 5 which respectively illustrate the behavior of cavity formation both under a static nozzle and under a traversing nozzle of constant translational velocity.

Figure 6. Jetting arm tool nozzle array for V shaped trench.

to “hydro-fracturing”. It results in a rough-looking and uneven cylindrical cavity surrounded by a wide zone of disturbance. It is often not fully cleared of soil lumps.

Figure 4. Observed cavity formation with static nozzle.

Figure 5. Observed cavity formation with translating nozzle.

Some notable observations that were reported from these test results can be summarized as follows: 1. During the static nozzle tests the “Quasiinstantaneous” cavity depth was found to correlate very closely with the threshold bearing pressure theory of Equation (4). The cavity appeared to form in a very short time interval indeed (estimated as a fraction of a second). 2. During static tests a full cavity depth generally develops over several seconds. It is greater than the depth of the initial “Quasi-instantaneous” cavity and its depth varies with clay properties. 3. During the traversing nozzle tests it is observed that the “Quasi-instantaneous” cavity depth consistently occurs once translational speeds exceed about 0.1 to 0.5 m/s, depending on clay properties. For lower translation speeds deeper cavities are formed. 4. A clear cylindrical cavity is usually formed by the water jet. It is either vertical or bell-shaped and has fairly smooth sides. Minimum diameter or width is always greater than the nozzle diameter and is often recorded at up to about 3 times the nozzle diameter. 5. Occasionally during testing a different type of failure is observed. It resembles a soil upheaval due © 2011 by Taylor & Francis Group, LLC

Based on the above observations, the authors have concluded that for conservative jetting assessment purposes a key design approach should be as follows: In cases where cutting of the soil is required the jet must be demonstrated as able to form a sufficiently deep cavity according to Equation (4) which defines a “Quasi-instantaneous” depth of cut. It could be argued that this rule is over-conservative in that it does not account for the additional timedependant cavity propagation effect that is always observed. However, it is tentatively suggested that this is a soil softening and particle erosion effect with a magnitude that depends on the property of permeability – a property which is often difficult to predict with accuracy. Another issue which needs to be addressed is the use of the constant bearing capacity factor (Nc = 6) irrespective of depth. This contrasts with established factors for deep penetration bearing (e.g. Nc = 9) and for CPT’s (e.g. Nc = 12). Here the authors’ supposition is that the local erosive nature of the jet failure mechanism results in a shear failure process that is more representative of a shallow penetration than a deep one even at the base of the cavity. However this is clearly an area not yet fully understood. 2.3

Dis-aggregation in clay soils

The process described above illustrates how the excavation face of a clay soil is penetrated and cut up with water jets to form cavities. The linking up of these cavities can then act to form a “cookie-cutter” grid that results in the cohesive soil being continuously extruded into blocks. In order for mass excavation to occur it is vital that these blocks must then disaggregate (separate) into lumps that can be efficiently transported. Figure 6 illustrates a jetting arm tool with a nozzle array designed to cut the soil into blocks and form a “V” shaped pipeline trench. Therefore, the second part the jetting model is to establish that the soil will dis-aggregate into lumps that are suitable for removal. The authors propose that

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there are in fact only two driving forces available for dis-aggregating the lumps of cohesive material and they are: 1. Self-weight/gravity. 2. Water flow that entrains the lumps with a viscous shear. On a vertical excavation face the clay extrusions will always tend to form blocks that cantilever outwards from the face. The depth and breadth section dimensions of the cantilever blocks will depend on the spacing between nozzles on the high pressure jet array. The effect of water flow around the clay bricks serves to provide an additional load that further contributes to their instability.

Table 1.

Clay Plasticity Classification

Plasticity Index (%)

Very Low Low Moderate High Very High Extremely High

100

2.5

Hydro-fracture and fluidization behavior

When pressurized water finds its way into voids within the excavation face a different type of soil failure © 2011 by Taylor & Francis Group, LLC

Water Content Range for Good Workability (%) 30

known as “hydro-fracture” can occur. Given the right conditions this mechanism will open up incipient fractures within the soil, thus forming voids and cavities. Hydro-fracture is a different mechanism to the bearing failure approach. Its occurrence during testing is described earlier in this report. In certain rare cases (e.g., the over-consolidated clays around the Gullfaks site in the North Sea), the authors have anecdotally observed that natural preexisting fractures seem to be present within the soil to such an extent that hydro-fracture mechanisms apparently occurs with hardly any need for bearing failure type jetting. Unexpectedly high overall excavation productivity appears to occur. However, more typically the authors have found that hydro-fracturing effects provide a welcome, if often unexpected, supplement to conventional mechanisms. Hydraulic fracture in cohesive soil is theoretically not time-dependent. In practice, a hydraulic fracture needs to be initiated by water flow into a cavity or highly permeable layer and this obviously requires a finite duration in which to occur. The fracture may then develop provided that the water pressure is maintained at a value sufficient to continue to expand the flow of water into the fracture. Simple mathematical relationships have been developed to assess the initiation of hydro-fracturing such as Farmer (1983). The applicable equation for the pressure required is:

2.4 Tensile strength and plasticity effects From a structural analysis view-point, a rigorous solution to the dis-aggregation problem therefore requires knowledge of the propensity of these cantilevering blocks of soil to collapse. It therefore requires knowledge of the tensile strength property of the clay. Unfortunately information about clay tensile strength is rarely provided in traditional geotechnical data surveys. Furthermore, the parameter seems to be rather difficult to predict. Literature cites field values ranging from a theoretical maximum of twice undrained shear strength for a highly plastic clay, to one tenth of undrained shear strength (or even less) for clay close to or below its plastic limit. Tensile strength relies on the negative inter-particle pore pressures that remain in the clay blocks after the excavation face has been cut up by the high pressure jets. These negative pore pressures will be released as pore water migrates, resulting in softening at a rate dependent on permeability and lump dimensions. As a consequence tensile strength is also influenced by the presence of natural fracturing, “toughness” and “brittleness” of the soil, factors which are difficult to quantify but are reflected to a degree in its plasticity description. Heavily over-consolidated stiff to hard clays with undrained shear strengths of about 100 to 170 kPa or greater can really only exist at or below their plastic limit (Skempton & Northey1952). In many ways they behave like weak rock and tensile strengths in the range one half to one tenth of undrained shear strength could be tentatively predicted. Such readiness of the soil to break up into small lumps upon loading is also known by agricultural engineers as its “workability index”. The authors have found that referencing studies of this property have also proved very useful in assessments of disaggregation and predictions of dis-aggregated soil lump sizes formed. Rosa et al. (2003) have tabled the following model:

Soil Workability Index, Rosa et al. (2003).

Where P = applied pressure, σ = minor principal stress, T = tensile strength of material. In the case of jetting, the orientation of the minor principle stress will affect the direction along which the fissure opens. Depending on the in situ stresses in the soil, the orientation of the minor principle stress would normally be vertical in a stiff over-consolidated clay. A higher stress will be present in the horizontal directions. However, as the excavation face advances, the horizontal stress along its axis will be relieved and this will be become the minor principal stress. Hence jetting into this face will open a fissure in a plane perpendicular and the face of the excavation will fracture as clay blocks are broken. It is recommended that for each jetting project a calculation using Equation (5) should be made to check the limiting value of soil tensile strength below which a

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to Hjulstrom (1935, 1939) and Shields (1936) previously cited, the authors use data provided by Wakefield (1997) and Wilson (1979). These papers include charts that give guidance on the effect of varying drag coefficient and shape factor for different soil particles in varying flow regimes.

4

HYDRAULIC FLOW EFFICIENCY

significant hydro-fracture event can be expected to initiate. This can be checked against the predicted value of Tensile Stress as discussed earlier. The pressure of the jetted water can also cause the soil to “fluidize”. Fluidization occurs when the soil pore pressures increase to equal the total external pressure, resulting in zero inter-locking shear forces between soil particles and complete dis-aggregation of the soil matrix. At this point the soil shear strength and tensile strength also reduce to zero. Fluidization may well occur during jetting in loose silts and silty sands, and soft clays near their liquid limit. The authors have observed fluidization during the jetting process at many different sites. A good example is the extremely “fat” seabed clay of the Green Canyon zone of the Mississippi Delta, in the deep water Gulf of Mexico.

The effect of nozzle geometry definitely makes a difference. Woodward (1985) has made an illuminating comparison of the pressure profiles of a number of commercially available water jetting nozzles. An efficient nozzle geometry is designed to minimize energy losses and give a discharge coefficient (Cd ) of typically 0.9 or better. The effect of straight section of the nozzle is also known to be important and both very short, or very long sections are likely to significantly increase energy loss. It is also of high importance that the hydraulic flow within the jetting arms and nozzles is well conditioned and streamlines are smooth and direct with no rapid changes of direction, eddy’s or other energy wasting diversions. The authors have found that the use of correctly designed flow straighteners is a highly effective way of improving the hydraulic flow efficiency of jetting arm designs. Flow straighteners are thin blades that are used to straighten hydraulic streamlines at the nozzle inlet.

3 TRANSPORTATION OF SOIL

5

After the phases of initial jet impact and soil disaggregation, the final stage that completes the analysis of the jetting process is an assessment of the transportation of the soil lumps and particles. A sketch of the overall jetting process is shown on Figure 7. Transportation of jetted soil particles and lumps is carried out by a turbid hydraulic flow. One source of this flow is from the water jets used to cut the cavities in the excavation face. Residual flow from these jets is deflected back from the excavation face and can then be harnessed for soil particle fluidization and lump erosion. At some distance away from the excavation face the turbid hydraulic flow will slow down to a quasiquiescent state. In this region soil particles and lumps are steadily re-deposited onto the seabed, at a rate controlled by the fall speed of the particles and lumps within the mixture. In between these two regions the fluidized layer always undergoes a steep and sudden shock-like transition as it changes from supercritical (high Froude number) to sub critical (low Froude number) flow. The transportation mechanism is well understood although, unfortunately, is rather complex to solve in closed form. However, there is a large amount of data and helpful charts in the literature. In addition

The authors have found that a soils jetting system can be systematically designed to provide a predictable and repeatable jet excavation performance in soil. A design procedure has been described. However, the authors have found that a great deal of uncertainty still remains in the detailed analysis of the soil jetting mechanism. Further work is definitely necessary if the industry wishes to further enhance the capability of soil jetting systems. In particular it is highlighted that soil softening and particle erosion effects, and their inter-relationship with permeability and plasticity, are currently not well understood. Also the role of the tensile strength parameter and its accurate determination is a matter of uncertainty. Additional testing programs and new types of in situ and laboratory testing are likely to be required.

Figure 7. Illustration of overall jet excavation process.


CONCLUSIONS

REFERENCES Blevins, R.D., 1984, Applied Fluid Dynamics Handbook, Van Nostrand Reinhold Company. Davies, T.W., and Jackson, M.K., 1981, Optimum Conditions for the Hydraulic Mining of China Clay, Proceedings of the First USWaterjet Conference, Golden Colorado. Water Jet Technology Association.

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Farmer, I., 1983, Engineering Behavior of Rocks. Chapman and Hall. Hjulstrom, F., 1935, Studies in the morphological activity of rivers illustrated by the River Fyris, Bulletin 25, Geology Institute of the University of Upsala, 221–258. Hjulstrom, F., 1939, Transportation of detritus by moving water”, Symposium on Recent Marine Sediments, ed.P.D.Trask, Specialist Publication of the Society of Economic Paleontology and Mineralogy, Tulsa USA, Vol. 4, 5–31. Machin, J.B., Messina F.D., Mangal J.K., Girard, J., Finch, M., 2001, Recent Research on Stiff Clay Jetting, Offshore Technology Conference, OTC13139, Houston, Texas. Rosa 2003, D. de la, Mayol, F., Diaz-Pereira, E., 2003, Aljarafe Model Soil Plasticity and Workability, Institute for Natural Resources and Agrobiology, Seville, Spain. Shields, A., 1936, Application of similarity principles and turbulence research to bed-Load movement, California


Institute of Technology, W.M. Kekck lab of Hydraulics and Water Resources, Report No 167. Skempton A.W. and Northey R.D. (1952) The sensitivity of clays. Geotechnique, Vol 3 30–53. Wakefield, A.W., 1997, The Complete Model of the Solidshandling jet Pump, 9th International Conference on Transport and Sedimentation of Solid Particles, Cracow, Proceedings Volume 2. Wilson, K.C., 1979, Deposition-Limit Nomograms for Particles of Various Densities in Pipeline Flow, Proc. Hydrotransport 6, BHRA Fluid Engineering, Cranfield, UK. Woodward, M.J., 1985, An Experimental Comparison of Commercially Available Steady Straight-Pattern Water Jetting Nozzles, Proceedings of the Third US Waterjet Conference, Pittsburgh, Pennsylvania. Water Jet Technology Association.

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Numerical modelling of soil around offshore pipeline plough shares W. Peng & M.F. Bransby University of Dundee, Dundee, UK

ABSTRACT: Plough share performance has been investigated by conducting a series of finite element analyses in which the process was simplified to that of a plane strain problem. Fully drained analyses have confirmed the relationship between tow force and trench depth and weight as observed previously during experiments. Analyses conducted to investigate rate effects confirmed that partially drained conditions were achieved and that the force-velocity relationships are unlikely to be linear over a wide range of velocities. 1

INTRODUCTION

Pipeline protection is often achieved through burial in the seabed sediment. One common way that this is achieved is by ploughing (e.g. Finch et al., 2000). A large (e.g. approximately 20 m long) plough (Fig. 1a) is dragged through the seabed, and the plough forms a ‘V’-shaped trench into which the pipeline is placed before usually being backfilled. A large support vessel is needed to conduct pipeline ploughing because as well as significant support infrastructure a large bollard pull is required to move the plough through the seabed. Significant tow force is required firstly because of the size and consequent weight of the plough needed to find a suitably deep trench (typically up to 1.8 m) and secondly because of rate effects. The latter mean that tow forces increase with plough velocity and will limit the achievable plough velocity for a given bollard pull. Accurate prediction of pipeline or cable plough performance, particularly velocity, is critical for offshore contractors as it affects the tender price and the choice of installation method and trenching tool. The variation of tow force with plough weight and trench depth is reasonably well known (e.g. Reece & Grinsted, 1986) and is partly analogous to the lateral capacity of a gravity retaining wall with a base friction and passive lateral pressure term. However, the variation of tow force with plough velocity (the plough rate effect) is less well understood, being believed to depend on the dilation potential of the soil, its permeability as well as the trench depth (Palmer et al. 1999; Cathie & Wintgens, 2001). Currently, methodologies exist which are based on, or calibrated from, backcalculation of previous trenching (e.g. Cathie & Wintgens, 2001). These have limited accuracy because of the uncertainty about and variability of soil conditions in the field. To examine the above issues, a series of finite element (FE) analyses have been carried out. These investigated directly how tow force varies with plough weight, trench depth and velocity. However, because of © 2011 by Taylor & Francis Group, LLC

Figure 1. Plough shape and 2D simplification.

the complexity of three dimensional shape of a pipeline plough, the problem was simplified here for the case of a plane strain (i.e. infinitely wide) share (Fig. 1b). Despite the simplification, the analyses were expected to capture the key aspects of soil behaviour around plough shares during ploughing which provides most of the tow resistance. The paper will explain the methodology used before presenting results for both ‘static’ (drained) and ‘dynamic’ (partially drained) plough speeds. 2 2.1

NUMERICAL METHODOLOGY Introduction

The plough share geometry was idealized to an infinitely wide wedge as shown in Figure 1b. The share

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is assumed to be cutting a trench of depth D and the share blade angle was denoted as β. The external forces W and F act on the blade where W is the vertical component of the plough weight (minus the vertical component of the tow force and load taken by the skids for a full plough) and F is the horizontal component of tow force taken by the share. FE analyses were carried out using the Plaxis program. The share blade was modeled as a stiff, linear elastic material with a rough interface (δ = φ) to model a rigid object, whereas the soil used effective stress constitutive models to describe its behaviour. A range of different blade angles were investigated starting with β = 90◦ to allow benchmarking. 2.2 Soil characterisation Soil was described using a linear elastic, perfectly plastic constitutive law using the Mohr-Coulomb failure criterion. In general, for fully drained analyses strength was modeled using a friction angle, φ and a separate specified dilation angle, ψ, with a small apparent cohesion, c given to prevent numerical errors at small effective stress. After using this constitutive model for both fully drained and undrained analyses, it was found to be less suitable for partially drained analyses. This was because there was no dilation cut-off so that soil continued to dilate after peak shear stress conditions were achieved. Consequently, later analyses used the ‘Hardening Soil’ model in Plaxis which has the ability to stop dilating after a critical state voids ratio has been reached. The type of calculation method used in Plaxis was ‘plastic’ for analyses of static resistance force both in drained and undrained condition, but ‘consolidation’ for the dynamic force analysis where the speed of model movement was controlled and partial drainage of pore water was modelled. 2.3

Consolidation analyses were required for the partially drained ‘dynamic’ loading conditions. Hence for these conditions sufficient time (one-year) was required after application of W to ensure full soil consolidation under the plough weight. The final calculation step involved lateral movement of the plough share through the soil. This was done using displacement control whereby only the horizontal displacement component of the vertical, left edge of the share was displaced. The method allowed possible vertical displacements of the share under the fixed vertical load but prevented rotation (which is unlikely to occur in steady-state ploughing conditions). The reaction forces were calculated for the incrementally applied displacement in order to obtain the soil resistance (F) – displacement (u) curve for the share. These confirmed that a steady state condition was achieved for all analyses. For the partially drained analyses, combinations of displacement magnitude and analysis time were selected to specify displacement velocities ranging from 0.0008 m/hr to 600 m/hr. 3 3.1

RESULTS: ‘STATIC’ ANALYSES Introduction

A series of analyses were first conducted for drained conditions. These were design to investigate the reaction forces on the simplified plough share under ‘static’ conditions. Following validation (with β = 90◦ ), the effects of varying the vertical reaction force (or plough weight), W and the embedment depth were investigated for one share geometry and soil condition before examining the effects of changing share blade angle and soil properties. For static analyses, F is expected to depend on a modified version Cathie & Wingens (2001) equation:

Geometry selection, mesh discretisation and loading

The geometry modeled is shown approximately in Figure 1b. Different analyses investigated the effect of varying both D and β. Both vertical and horizontal soil displacements were prevented at the base and lateral boundaries which were placed suitably remotely as not to affect the plough capacity. The mesh density was chosen following a validation exercise where results were benchmarked against retaining wall solutions. The mesh chosen used 15node elements which were concentrated in the area near the share tip and on the base. Share loading was achieved in three calculation stages. In the first stage, in situ soil stresses were fixed with no displacements occurring. In the second stage, the weight of the plough, W was applied. As this occurred the plough settled fractionally to reach an equilibrium position with realistic contact stresses between the plough and the share. For the case of fully drained analyses, this occurred instantaneously. © 2011 by Taylor & Francis Group, LLC

where Cw = base friction coefficient (Cw = tan δ, where δ is the interface friction angle); Cs = passive soil resistance (similar to Kp /2); and γ is the effective unit weight of soil. Note that the second term contains D2 rather than the D3 suggested for real ploughs because of the plane strain analysis conducted here. 3.2

Example problem

Results are shown first for a typical drained analysis where the soil displacements and reaction forces acting on the share are investigated in detail. In the analyses a blade of weight, W = 268 kN/m (27.3 tonnes per m width) with angle, β = 35◦ cuts a trench of depth, D = 1.5 m though soil with unit weight, γ = 17 kN/m3 , angle of friction, φ = 35◦ and dilation angle ψ = 10◦ . Note that this corresponds approximately to realistic conditions except that the buoyant unit weight would

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Figure 2. Displacement magnitudes at failure (γ = 17 kN/m3 ; φ = 35◦ ; β = 35◦ ; ψ = 10◦ ; D = 1.5 m).

be closer to 9–10 kN/m3 and the share would not be fully rough (i.e. δ = φ here). Contours of calculated incremental soil displacements are shown in Figure 2 at the point of soil failure (when peak soil resistance is achieved). As the share translates to the right (with no rotation because of the displacement fixity), a wedge of soil above and to the right of the blade is pushed upwards. A shear plane inclined at approximately 40◦ to the horizontal forms from the tip of the share blade to reach the soil surface. This agrees reasonably well with the angle = 90–φ/2– β/2 (=90–35/2–35/2 = 45◦ shown dotted in Fig. 2) predicted by Lewis & Coyne (1999), for an inclined blade. Both the normal and tangential reaction stresses acting on the inclined face of the share (Fig. 3a) and the share base (Fig. 3b) at failure were extracted from the output data. For the inclined face the two stress components were obtained from the horizontal, vertical and shear stresses in the plane using Mohr’s circle. These stresses were also integrated numerically to obtain the normal and tangential forces acting on the share base (Nb and Tb ) and share face (Ns , Ts ) as defined in Figure 1b. The vertical and horizontal components of these forces were added and compared to the applied W and measured total force F respectively with good agreement confirming that the method was accurate. Figure 3a reveals a triangular distribution of both normal and shear stresses acted on the inclined share face at failure. This is as expected as there is no normal stress at the soil surface because there in no vertical effective stress, but the vertical effective stress increases with length along the face as the depth in creases. The dotted line on Figure 3a gives an indication of the limiting shear stress acting on that plane calculated from the normal stress and the angle of friction (τn = σn tan φ). Soil is clearly shearing on the rough face with limiting shear stress conditions as the soil ‘slides’ past the blade. Figure 3b shows that the distribution of stresses on the base of the share is complex. There is a nonuniform distribution of normal stress and, surprisingly, the soil is not at limiting shear stress conditions on the share base. Note that the weight, W of the plough alone generates an average σn = 44.8 kPa. The additional normal stress is because the soil reaction forces on the inclined face (Ns and Ts in Fig. ab) give a net downwards reaction thereby increasing the vertical © 2011 by Taylor & Francis Group, LLC

Figure 3. Distribution of calculated normal and shear forces.

Figure 4. Contours of horizontal stresses σyy (γ = 17 kN/m3 ; φ = 35◦ ; β = 35◦ ; ψ = 10◦ ; D = 1.5 m).

force (W + Ns cos β – Ts sin β) applied to the share base. Finally, Figure 4 shows the contours of horizontal stresses in the soil when peak soil resistance is mobilized and confirms that the largest soil stresses occur near the tip of the share.

3.3

Effect of plough weight

A series of analyses were carried out in which the deadweight of the plough was varied for a fixed trench depth, share blade shape and soil parameters. A typical result of resistance force against plough weight is shown in Figure 5 for a plough with blade angle, β = 56◦ in soil with φ = δ = 35◦ and ψ = 10◦ . For an analysis with no plough weight, the reaction forces on the share face ensure that there is a tow force required (like a passive pressure on a retaining wall) and the

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Figure 7. Tow force against trench depth squared (γ = 17 kN/m3 ; φ = 35◦ ; ψ = 10◦ ; β = 56◦ ).

Figure 5. Tow force against plough weight (γ = 17 kN/m3 ; φ = 35◦ ; ψ = 10◦ ; β = 6◦ ).

Figure 6. Force-displacement curves for different trench depths (γ = 17 kN/m3 ; φ = 35◦ ; ψ = 10◦ ; β = 56◦ ; W = 600 kN/m).

total reaction force increases linearly as the plough weight increases, i.e. dF/dW = Cw (from Eq. 1). The analyses confirmed that Cw ≈ tan φ = tan δ for the rough plough shares considered for a range of drained conditions. Clearly real ploughs have smoother shares as Cathie & Wintgens (2001) suggested that Cw = 0.4 (or δ = 21.8◦ ). 3.4 Plough share depth Trench depth (or share depth) were investigated directly in a series of analyses where D was varied from 0 to 3 m. The tow force – share displacement curves (Fig. 6) show: (i) that tow force increases with trench depth, and; (ii) the distance required to mobilize peak load increases with D. Note however, that the true displacement to mobilize peak resistance for a plough will depend on the plough kinematics and the generation of steady-state spoil heaps around the share, neither of which is modeled here. Tow force appears to increase with trench depth squared (Fig. 7) as expected for plane strain conditions. The intercept of the trend line at D = 0 corresponds to the base shear condition (F = Cw W ) and the gradient of the line dF/dD2 = Cs γ (where Cs = 41.758/17 = 2.456 in Fig. 7). Note that the tow force result for D = 0 m (Fig. 6) is lower than expected © 2011 by Taylor & Francis Group, LLC

Figure 8. K2p and Kp tan φ plotted against deduced Cs for different angles of friction (β = 56◦ ).

and does not follow the trend shown in Fig. 7. This requires further investigation, but is not important for realistic plough geometries where D > 0. Additional analyses in which the unit weight of the soil was varied for a given D confirmed equation 1 and allowed separate calculation of Cs for given soil conditions with excellent agreement. 3.5

Change in angle of friction

A suite of finite element analyses were conducted in which Cw and Cs were deduced for set of FE analyses each with different angles of friction. Cw was confirmed to vary with the interface friction angle (Cw = tan δ). The deduced values of Cs are plotted against K2p and Kp tanφ in Figure 8, where the passive lateral earth pressure coefficient, Kp = (1 + sinφ)/(1 − sinφ). The linearity of both relationships suggests that these terms could be incorporated in expressions for Cs . 3.6

Change in share blade angle

Analyses were repeated for one angle of friction (φ = 35◦ ; ‘medium dense sand’) with different blade angles to determine how Cs varied with blade angle (Fig. 9). As expected, the tow force (and so Cs ) increases as the share blade become less sharp (i.e. β increases). The exact form of the relationship for

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Figure 9. Variation of Cs with blade angle.

different soil conditions requires further study (for example the large increase in Cs from β = 56.5 to 61.5◦ ), but Cs appears relatively insensitive to blade angle at the low blade angles used by ploughs. 3.7

Summary

In summary, drained analyses confirmed Grinsted’s (1985) and Cathie & Wintgens’ (2001) equation for static ploughing (Eq. 1). The base shear resistance term appeared to give dF/dW = tan δ as expected, but the Cs term was a more complex function of φ. From the study of one blade angle it appeared that Cs linearly depended on Kp tan φ or K2p with further investigation required to ascertain the generality of this finding. Finally, the effect of the blade angle, β is unclear and it not accounted for in Cathie & Wintgens’ (2001) model where it is implicitly assumed that this angle is constant for all ploughs. 4

RESULTS: RATE EFFECTS

Following the drained calculations, additional analyses were conducted to investigate how tow force changes with velocity and soil permeability. For these analyses the trench depth, D, plough weight, W and blade angle, β were kept constant together with the soil properties. Soil was modeled using the Hardening soil model with φ = 35◦ and a dilation cut-off was used. Consolidation analyses were conducted to allow calculation of excess pore pressure migration. A wide range of velocities from 0.0008 m/hr to 600 m/hr were investigated for soil with three different permeabilities, k = 0.42 m/hr, 0.042 m/hr and 0.0042 m/hr. These permeabilities represent soils ranging from fine sand to silt. Cathie & Wingens (2001) extended Eq. 1 to include rate effects:

where Cd = dynamic rate term; and v is the plough rate. For plane strain conditions, it is not clear whether the second term should contain D or D2 (the D2 term suggested is an empirical fit to data.) © 2011 by Taylor & Francis Group, LLC

Figure 10. Tow force against velocity for different soil permeabilities.

Figure 10 plots steady-state tow force against velocity as calculated using FEA. When velocity is plotted linearly (Fig. 10a) there is no clear similarity between the curves. In contrast to equation 2, all three are nonlinear, although each may be considered to contain an almost linear section at relatively low velocities. However, when plotted with a logarithmic velocity scale (Fig. 10b), the similarity of each curve becomes clearer. At low velocity F hardly varies with increased velocity (dF/dv = 0) and has the same value for each permeability: this is fully drained behaviour and the drag forces will be governed by the ‘static’ part of the tow force equation (Eq. 1). At very high velocity, the tow force approaches another, much higher plateau force where dF/dv = 0: this is fully undrained behaviour. In the section between the two extremes, partial drainage occurs. The form of the curve is similar to that observed for shallow foundations (Finnie, 1993), penetrometers (Silva et al., 2006) or pipelines (Bransby & Ireland, 2009) displaced at different, constant velocities through soil, where velocity was normalized by the coefficient of consolidation, cv , divided by the drainage path length, D viz. vD/cv . For the analysis with identical geometries (so D is constant) and soil stiffness (so Eo is constant), the dimensionless group vD/cv will vary directly with v/k. Alternatively, Palmer (1999) suggested that rate effects in ploughs depend on vS/cv , where S is the dilation potential. Again, this suggests that for otherwise identical soils apart from their permeability, the rate effect will depend on v/k. Consequently, Figure 11 re-plots the data with the velocity normalized by the permeability. The normalization is successful as the data reduce to one line.

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effects predicted (far larger than that observed experimentally or in the field) which may be due either to the soil constitutive model used or the two-dimensional nature of the analysis. Furthermore, additional analyses are required to investigate the effect of varying the trench depth, D. 5

CONCLUSIONS

A series of finite element analyses have been conducted to investigate the behaviour of plough shares ‘cutting’ soil during ploughing. Although the analyses have considered a simplified wide (plane strain) plough share only, the results have allowed investigation of soil behaviour during both ‘static’(drained) and ‘dynamic’ (partially drained) ploughing. The results show that the base friction component of loading is approximated well in existing solutions (e.g. dF/dW = tan δ is appropriate), but that the depth dependent term (Cs γD2 ), and particularly the prediction of the rate component of drag resistance may need improvement. REFERENCES Figure 11. Tow force against normalised velocity for different soil permeabilities.

The force-normalised velocity relationship shown in Figure 11a could be approximated by a linear relationship for v/k < 25 (see dashed line on Fig. 11a). However, this is a very low velocity range. For example for a soil with d10 = 0.1 mm (a value close to the mean of the field data shown by Cathie & Wintgens, 2001), v/k = 25 corresponds to a velocity, v ≈ 9 m/hr compared to likely real ploughing rates of several hundreds of metres per hour. Likely plough progress 100 m/hr < v < 1000 m/hr corresponds to 140 < v/k < 2800 for the example soil given above. This might suggest that the linear forcevelocity section fitted by field data may trend towards the almost undrained section of the line (dotted in Fig. 10a) and produce an over-estimate of the static component as indicated. The above findings should be treated with some caution. There appears to be very large dynamic rate


Bransby, M.F. & Ireland, J. 2009. Rate effects during pipeline upheaval buckling in sand. ICE Geotechnical Engineering 162(5): 247–256. Cathie, D.N. & Wintgens, J-F. 2001. Pipeline trenching using plows: performance and geotechnical hazards. Proc. Offshore Technology Conference OTC 13145, Houston, May 2001. Finch, M., Fisher, R., Palmer, A.C. & Baumgard, A. 2000. An integrated approach to pipeline burial in the 21st century. Deep Offshore Technology 2000. Finnie, I.M.S. 1993. The performance of shallow foundations in calcareous soil, PhD thesis, University of Western Australia. Grinstead, T.W. 1985. Earthmoving in submerged sands. PhD Dissertation, University of Newcastle-upon-Tyne, UK. Coyne, J.C. & Lewis, G.W. 1999. Analysis of plowing forces for a finite-width blade in dense, ocean bottom sand. Palmer, A.C. 1999. Speed effects in cutting and ploughing. Géotechnique 49(3): 285–294. Reese, A.R. & Grinstead, T.W. 1986. Soil mechanics of submarine ploughs, OTC 5341 (May 1986) Silva M.F., White D.J. & Bolton M.D. 2006 An analytical study of the effect of penetration rate on PCPTs in clay. Int. J. Num. Anal. Meth. Geomechanics 30: 501–527.

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Anchor–chain–rockfill–soil interaction: Evolution of design methods H. van Lottum & H.J. Luger Deltares, Delft, The Netherlands

ABSTRACT: Deltares (former GeoDelft and Delft Hydraulics) has carried out different kinds of research in the area of anchor – rockfill – soil interaction. The paper gives an overview of the research that was carried out during the last decade by Deltares on the encounters of anchors and rock fill berms and the employment and embedment of standard shipping anchors as well as high holding power offshore anchors in the soil. The paper presents a view on future research that is needed to improve the understanding of dragging anchors and pipeline protection systems 1

INTRODUCTION

Many offshore pipelines and cables are located in areas where they are exposed to damage by dragging or dropping anchors. To prevent pipelines and cables to fail in such events different kinds of protection measures are designed. For better understanding the mechanisms that occur in these protection measures and how anchors and chains approach the pipeline or cable, both physical modelling and analytical research is required. Such research leads to improvement of present protection measures and new protection techniques may be developed. In the last decade, Deltares carried out several research projects on anchor–chain–rockfil–soil interaction. Within this research, Deltares developed custom made design tools and gained new insights on mechanisms involved in this interaction. These new insights gave us our view on developments needed to improve description on behaviour of offshore anchors and design of pipeline protection covers. At first recent (model) research will be described, resulting in the authors perspective on the development of future design methods. 2

RECENT RESEARCH

2.1 Lamma pipeline cover verification For the Lamma gas-pipeline near Hong Kong, a verification of the designed pipeline protection was requested. To this end, Deltares developed the calculation model CoverCalc that aims to predict the protection level provided by an arbitrary rockfill cover over a pipeline. Calculations were accompanied by model tests carried out in a large test flume. 2.1.1 CoverCalc CoverCalc was developed in 2004 to enable the comparison of rockfill cover shapes and volumes. © 2011 by Taylor & Francis Group, LLC

CoverCalc calculates the cutting energy of the anchor chain into the moment up to the moment that the anchor arrives at the berm or the chain comes too close to the pipeline. This energy is expressed in the Integrated Protection Coefficient (IPC). Two effects are included in the IPC: the strength of the cover and the position of the area relative to the pipe (follows from the berm geometry). The cutting of the anchor through the cover is conservatively not included in the IPC. CoverCalc relates the IPC (the cutting energy) to the increase of the potential energy (the lifting) of the anchor. The model has been calibrated to experimental data of results of anchors through a rockfill berm in a pre-dredged berm. The maximum IPC, which can be mobilised by an anchor approaching at a certain depth, is determined. This is used to assess whether a given berm will provide enough protection when an anchor approaches. 2.1.2 Experimental research To investigate the behaviour of the anchor during the encounter with the rockfill berm, physical model test were carried out in the Dreding Flume of Deltares. In the Dredging Flume a test section with a length of 16.5 m and a width of 2.4 m was created. In this section, a clay layer was laid with a depth of 1.4 m. The anchors were pulled by the wagon present above the flume. The dragging speed during the tests was 0.44 m/s. Figure 1 shows a photo of a cover after testing in the dredging flume. These tests proved the adequacy of the cover that was designed for the Lamma pipeline, since the anchors passed over the pipeline without causing harm. By varying the berm geometry the tests were also used to assess a possible optimisation of berm width and height to reduce the total volume of required rockfill. As far as the damage mechanism was concerned they showed that most berm damage resulted when the chain and the anchor removed stones from

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Figure 2. Furrow caused by the moving chain.

Figure 1. Rockfill cover after testing.

the cover and created a furrow across the berm. It also became clear that for the investigated anchor (US Navy Stockless) the inherent instability of the anchor is an important parameter for the protection mechanism: it is not just the chain that is lifted and causes the anchor to pass over the pipeline. A case was recorded where the anchor made a 180 degree roll and surfaced on its own accord, not by the lifting of the chain. We believe that this instability of the anchor is more likely to occur when the berm is approached at an angle such that one of the flukes contacts the berm before the other or when the non-homogeneous character of the berm causes uneven forces on both flukes. At this stage numerical models to take these effects into account are not available.

2.2

Centrifuge test on interaction chain–rockfill

In 2005 Deltares carried out centrifuge tests to obtain both qualitative and quantitative insight in the processes which govern the cutting of a chain into a rockfill berm and the transport of rockfill created by the moving chain. In order to be able to investigate these processes in a representative and controlled manner a system was required which pulls an anchor chain with a controlled speed over and through a rockfill berm. A special designed test setup was designed to carry out these tests (Van Lottum et al. 2007). While moving the chain over the rockfill berm, the chain cut into the berm and was partially covered by berm material. The cutting action of the chain caused a significant amount of stone transport in the rockfill cover, which gradually led to the formation of a kind of trench across the cover and deposition of stones at the location where the chain exited the rockfill berm, see Figure 2. The interaction between chain and rockfill in this transport phenomenon depended on the size of the shackles in relation to the grain size distribution of the rockfill. Also the size of the shackles relative to the size of the berm plays a role as the capacity for transport is proportional with the size of the shackles. If the transported volume is negligible in comparison © 2011 by Taylor & Francis Group, LLC

to the volume of the cover, the transport phenomenon is not playing a significant role in the process and may be ignored. The tests showed that the chain behaves as a special case of long strip foundation. If one looks into the parallel movement (in the pulling direction) and the perpendicular movement (downwards, cutting into the berm) one can assume that per unit length of chain each ratio between these two displacement components will have a unique set of soil response forces. This concept is illustrated in Figure 3. The arrows indicate the ratio between parallel and perpendicular movement. Envelope A could represent the situation that the chain is lying on the berm, since no interaction force between berm and chain develops when the chain moves upwards (is lifted from the berm) and a relatively small force is needed for a pure perpendicular downward movement into the berm. Envelope B represents a situation where the chain is at some depth in the berm, since a breakout force is needed for an upward movement and in general higher forces are required for any of the movement directions, which is the result of the higher effective stress level around the buried chain. These tests showed also that the chain – rockfill size ratio is an important parameter in the penetration of the chain in the berm.The bigger the rockfill size the larger the resistance against penetration into the berm. The penetration of the chain is also characterised by the chain-saw like cutting and material transport process that disintegrates the berm. The correct modeling of material transport by the chain is therefore another essential part of any model, which is to predict movement of the chain trough the berm and to analyse the effect of a rockfill berm as a protective measure. Chain penetration rate is directly dependent on the curvature of the chain over the berm and the chain tension. The development of a model that predicts the performance of a rockfill berm as protective measure requires a proper model for the anchor forces, since these will determine the chain tension and therefore to a large extend the penetration speed.

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Figure 4. Shear planes in front of embedded anchor.

Figure 3. Concept of yield envelopes for a chain.

2.3 Anchor dragging tests in sand and clay in the centrifuge Deltares designed and constructed a model setup to be able to perform anchor pulling tests at model scale in a geotechnical centrifuge under controlled laboratory conditions. With this test setup anchor pulling tests can be performed up to 1.8 m in a predefined soil body of sand or clay. For a typical test at 100g this represents a dragging distance of 180 m in the prototype situation. During such tests, the soil phenomena and the behavior of the anchor can be monitored by video and various kinds of transducers. In 2008 two preliminary anchor dragging tests with a Vryhof Stevpris anchor were carried out to show the possibilities of the test setup. One test was carried out in sand and one in clay overlying a sand layer. These tests are described in detail byVan Lottum et al. (2010). In both tests, the anchor was placed on top of the sand and was dragged from this initial situation with a speed of 100 mm/min into the soil. The test in sand showed that the anchor slowly digs into the sand creating a pattern of shear planes in front of the flukes see Figure 4. The development of the pulling force shows small drops in the pulling force which correlates with the observed shear planes. The values measured in these test were in agreement with the values mentioned in the Vryhof Anchor Manual (2010). After the dragging test in clay (overlying a sand layer), the clay surface showed a small trough with some heave above the location of the anchor. After exposing the anchor it became visible that anchor had © 2011 by Taylor & Francis Group, LLC

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Figure 5. Scratch marks of the flukes on the sand layer after removing the clay layer.

not penetrated the sand layer under the clay. Probably the downward force exerted by clay on the flukes is too low to penetrate into the stiff sand. The scratch marks of the flukes showed that the anchor was dragged with the pointed ends of the flukes through the top of the sand, see Figure 5. The test in clay showed that the boundary between sand and the clay limited the penetration depth and therefore the anchor reached an equilibrium drag force at a reduced depth. This illustrates the need for an anchor-soil interaction model that takes the effect of strength variation between subsequent layers into account. 2.4 Anchor dragging test with SEA-anchor In 2008 model tests were performed on suction embedded anchors (SEA’s) in a stiff clay. The (steel) wire that was attached to the SEA cut through the clay and created a plane over which the clay could be split after the test. The strands of the wire left marks in the clay as shown in Figure 6. The path that an anchor follows depends, among other factors, on the interaction between the anchor chain and the soil. The available numerical models

the soil. This shape is derived by an analysis of the equilibrium of forces that act on the chain. The tension force in the chain may vary along the chain when the weight of the chain and the friction between the chain and the soil in the direction parallel to the chain are accounted for. In this type of model, when the soil is homogeneous in the horizontal direction, it can be assumed that the movement of the chain always has a component perpendicular to the chain and a component parallel to the chain. Consequently, it is often assumed that both the full perpendicular resistance and the full parallel friction are mobilized by the chain during the penetration and dragging process. The typical description of the anchor is based on a theoretical or experimental determined capacity that depends on the depth below the surface and the soil properties that are found at that depth. The orientation of the anchor follows from an equilibrium of the weight of the anchor, the chain force exerted at the shackle and the soil resistance.

Figure 6. Travel marks in the clay.

known to date assume that the chain always mobilises the full lateral resistance. The marks in the clay in Figure 6 show that during a phase that the force at the shackle is temporarily reduced, the direction of wire displacement changes to more parallel (slip) and less perpendicular penetration of the wire. In accordance with the concept of a yield-envelope, where parallel and perpendicular reaction forces vary and depend on the direction of displacement, the transitions from I → II → III → IV in Figure 6 coincides with the positions on the yield envelopes 1 – 2 – 2 – 1 in Figure 3.

3

EVOLUTION OF DESIGN METHODS

For the design of mooring systems on the one hand and protection systems for pipelines and cables on the other hand a good understanding of the relevant mechanisms is required. While significant progress has been made on the analytical and numerical descriptions of anchor installation, embedment and holding capacity (Neubecker & Randolph 1996, DNV 2007) physical testing remains very important (Ozmutlu, 2009) as the intricate geometries of the anchor systems do not lend themselves easily for analytical or numerical analyses. The authors feel that the way forward lies in the proper blend and use of numerical and experimental testing tools. Below a description of potential numerical tools and the supplemental experimental activities that provide an interesting perspective for the development of future design method is given.

3.1 “Simple” 2-D models The basic models that are available now describe the path of the anchor chain and the anchor during embedment in a laterally homogeneous soil. In these models, the interaction between chain and soil is characterized by the lateral (perpendicular) resistance that the anchor-chain experiences in the soil, which together with the weight of the chain and the tension in the chain determines the inverse catinary shape of the chain in © 2011 by Taylor & Francis Group, LLC

3.2

Improved chain modelling

Laterally non-homogeneous seabed geometries are needed to model a cover over a pipeline or cable.These, in turn, require an improved chain model: If the chain comes into contact with a stronger part of the subsoil (or a rock cover) the perpendicular resistance that is mobilized by the chain may prove to be well below the maximum resistance that might be mobilized in that soil. Simpler models that always assume the maximum perpendicular resistance to be mobilised will not suffice in that situation. Interaction between chain and soil (or chain and rockfill) needs to be modeled in such a way that not always the maximum perpendicular resistance is mobilized, but that also slip dominated movement can be described. Only then a chain that does not penetrate into the rockfill can be correctly modeled. Also situations where the resistance of the anchor is (temporarily) reduced are characterized by a (temporary) slip dominated chain movement. Reference is made to Figures 3 & 6 for illustration of this concept. This involves the implementation of a “failure envelope” that describes the interaction between the parallel and perpendicular movement of the chain in the soil. The full perpendicular capacity is only mobilized if the chain movement is purely perpendicular and the full parallel friction is only mobilized if the chain movement is purely parallel. Yield envelopes that describe these interactions will show that the soil reaction force on the moving chain is not in line with the direction of the chain movement, but that the ratio of perpendicular to parallel force tends to be larger than the ratio of the perpendicular to parallel movements. While quite experimental work is available that focuses on the lateral resistance of chains in the subsoil (Degenkamp & Dutta, 1989), more work is required to determine the complete yield envelopes of chains

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and wires in soils and rockfill and build proper yield envelope models. A feature that is more difficult to implement in this type of model is the effect of layer transitions in the direct vicinity of the chain on the yield-envelope. While for chains the over-all effect of such layer transitions may be relatively small, In view of the relative small diameter of the chain or wire the distance over which layer transitions have an influence is small and this effect may be neglected. For anchors (which are much larger) this is not the case. 3.3 Anchor models

3.4 Transport phenomena The effect of rockfill transport from the cover by the chain that is passing over or through the cover was illustrated by centrifuge tests (Van Lottum et al. 2007). If a chain is dragged over the berm for a long enough period, the berm deforms and may further and further be penetrated by the chain. Any structure under the berm, that needs protection, like a pipeline or a cable, would have to be placed in a trench, well below the ground surface. As the change in cover geometry will affect the yield envelope for the chain, the chain-induced transport needs to be accounted for in the models.

Many basic anchor models assume a direction of movement in the direction parallel to the flukes, combined with a rotation that enables the anchor weight and soil reaction forces on the anchor to make equilibrium with the chain force that is exerted on the anchor. The reaction forces may be derived in a theoretical manner, by analyzing the separate contributions parallel and perpendicular to the flukes, crown and shank. When considering the special shapes of some of the existing (and popular) anchors it will be clear that such theoretical analyses may have severe shortcomings. By comparison of actual anchor penetration and dragging tests with prediction of the models the description of the overall anchor-soil interaction may be optimized for a single anchor and a single soil type. Similar to the description of the interaction between chain and soil with a failure envelope a much more versatile modeling of the anchor-soil interaction is possible with a yield-envelope approach. Unlike the chain-soil model, which (in 2D models) only needs two degrees of freedom to describe the relation between lateral and parallel movements and forces the anchor yield-envelope model needs three degrees of freedom, since it also has to describe the interaction with the rotation of the anchor and the resisting moment against this rotation. As an alternative to theoretical derivations of the yield-envelope of the anchor model tests one could use to determine and construct such yield envelopes for arbitrary anchors and ideally create a database with tested and verified yield envelopes for various anchor types. A complication that the anchor yield envelope has is that, much more than the chain-soil models, the anchor “feels” the effect strength gradients, layer transitions and the seabed surface. One approach to deal with this is to “subdivide” the contributing parts of the anchor (e.g. the flukes and shank), into individual parts and assign these parts resistances which reflect the soil that the particular anchor piece is in contact with. An approach along these lines was described by Ruijnen (2005) for a clay that had a shear strength that increased with depth. That there is a need to deal with the effect of layer transitions on the anchor behavior is demonstrated by the anchor pulling experiments described earlier in this paper (Figure 5). © 2011 by Taylor & Francis Group, LLC

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3.5 The “ideal” 2-dimensional model From the descriptions above the ideal 2-D model for anchor-chain-soil-rockfill interaction can be defined. It contains provisions for: • An arbitrary subsoil geometry. • Chain-soil interaction that is based on a 2 D.O.F.

yield-envelope approach. • Anchor-soil interaction that is based on a 3 D.O.F.

yield-envelope approach and accounts for layer transitions and surface effects. • A way to assess the changes of the resistance in a rockfill berm due to the loss of stones that are being removed from the berm by the chain itself. 3.6

Extension to 3D and discrete element models

Most anchors that are being pulled and penetrate into the seabed will not do this in the perfect vertical plane that forms the space in which the 2-D models are described. Out of plane pulling and rotation or destabilization of the anchor is to be expected in practice. Clearly 2D models are inadequate to describe this. When considering a rockfill protection over a pipeline or cable a perpendicular approach by the anchor will be exception rather than rule. When cutting across a pipeline cover at an angle (not perpendicular) the effective width of the cover increases and the chances that the anchor damages the pipeline reduce. Moreover, apart from the actual resistance that the cover mobilizes against the chain one of the mechanisms that prevents the anchor from penetrating the berm and damaging the pipeline is the destabilisation of the anchor. This can be triggered by unequal forces acting on the flukes of the anchor, which would happen when one of the flukes reaches the cover before the other. Another cause for differential forces lies in the fact that the cover, consisting of relatively large stones, is not acting as a homogeneous continuum but has locally stronger and weaker spots. While the first effect might be handled by a true 3-D anchor model (which requires 6 degrees of freedom) and a true 3-D chain-soil model, for the latter effect the way forward would lie in the employment of Discrete Element Models (D.E.M.). Such models also have potential for a realistic analysis of the rockfill interaction and transport phenomena, which are believed

to have large influence on the protective properties of the cover (Gaudin et al. 2007). 3.7 The “ideal” 3-dimensional model The ideal 3-D model would enable the analysis of anchor break-out, which is typically effectuated by backwards pulling of the chain when done on purpose or sideways pulling when in severe loading conditions one or more mooring lines of e.g. a SBM-system have failed. It would also enable the analysis of a vessel riding on a single anchor in situations where the anchor loading direction changes due to variations in wind or flow direction. The main extension of the 2D model lies in the more complex description of the yield envelopes and the generalized numerical analysis of the anchor chain and anchor in three dimensions. 4

CONCLUSION

A number of improvements are possible in the numerical anchor-chain-soil-rockfill interaction models. These improvements require experimental input and verification, in particular to define proper yield envelopes for anchors and chain, both in soil and rockfill, and for the understanding of the transport phenomena that have been observed. Despite improvements in analytical techniques, verification of actual designs by means of physical modeling will be required in the time to come. There is an obvious conservatism in the assumption that all anchor–pipeline incidents occur at perpendicular tracks. Designers should try to explore this conservatism and verify the extra margin experimentally to arrive at more economic cover designs.


REFERENCES Degenkamp, G &Dutta, A. 1989, Soil Resistances to Embedded Anchor Chain in Soft Clay, Journal of Geotechnical Engineering, (115)10, 1420–1438. DNV 2007, “DIGIN User Manual”, Technical Report 20070882, Joint Industry Project, Det Norske Veritas, Hovik, Norway. Gaudin, C., Vlahos, G., Randolph M.F., 2007, Investigation in centrifuge of anchor-pipeline interaction. International journal of offshore and polar engineering , 17 (1) 67–73. Neubecker, S.R. & Randolph, M.F. 1996, The Performance of Drag Anchor and Chain Systems in Cohesive Soil. Marine Georesources and Geotechnology, 14: 77–96. Ozmutlu, S. 2009. The Value of Model Testing in Understanding the Behaviour of Offshore Anchors: Towards New Generation Anchors. Offshore Technology Conference, 4–7 May 2009. OTC-20035. Ruijnen, R.M. 2005, Influence of Anchor Geometry and Soil Properties on Numerical Modeling of Drag Anchor Behavior in Soft Clay. 1st ISFOG, Perth. Van Lottum H., Luger H.J., Bezuijen A., 2007, Interaction between anchor chains and rockfill tested in a centrifuge model. The Proceedings of the 17th International offshore and polar engineering conference, Lisbon, Portugal, July 1–6, 2007. Van Lottum, H., Luger H.J., Bezuijen, A., 2010. Centrifuge anchor dragging tests in sand and clay. Proceedings of the International Conference on Physical Modelling in Geotechnics 2010. Vryhof Anchors B.V. 2010, Anchor Manual 2010, Capelle a/d Yssel, The Netherlands.

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Development of a jet trenching model in sand J-F. Vanden Berghe Fugro Engineers SA, Belgium

J. Pyrah & S. Gooding CTC Marine Projects, UK

H. Capart National University of Taiwan, Taiwan

ABSTRACT: This paper presents a jet trenching model applicable to realistic jetting configurations that has been developed based on earlier research focussing on idealised jetting configurations (Vanden Berghe, 2008). Specifically, oblique and upright swords featuring multiple jets of different orientations are considered here. The swords are modelled after the jetting swords of the jet trenchers operated by CTC Marine Projects for cable and pipeline burial tasks. The jet trenching model has been validated and calibrated on the basis of a set of laboratory small scale tests. The tests were conducted at scale 1:30 with jetting tools modelled after the swords of CTC jet trencher. The experiment analysed the parameters influencing the trenching performance. It included the total jetting power, the progress rate of the trencher and the sand bed properties. Comparison between laboratory and model results were presented and discussed. The results of this research make it possible to better understand the mechanisms controlling the performances of pipeline burial operations by jet trenching. The proposed model helps the engineers in assessing the performance of the jet trenching machines. 1

INTRODUCTION

The main issue of a trenching process is to choose the right trencher and the right trencher configuration in order to optimise the trenching process duration, as well as to use a tool suited for given soil conditions. Usually the aim is to lower the product to a given target depth in the minimum time. The entire trench process duration is function of many parameters, which include: the number of passes required to lower the product, and the trencher speed of advance. This paper presents a jet trenching model assessing the burial depth achieved after each pass of the trencher taking into account the trencher performances (available jetting power and swords configuration), the soil condition (grain size and density) and the product characteristics (weight and stiffness). 2

GENERAL DESCRIPTION OF THE JET TRENCHING MODEL

3

The jet trenching model is based on the combination of the 2 following models: The first model assesses the shape of the trench created by the jet trencher. This model is called the multiple jets trenching model. The assumptions and formulation of the model are described in Section 3 together with the laboratory test results used to validate and calibrate the model. © 2011 by Taylor & Francis Group, LLC

The second model computes the likely shape adopted by the pipeline when the soil supporting it has been removed. This model is called the pipeline model and is described in section 4. The 2 models are combined in order to simulate the trenching process. The interaction between the two models allows the assessment of an appropriate trencher speed of advance and number of passes, for given soil, pipeline and trencher properties. The main assumption of the model presented above is the independence of the trenching model and the pipeline model. The trenching model ignores the presence of the pipeline and the pipeline model does not consider the hydrodynamic forces induced by the jetting process in the computation of the pipeline deflection. This assumption holds for small diameter or heavy products but may not for very large diameter or light products.

3.1

MULTIPLE JETS TRENCHING MODEL Model basis

Several laboratory observations (Su, 2007, Vanden Berghe, 2008) indicate that the interactions between the jet-induced current, ambient seawater and sand seabed can be characterised into five different processes: entrainment, erosion, deposition, breaching, and overspill (as illustrated in Figure 1).

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Figure 2. Experimental setup (Vanden Berghe, 2008).

Figure 1. Key physical process during jet trenching.

In the longitudinal direction, the dynamics of the jet-induced turbulent current are assumed to be controlled by entrainment of ambient water into the current, erosion of seabed material (water + sediment) into the current, and deposition of material due to the gravitational settling of sand grains back to the bed. From their initial erosion to their eventual deposition, suspended sand grains cause the current density to exceed the density of seawater, and the corresponding density stratification tends to damp turbulent mixing. Weakened eddies then lose their ability to entrain ambient water and erode bed material. The model needs to capture this mechanism. In the lateral direction, the trench cross-sectional shape is affected by the breaching of sidewalls. Breaching is a type of underwater slumping driven by reduced specific gravity and paced by the rate of infiltration of seawater required for sand bed dilatancy to take place. When the sediment-laden current rises above the trench walls, moreover, overspill of suspended sediment out of the trench occurs. This is important in practice because it causes a loss of sand cover over the cable or pipeline, decreasing the level of protection achieved even if a deep trench has been incised. The model development proceeded in the following steps. Based on prior experience with 2D horizontal plane jets (Perng, 2006; Perng and Capart, 2008), a model for a single 3D round jet travelling along the seabed was developed. This model incorporated all five processes (entrainment, erosion, deposition, breaching and overspill) and took stratification into account. This model was validated by comparison with a series of small-scale lab experiments involving point jets originating from a travelling needle (Su et al., 2007; Su, 2008). This model was then extended to multiple jets, injected at various locations and orientations, and coalescing into different currents inside the evolving trench. This in turn requires the model to be able to handle trench cross-sections of arbitrary shape.

the mechanism involved in the trenching process and to guide the development of the theoretical model that would represent these mechanisms as much as possible. The apparatus used for the experiments is illustrated in 3.2.2. A wheeled carriage travelling above the tank at a given speed supports an articulated arm ending with a jetting device. Several jetting devices were used going from a thin syringe needle to swords modelled after the jetting swords of the jet trenchers operated by CTC Marine Projects. More than 100 tests were conducted investigating mainly the influence of the jet configuration (sword inclination, jet orientation, rear jet,…), jetting power, travelling speed, grain size of the sediments and seabed density. All experiments took place under submerged conditions. The conversion from the laboratory scale to the prototype scale assumes a geometrical scale factor of 30. The non-geometrical variables (velocities, discharge and power) are scaled according to Froude similarity, which preserves the relative influence of inertial, density and gravity effects (see Vanden Berghe; 2008). In model tests governed by Froude similarity, problems may be encountered due to the effects of surface tension and/or viscosity. In the present case, these two effects are not expected to cause significant difficulties. The exception regards the possible influence of seabed permeability on the erosion and breaching response of the sand grains. In small-scale tests, it is not possible to simultaneously reproduce permeability and Froude number effects. For this reason, the experiments were conducted with 2 sands, allowing the possible effects of permeability to be estimated and corrected for in the numerical model. Two sand materials were used for the tests: very fine sand (d50 = 0.08 mm) and fine sand (d50 = 0.17 mm). Scaled to prototype size based on the fall velocity of the grains, the very fine sand is equivalent to medium sand (d50 = 0.17 mm) and the fine sand to coarse sand (d50 = 0.99 mm). Additional information on the laboratory tests are provided in Su (2008), Su (2007) and Vanden Berghe (2008).

3.2 Laboratory experiments 3.2.1 Experimental setup (Vanden Berghe, 2008) While much can be learned from trencher operation data, visualisation of the sediment motion is difficult in field conditions. This motivates the use of smallscale laboratory experiments. The focus is to examine © 2011 by Taylor & Francis Group, LLC

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3.2.2 Experimental results Many tests were performed with different testing parameters.An example of results is shown in Figure 4. For convenience and an easier comparison with actual operational performances, the results are presented at prototype scale.

Figure 5. Shallow layer jet-induced suspension flow model. • The length of open trench varies approximately lin-

early with the total flow rates supplied to the jetting tools. For a given jetting configuration, an increase in flow rate of about 25% associated with doubling the jetting power leads to an increase in trench length of about 20 to 25%.

Figure 3. Picture of the laboratory jetting tool.

3.3

Figure 4. Influence of progress speed and grain size on trenching performance (jetting power = 1.5 MW).

The results of Figure 4 are for inclined jetting swords with vertical downward facing jets. The jetting power is 1.5 MW. The figure compares the results for 2 grain sizes (d50 = 0.17 mm and 0.99 mm) and several progress speed varying between 200m/hr and 1,200 m/hr. The main findings of the experimental campaign can be summarised as follows: • The trench lengths achieved by a given set of jetting

tools depend greatly on sand size and to a lesser degree on the density of the seabed. Finer sands lead to longer trenches by slowing the pace of resedimentation. Denser beds lead to longer trenches by reducing the pace of sidewall breaching and slumping. • Regardless of conditions, giving a downward orientation to the forward jets leads to increased maximum trench depths. However, this increase in depth is limited to the immediate vicinity of the front jetting sword. An increase in localised trench depth does not translate into an overall increase in trench length. • For loose beds, the orientation of the forward jets appears to exert little to no influence on the ultimate length of open trench maintained behind the sword. Various configurations of the front sword lead to identical backfill profiles to the rear of the trencher. © 2011 by Taylor & Francis Group, LLC

Numerical model

3.3.1 Main assumptions The theory envisions a slender current of water and suspended sand flowing along a sand bottom of variable topography submerged in deep water ambient. The following vertical flow structure is assumed (see Figure 5). Motion is confined to a fully turbulent layer of thickness ht . The overlying ambient water is assumed quiescent, and the underlying solid-like sediment bed is assumed static. Embedded within the turbulent layer is a fluidised layer of sand and water of thickness hs , inside which the weight of the dilute sand phase is entirely supported by turbulent eddies. It is assumed that the interfaces between the layers are sharply defined and that the distribution of flow properties (longitudinal flow velocity u, turbulent kinetic energy k and sediment concentration c) is uniform within a given layer. Transfers of mass and momentum between the layers and across the interfaces are however allowed to take place. It is assumed that vertical accelerations are negligible, and that the vertical velocities are cinematically constrained by the thickness variations of the different layers. Vertical velocities are thus allowed, but not explicitly included in the description. The theory thus involves the following five variables, or degrees of freedom (see Figure 5): • Thickness of the turbulent bottom current (hs ); • Thickness of the fluidised layer in which sand is

entrained by the turbulent water (ht );

• Elevation of the loose, static sediment bed (zs ); • Sediment concentration inside the fluidised layer

(Ct );

• Longitudinal velocity of the water and sand inside

the turbulent layer (ut ). 3.3.2 Governing equations All five variables are function of both longitudinal coordinate and time, and their coupled evolution must be described by a set of five governing equations.

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These equations are obtained by applying mass conservation and momentum balance to the various layers. The different layers defined hereabove do interact. This interaction is characterised by the 3 following exchange fluxes: • The flux ew of the entrainment of quiescent water

into the turbulent current is computed on the basis of common model in hydrodynamics, i.e. proportionally to the current velocity:

where E is the entrainment coefficient, b is the effective width of the turbulent layer flowing along the trench bottom and u is the velocity of the turbulent layer • The flux et between the 2 turbulent layers only depends on the settling velocity of the grains:

Figure 6. Processes influencing the transversal section (a) vertical and lateral erosion of the bottom current (b) deposition of sediments, (c) breaching of the trench wall.

The evolution rules used to update the trench shape under the influence of the various processes are illustrated in Figure 6. Three processes can be identified:

where ωs is the effective fall speed of the sediment grains • The flux es between the seabed and the turbulent current is governed by the following equation

The first term is a turbulent erosion rate, formulated by analogy with turbulent entrainment (i.e. proportional to the current speed), and the second term is a deposition rate due to settling of sediment grains back to the seabed. Dimensionless coefficient ε (0 ≤ ε ≤ 1) depends on the size of the sediment grains and expresses the relative ease with which they can be eroded by the current. Upon erosion, the trenched seabed transfers sand to the current (es > 0), whereas the opposite occurs upon deposition (es < 0). The parameter θ denotes the angle below horizontal formed by the tangent to the bottom profile of the trench Turbulence, which controls both entrainment and erosion, is strongly damped by density stratification. In other words, the erosion flux and the entrainment flux decrease and could possibly be equal to zero when the density and the thickness of the turbulent layer increase. This is taken into account by expressing the entrainment coefficient E as a function of the Richardson number, which accounts for the increase of the potential energy relating to the kinetic energy. 3.3.3 Lateral variation of the trench The last component of the model is the procedure used to evolve the cross-sectional shape of the trench. Because multiple processes are considered, it was found that the trench shape can not be limited to a predefined geometry with limited degrees of freedom (e.g. a trapeze with variable top and bottom widths). Instead, the trench shape should be able to evolve freely. For these reasons, it was decided to characterise the trench transversal profile by a polygonal line allowing, for example, the trench to be wider at its bottom than at its surface, as observed in some experiments. © 2011 by Taylor & Francis Group, LLC

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• The first process that influences the shape of the

transversal section of the trench is the lateral erosion of the bottom current. As the turbulent current erodes the bottom of the trench, it also erodes the side walls. The lateral erosion is slower than the vertical one and it can reasonably be assumed that it varies proportionally with the trench depth as illustrated in Figure 6a. • The second process is the redeposition of the sediments that is assumed to be horizontal and uniform on the trench width (see Figure 6b) • The last process is the breaching i.e. the collapse of the side wall towards the natural equilibrium slope of the sand. This mechanism acts to widen the trench independently of jetting action and depends on the speed of retreat of a vertical trench wall λ0 , the natural angle of repose of the sand and the slope of the trench wall. Parameter λ0 is dependent on the properties of the seabed material (void ratio and permeability). The approach is similar to the breaching speed formulated in Mastbergen and Van den Berg (2003).

3.4

Comparison of model with experimental data

Model predictions and experimental results were compared. An example of comparison is shown in Figure 7. Both quantitatively and qualitatively, a good level of agreement is recorded between computation and experiment. Trench shapes and dimensions are well reproduced, and their responses to parametric variations are well captured by the model. The same model parameters are used for all jetting configurations, without case-by-case tuning. For the two different sands, only three parameters must be set differently: the single grain fall velocity ωs , the erodibility ε, and the breaching speed λ0 . Provided these values are set adequately, the contrast between the 2 sands analysed is well modelled. For the coarser sand, the trench is shorter, can be trenched at higher progress rates U and exhibits less sand loss out of the

Figure 9. Influence of the lay tension on the pipeline deflection.

Figure 7. Comparison between experimental and model results (jetting power = 1.5 MW – progress rate = 200 m/hr).

resolved analytically and the following equation of the pipeline deflection can be obtained:

where y = Pipeline Deflection (positive downwards) (m), q = Pipeline Submerged Unit Weight (N/m), T = Lay Tension (N), L = Beam Length (m), E = Elasticity modulus (N/m2 ), I = Moment of Inertia of the Beam Cross-section (m4 ) and x = Distance from beginning of Trench (m). As illustrated in Figure 9, taking the lay tension into account is very important to assess the likely shape of the product behind the trencher.

Figure 8. Equivalent pipeline model.

trench due to overspill. The model accurately captures these influences of sand size on the trenching response.

4

PIPELINE DEFLECTION MODEL

In order to assess the burial depth during trenching operations, it is required to compute the likely shape adopted by the pipeline when the soil supporting it has been removed by the trencher. The assessment of the pipeline deflection into the trench is based on an elastic beam model of the pipeline. The pipeline sinking into the trench is assumed equivalent to a hyperstatic cantilever beam uniformly loaded, in which the left-hand end is completely fixed and the right-hand end is restrained in rotation (as illustrated in Figure 8). The left-hand end simulates the pipeline lying on the seabed whereas the right-hand end represents the pipeline touchdown point in the trench. A lay tension is also applied at the right-hand position. After touching the re-sedimentation front, the pipeline displacements are assumed to be restrained. No pipeline settling is assumed in the re-deposited sand that still may be liquefied. The pipeline model described above can be resolved analytically. Because the lay tension in the pipeline applied a bending moment that depends on the pipeline deflection, the formulation required the used of a differential equation. This differential equation can be © 2011 by Taylor & Francis Group, LLC

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5

CASE STUDY

In an engineering environment the model can be used to either anticipate the performance of a given trencher in a variety of conditions or configure a trencher for optimal use in the soil conditions anticipated at the trenching location. The following case study is presented as an example of the use and effectiveness of the model in configuring a jet trencher for operations on a pipeline trenching project. The pipeline was located in the North Sea, the soil conditions were anticipated to consist of a layer (up to 0.85m thick) of very loose to loose fine silty sand overlying medium dense fine silty sand. The product was a 10” rigid steel pipe, which the client required to lower to 1.0 m below seabed. Several iterations of the model were run to identify the optimum jet leg configuration and the optimum trencher progress rate. The model was run with upper and lower bound soil conditions to account for the vertical and lateral variability of the soils. Figure 10 illustrates the output from the model at the optimal conditions. A depth of lowering of 1.0 m below seabed with at least 0.8 m cover was predicted at a forward progress rate of 275 m/hr. During operations at sea the trencher was operated at a progress rate of as close as possible to 275 m/hr and the product lowered to between 1.0 m to 1.2 m

Based on these observations, a model using the gravity- and jet-driven turbidity currents theory has been developed. This model is able to simulate the experiments quite well and highlights the importance of the different mechanisms. It has also been successfully used for several trenching projects in the North Sea and can be used to assess the best jetting configuration as illustrated in Figure 12. REFERENCES

Figure 10. Illustration of model use for burial assessment.

Figure 11. Depth of lowering achieved.

Figure 12. Illustration of model capability to investigate the benefit of rear jets.

below seabed with a cover thickness of between 1.0 m and 1.2 m. Hence, the model was found to provide a reasonably close but slightly conservative prediction for lowering in this instance (Figure 11). 6

Ahrens, J.P. (2000) A Fall-Velocity Equation. Journal of Waterway, Port, Coastal, and Ocean Engineering, Vol. 126, No. 2, pp. 99–102. García, M.H., and Parker, G. (1991) Entrainment of bed sediment into suspension. Journal of Hydraulic Engineering, Vol. 117, No. 4, pp. 414–435 Hughes, S.A. (1993) Physical Models and Laboratory Techniques in Coastal Engineering. World Scientific. Mastbergen, D.R., and Van den Berg, J.H. (2003) Breaching in fine sands and the generation of sustained turbidity currents in submarine canyons. Sedimentology, Vol. 50, No. 4, pp. 625–637. Perng, A.T.H. (2006) Trenching of underwater sand beds by steadily moving jets. PhD thesis, Graduate Institute of Civil Engineering, National Taiwan University. Perng, A.T.H., and H. Capart (2008) Underwater sand bed erosion and internal jump formation by travelling plane jets. Journal of Fluid Mechanics, Vol. 595, pp. 1–43. Spence, K.J., and Guymer, I. (1997) Small-scale laboratory flowslides. Géotechnique, Vol. 47, No. 5, pp. 915–932 Su, J.C.C., A.T.H. Perng, and H. Capart (2007) Underwater trench incision and turbid overspill due to moving point jets. Proc. XXXII Congress IAHR, Venice, Italy, July 2007. Su, J.C.C. (2008) Seabed trenching by moving point jets: experiments and theory. MSc thesis, Graduate Institute of Civil Engineering, National Taiwan University. Vanden Berghe J-F., Capart H. and Su J.C.C., (2008), Induced Trenching Operations: Mechanisms Involved, OTC-19441, proceeding of 2008 Offshore Technology Conference held in Houston, Texas, U.S.A., 5–8 May 2008, Van den Berg, J.H., Van Gelder, A., and Mastbergen, D.R. (2002) The importance of breaching as a mechanism of subaqueous slope failure in fine sand. Sedimentology, Vol. 49, No. 1, pp. 81–95.

CONCLUSIONS

The laboratory experiments showed that the jetting performance of a jet trenching tool depends mainly on five mechanisms: entrainment, erosion, deposition, breaching, and overspill. These mechanisms control both the trench depth and the trench length.


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12 Design and risk



Structural factors affecting the system capacity of jacket pile foundations J.Y. Chen & R.B. Gilbert The University of Texas at Austin

J.D. Murff Consultant

A.G. Young Geoscience Earth & Marine Services

F.J. Puskar Energo Engineering

ABSTRACT: This paper presents an analysis of various structural factors affecting the system capacity of jacket pile foundations. We use a simplified foundation model based on the upper-bound plasticity theory and a three-dimensional finite element model of the platform structure to analyze these factors. Well conductors can contribute significantly to the shear capacity of a foundation system. Conductors can also contribute to the overturning capacity when their moment capacities are large relative to the piles. The foundation capacity in shear is generally more sensitive to the strength of the steel in the piles than the strength of the soil. Jacket leg stubs penetrating below the seafloor can increase the shear capacity of the foundation system. The rigidity and strength of the jacket can limit the full mobilization of foundation system capacity. Foundation system redundancy depends on the number of piles, the geometry of the pile system, the direction of the load, and the ratio of shear force to overturning moment in the load. Collectively, these structural factors have a significant impact on foundation performance. Piles supporting the platform and the platform structure itself need to be considered as a complete system to achieve an effective and consistent level of foundation performance. 1

INTRODUCTION

Table 1.

This paper presents an analysis of various structural factors affecting the system capacity of jacket pile foundations. Jacket platforms are fixed base offshore structures used in relatively shallow water (less than 450 m) to produce oil and gas worldwide. These platforms are generally supported by open-ended steel pipe piles driven through the legs of the jacket and connected to the jacket above the sea level. Occasionally, the annuli between the jacket legs and piles are grouted to enhance their connections. These platforms are often equipped with well conductors. Geotechnical engineers often pay more attention to the axial and lateral capacities of a single pile than the overall behavior of the pile system. Furthermore, structural factors, which are as important as geotechnical factors, are sometimes overlooked in determining the capacity of the foundation system. The effects of well conductors, jacket leg stubs, yield stress for piles, structural rigidity and system redundancy on the capacity of the foundation system are investigated in this paper. This study is part of a research project undertaken to compare the predicted and observed performance of jacket pile foundations in recent hurricanes in the Gulf of Mexico (OTRC 2009 and Gilbert © 2011 by Taylor & Francis Group, LLC

Key parameters for referenced platforms.

Platform number

Water depth (m)

Number of piles

Number of conductors

Foundation plan geometry

9 10 22 30 31

19 109 34 47 30

4 3 4 6 8

1 1 None 12 None

Square Triangular Square Rectangular Rectangular

et al. 2010). Platform numbering used in this paper is kept the same as that in OTRC (2009) to facilitate reference to the report. Key parameters for the platforms referenced in this paper are presented in Table 1.

2

MODELS FOR ANALYSIS

In this study we used a simplified foundation model based on the upper-bound plasticity theory (Murff & Wesselink 1986, Murff 1987, Tang & Gilbert 1992, Murff 1999, and OTRC 2009) to analyze various structural factors affecting foundation system capacity. We

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Figure 1. Schematic of the simplified upper-bound plasticity model for the foundation system.

also used a three-dimensional (3-D) structural model, SACS™, based on the finite element method to perform pushover analysis to estimate the capacity of the structural system including the foundation. The 3-D model was also used to infer the hurricane load on the foundation from the hindcast wave, wind and current conditions in the hurricane. Figure 1 presents a schematic of the simplified foundation model. This model assumes a plastic collapse mechanism, where all elements of resistance are characterized as rigid and perfectly plastic. This model incorporates both the structural capacities of piles and conductors and the axial and lateral capacities of soils. The piles and wells in the system collapse when one hinge forms at the head, where it is constrained by the bottom of the jacket leg for a pile or the bottom row of conductor guide framing for a well, and a second hinge forms at some depth below the head. The collapse of the entire system occurs when two hinges form in each of the piles and wells in the system due to the translation and rotation of the platform base (Fig. 1). The structure supported by the piles is assumed to be perfectly rigid and infinitely strong so that it will not fail and can distribute the load as necessary to develop the full foundation collapse mechanism. Well conductors are modeled as piles that are connected to the structure with rollers so that they can only be loaded horizontally. The solution provides an upper-bound approximation to the system capacity because it does not explicitly satisfy force and moment equilibrium. The best upper-bound solution, hereinafter designated as the solution, is the mechanism that incorporates a combination of base translation and rotation that gives the minimum system capacity. Figure 2 is an interaction diagram showing the upper-bound foundation system capacity of Platform 22 in the end-on direction. This platform has four legs, each supported by a 1067-mm (42-inch) diameter pile battered in two directions. Its foundation system capacity is represented by a base shear versus overturning moment interaction curve. Comparisons between the upper-bound solutions and results from more rigorous 3-D pushover analyses of jacket structures indicate that the upper-bound model overestimates the base shear and overturning moment causing foundation failure by approximately 10% (Murff & Wesselink 1986 and OTRC 2009). As such, the upper-bound © 2011 by Taylor & Francis Group, LLC

Figure 2. Upper-bound and expected foundation system capacities of Platform 22 in the end-on direction as compared to the hurricane load and pushover capacity according to the 3-D model.

capacity was reduced by 10% to obtain the expected foundation system capacity. The pushover failure of this platform is dominated by piles plunging and pulling out. As expected, the pushover failure load matches up very well with the expected foundation system capacity (Fig. 2). The potential foundation failure can be dominated by shear, where the lateral capacities of the piles and conductors contribute mostly to the foundation system capacity, or by overturning, where the axial capacities of the piles contribute mostly to the system capacity. Both structural and soil capacities work together to provide axial and lateral capacities for the piles and conductors in the system in these failure modes. The failure can also be dominated by a combination of shear and overturning. These distinct regions of foundation behavior are shown along the interaction curves (Fig. 2). In the shear failure mode, the base shear increases with increasing overturning moment due to pile batters. The potential failure mode of the foundation of Platform 22 is a combination of shear and overturning, which also agrees with the results from 3-D pushover analysis.

3 WELL CONDUCTORS Well conductors can contribute significantly to the shear capacity of a foundation system. The structural capacity of the outer casing for the well conductor is modeled herein to represent the minimum contribution of the conductor to foundation system capacity. Figure 3 presents the foundation system capacity interaction curves of Platform 30 showing the effect of the twelve 610-mm (24-inch) diameter conductors on the foundation system capacity. This platform has six legs, each supported by a 1219-mm (48-inch) diameter pile. The shear capacity of the foundation system (i.e. the base shear at a low overturning moment) including the conductors is about 30% higher than that excluding

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Figure 3. Expected foundation system capacities of Platform 30 in the diagonal direction.

Figure 5. Increase in foundation system capacity of Platform 22 in the end-on direction due to increasing steel yield stress.

including the only conductor is more than 2.5 times the shear capacity excluding the conductor (Fig. 4). Furthermore, the conductor increases the overturning capacity of the foundation system by about 20% due to its high moment capacity. The maximum load in Hurricane Katrina on the foundation is higher than the foundation system capacity excluding the conductor. Failures of the foundation system could have occurred but no sign of foundation failures was found, suggesting that the conductor probably contributed to the survival of this foundation. The potential failure mode of this foundation excluding the conductor is dominated by combined shear and overturning, while it is dominated by overturning when the large conductor is included.

Figure 4. Expected foundation system capacities of Platform 9 in the end-on direction.

the conductors (Fig. 3). In contrast, the conductors increase the overturning capacity of the foundation system minimally because they are not constrained to contribute their axial capacities and can only contribute their relatively low moment capacities. The maximum load in Hurricane Katrina on the foundation is about the same as the foundation system capacity excluding the conductors. The potential failure mode of this foundation is dominated by shear. The foundation system could have shown significant distress since it was nearly loaded to its capacity. However, no sign of distress was found on the foundation during underwater inspections, suggesting that the conductors probably contributed to the shear capacity of this foundation. Figure 4 presents the foundation system capacity interaction curves of Platform 9 showing the effect of the 1829-mm (72-inch) diameter conductor on the foundation system capacity. This platform has four legs, each supported by a 762-mm (30-inch) diameter pile. The conductor is more than two times larger in diameter than the piles, which are battered in two directions.The shear capacity of the foundation system © 2011 by Taylor & Francis Group, LLC

4

STEEL YIELD STRESS

The yield stress for steel tubular members used in offshore construction is higher than its nominal value on average (Energo 2009). The higher yield stress leads to higher shear and overturning capacities of the foundation system. Its effect is generally greater on shear capacity than on overturning capacity. Figure 5 presents the foundation system capacity interaction curves of Platform 22. The shear capacity of the foundation system assuming a yield stress of 345 MPa for the piles is about 30% higher than that assuming 248 MPa (Fig. 5). The increase in shear capacity is nearly proportional to the increase in steel yield stress. In contrast, the increase in steel yield stress only increases the overturning capacity of the foundation system minimally, because the overturning capacity is dominated by the axial capacities of piles in tension and compression rather than their moment capacity. The effect of steel yield stress is more significant in shear dominated failure mode when stronger soils are present near the mudline.

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Figure 7. Comparison of pushover failure load and foundation system capacity of Platform 31 in the broadside direction.

Figure 6. Increase in foundation system capacity of Platform 31 in the diagonal direction due to higher lateral soil resistance, increasing steel yield stress and modeling jacket leg stubs.

6 5

JACKET LEG STUBS

Jacket leg stubs penetrating below the seafloor can increase the shear capacity of the foundation system by constraining the first plastic hinges to form at the bottom of the leg stubs and pushing the second plastic hinges deeper below the mudline. These leg stubs allow the foundation system to mobilize more lateral soil resistance. Figure 6 presents a series of capacity interaction curves for Platform 31 showing the effects of pile yield stress, lateral soil resistance and jacket leg stubs on the system capacity. This platform consists of two structurally connected 4-leg jacket structures. The 914-mm (36-inch) and 1067-mm (42-inch) diameter piles supporting these jackets are battered in two directions. The base case foundation system capacity interaction curve was developed using a nominal yield stress of 248 MPa and cyclic p-y curves (Fig. 6). The maximum load in Hurricane Ike on the foundation is higher than the base case foundation system capacity and yet failures of the foundation did not occur. Rather, a few structural members of this platform above the mudline suffered damage. When piles are loaded laterally to failure under extreme loading conditions, such as those in a hurricane, they are pushed into undisturbed soils at large lateral displacements. Static p-y curves, which represent higher lateral resistance than cyclic p-y curves, are most suitable to model the lateral soil resistance (e.g. Jeanjean 2009). The intermediate capacity interaction curve developed using static p-y curves and a yield stress of 285 MPa, which is 15% higher than the nominal value, nearly surpasses the maximum load in Hurricane Ike. Furthermore, if the jacket leg stubs are modeled by constraining the first plastic hinges to form at the bottom of the leg stubs, the foundation capacity increases further (Fig. 6) and the survival of this foundation can well be explained. The effect of jacket leg stubs increases as the strength of the soil near the mudline increases. © 2011 by Taylor & Francis Group, LLC

RIGIDITY OF THE STRUCTURAL SYSTEM

The rigidity and strength of the jacket can limit the full mobilization of foundation system capacity. The simplified foundation model assumes a perfectly rigid and infinitely strong structure. Therefore, the foundation system capacity determined from this model represents the highest capacity that a structure can mobilize. Figure 7 presents the foundation system capacity interaction curve of Platform 31 and compares it with the pushover failure load in the broadside direction. Several mudline framing members were failed in the pushover analysis, resulting in a pushover capacity that is dominated by structural failures and lower than the foundation capacity. The rigidity and strength of connections between well conductors and the jacket can particularly affect the contribution of conductors to the overall capacity of the foundation system in shear (OTRC 2009).

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7

SYSTEM REDUNDANCY

Foundation system redundancy depends strongly on the number of piles, the geometry of the pile system, the direction of the load, and the ratio of shear force to overturning moment in the load. The sensitivity of foundation system capacity to the variation in the capacity of a single pile in the system reflects qualitatively the redundancy of the system. The more sensitive the system capacity is to the capacity of a single pile, the less redundant it is. Sensitivity analyses were performed by increasing and decreasing the axial and lateral capacities of one pile by 30% for Platform 10. The ±30% variation reflects possible variations in soil properties, driving conditions and structural properties between piles; it corresponds roughly to the coefficient of variation between measured and predicted capacities of individual piles (e.g. Tang and Gilbert 1992). Platform 10 is a 3-leg jacket structure supported by a system of

Figure 8. Foundation plan of Platform 10. Figure 11. Effect of Pile C on the system capacity of Platform 10 in the hurricane loading direction.

Figure 9. Effect of Pile A on the system capacity of Platform 10 in the hurricane loading direction. Figure 12. Possible range of foundation system capacity of Platform 10 due to increasing or decreasing the axial and lateral capacities of one pile by 30% in the hurricane loading direction.

Figure 10. Effect of Pile B on the system capacity of Platform 10 in the hurricane loading direction.

three 1219-mm (48-inch) diameter piles with a 508mm (20-inch) diameter well conductor (Fig. 8). The length of Pile A is 80.8 m (265 feet) and the length of Piles B and C is 67.1 m (220 feet). Figures 9–11 present the capacity interaction curves of this foundation system due to increasing and decreasing the axial and lateral capacities of Piles A, B © 2011 by Taylor & Francis Group, LLC

901

and C, respectively. The base case foundation system capacity interaction curve is also presented in these figures to provide a reference for comparison. Pile C is the most influential pile in this foundation system as shown by the large variation in foundation system capacity (Fig. 11) and Pile A is the least influential pile (Fig. 9). These results are because Pile C is further away from the axis of rotation than Piles A and B and the direction of batter for Pile C is more effective in resisting overturning moment in the hurricane loading direction. Figure 12 presents the bounds of all capacity interaction curves shown in Figures 9–11. It shows the possible range of foundation system capacity of this platform due to ±30% variation in the capacities of each pile. The overturning capacity of the foundation system is much more sensitive to the axial capacity of a single pile than the shear capacity of the system is to the lateral capacity of a single pile as shown by the wider range of variation (Fig. 12). Similar sensitivity analyses were performed for the foundation of Platform 30 (Fig. 13). Figure 14 presents the bounds of all capacity interaction curves from the

8

CONCLUSIONS

Structural factors are significant in estimating the system capacity of jacket pile foundations. Well conductors, jacket leg stubs and steel yield stress are important to the shear capacity of the foundation system while foundation system redundancy is critical to the overturning capacity of the foundation system. The best practice in offshore platform design and assessment is for structural and geotechnical engineers to work closely together. Piles supporting the platform and the platform structure itself need to be considered as a complete system in order to achieve an effective and consistent level of performance for the foundation. ACKNOWLEDGMENTS We acknowledge Minerals Management Service and American Petroleum Institute for providing financial support through Offshore Technology Research Center for the research projects upon which this paper is based. We also acknowledge Engineering Dynamics, Inc. for providing use of their Structural Analysis Computer Software (SACS™). Britain Materek, Justin Carpenter, Sean Verret, Aditya Hariharan and Matthew Garcia contributed significantly to the research projects. The views and opinions presented herein are ours alone and do not necessarily reflect those of our sponsors.

Figure 13. Foundation plan of Platform 30.

REFERENCES

Figure 14. Possible range of foundation system capacity of Platform 30 due to increasing or decreasing the axial and lateral capacities of one pile by 30% in the diagonal direction.

sensitivity analyses. PilesA2 and C1 are the most influential piles of this foundation system in the diagonal loading direction because they are further away from the axis of rotation and their batters are more effective. The overturning capacity of the foundation system is still much more sensitive to the axial capacity of a single pile than the shear capacity of the system is to the lateral capacity of a single pile (Fig. 14). However, comparing Figures 12 and 14, the variation in the capacity of the 3-pile foundation system is greater than the variation of the 6-pile foundation system because the 3-pile system is less redundant than the 6-pile system. The 3-pile foundation system was failed in overturning in Hurricane Ike. The 6-pile foundation system survived the loading in Hurricane Katrina even though it was nearly loaded to its shear capacity. The effect of redundancy on foundation system capacity likely contributed to the actual performance of these platforms when they were loaded to their capacities. © 2011 by Taylor & Francis Group, LLC

Energo. 2009. Evaluation ofYield Stress for Steel Members in Gulf of Mexico Fixed Platforms. Final Report to American Petroleum Institute. Gilbert, R.B., Chen, J.Y., Materek, B., Puskar, F., Verret, S., Carpenter, J., Young, A. and Murff, J.D. 2010. Comparison of observed and predicted performance for jacket pile foundations in hurricanes. Offshore Technology Conference. OTC 20861. Jeanjean, P. 2009. Re-assessment of p-y curves for soft clays from centrifuge testing and finite element modeling. Offshore Technology Conference. OTC 20158. Murff, J.D. 1987. Plastic collapse of long piles under inclined loading. International Journal for Numerical and Analytical Methods in Geomechanics 11: 185–192. Murff, J.D. 1999. The mechanics of pile foundation collapse. In Jose Roesset (ed.), Analysis, Design, Construction and Testing of Deep Foundations; Proceedings of the OTRC ’99 Conference, Austin, Texas, 29–30 April 1999. Murff, J.D. & Wesselink, B.D. 1986. Collapse analysis of pile foundations. 3rd International Conference on Numerical Methods in Offshore Piling, Nantes, France: 445–459. OTRC. 2009. Analysis of Potential Conservatism in Foundation Design for Offshore Platform Assessment. Final Project Report Prepared for the Minerals Management Service. MMS Project Number 612. Tang, W.H. & Gilbert, R.B. 1992. Offshore Pile System Reliability. Final Report to American Petroleum Institute.

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The new API Recommended Practice for Geotechnical Engineering: RP 2GEO P. Jeanjean BP America Inc., Houston, TX, USA

P.G. Watson Advanced Geomechanics, Perth, Australia

H.J. Kolk Fugro Engineers, B.V., The Netherlands

S. Lacasse Norwegian Geotechnical Institute, Oslo, Norway

ABSTRACT: The American Petroleum Institute (API), through the sub-committee on offshore structures SubCommittee 2 (SC2) and its Resource Group 7 (RG7) on geotechnical engineering, will soon publish a new recommended practice (RP) on geotechnical engineering, entitled RP 2GEO. This new RP covers both shallow foundations and pile foundations and is aligned to the greatest extent possible with the current ISO 19901-4 (for shallow foundations) and ISO 19902 (for pile foundations for fixed steel structures). This paper describes the key features of the new code and how its compares both with existing ISO guidance and with previous guidance in API RP 2A WSD 21st Ed. The philosophy behind the technical changes in RP 2GEO will also be explained. The authors of the papers are all members of RG7 and its ISO equivalent committee, Working Group 10 (TC7/SC7/WG10). 1

INTRODUCTION

Recommendations on geotechnical practice in API codes pertaining to offshore structures are prepared and written by the geotechnical Resource Group 7 (RG7) (formally known as RG5). In the International Standard Organization (ISO), the group equivalent to RG7 is Work Group 10 (WG10) on Foundations, formally referred to as TC67/SC7/WG10. Recognizing the need for seamless integration of geotechnical practice around the world, RG7 and WG10 have functioned as a joint committee preparing and jointly approving API and ISO foundation guidance. RP 2GEO is aligned with ISO 19901-4 (2003) and ISO 19902 (2007). At the time of writing this paper, the recommended practice “RP 2GEO (1st Edition)” is pending approval for issue by API. Its title will be “ANSI/API Recommended Practice 2GEO Geotechnical and Foundation Design Considerations”. This paper describes the key features of RP 2GEO and how its compares with existing guidance in ISO 19901-4 (2003) for shallow foundations and ISO 19902 (2007) for pile foundations, and with previous guidance in API RP 2A 21st Edition. The philosophy behind the technical changes in this new RP will also be explained. The “lumped” safety factor approach in use in API RP 2A WSD 21st Ed was retained in RP 2GEO. Key © 2011 by Taylor & Francis Group, LLC

changes were implemented in the shallow foundation section including adopting the Brinch Hansen method instead of the Vesic approach currently used in API RP 2A. In the deep foundation section, recommendations on the use of cone penetration test based methods for pile design in sands are now provided, while the soil structure interaction curves, so called t-z curves, for sands have also been modified. Because ISO 19901-4 uses a load and material factor approach, and RP 2GEO a lumped safety factor, the two codes do not compare exactly. Where possible, care was however taken so that the two codes provide foundations of comparable sizes for given load and soil profile. Because of space limitations, this paper presents an abbreviated discussion and a more detailed description of RP 2GEO can be found in Jeanjean et al. (2010). 2

DESIGN OF SHALLOW FOUNDATIONS

2.1 API 21st edition (2000) The framework proposed by Vesic (1975) was included in the RP 2A 10th Edition (1979) and was unchanged through the 21st edition (2000). In Vesic (1975), the bearing capacity factors Nc and Nq were the same as those recommended by Meyerhof (1951). The Nγ factor was approximated with

903

an empirical relationship (Nγ = 2(Nq + 1) tanφ ) that closely matched the results of Caquot and Kerisel (1953).

friction angle, the ultimate load envelopes are significantly smaller when using RP 2GEO compared to ISO 19901-4. No effort was made to harmonize these envelopes. • However, due to differences in the way safety factors in RP 2GEO and load/material factors in ISO 19901-4 are applied, envelopes of allowable loads are similar for both codes.

2.2 The ISO 19901-4 (2003) approach The Brinch Hansen (1970) method included bearing capacity factors Nc and Nq that were the same as those recommended by Meyerhof (1951) and Vesic (1975). The Nγ factor was approximated by an empirical relationship Nγ = 1.5 (Nq − 1) tanφ . Values of φ used for the derivation of Nc and Nq are based on plane strain conditions, φPS (rather than triaxial conditions, φTX ), and it was suggested that φPS may be (conservatively) . taken as 10% higher than φTX The Brinch Hansen (1970) approach and the load and material factor framework was adopted for the ISO 19901-4 (2003) standard on shallow foundations, with modified depth and shape factors, as summarized in Table 1.

Overall, the calculations performed with RP 2GEO and ISO 19901-4 for foundations on soils behaving as drained will result in comparable foundation size to resist a given set of allowance (unfactored) load. Going forward, the greatest challenge to align the RP 2GEO and ISO 19901-4 texts lies in how to reconcile the lumped safety factor, working stress design (WSD) API approach with the partial load factor and material factor design ISO approach. 3

CALCULATION OF PILE CAPACITY

2.3 The RP 2GEO approach

3.1

The RG7/WG10 group decided that the differences between ISO 19901-4 (2003) and API RP 2A 21st Edition (2000), while not entirely reconcilable, would be minimized in RP 2GEO. The key differences between the adopted recommendations of RP 2GEO and those of RP 2A 21st Edition, and ISO 19901-4 (2003) are summarized in Table 1.

Design practices for the axial capacity of piles in sand have been and still are mostly aligned around the world, except for the choice of the value for K, the coefficient of lateral earth pressure. Although a K value of 0.8 for compression and tension is used in the Gulf of Mexico, K values of 0.7 in compression and 0.5 in tension are still often applied for North Sea projects and is recommended in DNV (1992) and Lloyd’s Register (1989, 2008). These values can be upgraded for North Sea projects for dense sands, if high quality CPT data is available (Jeanjean et al., 2010).

2.4 Comparison of RP 2GEO and ISO 19901-4 for jacket mudmats In the example of Fig. 1, failure envelopes are developed in V-H space. The curves depict the combinations of allowable (i.e. unfactored) vertical loads, V, and horizontal loads, H, that satisfy the requirements of the RP 2A 21st Ed., RP 2GEO, and ISO practices. Failure envelopes are developed for a surface (no skirts) 10 m square mudmat founded on soil with constant su = 10 kPa. The material factor was taken as 1.25 in the ISO 19901-4 (2003) calculations. Note that the two allowable envelopes generated using ISO 19901-4 (2003) represent different (assumed) combinations of dead and environmental load: a. All vertical and lateral loads are assumed to be environmental and the load factor is 1.35. b. 50% of the vertical load is assumed to be dead load, with the remainder (and all lateral load) assumed to be environmental. The load factor is therefore taken as 0.5(1.1 + 1.35) = 1.225 for the vertical load and as 1.35 for the horizontal load. It can be seen that comparable envelopes of ultimate capacity are generated using ISO 19901-4 (2003) and RP 2GEO. However, differences in factoring between WSD and load/material factor design approaches leads to differences in allowable load. In comparing drained envelopes from ISO 19901-4 (2003) and RP 2GEO (Fig. 2), it is typically seen that: • Because RP 2GEO uses a smaller triaxial friction

angle and ISO 19901-4 uses a larger plane strain © 2011 by Taylor & Francis Group, LLC

3.2

Gulf of Mexico and North Sea practices

Key studies on pile capacity in sands 1993–2008

Key studies between 1993 and 2008 that influenced the recommendations in RP 2GEO include the following (a summary of the findings of each study can be found in Jeanjean et al., 2010): Offshore Technology Research Center (OTRC) – Fugro McClelland (1990–1994) study; Hossain and Briaud (1992) and (1993); Fugro McClelland (1994); The EURIPIDES pile load tests (1995); The MTD Method: Jardine and Chow (1996) and the ICP method: Jardine et al. (2005); The Ras Tanajib II pile load tests (1996–1999); APIsponsored Fugro Engineers Pile Study (2003–2004); 2004 seminar on “Piles in Sand” – Houston TX; Studies at the Norwegian Geotechnical Institute (NGI): 1995 to 2005; The University of Western Australia (UWA) study - Lehane et al. (2005). Detailed references to these studies can be found in Jeanjean et al (2010). 3.3 The RP 2GEO approach for pile design The modifications to the RP 2A 21st Edition (2000), as implemented in RP 2GEO, are now summarized. These modifications were initially published as part of the RP 2A Addendum and Errata #3 (2007) in October 2007. Unfortunately, several key typographical

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Table 1.

Key differences between RP 2A 21st Ed., RP 2GEO, and ISO 19901-4 for shallow foundations.

Topic Framework Factors

RP 2A 21st Ed. (2000)

RP 2GEO

ISO 19901-4

Lumped safety factors, independent of type of loads (i.e. gravity vs environmental)

Lumped safety factors, independent of types of load (i.e. gravity vs environmental)

Load and material factors, with load factors varying for gravity loads and environmental loads.

Not included

Included with unfactored soil effective unit weight Horizontal load is transferred to foundation base to calculate inclination factors and safety factor against sliding. Vertical load is transferred to foundation base.

Undrained bearing capacity Overburden term in Included with soil bearing capacity total unit weight equation Horizontal load No recommendations transfer to foundation base Vertical load transfer to foundation base

No recommendations

Bearing capacity factors Inclination factors

Nc = Meyerhof (1951) = Prandtl (1920) = 5.14 Vesic (1975)

Shape factors

Vesic (1975)

Depth factors Base inclination and seafloor slope factors Linearly increasing shear strength profiles

Vesic (1975) Vesic (1975)

Horizontal load is transferred to foundation base to calculate inclination factors and safety factor against sliding. Vertical load is transferred to foundation base only to calculate load eccentricity and effective area (if applicable), not to calculate bearing safety factors. Nc = Meyerhof (1951) = Prandtl (1920) = 5.14 Brinch Hansen (1970) with unfactored horizontal load (where sliding outside the effective area used to reduce inclination factor, this is to be factored) DNV (1974) and Salençon & Matar (1983) with the shape factor for circular foundations and pure vertical loads changed from 0.2 to 0.18 for uniform soil conditions DNV (1992) = DNV (1977) Brinch Hansen (1970)

No recommendations

Davis and Booker (1973)

Davis and Booker (1973)

Included with soil total unit weight

Not included

No recommendations

Horizontal load is transferred to foundation base to calculate inclination factors. Not included Nq = Meyerhof (1951) = Reissner (1924) Nγ = Brinch Hansen (1970)

Included with unfactored *soil effective unit weight Horizontal load is transferred to foundation base to calculate inclination factors. Vertical load is transferred to foundation base. Nq = Meyerhof (1951) = Reissner (1924) Nγ = Brinch Hansen (1970)

Brinch Hansen (1970) with unfactored loads Brinch Hansen (1970) DNV (1977)

Brinch Hansen (1970) with factored loads Brinch Hansen (1970) DNV (1977)

Brinch Hansen (1970)

No recommendations

Not included in bearing capacity equation Triaxial friction angle

Included in bearing capacity equation Plane strain friction angle (10% greater than triaxial)

Drained bearing capacity Overburden term in bearing capacity equation Horizontal load transfer to foundation base Vertical load transfer to foundation base Bearing capacity factors

Inclination factors Shape factors Depth factors Base inclination and seafloor slope factors Effective cohesion c’ Choice of drained friction angle for sand

No recommendations Nq = Meyerhof (1951) = Reissner (1924) Nγ = Vesic (1975) (approximated from Caquot and Kerisel (1953)) Vesic (1975) Vesic (1975) Vesic (1975) = Brinch Hansen (1970) Vesic (1975) Included in bearing capacity equation No recommendations


905

Nc = Meyerhof (1951) = 4 Prandtl (1920) = 5.1 Brinch Hansen (1970) with factored loads

DNV (1974) and Salençon & Matar (1983)

DNV (1992) = DNV (1977) No recommendations

Figure 3. Ratio of Calculated-to-Measured pile compression capacity, according to the API main text method, as a function of sand relative density.

Figure 1. V-H envelope for 10 m square mudmats founded on soil with su = 10 kPa (0.21 ksf).

Figure 4. Ratio of Calculated-to-Measured pile compression capacity, according to the API main text method, as a function of pile Length-to-Diameter ratio.

Figure 2. V-H envelope for 10 m square mudmats founded on soil with triaxial φ = 32◦ .

mistakes occurred in the Commentary and this document is not to be used in design. The same modifications are also included in ISO 19902 (2007), but without typographical errors. The bias of the API (1993) method, as documented by some of the above studies, is illustrated in Figures 3 to 6, as a function of the relative density of sand and the pile-length diameter ratio, both for compression and tension loading. A key question is how the RP 2A calculated pile capacities in sand should be viewed. Are they (1) an accurate and unbiased estimate, (2) a conservative © 2011 by Taylor & Francis Group, LLC

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Figure 5. Ratio of Calculated-to-Measured pile tension capacity, according to the API main text method, as a function of sand relative density.

estimate or (3) an estimate for prudent design based on the (best) available experience? Murff (2005) stated that the API method is a design method and not a prediction method. The track record of the method has been good. However, there is probably only a fraction of the API-designed foundations that have been subjected to their design loads. It is the opinion of the four authors that the pile capacities calculated with the API approach provide

Figure 6. Ratio of Calculated-to-Measured pile tension capacity, according to the API main text method, as a function of pile Length-to-Diameter ratio. Figure 7. Measured cone profile in West Delta area.

good, appropriately conservative estimates for the design of piles for offshore installation in medium to dense sands. Its application is simple, therefore unambiguous, and should minimize user errors. 3.4 Modifications to the main text method The main text method was modified as follows: • The symbols K and δ were removed and replaced

by the product “β = K·tan (δ)” in the pile design method to reduce confusion and lower the risk of errors (see Jeanjean et al., 2010, for more details). • To remove unconservatism, the API main text method is no longer recommended for very loose and loose sands, loose sand-silts, and medium dense silts, and dense silts. The CPT-based methods in the Annex are recommended instead. The designer should be aware that areas of potential unconservatism still remain in the RP 2GEO main text method. In comparison to the CPT-based methods, the method may not be conservative in tension for thick deposits of medium dense sands (Fig. 5). The description of where the main text method is conservative (short piles less than 45 m (150 ft) in compression in dense to very send sands) and where the method may be unconservative (all other cases) has been changed from previous RP 2A text. 3.5 Addition of CPT-based methods in the Annex The RG7/WG10 committee decided after long deliberations to include four CPT-based methods in the Commentary/Annex section of RP2GEO. No agreement was reached on which method was the preferable one to include in the main text. The four methods in the annex are: the Simplified-ICP, Fugro-05, NGI-05, and Offshore UWA-05 methods. All methods arediscussed in Lehane et al. (2005). 3.6 Example of pile capacity calculations. To illustrate the use of the CPT methods, the capacity of a driven pile is calculated for a site in the Gulf of Mexico West Delta area (Fig. 7). Pile capacities © 2011 by Taylor & Francis Group, LLC

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Figure 8. Comparison of friction capacity in tension, as a function of pile penetration depth.

were calculated by using the API main text and the CPT methods with the cone resistance profile shown in red for sand layers. For clays layers, the API main text method was used with the shear strength profile shown in green. The capacity curves, calculated for a 1.07 m (42in) pile with 38.1 mm (1.5 in.) wall in Fig. 8 and 9 reflect the biases of Fig. 3 to 6 and illustrate the dilemma of RG7/WG10 noted in section 4.5. The example herein is only one example of calculations with the four CPT and API main text methods. They are for a slender pile in a highly layered profile. Other soil profiles can give other relative pile capacity profiles for each of the different analysis methods. For clarity of comparison, the calculated end bearing is not averaged over a few pile diameters to account for soil layering. This averaging of the end bearing when the pile tip approaches a sand-clay layer interface should be included in an actual design.

Brinch-Hansen, J. 1970: “A Revised and Extended Formula for Bearing Capacity”, Geoteknisk Institut, Bulletin No. 28, p. 5–11, Copenhagen. Caquot A., and Kerisel, J. 1953: “ Sur le Terme de Surface dans le Calcul des Fondations en Milieu Pulvérulent” Proc. of the 3rd Conference on Soil Mechanics and Foundation Engineering, Vol. 1, Zürich. Davis, E.H. and Booker, J.R. 1973: “The Effect of Increasing Strength with Depth on the Bearing Capacity of Clays”, Géotechnique, 23 (4), 551–563. DNV 1974: “Rules for the Design, Construction and Inspection of Fixed Offshore Structures”, DetNorskeVeritas, Oslo. DNV 1977: “Rules for the Design, Construction and Inspection of Offshore Structures – Appendix F – Foundations”, Det Norske Veritas, Oslo. DNV 1992: “Foundations”, Classification notes No. 30.4, Det Norske Veritas, February Fugro Engineers 2004: “Axial Pile Capacity Design Method for Offshore Driven Piles In Sand”, Fugro Report No.: P1003 by A. Baaijens and H.J. Kolk to the American Petroleum Institute, August 5. Fugro McClelland 1994: “Axial Capacity of Piles in Sand”, Report No. 0201-1485 by T. Hamilton and J.S. Templeton Figure 9. Comparison of total capacity in compression, as to the American Petroleum Institute, August. a function of pile penetration depth. Hossain, M. K., and Briaud, J.L. 1993: “Improved Soil Characterization for Pipe Piles in Sand in API RP2A”,Proceedings, Offshore Technology Conference, 4 CONCLUSION Houston,TX,Paper 7193. ISO 19901-4 2003: Petroleum and natural gas industries — The API and ISO committees working to develop Specific requirements for offshore structures — Part 4: guidelines for offshore foundations recently introGeotechnical and foundation design considerations, 1st duced RP 2GEO. Except for its working stress design Edition ISO 19902 2007: Petroleum and natural gas industries — (WSD) framework, RP 2GEO incorporates pile design Fixed Steel Offshore Structures, 1st Edition recommendations similar to those of ISO 19902 and Jardine, R., Chow, F. 1996: “New Design Methods for Offrecommendations on shallow foundations that are shore Piles”, MTD Publication 96/103 by The Marine largely aligned with those of ISO 19901-4, thereby Technology Directorate Ltd, constituting a great step in harmonizing offshore Jeanjean, P., Watson, P.G., Kolk, H., and Lacasse, S. 2010: geotechnical practices around the world. “RP 2GEO: The new API Recommended Practice for Geotechnical Engineering.”, Proceedings, Offshore Technology Conference, Houston, TX, Paper 20631 Lehane, B., Schneider, J.A., and Xu, X. 2005: “A Review ACKNOWLEDGEMENT of Design Methods of Offshore Driven Piles in Siliceous Sands”, University of Western Australia, Report GEO The members of theAPI RG7 and ISOTC67/SC7/WG10 05358, September committees are acknowledged for their many passionLloyd’s Register 1989: “Rules and Regulations for the Classiate discussions on geotechnical matters over the years fication of Fixed Offshore Installations’, Part 3, Chapter 2. and their contributions to the writing and editing of RP Lloyd’s Register 2008: “Rules and Regulations for the Clas2GEO. sification of a floating offshore installation at a fixed location”, Part 3, Chapter 12., April The authors are grateful to their respective compaMeyerhof, G.G. 1951: “The Ultimate Bearing Capacity of nies for the permission to publish. Foundations”, Géotechnique, Vol. 2, 301–332 Murff, J.D. 2005: Personal communication to P. Jeanjean and the RG7/WG10 resource groups, June. REFERENCES Salençon, J., and Matar, M. 1983: “Bearing Capacity of Circular Shallow Foundations in Foundation Engineering, ANSI/API Recommended Practice 2GEO “Geotechnical and Soil Properties”, Foundation Design and Construction, Foundation Design Considerations”, 1st Edition, (under Vol. 1, pp.159–168, Presses de l’Ecole Nationale des Ponts development). et Chausées. API RP 2A 1st Edition (1969) to 21st Edition (2000): “RecVesic, A.S. 1975: “Chapter3 – Bearing Capacity of Shalommended Practice for Planning, Designing and Conlow Foundations”, Foundation Engineering Handbook, structing Fixed Offshore Platforms – Working Stresses Edited by Winterkorn, H.F. and Fang, .H.Y., Von Nostrand Design”. Reinhold Company, NewYork, pp.121–147. API RP 2A-WSD Errata 2007: “Errata and Supplement 3 Wisch, D., and Mangiavacchi, A. 2010: “Strategy and Structo API Recommended Practice 2A-WSD, Recommended ture of the API 2 Series Standards, 2010 and Beyond”, Practice for Planning, Designing and Constructing Fixed Proceedings, Offshore Technology Conference, Houston, Offshore Platforms – Working Stresses Design”, 21st TX, Paper 30831. Edition, December 2000.”, October 2007. © 2011 by Taylor & Francis Group, LLC

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Comparison of ISO 19901-2 and API RP 2A seismic design criteria for a site in the Caspian Sea, Turkmenistan Z.A. Lubkowski Arup, London, UK

J.E. Alarcon AIR, London, UK

Z.A. Razak Petronas, Kuala Lumpur, Malaysia

ABSTRACT: Due to unavailability of zoning data, as per clause 2.3.3c of the API RP-2A WSD 22nd Edition, a probabilistic seismic hazard assessment was carried out for the proposed offshore facilities by the Operator of Block 1, Turkmenistan, Caspian Sea. The purpose of the study was to derive strength level event (SLE) and ductility level event (DLE) seismic design criteria for the facilities. These seismic design criteria are compared to those derived using the potentially more rigorous approach presented in ISO 19901-2, where the concepts of “consequence of failure” and “exposure level” are directly accounted for. ISO 19901-2 allows the designer to consider the risks and uncertainties in making decisions to come up with the abnormal level earthquake (ALE) and the extreme level earthquake (ELE) ground motion values for design. This paper presents a comparison of these two seismic design criteria and also examines the influence of different seismic hazard models on the resulting design criteria.

1

BACKGROUND

The site, known as Block 1, is located in the Turkmenistan sector of the Caspian Sea, approximately 70 km offshore in water depth ranging from 30 m to 90 m. The design code selected for the proposed gravity based platform is API RP-2A (2000). However, since API does not provide seismic criteria outside of the USA the operator of Block 1 embarked on a study to determine values for the design of their facilities. The principal findings from that study were peak ground acceleration (PGA) values of 0.2 g and 0.4 g for return periods of 200 years SLE and 2000 years DLE criteria respectively. Subsequently a more detailed probabilistic seismic hazard assessment (PSHA) was carried out (Fugro, 2006) based on the latest available data at that time. This more detailed study found that the PGA values to be considerably higher, ranging from 0.34 g to 0.73 g for the 200 years SLE and 2000 years DLE respectively. This magnitude of increase in the seismic criteria had significant implications for the seismic design of the platform. The operator therefore decided to implement performance based seismic analysis methods together with revised seismic criteria based on the newly released ISO 19901-2 (2004) code, where the concept of “consequence of failure” and “exposure level” together with platform type and function are © 2011 by Taylor & Francis Group, LLC

directly accounted for in defining appropriate seismic design criteria. This paper presents a comparison of these two seismic design criteria for this project and also examines the influence of different seismic hazard models on the resulting design criteria.

2

OFFSHORE DESIGN GUIDELINES

API RP-2A (2000) defines two levels of earthquake ground motion for design. Firstly, those which have a reasonable likelihood of not being exceeded at the site during the platform life span, and secondly those ground motions from a rare intense earthquake (RIE). The first level of ground motions is named the strength level earthquake (SLE) while the latter is named the ductility level earthquake (DLE). API requires, from the structural and geotechnical design, that the offshore platform should have no significant damage for the SLE and that collapse is to be prevented during the DLE. API also presents exposure categories for selection of design levels; these exposure categories depend on a life-safety level (i.e. is the structure manned and non-evacuated, manned and evacuated or unmanned) and on a consequences of failure level (i.e., high, medium or low).

909

However, API does not specify the actual level of ground motion to be used and thus, the selection of return periods for the SLE and the DLE are left at the owner’s descretion. Furthermore, there is a potential inconsistency in the use of the API guidelines, which is related to the widespread selection of the DLE ground motions as twice of the SLE level (i.e., DLE = 2.0 × SLE). API mentions in Section 2.3.6.d that “No ductility analysis of conventional jacket-type structures with 8 or more legs is required if the structure is to be located in an area where the intensity ratio of [DLE] to [SLE] is 2 or less…and the following conditions are adhered to in configuring the structure and proportioning members:…” after which the “DLE = 2.0 × SLE” rule has developed. To the author’s best knowledge, API does not state this as a general requirement for defining the DLE ground motion. ISO 19901-2 (2004) also proposes two levels of earthquake ground motion for design. The extreme level earthquake (ELE) is that at which the structure should be able to sustain no major damage while the abnormal level earthquake (ALE) is that at which the structure should not suffer complete loss of integrity. Three levels of exposure L1, L2 and L3 (where L1 represents facilities that require higher standards of design) are considered in the definition of ground motion design levels (see Section 6.4 in ISO 19901-2 for further details). The potential advantage of ISO over API lies in the fact that the recommended return periods for design are clearly determined based on the seismicity characteristics of the region surrounding each particular site. The ALE and ELE return periods are defined by following a step-by-step procedure that makes use of the part of the hazard curve corresponding to a certain response period. 3

SEISMIC HAZARD MODEL

The hazard curves produced from a PSHA are the primary inputs for defining the design levels in both the API RP-2A and ISO 19901-2 approaches. At the site, two independent PSHAs were carried out, the first one was by Fugro in 2006 and the second by Arup in 2007. Both studies implemented state-of-practice methods and thus the results can be considered to be mutually complementary. PSHAs combine seismic source zoning, earthquake recurrence and ground-motion predictive (attenuation) equations to produce hazard curves that represent the level of ground motion, and an associated annual frequency of being exceeded. These key elements are explained briefly in the subsequent sections, and include the description on how these elements were implemented in the Fugro and Arup models. 3.1 Tectonic setting The Caspian Sea region is located in the northern most collision boundary between the Arabian and Eurasian © 2011 by Taylor & Francis Group, LLC

plates. The predominant northward movement of the Arabian plate creates a compressive environment in the Caucasus, which induces predominantly westward movement of the Anatolian (Turkish) block and northeastward movement of the Iranian block. The detailed tectonic setting in the project area is complex and describing its details lies beyond the scope of this paper. However, it is important to highlight that the denser oceanic crust is being subducted beneath the lighter central Caspian Sea continental crust (Knapp et al., 2004; Jackson et al., 2002). The deformation along this complex plate margin is believed to be partitioned onto both strike-slip faults and fold and thrust structures including the Apsheron sill or Apsheron anticline. This is a linear uplifted structure of faults and folds, the axis which extends from Baku in Azerbaijan to Cheleken in Turkmenistan.

3.2

Seismic source models

The geographical variation of earthquake activity is represented in a seismic source model. They are based on the distribution of observed seismic activity together with geological and tectonic understanding of the region. The zones represent areas where the seismicity is assumed to be homogenous; i.e. an equal chance that a given earthquake will occur at any point. Fugro and Arup used a similar database of historic and instrumental recordings in their respective analyses. Replicated events were removed from the compiled database considering: the quality of information (some were given a higher confidence level due to better documentation), the completeness of information and the magnitude scale used (a uniform magnitude scale is required for the composite catalogue, so preference was given to entries that minimise adjustments). Foreshocks and aftershocks were removed from the catalogue following the Gardner and Knopoff (1974) methodology. Figure 1 shows the idealised seismic source model developed by Arup (2007). A similar model was developed by Fugro (2006). The seismic source models included shallow crustal sources (10 in the Fugro model, 20 in the Arup case) and subduction zone sources of the South Caspian Sea. The model also includes the Apsheron sill or Apsheron anticline, modelled as a 72 km long planar source located about 10 km from the site. Differences in the models arise largely from the range of expert opinions embodied in the PSHA source zonation process. However, a detailed discussion is beyond the scope of this paper.

3.3

Recurrence relationships

Recurrence relationships in the form of the “Gutenberg-Richter” equation (Log10 N = a − bM) were assigned to each of the Arup and Fugro seismic source zones. Table 1 presents the values for the Arup model, as a detailed comparison and discussion is beyond the scope of this paper.

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Table 1. Mean magnitude-recurrence parameters and maximum magnitude per zone for the Arup seismic source model.

Figure 1. Arup seismic source models, showing shallow crustal (upper) and deep source zones (lower).

Source zone ID

b

Activity rate, a (N > 4.0/ year)

Maximum magnitude Mw

AZB1 AZB2 AZB3 AZB4S AZB5S AZB6S AZB7S AZB8 AZB9 AZB10 AZB11 AZB12S AZB13S AZB14 AZB15 AZB16 AZB17 AZB18 AZB19 AZB20 AZB4D AZB5D AZB6D AZB7D AZB12D AZB13D

0.74 1.40 1.00 0.83 1.84 1.30 1.00 0.93 1.40 1.20 0.75 0.83 0.83 0.75 0.75 1.20 1.00 0.83 1.00 1.30 0.75 0.93 0.93 1.20 0.93 0.83

0.095 4.837 0.083 1.196 7.652 3.755 0.486 0.458 1.326 0.231 0.349 3.783 0.543 0.45 0.528 0.19 1.852 1.445 1.717 2.695 0.691 4.183 1.598 0.86 2.246 0.816

7.5 7.0 8.0 8.0 7.2 7.2 8.0 8.3 7.0 6.5 8.0 8.0 8.0 8.0 8.0 6.5 8.0 8.0 8.0 7.2 8.0 8.3 8.3 6.5 8.3 8.0

Table 2. PSHAs.

Ground-motion predictive equations used in the

3.4 Ground motion predictive equations

Tectonic regime Fugro

A set of ground-motion predictive (attenuation) equations (GMPEs) is required to define what ground motion could be expected at location A due to an earthquake of known magnitude occurring at location B. These equations are generally derived from observations from past earthquakes, and play a significant role in a PSHA. The GMPEs used in the Fugro and Arup studies are summarised in Table 2 The large number of shallow crustal zones with respect to the subduction ones in the Arup model is the fundamental reason for the selection of two equations applicable to this kind of region. Epistemic uncertainties were considered in both studies using a logic tree approach.The logic trees were applied on parameters such as the maximum magnitude, recurrence parameters and GMPEs. The different branches of the logic trees and their relative weightings (i.e., the values given that represent the degree of confidence in each branch) are not presented herein.

Shallow Crustal Abrahamson & Silva (1997)

4

SEISMIC HAZARD RESULTS

The basic methodology adopted in the PSHA is based on that originally proposed by Cornell (1968) from © 2011 by Taylor & Francis Group, LLC

Subduction

Arup

Ambraseys et al. (1996) Sadigh et al. (1997) Youngs et al. (1997) Youngs et al. (1997) Crouse (1991)

the work of Cornell (1964) and Esteva (1967). The analyses presented here calculate the peak ground acceleration (PGA) and the uniform hazard response spectra (UHRS) for given return periods. For the purpose of comparing ISO and API methodologies using two different, but very robust seismic models, the Fugro model was re-run using the Arup program SISMIC. A comparison of the UHRS from the Fugro and Arup models is presented in Figure 2 for return periods of 200 and 2000 years. The results presented in Figure 2 show a good agreement for the response spectra with a return period of 200 years, whilst significant differences for the response spectra with a return period of 2000 years. The observed differences in the Fugro and Arup models do not come as a surprise since two different seismic hazard models are not expected to

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Figure 2. Comparison of UHRS for 200 and 2000 years return period from the Fugro and Arup models. Table 3. Comparison of Arup and Fugro PGA values for different return periods. The ratios between Arup and Fugro values are given in the last row. Return period (years)

SLE in (g)

DLE in (g)

100

150

200

1000

1500

2000

Arup Fugro Ratio A/F

0.18 0.30 0.60

0.22 0.34 0.65

0.27 0.38 0.71

0.65 0.66 0.98

0.79 0.75 1.05

0.88 0.81 1.09

From the data presented in Table 3 it is observed that the SLE ground motion accelerations are highly dependent on the seismic model employed, with the DLE ground motions showing slightly more stable results. Since the SLE requires the structure to sustain a low level of damage, a difference of around 30% in the structural demand may impose significant construction costs and large variability in the safety levels evaluated. If the widely implemented practice of scaling DLE from SLE, as previously described, is applied to the results presented in Table 3 for the 200 year return period case, the DLE ground motions for Arup and Fugro models would be 0.54 g and 0.76 g respectively. These ground motions equate to return periods of about 630 years and 1600 years for Arup and Fugro models respectively, which are lower than the 2000 year specified for the facilities design. These differences not only point out the poor practice of scaling DLE from SLE, but also highlights the dependence of PSHA results on the seismic model used. 5.2

ISO defines the ELE and ALE return periods. The key input parameters are the slope of the seismic hazard curve (aR ) which best represents the principle vibration mode of the platform being assessed and the probability of failure (Pf ). However, since the vibration mode may not be available at the time when the selection of ALE and ELE levels is carried out, ISO recommends the use of the 1.0 second response-period hazard curve as a default. Even though no specific reason is given for the selection of 1.0 second period for analysis, this may be related to the fact that most offshore structures have a mode of vibration greater than 1.0 second, and 1.0 second is usually corresponds to to the constant velocity portion of the response spectrum. This seems reasonable since peak ground velocity (PGV) has been shown to be better correlated with damage than PGA (Bommer & Alarcon, 2006). Table 4 shows the resulting ELE and ALE return periods, and their corresponding ground motions for the Arup and Fugro models. From these results it is observed that the influence of the seismic hazard model on the PGA values is not as significant as in the case of the API procedure. This effect seems to be particularly true for the lower return periods (i.e., ELE vs SLE levels) in which the ISO results show a minimum ratio of 0.91 whilst the API results show a minimum ratio of 0.42. For the longer return periods (ALE vs DLE levels), there does not seem to be a significant difference in the ISO and API results.

produce identical results. However, the significant differences observed at the 2000 years return period may be unusual. These differences come from the source zoning, earthquake data processing and predictive equations selected in each model. Considering that each model has gone through a comprehensive review, the observed differences highlight the large uncertainties enveloped in any PSHA. 5 API AND ISO DESIGN GUIDELINES TheAPI RP-2A and ISO 19901-2 design guidelines are implemented using the seismic hazard results from the Arup and Fugro models. These results and sensitivity analyses carried out are presented herein.

5.1 API RP-2A As introduced above API defines the SLE and DLE ground motion levels for design. API does not define concrete return periods, but the values of 200 and 2000 years have become accepted by industry. Table 3 presents the PGA values calculated from the Arup and Fugro models for these two levels. Additionally, and for sensitivity analysis purposes, other return periods have also been included for the following return periods: 100 and 150 years for the SLE and 1000 and 1500 years for the DLE. These SLE and DLE values have been selected arbitrarily, but 50 years was selected for compatibility with the minimum value recommended in ISO 19901-2. © 2011 by Taylor & Francis Group, LLC

ISO 19901-02

Sensitivity assessment considering the response period hazard curve As previously introduced, ISO recommends the use of the seismic hazard curve at 1.0 seconds response period, if project specific data are not available. Since the offshore facility considered has a natural period of vibration of about 2.5 seconds, a sensitivity assessment

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5.2.1

Table 4. Comparison of Fugro and Arup return periods and PGA values (in g) obtained when applying ISO 19901-2. The ratio on PGA between the Arup and Fugro models is given in the last row. Exposure & L1 Design Levels ELE Arup Fugro Rat. A/F

475 0.45 g 350 0.46 g 0.98

L2

L3

ALE

ELE

ALE

ELE

ALE

3125 1.04 g 3620 0.94 g 1.11

270 0.32 g 157 0.35 g 0.91

1175 0.70 g 1260 0.71 g 0.99

125 0.20 g 60 0.22 g 0.91

450 0.44 g 475 0.51 g 0.86

Table 5. Fugro ELE and ALE return periods (in years) calculated from hazard curves at different response periods. Exposure & Design Levels

L1

L2

L3

ELE

ALE

ELE

ALE

ELE

ALE

1.0 s 2.0 s 3.0 s

350 470 560

3620 3050 3000

157 212 265

1260 1180 1150

60 80 101

475 470 465

Figure 3. Derivation of the slope aR of the seismic hazard curve for a given response period. Table 7. Fugro ELE and ALE return periods (in years) calculated from different aR values.

Table 6. Arup ELE and ALE return periods (in years) calculated from hazard curves at different response periods. Exposure & Design Levels

L1

L2

L3

ELE

ALE

ELE

ALE

ELE

ALE

1.0 s 2.0 s

475 490

3125 3220

270 280

1175 1125

125 135

450 450

on the influence of the response period hazard curve used in the ALE and ELE evaluation is carried out. For this comparison hazard curves for 2.0 and 3.0 second response periods are used to evaluate the variability of the results. Table 5 and Table 6 present the results of this analysis. The results from the Fugro model show a degree of dependence of the ALE and ELE levels on the response period employed in the calculation, whilst the results from the Arup model do not seem to be significantly affected by it. The results presented in Tables 4 and 5 require further exploration before reaching a more definitive conclusion, since basic factors such as the GMPEs employed in each model may have had a significant influence on the results. 5.2.2 Sensitivity assessment considering the slope aR of the hazard curve A sensitivity assessment on the methodology used to calculate the hazard slope aR was carried out.The slope aR is defined as the ratio of the spectral accelerations corresponding to two probability values, one at either side of Pf , where Pf is the “Target annual probability of failure” as given in ISO 19901-2. ISO recommends © 2011 by Taylor & Francis Group, LLC

Exposure & Design Levels

L1

L2

L3

ELE

ALE

ELE

ALE

ELE

ALE

50 – 50 40 – 60 30 – 70

350 350 340

3620 3620 3600

157 155 155

1260 1250 1250

60 60 60

475 475 475

that the two probability values used to calculate aR (P1 and P2 ) should be one order of magnitude apart, with P1 being preferably close to Pf , as shown in Figure 3. The ALE and ELE variability on the selected distribution of the P1 and P2 values was examined. Three distributions were considered: when P1 and P2 are at one order of magnitude apart but each is at equal distance from Pf , this case is named as the 50-50 option. When P1 is moved closer to Pf in a condition where 40% of the “one order of magnitude apart” distance corresponds to the Pf to P1 separation, and thus the Pf to P2 distance is 60% of the total separation, then this case is named as the 40–60 option. The same principle applies for the 30–70 option. The resulting ALE and ELE return periods for the three conditions are presented in Table 7 and Table 8 for the Fugro and Arup models respectively. Negligible variations are observed in the ELE and ALE results for the three P1 - P2 distributions used to define aR . These results may be explained by the fact that normally a seismic hazard curve is plotted on a Log-Log scale which shows a linear trend from medium to low probabilities of exceedances (i.e. from intermediate to large return periods). It can therefore be concluded that for this case study the ISO methodolgy represents a consistent procedure that does not depend on the segment of the hazard curve used to define aR .

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Table 8. Arup ELE and ALE return periods (in years) calculated from different aR values. Exposure & Design Levels

L1

L2

ELE

ALE

ELE

ALE

ELE

ALE

50 – 50 40 – 60 30 – 70

475 475 470

3125 3140 3180

270 273 273

1190 1190 1190

130 129 129

450 450 450

levels in the world. The conclusions reached herein may point to recommending the use of ISO 19901-2 over API RP-2A as a more robust design guideline. A review on the extensive practice of scaling the API DLE ground motions as twice the SLE level, shows that unconservative ground motion criteria may be obtained for design. This practice appears to be the result of a misinterpretation of API guidelines.

L3

REFERENCES Table 9. Design PGA from the combination of seismic models and design guidelines. Values in brackets present the ground motions return periods in years. API Guideline & Design Level Arup Fugro

6

ISO

SLE

DLE

ELE

ALE

0.27 g (200) 0.38 g (200)

0.88 g (2000) 0.81 g (2000)

0.32 g (270) 0.35 g (157)

0.70 g (1175) 0.71 g (1260)

COMPARISON OF DESIGN LEVELS

A brief comparison of design PGA levels between Arup and Fugro with API and ISO is presented in this section. The design levels from the possible combination of seismic models and design guidelines are presented in Table 9. From these results it is observed that ISO procedure produces smaller ALE PGA values than API DLE. For the API results the Arup model gives higher DLE PGA values. The Arup results using the ISO procedure are very similar.

7

CONCLUSIONS

The variability of design levels from the application of API RP-2A and ISO 19901-2 guidelines have been reviewed in the light of the offshore project currently under development by the Block 1 Operator in the Caspian Sea, Turkmenistan. Two independent albeit robust seismic models, developed by Fugro and Arup, were used in order to determine the sensitivity of the design guidelines with respect to the different seismic hazard models. From the analysis carried out, it may be concluded that for the particular case under study the API RP-2A approach presents a significant degree of dependence in the seismic model employed when compared to the ISO 19901-2 results. Therefore for this site, ISO 19901 represents a more robust procedure in terms of the definition of the ground motions for design. The extension of the findings presented in this study to a more generalised conclusion may be obtained by carrying out similar studies for regions with different seismicity © 2011 by Taylor & Francis Group, LLC

Arup (2007). Turkmenistan Block 1 Gas Development Project – Probabilistic Seismic Hazard Assessment. A report for Petronas Carigali (Turkmenistan) Sdn. BhD. Abrahamson, N.A. & W.J. Silva (1997). Empirical response spectra attenuation relations for shallow crustal earthquakes. Seismological Research Letters 68(1), 94–127. Ambraseys, N.N., K.A. Simpson & J.J. Bommer (1996). Prediction of horizontal response spectra in Europe. Earthquake Engng and Structural Dynamics 25, 371–400. API RP-2A-WSD (2000). Recommended practice of planning, designing and constructing fixed offshore platform – Working stress design. 21st edition. American Petroleum Institute. Bommer, J.J. & J.E. Alarcon (2006). The prediction and use of peak ground velocity. Journal of Earthquake Engineering 10(1), 1–31. Cornell, C.A. (1964). Stochastic processes in civil engineering. Ph.D. Thesis, Department of Civil Engineering, Stanford University. Cornell C.A. (1968). Engineering seismic risk analysis. Bulletin of the Seismological Society of America 58(5), 1583–1606. Crouse, C.B. (1991). Ground motion attenuation equations for earthquakes on the Cascadia subduction zone, Earthquake Spectra 7, 201–236. Esteva L. (1967). Criterios para la construcción de espectros para diseño sísmico. Proceedings of the XII Jornadas Sudamericanas de Ingeniería Estructural y III Simposio Panamericano de Estructuras, Caracas. Fugro West Inc (2006). Design Ground Motions for Platform and GBS Locations, Turkmenistan, Caspian Sea. A report for Petronas Carigali (Turkmenistan) Sdn Bhd. Gardner, J.K. and Knopoff, L. (1974). Is the sequence of Earthquakes in Southern California, with Aftershocks Removed, Poissonian?, Bulletin of the Seismological Society of America, 64(5), 1363–1367. ISO 19901-2 (2004). International Standard ISO 19901-2, Petroleum and natural gas industries – Specific requirements for offshore structures- Part 2: /seismic design procedures and criteria. Jackson, J., Priestley, K., Allen, M. and Berberian, M. (2002). Active Tectonics of the South Caspian Basin. Geophysical Journal International, 148, 214–245. Knapp, C.C., Knapp, J.H., and Connor, J.A. (2004). CrustalScale Structure of the South Caspian Basin Revealed by Deep Seismic Reflection Profiling. Marine and Petroleum Geology, 21, 1073–1081. Sadigh, K., C.Y. Chang, J.A. Egan, F. Makdisi & R.R. Youngs (1997). Attenuation relationships for shallow crustal earthquakes based on California strong motion data. Seismological Research Letters 68(1), 180–189. Youngs, R.R., S.J. Chiou, W.J. Silva & J.R. Humphrey (1997). Strong ground motion attenuation for subduction zone earthquakes. Seismological Research Letters 68(1), 58–73.

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Offshore geotechnics – safe and sustainable J. Peuchen & J. Haas Fugro, Leidschendam, The Netherlands

ABSTRACT: Offshore Geotechnics is a service industry supporting the installation, operation and decommissioning of offshore facilities. Client industries, particularly those that extract energy from offshore regions, regard offshore geotechnics as providing “added value”. What specifically comprises “added value” changes over time, as industry and societal values change. The growing awareness of the importance of safety and sustainability has affected offshore geotechnics in the past and will continue to offer challenges in the future. This paper describes trends in thought about safety and sustainability in offshore geotechnics within a pragmatic context of adding value to the industries it serves. Future directions for the industry are suggested.

1

INTRODUCTION

Offshore Geotechnics fits within a group of service industries supporting the installation, operation and decommissioning of offshore facilities. Client industries, particularly those that extract energy resources from offshore regions, regard offshore geotechnics as providing “added value”. What specifically comprises “added value” changes over time, as industry and societal knowledge bases and values change. Historically, the start of offshore geotechnics as an industry may be traced to the first offshore sample borings in the Gulf of Mexico. The acquired information and clever engineering calculations helped development of a significant offshore hydrocarbon industry. Traditionally, economy, and its derivative “technology”, were the key drivers. Particularly, offshore geotechnics achieved added value by adopting or adapting technologies and practices from diverse sources: ocean/earth research, oil and gas exploration and production organisations, resource mining operations, the military, and traditional onshore geotechnics. Lord Cullen’s report (1990) on the Piper Alpha disaster of 6 July 1988 changed perceptions about offshore operations. As a result, another driver was progressively and successfully adopted for offshore geotechnics: safety. At about the same time, the Brundtland team (1987) published on global sustainability, highlighting social responsibility.They defined sustainability as “economic development that meets the needs of the current generation without endangering the opportunities for meeting the needs of future generations”. The sustainability concept is broad and complex. Particularly, it is difficult to agree performance indicators that are consistent and transparent (Boyle & Depraz 2006). Some industry segments recognised this difficulty and developed guidelines on voluntary © 2011 by Taylor & Francis Group, LLC

915

corporate reporting on sustainability, for example IPIECA and API (2003). A more general approach is offered by the draft version of ISO 26000 (ISO 2009a). This standard provides guidance on ways to integrate and manage socially responsible behaviour into existing organisational strategies, systems, practices and processes. For service industries the concept of sustainability is slowly emerging and only now (2010) increasingly mandatory for individual service organisations (for example SenterNovem 2008). Yet, sustainability issues for an industry are more complicated than simple “stacking” of sustainability results achieved by individual organisations contributing to the industry. Voluntary collaboration and global and regional incentives are believed necessary for genuine advancement at industry level. 2 TERMINOLOGY It would seem that much still has to be developed to effectively implement sustainable practice in offshore geotechnics. For offshore geotechnics, the denotation given in Figure 1 offers some pragmatic guidance. It is clear from Figure 1 that sustainability encompasses safety and economy values. This should be no surprise as sustainability ultimately converges to some fundamental attitudes of individuals, organisations and societies toward man, things and nature. From this point, any usage of “sustainability” will include “safety”. A further split into “service” and “facility” issues is convenient. “Service” refers to activities in offshore geotechnics, including any on-site geotechnical operations. “Facility” refers to Client facilities and operations affected by information and advice produced by the offshore geotechnics industry i.e. the service product. The sustainability of offshore geotechnics must consider the interaction between service and facility

Figure 3. Illustrative example approaches for safe and sustainable practice: (E) modify and/or re-use, (I) improve, (A) use alternative.

2002). The scorecard measures performance in a number of important dimensions or indicators. The overall score is less important than the score compared to a reference score. The reference score is often indicated by a “footprint” in a radar diagram. The scorecard appears to suit offshore geotechnics, because offshore geotechnics adds value primarily in the early phases of the life cycle of a facility: site selection, development of facility options and detailed design. These early phases are especially influential for the sustainability of the facility (SISG 1993). The tools for offshore geotechnics typically comprise: (1) desk study and re-processing of available data, (2) geophysical survey, (3) in-situ testing, (4) sampling and laboratory testing, (5) empirical estimation, geotechnical modelling and risk assessment (ISO 2003, ISO 2009b, Van Staveren 2006). For the service scorecard, the dominating activities are those that involve offshore crews, vessels and equipment, i.e. (2), (3) and (4). Figure 4 shows proposed scorecards for assessing the sustainability of service and facility with respect to offshore geotechnics. Issues of particular importance to offshore geotechnics are indicated in italics. The principal topics are the same for both scorecards, the subcategories differ.

Figure 1. Sustainable practice according to IPIECA andAPI (2003).

Figure 2. Interaction between Service (S) and Facility (F): the total surface of both S and F relates to the overall improvement of sustainability.

aspects, as illustrated in Figure 2. Focus on service only may be at the expense of sustainability for the facility, resulting in net loss to society. Generic approaches for improving safe and sustainable activities for a facility or service are: (E) modify/re-use an existing facility or existing service data, (I) improve on existing concepts, (A) use alternative/new concepts. Figure 3 gives illustrative examples of the three approaches. In some cases it may be possible to evade the facility altogether, a special case ofApproach (E).Technology is often an important driver for improvement.

3

SCORECARDS: ASSESSING SUSTAINABILITY

4

The Infield Report (2009) forecasts 81,300 km of offshore pipeline to be installed globally over the 2009 to 2013 period. These pipelines may be placed in the following groups: in-field flowlines, tieback pipelines and grid interconnectors. As explained in Section 2 and Figure 3, three generic approaches are proposed (E), (I) and (A) for improving safe and sustainable practice of service and facility. The (E) approach is a difficult but obvious possibility for adding facility value to tieback pipelines and grid interconnectors. Transport of gas/fluid/solids by vessel instead of a pipeline would be an (A) example. Examples for improving (I) an existing facility concept could include:

Any assessment of alternatives requires benchmarking. Common tools for assessment are (a) scorecards, (b) life cycle techniques and (c) full valuations (Fisk © 2011 by Taylor & Francis Group, LLC

PIPELINES EXAMPLE

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– Improved routing to suit seabed conditions and geohazards. – Improved site-specific design resulting in reduced use of pipeline materials.

Figure 4. Proposed score card for assessing the sustainability of service (a) and facility (b).

Figure 5. Example of using score cards to assess the safety and sustainability for both service and facility aspects of routing.

The following subsections clarify these examples, including the use of the proposed scorecards and the interaction between service and facility.

4.1 Improved routing Important trends observed for improving the routing of pipelines are (a) the role of Autonomous Underwater Vehicles (AUVs) for deepwater site characterization by geophysical techniques (e.g. Campbell & Burrell 2003) and (b) the increased use of knowledge repositories or databases. Figure 5 shows the resulting score © 2011 by Taylor & Francis Group, LLC

cards, comparing conventional techniques (dashed red line) and improved techniques (solid blue line) for (a) and (b) combined. In comparison with conventional towed systems, AUVs conduct geophysical survey close to the seafloor, improving data resolution and accuracy.AUV operations are not constrained to long tow cable, making curved survey lines easier to execute. An AUV survey is typically faster and results are more accurate. Service sustainability improves for the categories “Economy”, “Material Use, Waste and Recycling” and “Energy Efficiency and Emissions”. Facility sustainability improves for the same categories, primarily as a result of the higher data resolution allowing optimisation of facility design and risk management. Increasing use is made of knowledge repositories held by national geological surveys (public domain) and the regional geoscience industry (private domain) for routing of pipelines (Monteiro da Costa et al. 2009, Tervoort & Peuchen 2010). For offshore geotechnics as service, this leads to more efficient feasibility and planning of offshore surveys, predominantly adding value for the category “Economy” but also favourably affecting most of the other categories including safety. The use of knowledge repositories improves facility sustainability primarily for the design phase. The categories “Economy”, “Material Use, Waste and Recycling” and “Energy Efficiency and Emissions” have potential for significant improvement.

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Figure 7. Example of using score cards to assess the safety and sustainability for both service and facility aspects of site-specific design.

Figure 6. SMARTSURF® operated in combination with a LARS.

It is suggested that “sparse data” technologies developed for oil and gas reservoir characterisation can add further value in the future. Surprisingly, some of these technologies rely on shallow CPT results (Geel & Donselaar 2007).

4.2

Improved site-specific design

The (I) approach for improved site-specific design is effective once a pipeline route has been selected. The observed trend for added value is one of integrated use of more advanced pipeline-soil models and specific geotechnical tools for on-site acquisition of important parameter values required for these geotechnical models. Examples of more advanced geotechnical models for surface-laid pipelines are given by White & Randolph (2007) and others. Examples of specific geotechnical tools are the FUGRO SMARTPIPE® (Hill & Wintgens 2009) for in-situ pipeline section testing and the companion SMARTSURF® for rapid in-situ testing and sampling of shallow seabed. The FUGRO SMARTPIPE® has been operational since 2008. It deploys an instrumented pipeline section, subjected to static and cyclic soil/pipe interaction forces in 3 dimensions (vertical, axial and lateral) in water depths up to 2,500 m. The equipment also includes supplementary cyclic mini T-bar and/or mini ball testing for characterising the upper portion of the seabed. The SMARTSURF® (Figure 6) allows efficient extrapolation of data acquired by pipeline section testing. This lightweight tool has been in use since 2009. It is equipped with retractable mudmats and © 2011 by Taylor & Francis Group, LLC

seafloor detection tools for operation on extremely soft seabed. Geotechnical tools include (a) 3 m push system for conventional CPT/T-bar/ball penetrometers, (b) a 1 m push system for pressure-compensated, cyclic mini T-bar or ball penetrometers and (c) 2 m PISTON SAMPLER. A single deployment permits use of each of these tools. Fugro’s LARS (Launch And Recovery System, Figure 6) suits the FUGRO SMARTPIPE® and the SMARTSURF® . It provides a step-change in service safety for deployments by A-frame. This is because of full remote control of back-deck operations. The scorecard of Figure 7 illustrates the interaction between the service and facility aspects according to Figure 2a. It can be inferred that improved sitespecific design increases on-site (vessel) operations. This has an adverse effect on service sustainability scores in the categories “Economy”, “Material Use, Waste and Recycling” and “Energy Efficiency and Emissions”. However, the facility score improves in the same categories, leading to significant overall benefits. It is suggested that a future scenario could possibly include regional “soil-pipeline interaction catalogues” which add strategic value to on-site testing.

5 WIND FARMS EXAMPLE The offshore geotechnics industry in Europe is showing rapid growth in line with rapid growth of offshore renewable energy, notably many wind farms. A single wind farm may comprise up to hundreds of identical bottom-founded offshore structures, spaced at about 500 m. Typically, the cost of offshore substructures, including foundation elements, accounts for about 15% to 25% of the total initial wind farm investment (Junginger et al. 2004). This unprecedented scenario is generating service and facility concepts for each of the categories (E), (I) and (A), including the topics of pile driving noise for bio-diversity and modular structure build-ups for social investments (Carbon Trust 2009). Up to about 2007, offshore geotechnics for wind energy converters saw approaches applied in the offshore oil and gas industry competing with traditional onshore geotechnics. The following (I) trends can

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Offshore geotechnics offers added value by service. It is therefore essential to consider the interaction between service and facility aspects, i.e. the Client facilities and operations affected by information and advice produced by the offshore geotechnics industry. Focus on service only may be at the expense of sustainability for the facility, resulting in net loss to society. Scorecards appear practical. Specific criteria for scoring are proposed for offshore geotechnics as service and for facility sustainability. A brief examination of offshore geotechnics for pipelines and wind farms shows ongoing improvements and future opportunities for added value. These include: (a) enhanced integration of information from knowledge repositories, soil catalogues and advanced on-site testing; (b) implementation of “sparse data” technologies developed for oil and gas and gas reservoir characterisation; (c) introduction of rapid/robust probing techniques in combination with enhanced integration; and (d) development of alternative foundation concepts.

now (2010) be observed for tools used in offshore geotechnics: – As for pipelines, increasing use is made of knowledge repositories for characterising a wind farm site. Effective exploitation of this desk study tool greatly improves most factors of the service scorecard. – Most jack-up and one-off operations for in-situ testing and sampling are now replaced by specialist vessels, safe and with good weather endurance, with continuous operations only interrupted by periodic port calls and dry-docking. This “industrialisation” significantly improves service safety, but also improves scores on economy, energy efficiency and material use. – As evident elsewhere in offshore geotechnics, parameter values for geotechnical models are increasingly inferred from in-situ tests correlated to published, well-defined “soil catalogues”, for example based on results from over 60 International Geotechnical Experimentation Sites (Mayne et al. 2009). In practice, reliance on site-specific supplementary laboratory test results is limited. Insitu tests such as the Cone Penetration Test (CPT) provide results with reasonably well-defined accuracy (ISO 2010) compared to the usually overrated (Barends 2005) and lower accuracy achieved by a process of sampling, sample handling, laboratory testing (CUR 2002). This trend for in-situ testing adds significant value in terms of time schedule for data delivery and usually provides improvement on most other points of the service scorecard. It is suggested that extremely rapid/robust probing techniques can provide future potential. This should be in combination with knowledge repositories, advanced geophysical survey interpretation and regional soil catalogues. – Geotechnical modelling increasingly focuses on fine-tuning of soil-structure interaction. A brief literature survey indicated no less than 63 publications in three years on optimising the PY model for monopile design. However, limited future potential is foreseen here for further increase of safe and sustainable practice. The most significant offshore geotechnics service values for facility sustainability should probably be sought in helping to develop alternative (A) concepts for (a) offshore structures that can carry wind energy converters (Carbon Trust 2009) and (b) offshore structures for transport and storage of the generated energy (Carbon Trust 2009, Diepeveen 2009).

6

CONCLUSION AND DISCUSSION

Safe and sustainable practice is broad and complex. It is difficult to agree performance indicators that are pragmatic, consistent and transparent. For service industries the concept of sustainability is slowly emerging and only now (2010) increasingly mandatory for individual service organisations. © 2011 by Taylor & Francis Group, LLC

ACKNOWLEDGEMENTS The authors gratefully acknowledge Fugro’s commitment and support to improving safe and sustainable practice. The opinions expressed in this paper are those of the authors. They are not necessarily shared by Fugro. REFERENCES Barends, F.B.J. 2005.Terzaghi Oration 2005;Associating with advancing insight. In Proceedings of the 16th International Conference on Soil Mechanics and Geotechnical Engineering; 16ICSMGE, 2005, Osaka, Vol. 1:217–248. Rotterdam: Millpress. Boyle, B. & Depraz, S. 2006. Oil and gas industry guidance on voluntary sustainability reporting. In SPE international conference on health, safety, and environment in oil and gas exploration and production held in Abu Dhabi, U.A.E., 2–4 April 2006. Paper no. SPE 98585. Brundtland, G.H. (chairman), World Commission on Environment and Development 1987. Our common future. Oxford: Oxford University Press. Campbell, K.J. & Burrell, R. 2003. Deepwater development fast-tracking: the critical role AUV surveys play in integrated site investigation and geohazard assessment. In 7th Annual Offshore West Africa Conference and Exhibition, 11–13 March 2003, Windhoek, Namibia. Carbon Trust. 2009. Carbon Trust and Ed Miliband launch global competition to cut the cost of offshore wind energy. http://www.carbontrust.co.uk/news/news/presscentre/2009/pages/2009.aspx (accessed January 18th, 2010). Cullen, W.D. (Lord) 1990. The public inquiry into the Piper Alpha disaster. London: HMSO. CUR Civieltechnisch Centrum Uitvoering Research en Regelgeving. 2002. Research of compression tests. CURrapport 2002-2. Gouda: CUR. (in Dutch).

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Diepeveen, N.F.B. 2009. Closed-loop fluid pumping as a means to transfer wind energy. In European Offshore Wind Conference & Exhibition Stockholm, Sweden, 14–16 September 2009. Fisk, D. 2002. Sustainability & diversity – after the World Summit. In M. Cook, M. et al. (eds.), Offshore site investigation and geotechnics; ’Diversity and sustainability; Proceedings of an international conference held in London, UK, 26–28 November 2002: 9–12. London: Society for Underwater Technology. Geel, C.R. & Donselaar, M.E. 2007. Reservoir modelling of heterolithic tidal deposits; sensitivity analysis of an object-based stochastic model. Netherlands Journal of Geosciences – Geologie en Mijnbouw 86(4): 403–411. Hill, A.J. & Wintgens, J.F. 2009. In-situ measurement of pipesoil interaction in deep water – results of a successful offshore campaign. In Proceedings of the SUT Annual Conference 2009, Perth, Western Australia. Infield Systems Limited. 2009. Offshore pipelines & control lines market update report 2009/13. London: Infield Systems Limited. IPIECA International Petroleum Industry Environmental Conservation Association & API American Petroleum Institute. 2003. Compendium of sustainability reporting practices and trends for the oil and gas industry: joint corporate reporting project. London: IPIECA, & Washington: API. ISO International Organization for Standardization. 2003. Petroleum and natural gas industries: specific requirements for offshore structures – Part 4: geotechnical and foundation design considerations. International Standard ISO 19901-4. Geneva: ISO. ISO International Organization for Standardization. 2009a. Guidance on social responsibility. Draft International Standard ISO/DIS 26000. Geneva: ISO. ISO International Organization for Standardization. 2009b. Risk management – Principles and guidelines. International Standard ISO 31000. Geneva: ISO.


ISO International Organization for Standardization. 2010. Geotechnical investigation and testing – Field testing – Part 1: Electrical cone and piezocone penetration tests. ISO/NP 22476-1. Junginger, M., Faaij, A. & Turkenburg, W.C. 2004. Cost reduction prospects for offshore wind farms. Wind Engineering 28(1): 97–118. Mayne, P.W., Coop, M.R., Springman, S.M., Huang, A.B. & Zornberg, J.G. 2009. Geomaterial behavior and testing. In M. Hamza, M. Shahien, & Y. El-Mossallamy (eds.), Proceedings of the 17th International Conference on Soil Mechanics and Geotechnical Engineering, Alexandria, Egypt, 5–9 October 2009: 2777-2872. Rotterdam: Millpress. Monteiro da Costa, R., Lopes Teixeira, J., Des Déserts, L., Harris, R.J.S. & Burban, P.Y. 2009. The intentions and achievements of the Angola Deepwater Consortium. In OTC.09: Proceedings 2009 Offshore Technology Conference, 4–7 May, Houston, Texas, USA. OTC Paper 19940. SenterNovem. 2008. How to purchase sustainable external consultancy. The Hague: SenterNovem. (in Dutch). SISG Site Investigations Steering Group. 1993. Site investigation in construction. 4 Vols. London: Thomas Telford. Tervoort, E. & Peuchen, J. 2010. Exploring the shallow subsurface of the North Sea. In First International Conference on Frontiers in Shallow Subsurface Technology, 20–22 January, 2010, Delft, The Netherlands. Van Staveren, M.Th. 2006. Uncertainty and ground conditions: a risk management approach. Amsterdam: Elsevier. White, D.J. & Randolph, M.F. 2007. Seabed characterisation and models for pipeline-soil interaction. In The proceedings of the Seventeenth (2007) International Offshore and Polar Engineering Conference ISOPE, Lisbon, Portugal, July 1–6, 2007, Vol. 2: 758–769. Cupertino: International Society of Offshore and Polar Engineers.

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Frontiers in Offshore Geotechnics II

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Geotechnics and Earthquake Geotechnics Towards Global Sustainability

Frontiers in Biomedical Polymer Applications, Volume II

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New Frontiers in Angiogenesis

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Frontiers in Geofluids

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New frontiers in economics

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Rockets (Frontiers in Space)

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Frontiers in Offshore Geotechnics II

Offshore

Geotechnics and Earthquake Geotechnics Towards Global Sustainability

Frontiers in Biomedical Polymer Applications, Volume II

Soil Behaviour in Earthquake Geotechnics

Offshore Finance

Offshore Hydromechanics

Plasticity and Geotechnics

Geotechnics for building professionals

Offshore Pipelines

Offshore Finance

Forging New Frontiers: Fuzzy Pioneers II

Wstęp do offshore

Offshore Geotechnical Engineering

Offshore Magazine - January 2011

New Frontiers in Economics

Frontiers in Pension Finance

Satellites (Frontiers in Space)

Frontiers in Viral Hepatitis

New Frontiers in Angiogenesis

New Frontiers in Angiogenesis

Frontiers in Polymer Research

Frontiers in Geofluids

Frontiers in adaptive control

New frontiers in economics

Frontiers in Tissue Engineering

Frontiers in Crystal Engineering

Frontiers in Tissue Engineering

Offshore Magazine - February 2011

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