Embankments on Organic Soils (Developments in Geotechnical Engineering)

Developments in Geotechnical Engineering, 80

Embankments on Organic Soils

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Developments in Geotechnical Engineering, 80

Embankments on Organic Soils

Edited by J. Hartlen

Swedish Geotechnical Institute, S-581 93 Linkoping, Sweden and W. Wolski

Department of Geotechnics, Warsaw Agricultural University, ul. Nowoursynowska 166,OZ-766 Warsaw, Poland

1996

ELSEVIER Amsterdam - Lausanne - New York - Oxford - Shannon -Tokyo

Further titles in this series: VOlUmeS 2, 3, 5-7, 9, 10, 12, 13, 15, 16A, 22 and 26 are out of print G. SANGLERAT - THE PENETROMETER AND SOIL EXPLORATION R. SILVESTER - COASTAL ENGINEERING. 1 AND 2 L.N. PERSEN - ROCK DYNAMICS AND GEOPHYSICAL EXPLORATION Introductionto Stress Waves in Rocks 11. H.K. GUPTA AND B.K. RASTOGI - DAMS AND EARTHQUAKES 14. B. VOIGHT (Editor) - ROCKSLIDES AND AVALANCHES. 1 and 2 17. A.P.S. SELVADURAI - ELASTIC ANALYSIS OF SOIL-FOUNDATION INTERACTION 18. J. FEDA - STRESS IN SUBSOIL AND METHODS OF FINAL SETTLEMENT CALCULATION 19. A. KEZDI - STABILIZED EARTH ROADS 20. E.W. BRAND AND R.P. BRENNER (Editors) - SOFT-CLAY ENGINEERING 21. A. MYSLIVE AND 2 . KYSELA -THE BEARING CAPACITY OF BUILDING FOUNDATIONS 23. P. BRUUN - STABILITY OF TIDAL INLETS -Theory and Engineering 24. Z. BAZANT - METHODS OF FOUNDATION EGlNEERlNG 25. A. KEZDI - SOIL PHYSICS - Selected Topics 27. D. STEPHENSON - ROCKFILL IN HYDRAULIC ENGINEERING 28. P.E. FRIVIK, N. JANBU, R. SAETERSDAL AND L.I. FINBORUD (Editors) -GROUND FREEZING 1980 29. P. PETER -CANAL AND RIVER LEVEES 30. J. FEDA - MECHANICS OF PARTICULATE MATERIALS -The Principles 31. Q. ZARUBA AND v . MENCL - LANDSLIDES AND THEIR CONTROL Second completely revised edition 32. I.W. FARMER (Editor) -STRATA MECHANICS 33. L. HOBST AND J. ZAJiC - ANCHORING IN ROCK AND SOIL Second completely revised edition 34. G. SANGLERAT, G. OLlVARl AND B. CAMBOU - PRACTICAL PROBLEMS IN SOIL MECHANICS AND FOUNDATION ENGINEERING, 1 and 2 35. L. RETHATI-GROUNDWATER IN CIVIL ENGINEERING 36. S.S. VYALOV - RHEOLOGICAL FUNDAMENTALS OF SOIL MECHANICS 37. P. BRUUN (Editor) - DESIGN AND CONSTRUCTION OF MOUNDS FOR BREAKWATER AND COASTAL PROTECTION 38. W.F. CHEN AND G.Y. BALADI -SOIL PLASTICITY -Theory and Implementation 39. E.T. HANRAHAN -THE GEOTECTONICS OF REAL MATERIALS: THE eg'k METHOD 40. J. ALDORF AND K. EXNER -MINE OPENINGS - Stability and Support 41. J.E. GILLOT - CLAY IN ENGINEERING GEOLOGY 42. AS. CAKMAK (Editor) -SOIL DYNAMICS AND LIQUEFACTION 43. A.S. CAKMAK (Editor) - SOIL-STRUCTURE INTERACTION 44. A.S. CAKMAK (Editor) -GROUND MOTION AND ENGINEERING SEISMOLOGY 45. AS. CAKMAK (Editor) - STRUCTURES, UNDERGROUND STRUCTURES, DAMS, AND STOCHASTIC METHODS 46. L. RETHATI - PROBABILISTIC SOLUTIONS IN GEOTECTONICS 47. B.M. DAS -THEORETICAL FOUNDATION ENGINEERING 48. W. DERSKI, R. IZBICKI, I. KlSlEL AND Z. MROZ - ROCK AND SOIL MECHANICS 49. T. ARIMAN, M. HAMADA, A.C. SINGHAL, M.A. HAROUN AND A.S. CAKMAK (Editors) - RECENT ADVANCES IN LIFELINE EARTHQUAKE ENGINEERING 50. B.M. DAS - EARTH ANCHORS 51. K. THIEL - ROCK MECHANICS IN HYDROENGINEERING 52. W.F. CHEN AND X.L. LIU -LIMIT ANALYSIS IN SOIL MECHANICS 53. W.F. CHEN AND E. MIZUNO - NONLINEAR ANALYSIS IN SOIL MECHANICS 54. F.H. CHEN -FOUNDATIONS ON EXPANSIVE SOILS 55. J. VERFEL - ROCK GROUTING AND DIAPHRAGM WALL CONSTRUCTION 56. B.N. WHITTAKER AND D.J. REDDISH - SUBSIDENCE - Occurrence, Prediction and Control 57. E. NONVEILLER - GROUTING, THEORY AND PRACTICE 58. v . KOLAR AND I. NEMEC -MODELLING OF SOIL-STRUCTURE INTERACTION 59A R.S. SINHA (Editor) - UNDERGROUND STRUCTURES - Design and Instrumentation 598 R.S. SINHA (Editor) - UNDERGROUND STRUCTURES - Design and Construction 60. R.L. HARLAN, K.E. KOLM AND E.D. GUTENTAG - WATER-WELL DESIGN AND CONSTRUCTION 61. I. KASDA - FINITE ELEMENT TECHNIQUES IN GROUNDWATER FLOW STUDIES 62. L. FIALOVSZKY (Editor) -SURVEYING INSTRUMENTS AND THEIR OPERATION PRINCIPLES

1. 4. 8.

63. 64. 65. 66. 67. 68. 69. 70. 71. 72 73 74. 75. 76. 77. 70. 79.

H. GIL - THE THEORY OF STRATA MECHANICS H.K. GUPTA - RESERVOIR-INDUCED EARTHQUAKES V.J. LUNARDlNl - HEAT TRANSFER WITH FREEZING AND THAWING T.S. NAGARAI - PRINCIPLES OF TESTING SOILS, ROCKS AND CONCRETE E. JUHASOVA - SEISMIC EFFECTS ON STRUCTURES J. FEDA - CREEP OF SOILS - and Related Phenomena E. DULACSKA - SOIL SETTLEMENT EFFECTS ON BUILDINGS D. MlLOVlc - STRESSES AND DISPLACEMENTS FOR SHALLOW FOUNDATIONS B.N. WHITTAKER, R.N. SINGH AND G. SUN - ROCK FRACTURE MECHANICS - Principles, Design and Applications M.A. MAHTAB AND P. GRASS0 - GEOMECHANICS PRINCIPLES IN THE DESIGN OF TUNNELS AND CAVERNS IN ROCK R.N. YONG, A.M.O. MOHAMED AND B.P. WARKENTIN - PRINCIPLES OF CONTAMINANT TRANSPORT IN SOILS H. BURGER (Editor) - OPTIONS FOR TUNNELING 1993 S. HANSBO - FOUNDATION ENGINEERING R. PUSCH - WASTE DISPOSAL IN ROCK R. PUSCH - ROCK MECHANICS ON A GEOLOGICAL BASE T. SAWARAGI - COASTAL ENGINEERING - WAVES, BEACHES, WAVE-STRUCTURE INTERACTIONS 0. STEPHANSSON, L. JlNG AND CHIN-FU TSANG (Editors) - COUPLED THERMO-HYDRO-MECHANICAL PROCESSES OF FRACTURED MEDIA

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ELSEVIER SCIENCE B.V. Sara Burgerhartstraat 25 P.O. Box 21 1,1000 AE Amsterdam, The Netherlands

ISBN: 0-444-88273-1

0 1996 Elsevier Science B.V. All rights reserved. No part of this publication may be reproduced, stored in a retrieval system or transmitted in any form or by any means, electronic, mechanical, photocopying, recording or otherwise, without the prior written permission of the publisher, Elsevier Science B.V., Copyright & Permissions Department, P.O. Box 521,1000 AM Amsterdam, The Netherlands. Special regulations for readers in the USA - This publication has been registered with the Copyright Clearance Center Inc. (CCC), 222 Rosewood Drive, Danvers, MA 01923. Information can be obtained from the CCC about conditions under which photocopies of parts of this publication may be made in the USA. All other copyright questions, including photocopying outside of the USA, should be referred to the copyright owner, Elsevier Science B.V., unless otherwise specified. No responsibility is assumed by the publisher for any injuty and/or damage to persons or property as a matter of products liability, negligence or otherwise, or from any use or operation of any methods, products, instructions or ideas contained in the material herein. This book is printed on acid-free paper Printed in The Netherlands

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1x

The Authors w. Wolski

Department of Geotechnics, Warsaw Agricultural University, Warsaw, Poland

Z. Lechowicz

Department of Geotechnics, Warsaw Agricultural University, Warsaw, Poland

A. Szymanski

Department of Geotechnics, Warsaw Agricultural University, Warsaw, Poland

T. Baranslu

Department of Geotechnics, Warsaw Agricultural University, Warsaw, Poland

J. Hartlen

Swedish Geotechnical Institute Linkoping, Sweden

U. Bergdahl

Swedish Geotechnical Institute Linkoping, Sweden

P. Carlsten

Swedish Geotechcal Institute Linkoping, Sweden

R. Larsson

Swedish Geotechcal Institute Linkoping, Sweden

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Preface Development of regions and countries is often associated with the necessity of civil engineering works on soft soils. This happens more and more frequently because of earlier utilization of areas with better foundation conditions. The most troublesome of soft soils are organic soils, mainly due to their high compressibility, much higher than in mineral soils and very low shear strength. The large diversity of organic soils with respect to their origin as well as properties make their classification, testing and engineering prediction of behaviour very difficult. That is why engineers try, in general, to avoid constructing on deep layers of organic soils; if forced by the necessities, they manage with light structures, e.g. embankments or low buildings. The authors of this book have been involved in a joint research project of test embankments on organic soils carried out in the North-Western part of Poland by the Swedish Geotechnical Institute and the Department of Geotechnics of the Warsaw Agricultural University. These studies prompted us to write the present book. The authors wish to gratefully acknowledge the help of the Polish Local Land Reclamation Office WZMUW in Pila in the performance of field works. The authors are very grateful to professor Sven Hansbo at the Chalmers University of Gothenburg, who has reviewed this book and who has given many valuable comments on the text. The authors also want to express their gratitude to the Swedish Road Administration, for them letting us use their experience and publications on the subject. The authors will also greatly acknowledge the work done by Mr Eugeniusz Koda, making the computation of examples in chapter 9, and Mr Ryszard Zycnowicz, making most of the drawing, both working at the Warsaw Agricultural University, and to Mr Jan Lindgren, the Swedish Geotechnical Institute, who has done the large job of reviewing and editing the whole book. Spring 1996 The authors

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xiii

Contents PREFACE .......................................................................................... NOTATION AND SYMBOLS ........................................................... INTRODUCTION (J. Hartlen & W. Wolski) ..........................................

xi Xxii 1

PART I: TESTING AND ANALYSIS

1. 1.1

1.2 1.3

ORGANIC SOILS (R. Larsson) ............................................................ 4 Geological origin ................................................................................... 4 General Biogenic matter Organogenous soil Chemical sediments Organic soils Engineering properties ....................................................................... 15 Soil classification ................................................................................ 16 1.3.1 Identification of soil type * Gyttja-bearing soils . Dy-bearing soils Peat Topsoils * Marl and shell soils 1.3.2 Classification according to composition * Organic soils * Calciferous soils * Sulphide-rich soils 1.3.3 Other classification systems for organic soils 1.3.4 Detailed classification of peat * Humification . Water content * Sedge fibres * Root threads

xiv

. Wood remnants Designation Organic content Tensile strength - Smell - Plasticity Acidity 1.3.5 Other classification systems for peat 1.3.6 Geotechnical classification of peat References .................................................... *

*

1.4

2. 2.1 2.2 2.3

2.4

2.5

2.6 2.7

2.8

.....................................

29

SITE INVESTIGATIONS (U. Bergdahl) ........................................... General ................................ ...................... ...................... Mapping, general survey ........... Soil layer sequence ......... .............................................. 2.3.1 Soil radar 2.3.2 Penetration testing 2.3.3 Dilatometer testing 2.3.4 Sampling Groundwater ...................................................................................... 2.4.1 Pore pressure measurement 2.4.2 Permeability measurements Strength and deformation characteristics ......................................... 2.5.1 General 2.5.2 Field vane test . . Monitoring equipment ....................................................................... Test embankments for design purposes .......................... 2.7.1 Introduction 2.7.2 Preparatory investigations 2.7.3 Location of the test embankment 2.7.4 Design of test embankments 2.7.5 Test embankment monitoring 2.7.6 Construction of the test embankment 2.7.7 Presentation of test results 2.7.8 Analvsis of test results and recommendations References .........................................................................................

31 31 32 33

53

60

66

82

xv

3. 3.1 3.2

3.3 3.4

3.5

3.6

3.7 3.8 4. 4.1

4.2

LABORATORY INVESTIGATIONS (2 Lechowicz, A. Szymanski & T. Baranski) ................................................... .............................. 85 General ............ ....................................... Routine tests ....................................................................................... 85 3.2.1 Soil density 3.2.2 Consistency limits 3.2.3 Organic content 3.2.4 Carbonate content 3.2.5 Ferrous sulphide Determination of stress history .......... .......................................... 96 3.3.1 Preconsolidation pressure 3.3.2 Coefficient of earth pressure at rest Determination of deformation and consolidation parameters by oedometer tests .................................................... .......... 101 3.4.1 Incremental loading oedometer tests Testing procedure . Deformation and consolidation parameters 3.4.2 Continuous loading oedometer tests Testing procedure Deformation and consolidation parameters Determination of deformation parameters by triaxial test ............. 112 3.5.1 Testing procedure 3.5.2 Young’s modulus and Poisson’s ratio 3.5.3 Bulk modulus and shear modulus 3.5.4 Yield envelope and creep characteristics Determination of shear strength .............................................. 3.6.1 Swedish fall-cone test 3.6.2 Laboratory vane shear test 3.6.3 Direct simple shear test 3.6.4 Triaxial test Determination of permeability......................................................... 127 References ...................................................................... ..... 131 STABILITY ANALYSIS (Z. Lechowicz) ......................................... General ....................................................................................... Shear strength used in stability analysis .......................................... 4.2.1 Problems in the evaluation of shear strength 4.2.2 Undrained shear strength

137 137 138

XVl

4.3

4.4 4.5

4.6

4.7

5. 5.1 5.2

5.3

4.2.3 Effective shear strength Methods of stability analysis ............................................................ 4.3.1 Types and scope of analysis 4.3.2 Simple procedures of stability assessment * Preliminary estimation of safe embankment height Steady seepage conditions * Sudden drawdown conditions 4.3.3 Swedish circle method 4.3.4 Simplified Bishop method 4.3.5 Janbu generalized method 4.3.6 Janbu simple routine procedure 4.3.7 Morgenstem-Price method 4.3.8 Non-circular slip surface Stability of single-stage embankment .............................................. Stability of stage-constructed embankments ................................... 4.5.1 Types and scope of analysis 4.5.2 Evaluation of the increase in undrained shear strength based on empirical relations 4.5.3 Evaluation of the increase in undrained shear strength based on laboratory testing Other approaches in stability analysis ............................................. 4.6.1 Three-dimensional analysis 4.6.2 Stability asessment by f h t e element analysis References .......................................................................................

ANALYSIS OF SUBSOIL DEFORMATIONS (A. Szymanski) .................................................................................. General ....................................................................................... Deformation and consolidation parameters .................................... 5.2.1 Selection of parameters 5.2.2 Parameters used in deformation analysis 5.2.3 Consolidation parameters Analysis of “final” deformation ........................................................ 5.3.1 Type and scope of analysis 5.3.2 Initial settlement and horizontal movement 5.3.3 Empirical prediction of ‘‘final” settlement 5.3.4 Prediction of settlement in one-dimensional consolidation

145

160 162

171

176

181 181 183

192

xvii

5.4

5.5 5.6 5.7 5.8

Consolidation analysis .......................... ...................................... 203 5.4.1 Type and scope of analysis 5.4.2 Empirical prediction of the consolidation course 5.4.3 Prediction of one-dimensional consolidation at small strains 5.4.4 Prediction of one-dimensional consolidation at large strains 5.4.5 Layered soils Consolidation analysis of subsoil with vertical drains .................... 223 Swelling analysis ..................................... ................................. 227 Development trends in deformation a olidation analysis .... 230 References .......................................... .................................... 233

PART 11: DESIGN AND CONSTRUCTION METHODS

6. 6.1 6.2 6.3

7. 7.1

7.2

METHODS OF CONSTRUCTION (J. Hartlen) .............................. General ....................................................................................... Choice of method .............................................................................. Review of basic concepts of embankment construction on organic soils ...................................................................................... 6.3.1 Load adjustment 6.3.2 Soil replacement 6.3.3 Soil improvement 6.3.4 Other techniques

240 240 241

LOAD ADJUSTMENT (P. Carlsten) ................................................ Profile lowering ................................................................................. 7.1.1 Introduction 7.1.2 Design considerations 7.1.3 Limitations Pressure berms .................................................................................. 7.2.1 Introduction 7.2.2 Design considerations and dimensioning 7.2.3 Limitations 7.2.4 Construction aspects 7.2.5 Calculation example

249 249

246

250

xviii 7.3

7.4

8. 8.1 8.2

8.3

8.4

9. 9.1 9.2

9.3

Lightweight fills ................................................................................ 259 7.3.1 Introduction 7.3.2 Different lightweight fill materials 7.3.3 Design considerations 7.3.4 Construction aspects 7.3.5 Limitations 7.3.6 Case history from lightweight fill in a transition to a bridge 7.3.7 Case history from use of lightweight fill in repairing an existing road References ....................................................................................... 272 REPLACEMENT (P.Carlsten) ........................................................ General ....................................................................................... Excavation and backfill ..................................................................... 8.2.1 Description of the method 8.2.2 Design considerations 8.2.3 Limitations 8.2.4 Construction aspects Progressive displacement .................................................................. 8.3.1 Description of the method 8.3.2 Design considerations 8.3.3 Limitations 8.3.4 Construction aspects 8.3.5 Case history References .......................................................................................

274 274 274

STAGED CONSTRUCTION (W. Wolski) ...................................... General ....................................................................................... Precompression technique ............................................................... 9.2.1 Introduction 9.2.2 Design considerations . Surcharging to eliminate primary consolidation settlement * Surcharging to compensate for secondary compression 9.2.3 Design parameters 9.2.4 Limitations 9.2.5 An example - the Dalarovagen road Vertical drains .................................................................................. 9.3.1 Introduction

293 293 294

281

292

304

XiX

9.4

9.5 9.6

9.7

9.8

9.3.2 Installation of vertical drains 9.3.3 Design considerations 9.3.4 Design parameters 9.3.5 Limitations. Antoniny case history Construction monitoring ............................................... 318 9.4.1 Introduction 9.4.2 Instrumentation 9.4.3 Interpretation of the monitoring results Construction aspects ............................. 322 Design example for staged embankment with the use of vertical drains ........................... ...... ........ 323 9.6.1 Introduction 9.6.2 Stress distribution under the embankment axis during the first stage 9.6.3 Prediction of the immediate and consolidation settlements at the first stage of embankment loading 9.6.4 Consolidation performance at the first stage of loading 9.6.5 Total settlement at the end of the first stage 9.6.6 Shear strength increase due to the first stage of loading 9.6.7 Stress distribution in the subsoil under the embankment centre line, due to the second stage of loading 9.6.8 Prediction of the total settlement S, under the second stage of loading 9.6.9 Consolidation performance at the second stage of loading 9.6.losettlement of the subsoil after a lapse of two years from the beginning of consolidation 9.6.11Final remarks Design example for the staged embankment with surcharging ...... 345 9.7.1 Introduction 9.7.2 Required height of the embankment at the second stage of loading 9.7.3 Prediction of the settlements under the second stage embankment (with surcharge) 9.7.4 Vertical stress distribution after removal of the surcharge 9.7.5 Prediction of the swelling behaviour after removal of the surcharge 9.7.6 Prediction of the secondary compression behaviour of the subsoil 9.7.7 Final remarks References .....................

xx 10. LIME AND LIMECEMENT COLUMNS (P. Carlsten) ................. 355 10.1 Description of the method ................................................................. 355 10.1.1 What happens inside a lime/cement column? 10.1.2 Installation of limekement columns 10.2 Requirements for field and laboratory investigations ........ 10.2.1 Field investigations 10.2.2 Laboratory investigations 10.2.3 Test columns ................................... 360 10.3 Design considerations....................... 10.3.1 Introduction 10.3.2 Demands on the admixtures 10.3.3 Choise of admixture for different kinds of soils 10.3.4 Stability calculations * Embankments on horizontal ground surfaces or ground surfaces with slight lateral inclination . Embankments along a slope 10.3.5 Settlement calculations - Magnitude of settlements Distribution of load - limekement column and unstabilised soil * Settlementhime relationship ............................ 373 10.4 Limitations ......................... Construction aspects ........... ................................................... 374 10.5 ............................ 375 10.6 Requirements for field mea 10.6.1 Determination of the shear strength of lime/cement columns 10.6.2 Inspection of settlements 10.7 Example: Dimensioning of lime columns for reduction of settlements and for stabilisation of a road embankment .................................... 377 on soft and organic clay ................... 10.7.1 Introduction 10.7.2 Dimensioning of lime column reinforcement Calculation of settlements in stabilised clay . Settlementhime relationship 10.7.3 Stability calculation * Stability during construction . Stability of the road when in use 10.7.4 Results from dimensioning . Transition to the culverts * Construction schedule . Inspection and follow-up . Comments

xxi Case history - Bridge foundation on soft clay stabilised with lime columns 10.8 References .......................................................................................

396

. 400 11. OTHER METHODS (P. Carlsten) ..................................... ............................ ..................... 400 11.1 Reinforcement ... 11.1.1 Tradition S 11.1.2 Geotextiles . Introduction and description of the method * Construction aspects . Applications .................................. ...................... 406 11.2 Pile foundation ........ 11.2.1 Description o ethod and constructi ................................... 11.3 References ...............

AUTHOR INDEX ..................................................................................... .410 SUBJECT INDEX .....................................................................................

4 17

xxii

Notations and symbols -

area ratio (constant for a specific cone)

-

drain thickness

- width of embankment slope area of column base

A

-true

AQC

- failure surface

b

- load factor -

width of slice

- width of embankment -

B

width of drain

-width

of loaded area

-reduced width of embankment C r

- effective cohesion intercept

%

- coefficient of consolidation at horizontal drainage

%~ Ca

-effective field coefficient of consolidation at horizontal drainage

Cv C

-

apparent cohesion intercept

-

coefficient of consolidation at vertical drainage

-

constant for the fall cone test depending on the apex angle of the cone and the condition of the soil (undisturbed or remoulded) K0-consolidated test

CKoU

-undrained

Co

- compression index

Cee CC

- modified compression index -

continuous consolidation oedometer test pore pressure gradient oedometer test

CG

-constant

Cr

- recompression index

Cre CRL

-modified recompression index -

constant rate of loading oedometer test

xxiii constant rate of strain oedometer test

CRS

-

Cs

- swelling index

Cse

-modified

C~

-coefficient of secondary compression

C~

-modified secondary compression index

Cs

-coefficient of secondary compression at recompression

swelling index

-diameter of drainage well -depth of cone penetration

d/L

-value defining the location of the failure surface in Janbu's simplified method

dA

-bottom area of vertical column

ds

-

diameter of disturbed zone

dw D

-

equivalent diameter of drain

- depth factor -

diameter of sample

-diameter

of soil cylinder with drain

direct simple shear test

DSS

-

Dv

- diameter of vane - void ratio void ratio

e0

-initial

cp

-void ratio at the end of primary consolidation

E

- deformation modulus - normal component of interslice force - Young's modulus

g

r

- drained modulus of elasticity tangent modulus

Ei E

-

secant modulus

Et

-

tangent modulus

E

-

undrained modulus of elasticity

ED

-

dilatometer modulus

ESL

-effective stress level

S

-initial

xxiv f

f fl

offset of the normal force from the center of rotation

-perpendicular -correction

factor in Janbu's simplified method

-coefficient in Steinbrenner's formula -coefficient for Hansbo's diagram -coefficient

in Steinbrenner's formula

f(x)

- function describing the way in which the ratio T/E varies in a slope

F

- factor of safety

Fo

-

initial factor of safety in Janbu's simplified method

F2D

- two-dimensional factor of safety

F 3D

- three-dimensional factor of safety

g

-acceleration due to gravity

G

-

h

-depth below ground surface of the point where u is estimated

he

- final height of embankment

ho h

- initial height of embankment

h+f

- thickness of embankment with surcharge

H

-

thickness of subsoil

-

degree of humification

shear modulus

- thickness of surcharge

- height of embankment -

height of sample

Hs

- safe height of embankment

Ht Hv

-factor -

height of vane

H0

-

initial height of sample

i

-

hydraulic gradient

I

-

influence factor

depending on decomposition

ID

- material index

Ih

-horizontal influence displacement factor

IL

- liquidity index

XXV

Ip

- plasticity index

I

-vertical

v

influence displacement factor

IL

-

k

- coefficient of permeability

incremental loading oedometer test

- coefficient of horizontal permeability

\

-coefficient of horizontal permeability in disturbed soil - coefficient of vertical permeability

K

-

bulk modulus of shear strength increase (depending on loading case and overconsolidation ratio)

-coefficient

Ko

-coefficient -

of earth pressure at rest

horisontal stress index

- length of failure surface at the base of a slice

1

-

length of drainage path

-

length of drain

- reduced length of embankment L/H

- draw-down ratio

Ls

-total

m

-

length of a slide

mass

- slope of ~fu/CY'v - OCR relation in log-log scales slope of the relation between log0;fu/Cr'v) and log(ESL) in the normally consolidated state (ESL _< 1)

mnQ

-

Ill o o

- slope of the relation between log0;fu/~'v) and log(ESL) in the overconsolidated state (ESL > 1)

mv M

- coefficient of volume change -

oedometer modulus defining the location of the critical slip circle in Taylor's stability chart

-value

-parameter in Cam Clay model M

r

ME

-

modulus number

-

resisting moment due to end area effects for each end plane

xxvi initial oedometer modulus

M0

-

n

-porosity

N

-total

normal force on the base of a slice

-

critical stability number

-

normally consolidated

NF

-

Cousin's stability number

NKT

-empirical

NT

- Taylor's stability number

0C

- overconsolidated

0CR

-

overconsolidation ratio

P

-

mean normal stress

P~

-equivalent isotropic pressure

Pa PSC

- atmospheric pressure

No NC

-

cone factor

Plane strain compression test

- deviatoric stress -

load applied to the subsoil

- embankment load

qQ

-measured

value of cone resistance

qf

-final embankment load

qj

-component of the specific discharge vector of pore water

%

-

%

-total

qw

-discharge

Q

-outflow of water from a drain

-

surcharge load cone resistance capacity of a drain

drain radius

r

-pore pressure ratio

R

-radius of dewatered soil cylinder

xxvii

-

R

radius (moment arm) associated with mobilized shear forces S m percentage of humification

-radius of cylindrical part of failure surface - radius of failure surface -normalized undrained shear strength in the normally consolidated state (OCR = ESL = 1) - s p a c i n g

-

of vertical drains

settlement

So

-settlement associated with primary consolidation

S~

-

S c(f+sr)

Sf Sh

final consolidation settlement

-final consolidation for embankment and surcharge load -

final settlement

- horizontal movement

Si

-

immediate settlement

-

initial settlement

Sm

- shear force mobilized at the base of a slice

Ss

S sw

St

- settlement associated with secondary compression - swelling value -

-

total settlement settlement at time t

t

- time

tf

-time to failure in a vane shear test -time at the end of the settlement process

tp

-time at the end of primary consolidation

ts

-time for swelling

tso

-designed life-time of the embankment -time at removal of the surcharge

tso

-time for 50 % primary consolidation

tgo

-time for 90 % primary consolidation

boo

-time for 100 % primary consolidation

tL

- vertical distance from the base of a slice to the line of thrust on the

xxviii left side of the slice

tR

-

T

-tangential component of interslice force

vertical distance from the base of a slice to the line of thrust on the right side of the slice

- applied peak torque during a vane test (TC)

-

triaxial compression test

(TE)

-

triaxial extension test

Th T,,

-time factor for consolidation at horizontal drainage -time factor for consolidation at vertical drainage -excess pore pressure - pore pressure pressure measured behind the conical tip during cone penetration

Uc

-pore

ub

-pore

pressure acting at the centre of a column base

-pore

pressure at an undrained bottom end in an oedometer test

uj

-

U

- degree of primary consolidation

Uh Up

-

UV

-degree of consolidation at vertical drainage

U(f+sr)

-degree of consolidation at surcharge removal

(V1)max

-highest

W. 9

-component

W. J

- displacement vector

WE

-

WN

-natural

WV W

-

WO

-total weight of a column

l,j

gradient of pore water pressure degree of consolidation at horizontal drainage of consolidation at surcharging required to produce settlements equal to primary consolidation for the embankment load

-degree

allowable rate of axial displacement in a drained triaxial test of the displacement gradient

liquid limit water content

plastic limit

- weight of a slice

xxix -horizontal distance from a slice to the center of rotation -vertical space coordinate -depth - d i s t a n c e

from the open end of a drain

- vertical coordinate

-

-

angle between the horizontal plane and the tangent to the centre of the base of a slice angle between the horizontal plane and the line of thrust on the right side of a slice

- angle of embankment slope -compressibility of the pore water -permeability change index - convective coordinate 8ij

- Kronecker's delta

Ae~,Aey, Ay~y - increments of strain components A~x,A~y,A % - increments of stress components Aq

- load increment

Ap

- pressure difference

E

-

strain

-

lateral strain

l~ij(e)

-

elastic strain increment

Eij(P)

- plastic strain increment

Eo

- strain at start of primary compression

E1

- major principal strain

e3

- minor principal strain

E1

-vertical creep rate

e~

XXX

Elf El00

-

axial strain at failure in a triaxial test

- strain at the end of primary consolidation

-

angle of internal friction

-effective angle of internal friction

7o %

- unit weight of embankment material -

unit weight of water representing the portion of the function f(x) which is used when solving for the actual factor of safety in the MorgensternPrice method

- c o n s t a n t

X,,p

-

dimensionless parameter

g

- correction factor

gl la2

-

factor including the effect of drain spacing

-

factor including the effect of soil disturbance

la3

- factor including the effect of well resistance

V

-

V'

-Poisson's ratio in drained conditions

Vu

- P o i s s o n ' s

P

-

radial coordinate

-

density of soil

Poisson's ratio ratio in undrained conditions

Pd

- dry density

P~

-density of solid particles

Pw

- density of water

%

-horizontal stress component

(Jv

-vertical stress component

Gvo

- i n

situ overburden pressure

- major principial stress (Y3

-

minor principal stress

xxxi -

0""h ij I~' N (y' p

effective stress

-effective horizontal stress -

effective stress tensor

-effective -

preconsolidation pressure

-initial

(y'

11

(Y'v l~ "VO

normal stress preconsolidation pressure

-effective pressure at%r pore pressure equalization -effective vertical stress -in

situ effective vertical stress

(Y'vf

- final effective vertical stress after removal of surcharge

(Y'vf+s

- final effective vertical stress due to permanent load and surcharge after pore pressure equalization -decrease of effective vertical stress due to surcharge removal

q~

"Cfd %

-

shear stress

-

available shear strength

-

drained shear strength

-

undrained shear strength

%c

-undrained

shear strength in triaxial compression

'l~fuD

-undrained

shear strength in direct simple shear

q~fuE

-

~fu(LAB) q~fv q7N

undrained shear strength in triaxial extension

- average undrained shear strength in laboratory tests, [0:f~C+q:f~D+q:f~E)/3l strength value obtained in field vane shear test

-shear -

shear stress

This Page Intentionally Left Blank

Introduction J. Hartldn, Swedish Geotechnical Institute W. Wolski, Department of Geotechnics, Warsaw Agricultural University

The purpose of this book is to introduce up to date knowledge of how to construct embankments on organic soil. Organic soil is a conception, which involves several types of soil from pure organic forms as peat and gyttja to transition forms towards the mineral soils, as clayey gyttja and organic clay. In the geotechnical practice, there is however not a generally accepted rule how to classify the organic soils into main groups. This book only covers questions relevant to organic soil. This means that much of the achievements reached during the last years about soft clay are not dealt with here as long as they are not relevant for design calculations associated with construction on organic soil. Embankments on organic soils are most often constructed for roads or for flood control dikes. There are dams for water retention as well as waste tailings dams founded on organic soils. In recent years temporary embankments have been utilized to preload the sot~ subsoil, including organic subsoil and in this way improving the bearing capacity before the structure is built. Such procedures are used not only for easily adjustable structures like e.g. coal yards but also for buildings. When an embankment is erected directly on soft organic soil layers, both stability and settlement problems will generally arise. The load increase and geotechnical properties of soft soil together with schedule of construction (available construction time) and acceptable future settlements are the important factors that govern the choice of construction method. Improvement of existing structures founded on organic soil create special problems. Formerly, the settlements of e.g. a road sometimes were of little importance and thus many roads were built as "floating" structures (on timber fascines) with a relatively low factor of safety. Organic soil occurs in many forms and with varying thickness. Due to this, the problems differ from site to site and the construction methods must be adopted to the conditions in each specific case. One also has to consider the experience of an individual country and the machinery available by the contractors.

The progress in understanding the behaviour of organic soils under load as well as in utilization of new techniques and materials makes it possible to undertake successfully projects which could not be performed a couple of years ago. This textbook presents the experience gained by the authors from practical consuiting work and from research projects. A research co-operation has been going on between the Swedish Geotechnical Institute and the Warsaw Agricultural University. This close cooperation has resulted in several reports and papers. The authors are all involved in the total text although the great work of making the manuscript to each chapter was divided. The textbook consists of two parts. The first part is dedicated to the testing, calculation and general behaviour of organic soils under load and the second part to design and construction methods. The first part of the textbook thus presents 9 Classification 9 Field tests 9 Laboratory testing 9 Stability analysis 9 Deformation analysis Based on these properties (basic information) different foundation methods are presented in the second part of the textbook. The methods are applicable to construction of 9 Road embankments 9 Dikes e.g. for flood control 9 Dams for water retention 9 Preloading of foundations for structures 9 Waste tailing dams on organic soils 9 Widening of existing embankments 9 Transition zones between settling embankments and stiff structures 9 Land reclamation with soft sea beds. The second part of this textbook is written in such a way that it shall be possible for a designer/contractor to pass over the basic chapters in part one. To improve the usefulness of the textbook, detailed examples are shown how to make evaluation and calculation.

PART I:

Behaviour of Organic So~lsTesting and Analysis

Chapter I

Organic Soils R. Larsson, Swedish Geotechnical Institute

1.1

GEOLOGICAL ORIGIN

General "Organic soils" or "soils with an organic content" has often been a concept with various meanings in geotechnical engineering and the rules for division into different groups have often been rather diffuse. Apart from purely organic forms of peat and gyttja, there are a large number of transitional forms towards the mineral soils. Examples are clayey gyttja, organic clay and flood- plain sediments. Soils rich in sulphides and so called "svartmocka", which mainly consists of sulphide rich silt, are often designated as organic soils. Sediments rich in calcium carbonates, such as marl and diatomaceous soil, are sometimes included in the concept and when the limits are extended very far, also shell soils are included. Topsoils which have some kind of organic content are also included among the organic soils.

Biogenic matter The organic matter in soils originates from living plants, animals and organisms, fonning biogenic matter in contrast to mineral matter. Animal life on the continents plays a relatively small role from a geological point of view. Marine animals, on the other hand, have a major role in the formation of coral reefs and shell soils, of which many are wave-washed sediments. Aquatic animals and organisms also form the basis for many other more or less biogenic sediments. A large number of formations originate directly or indirectly from plants. During transformation processes of the plants, organic products such as peat and coal are created, as well as various inorganic products.

Geological Origin

5

Organogenous soil The term "organogenous soil" is used to denote matter consisting of plant and animal remains which cannot decay into humus or gyttja. Especially in the tropical seas, this type of matter occurs in the form of coral reefs. More globally, the organogenous soil occurs in the form of shell soils. Diatomaceous soil consists of the small but resistant SiO 2 skeletons of diatoms and needles of sponge animals. Various forms of inorganic muds also occur in the deep seas.

Chemical sediments The chemical sediments, mainly calcareous soils, have been formed by a combination of mechanical accumulation and precipitation of water soluble calcium and iron compounds, with or without the aid of organisms. Marl occurs in areas with calcareous bedrock, usually in connection with peat deposits.

Organic soils Organic soils can easily be identified by their combustibility. They are formed during the decomposition of dead organic substances i.e. remnants of plants and animals. This process takes place in different ways, mainly through bacterial activity, and is intensified by a hot climate, suitable humidity and access to oxygen from the air. The processes are schematically as follows: Living plants and animals

Microscopical plants and animals

Dead organic substances

Dead organic substances

Complete oxidation mainly in the tropics

Incompleteoxidation and decomposition

Especially anaerob reduction in water

Humus CO 2, H20

etc.

#

In dry areas: Topsoils k

Relatively fast

Fig. 1.1.

In swampy areas and lakes: Peat and Dy

Gyttja

Y

Retatively slow

Schematic process during decomposition of biogenic matter. After Hallden (1961).

6

Organic Soils

Humus is a dark substance with a colloidal structure. The destruction process from dead organic substances to humus, which is the solid product of this process, is called humification. The process takes place with the aid of fungi, bacteria and other organisms. In dry areas, the further decomposition processes are so rapid that the thickness of the topsoils seldom exceeds a few tenths of a metre. In swampy areas, the processes are slower as lack of oxygen delays the oxidation. Access to air is prevented and the water in these areas may be almost totally devoid of free oxygen from dissolved air. In the upper layers, there is usually a certain amount of oxygen from precipitation, contact with the air, flowing water and fluctuating water levels. Some of this oxygen may penetrate to deeper layers. Also in the absence of free oxygen, the decay proceeds in the form of fermentation and putrefaction. This is evidenced by the evolution of gaseous products, such as methane and sulphuretted hydrogen. Other sulphides and compounds devoid of oxygen are also produced. These anaerobic processes are much slower than the decay process when there is access to free oxygen. The formation of peat areas mainly occurs in humid parts of the temperate climate zones. Mires will form and peat accumulate wherever the conditions are favourable, irrespective of altitude or latitude, but mainly in those parts of the world where the climate is relatively cold and wet and where suitable conditions are consequently most frequent, Table 1.1. The list is not complete as detailed information from several areas is unavailable - from all African countries, from several South and Central American countries, from parts of Southeast Asia and from parts of Australia/Tasmania.

Peat originates from plants and denotes the various stages in the humification process where the plant structure can still be discerned. Dy denotes the stage where the plant structure is completely destroyed. Peat is a sedentary soil which has been formed in situ from the original material. Some types of dy are also formed in situ, where they constitute the highest degree of humification of the peat. Other types of dy have been transported by water and precipitated in a colloidal form in environments with low contents of calcium. Gyttja originates from remains of plants and animals rich in fats and proteins, in contrast to peat which is formed by remains of plants rich in carbohydrates. Dead microscopic aquatic animals are dissolved and decomposed with the aid of bacteria to a flocculent substance, in which mineral particles and less decomposed remains of plants and animals are embedded. Further decomposition occurs with the aid of organisms living in the substance, such as worms and larvae. Fermentation processes generating sulphuretted hydrogen and methane complete the formation of gyttja. Gyttja formed in nutritious water is greenish in colour. In less nutritious environ-

Geological Origin Table 1.1. Rank

Percentage of national area covered by peat in different countries in rank order (after Kivinen and Pakarinen 1980 and Taylor 1983). Country

Peatland area mill. ha > 0.3 m peat

Peat area order %

Finland Canada Republic of Ireland Sweden Indonesia

10.4 170 1.2 7.6 26

33.5 18.4 17.2 17.1 13.7

6 7 8 9 10

Northern Ireland Scotland Iceland Norway Wales

0.2 0.8 1.0 3.0 0.2

12.7 10.4 9.7 9.4 7.7

11 12 13 14 15

The Netherlands Malaysia USSR Germany (GDR) Poland

0.3 2.4 150 0.6 1.4

7.4 7.2 6.7 5.1 4.4

16 17 18 19 20

Germany (GFR) Cuba USA (including Alaska) England Austria

1.1 0.5 30 0.4

4.4 3.9 3.3 2.8 2.8

21 22 23 24 25

Denmark Switzerland Hungary New Zealand Belgium

0.1 0.1 0.1 0.2

2.8 1.3 1.1 0.6 0.6

26 27 28 29 30

Uruguay Japan Yugoslavia China Italy

0.1 0.2 0.1 3.5 0.1

0.5 0.5 0.4 0.4 0.4

31 32 33 34 35

Israel Czechoslovakia France Greece Romania

0.1

0.25 0.2 0.2 0.04 0.03

36 37 38 39

Argentina Spain Australia Bulgaria

0.016 0.012 0.002 0.001

8

Organic Soils

ments, the gyttja becomes brown from mixing with brown-black dy. Gyttja has a more or less elastic consistency which is sometimes almost jellylike. Dy has a stickier consistency. Gyttja formed in areas with calcareous soils often occurs as a transitional form between gyttja and marl, called calcareous gyttja. Depending on the content of mineral particles a number of different soils occur, such as clayey gyttja, organic clay etc. Due to the increasing biological activity during deglaciation, postglacial clays in general contain some organic matter. Another type of sediment with a highly variable organic content consists of the

flood-plain sediments. These have been deposited when streams, lakes and seas at high water overflowed their natural boundaries. The flood-plain sediments thereby contain mixtures of coarse and fine mineral particles and organic matter. A special type is flood-plain clay, which is considered to have become more common as cultivation has progressed. The run-off water from the fields has thereby transported away a large amount of fine grained material previously confined by the vegetation. Flood-plain clay mainly occurs in the low-lying parts of the cultivated areas. The peat areas and the thick deposits of more or less organic soils occur to a large extent in the northern parts of the world. In northern Europe, the deposition of postglacial clay started at a time when large parts of the land were submerged under the sea. At the same time as the inland ice retreated, the temperature rose and thereby also the biological activity in the areas now free from ice. This brought an increase in the amount of plant and animal remains in the sediments that were deposited. Lakes and seas slowly became shallower through the accumulation of a mixture of mineral and organic material at the bottom and because the ongoing land-heave bays were cut off from the sea, becoming lakes and marshes. Soils such as organic clays, clayey gyttjas and gyttjas were formed as the organic content increased. The postglacial clays, which are located in topographically lower areas, thus became overlaid to a large extent by soils with increasing organic contents. The thickness of the deposits varies from place to place. Deposits of postglacial clay 10 to 20 metres thick are common, but may in extreme cases exceed 100 metres. The overlying deposits of soils designated as organic mineral or organic soils may be up to 10 metres thick. In some areas, the lakes were overgrown, becoming swamps and fens where different types of peat were formed. Such areas were also created in higher regions by topographical conditions leading to high ground water levels and marshes. The formation of peat bogs has been described in detail by Hobbs (1986). The development of a mire from open water to a raised bog is shown schematically in Table 1.2. In the first stage, the mires develop in moving water in lakes, basins and valleys

Mire stages, morphology, flora and some associated properties of some British peak (Hobbs, 1986).

Geological Origin Table 1.2.

9

10

Organic Soils

under the control of the water level, nutrients being brought in by stream flow, runoff and percolating groundwater. The nutrient supply is generally rich. The landscape at the completion of this stage is marsh-like and is commonly referred to as "fen" and the peats as "fen peat". Such mires are generally underlain by very soft organic muds of gyttja and organic clay, which can cause severe engineering problems. In the second, transitional stage, the mire relies to an increasing extent on precipitation. It continues to receive nutrients via fluctuating ground water, although to a decreasing extent. In the third stage, the mire has grown beyond the maximum physical limits of the ground water and relies solely on direct precipitation for its water supply. The peat itself acts as a reservoir holding water above the level of the ground water. The nutrient conditions are deficient since rainfall or snow contains only minute quantifies of salts, dust and dissolved gases. These mires are called raised bogs and are acid in character. The third stage does not have to be preceded by the earlier stages, but mires can develop directly on the land surface provided that the climatic and topographic conditions are favourable. Such mires, which may be extensive, are known as blanket bogs. Another presentation of the normal development of a mire from a lake to a raised bog with emphasis on vegetation and type of organic soil is shown in Table 1.3. The corresponding stratigraphy is shown in Fig. 1.2. It should be observed that the presentation in this figure is made for convenience. The different stages do not develop simultaneously. Bog raising normally starts at the centre of the old lake and only after filling of the lake is complete. The development of a typical Swedish peat bog is shown in Fig. 1.3. The height to which a bog can be raised depends mainly on the topographic and climatic conditions. The relationship between raised height, diameter of the bog basin and mean annual precipitation for some raised bogs in southern Sweden is shown in Fig. 1.4. This relationship probably varies with the average temperature. Mires are highly complex systems and changes in their environments concerning water supply, temperature and supply of nutrients alter the course of their development. The different soil layers in a profile contain various amounts of calcium. Apart from the clay particles originating from lime-rich rock, also dissolved calcium was precipitated and already in the early stages various amounts of shell bearing organisms were deposited. The remains of these may be microscopic and evenly distributed in the soil, but also wave-washed layers of shell soil occur. The carbonate contents in different layers thus vary depending on the environment at deposition. In the

Geological Origin

Table 1.3. Vegetation and peat succession (Hobbs 1986).

~~

Notes

1 lherr

FIG 1. Lake filling-the hydroseral succession. Note that bog raising normally starts at the centre of the old lake after filling is complete but is shown in the lake margin here for convenience.

11

wll b e romr natural ovsrloppng at the vortous rloqer 7 Layers o f silt brought In dutmg tloodmq may be prestnl on t h e t e n stapes 3 Condittanr 1 1 1 1 b r partuularly f t r h In calcareous regions. rich m species

12

Fig. 1.2.

. . . . .

. . . . .

. . . .

_>

__>

.__>

->

I'

L

I-

! I

L~_

i

I'.< LU

0. UJ 14.

co

10

(D

_~

(3.

to

::3 (3

43

to

E ,,.* 9

9

.c t~ 43

(D G)

10

. . . .

......

-.->

__>

....-~

-->

o L

E

r-

.,==

o

43

u~

.c

Organic Soils

I I

I i II li li

I I I I

I I II II

i

~

to

43

rr

._~

10 a)

0 UO

(D L O.

10

.c

O~

:3 0

3=

--

10 L

--

r 10 L

~)

0

-.

.=.

10

Ck ::3

.c

__

___>

10

.c c 0 to o~ 0 43 10

.to nrv

. . . .

q)

to

E

~,

43

Lake filling - the hydroseral succession. Note that bog raising normally starts at the centre of the old lake after filling is complete, but is shown in the lake margin here for convenience. (Hobbs 1986).

13

Geological Origin

3

4 .s

Fig. 1.3.

Four stages during the development of a typical Swedish raised bog. 1) A lake with gyttja deposited at the bottom. 2) The lake has become a fen with sedge peat. 3) Sphagnum mosses have invaded and the fen has been transformed into a bog. The bog has grown in height, expanded over surrounding firm ground and finally become a vegetation of pine trees on the surface. Only a small pond remains of the former lake. 4) The bog has grown further and a typical bog for central Sweden has developed, surrounded by birch and alder vegetation. Pine trees grow on the outskirts of the bog and scattered dwarf pines grow in the central parts, (Magnusson et al 1963).

14

Organic Soils

1

I

I

I

I

I

I

I

I

I

I

I

I

I

(I) (D

E

r

J~.

I I

nn nn l

-r

--__.__

~O

9

Rainfall

s9

i

D

t

o

I

o

z~

I

550 - 600 700 - 800

1

1

1

1

Diameter

9 i

1

1000

500

Fig. 1.4.

450 - 500 9

I

i

( mm )

( me~

!

900 - 1000 1

1

1500

)

The relationship between degree of convexity of raised bogs in southern Sweden (plotted as height of cupola against diameter of bog basin) and the mean annual rainfall. The data are replotted from Granlund (1932) and many of the values to the left of the broken line have been omitted for clarity. The continuous lines define so-called limiting heights for a given rainfall range, (Tallis 1983).

dry crust and other zones affected by climatic conditions, the carbonate contents have later often been greatly reduced due to weathering and leaching. In lakes with a high content of calcium in the water, various forms of calcareous gyttja and marl were formed. Weathered calcium was transported long distances from the lime-rich bedrock and occurs in large areas. In mires formed under such conditions, a layer of calcareous marl is sometimes found between the peat and the gyttja layers. This layer has been deposited through photo- reduction of dissolved

Engineering Properties

15

carbon dioxide by submerged green plants in clear shallow water penetrated by sunlight. Such layers may be up to 1.5 metres thick and may constitute engineering hazards if the underlying very soft layers are not detected in field investigations. Other types of marl layers may occur in alterations of the normal sequence for mire formation. In clay sediments, there are also often layers containing remnants of decomposed organic material, which have been transformed during processes in a reducing environment to ferrous sulphide, among other substances. The ferrous sulphide, which in pure form is completely black, occurs as dark spots, patches, bands or completely colours the soil even at moderate contents. In Finland and northern Sweden, a special type of mass transport took place. Because of the land heave and the narrow fjord-like creeks, the rivers started to erode through the earlier deposited deltaic sediments. The sediments were redeposited further and further away as the coast line moved. The redeposition occurred in still water along the rivers and the coast. Thick layers of silt and clay mixed with dead algae and remains of animals were formed. The deposition took place in a reducing environment. The decomposition of the organic matter in this environment created a special *ype of soil called "svartmocka", which covers large parts of the coastal areas of the Gulf of Bothnia but is confined to this region. It consists of silt and clay with a relatively high content of amorphous ferrous sulphide and is black in colour. The content of gyttja varies, but is usually relatively low.

1.2

ENGINEERING PROPERTIES

The engineering properties of organic soils show a great variation depending on the type and amount of organic matter. The organic matter may occur in many forms from small amounts of amorphous or colloidal substance embedded in the pores of a mineral soil to fibrous peat with a structure resembling a coarse, loosely woven mat. The effect of the organic content on the engineering properties in relation to the properties of a pure mineral soil is in the former case mainly confined to a decreased permeability and a somewhat increased tendency to creep. In the latter case, the properties are quite different in most respects. As in mineral soils, the strength of organic soils is a function of the effective stresses acting in the soil, as well as its loading history. When normalized towards the stress history, the shear strengths of organic soils are usually higher than for mineral soils. Most organic soils, however, have no significant loading history as they are fairly recent deposits in waterlogged areas. Many of these profiles do not even have a dry crust and most of the soil layers have not been subjected to any load

16

Organic Soils

other than the weight of the overlying soil. The resulting effective stresses are relatively low because of the high ground water levels and the low densities of the organic and uncompressed soils. Consequently, most organic soils have very low strengths and are extremely compressible. They also exhibit large creep effects. Highly fibrous and undecomposed organic soil has a pronounced structural anisotropy. The fibres and the plant remains usually have a horizontal orientation. The fibres constitute a horizontal reinforcement and failure surfaces in such materials usually occur as vertical fractures or horizontal shear planes parallel to the fibres. The distribution of stress from loads on the ground surface with depth is relatively small because of the fibres. The permeability of the soil is relatively high and is often many times higher horizontally than vertically. Due to the high permeability and the fibres, stability is usually not a problem in the fibrous peat itself, provided that measures are taken to prevent punching or cracking under the loaded area and that the loading is not extremely rapid. Many fibrous peats, however, are underlain by other very soft soils and serious stability problems often occur. The compressibility of the peat is very high. Even for small extemal loads it is common for the settlements to amount to more than half of the original thickness of the peat layer. On top of this, there are considerable creep deformations with time, which for most engineering tasks cannot be accepted and have to be stopped. The effects of structural anisotropy decrease with decreasing content of fibres and increasing degree of humification. The permeability also rapidly decreases as the soil becomes more humified. The engineering properties of organic soils thus depend on the type of organic matter as well as the organic content. Structural anisotropy is important not only in peat but also in other organic soils such as gyttja, where significant effects may occur even at relatively low organic contents. The engineering problems in the more humified organic soils with low permeabilities resemble the problems encountered in soft mineral clays, but are often more accentuated because of the higher compressibility, the enhanced creep effects, the very low effective stresses and strengths and the sometimes very low permeabilities.

1.3

SOIL CLASSIFICATION

Different classification systems are used in different countries. The rules for classification presented here conform with the Swedish geotechnical classification system worked out by Karlsson and Hansbo (1981) in cooperation with the Laboratory Comittee of the Swedish Geotechnical Society.

Soil Classification

17

Understanding of the stratification and properties in a soil profile is made easier if the geological history and the environmental conditions at deposition of the sediments are known. When possible, the soils should therefore first be classified in this respect as, for example flood-plain sediments, wave-washed sediments, lowmoor peat (deposited in fens) or highmoor peat (formed in raised bogs). In order to classify the soils correctly in the laboratory, an initial requirement is a determination of the organic content, the content of carbonates and possibly the content of ferrous sulphide. The natural water content, the consistency limits and the density are also valuable aids to classification. For detailed classification of peat, a number of other determinations are required. As a rule, fresh samples of organic soils can be distinguished from pure mineral soils by their odour, which originates from decayed organic substances. The odour from a sample with a low water content can be accentuated by wetting and heating of the sample.

1.3.1

Identification of soil type

Gyttja-bearing soils: 9 Gyttja is normally greenish in colour, but may be brown or red. It bleaches on drying, usually to a grey colour. In the wet state, gyttja has an elastic, rubbery consistency. It has a brittle rupture. It shrinks strongly on drying to form hard lumps with low density. 9 Clayey gytt]a in the damp state has a green-grey colour. It differs from gyttja in the wet state in that it feels sticky, due to the clay content. 9 Gyttja-bearing clay in the damp state has a dull, slightly greenish, often dark colour, sometimes brown due to the presence of dy, sometimes black or with black patches due to ferrous sulphide. Gyttja-bearing clay is less elastic and less brittle than gyttja. On the soil surface, it often cracks in a characteristic cubic pattem. 9

Gyttja-bearing silt and sand are seldom encountered.

Alkali extracts of gyttja-bearing soils are light yellow or light green in colour. 9 Calciferous gyttja can be distinguished from marl by lowering a sample into a beaker containing dilute hydrochloric acid. If the sample consists of calciferous gyttja, it will retain its gyttja skeleton.

Dy-bearing soils: 9 Dy consists of a dense, black or dark brown soil which, besides dy matter, also contains peat or gyttja matter and mineral particles. Pure dy is seldom seen. On

18

Organic Soils

drying, dy retains its dark colour. In contrast to gyttja, dy is relatively inelastic and has a mushy consistency. Like gyttja, it shrinks strongly on drying to form hard, very light lumps. Sand or silt may often be mixed with dy, giving rise to intermediate forms, such as sandy or silty dy or dy with sand or silt layers. Dy-bearing clay is uncommon. Alkali extracts of dy have a dark colour.

Peat: In practice, the classification and division of peat is based on ocular inspection of the structure and consistency and on the squeezing test according to von Post (1924), (see Table 1.7). For many engineering purposes, only a coarse division is made into three types, (see Table 1.4). 9 Fibrous peat is low-humified and has a distinct plant structure. It is brown to brownish-yellow in colour. If a sample is squeezed in the hand, it gives brown to colourless, cloudy to clear water, but without any peat matter. The material remaining in the hand has a fibrous structure. (Degree of decomposition on the von Post scale; H l-H4.) 9 Pseudo-fibrous peat is moderately humified and has an indistinct to relatively distinct plant structure. It is usually brown. If a sample is squeezed in the hand, less than half of the peat mass passes between the fingers. The material remaining in the hand has a more or less mushy consistency, but with a distinct plant structure, (H5H7).

9 Amorphous peat is highly humified. The plant structure is very indistinct or invisible. It is brown to brown-black in colour. If a sample is squeezed in the hand, more than half of the peat mass passes between the fingers without any free water running out. When squeezing, only a few more solid components, such as root fibres, wood remnants, etc. can be felt. These constitute any material remaining in the hand, (H 8-H 10).

Topsoils: In the classification of topsoils, the humus content and the composition of the mineral components are stated, e.g. somewhat humus-bearing sand, humus-bearing clay (see Table 1.5). The colour of the topsoil may be darker or lighter depending on the humus content. For one and the same mineral soil, the colour will be darker with higher humus content. Even a somewhat humus-bearing sand has a fairly dark colour. On the other hand, a humus-bearing clay has about the same colour as a pure clay.

Soil Classification Table 1.4.

19

Classification of peat on the basis of decomposition on the von Post scale. After Karlsson and Hansbo (1981).

Designation

Group

Description

Fibrous peat

H1-H4

Low degree of decomposition. Fibrous structure. Easily recognizable plant structure, primarily of white mosses.

Pseudo-fibrous peat

H5-H7

Intermediate degree of decomposition. Recognizable plant structure.

Amorphous peat

H8-H10

High degree of decomposition.No visible plant structure. Mushy consistency.

Alkali extracts of topsoil humus (mull) are normally colourless or faintly brown. Marl and shell soils:

Marl and shell soils can normally be identified on the basis of colour, structure, mode of formation and place of formation. Diatomaceous soil is best identified with the aid of a microscope. Marl is almost entirely soluble in dilute hydrochloric acid. 1.3.2

C l a s s i f i c a t i o n a c c o r d i n g to c o m p o s i t i o n

Organic soils:

On the basis of composition, the organic soils are divided into three main types: gyttja, dy and peat. To these are added topsoils. Organic mineral soils and medium organic soils are classified on the basis of the content and nature of organic material, as well as the composition of the mineral material, Table 1.5.

20 Table 1.5.

Organic Soils Guiding values for the classification of soils on the basis of organic content. After Karlsson and Hansbo (1981).

Soil group

Organic content in weight % of dry material (< 2 mm)

Examples of designations

Low-organic soils

2-6

Gyttja-bearing clay Dy-bearing silt Humus-beating, clayey sand

Medium-organic soils

6-20

Clayey gyttja Silty dy Humus-rich sand

High-organic soils

>20

Gyttja Dy Peat Humus-rich topsoil

Calciferous

soils"

On the basis of the calcium carbonate content, the fine-grained soils can be classified according to Table 1.6. Table 1.6.

Guiding values for the classification of fine-grained soils on the basis of carbonate content. After Karlsson and Hansbo (1981).

Designation

Calcareous soils Clayey or silty marl Very marly or very calciferous clay or silt Marly or calciferous clay or silt

Carbonate content in % of material < 0.06 mm

>80 80-40 40-20 20- 5

For lime content below 5 %, the modifiers "somewhat marly" or "somewhat calciferous" can be used. S u l p h i d e - r i c h soils:

At present, there are no general guidelines for the classification of soil on the basis of sulphide content.

Soil Classification 1.3.3

21

Other classification systems for organic soils

A number of classification systems for organic soils are used in various countries and are based on similar grounds. Most of them, however, have not been specially designed for geotechnical purposes. Some classification systems used or suggested in context with soil mechanics in the USSR, Poland, Canada and USA are compared to the Swedish system in Fig. 1.5. O. 10.

peats

20" ,==,,

peats

30. 40.

F-

z

50"

z

60"

LU l--

0 (0

r

ch

70" 80. 90. 100.

highly organic 0

u~ >LLI

o_

__

peaty organic soils

high organic ( gyttja, dy, peat, humusrich topsoil}

organic soils

organic

Low organic

medium

organic

mineral with org. cont. Low organic mineral mineral

high ash

i, U3

C3

medium

O

Fig. 1.5

tow ash medium ash

IIIe]l[(lI

|

mineral

|

low

ash

Low

QI,

ash ~) Z

< (.9

medium ash

>,,

vl

o medium ash

mineral sediments

high

u:~

ash

mineralorganic

0co

~ ~

____m~,r.J~C

9

8

==.__ _ _ =

(9

Comparison of some classification systems for organic soils on a basis of ash contents, (Wolski 1988).

1. Konvalov (1980) (USSR) 2. Karlsson and Hansbo (1981) (Sweden) 3. Landva et al (1983) (Canada) 4. Andrejko et al (1983) (USA) 5. Classification used at the (Poland) Department of Geotechnics, WAU, based on the reports of Okruszko (1969, 1984) and Zawadzki (1970)

22

Organic Soils

1.3.4

Detailed classification of peat

According to Hobbs (1986), a detailed classification system for peat should include the following characteristics: -

T h e colour o f the p e a t in situ, which may change rapidly on exposure to air.

-

T h e degree o f decomposition (or humification).

Wetness, normally replaced by water content and, if possible, consistency limits in geotechnical engineering. -

-

M a i n constituents; fibres, wood remnants, amorphous and granular material.

-

M i n e r a l content a n d p o s s i b l e layers. The former is usually expressed in terms

of organic content, ash content or mineral content in geotechnical engineering. - Smell. Sulphides smell strongly but may be unevenly distributed. Methane requires a detector. - Chemistry. Measurement ofpH to detemame whetherthe peat should be described as alkaline or acid. -

Tensile strength. The resistance to tensile forces is an indicator of the structure

and the condition of the fibres. W h e t h e r a plastic limit test is possible or not. This is possible on fen and some transitional peats, but not on bog peat unless it is almost completely humified. - S p e c i a l characteristics including plant types if they can be identified. -

An example of such a classification as carried out in the field would be: Dark brown, moderately decomposed (Hs), wet, mainly fine fibrous PEAT with some amorphous granular matter and occasional rotten woody remnants, odourless, neutral, horizontal resistance slightly greater than vertical. Plastic limit test possible. Plant types not identified. A similar but even more detailed system was suggested by Landva et al (1983). A classification system covering some of these points was introduced by yon Post (1924). It has been much used in Europe and has lately also been introduced in Canada. The von Post system uses the following characteristics:

Humifieation (H) The degree ofhumification is graded on a scale from 1 to 10 and designated H 1 t o H10. The various degrees ofhumification are recognized as shown in Table 1.7.

23

Soil Classification Table 1.7.

Degrees of humification according to yon Post (1924), (Landva and Pheeny 1980). D e g r e e s of h u m i f i c a t i o n

Degree of humification

Decomposition

Plant structure

Content of amorphous material

H! H2 Ha

None Insignificant Very slight

Easily identified Easily identified Still identifiable

None None Slight

H4

Slight

Some

H.s

Moderate

H,

Moderately strong

H7

Strong

Not easily identified Recognisable. but vague Indistinct (more distinct after squeezing) Faintly recognizable

H~

Very strong

Very indistinct

High

H9

Nearly complete

Almost unrecognisable

H~,

Complete

Not discernible

Considerable Considerable

High

Material extruded on squeezing (passing between fingers) Clear. colourless water Yellowish water Brown. muddy water; no peat Dark brown, muddy water; no peat Muddy water and some peat About one third of peat squeezed out" water dark brown About one half of peat squeezed out; any water very dark brown

Nature of residue

Not pasty Somewhat pasty Strongly pasty

Fibres and roots more resistant to decomposition

About two thirds of pcat squeezed out; also some pasty water Nearly all the peat squeezed out as a fairly uniform paste All the peat passes between the lingers; no free water visible

Water content (B) In the field, the water content of the peat is estimated on a scale of 5 degrees. B 1 represents air dried peat, B 2 somewhat dried peat, B 3 peat with normal water content, B 4 v e r y w e t peat and B 5 largely free water with slime. In the case of actual water contents, Landva and Pheeney (1980) have later suggested the following range s B 2 less than 500 %; B 3 500 to 1000 %; B 4 1000 to 2000 % and B 5 greater than 2000 %.

Sedge fibres (F) The content of fibres and stems from sedge is given as F. Roots should not be included. F 3 denotes a peat entirely or mainly consisting of such fibres, F 2 a high but not predominant fibre content, F 1 a low fibre content and F 0 no macroscopically discernible fibres. Landva and Pheeney suggest that also fibres from mosses and shrub rootlets should be included, provided that they are properly specified as F(H) or F(S) for Hypnum and Sphagnum mosses and F(N) for shrub rootlets.

24

Organic Soils

Root threads (R) The content of root threads, R, was given as R 3 for almost pure root mat (felt), as R 2 for high and R 1 for low contents of root threads and R 0 ifthe content is nil. If the species of root threads can be determined, it should be given in brackets after the R symbol. Landva and Pheeney suggest that "sedge fibres" and "root threads" should be exchanged to "fine fibres" and "coarse fibres". The division between fine and coarse should then be at a diameter or width greater or smaller than 1 mm. No distinction should be made between fibres, stems and rootlets, but the plant origin should, if possible, be given in brackets after the F and R symbols.

Wood remnants (V) The content of wood remnants was given by the symbols V3, V2, V 1 and V 0 along the same lines as for symbols F and R. The species of wood and its consistency should, if possible, be given in brackets after the R symbol. Landva and Pheeney suggest a division of the V symbol into W for wood remnants and N for shrub remnants. Landva and Pheeney further suggest that the peat should be designated according to its plant origin:

Plant types Bryales (moss) Carex (sedge) Equisetum (horse tail) Eriophorum (cotton grass) Hypnum (moss) Lignidi (wood) Nanolignidi (shrubs) Phragmites Scheuchzeria (aquatic herbs) Sphagnum (moss)

= = = = = = = = = =

B C Eq Er H W N Ph Sch S

Designation With few exceptions, natural peats consist of a mixture of two or more plant types. The designation adopted is to list the plant types in the descending order of content, i.e. the first symbol represents the principal component. For example, a peat classified as ErCS consists mainly of Eriophorum remnants, while the content of Carex remnants would be lower and that of Sphagnum remnants relatively low. The designation is omitted when plant types cannot be identified.

Soil Classification

25

Hobbs (1986) suggests that the von Post classification system should be further extended through the addition of symbols for organic content, anisotropy, smell, plasticity and acidity as required:

Organic content (N)A It is not possible to estimate the organic content in the field unless the peat is obviously clayey, when the von Post humification test would not be realistic. Following ignition loss determinations, the organic content may be graded as follows: N 5 greater than 95 % organic matter; N 4 95 to 80 %; N 3 80 to 60 %; N 2 60 to 40 %; N~ 40 to 20 %. A) The suggested notation N is unfortunate as it may be confused with the notation for shrub remnants suggested by Landva and Pheeney (1980).

Tensile strength (TV and TH) The tensile strength in the vertical and horizontal directions may be judged by pulling specimens apart in these directions. The following scale may be used" T o zero strength; T 1 low, say less than 2 kN/m~; T 2 moderate, say 2 to 10 kN/m~; and T 3 high or greater than 10 kN/m ~.

Smell (A) The smell, which is an indication of fermentation under anaerobic conditions, may be scaled as follows: A 0 no smell; A 1 slight; A 2 moderate; A 3 strong. Note, methane, CH 4, the main indicator of anaerobic activity, has no smell. If specially detected, it should be reported.

Plasticity (P) Plastic limit test possible P1, not possible P0

Acidity (pH) Acid pilL; neutral PH0; alkaline pH H. An example of the use ofthe von Post classification in the field extended according to Hobbs is given below:

Soil description- Dark brown, oxidizing to black, moderately decomposed H 5, mainly fine fibrous PEAT with some coarse fibres and amorphous material. Low vertical tensile strength, moderate horizontally. No smell. Plastic limit test possible. Genera not identified.

26

Organic Soils

After determinations of ignition loss and pH, this becomes H5 B 2 F 3 R 1 V 0 /

N3 TV1 TH2 Ao P1 PHo

Lvon Post

Extension proposed by Hobbs

The classification in the example can be further elaborated if the modification of the von Post classification suggested by Landva and Pheeney (1980) is used. Peats are composed of the partly decomposed remains of plant communities containing varying morphology and texture. It is this structure that affects the retention or expulsion of water in the system, provides tensile strength and ultimately differentiates one type of peat from another. The von Post classification attempts to describe peat and the structure in quantitative terms and the extended system suggested by Hobbs is designed to provide a means of correlating the types of peat with their physical, chemical and structural properties.

1.3.5

Other classification systems for peat

The Russian handbook for peat (Lazarev and Kortjunov 1982) uses percentage of humification instead of the von Post scale. The following relation between the percentage ofhumification (R) and the von Post scale (H 1to H10) is given by Schneider (1967)" H (von Post)

R%

1

5

2 3 4 5 6 7 8 9 10

10 15 20 25-30 35 45 55 65

Radforth (1969) proposed a classification system for Canadian peat (muskeg) based on the structure of peat rather than its botanical origin. The system is described in Muskeg Engineering Handbook (MacFarlane 1969). According to Radforth, this approach makes it easier to classify structure and also leads to a better

Soil Classification

27

basis for estimating mechanical properties than a purely botanical classification system. Radforth divides peat into the main categories amorphous- granular, fine-fibrous (fibre diameter _< 1 mm) and coarse-fibrous according to the main character of the structure. Besides these main features, the concepts "woody" and "non-woody" are used to characterize the peat. A subdivision can be made into 17 categories, Table 1.8. Table 1.8. Classification of peat (The Canadian system after Radforth 1969). Predominant characteristic Amorphous-granular

Category 1

2 3 4 5

~~ 7

Fine-fibrous

8 9 10 11

Coarse-fibrous

12 13 14 15 16 17

Name Amorphous-granular peat Non-woody, fine-fibrous peat Amorphous-granular peat containing non-woody fine fibres Amorphous-granular peat containing woody fine fibres Peat, predominantlv amorphous-granular. containing non-woody fine fibres, held in a woody, fine-fibrous frame work Peat, predominantly amorphous-granular containing woody fine libres, held in a woody, coarse-fibrous framework Alternate layering of non-woody, linefibrous peat and amorpht~us-granular peat containing non-woody fine fibres Non-woody, fine-fibrous peat containing a mound of coarse fibres Woody, fine-fibrous peat held in a woody. coarse-fibrous framework Woody particles held in non-woody, linefibrous peat Woody and non-woody particles held in fine-fibrous peat Woody, coarse-fibrous peat Coarse fibres criss-crossing fine-fibrous peat Non-woody and woody fine-librous peat held in a coarsc-librous framework Woody mesh of fibres and particles enclosing amorphous-granular peat containing fine fibres Woody, coarse-fibrous peat containing scattered woody chunks Mesh of closely applied logs and roots enclosing woody coarse-fibrous peat with woody chunks

Organic Soils

28

The classification is facilitated by useful photographs of the various categories in the Muskeg Engineering Handbook. In the Radforth system, no mention is made of colour, wetness, degree ofhumification or organic content and these characteristics have to be supplemented in some way. Landva and Pheeney (1980) and Hobbs (1986) have found that the system is not generally applicable, even if it is useful for large areas in Canada. Radforth suggests that also vegetal cover and topsoil should be classified on the basis of structure rather than gefiesis. The vegetal cover is divided into nine different classes, Table 1.9. Since there is often more than one type of cover, the classification can be combined, e.g. ADE. The symbols should list the types of cover in descending order of amount, i.e. the first symbol represents the principal component, the second symbol represents the next largest component and so on. Types that cover less than 25 % of the area are not considered. Table 1.9.

Properties designating nine coverage classes (The Canadian system after Radforth 1969).

Coverage type (class)

Woodiness vs. nonwoodiness

Stature (approximate height)

A B

Woody Woody

15 fl or over 5-15 fl

C D

Non-woody Woody

2-5 fl 2-5 ft

Texture (where required)

Woody Up to 2 fl Non-woody Up to 2 ft G

Non-woody Up to 2 ft

H

Non-woody Up to 4 in. Non-woody Up to 4 in.

Leathery to crisp Soft or velvety

Growth habit

Tree form Young or dwarf tree or bush Tall, grasslike Tall shrub or very dwarfed tree Low shrub Mats, clumps or patches sometimes touching Singly or loose association Mostly continuous mats Often continuous mats, sometimes in hummocks

Soil Classification

1.3.6

29

Geotechnical classification of peat

In many cases, only a simplified version of the von Post scale for degree of humification is used in geotechnical engineering for the classification of peat, together with the normal geotechnical parameters used for all soils, such as water content, consistency limits, organic content, bulk density etc. This procedure has been recommended by Helenelund (1975) and Karlsson and Hansbo (1981), among others, and is normally used in Fenno - Scandinavia. In Canada, the Radforth system has been used to some extent and in other countries regional classification systems, often related to von Post in one way or another, have been used. In Scandinavia and Canada, as well as elsewhere, strong recommendations have been made to introduce the full von Post classification system also in geotechnical engineering, with or without suitable modifications and extensions, e.g. Landva and Pheeney (1980), Hobbs (1986), Landva et al (1983), Carlsten (1988).

1.4 REFERENCES Andrejko, M.J., Fiene, F. and Cohen, A.D. (1983) Comparison of ashing techtuques for determination of the inorganic contents in peats. Testing of peats and organic soils. ASTM Special Technical Publication 820, pp. 5-20.

Carlsten, E (1988) Geotechnical properties of peat and up-to-date methods for design and constrtiction on peat. State of the Art Report. International Conference on Peat. Tallinn.

Godwin, H. (1978) Fenland; Its Ancient Past and Uncertain Future. Cambridge University Press.

Granlund, E. (1932) De svenska h6gmossarnas geologi. Deras bildningsbetingelser, utvecklingshistoria och utbredning jfimte sambandet mellan h6gmossbildning och f6rsumpning. (The geology of Swedish raised bogs). Swedish Geological Survey. Ser. C. Nr. 373. Hallden, B.E. (1961) Allm~n Geologi. Kompendium Nr. 83. Tekniska h6gskolans studentkfir. Stockholm.

Helenelund, K. V. (1975) Geotechnical peat investigations. Proceedings, Baltic Conference on Soil Mechanics and Foundation Engineering, 1. Gdansk. Vol. 1. pp. 105-123.

Hobbs, N.B. (1986) Mire morphology and the properties and behaviour of some British and foreign peats. The Quartemary Journal of Engineering Geology. Vol. 19. No. 1.

30

Organic Soils

Karlsson, R. and Hansbo, S. (1981) (in collaboration with the Laboratory committee of the Swedish Geotechnical Society). Soil classification and identification. Swedish Council for Building Research. D8:81. Stockholm. Kivinen, E. and Pakarinen, P., (1980) Peatland areas and the proportion of virgin peatlands in different countries.In: 6th International Peat Congress, Duluth, Minnesota. Konovalov, P. A. (1980) Ustojstvo fundamentov na zatorfovannych gruntach. Moskwa. Stroizdof. Landva, A.O., Korpijaakko, E. O. and Pheeney, P. E. (1983) Geotechnical Classification of Peats and Organic Soils. Testing of Peats and Organic Soils. ASTM Special Technical Publication 820. pp. 37-51. Landva, A. O. and Pheeney, P. E. (1980) Peat fabric and structure. Canadian Geotechnical Journal, Vol. 17. pp. 416-435. Lazarev, A. V. and Kortjunov, S. S. (Editors) (1982) Spravochnik po torfy, Moskva Nedra, 1982. MacFarlane, I. C. (1969) Muskeg Engineering Handbook. University of Toronto Press. Magnusson, N. H., Lundqvist, G. and Regnell, G. (1963) Sveriges Geologi (Swedish Geology). Svenska Bokforlaget. Stockholm. Okruszko, H. (1969) Powstawanie mulrv i gleb mulowych. Roczniki gleboznawcze, Vol. 20, No. 1. Okruszko, H. (1984) Zaktualizowana klasyfikacja grtmtrw organicznych dla potrzeb budownictwa ziemnego. Mat. niepublikowane. yon Post, L. (1922) Upplysningar rOrande Sveriges Geologiska UndersOknings torvmarksrekognosering (Information on SGU peat inventory). Swedish Geological Survey. Ser. D. yon Post, L. (1924) Das genitische System der organogenen Bildungen Schwedens. Mrmoir. Nomenclat. et Classific. Sols 1924 pp. 287-304. Comit6 Intemat. Prdol. 4. Commiss. Nr 22. Radforth, N. W. (1969) Classification of muskeg. In: MacFarlane, I. C. (ed.) Muskeg engineering handbook. Canadian Building Series. University of Toronto Press.

31

Chapter2

Site Investigations U. Bergdahl, Swedish Geotechnical Institute

2.1

GENERAL

The results of a site investigation for an embankment project on organic soils must contain all the basic data for the design and construction of the embankment. Thus it should be possible to evaluate the following: 9 Local and total beating capacity of the soil. 9 Amount of settlement and time-settlement distribution. 9 Need for and design of reinforcement or improvement. 9 Design parameters for the embankment. 9 Choice of construction method and machines. 9 Costs for foundation and construction. 9 Influence on neighbouring area or structures. The field investigations must have such an extent that the whole area involved is covered regarding stability, settlements and the influence on neighbouring objects, Fig 2.1.

Investigation

area

_j

Fig. 2.1. Example of the area which it is necessary to investigate close to a creek.

Site Investigations

32

The site investigations are often performed in stages depending on the need for information in the different phases of planning, design, construction and maintenance of the embankment.

2.2

MAPPING,GENERAL SURVEY

At an early stage of a localization for a large embankment, it is important to have an overview of the soil* and groundwater conditions in the whole area involved. For this purpose, the aerial photo interpretation method may be recommended. Such an investigation is normally divided into three main parts (Viberg 1984): 9 inventory phase 9 aerial photo interpretation 9 field inspection

The inventory phase contains a compilation of existing knowledge from the areas of interest, e.g. maps and earlier soil investigations. From these sources, the aerial photo interpreter can obtain an overview of the geological and hydrological conditions of the area. The most common maps are topographical, geological, hydrological, economic and engineering geological maps. As most maps indicate the conditions close to the ground surface, previous borings in the area provide important additional information for understanding the soil profile at the site. Data from soil investigations can be obtained from different authorities, such as the national geological survey, road and railroad administrations, regional planning and building departments, geotechnical and geological consulting firms etc.

Aerialphoto interpretation is based on the fact that a number of indicators can be recognized in different areas of the site. These indicators may be topography, contours, vegetation, texture, colour or grey tone (black and white air photographs) and land use. Normally, the topography changes when the geological conditions change. Therefore, aerial photo interpretation should always be made in a threedimensional model using a stereoscope. During interpretation, areas of similar appearance as well as the limits between different areas are defined. The following areas can normally be identified: 9 rocky outcrops or rock covered by 0.5-1.0 m of soil. 9 tills or moraines 9 coarse sediments (sand, gravel) *) The word "soil" here means earth material

Mapping, General Survey

33

* fine sediments (silt, clay) . organic soils (peat, gyttja, mud). Areas of fine sediments can often be subdivided into areas of shallow and deep silt and clay layers. In areas with organic soils, the following indications can be recognized: 9 The ground surface is often fiat and nearly horizontal. The only exception is raised peat bogs, which are dome shaped. .

The vegetation is composed of plants requiring a great amount of water. Normal plants in the area often have poor growth.

9 The texture in peat areas is often tufty with closely spaced ditches. However, in areas of gyttja the soil often cracks, which makes the ditches unnecessary. 9 The colour or greyish tone is brown to black or dark to black in cultivated areas. However, dried gyttja is often light-grey. .

Organic soils are often found in the lowest part of an area, except for the peat bogs mentioned above. 9 Occasionally, peat bogs are partly excavated for fuel or soil improvement.

.

In populated regions, areas with organic soils are often cultivated also when they are occasionally flooded.

Good aerial photo interpretation requires overlapping aerial photos of high quality. Black and white films as well as colour films or infrared films can be used. The optimum scale of photographs for this purpose has been found to be 1:10.0001:15.000. In this scale, both an overview and a picture of ground detail can be obtained. After interpretation, it is necessary to make afield inspection on the site in order to check the composition and limits of the different areas indicated. For this purpose, lightweigt sounding and soil sampling equipment is used.

2.3

SOIL LAYER SEQUENCE

After the general survey of the area in question, it is important to obtain mformation on the soil layer sequence. Especially in organic soils, the characteristics may vary greatly from the ground surface to the denser bottom layers. To obtain a relevant profile of the soil layers, soil radar, some type of penetration testing and sampling can be used. Experience of soil radar is so far limited, except for investigation of peat layer thickness. Experience of penetration testing is good for estimation of the soil profile, but the possibilities for evaluation of the soil properties are

34

Site Investigations

very limited. Soil sampling is a necessity in organic soils, both for detailed soil identification and classification and also for determination of properties in laboratory investigations. 2.3.1

Soil

radar

Soil radar can be used to investigate the thickness of soil strata with quite different electrical resistivity or, more correctly, dielectric constant at shallow depths (< 10 m). By transmitting electromagnetic waves through the soil from the ground surface, different layers can be identified (Johansson 1987). Radar stands for radio detection and ranging. The georadar equipment consists of an antenna for frequencies between 80 and 1000 MHz, receiver and control unit. A low frequency antenna is used for depths greater than 5 m. The control unit contains an oscilloscope, amplifier, filter and recorder with plotter.

During the investigation, the antenna is moved over the ground surface with a speed of 1 to 10 kin/h, while the reflection of the electromagnetic waves is recorded on radargraphs, Fig. 2.2. Ground

surface

Road

~9 .~_~.2:'~: - :.:~ .- -- , ~ t ' , , . . . . 9 -::r." - ~ , "~ ~, ; ; ' .--- ---~';--...'---t-'~'_ ~ " - -=. . .9~ " ~ .=.,~-__--== = ~.~2-". -el. .. . . . .. .. .

--

.'::

- --

---

."

. . . . . .

-

9--..~e'Z-':.;...m,=: --* .x ~-,+w--

., . .9 ." " .

~. . . . . . .

,/'~-~,,

9

9 ,~: - ~ : -__ ~ : . . - ~ - - --~,---=, - ~ , ~ . .. . ~ . . -~ -~ 9"r ~ ~ . . . ..-~__ . . . ~: - . . L _,_ ~3 ._. . _ _ : .... 9 ...-

"- .~F..~---,,,. ~: ~ " .. r. -. -.- - - , - ~ . -

9" . . . ~ ~

~=~..~,.~-,~.4"-~.~,~..-,_--_-=-.:..'.."

....

. . . . . . . . . . .

Peat

"~

9

,-# "."

.,, o | * .... 9

;.~ .,..:.:...,r162 '

9 9 I.: t

'

99

" ' ' '

..

$'

Fig.

.. : , ~ , : 9 9-o "--''..~.., ' ....

~ "'"w'

2.2.

~ ,.~.

,~. ~

9 9. , .'." "'~"

"~"

Example

-. ..,

: ~,: :'. ~ ; . : "

, : ,,,,~." 9

Ti[[ ."

' :::

, * ; ~9 ' , ' - : ,

,.e :. " ." ~

~ ~ , -.

...' ." . . . . . '

.~.,~--,,.,;,, ,, " .,

of radargraph

:'. 9

9. .

from

peat

. .,,

:p~

~. .

~,--::. 9.,.: .4: -.,,.

.. ~

soil

9

, ;;.~=~y.~,~ ' :,~;-.,'.'.'-~"~"

:~'-~:~.-

9 9

' , ~9 - , . "9 . -,."..", , ,

~-,

k

:...# :.. -. ."...;. - . : ~ . ~ - - : : . . . .

. :..

9. . . . . . . .

. , . ,

.......

with

. . . . . . . . . .9. . . . . . . . . . . . .

9

" ..

.

'....r.;--.. ,.: : ..

. ,..

..

~ .

.

.

. , .{.'..,..

"':i

,....

underlying

.

. ,

". JI ~ .. , - t - ~ 9 . . I " -9 --. :. , " .

.-7 , . , ~ '.i l.e,..

-',&-.

t.

glacial

""

" .. :'." % " " :

.....

."

.--%

till.

For interpretation of the radargraphs, it is important to identify the different soil layers on the site by penetration testing and sampling, and to have thorough knowledge of the geology and groundwater conditions in the area. In the evaluation of the radargraphs, it is important to consider that the ground surface is shown as a horizontal line also when the terrain is hilly. Investigations (Bjelm et al 1982) indicate that it is possible to measure the thickness of organic soil layers, especially peat, and the depth to firm bottom layers or rock with the soil radar.

Soil Layer Sequence

35

Soil radar has been used for surveying peatlands. In Sweden, it has also been used to establish the thickness of existing roads over peat bogs. The method is quick and has shown good agreement with results obtained with sampling technique. It is essential to know the thickness of an existing road when widening and strengthening roads with low bearing capacity. Fig. 2.2 shows results from such measurements in Sweden. The radargrams show a clear boundary between the road material and the peat. The measurements by impulse radar were followed by soundings and sampling in the embankment. Very good agreement was obtained in this case between depths estimated from radar and actual depths. When widening existing roads, this method provides information on how the peat has reacted under a load, and it is thereby possible to gain experience from what could be considered as a very durable test embankment. In Finland and in Sweden, soil radar is also used to determine peat thickness and the topography of the mineral soils beneath the peat. In Finland, use is also made of a "radiowave moisture probe" to determine the dielectric constant, water content and dry matter content of peat. When comparing the measuring data with conventional methods in the laboratory, very good correlations have been obtained.

2.3.2

Penetration testing

Penetration testing or sounding is normally used to determine the thickness of different soil layers and the relative density or stiffness of the soil. A number of penetration testing methods are available. For the investigation of soft organic soils in general, static methods or light dynamic methods are preferred. A proposal for international Reference Test Procedures for four different methods has been produced by the ISSMFE Technical Committee on Penetration Testing of Soils and has been published by the Swedish Geotechnical Institute (Bergdahl 1988) among others. The four methods are: 9 Cone penetration test (CPT) 9 Standard penetration test (SPT) 9 Dynamic probing (DP) 9 Weight sounding test (WST)

The conepenetrometer was developed in the Netherlands at the beginning of the thirties. Originally, the cone penetrometer was mechanically operated, but today most penetrometers measure the resistance electrically. The penetrometer has a conical tip with a cross sectional area of 1000 mm 2, Fig. 2.3. Normally, a friction sleeve is mounted above the base of the cone with a shaft area of 15000 mm z. In addition, a pore pressure transducer may be mounted in the penetrometer tip with a

s filter located just above the cone base (a piezocone). Thus it is possible to measure simultaneously with penetration the cone resistance, local skin friction and generated pore pressure. Measurement of the generated pore pressures is of special importance in soft fine-grained soils, where high excess pore pressures develop during penetration. A new standard for the cone penetration test with special consideration to tests in very soft soils has been produced in Sweden (SGF 1993). Thrustmachine

I I

'1

I

I

-"----I Probe

Friction sleeve

I j

e--j

031

Fig. 2.3. Set-up for cone penetration using the piezocone principle.

The penetrometer is pushed into the soil with a constant rate of penetration of 20 mm/s using a thrust machine. The measured resistances are recorded continously. An example of resistance and pore pressure curves is shown in Fig. 2.7. The cone penetrometer is the most accurate sounding method and gives the best information on soil layer sequence and soil characteristics. When pore pressure measurement is added, it is possible to indicate also small variations in the soil. If the measured resistance is to be used to evaluate the soil characteristics, especially in soft soils, the measured cone and friction resistances must be corrected to total resistances according to Fig. 2.4. In estimating the soil layer sequence, the charts in Fig. 2.5 may be useful. However, the observations in organic soils are limited. According to Robertsson et al (1986), the corrected cone resistance qT, the friction ratio Rf and the pore pressure ratio Bq may be used to evaluate the type of soil penetrated.

37

Soil Layer Sequence

I

A N -Net area A T -Total area F c - M e a s u r e d cone resistance F T - Total cone resistance u

- Generated pore pressure

FT :F c +u(A T -A N)

Principle of correcting measured cone resistance to total resistance for the piezocone.

Fig. 2.4.

RU

(2.1)

~

qT U - Uo B

(2.2)

q _.. qT- (Yvo

where f qT = u = uo = ~vo =

unit skin friction resistance total cone resistance measured pore pressure static pore pressure total overburden pressure

The charts in Fig. 2.5 are to be used in parallel and may give contradictory indications of soil type. Another way of estimating soil type using similar parameters has been suggested in Sweden (Larsson 1992). In this method, the net cone resistance (qx- CYvo)and the total unit sleeve friction (fx) are normalized against the effective overburden pressure (~'vo) and the net cone resistance respectively. They are then plotted in the chart in Fig. 2.6 a. If the soil according to this chart is classified as "clay or organic soil", the second chart in Fig. 2.6 b is used to obtain a more detailed classification for this type of soil. In practice, it is very difficult to separate clays from organic soils on the basis of results from cone penetration tests alone.

Site Investigations

38

,,oo i oj/~.l, I

I

I

i

I

.,~ 8 z

10

0.1

I

I

11

Zone

1 -.-~....

3 _.....-1

'

0

~

.

3000-

Very stiff

2000"

Stiff

-

Normally Low plastic ;consolidated ~andlor highly clays or ~tsensitive clays slightly 't ov~-consotidat~ silty clays

I.-

. . . . .

Medium stiff . . . . .

-0.2

Fig. 2.6.

t t

__

1000-

0 I

"---.

.

.

.

,

. . . . . . . . . . . . . . .

.

......

0

.

0.2

i

. . . . .

1

--tI t

I -"-l-

- - = . .. .. .. .and . . . . . .- . . . . .Soft . . . . . . . . ~ _E -Gzttj.qs I - -. . . . . . 1' . . . . Very soft organtc clays,

~'*..,

. .

. . . . . .

0.4 0.6 Parameter Bq

I -,4-. . . . . . . . . i___ ~ ..... i ........ 0.8 1.0

1.2

a) Chart for evaluation of soil type according to Larsson (1991). b) Special classification chart for clay and organic soil (Larsson 1992).

40

Site Investigations

According to Landva et al (1986) the use of the static cone penetration test in fibrous peat is questionable owing to the fact that negative pore pressure is induced and that the peat in front of the cone is compressed before failure.

.10 0

CONE

RESISTANCE

RESSUREin

HPa

IN H P a 20

~,0

60

LOCAL FRICTION ft 0.0 0.2

IN H P a 0/,

FRICTION 0

R A T I O R f IN "/,. S tO

02

GL=S 09re.NAP

_ . , . u ,~-_

o. < z

- ~_--..=_~ - _ _.,_ _ _

.-= = == .. .=== . . . . . .

o I.cI ,., - I 0 n, uJ u. IL

---- "----

-10

-~__Z . . . .

Q: E

_z Q. 121

20

-20

-30

-30

Paezo cone nr : 1 0 / 1 - 3 5 S size of t i l t e r : h e i g t h 3 . 0 r a m . t h i c k n e s s 3.0ram l o c a t i o n of f i l t e r in t h e c y l i n d r i c a l e x t e n s i o n o f t h e cone m a t e r i a l of f i l t e r : s i n t e r e d s t a i n l e s s s t e e l " b e f - o r e ' t e r , t-T a f t e r test I capacity i Z. . . . . . . d,ng - ~ n e o~ ~ ! -o.o,o NPa I J o _ o _ - - E ~

i'~;~r~---~l~-;~i -~

DEL F T GEO TIcCliNICS WONINGEN TE HAASSLUIS CONE PENETRA T/ON TEST

Fig. 2.7.

1 -~.~6~ MPa lo.7 MP~__l

GO O2(RE)

date of t e s t

hme

: :

Remarks : fr,cr,on reducer : not apple,d abnormal interruptions : none observations : no s p e c i a l o b s e r v ~ t , o n s fall/excavation : old f i l l /.m t h , c k n e s s i n c l i n o m e t e r : no r e a d i n g s t a k e n c o n d i t i o n of p u s h r o d s / p e n e t r o m e t e r lip a f t e r t e s t : good waterievet in sounding h o l e : hole c o l l a p s e d near s u r f a c e backfilhnq - none

87-02-19 14-15 hrs

E x a m p l e o f t h e p r e s e n t a t i o n o f C P T t e s t results, a c c o r d i n g p r o p o s e d r e f e r e n c e test p r o c e d u r e , ( B e r g d a h l 1 9 8 9 ) ,

to t h e

The Standard Penetration Test (SPT) is the most common penetration testing method in the world. It is used to determine the bearing capacity of both piles and shallow foundations. As it is a heavy dynamic penetrometer, it is difficult to obtain minor variations in the characteristics of soft organic soils. However, disturbed samples are also obtained with this test procedure.

Soil Layer Sequence

41

The Standard Penetration Test is performed with a driving rig and a 63.5 kg hammer falling 760 mm onto the anvil fixed to the driving rods. The penetrometer itself contains a thick-walled sampler and the driving rods. The sampler has an inner diameter of 35 mm and a sample length of 457 mm, Fig. 2.8.

a]

Lifting head

b)

~ -~ ~

63.5kg hemmer----! I I .~,r

\\

V . m

o

,~ ~

,~ a

E C,4

L.

SPT sampler~~

Fig. 2.8. Arrangements (a) and cross section (b) of the Standard Penetrometer.

The SPT is normally performed in a casing. The sampler is driven into the soil by blows of the hammer. At each test level, there is first a seating drive of 0.15 m, followed by the actual test in the following 0.3 m of penetration. The number of blows required for 0.3 m of penetration is recorded as the penetration resistance, Fig. 2.9.

Site Investigations

42 7_

SPT +9.35 S< nd ;W

2 §

--z~

~735

P~ at

S~ind

i"~ "" z/i/i//2

Ground water level

68

"

S(Ind

i " ~ if/i//, "Ilia

50

!/i//ilL

~'/ii1111.

68

"lilllli,

Zlllll~

7i/.~

(50)cone

i

!

20

Number of blows outside range of scale

~lliillz

0

Fig. 2.9.

surface

79.09.05. Dote of ground water measurements

Sand

Gn vel

Elevation ground

GW§

~.o~ ....

79. Sand /

Bore hole number

Uncertain blowcount

n

/+0

60 bl/O.B0m

Example of results from a Standard Penetration Test.

The dynamic probing method includes a number of penetrometers with different weight of hammer, height of fall, cone and rod diameters. In the proposed reference test procedure, four types of dynamic probing methods are presented, Stefanoff et al (1988). As the driving energy for three of the penetrometers is rather high, they are normally not suitable for the indication of soil characteristics in soft organic soils. Only the Light Dynamic Probing method (DPL) can indicate small variations in the soil resistance. However, in deep and stiff organic soil layers, also the heavier versions of the dynamic probing method may be useful, especially for indication of depths to denser bottom layers. In this report, only the light dynamic probing method is presented. Information on the heavier method can be found in the report by Stefanoff et al (1988). The light dynamic penetrometer may be either manually or mechanically operated. It consists of a conical tip with an apex angle of 90 ~ and a cross sectional area of 10 cm z, Fig. 2.10. The rods should be o 22 mm hollow rods. The penetrometer is driven into the soil by blows from a 10 kg hammer falling freely from 0.5 m height. During the test, the number of blows per every 0.1 m of penetration (N10) is recorded. The results of such a penetration test are shown in Fig. 2.11.

43

Soil Layer Sequence

BLOWS PER 0.I m. N

OPL

10

20 ELEVATION.

12.3mNAP

E 1.0

6ROUNOWATER LEVEL:

I0.] mNAP

z 1.5 -'Jr).a. 2.0

PROJECT-NUHBER:

ZW "/

OAIE OF TESf

80-03-12

0.5

TYPEOFPENETROHETER: R DPL

O 2.5

Fig. 2.10. Scheme of cone for the light dynamic penetrometer.

:

NUtIBEROF TEST:

13

LOCATION.

XBOURG

Fig. 2.11. Example of presentation and rod from a DPL test, (Bergdahl 1989).

The weight sounding test was developed in Scandinavia in about 1915 as a tool for investigating the risk of landslides, mainly in soft soils. The weight penetrometer consists of a screw-shaped point, a number of weights (5, 10, 10, 25, 25 and 25 kg) a number of rods e 22 mm and a handle. Fig. 2.12. _~C~~L,~.~_~

I

=4s

-'~ To p

,.o?~ ~ - w ~ , g h t ,

,o'om

II

Pt

Weights 2,5 kg

,0 kg

E z 3 -xl-Q- S uJ c)

W S T 22

i ~ ~,~

,b(sp,,8o ~-i

4

(Rubber) 010 k~d

W S T 22

ht/GL2 m

0.80m J ~t

II [ i ~ R o d ~

22 rnrn

-~, ~ S c . r e w

point

V

Fig. 2.12. Details of the manually operated weight penetrometer, (Bergdahl 1989).

P! fb(.~oe00}

201,060

kN. ht/0.2

m

W E I G H T SOUNOING TEST. 22

mm

RODS

NUMBER OF HALFTURNS PER 0.2 m OF PENETRATION DRY CRUST OF ClAY PREBO~ING TO THIS LEVEL WITH 8Omm OIAM AUGER

DIAGRAM TO THE LEFT INOICATE LOADS APPLIEO IN kN

Fig. 2.13. Example of test results from a weight penetrometer test.

Site Investigations

44

The weight penetrometer is used as a static penetrometer in soft soils where the penetration resistance is less than 1 kN. If the penetrometer does not sink with this load, the penetrometer is rotated and the number of half turns for every 0.2 m of penetration is recorded. Nowadays, both petrol driven and hydraulic machines are used for the weight sounding test. In this equipment, the penetration resistance during the static phase is measured with a dynamometer. The penetrometer resistance from a weight sounding test is presented in the diagram shown in Fig. 2.13, with loads to the left and the number of half turns per 0.2 m of penetration to the right. The results of the weight sounding test are used to obtain a continuous soil resistance profile and indications of the layer sequence and the lateral extent of different soil layers. They are also used to determine the relative density of cohesionless soils and to estimate the relative strength of cohesive soils. Another type ofpenetrometer is the lime column penetrometer, which is used to check the homogeneity and strength increase of lime columns, c.f. Chapter 10. This penetrometer is a static mechanical penetrometer with a ~ 50 mm tip, which is designed to follow the centre hole in the column. About 450 mm from the tip the penetrometer is provided with 400 or 500 mm wide wings (depending on the column diameter) in order to cut the column into two parts, Fig. 2.14. 5001L.00

mm

.

10 o r 15 m m

Section A- A

L15 or 20 mm

36ram ~-

r~

50 mm

Fig. 2.14. Lime column penetrometer.

45

Soil Layer Sequence

The penetrometer is pushed into the column at a constant rate of penetration, 20 mm/s, and the total force is recorded. The shear strength of the column is taken as 1/ 10 to 1/11 of the unit penetration resistance. Fig. 2.15. The evaluation may be calibrated in undisturbed soil beside the lime columns.

Orga

-1-

da

o

Clay --

Clay

on/5, o, mn_l o lOO 200 Shear strength (kPa}

Fig. 2.15. Results of lime column penetrometer test in a column and, for comparison, also in the unstabilized clay.

The penetration test normally comprises 0.5-2.0 per cent of the total number of columns, cf Chapter 10. 2.3.3

Dilatometer

testing

Dilatometer testing is a relatively new in situ method of determining soil stratigraphy and soil properties. It was developed in Italy by Marchetti (1975) and has rapidly gained worldwide acceptance. The dilatometer consists of a spade- shaped instrument, which is pushed into the ground with a constant rate of penetration of 20 mm/s. The instrument, Fig. 2.16, is supplied with a flexible membrane on one side. At regular intervals, usually every 0.2 m, the penetration is stopped and the membrane is expanded and pushed out into the soil by the application of a regulated gas pressure on the inside of the membrane. The pressures required to overcome the

46

Site Investigations

earth pressure and make the membrane lift off from its base (P0) and the pressure required to expand the membrane 1.1 mm into the soil (P l) are recorded. The membrane is then deflated and the penetration is resumed. By using inflation tests at every 0.2 m depth, almost continuous curves of the variation ofthe pressures P0 and p~ are obtained, Fig. 2.17.

Fig. 2.16. The Marchetti dilatometer.

47

Soil Layer Sequence

Pressure

Pressure

(MPa}

Slgeffv

--UO 0

--PO

0.5

---PO

~Pl

0 0

1.0

0 ~

2

(MPa}

~Pl

0.5

1.O -

.

.

.

2.0

:1..5

1

.

I

~ i

, ,

i

9

2.5 1

-t - - - - ~ t

'

i i

,--,

~

k~/

t0

I

__)

,

t2

l,~

i4

16

16

18

18

20

20

a)

t

b)

Fig. 2.17. Results from dilatometer tests (a) soft slightly organic clay (b) sand with a clay layer.

The measured pressures are used together with the in situ stresses in terms of effective vertical stress (r'vo and pore pressure u 0 to calculate the parameters Material index

ID =

Horizontal stress index

KD = (P0 - u0)/Cr'vo

(2.4)

Dilatometer modulus

ED = 34"7(Pl - Po)

(2.5)

(Pl

-

P0)/(P0 - u0)

(2.3)

These parameters are then used for classification of the soil and estimation of soil properties. The internationally most widely used classification chart was presented by Marchetti and Crapps (1981) and was slightly modified by Schmertmann (1986), Fig. 2.18.

Site Investigations

48 2000.

EO . IOOO

iO(n

LINE A S C D

§

9

S..TY I ~ ' e " ~

lo| I 0) n

9

0.58S 0.621 0.6H? 0.696

S'ANDI

SILT

E.(~UATION OF THE LINES

1.737

2 013 2.289

2 ~64

CLAY

W o,;..:::3/ LId

,

p=(043

)~

> 0.5

(2.17)

WL

and w E - liquid limit of the soil For organic soils with a liquid limit > 200 %, the correction factor ~ is equal to 0.5. According to Landva (1980), the shear failure in peat was found to occur along a cylindrical body 7-10 mm larger than the diameter of the vane. This is one of the reasons why Landva does not recommend field vane tests in fibrous peat. However, for other types of organic soils the field vane test is considered to be a useful tool e.g. Schwab (1976), Wolski et al (1988) and Wolski et al (1989). Schwab pointed, out the necessity of making a greater reduction ofthe ~v values than that used at that time in Sweden for organic sulphide soils. However, the reduction rules have been changed since then. The selection of shear strength parameters for design is discussed in Chapter 4.2.

66

2.6

Site Investigations

MONITORING EQUIPMENT

Monitoring equipment is often used for embankments on organic soils both during normal construction and especially for test fills, Chapter 2.7. The reasons for this are mainly the large deformations and low bearing capacity of the soil, which make it necessary to use special foundation and improvement methods. An extensive description of different instruments for various purposes and constructions has been made by Hanna (1985). This book deals only with equipments that can be used for monitoring embankments on organic soils. Pore pressure measurements in the soil beneath an embankment can be made with the same equipment as described in Chapter 2.4.1. However, closed piezometers ought to be used. It is important to consider that large settlements in organic soils may cause increased pore pressures around the tip because of penetration of the piezometer, especially during filling, when the pore pressure may be critical for the stability of the embankment. To avoid this, the piezometer pipe can be provided with a casing to about 1 m above the filter. Also closed systems without pipes may be used or the piezometers may be installed with an inclination of, for example, 1:1. For the measurement of settlements, different kinds of benchmarks have been used, Fig. 2.33, both with steel plates on the ground surface and with screw tips that can be installed at certain depths. Benchmarks at greater depths ought to be provided with casings to protect the rods from the friction of the settling soil. In order to obtain continuous settlement distribution curves across the embankment and to prevent disturbance of the construction work, the hose settlement gauge can be used, Fig. 2.34. To perform these measurements, flexible tubes or hoses are placed in shallow ditches or under small sand fills on the ground surface across the embankment in the sections to be measured. If the tubes are extended outside the base of the embankment, also the heave of the ground surface outside the embankment can be measured. The measuring unit consists of two plastic hoses, one inside the other. The smaller tube contains air and an electric cable. The annular space between the two hoses is filled with a liquid (normally water). The lower ends of the hoses are connected to the measuring head containing the pressure transducer. This transducer measures the liquid pressure in relation to th e atmospheric pressure. The upper ends of the plastic hoses are connected to an open standpipe. Thus the difference in level between the measuring head, inserted in the flexible tube under the embankment, and the liquid level in the standpipe can be measured. The liquid level in the standpipe is in turn levelled in relation to a fixed point located outside the test area. An example of such measurements is shown. Fig. 2.35.

6?

Monitoring Equipment DEEP SETTLEMENT GAUGE {SCREW) -- [

[

SUPERFICIAL SETTLEMENT GAUGE (PLATE)

- Extension rod

Extension rod

J25 mm

25ram

lSOmm

50ram Protectinq tube

I Protect~nQ_ tube

I J

. .

.

I.

.

t 1t: / ....... ""

,.. i " .".. . /~. ' " " I ~3~176 j ~

Plate

.7

s~w

~,p

0 SOOmm

[

Fig. 2.33. Benchmarks for superficial and deep settlement measurements.

Read out unit

Atmospheric pressure Fluid level

--Cable

Atmospheric pressure

Fluid level regulator Pressure

Measuring head

Fig. 2.34. Principle of the measuring unit for the hose settlement gauge type SGI II.

Site Investigations

68 a)

Embankment

No.1

g_

! 4-

3.6

4-

3.9

L 2 ~. "

11 6 "..

4,

i

/..2

4-

III

J

. #

3.9

J-

3.6

4"

1/.

~

II

~

1.3

-0.2 0.0 u',

T ~r x

!

~

m

--4

~('~x _ . - - - - - - x

0.t. --4

._

rfrl

L_ !

"

,'--"z~

.I.

~.

~

0.8 m z

--4

..-

12 16

i

i

. .............................

. . . . . . . . . . . . . . . .

!

.. ~....-" L

z,5

I D (at e

i0907

{03 071 - - ~ ' - T ~ r * ' - q

---20 Ira)

11a4" ~ +,#~" i,,,~rF-T-nqar--T-~a~---1-~-,j-q-v~-T-T,]~r'r-q

io-ol - I:- to 4o, q-, -l-:-]----x IFig. 2.35. Example of settlement measurement obtained with the Swedis hose settlement gauges under a test embankment at Antoniny, Poland.

In order to measure the compression in different soil layers, the bellows hose settlement gauge or the magnetic screw settlement gauge may be used. This may be important in field tests where soil layers of different strength and deformation characteristics are found beneath the embankment. The bellows hose settlement gauge consists of a compressible, reinforced plastic tube 25 mm in diameter. The hose is cut into pieces corresponding to the distances between the observation points. The pieces are joined together again with special inner casings provided with a metal ring. This metal ring closes an electrical circuit when the sensor passes the ring, Fig. 2.36. The sensor is lowered in the pipe by a cable which at the same time is used as a measuring tape. In this way, the depth below the upper end of the pipe to the different metal rings can be measured from time to time.

69

Monitoring Equipment

"~-... Ir~dacafng casing with cap z m

~"--'--'-~

Measuring t a p e

One of ~ - . r a l r n e a ~ r i r ~ poin= =t pn~letermined c~=ntr~

Sounding rods fused during inr~dladonl

Lowes~ "point -

measuring ruled

35 m m i.d. ca~ng lu.~=d during in.~allaEon)

Steel 6p for driving and for hose at the bonom k=.vd

9fi~ng die

Fig. 2.36. Components of the bellows hose settlement gauge.

The installation process uses a system of inner rods and sometimes a casing. When the full depth is reached, the casing and inner rods are pulled out. One limitation of this method is that the compression of the plastic hose must be less than about 10 %. In a similar way, settlement measurements can be performed with the magnetic screw settlement gauge, which consists of a plastic pipe and a number of screw

70

Site Investigations

plates, Fig. 2.37. The screw plates are turned down to the desired depths around the plastic pipe. Magnetic tings are inserted in the plates close to the pipe. The level of the magnetic tings can be measured from time to time with a sensor on a measuring tape. Normally, the readings are taken in reference to the lowest plate where small or no settlements occur.

Settlement

indicator

_ -

r

,x.,u.t.~

Guiding plastic tube 9 ,x" t , , ~

.

,. . . . . .

Settling

screw plate m

~, ~

Reference plate Distance pipe

8 4

9 ,s

~

Screw tip

AAh

Fig. 2.37.

Magnetic screw

settlement gauge.

The main advantage of this type of equipment is that it can be used for larger compressions in the layers between the plates and that it is certain that the screw plates follow the settling soil layers. Fig.2.38 shows an example of magnetic screw settlement measurements where the two lower plates could not be measured atter 600 days due to buckling of the plastic pipe caused by the large settlements in the upper layers.

Monitoring Equipment

71

0.0 '~.- ~ = - ~ . - - . "- ~ ~ , _ . . . _ ~9.

0.2 0.4

.-. E E ~

~-.~.

O)

~ ~.,~,~

M-1

tube buckted

\ " ~

. ~

0.6

0.8

c

"~.

EMBANKMENT No 1.

H-2/__

1.0

l]J

4""

1.2

~'~

M-1

~- ~ ~

..

I 4"" Peat

"~,

"3.1

\.

co,c.,o,,

1.4

Sand 0

120

240

360

480

600

720

Time, (days)

840

Fig. 2.38. Results of settlement measurements with magnetic screw plates at different levels under the test embankment at Antoniny, Poland.

Considerable horizontal displacements also occur in the construction of embankments on organic soils due to shear deformations in the ground. The displacement of benchmarks at the ground surface can be measured with optical or electrooptical instruments or with a measuring tape using a fixed point outside the influenced area as reference. However, in order to measure the horizontal displacement below the ground surface, inclinometers have to be used. Various inclinometers are described by Hanna (1985) and here the inclinometer developed at the Swedish Geotechnical Institute is briefly described. The SGI inclinometer measures the horizontal displacements in a flexible plastic pipe which is installed in the ground at the point of interest. The pipe has an inner diameter of 42 mm without any tracks and is provided with a telescopic tip in order to avoid the influence of settling soil. The measuring unit consists of a cylinder with pairs of guide bosses, which are pressed against the plastic tube. Inside the cylinder, a pendulum is suspended in a plate-spring equipped with electrical resistance strain gauges. In this way, the deviation of the tube from the vertical axis can be measured accurately by means of a Wheatstone bridge.

72

Site Investigations

The inclination of the pipe, or change in inclination, is normally measured at every one or two metres of depth in two perpendicular directions, of which one is parallel to the direction of the expected main movements. To be able to measure in a certain direction from time to time, the inclinometer is inserted in the plastic tube with a series of torsionally rigid rods. The horizontal direction of the inclination measured is determined on a horizontal scale at the top of the pipe. Changes in the inclination in a certain direction correspond to the angular strain in the soil. The position of the pipe in relation to the tip can be calculated by integrating the inclination from the tip to the level considered. Examples of such inclinometer measurements are shown in Fig. 2.39.

ET

I-~

--

~

t'ttr i

2

PEAT

3

,,~

~'

2

2

3

3

II ilI:

~

~

" ~ ; " "7"

/r162

II

/S/f7 t'

/;12

I I ill i i-

\\ ~\ 11 ~

~~

. . . . . . . . . . . . . . . . . cm

cm

"

3

\ \ \\\\ \ I 1

Stoge

.

.

.

.

.

i; ~176

1;~o~o~o---o "o,~

,----_

i ~

smelt toad steps~ short duration

iI.

Fig. 3.13.

doubted steps standard duration

Compression curve illustrating the loading procedure suggested by Bjerrum (1972).

Any change in the loading procedure, however, entails a change in the stressstrain curve (Fig. 3.14).Thus the use of the Bjerrum procedure brings a curve which, if the Casagrande evaluation is used, gives too high a preconsolidation pressure, requiring correction, and a modulus that is lower than usual. If smaller load steps are used than in the standard procedure, it often becomes difficult to evaluate the end of primary consolidation and the rate of secondary consolidation. Taking these aspects into account, it is often better to use the normal loading procedure and more tests rather than more load steps m one test. The permeability of the soil is often estimated from the shape of the time- settlement curves for

103

Oedometer Tests

Vertical stress, 6"v (tog scale) I

v

~

E

121

\

~__~-r

I--

I/1

\

_=..,

0 u

-E >

Fig. 3.14.

\\ \ \\\ steps

A-doubted toad short duration (not futt consolidation ) B- Bjerrum method C- standard method D- smart toad steps tong duration

\ "\~\\

Compression curves obtained in oedometer tests IL performed with different loading procedures.

the individual load steps. This estimation is very crude. If the permeability is to be evaluated from this type of test, it is better to perform a number of falling - head permeability tests at the end of selected load steps and before the following load increments are applied (Tavenas et al., 1983 ). The tests are performed at such deformations that the relation between permeability and deformation can be evaluated for the range of deformations of interest. In oedometer tests, using fixed or floating rings, friction occurs between the sample and the ring. It is very important to minimize this friction, which is done by using very smooth rings and grease, or, in the case of fibrous peat, by using a compressiometer (Caflsten 1988). In the compressiometer (Fig. 3.15), the confining ring is replaced by a rubber membrane and lateral deformations are prevented by a number of thin rings spaced so that no vertical load can be transferred between the rings.

104

Laboratory Investigations P CI am .

.

.

~ .

Rubbermembrane

.....

Coarsefilter

.

. . . . .

Sample

~

Confining rings_

-

I:-

"

'

"

.-:-!

Fig. 3.15. The compressiometer apparatus.

(b) Deformation and consolidation parameters: The results from incrementally loaded oedometer tests are presented as the relation between void ratio e or strain ~ and effective vertical stress 6". The results are often first plotted on linear scales to enable a closer study of the shape of the curve. The oedometer modulus M, the permeability k, the coefficient of permeability change 13k, and coefficient of consolidation %, may then be evaluated in a similar way as in the CRS-tests (see Chapter 3.4.2). The results are then plotted with the effective vertical stress on a logarithmic scale to enable evaluation of the preconsolidation pressure G'p. From this plot the compression compression index Coot C e ,where Coe = Co/(l+e0),(Fig. 3.16 a), can also be evaluated. If the test is performed with a cycle of unloading and reloading, the swelling characteristics can be determined. These may be expressed as moduli or as swelling and recompression indices, The swelling and recompression indices are evaluated as shown in Fig. 3.16 b. The deformation process in organic soils often strongly deviates from the simple model used in Terzaghi's consolidation equation, which is the basis for the Casagrande and Taylor evaluations of primary consolidation and coefficient of consolidation.

105

Oedometer Tests a)

b) Vertio3[ pressure 6 v (log SCQ[e}

Vertical pressure 6v (log scale)

CO

(aJ

d

E] I..

I...

E~

U .m -i.., I_

A e or E 2

I._ (b >

> /

I_. 0

._d E~ L.

~" E l 0

E] .u_

O

z~e Cc- a log 6'v

o" E] I..

o~ -i..,

A

>

sE

E2

CcE : log 2

Ae r

Cr = A - l o g ~'"v

Fig. 3.16.

or

aEs

--'----

~Iog6"v

txEr CrE = ---------

~[ogB'v

Evaluation of compression index Cc, swelling index C s and rr index Cr"

Results of tests performed by several authors (Ozden and Wilson, 1970; Berry and Poskitt, 1972; Edil and Dhowian, 1979; Szymanski et al., 1983) indicate that in organic soils (a mixture of organic fibres, colloidal particles, water and gas bubbles) the deformation process involves elastic strains of gas, elasto- plastic primary strains of the soil and viscoplastic creep of the soil skeleton. These processes take place simultaneously, which significantly influences the stress-strain-time characteristics. Evaluating primary consolidation, coefficient of consolidation and permeability from the shape of the time-settlement curves may therefore be very misleading. In calculations of settlement, the Terzaghi equation is still normally used with calculation procedures that allow for large deformations and changing parameters and boundary conditions. In these calculations it is important that the degree of non-saturation be accounted for and that the changing permeability be accurately determined. The creep effects, which also can be accounted for are defined by the coefficient of secondary compression C a and can be determined from the time-settlement curves in the incremental load tests. Other rheological models and calculation methods for the process of on consolidation in organic soils have been suggested (Gibson and Lo 1961, Barden 1968, Berry and Poskitt 1972). Relevant parameters used in these models are evaluated from incremental oedometer tests.

Laboratory Investigations

106

The coefficient of secondary consolidation is evaluated for each load step. Fig. 3.17. This parameter is a function of stress history and deformation and the variation within the particular range of stresses and deformations to be applied in the field should be determined. r-

.,.--.

Cl

-,&

t 100

Time (tog scare) v

I,..

~ .,I,-, I..

primary consotidation described by c v

(P > I,-

0 0 ,,.I-, I,.,_

secondary compression .4~.~escribed by Co,,.

13 0

~

Fig. 3.17. Consolidation curve obtained in incremental loading oedometer tests.

The coefficient of secondary compression may be expressed as Ca= de/dlog t or as C e= de/dlog t, (C~ =C~e (1 +e0)). In long-term tests there is sometimes a downwards curvature of the log time-settlement curve m the secondary compression phase. This phenomenon is someteimes called tertiary compression. There is no field evidence of tertiary compression and it may therefore be considered a laboratory effect which need not be included in the test evaluation. The coefficient of consolidation cv is a function of modulus M and permeability k; cv = M. k / g" Pw and should preferably be calculated from directly determined permeabilities. Both M and k vary with stress and strain and consequently cv is a variable. Methods of estimating the coefficient of consolidation and indirectly also the permeability have been presented by Casagrande and Taylor. The results should, as mentioned above, be treated with great caution for organic soils.

107

Oedometer Tests

Continuous loading oedometer tests

3.4.2.

(a) Testing procedure: The incremental loading tests are time-consuming (one to two weeks). The need for time-saving tests and more accurate descriptions of the consolidation parameters and their variation has led to the development of oedometer tests with continuous loading CL, which are performed as: constant rate of strain CRS (Smith and Wahls, 1969), constant rate of loading CRL (Aboshi et al., 1970)~ constant gradient CG (Lowe et al., 1969) or continuous consolidation test CC (Tokheim and Janbu, 1980). These tests give continuous stress- strain relations and the continuous variation of the consolidation parameters. All of these tests give the consolidation parameters (modulus and permeability) and hence also the coefficient of consolidation, but the influence of secondary consolidation can not be separated or evaluated. As the loading procedures are different for the different types of tests, the strain rates become different and therefore also the stress-strain relations obtained in the various types of tests become different. (Fig. 3.18 and 3.19).

0

"-" 10

Effective vertical stress, 6 v (kPa} 100 200 300 !

i

i

400 ,

7.,~ Range for CRS test with tow rote of strain

E -p ~ 20. a (...) I,,.

Ub/6' v

30

Fig. 3.18.

Correlation between CC test with different relations UDI6'v and CRS test (Larsson and S~illfors 1985).

Laboratory Investigations

108 t

Effective vertical stress. 6 v

=10n-2 ~= 10n I~=10n+2 CO

d 0 L.

0 t.) .m .,4-., t_ Q)

II

CRS

- - - - - CGT CC Stress-strain relation for a constant rate of strain Fig. 3.19.

Schematic relation between a CRS test at low rate of strain, a CG test with a low gradient and a CC test with low relation UblG' v (Larsson and S~Ulfors 1985).

The only type of continuous test that to a large extent has been applied to organic soils is the constant rate of strain test. For this test it has been shown that the S~illfors evaluation of the preconsolidation pressure is applicable also for organic soils and that the stress-strain relations evaluated according to Larsson (1981) are directly compatible with the results from standard incremental tests, (Larsson and S~illfors 1985) Fig. 3.20. One condition for this type of test is that the pore pressure should not at any stage of the test exceed 15 % of the applied load. Therefore, also this test may require several days of testing time in highly compressible and low permeable organic soils, (Larsson 1990). The CRS-test is a standard test in Sweden and its use is spreading in other countries. No estimation of the secondary consolidation can be made from this type of test. Supplementary incremental load tests therefore have to be performed unless the empirical relations for C~ are considered sufficient.

Oedometer Tests

109

0

Effective vertical stress,

0

100 l

200

300 I

I

6"v (kPa) L.O0 I

Oo~\ 10.

Range for 24hour readings in standard incremental tests

20-

d ~

30-

0 u

1= > 40-

50-

8m Bisckebol 4 m

B~sckebol

. . . .

Vallda

Fig. 3.20.

Incremental loading 24 hours reading v . o

Correlation between standard incremental test results and results from CRS test (Larsson and S~illfors 1985).

(b) Deformation and consolidation parameters: The oedometer curves from CRS-tests are obtained as continuous relations between effective stress and strain, between modulus and effective stress and between permeability and strain. The average effective stress in the specimen is calculated, on the assumption of a parabolic distribution of pore pressure, according to the equation:

6"v = P / A - 2Ub/3 where P = applied vertical force A = cross-sectional area of specimen u b = pore pressure at undrained bottom

(3.17)

Laboratory Investigations

110

Taking into consideration the additional requirement for this test, i.e. that the pore pressure does not exceed 15 % of the applied vertical stress, the possible error resulting from Eqn. 3.17 is limited. The permeability is calculated from: g 9Pw 9H

de

2 ub

dt

kv =

(3.18)

and the coefficient of consolidation

kvM cv =

(3.19) g . [3w

where g Pw H de/dt M

= = = =

gravity (9.81 m/s 2) density of water height of specimen rate of vertical strain oedometer modulus.

From the continuous stress-strain curve obtained from a CRS-test (Fig 3.21) the preconsolidation pressure Cr'p is first evaluated according to S/fllfors (1975). The curve for stresses higher than cy p is then moved a distance c to the left in the diagram to compensate for rate effects and to become compatible with results from standard incremental tests. To describe the variation in modulus, the curve is divided into three parts. For stresses up to the preconsolidation pressure the modulus is constant, M o. For the part of the curve above the preconsolidation pressure where the stressstrain curve is a straight line (G'p - 6"L ) the modulus has another constant value, M L and for stresses above or"L the modulus increases linearly with the modulus number M'.

The permeability is evaluated by simplifying the log permeability - e curve to a straight line (Fig. 3.21 c). The initial permeability k i is evaluated at the intersection of the straight line and the horizontal line e = 0. The decrease in permeability with compression is expressed by the parameter 13k = -Alog(kv)/Ae. The coefficient of consolidation c v is calculated from Eqn. 3.19 and is thereby corrected by the correction of M (Fig. 3.2 ld).

Oedometer Tests

lll

PermeQbility, k v (m/s)

Effective vertica[ stress,~ (kPa)

I0-I0i~

1

201 p 80 120 '1 . . . . .

~'

0

160

I06, I04

,

gkv

N 10

.

A

CLJ

c

tJ c"

10,

I

~ 2o -~

0 L.

~E

m 15. 121 u

J'k =" A i ~ k v

I

30

L-

> 20-

40

25

2000, 1600

_

,' I/1 c,4

C

.o

%

E

/z

u

>

z 1200 ID

i 400

ML

Fig. 3.21.

. . . .

|

i

/]

0

0 u

A~v

0

/5~/ M'_aM

a csL

i

1

~v l

.

l

o Q;

0

CV=

I07

Mo kv gJ~w

168 169 0

,

Ii ' 80 ' 120

Effective vertical stress, 6v (kPa)

Interpretation of the CRS test results (Larsson and S~illfors 1985). Dashed lines indicate curves corrected for strain rate effects.

Laboratory Investigations

112

3.5

DETERMINATION OF DEFORMATION PARAMETERS BY TRIAXlAL TEST

3.5.1

Testing procedure

For calculations of two-or three-dimensional deformations, the parameters used in the applied soil models have to be determined. For calculations according to the theory of elasticity, Young's modulus E and Poisson's ratio v have to be determined. For elasto-plastic calculations, the yield criteria have to be established and for viscoelastic-plastic calculations, the creep parameters have to be determined. All these parameters are determined in triaxial tests designed to simulate in situ conditions and to measure the relevant parameters for the model to be used in the calculations. Depending on which case, drained or undrained, the calculations aim at, the tests should be carried out as drained or undrained tests. The difference between undrained and drained parameters should be distinguished. Undrained parameters are denoted here as E , v~ and drained parameters as E', v'. The symbols E, v are used in a general sense. Elastic parameters E, v, are evaluated from the results obtained in triaxial tests. The testing method is similar to that used to determine soil shear strength (Chapter 3.6).There are four important factors which have a significant influence on the accuracy of the evaluated E and v values 9 measurement of the vertical strain e~ 9 measurement of the horizontal strain e3 9 reconsolidation of the tested sample (to simulate "in situ" conditions) 9 testing procedure The first and second factors depend mostly on the type of the triaxial apparatus and the method of measurement. There are many methods for determination of the horizontal strain ~3" The methods can be divided into two main types: direct measurement of change in sample diameter reside a triaxial cell (Fig. 3.22; Felix, 1982; Paute, 1983) or measurements of change in distance between the cell wall or some other fixed point on the specimen perimeter (Fig. 3.23; Baranski and Wolski, 1985). Both types of method make it possible to determine volumetric and shear strains as well as dilatancy and creep characteristics in unsaturated soils and in soils saturated by pore water with gas bubbles, such as many organic soils. The compression characteristics of the soil can also be studied during the consolidation phases prior to tests performed to obtain the shear strength. This consolidation is usually performed in small steps with anisotropic stresses adjusted so that drained K conditions (i.e. no lateral strains) prevail throughout the consolidation process. These conditions are similar to those in an oedometer test.

Triaxial Test

ll3

~ ~1 ...... i!

I

7-

Gauges for lateral displacements

"l ~____ J

ik._

r'-

Fig.3.22.

Measurement of the lateral displacements of the sample inside the triaxial cell.

RESULT ON SCREEN

TRIAXIAL CELL

Atternate positions ~of ultrcL~nic head . . . . . ~ ~

-1" T

v//~'/)///~ /

V///////~

\\\

\\\ W--.c~*-~

iiiiiil

:--:---:'

/T~,ox,o~ k~o,, ceil

Fig. 3.23.

specimen

A

1///I 1~ 1

j: j. --

''o~k

on the screen

INTERPRETATION Disturbance area /~. . . . . . rlnl~lat snape

up spot ,.//of specimenEr

, VIII ~/7-~,//~

! I

A'

I!

~oo,ou~o,k

specimen

s ~0

The ultrasonic method for measuring the lateral displacements of the sample in the triaxial cell (Baranski and Wolski, 1988)

Laboratory Investigations

114

The values of the strength and deformation parameters obtained in a triaxial test depend to a large extent on the mode of loading during the test. In order to simulate the field conditions the tests can be performed as four main types: * active tests with increasing vertical stress (rv and constant horizontal stress 6a 9 active tests with constant ~v and decreasing crh . passive tests with decreasing 6v and constant 6h * passive tests with constant cyv and increasing ~h More complicated tests with simultaneous variation of both cy~ and ~h are also possible. The mode of loading is selected so that it closely corresponds to the loading conditions in the field in the particular zone of subsoil beneath the embankment (Fig. 3.24). a) loading

b) unloading zxq > 0

r

///,&,.7/,2A::~///,

zxq < 0

///,&-,7"//4,.'9"///

~

: ',F, Passwe zone

Active zone

7 / / . / / / / / / / / / / / / / / / / / /

/////

6v = const

Active zone

//////////////////////I/II/,,

///Z.,"7//,~//,

la6"v

a6"v < 0

Passive zone

Fig. 3.24. Stress conditions beneath embankment. (a) Loading, (b) Unloading. 3.5.2. Y o u n g ' s m o d u l u s a n d P o i s s o n ' s ratio The results from the triaxial tests are usually presented as the relation between deviatoric stress and vertical strain (Fig. 3.25). Young's modulus of elasticity E is expressed as: A E =

where (CYl-Cr3) = deviatoric stress E1 = vertical strain.

(3.20)

115

Triaxial Test

bl q = (61-63)

q =(61-6 3)

I

/!

/ / i ~(61-6 3)

//

.

.

.

.

.

aE I

J

li

il ~q

a(61-63) E t = ~

[

aE I

--

E. = a(61-631 -= - aE 1

Strain E 1

E1

c) q=(61 -63)

q =61-6 3

':/25 +`

tangent at

/ ~

'

"

n, AB

AF/' /

//

///

Ci~secant OB I _ ,cc~-G3 llaDI

i E~:'"~I{OO,

!

D

Fig. 3.25.

E1

S'" corrected origin

E1

v

Determination of Young's modulus from non-linear stress-strain relations (Head, 1986)

Depending on the method of calculation, three different types of elastic modulus can be evaluated: 9 initial tangent modulus E i - the slope of the tangent at the initial point (the origin) of the stress-strain curve, 9 secant modulus E - the slope of the line connecting two specified points in the curve, the lower of which could be the origin, 9 tangent modulus E t- the slope of the line drawn as a tangent to the stress-strain curve at selected stress level or strain. As has been mentioned above, precise strain measurements during the initial state of stresses are very important in modulus estimation. If the stress-strain characteristics in the initial phase of the test are incorrect, as is shown in Fig. 3.25, that part of the curve should be omitted.

Laboratory Investigations

116

The initial modulus E i has been found to vary with the confining pressure (r3, and this dependence may be expressed according to Janbu (1963) as" E i = m P a ( 0 " 3 / P a )n

(3.21)

where pa = the atmospheric pressure m and n = constants determined in the tests. A vertical compressive stress (Yl applied to an elastic material produces not only a vertical compressive strain el, but also a lateral expansion. The horizontal strains in two perpendicular directions are denoted by e 2 and e 3 and for an isotropic material e 2 = e 3. These transverse strains are related to the longitudinal strain by the equation:

g 2 = g 3 -- - VE 1

(3.22)

in which v is Poisson's ratio for the material. The minus sign indicates lateral expansion corresponding to longitudinal compression. Over a particular stress range, Poisson's ratio is calculated from measured stratus by the equation: AE 3

v -

~

(3.23) Ae~

Poisson's ratio v is numerically the slope of this curve and the value is calculated from a tangent or a secant using the same criteria as for determining Young's modulus (Fig. 3.26). For undrained fully saturated conditions, in which the volume remains constant, Poisson's ratio v has a value of 0.5.

3.5.3.

Bulk m o d u l u s and s h e a r m o d u l u s

Bulk modulus K and shear modulus G are often determined on the basis of E and V as

K=

(3.24) 3(1-2v)

and

117

Triaxial Test Stress 61

/

or

stress level

// ] //

//

I

I

46"1 Es

I

Lateral strain E3

Fig. 3.26.

I

-

a6"1 AE I

Axiat strain

E1

E3 E1

Typical graphical plots of axial stress and lateral strain against axial strain, showing determination of secant modulus and Poisson's ratio (Head, 1986)

E G =

(3.25)

2(l+v) Both moduli can also be determined from laboratory tests: the modulus K on the basis of isotropic compressive stress - volumetric strain characteristic and G from the deviatoric stress - deviatoric strain curve.

3.5.4

Yield envelope and creep characteristics

To predict deformations with an elasto-plastic soil model, the stress ranges for small strains (elastic) and large strains (plastic) should be distinguished. For this purpose, a yield envelope (locus of stress states in q-p' stress space which cause large plastic strain responses to loading) can be determined. Laboratory tests performed on organic soils (Lechowicz and Szymanski, 1988) show that the shape of the yield envelope is anisotropic, as for other natural soft soils. (Fig. 3.27). The yield envelope can be determined by a number of consolidated drained triaxial tests with varying stress paths.

Laboratory Investigations

118

150 -6 "-"

100

"/~\~ r I \

1,.0 ~'-

II

o0V Fig. 3.27.

, ,J/

50

~

modet

l

, 100

.Modified Cam-CLay

150

',tl

:

200

o

.

p'=11316~.26~1 (kPa)

Yield envelope obtained for calcareous soil from the Antoniny site (Lechowicz and Szymanski, 1988).

In organic soils and other soft soils, the deformations are of an elastic-viscoplastic character because of creep effects. To estimate parameters describing the visco-plastic behaviour it is necessary to use suitable laboratory tests. One of the tests employed to describe this behaviour is the triaxial creep test, which can be performed as drained or undrained with different modes of loading. These tests are performed in standard triaxial cells (Singh and Mitchell, 1968; Larsson, 1977). The test results obtained in creep tests depend on the applied stress conditions. Tests performed on organic soils by Lechowicz and Szymanski (1988) show a wide range of deformation rates depending on stress conditions and time after load application (Fig. 3.28). When the deviatoric stress is smaller than the stress that ultimately leads to creep failure the deformation rate regularly decreases, but when the deviator stress is higher than this value, the rate of deformation first decreases and then rapidly increases when failure is approached. The test results show that unless failure is approached, the relation between deformation rate and time for a given constant shear stress in log-log scales is linear. The relation between deformation rate and stress for a given time is described by an exponential function. This fact makes it necessary to use stresses in the creep tests which closely simulate the stress paths occuring in the subsoil.

Triaxial Test

119

a) I

I

I

i

Peat

60

60 0 13.. v

i

Calcareous soil

I

e21 200

o10 8~ 9

Lo 40i

-

e19

40

e7

LO

ii

o18 16217

e6 e5

1:31-

20-

14~15 o12 "13 o11

20

20 ~34 1o

_

1

0

I

I

20

!

I

-

/..0

0

p" 1/3 (6i +263) (kPa)

I

1

20

l

4O

-

I

60

p': I13 (6i+ 26"3} (kPa)

b) 1

t

i

I

9 ~ 101

9~

1

_

10-2

10-3

lCi3

/

~6 51

-,-.,_ l d 6

_

>

i0-I

Fig. 3.28.

!

16 ~ ~3

10-5

io .7 _

!

Calcareous soil

19

10-1

10-2

"~ 16/._ E 0 o

I

Peat

I 1

I 10

I 10 2

J , 10 3 104 105 Time (rain)

~6 e_

11~7 104

I

10

l

102

I

103 104 105 Time (rain)

Creep test results obtained for organic soils from the Antoniny site. Relation between strain rate and time. Higher numbers indicate higher shear stresses. (Lechowicz and Szymanski, 1989).

Laboratory Investigations

120

3.6

DETERMINATION OF SHEAR STRENGTH

3.6.1

Swedish fall-cone test

A simple method to determine undrained shear strength is the Swedish fall-cone test (Fig. 3.2). The fall-cone apparatus is equipped with four different cones, the mass and apex angle of which are 400g/30 ~ 100g/30 ~ 60g/60 ~ and 10g/60 ~ The 60g/60 ~ cone can also be used to determine the fall-cone liquid limit (Chapter 3.2.2). Based on the correlations with vane shear tests by Hansbo (1957), the undrained shear strength l:fomeasured by fall-cone depends on the depth of cone penetration i, the cone apex angle and the cone mass m as follows: ~fo = K m g/i 2

(3.26)

where K= constant, dependent on the apex angle of the cone and to some extent on the type of soil g = acceleration of gravity. Based on tests on remoulded clay, Hansbo (1957) suggested a constant K=0.25 for 60 ~ cones and 0.8 for 30 ~ cones. For undisturbed samples, the value of K for the 30 ~ cone was found to be 1.0. Ranges for this value between 0.8 and 1.2 have been suggested by Gameau and Lebihan (1977) and Wood (1981). A comparison of undrained shear strength values measured by the laboratory vane shear test and the fall-cone test presented by Wasti and Bezirci (1986) indicates that for 30 ~ cone the value of the constant K is equal to 1. Determination of undrained shear strength using Swedish fall-cone test can be performed together with the evaluation of other index properties. However, an application of this test for peat and especially fibrous peat is questionable. According to Swedish practice, the undrained shear strength values zfo obtained from fall-cone tests should be corrected in the same way as vane shear strength values.

3.6.2

Laboratory vane shear test

Another simple method used for determining the undrained shear strength of organic soils is the laboratory vane shear test. The vanes commonly used in laboratory tests have a height/diameter ratio of 2, as in most field vanes, or 1 (Hanrahan, 1954).The undrained shear strength value zf~ is typically calculated assuming full strength mobilization along a circular cylinder circumscribing the vane as

121

Determination of Shear Strength

....

L a b o r a t o r y v a n e s h e a r test 9F a l l - c o n e

10.

test

J

x ~9 05"IJ

.J

0 v

,

i

,

,

,

) , j

1

Fig. 3.29.

q~fv----2

10

!

,

,

u

vv

i

,

...----

100 r f c ' (kPa)

Comparison of shear strength values from laboratory vane and fallcone tests (Wasti and Bezirci, 1986).

Mma x / ( r t D

e (H + D/3))

(3.27)

where Mmax = applied peak torque Dv = vane diameter Hv = vane height The above equation includes the assumption that the soil has isotropic strength properties and that the distribution of shear stress around the sheared cylinder is uniform. The laboratory vane test is usually carried out on a sample remaining in the sampling tube. It requires a minimum of sample preparation, but the testing is performed in uncontrolled stress and drainage conditions. Because of this, the vane shear strength values can be considered as index values only. The results from laboratory vane shear tests should be corrected, as fall-cone tests and field vane tests. By performing laboratory vane shear tests m a triaxial cell, the vane shear test can be conducted in axi-symmetrically loaded samples under known stress conditions with controlled drainage conditions at the boundaries of the tested sample (Law, 1979). The laboratory vane apparatus can also be combined with the Rowe oedometer to examine the change in undrained shear strength during one-dimensional consolidation.

Laboratory Investigations

122

Also the results of laboratory vane shear tests are highly questionable in peat and especially in fibrous peat. The results of vane shear tests depend on the test procedure and it is important that the standard procedure be followed. Therefore the test should start directly after insertion of the vane and the torque should be applied at such a rate that failure occurs within 1 to 3 minutes after the start of the test. If the test is used in conjunction with triaxial apparatus or a Rowe oedometer, this procedure may not be feasible. In this case, correction factors have to be produced for the particular procedure or the results may be used to illustrate relative changes in strength only.

3.6.3

Direct simple shear test

The direct simple shear test has been extensively used for over 50 years to determine shear strength of soft soils. It has been criticized from time to time for nonuniformity of stresses and boundary conditions that hinder a closer evaluation in terms of stress-paths. The test is still very commonly used partly because of the relatively simple testing procedure, but mainly because of the well documented experience. The evaluated undrained shear strengths from this type of test are directly applicable to stability calculations, e.g. Ladd 1981. The results of direct simple shear tests are usually applied to the central part of a slip surface where it is more or less horizontal. As the results of direct simple shear tests are close to the average of active and passive triaxial tests the results are sometimes used for the entire slip surface. Furthermore, the results of direct simple shear tests on undisturbed samples reconsolidated to in situ conditions are normally very similar to the results of corrected field vane shear tests. In direct simple shear tests, a soil specimen is laterally confined by a rubber membrane and a series of thin and evenly spaced tings (SGI device) (Kjellman, 1951; Larsson, 1977) or by a cylindrical wire-reinforced rubber membrane (NGI device) (Bjerrum and Landva, 1966) in such a way that the diameter of the specimen is kept constant. The specimen is consolidated one-dimensionally for a vertical load. The sample is then sheared by moving the top cap horizontally at a constant rate or by incremental horizontal loads while the bottom pedestal is fixed (Fig. 3.30). The pedestal and the top cap may be supplied with short pins which protrude into the specimen to prevent sliding at these interfaces during shear. Shearing of the soil specimen can be conducted in undrained conditions with constant volume or in drained conditions. It is not possible to perform truly undrained tests with pore pressure measurements in the normal direct simple shear device. A comparison between truly undrained direct simple shear tests and conventional constant volume tests presented

123

Determination of Shear Strength

by Dyvik et al. (1987) shows that the test results obtained from these two methods are very similar.

a)

bl

Rejnmf~:C:de

6hc :t~6'vc

i n '

"'S

Filter

=

=

=

1'h

lttttltttI

sliding

Fig. 3.30.

0.4 U

0.5

Peat 6"b =15 kPa

.~'vc: 20kPa

OLo .4.-" O

c_ 9

0 ~-

f

C"

Nb-,

0.3

N Ul 0. 2

0

c"

z m O.1

0 N

X

80kPa

/S~~'~~x 60kPa

//#//"

.Ne-~ 0.2

80 kPo

Z m 0.1

// r I

I

!

Shear strain, ~ 1%) Fig. 3.31.

/ Z c-2_0kP_o._ _

~

c

\60kPo

//

O E ~

~=20kPa

-e. ~ 0.z.

g ~ 0.3

/

~n ul

~P

Calcareoussoil

)

l

2

A

z,

l

l

10 12 Shear strain, ~ (%) 6

Results of undrained direct simple shear tests on normally consolidated organic soils from the Antoniny site,

8

Laboratory Investigations

124

Effective shear strength parameters obtained from drained direct simple shear tests carried out on organic soils from the Antoniny site are shown in Fig. 3.32. To estimate the corrected angle of internal friction (1)' the test results were corrected for volumetric changes according to Bishop (1954). The values of obtained (1)' from drained direct simple shear tests are somewhat lower than the corresponding values obtained in triaxial tests. Like all other tests, the results of direct simple shear tests are rate dependent. According to the standard procedure used in Sweden, the specimens are consolidated for 24 hours and the rate of shear in undrained test is about 0.6 % of the specimen height per hour. This rate can normally be used also for drained tests when the specimen height is 10 - 20 mm, but especially for highly compressible and low permeable soils, such as gyttja, it is necessary to check whether this rate is sufficiently low for the test to be drained. Tfd (kFb

Peat

i

e r = 27*

?fd

Calcareous

soil

e~- = 25.5"

(kPa) 30

30 20

l'fu = 5.2 kPa

10

9

20 Tfd = 0.35 6v

"rf~ : 6.6 kPa

10.

0 Fig. 3.32.

3.6.4

20

/,.0

60 -6"~, (kPa)

o

~

,

y

~

:0.3~.e~,

I_ _i._~.,....~"*" ..I

~(~;:20kPa

20

'

Lo

'

Co

6'v (kPa)

-

Results of drained direct simple shear tests on normally consolidated organic soils from the Antoniny site.

Triaxial test

In the conventional triaxial test, a solid circular specimen is loaded axi-symmetrically. The fluid pressure acting in the triaxial cell provides two of the principal stresses, while the third is provided by both the fluid pressure and the axial force imposed by the piston. Consolidation of the sample can be applied three-dimensionally at any given ratio of axial to lateral stress. Although in situ stress conditions are usually anisotropic, isotropic stress conditions are often used in routine triaxial tests for soils with the coefficient of earth pressure at rest higher than 0.8. Testing of the sample can be conducted at any ratio of principal stresses or at a constant mean stress (Bishop and Wesley, 1975). To model the loading conditions imposed by the construction of an embankment, active triaxial compression tests can be performed

125

Determination of Shear Strength

by increasing the axial stress or decreasing the lateral stress, and passive triaxial extension test can be performed in the opposite way (Fig. 3.33). To simulate the field drainage conditions, triaxial tests can be carried out in drained or undrained conditions (Head, 1986).

compression

compression

!

1461

I

U~

c5

c~

II

II

A~I

extension

Fig. 3.33.

9

extension

Various loading paths in triaxial compression and extension tests used to model the loading conditions beneath embankments.

Because organic soils are often extremely loose and sensitive to disturbance, special equipment should be used for mounting specimens in the triaxial cell. Due to difficulties in the preparation of organic soils with a high fibre content, the triaxial tests should be performed on specimens with a larger diameter than that used in standard tests (Landva et al., 1986). The specimen in a triaxial cell should be confined by a rubber membrane with very good resistance to water diffusion to avoid leakage problems. To reduce problems with leakage through the rubber membrane, the triaxial cell can be filled with liquid paraffin (Berre, 1981) or castor oil (Tavenas et al. 1983). Filter paper drains fitted to the side of the sample help to equalize pore pressures as well as provide short radial drainage paths. In the case of low permeable organic soils, they reduce the testing time both for consolidation and for the test phase. To reduce the effect of the side drains on the stress conditions, the filter paper drains should be placed in spirals around the specimen with inclination 1:1.3 for compression tests and 1:1.5 for extension tests (Berre, 1981). The results of triaxial tests on heterogeneous soils (mixed soils, coarse soils and fissured soils) are sensitive to the size of the specimen and in such materials large specimens should be tested.

Laboratory Investigations

126

The results of all triaxial tests are sensitive to the testing procedure and the imposed strain rate. The standard rate used in Scandinavian laboratories for undrained tests on specimens with a diameter of 50 mm and a height of 100 mm is a deformation of 0.6 % per hour. This rate has been found to give appropriate results for calculations of stability in soft clays. Investigations by Graham et al. 1983 have shown that in this rate region a tenfold decrease in strain rate results in a decrease in shear strength of 10 - 20 % for clays (Fig. 3.34). Similar results have been obtained for organic clays. Considering .that some organic soils such as gyttja and clayey gyttja often have coefficients of consolidation about one order of magnitude lower than soft clays, it would be prudent either to reduce the testing rate to 0.06 % per hour or to use a correction factor of 0.8 to 0.9 on results obtained with standard tests. Belfast clay___,._ ~

9

u

~,~ 0.5 -4--"

0.4

tn

i.. ETI r

0.3

_- - - o ~ ~ ~

~

.m

t_ r-

0.2

"a

~-

C

~__._.~a

0.1 -

L,. C

0

i

0.003

Fig.3.34.

0.01

I

0.03

1

0.1

9CAU o CAU 9CAU a JCAU 0.3

Triaxiat Triaxiai Triaxial Triaxiat J

compression-constant rate compression-relaxation compression-step changing extension-relaxation i I

1 3 Strain rate ( % / h )

10

Influence of strain rate on undrained shearing resistance from undrained triaxial compression and extension tests (Graham et al., 1983).

The results of undrained triaxial tests can be evaluated as undrained shear strengths from the stress-strain curves and as effective strength parameters at constant volume from the stress-paths in the tests (Fig. 3.35). Drained tests must be run slowly enough to allow water to drain out of the sample so that no appreciable pore pressure can build up. Filter paper side drains and drainage at both ends are used to speed up the water flow from the specimen. The highest permissible rate of axial displacement (VlmJ during drained tests in such condition can be calculated from the following expression:

Determination of Shear Strength

g.

L,O

127

~ Catcareous soil

Peat c'= 2kPa

"-" z'Oi

c'=3kPa

I

'

~ 2o 0

..-

Fig. 3.35.

0

20

o.s

40

60

I lkPo !

80

'-'

1

0

I,

,

|

,

20

,~

LO

,

,

60

,

,

~.

~_____

80

s':0.5(6'~-6'3 ) (kPa)

Stress paths in undrained triaxial compression tests performed on organic soils from the Antoniny site.

(V1 max ) = (H Elf )/(15 tl0 0 )

(3.28)

where H = height of sample e~f = expected axial strain at failure tl00 = time required for 100 % of primary consolidation. Data from the consolidation stage can be used to determine h00 and a suitable rate of strain for the drained tests can be calculated. For organic soils with low permeability, shearing to failure sometimes takes several weeks. Triaxial tests on fibrous peat are very difficult to interpret. The fibres act as a horizontal reinforcement, so failure is seldom obtained in a drained test. Only large compressions occur. In undrained tests, failure usually occurs when the pore pressure build-up is so large that tensile stresses occur and the sample cracks. This is very different to the behaviour of granular materials and clays and should not be interpreted in the same way.

3.7

D E T E R M I N A T I O N OF P E R M E A B I L I T Y

The permeability of a soil is a function of its void ratio. Many authors (Berry and Poskitt 1972, Larsson 1981, Tavenas et al. 1983) have found that for the working range of compressions the permeability can be simplified to a straight line in the log permeability - strain plot. The logarithm of the permeability can thus be written as: log k e - log k i - I]k" e

(3.29)

Laboratory Investigations

128 where ke = permeability at strain e k i = initial permeability ~k = permeability change index e = strain

The initial permeability is determined by field or laboratory tests. The permeability change index is determined by laboratory tests or may, for some types of soil, be estimated from empirical correlations with water content or void ratio. For some types of peat, both the initial permeability and the permeability change index can be estimated from empirical correlations with the natural water content (Carlsten 1988) Fig. 3.36. These values are to be used in preliminary calculations only and have to be verified by laboratory tests before they are used in design calculations. -7

I

I

I

I

w

I

V////~

I

I

8

I

Peat H5 - H 10

o ._~

-

- 50

1000

2000

Water content, w (%)

Fig. 3.36.

z, 0

1

1000 Water content, w

1

2000 (%}

Relation between permeability and water content for peat. (a) Initial permeability coefficient kvo. (b) Permeability change index 13k (Carlsten, 1988)

Depending on the type of design method to be used for the embankment and type of calculations to be performed, the horizontal permeability k h may have to be determined as well as the vertical permeability k v. The vertical permeability is normally determined by CRS-tests or by permeability tests at various deformation levels in the incremental oedometer tests. In special cases, where it may be assumed that creep effects have a negligible influence on the course of the primary consolidation and the soil is fully saturated, the vertical permeability may be roughly estimated from the time settlement curves in the incremental tests.

Determination of Permeability

129

Permeability tests in both directions can be performed on soil specimens in the Rowe cell (Fig. 3.37; Rowe and Barden, 1966) either with flow of water in the vertical direction (upwards or downwards) or with radial flow horizontally (inwards or outwards). The permeability tests are carried out after vertical consolidation for incremental loads at the end of each loading stage with flow and drainage conditions designed to simulate the field conditions.

,• Y

o)

II

!1

Back pressure system (1)

"~

......

(c tosed ) ~x'~X'~'~NxNXx~ ~.XXXXx~lj~NXXXx~~'~XXXXxN~ outflow~ p I) ~ (with volume change gauge) ~~--..~F/'~ rigid plate with centrat hole

|

~------~11 _---:----~

!'i:!i imi; i,' t'e

r upward flow (shown)... P2 > Pl for downward flow ........... Pl > P2

Back pressure system (2)

Back pressure system { 1 ) ~n flow ~)

(closed)

F~o~ou~plo~tio F~--;-~ ! /

~ r-lip t e~ , ' ~ ' ~

d:t::age hole plugged

K~,\\\\\\\\\\~

out flow (

sand drain

\

]

For flow radiaUy inwards (shown)..p1 > P2 For flow radially outwards ........ P2 > Pl

Fig. 3.37.

Back pressure system (2) (with volume change gauge)

Arrangement of Rowe cell for permeability test. (a) Vertical flow. (b) Radial flow (Rowe and Barden, 1966).

Laboratory Investigations

130

Measurement of permeability in the Rowe cell can be performed with four different types of testing procedures (Fig. 3.38): 9 with the flow of water vertically upwards, 9 with water flow vertically downwards, 9 with horizontal flow radially outwards, 9 with horizontal flow radially inwards. Back pressure

Diaphragm

Back pressure

b)

..............................;~//. Back pressure (2) ~ ~ (Inflow) ......

/, ~W"7)'7"~ ~YJ7);2J /, Back pressure --------d;b(... Outflow) (2) b.p.(1) > b.p.(2)

b.p.(2) > b.p.(1)

d)

/7 S,

"//////,

Back pressure (1)

S,, , @ (Inflow) Back pressure b.p.(2) > b.p.(1) (2)

Fig. 3.38.

| ~

.

[

~

Back /~ pressure 7

':i~,

(~ ,(Outflow) b. p.(1) > b.p.(2)

Back pressure (2)

Flow conditions for permeability test in Rowe cell. (a) and (b) Vertical flow. (c) and (d) Radial flow (Head, 1986).

Permeability tests are usually carried out under "equal strain" loading, i.e. with a rigid plate on top of the specimen to maintain a uniform sample thickness. The calculation of permeability from vertical and horizontal flow tests is performed for each of the tests at various deformations from measurements of water flow rate and hydraulic gradient across the sample. The vertical permeability k is calculated from:

Determination of Permeability

131

(3.3o)

k v - q/(A 9i) or

kv - q .

H/(A.

Ap)

(3.31)

where A = cross section area of sample i = hydraulic gradient q = flow rate of water H = height of sample Ap = in difference water pressure at the ends of the sample. The horizontal permeability is evaluated assuming axi-symmetrical radial flow of water. The coefficient of horizontal permeability k h is calculated from:

kh =

q ' P w "g

In (D/d)

(3.32)

2 7 z ' H " Ap where D = diameter of sample d = diameter of central drain Ap = applied pore pressure difference Pw = density of water g = 9.81 m/s 2 The relaton between the vertical and horizontal permeabilities is the permeability anisotropy of the soil. The anisotropy of the permeability is especially pronounced in peats with low degrees of humification and some varved and layered soils which have very high horizontal permeabilities k h as compared to the vertical k v. The anisotropy decreases with increasing degree ofhumification and, as in other homogenous soils, it becomes negligible for highly decomposed peats.

3.8 REFERENCES Aboshi, H., Yoshikumi, H. and Maruyama, S. (1970). Constant loading the consolidation test. Soils and Foundations, Vol. 10, pp.43-56.

132

Laboratory Investigations

Adams, J.I.(1965). The engineering behaviour of Canadian Muskeg. Proc. of the 6th International Conference on Soil Mechanics and Foundation Engineering, Montreal, Vol. 1, pp.3-7. Baranski, T. and Wolski, W. (1985). Ultrasonic testing for the strain characteristics of soil, ASTM Symposium on Consolidation of Soils, Fort Lauderdale, Florida, pp.516-525 Barden, L. (1968). Primary and secondary consolidation of clay and peat, Geotechnique, Vol. 18, pp.345-362 Berre, T., (1981). Triaxial testing at the Norwegian Geotechnical Institute. Norwegian Geotechnical Institute, Publication No. 134. Berry, EL. and Poskitt, T.J. (1972). The consolidation of peat. Geotechnique, Vol.22, pp.27-52. Bishop, A.W. (1954). Correspondence on shear characteristics of a saturated silt measured in triaxial compression. Geotechnique, Vol. 4, No. 1, pp. 43-45. Bishop, A.W. and Wesley, L.D. (1975). A hydraulic triaxial apparatus for controlled stress path testing, Geotechnique, Vol. 25, No. 4, pp. 657-670. Bjerrurn, L. (1972). Embankments on soft ground. Proc. of the ASCE Specialty Conference on Performance of Earth and Earth Supported Structures, Purdue University, Lafayette, Vol. 1, pp. 1-54. Bjerrurn, L. and Landva, A.O. (1966). Direct simple shear tests on a Norwegian quick clay. Geotechnique, Vol. 16, No. 1, pp. 1-20. Casagrande, A. (1936). The determination of the preconsolidation load and it's practical significance. Proc. 1st International Conference on Soil Mechanics and Foundation Engineering, Cambridge, Mass, Vol. 3, pp. 60. Carlsten, P. (1988). The use ofpreloading when building roads an peat. Proc. of the 2nd Baltic Conference on Soil Mechanics and Foundation Engineering, Tallinn, pp. 135-143. Carlsten, P. (1988). Geotechnical properties of peat and up-to-date methods of design and construction on peat. State-of-the-art report. 2nd Baltic Conference on Soil Mechanics and Foundation Engineering, Tallinn. Also in Swedish Geotechnical Institute Varia No. 215. Link6ping. Dyvik, R., Berre, T., Lacasse, S. and Raadirn, B. (1987). Comparison of truly undrained and constant volume direct simple shear tests, Geotechnique, Vol.37, No.l, pp. 3-10. Edil, T.B. and Dhowian, A.W., (1979). Analysis of long therm compression of peats. Geotechnical Engineering, Vol. 10, No. 2, pp. 159-178.

References

13 3

Edil, T.B. and Dhowian, A.W. (1981). At-rest lateral pressure of peat soils, Journal of the Geotechnical Engineering Division, ASCE, Vol.107, No. GT2, pp. 201-217. Edil, T.B. and Mochtar, N.E. (1984). Prediction of peat settlement. Proc. of the Symposium on Sedimentation/Consolidation Models, San Francisco ASCE, pp. 411-424. Felix, B. (1982). Systeme de mesure des deformations radiales pour eprouvettes de sol. Laboratoire Central des Ponts et Chaussees, Paris. Fredriksson, D. and Kjellin, B., (1973). Meddelande till kunder vid jordartslaboratoilet. Swedish Geological Survey. Garneau, R. and Lebihan, J.P. (1977). Estimation of some properties Champlain clays with the Swedish fall cone. Canadian Geotechnical Journal,, Vol. 14, No. 4, pp. 571-581. Gibson, R.E. and Lo, K. (1961). A theory of consolidation for soils exhibiting secondary compression. Acta Polytechnica Scandinavica No. 41, pp. 1-15. Graham, J., Noonan, M.L. and Lew, K.V. (1983). Yield states and stress-strain relationships in a natural plastic clay. Canadian Geotechnical Journal, Vol. 26, pp. 502-516. Hansbo, S. (1957). A new approach to the determination of the shear strength of clay by the fall-cone test. Swedish Geotechnical Institute, Proceedings No. 14, pp.5-47. Hanrahan, E.T. (1954). An investigation into some physical properties of peat. Geotechnique, Vol. 4, pp. 108-123. Head, K.H. (1986). Soil laboratory testing, Vol. 3, Effective stress tests, Pentech Press. Hobbs, N.B. (1986). Mire morphology and the properties and behaviour of some British and foreign peats. Quaternary Journal of Engineering Geology, Vol. 19, pp. 7-80. Janbu, N. (1963). Soil compressibility as determined by oedometer and triaxial tests. Proc. 3rd European Conference on Soil Mechanics and Foundation Engineering, Vol. 1. Wiesbaden. Karlsson, R. (1981). Consistency Limits. Swedish Council for Building Research, Document D9:1981, Stockholm. Kjeliman, W. (1951). Testing the shear strength of clay in Sweden. Geotechnique, Vol. 2, No. 3, pp.225-232.

134

Laboratory Investigations

Ladd, C.C. (1981). Discussion on laboratory shear devices. Laboratory shear strength of soil, ASTM STP 740, pp. 643-652. Landva, A.O., Korpijaakko, E.O. and Pheeney, P.E. (1983). Geotechnical classification of peats and organic soils. Symposium on testing of peats and organic soils, ASTM, pp. 37-51. Landva, A.O., Pheeney, P.E. and Mersereau, D.E. (1983). Undisturbed sampling of peat. Symposium on testing of peats and organic soils, ASTM, pp. 141- 156. Landva, A.O., Korpijaakko, E.O. and Pheeney, P.E. (1986). Notes on the original von Post peat and peatland classification system. Proc. Advances in Peatlands Engineering, Ottawa, pp. 17-29. Larsson, R. (1977). Basic behaviour of Scandinavian soft clays. Swedish Geotechnical Institute, Link6ping, Report No. 4. Larsson, R. (1981). Drained behaviour of Swedish clays. Swedish Geotechnical Institute, Linkrping, Report No. 12. Larsson, R. (1986). Consolidation of soft soils. Swedish Geotechnical Institute, Link6ping, Report No. 29. Larsson, R. (1990). Behaviour of organic clay and gyttja. Swedish Geotechnical Institute, Link6ping, Report No. 38. Larsson, R. and S~illfors, G. (1985). Automatic continous consolidation testing in Sweden. ASTM Symposium on consolidation of soils: Testing and evaluation, Fort Lauderdale. ASTM STP 892 Consolidation behaviour of soil. Larsson, R., Niison, G. and Rogbeck, J. (1987). Determination of organic content, carbonate content and sulphur content in soils. Swedish Geotechnical Institute, Link6ping, Report no. 27E. Law, K.T. (1979). Triaxial-vane tests on a soft marine clay. Canadian Geotechnical Joumal, Vol. 16, pp. 11-16. Lechowicz, Z. and Szymanski, A. (1988). Deformation analyses of organic subsoil in anisotropic stress conditions. Archiwum Hydrotechniki, Gdansk. Vol. XXV, pp. 125- 133. Lechowicz, Z. and Szymanski, A. (1988). Creep behaviour of organic soils, Annals of Warsaw Agricultural University, Land Reclamation, No.24, pp. 99 - 106. Leroueil, S., Collins, G. and Tavenas, E, (1983). Total and effective stress analyses of slopes in Champlain sea clays. Symposium on Slopes on Soft Clays. Swedish Geotechnical Institute, Link6ping, Report No. 17, pp. 293-321.

References

135

Lowe, J., Jonas, E. and Obrician, V., (1969). Controlled Gradient consolidation Test. Journal of the Soil Mechanics and Foundation Division, Vol. 95, No. SM 1, pp. 77-97. Mayne, P.W. and Kulhawy, EH., (1982). K0-OCR Relationships in soil. Journal of the Geotechnical Engineering Division, ASCE, No. GT6, pp. 851-871. Moum, J., (1967). Determination of inorganic and organic carbon in soil samples. Internal report, Norwegian Geotechnical Institute, Oslo. Ozden, Z.S. and Wilson, N.E. (1970). Shear strength characteristics and structure of organic soils. Proc. of the 13th Muskeg Research Conference, Fredericton, New Brunswick, National Research Council, Technical Memorandum No. 99, Ottawa. Paute, J.L. (1983). Comportement des sols supports de Chaussees a 1' appareil triaxial a'chargements repeter. Bulletin 123, Laboratoire Central de Ponts et Chaussees, Paris. Rowe, P.W. and Barden, L. (1966). A new consolidation cell. Geotechnique, Vol. 16, No. 2, pp. 162-170. S~ilifors, G. (1975). Preconsolidation pressure of soft, high-plastic clays. PhD Thesis, Chalmers University of Technology, Gothenburg, Sweden. Schmidt, B. (1966). Earth pressure at rest related to stress history. Canadian Geotechnical Journal, Vol. 3, pp. 239-343. Singh, A. and Mitchell, J.K. (1968). General stress-strain-time function for soils. Journal of the Soil Mechanics and Foundations Division, ASCE, Vol. 94, No. SM 1, pp. 21-46. Skempton, A.W., (1954). Discussion of the Structure of Inorganic Soil. Journal of the Soil Mechanics and Foundations Division, ASCE, Vol. 80, p. 478, New York. Smith, R.E. and Wahls, H.E. (1969). Consolidation under constant rate of strain. Journal of the Soil Mechanics and Foundation Division, ASCE, Vol. 95, No. SM 2, pp. 519-539. Szymanski, A., Fiirstenberg, A., Lechowicz, Z. and Wolski W. (1983). Consolidation of organic soils. Proc. of the 7th Danube European Conference on Soil Mechanics and Foundation Engineering, Kishinev, Vol.II, pp. 273-278. Tavenas, E, Leblond, P., Jean, P. and Leroueil, S. (1983). The permeability of natural soft clays. Canadian Geotechnical Journal, Vol. 20, No. 4, pp. 629- 644.

Tavenas, F., Jean, P., Lebiond, P. and Leroueil, S. (1983). The permeability of natural soft clays. Canadian Geotechnical Journal, Vol. 20, No. 4, pp. 645- 660.

136

Laboratory Investigations

Terzaghi, K. (1925). Principles of Soil Mechanics, Engineering News-Record 26. Thalme, O. and Almen, K-E. (1975). Jordartsanalys. Laboratorieanvisningar, Del 1. Kvart~rgeologiska institutionen, Stockholm University. Tokheim, O. and Janbu, N. (1980). A continous consolidation test. Norwegian Institute of Technology, Geotechnical Division,. Meddelelse No. 9. Wasti, Y. and Bezirci, M.H. (1986). Determination of the consistency limits of soils by the fall cone test. Canadian Geotechnical Journal, Vol. 23, pp. 241- 246. Woiski, W., Baranski, T., Garbulewski, K., Lechowicz,Z. and Szymanski, A. (1985). Testing of anisotropic consolidation in organic soils. Proc. l lth International Conference on Soil Mechanics and Foundation Engineering, San Francisco, Vol. 2, pp. 699-702. Wood, D.M. (1981). Cone penetrometer and liquid limit. Geotechnique, Vol. 32, No. 2, pp. 152-157. Zawadzki, S. (1970). Relationship between the content of organic matter and physical properties ofhydrogenic soils. Polish Journal of Soil Science, Vol. III, No. 1.

137

Chapter4

Stability Analysis z. Lechowicz, Department of Geotechnics, Warsaw Agricultural University

4.1

GENERAL

For embankments and dikes on organic soils, an evaluation of the stability during construction is of major importance. Stability analyses may also have to be carried out for other particular loading situations, such as: (a) steady seepage conditions (b) sudden drawdown of the water level and (c) application of additional load. In estimation of stability of embankments on organic soils, two particular cases may be identified: single-stage loading - when the initial shear strength of the soil is sufficient to support safely the maximum embankment load, and multi-stage loading (stage-constructed embankments) - which requires evaluation of the shear strength increase with load and time for the design of a safe loading rate. In evaluation of the stability of embankments on organic soils, more attention should be paid to the appropriate selection of initial values of soil parameters and their changes due to consolidation than to the application of more complex methods of stability calculation. Therefore, this chapter gives an outline of the approach for estimation of stability used in current design practice with special focus on the selection of initial values of shear strength and prediction of the increase in shear strength, as these parameters are of major concern when loading organic soils. In geotechnical practice, the methods of slices, based on the limit equilibrium approach, are commonly used for estimation of stability. The presentation in this chapter is mainly concemed with these methods, but simple procedures of stability evaluation and trends in advanced techniques for stability assessment are also presented. Taking into account the way in which the available shear strength on a potential failure surface acts and how the resulting factor of safety is computed, two types of analysis are usually considered: (a) total stress analysis and (b) effective stress analysis. The total stress analysis assumes undrained conditions and consequently uses undrained shear strength parameters. The effective stress analysis can consider undrained or drained conditions using effective stresses and effective strength parameters.

13 8

Stability Analysis

4.2

SHEAR STRENGTH USED IN STABILITY ANALYSIS

4.2.1

Problems in the evaluation of shear strength

For design of embankments on organic subsoil, the information concerning in situ shear strength is often insufficient. Moreover, changes in shear strength resulting from future changes in the stress conditions are required. The shear behaviour of organic soils is strongly affected by the pore pressure, the effective stress state and the stress history. In engineering practice several different laboratory and field tests are used to estimate the shear strength. In situ methods, such as the vane shear test, dilatometer test and penetration tests, are often used with preference as they are costeffective and also avoid some of the problems associated with disturbance of soil samples during extraction from the subsoil. However, these methods can only be used to determine the shear strength at the prevailing stress conditions and the interpretation of the results often has to be calibrated against qualified laboratory tests. On the other hand, sampling and laboratory testing permit modelling of almost any stress condition and evaluation of the change in shear strength with changing stress conditions. Both in situ and laboratory testing are normally used simultaneously. The testing rate in in situ tests is usually so high in relation to the low permeability of soil and organic soils that the tests can be considered as completely undrained. Fibrous peat is an exception as the permeability may be so high that the tests become at least partly drained. In most laboratory tests, the drainage conditions can be controlled and therefore the undrained as well as the effective shear strength parameters can be estimated.

4.2.2

Undrained shear strength

It has been widely recognized that the shear strength value 1;f~ obtained from field vane shear tests cannot be used directly in the calculation of the stability of embankments. To evaluate undrained shear strength of organic soils from vane shear tests, the measured values have to be corrected according to: % = ~t % where Zfu = corrected undrained shear strength g = correction factor Zfv = shear strength value from vane shear test

(4.1)

139

Shear Strength used in Stability Analysis

Bjerrum's correction factor g is related to the plasticity index Ip and is often used, although it is mainly valid for normally consolidated low plastic clay deposits (Bjerrum, 1972) (Fig. 4.1). A revised version of this correction, which attempts to consider the influence of end effects in the full scale failures forming the empirical basis of the factor, gives corrected field vane shear strengths about 10 % lower than those previously recommended by Bjerrum (Azzouz et al., 1983).

:~. {.,.

1.2

I

I

I

I

1

- 1.1-

0

U

o 1 0 -. IE O

u

Bierrum (1972)

0.9-

I... I...

o

o

0.80.70.60.5

0

Azzouz

11983) I

20

et

at./ I

40

I

60

I

80 Plasticity index,

Fig. 4.1.

1

100 Ip

120 (%)

Field vane strength correction factor proposed by Bjerrum (1972) and revised version presented by Azzouz et al. (1983).

Based on Swedish experience in normally consolidated and slightly overconsolidated high plastic soft soils, the Swedish Geotechnical Institute (Larsson et al., 1984) has recommended the correction factor g calculated as a function of the liquid limit w L (Fig. 4.2). Correction factors higher than 1.2 should not be used without supporting evidence from complementary ilwestigations. The lowest correction factor is 0.5. All these correction factors have been derived from empirical experience and are averages of data with a considerable scatter. If there is any doubt about the validity of the correction factors in a particular soil of interest, the field vane test should be calibrated against results from triaxial and direct simple shear tests. Several investigations have been carried out to observe the mode of failure during field vane shear tests in organic soils. Field vane shear tests (Landva, 1980) performed by the 65xl 30 mm vane in sphagnum peat showed that the diameter of the

Stability Analysis

140

1.3

I

I

I

i

I

I

I

:~,

I

I

0.45

~

=

U

Ol. 1 C

"

.s "~ 1.OL. 0

"

0.9O.80.70.60.5

0

I

I 0.4

I

I 0.8

~

J 1.2 Liquid

Fig. 4.2.

l

J 1.6

2.0

limit, WL

C o r r e c t i o n f a c t o r f o r s h e a r s t r e n g t h v a l u e s o b t a i n e d by v a n e s h e a r t e s t or S w e d i s h f a l l - c o n e test ( L a r s s o n et al., 1984).

observed failure surface was 7-10 mm greater than that of the vane. Based on the measurements of the failure zone during vane shear tests, Golebiewska (1976, 1983) has proposed the correction factors for peats ~t = 0.5-0.55 and for gyttja and calcareous soil ~t = 0.6-0.8. In organic soil, the scatter in the results from field vane shear tests is often quite large. To obtain the undrained shear strength profile for design, the average of several corrected shear strength values should be calculated (Larsson et al., 1984; Baecher and Ladd, 1985). An example is shown in Fig. 4.3, where shear strength values obtained in field vane shear tests corrected according to the SGI recormnendations are presented. Fig. 4.3 also shows undrained shear strength obtained from triaxial compression tests (TC), direct simple shear tests (DSS) and triaxial extension tests (TE) as well as the average undrained shear strength obtained from laboratory tests. A comparison between the average undrained shear strength obtained from laboratory tests and the corrected shear strength from field vane shear tests indicates that for the organic soils at the Antoniny site, the correction factors estimated according to the SGI recommendations are in general agreement. Differences were obtained for calcareous soil where a reduction in addition to the general correction factor was necessary. The calibrated correction factor for calcareous soil estimated on the basis of the average laboratory undrained shear strength and field vane shear strength

Shear Strength used in Stabili.tF Analysis

0

0

~

I

I,I~L.

corrected

.c:: 4 r'~

field

vane

_.

5 - o (/) (/)

ov,ro,,

Lob.

I

//~J/clverag e f ~ / " ,measured I,,~'C field v a n e

~,~t,

2-a_

6-

I

,•/

0

0

Undrained shear strength, (kPa) 10 15 20 25

5

_

-E i3 -

141

II

"~x ~"~~>

Cr

l

"-"~-,.~>

.,,.p

,'Y/-

#1c if!

O (b 1..

8

7--~

O

_+one standard deviation I

1

I

I

Fig. 4.3. Undrained shear strength profile obtained from field vane shear tests at the Antoniny site.

values equals 0.6, while a correction factor obtained from the SGI procedure was 0.7. In order to estimate calibrated correction factors which correspond to different modes of shearing, the undrained shear strength obtained from different types of laboratory tests can be related to field vane shear strength values. In this way, more reliable correction factors for field vane shear strength values can be obtained. The cone penetration test with simultaneous measurement of pore pressure, i.e. the piezocone test, can be used to estimate the undrained shear strength, (see Chapter 2.5). Similar to the correction factor p to the vane shear test, the cone factor NKT can be calibrated locally for a specific type of soil in order to obtain more reliable values. An exalnple of a such calibration in a soil stratum with peat and calcareous soil is given in Fig. 4.4. In this particular case the cone factor was found to be 22 for the peat and 16 for the calcareous soil. The dilatometer test can also be used to obtain the undrained shear strength (see Chapter 2.5). Alternatively, the undrained shear strength ~fu can be estimated from

142

Analysis

Stability

Total point resistance (kPa)

qT 0

!

200 !

0(b CL

I

600

400 i

outside the ill[

.c.

10

I

I

,, 20

I

I

1

i,x /'t'fu (LAB}

~)3

x1~,

~'///////Z

--3 E

'

t/~"

~

o

it ],

N

v

40

i corrected field

vane test

,,~.

~k ~ e

trati~

test

0

-

-

-

under the fill i

qTfu

v

//~./calibrated corrected

rl M" ~

I.. 0

~=

30

::. :...:.~ /Ill~l/,

~

,

5-

Fig. 4.4.

w

J:

u

o

30

Estimated undrained shear strength (kPa)

1

I

1

I

I

I

I

i

I

Undrained shear strengths beneath and outside the test fill at the Antoniny site.

S (0.45 9KD)"2~

(4.2)

(Y'v where ~'v = vertical effective stress S = normalized undrained shear strength for the normally consolidated state ( O C R - 1) K D- horizontal stress index An example of a comparison between calibrated corrected field vane shear strength values and undrained shear strength from dilatometer test evaluated based on Eqn. 4.2 for a soil stratum with peat and calcareous soil is given in Fig. 4.5. In this case the Sparameter was found to be 0.5 for the peat and 0.45 for the calcareous soil (see Chapter 4.5.2). The relevant range of undrained shear strength of organic soils for design of small structures or as preliminary low-cost investigations can often be estimated from Swedish fall-cone tests or from laboratory vane shear tests. However, these

Shear Strength used in Stability Analysis

143

Horizontal Materiel index stress index ID KD 0;4 0.8 0 4 8 00 '

I

'

9

I

,

'

!

'

1

'

i

I

i

'

Ditatometer modulus ED, MPa 0 0.4 0.8 1.2 '

I

v

!

'

9

J

Undrained shear strength Tfu, kPo 0 5 10 15 v

!

I

v

|

_

Q.

E

,s I

11

i

!

|

i

v

i

i

) (

E3

1

1

bU

o (..) 9

Fig. 4.5.

I

n

I

I

I

J

9

i

,

I

9

I

,

i

,

,

!

,

i

l

.~._c

1,

.CI .1O

I~

l

i

=

1

l

,

,

"6

,

i

l

l

1

I

Undrained shear strength profile obtained from dilatometer test at Antoninv site.

tests are obviously not applicable for highly fibrous soils. According to Landva et al. (1986) the application of these methods is doubtful when the fibre content amounts to 40 %. From the Swedish fall-cone test, not only the undrained shear strength can be roughly estimated but also the remoulded shear strength and the liquid limit. Because of this, the fall-cone test is used as a routine test, mainly in the Scandinavian countries but to an increasing extent around the world. The appropriate mode of shear at different locations in the soil beneath and outside an embankment can be simulated by triaxial compression, direct simple shear and triaxial extension tests for evaluation of undrained shear strength. In order to simulate the initial in situ stress conditions, the specimens should be consolidated under conditions of no lateral strain, that is, under K 0- conditions. The results obtained from shear tests are often presented as normalized values (i.e. divided by the pre-shear vertical consolidation stress Or'v) (Fig. 4.6). Because the designer of stage-constructed embankments must estimate not only the initial undrained shear strength but also how it will change throughout the period of construction, the laboratory shear tests should provide the data to evaluate the normalized undrained shear strength versus the overconsolidation ratio. In order to estimate correctly the initial undrained shear strength at stresses below the preconsolidation pressure, it is important that the preconsolidation stresses are not exceed-

Stability Analysis

144 a) 2"01 Test Sym S m 1.5 TC 0.325 0.78 DSS o 0.2550.78

:~ TE 1.0 6

v

I

I

I

L..I

I

/~-

/

Intact'

l'~

s Iml

IDSSI 0.29 10.69510:2S /

'

'

'

~ ~ -

-[TE ] 0.235l0,82 1".0,20 A ~I ~ ' / ~ -

-

~

0.4

i " 7 -r ~

'rfu=qf for TC end TE Tfu=Tma x for DSS

.

1 2 3 456 810 Overconsolidation ratio OCR= 6p / 6'v !

Fig. 4.6.

s ]

-ITc 10.t.5 10. ! 0.33sl

0.2ooo.8~,,,~/

0.6

"~ 0 2 l ~ ::3 ,

1Desth/

- ~ DSS

o

-

o~

]

Condition ,IOCRITClDSSlTE

lhs~l

o I~

: D,estructumdl , / A I 9I 9 1 1.5 2.0 2.5 3.0 3.5 4.0 OverconsolidQtion rQtio OCR= 6';::)/~v i

Relation between normalized undrained shear strength versus overconsolidation ratio relationship for marine clay. (a) SHANSEP CKoU tests. (b) CKoU tests on intact and destructured samples (Jamiolkowski et al., 1985).

ed and that the soil is allowed to keep its structure. Once the preconsolidation pressure is exceeded, the soil becomes "destructured", which in soft soils entails large compressions. In some materials, this also entails a breakdown of a strength component termed bonding or cementing of the structure. In the case of embankments on soft soils, loading involves exceeding the preconsolidation pressure with consequent large deformations, and the destructured strength thereby constitutes the relevant strength. 4.2.3

Effective

shear strength

The effective shear strength used in effective stress stability analysis is expressed by the Mohr-Coulomb effective shear strength parameters c" and @'. To simulate the appropriate mode of shear behaviour below an embankment, the effective shear strength parameters can be investigated by triaxial and/or direct simple shear tests. The strength parameters c' and @" are normally determined in totally drained triaxial tests, but may also be estimated from the effective stress paths in undrained triaxial tests. Undrained triaxial tests give the effective shear strength parameters at constant volume, which underestimate the shear strength somewhat in overconsoli-

Methods of Stability Analysis

145

dated state and overestimate the shear strength that can normally be used in the normally consolidated state. The parameters for constant volume are used in calculations of stability in the undrained case in terms of effective stress analyses. Effective shear strength parameters c" and ~" can also be evaluated from drained direct simple shears. Parameters c" and ~" obtained from direct simple shear tests are normally somewhat lower than the values from triaxial compression tests.

4.3

METHODS OF STABILITY ANALYSIS

4.3.1

Types and scope of analysis

In the preliminary design stage, the stability of an embankment may be estimated with the aid of simple rule of thumb or stability charts. The estimation of a safe embankment height during the first stage of construction can thus be made using e.g. Terzaghi's equation or Taylor's stability chart. Having established the pore pressure conditions, it is also possible to estimate the embankment stability at steady seepage using Cousins' stability charts (Cousins 1978). Preliminary assessment of embankment stability during rapid drawdown of water level, e.g. for dykes can be carried out using Morgenstem's stability charts (Morgenstem 1963). For the design of embankments under complex subsoil and pore pressure conditions the two-dimensional methods of slices with assumed plane strain conditions are often used (Fredlund and Krahn, 1977). In the limit equilibrium analysis, the average factor of safety F may be defined as follows: F = ~f / ~

(4.3)

where 1:f = available shear strength along a shear surface I: = equilibrium shear stress along the same shear surface In the methods of slices, the potential sliding mass is subdivided into a number of imaginary vertical slices with width b. Each slice is acted upon by its own weight W and by the boundary interslice forces which have both a tangential component T and normal component E (Fig. 4.7). The forces acting on the basis of a slice with inclination cz and length 1 are shear resistance S~ and normal force N, Fig 4.7. The methods of slices most commonly used for stability calculations are: the Swedish circle method; the simplified Bishop method; the Janbu generalized and simple methods; and the Morgenstem-Price method. The Swedish circle method is

Stability Analysis

146

~---

I

I

~---

.....

-1-1

\ \

"x

.

Fig. 4.7. Forces involved in the methods of slices.

the simplest method since no interslice forces are considered. Because the Newtonian force principles at the interslices are not satisfied, this method may in extreme cases give errors in the factor of safety of as much as 60 % (Whitman and Bailey, 1967). Of the other methods listed above, in which interslice forces are considered, the simplest and probably the internationally most commonly used is the simplified Bishop method. The factors of safety calculated by the above mentioned methods (with the exception of the Swedish circle method) normally differ only marginally for low embankments on soft soils (Wright et al., 1973). The simplified Bishop method may also give errors in the order of 10 % as not all interslice forces are accounted for. The errors are on the "safe" side, i.e. the calculated safety factor is too low. 4.3.2

Simple procedures

of s t a b i l i t y a s s e s s m e n t

(a) Preliminary estimation of safe embankment height: A simple equation commonly used to estimate the safe height for an embankment on homogeneous subsoil in undrained conditions is: H~ = (NdXfu)/(F-~(~)

(Terzaghi, 1956)

(4.4)

Methods of Stability Analysis

147

where H = height with acceptable factor of safety F N~ = stability number l:f~ = undrained shear strength of subsoil F = factor of safety y~ = unit weight of embankment. For a circular failure surface the stability number is N~ = 5.52, on the assumption of no internal strength of the embankment soil. The safe height of an embankment on soft subsoil can also be set, ~ from stability charts. One such chart, applicable in total stress analysis of the stability of an embankment on homogeneous subsoil, was developed by Taylor (1956). In this chart, Taylor's stability number N T is related to the depth factor D for given values of the slope angle 13and the parameter M which defines the location of the considered slip circle (Fig. 4.8). The safe height of an embankment can be calculated as: H = ~ f . / ( F . NT'Tc) 0.19 0.18

(4.5)

13=530

-.

0.17 T

~>

~ ~

~'

\\

/...._...~~.-~ ~

-

0.16

"I-- 0.15

z

d

~

0.14

\

/

J3

E c-

0.13

..~

0.12

\ \\0

\

u')

0.11 l i l i l i l i l l l i l i l i i l

0.10

0.09

1

2

3

i/

4

Depth factor, D Fig. 4.8. Taylor's stability chart for ~u = 0 conditions (Taylor, 1956).

3

Stability Analysis

148

Nc.Tfu 5.52.8.5

H s = ~

I I

:J

~

-

".'":'H.s'=?.''."

....

:

=18kNl'm

7. s

,,

MH,

s

F

a)

"

, OH s

~'e 9

=

~

1.3.18

11=18.4'; D = 2 . 5 ;

'*

=2m

M=1.6;

N.,.:o.,+', H s =Tfu/(F N T ~'e)=

EI G y t t j a , ~

~ T f u

=9 kPa

9. S a n d . 9. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .

..:;

= 8.5/(1.3- 0.167.18) = 2.2 m b}

13=18.4'

D=5

M=4

N T = 0.179 H s = 9/(1.3.0.179.18) = 2.1m

Fig. 4.9.

Example of estimation of the safe height for first stage of a dyke based on Taylor's chart.

The use of Taylor's chart is demonstrated by estimation of the safe height of a dike considered in Chapter 9 (Fig. 4.9). In order to find the approximate values of the parameters D and M in the first stage, the embankment height was calculated from Eqn. (4.4).

(b) Steady seepage conditions: Several types of stability charts designed to account for known pore pressure conditions and using effective stress analysis have been introduced. On the basis of the stability charts, an estimation of the factor of safety for known soil parameters, pore pressure conditions and geometry of embankment slope can be made, or a suitable slope inclination having a prescribed safety factor can be selected. To assess embankment stability under steady seepage conditions, Cousins' stability charts can be used (Cousins, 1978). Based on Cousins' stability charts, the safety factor F can be obtained from the following equation: F = N F [c'/(T 9H)]

(4.6)

Methods of Stability Analysis

149

where Nv = stability number; c" = effective cohesion intercept 3' = unit weight of soil (average) H = height of embankment. Cousins' stability charts are given in terms of stability number N F as a function of slope angle 13, depth factor D, dimensionless parameter ~ and pore pressure ratio r~. Dimensionless parameter ~,~ and pore pressure ratio r~ are defined as follows" 9H ) ]

=

(4.7)

and r. =

h)

(4.8)

where ~" = effective angle of internal friction u = estimated pore pressure (average) h = depth of the point below the soil surface for which u was estimated Fig. 4.10. shows stability charts considering critical toe circles and circles with specified depth factor D = 1; 1.25 and 1.5 and different values of pore pressure ratio r = 0, 0.25 and 0.5. (Cousins 1978). For the pore pressure ratio r > 0.5 the estimation of stability number N r can be made by using a new dimensionless parameter defined as follows: ~,'~ = (1 - r ) ~,~

(4.9)

In this way, the stability number N F can be estimated from stability chart for r = 0 by using parameter ~,'~ instead of~,~r The use of parameter g ' r together with the chart for r~ = 0 is recommended only for a slope angle less than 17.5 ~ An example of estimation of the factor of safety for a dyke designed in Chapter 9 under steady seepage conditions by using Cousins' charts is shown in Fig. 4.11. The pore pressure ratio r is calculated using the pore pressure at the lowest point of the slip surface and the average of the unit weights in the embakment material and the peat layer.

Stability Analysis

150

500

I

I

I

I

I

I

I

I

/

ru =0

u_ 300 z

,.5

0 .13

E :3

c

100

-

,,\

~

\Oi.4

1

50

oHIH

,

~

20

""

~

6

10_" " ' ? ~:cc.- - -...._._. ===__.=_~. 0

o,~ r-.,o,:,,,~,~or~~ ~.3oo I

"'~~

......

~.~_~1001|

~\

ru='o.~',' t '~176 3o0 =50

.,oo ,o

~~'~.~--~~ ~,o~ ,

, lo

, ~ - ~ - - - . ~ 2~ 3o

, ,..o

Slope angle, J3 (Degrees)

, o

, ',o

,

, 2~

, 30

,

, ,..o

SLope ongte, J3 (Degrees)

Fig. 4.10. Stability number N F considering specified depth factor D = 1; 1.25 and 1.5 when r u = O; 0.25 and 0.5 (Cousins, 1978).

Methods of Stability Analysis H=3m;D=2.0

151

~=12.5'.

ru>0.5

~

c'/(~' H)=0.042 Xc~) = 13.7

ru =0.6.

~.'cr162

I

f

.

.

NF =38

"

.

.

.

.

.

H=3m H

/ Gyttj(a 9

9S a n d .

.

I F = NF (c'/Ct~" Hi) = 38" 0.042 = 1.59 .

.

9. . . .

. ;...-

.

.

9

..

.

. . . .

i

.

9

.

. . - . .

,

,

.

.

9

.

.

".'.

E]

.

:

.

."

''

"

.

. ..

J-

9

. .

:...

Fig. 4.11. Example of estimation of the safety factor for a dyke under steady seepage conditions based on Cousins' charts.

(c) Sudden drawdown conditions: In cases with dykes or embankments where the fill material has a low permeability, it may be important to investigate the stability under sudden drawdown conditions. For preliminary design purposes, stability charts presented by Morgenstem (1963) can be used. These charts were elaborated for the special loading condition of sudden drawdown from full submergence with the assumption that no dissipation of pore pressure occurs. The charts are given in Fig. 4.12. The values of the safety factor have been plotted against drawdown ratio L/H for ~ '-values ranging from 200 to 40 ~ with values of c'/(y ~ H) equal to 0.0125, 0.025 and 0.05, and slope inclinations from 1:2 to 1:4.

4.3.3

Swedish circle method

The Swedish circle method is the simplest method of slices and is based on the assumption that the interslice forces can be neglected. This assumption causes an error in the estimated safety factor but, due to simplicity in calculation, the method is often used for approximate calculations. The safety factor F can be calculated using the following equations:

Stability Analysis

152

3.5| --

u_

~

/I-

'~

,

,

~

1:2

3~

.

.

.

.

O owOown

~

c'/( ~'e H) = 0.0125

.._

c./(~,eH}=O.025

"-

~'(lr~ H'= 0.05

o

prior to drawdown

.

,

- --

2.s ~. \ \ Level

.... i "'

y

-'-

2

,?

,

1.5:

-

.

"-~-

~ - ~ ~

-

-

0.5O-

J

0

I

0.2

I

I

I

0.4

I

i,.

0.6

Drawdown 7

T

-

LL

~6-

T

I

I

I

I

1

I

# ....

71-

I

LL

1"3 ,

l

|

!

I

I

I

0.8

I

1.0

ratio. L I H !

I'

'1

1"4

>~ _

"-"O

5-

"6

i.. O ..i-.

u 4 El

LL

_\

LL

3

00

i

~

'

L

0.2

I

I

oz,

i

I

06

Drawdown

I

i

0.8

1.

ratio. L/H

1.o

00

" "~ ,,, ,~

I

0.2

.

I

i

0,4

~'~.

I

"

-9- - - - ~ . . _ . . . _ .

I

I

1

,1

0.6 0.8 1.0 Drawdown ratio, LIH

Fig. 4.12. Drawdown stability chart for specified slope inclinations 1:2, 1:3 and 1:4. (1) ~'=20~ (2) ~'=30~ and (3) ~'=40 ~ (Morgenstern, 1963).

Methods of Stability Analysis

153

9 for the total stress analysis E,f~ 1

(4.10)

F

Z W sm~ 9

for the effective stress analysis E(W coscz - u 1) t a n r

+ E c' 1 (4.11)

F _~

E W simz where zf~ = undrained shear strength for cohesive soils u = presumed pore pressure at failure t~" = effective angle of internal friction c" = effective cohesion intercept. For the total stress analysis of the stability of embankments on soft soils, the normal procedure is that the material in the granular fill or granular layers in the subsoil is assumed to be drained and the soft soils undrained (Fig. 4.13).

~ ~176

~

e

.G,I

i|174c : Q " ......." ~W

"

"

c o s oC

Fig. 4.13. Stresses along the failure surface for granular and cohesive soils, which should be considered in total stress analysis.

In such a case, the available shear strength along a failure surface consists not only of undrained shear strength in the soft soils, but also of effective shear strength in the granular soils. When large deformations are expected, the friction angle as-

Stability Analysis

154

sumed for the granular fill material should be related to the critical state, i.e. shear strength at constant volume. In rough estimations of stability for low embankments, such as when using Terzaghi's equation (4.4), the shear strength in the fill material is neglected, as it is in Eqn. (4.10).

4.3.4

Simplified Bishop method

In the simplified Bishop method (Bishop, 1955) only the normal component E of the interslice forces is taken into account, and the variation in the tangential component T is ignored (Fig. 4.7). The resulting equation, from moment equilibrium about the centre of rotation, for the safety factor F is: 9 for the total stress analysis

F=

(4.12) X W sinc~

9 for the effective stress analysis Z[(W- u b) tan~" + c'bl/m~ F=

(4.13) Z W sinc~

where m~ = coscz (1 + tanc~ tank'/F) Since F appears on the left as well as the fight hand side in the equation, calculations start with an assumed value of F and a new value of F is calculated. With this new value, the calculations are repeated and such iterations proceed until there is little difference in successive values of F. An example of estimation of the safety factor for the 3rd stage of a test fill at the Antoniny site is shown in Fig. 4.14 and in Table 4.1. In this example, the organic subsoil was divided into three zones with different magnitudes of undrained shear strength. The soils in these zones were divided into layers with characteristic undrained shear strengths. In the calculations, the effective shear strength in the fill material was also taken into account. The calculations were made for a third stage of an embankment when the soil below was partially consolidated for the previous stages. The undrained shear strength has therefore become unevenly distributed across the embankment and the centre of the critical circle is not necessarily located above the midpoint of the slope.

Methods of Stability Analysis

155

t

I

R=20m

Zone

,

r

C

S6nd.

~ '

~ "

"

Gyttja

Peat

A

16

12

B

13

10

C

6

8

" "..'.

" " ::

I

Gyttja

(kPa)

fu

~ ""

. "..

" . i

"'.

. .-'.

Fig. 4.14. Example of estimation of the safety factor for the 3rd stage of a test fill at the Antoniny site based, on the simplified Bishop methods.

Table 4.1.

Example of estimation of the safety factor for the 3rd stage of a test fill at the Antoniny site.

Slice W No.

r

122 370 202 448 415 181 316 249 84 56

I

Wtan~" m a

Assumed F = 1.25 WtanO'/mot m a WtanO'/mot or

(o) 1 2 3 4 5 6 7 8 9 10

Wsin(x Xfu

48.6 36.9 26.4 17.5 5.74 -2.87 -11.5 -23.6 -33.3 -40.9

~ 91.5 222.2 89.8 134.7 41.5 -9.1 -63.0 -99.7 -46.1 -36.7 E 325.1

(1) F = 393.5/325.1 = 1.21

(kPa) (m)

(k~

14.8 10 10 10 10 8 8 8 6

85.4 -

5.1 5.4 1.8 4.2 4.1 2.0 4.1 4.3 2.4 4.5

or

xf. ~) 1.08

(2) F = 392.7/325.1 = 1.21

79.1 80 18 42 41 20 32.8 34.4 19.2 27 E 393.5

Xfu~ 1.09

78.3 80 18 42 41 20 32.8 34.4 19.2 27 E 392.7

Stabili .ty Analysis

156

4.3.5

Janbu generalized method

In the Janbu generalized method (Janbu, 1973), which can be used for slip surfaces of arbitrary shape, both components of interslice forces are taken into account by making an assumption that the points at which the interslice forces act can be defined from a "line of thrust" (Fig. 4.15). The position ofthe line of thrust is defined by vertical distances from the base of the slice to this line i.e. t c on the left side and t R on the right side of the slice, and angle c~t between the line of thrust on the right side of the slice and the horizontal.

w

E

-.

,,

line

of

thrust

b/2 tan o( l/tl_/ / / / / /

/

/

Fig. 4.15. Forces acting on each slice in the Janbu generalized method (Janbu, 1973).

The normal force N on the base of the slice is derived from the SUlnmation of vertical forces: 9 for the total stress analysis N = [W - (T R - T L ) - -CfuI sin~/F]/cosc~ (for cohesive soils)

(4.14)

Methods of Stability Analysis

157

9 for the effective stress analysis N = [W - (T R - T L ) - c ' l smt:x/F + u 1 tan~' sinct/F]/moc

(4.15)

The safety factor equation based on force equilibrium derived by summing forces in the horizontal direction is" 9 for the total stress analysis

X%b F=

(4.16) Z N sint~

9 for the effective stress analysis Z [(N - u 1) tan~" + c" 1] cosct F=

(4.17) E N sin~

Forces, directions and their location are evaluated by iterative procedure.

4.3.6

Janbu simple routine procedure

The Janbu simple routine procedure (Janbu et al., 1956) assumes that in Eqn. (4.14) and (4.15), where the normal force N is defined, the interslice tangential forces can be ignored. A correction factor fo accounts for the effect of the interslice tangential forces and is used to correct the initial safety factor F o. The correction factor is related to cohesion, the angle of internal friction and the shape of the failure surface cl~ (Fig. 4.16). An initial safety factor is calculated on the basis of Eqn. (4.16) for total stress analysis or Eqn. (4.17) for effective stress analysis using a reduced form of Eqn. (4.14) or (4.15). The corrected safety factor is obtained as: F = f o Fo

(4.18)

15 8

Stability Analysis 1.2

i

1

i

1

i

I

i

L

O ~--

_

.s 1.1 _ s u r f a c e

r = 0

U 0

LL

-

_

-

5......-- c=0 1.0

0

I 0.1

~ Ratio

I 0.2

I

I 0.3

O.Z,

d/L

Fig. 4.16. Correction factor f0 as function of curvature ratio d/L and type of soil (Janbu

4.3.7

et al., 1 9 5 6 ) .

Morgenstern-Price

method

The Morgenstem-Price method (Morgenstern and Price, 1965) which is the general approach of limit equilibrium analysis assumes that there is an arbitrary mathelnatical function to describe the direction of the resultant of the interslice tangential component T and nomlal component E of the following form: T/E = )v f(x)

(4.19)

where f(x) = a function that describes the rammer in which T/E varies across the slope X = a constant representing the portion of the function f(x) used when solving for the safety factor The equations for the safety factor are obtained using the summation of forces tangential and normal to the base of a slice and the SUlmnation of moments about the centre of the base of each slice. The equation for the safety factor at force equilibrium is the same as Eqn. (4.16) in the Janbu generalized method for total stress analysis and as Eqn. (4.17) for effective stress analysis. The application of the Morgenstem-Price method unavoidably involves the use of an appropriate numerical programme which, as well as more details concerning the Morgenstem-Price method, is out of scope of this book.

Methods of Stability Analysis 4.3.8

159

N o n - c i r c u l a r slip s u r f a c e

In soils containing weak layers of limited thickness or for embankments with pressure berms, the safety factor for the case of non-circular failure surfaces has to be considered. Both the Janbu method and the Morgenstem-Price method can be used for failure surfaces of arbitrary shapes. Fredlund et al., (1981) presented another method of slices commonly used for non-circular composite slip surfaces. The composite slip surface starts and ends with a circular portion and has a central linear portion (Fig. 4.17).

i R/

\\

Fig. 4.17. Forces acting in the method of slices with composite slip surface.

In such a case, the equation of the two-dimensional safety factor for moment equilibrium in the methods of slices presented above becomes: 9 for the total stress analysis Z%

1R

F-

(4.20) ZWx-ZNf

9 for the effective stress analysis ~ [ ( y - u l) tan~" + c'l] R V=

(4.21)

ZWx-ZNf

Stability Analysis

160

where R = radius or the moment arm associated with mobilized shear forces Sm x = horizontal distance from the slice to the centre of rotation f = perpendicular offset of the normal force from the centre of rotation In the case when the critical failure surface has a long horizontal portion, the wedge method may be used. The shape of the wedge failure surface can be estimated as in Fig. 4.18.

9

or

9

9

9

9

9

9

9

9

I

9

S a n d

~

I .

9

~

o

"

o

.

.

.

9

.

~

.

.

.

.

.

~

9

o

.

"9

~

9

9

~

9

"-

" "

~

I

oq>

9

-

"

'

"

.

."

.

"

.

I

.

"

9

"

"

~

,

,

9

9

"

. . . .

"i

Fig. 4.18. Failure surface in wedge method of stability analysis.

4.4

STABILITY

OF SINGLE-STAGE

EMBANKMENT

When the initial shear strength of the soil is sufficient to ensure the required stability for the maximum embankment load, the embankment can be raised to the total height in one stage. In most cases with organic soils, only low embankments can be constructed in a single stage because of the low initial undrained shear strengths. The permeability of organic soils becomes relatively low regardless of whether the initial value is high or not as it decreases significantly at load application and subsequent compression. If the permeability is low with reference to the rate of loading, the stability analysis of an embankment during construction should be carried out as an undrained analysis. At loading of overconsolidated soils, some partial drainage may occur, whereby the effective stresses increase. When the vertical stress reaches the preconsolidation pressure, no further increase in effective stress will take place from a practical point of view. Additional loading brings about increases in pore pressure equal to the increase in total vertical stress and in those zones of the subsoil where the principal stresses rotate, the increase in pore pressure may be even higher.

Stability of Single-Stage Embankment

161

A drained analysis assumes that all excess pore pressure has dissipated. An effective stress analysis using measured or predicted pore pressures often does not account for the further generation of pore pressure at loading up to failure (Fig. 4.19). Therefore, modified total stress analysis so-called Undrained Strength Analysis (Ladd 1991) using undrained shear strengths should be used. This entails estimation of the in situ undrained shear strength 1:fu. Methods commonly used for determination of undrained shear strength, summarized in Chapter 4.2, are presented in Chapters 2 and 3. The estimation of undrained shear strength on the basis of empirical relations is discussed in Chapter 4.5.2.

~ d

: c'+ (G~p*AG) tan r

. . . . . _._. ~ . . ~ . '

.Tff = c', 6"vptan r

-

/

b., d ul

Effective Stress Anatysis

/

/

. .~ . . .

~,~..._~

~),

'

/ /Drained

f-L. rL/3

Undrained Strength Anatysis

..__

Effective norma[ stress, '6' Fig. 4.19. Assumed shear strength at failure r in Drained Analysis, Effective Stress Analysis and Undrained Strength Analysis (Ladd 1985b).

For preliminary studies, rapid stability estimations for embankments during construction can be made using Terzaghi's equation or Taylor's stability chart. For final design, an appropriate method of slices should be applied using the undrained shear strength m layers with organic soils or other soft low-permeable soils. In undrained strength analyses, the required safet7 factor during the construction period is typically of the order of 1.3-1.5, although values as low as 1.2 are sometimes allowed when the site conditions and undrained shear strength are very well established. Much higher values, in the order of 2 or more, are employed if significant shear deformations must be avoided. In design of an embankment, also other loading conditions with different acting forces and drainage conditions than the end-of-construction conditions may have to

Stability Analysis

162

be checked. For constructions such as flood control levees, pond dykes and waterretention dams, additional stability analyses for steady state seepage conditions and conditions of sudden drawdown of the water level should be carried out. For preliminary studies, stability charts presented in Chapter 4.3.2 can be used. For final design, effective strength analyses with computed pore pressures should be made. In analysing stability of road embankments, additional loads arising from road construction and traffic should be considered. The influence of additional loads on embankment stability is significant for low embankments on very soft soils.

4.5

STABILITY OF STAGE-CONSTRUCTED EMBANKMENTS

4.5.1

Types and scope of analysis

When the stability analysis for the end-of-construction stage indicates that the initial shear strength of the soil yields an inadequate safety factor, construction may be divided into two or more stages. The estimation of the maximum load for the first stage is then carried out as for a single-stage embankment, but the stability analysis for the following stages of construction requires prediction of shear strength increase developed with time due to partial consolidation. For most cases of stage-constructed embankments, the most critical stability condition occurs during actual construction. All stability calculations for stage-constructed embankments during construction should be carried out as undrained analyses. However, as for single-stage embankments, the stability for conditions after construction should also be ensured. Different levels of sophistication can be employed in estimating and using the increase in undrained shear strength in design of stage-constructed embankments. Firstly, they depend on the method of stability analysis and secondly on the method used to estimate the initial shear strength and the increase in shear strength during consolidation (Ladd, 1991). The methods for estimation of the increase in shear strength can be divided into two groups of sophistication and expense : 9 empirical relations, and 9 laboratory testing to obtain complete strength-stress data.

Stability of Stage-Constructed Embankments

4.5.2

163

Evaluation of the increase in undrained shear strength based on empirical relations

Past experience, summarized by Larsson (1980) and Ladd (1985b), shows that for soft normally consolidated mineral soils the following empirical relation can be used to predict shear strength increase: 1:f. = K "o'"v

(4.22)

where zf, = undrained shear strength K = coefficient of shear strength increase (depending on the loading case) (r"V = vertical effective stress. For overconsolidated soils, the coefficient of shear strength increase (Ladd, 1985b) can be expressed as: K = S (OCR) m

(4.23)

where S = zf,/cr' - normalized undrained shear strength for the normally consolidated state (OCR= 1) OCR = C'p/6'v - overconsolidation ratio cr p = preconsolidation pressure m = slope of l:f~/6"v - OCR relationship on log-log scale. Examples of values of coefficients S and m for soft mineral soils presented by Jamiolkowski et al. (1985) are shown in Fig. 4.6. The coefficient of shear strength increase for normally consolidated soft mineral soils varies with the plasticity, but can be assumed as constant for a given soil (Larsson, 1980; Ladd, 1985a) (Fig. 4.20): K = constant = S

(4.24)

For normally consolidated mineral soils, the undrained shear strength can then be calculated as: 1:f~ = S - ~ p

(4.25)

Stability Analysis

164 .4 . . . . &&

~- 0.3

_

lX

IX

~

I,.

Ix

TC &

IX

-

&

&

IX

O

oiX

IC

'-- 0.2

&

0

0

0

v

DSS

-~.... ~

_ _ _

_ - -

~

- - -

~--o

v

~ g~v~--" "~ V

.N _

,~ Triaxial compression (TC). o Direct simple shear (DSS). v Triaxial extension (TE).

0.1

E L

O Z

0C

I

10

.-

l

20

I

30

1

1,0

,1

50

I

60

Tfu =qf 'f'fu ='trh(max) Tfu = qf 1

70

I

80

90

Plasticity index, Ip (%) Fig. 4.20. Normalized undrained shear strength versus plasticity index (Ladd, 1985a).

Similar relations have been found for organic soils (Larsson 1990), and the normalized shear strength then varies with the organic content, (Fig. 4.21). Experience from organic soils (Bergdahl et al., 1987) indicates that the normalized undrained shear strength changes not only in the overconsolidated states but also in the normally consolidated state. For a better description of the change in undrained shear strength with stress, especially in the normally consolidated state, the concept of the effective stress level ESL (Lechowicz, 1986) was introduced: ESL = (O"p)o/(Y'v

(4.26)

where (6'p)o = initial preconsolidation pressure. The increase in undrained shear strength with effective stress level ESL can be estimated according to Eqn. (4.22) for which the coefficient of shear strength increase is obtained from the following relations"

165

Stability of Stage-Constructed Embankments

3 0. 1

~,,-.

/

e/ I

c O./.,

o

/

I,,. ,.m,-,,

~."

K.,,

u~ 0.3

~f s~

70

/

&,-

"0 C

/ ,or

l

l

9Triaxia[ compression o Triaxial extension o Direct simple shear

Mineral soil

"o 0.1

o._Eganic - Mineral soil

N

Z

/

O

.c_ a 0.2-

L.

0

l

0

JE

E o 0.0

9

/

....

/__

-T

7-

mineral-organic soil

organic soil

l-

20

8 Organic

3b

content,

(%)

~o>0.5 ;9

t..,'+-

.E

,-- O.Z, C I...

~O

4

0.3

/

I I

0.2. OJ

-o 0.1

Ba _,,1o

//

~ ~ " ~

0

"~ " -

m

m

0

o

O/

/

/

l

/

//

l

/

/

115~, /

/

/

/

/

0

Empirical relations for inorganic clays (Larsson 1980)

N

o~

O

E o00 z

0 ,..,..~.

t /

JE cu~ O K_ "10 c

0

I

0

2Go

Liquid limit, (%)

9Triaxia[ compression o Triaxia[ extension o Direct simple shear

3oo

Fig. 4.21. Normalized undrained shear strength in organic soils (a) as a function of organic content (b) as a function of liquid limit. (Larsson 1990).

Stability Analysis

166 K = S (ESL) m~162

ESL> 1

K = S (ESL) ~r

ESL_ I

Sym

Soil

~c~ 3u~

I

I

~'-

0.1

I

0.3

0.2

1

I

i

0.4 0.50.6

I

I

I

I

I

0.8 1

2

1

3

4

5

J

I

6 7 8

10

Effective stress level ESL = (6;]o/6"v Fig. 4.23. Normalized undrained shear strength versus effective stress level from in situ and laboratory tests (Bergdahl et al., 1987). Undrained

Vertical effective stress. (kPa) 0

0

\

'

=-

'

40 '

'

/

~

.

0

60 ' I

'

, .,~vo I ~ I/ / 6"v before ' I , ' 1(6P )o I / stage 2

I

-Q_

20 '

~

.

shear strength,

Tfu (kPa) 10 2O

/

-..-.-I

~~,. ~'t

31

~ ~ o

.~_

56-~

7~

,r i

I

- -

',

I I -

-/- 4 --

-

-

a

..

-?--4

k]

Fig. 4.24. Calculated effective stresses and predicted shear strengths at different stages of a test fill. Assumed values of coefficients; Peat: S=0.5, moc=0.85, m,c=0.2; Calcareous soil: (1) S=0.45, moc=0.85, mnc =0.2; (2)S=0.45, moc=0.8, mnc=0.2.

Stability of Stage-Constructed Embankments

169

"I'fu = 0.45 9if;

O

3 II

1_o :3

2

0.2 1"fu = 0.45.6"v (ESL)

1

'Tfu = 0.45.6'v (OCR) 0.8 I

o o12 0:5

;, I

,

i

~

i

o:s

3

2,

o13

ESL= (6"~))o16'{,

i

Fig. 4.25. Predicted undrained shear strength versus effective vertical stress (Wolski et al., 1988).

Finally, it should be mentioned that a more global description of currem stress state includes both vertical and horizontal effective stresses (Wroth and Houlsby, 1985; Becker et al., 1987; Lechowicz 1994). Therefore, it would be better to define the increase in shear strength as a yield surface changing in both shape and orientation in stress space (Runesson, 1978; Larsson and Sfillfors, 1981). In this way, a more accurate description of the stress state and the undrained shear strength would be obtained. To estimate undrained shear strengths to be used in the stability analysis, the subsoil is divided into several zones. The division of the subsoil into zones should model the layering of different types of soil, the distribution of effective stresses and the overconsolidation ratio. Such a system of zones may be further elaborated by allowing for the effects of principal stress rotation on the distribution of undrained shear strength.

4.5.3

Evaluation of the increase in undrained shear strength based on laboratory testing

The undrained strength-deformation behaviour of soils below and outside the loaded area during stage-construction can be simulated by using three types of lab-

Stability Analysis

170

oratory tests. Undrained shear strength for different areas beneath an embankment can be estimated from: 9 triaxial or plane-strain compression shear tests for the area directly beneath the embankement; 9 simple shear tests for the area where the shear surface is nearly horizontal usually below the slope of the embankment; 9 triaxial or plane-strain extension shear tests for the area outside the toe of the embankement. The tests require detailed knowledge of the initial stress history of the soil and when they are to be used for prediction of future increases in shear strength, an accurate prediction of the future stress history is also essential so that real soil conditions can be simulated. Based on the knowledge of existing and expected stresses in soil, shear strength tests should be carried out on soil samples with prevailing stresses and with expected future stress levels. In this way, the normalized stress-strength relations for current and future stress levels are obtained and undrained shear strength parameters S and m may be evaluated. A simple approach for estimating the shear strength increase m soft subsoils beneath embankments was proposed by Aas (1976) and Larsson et al. (1984). The failure surface is divided into three different segments. The increase in shear strength for each of them is evaluated in a different way as shown in Fig. 4.26. For the steep part of the slip surface beneath the embankment, the shear strength increase is estimated from plane-strain compression or triaxial compression tests and for the next flat part beneath the embankment out to the mid-point of the slope the increase is estimated from DS S tests.

J

T "

A'rfu =0 ~

"

!

9

~176176 ". ~ ~

~ o

~ ~

~ " ,

i ~

!

6v: (Gv)o*"6v

~6p

.6;= 6v C%1o

BV

Fig. 4.26. Simplified estimation of increase in shear strength due to consolidation (Larsson et al., 1984).

Other Approaches in Stability Analysis

171

Most geotechnical laboratories do not possess plane strain devices, but triaxial tests are routinely used. The simplest approach to the estimation of initial strength and the prediction of shear strength increase for evaluation of stability is to use the data from CKoU direct simple shear tests for the whole failure surface.

4.6

OTHER APPROACHES IN STABILITY ANALYSIS

4.6.1

Three-dimensional analysis

The methods of stability analysis presented above are based on the assumption of two-dimensional (plane strain) geometry. In most practical problems, the length of the embankment is limited and, within this length, the geometry, load and shear strength are also often not uniform. For a more economic and accurate design, the stability of embankments with complex geometries and variable soil and pore pressure conditions may be estimated with three-dimensional analyses (Azzouz et al., 1981). In such cases, instead of the infinite cylinders assumed in two-dimensional circular arc analysis, the stability analysis may consider failure surfaces consisting of cylinders with ends of different shapes (Baligh and Azzouz, 1975) (Fig. 4.27).

al

b)

c)

Fig. 4.27. Cylindrical failure surface with different ends" (a) plane end surface (b) conical end surface (c) ellipsoidal end surface (Baligh and Azzouz, 1975).

The simplest way to account for the end effects is to assume a circular slip surface with plane ends (Gens et al., 1988). To calculate the three-dimensional safety factor, the resisting moments from each end plane are then added to those from the cylindrical surface (Fig. 4.28). In this case, the three-dimensional factor of safety F3D can be expressed in terms of the two-dimensional factor F2D (Gens et al., 1988) as:

Stability Analysis

172

F3D = F2D (1 +

(4.28)

0) Ls

where Ro

L= 0 ME AQC

= = = = =

radius of the cylindrical part of the failure surface overall length of the slide 2 M E/(AQC 9Ro2 ) resisting moment of area of each end plane failure surface.

b)

o)

'/ /;J

0

d

H

Fig. 4.28. Geometry of three-dimensional cylindrical slide with plane ends. (a) oblique view (b) cross-section (Gens et al., 1988).

The assumption of plane ends generally overestimates the end effect because the most critical case involves curved ends. Gens et al. (1988) have analyzed the end effects using a large variation of the shape of the end surfaces. Minimum safety factors were obtained using a family of power curves to generate the end shapes. According to Gens et al. (1988) the errors involved in neglecting end effects by treating slope stability problems two-dimensionally can be as high as 30 %. When the considered length is small, the three-dimensional safety factor is always significantly greater than the two-dimensional factor. An extension of the simplified Bishop method of slices to three-dimensional analysis was presented by Hungr (1987). Thus the existing computer programs for twodimensional analysis can be easily modified for three-dimensional analysis. In this approach, the sliding body is divided into a series of vertical columns of a rectangular cross-section. The base shape of the individual column results from the compound shape of the sliding body consisting of cylindrical and semi-ellipsoidal parts (Fig. 4.29).

Other Approaches in Stability Analysis

173

z'

z[

c)

d) Axis of rotation 9- - - . I - T - , . ; I i:l =

I I

I

I ; I I ', I

--r-----T

J I I

Axis of rotation

9- - r -

~ , I ,/ I ."

a I I, ' ] I ! I./Crest ~-

H-

Crest

Ic

E V) ,...

o

~X

~"

/

1

Toe

E: o Q.

x

y

Fig. 4.29. Failure surface for three-dimensional simplified Bishop method.(a) one half of the sliding body (b) forces acting on a single c o l u m n (c) vertical cross-section of the sliding body in the x-z plane (d) and in the y-z plane (Hungr, 1987).

The resulting equation from moment equilibrium for the safety factor F3D is: 9 for the total stress analysis Zzf~ A F3D =

*

X,W~ sin%

(4.29)

for the effective stress analysis Z[(W o - u b A cOSyz) tanr + c" A cOSyz]/M= (4.30)

F3D =

x w o sin%

17 4

Stability Analysis

where W ~ = total weight of the column A

=Ax Ay (1 - sin20~x sin2C~y)~

coSC~y) - true area of the column base

cOS~z = [ 1/(tan2C~y + tan2% + 1)]~ Mu

=cOSyz [1 + (sin~y tan@')/(F cOSyz)]

ub

= pore pressure acting in the centre of the column base

An example of the use of the simplified Bishop method modified for three-dimensional analysis to estimate the safety factor of test fill during the failure test is shown in Fig. 4.30. The fill constructed in the failure test had approximately the shape of a truncated pyramid. Half of the sliding body of the fill was divided into cylindrical part 1o wide and a semi-ellipsoidal part 1 wide. Both parts were divided into a series of colunms arranged in rows of uniform width. The organic soils were divided into zones with different magnitudes of undrained shear strength created by previous load stages. The soils in these zones were also divided into layers with specified undrained shear strength. In stability calculations, the shear strength in the fill material was also taken into account. The three-dimensional calculations of stability for the final stage in the failure test indicate that the factor of safety was about 1. The two-dimensional calculations carried out for the cross- section of the cylindrical part yielded a safety factor of about 0.75. F3D =1.04 Zone of subsidence \ ",,

~ ~ r

F

o~k ,~,"1,I, I./{.

'",N

IJ

CyLindricaL

I \part

Fig. 4.30.

Y

Z~B-B

iA

.....

I F

!i/,~ 7-" Crack ari~. due to failure 1X

F20 =0.TS

Zone of heave

I/c-I1! c.I

~4 g: ~

BIX

zl A-A

Axis of l/rotation

_J._\Semi-ellipsoidal - - part

u

Z C-C

9 9

.

9 9

~

~

Y

~ v

Example of estimation of the safety factor for test fill in the failure test based on the three-dimensional simplified Bishop method.

175

Other Approaches in Stability Analysis

4.6.2

Stability assessment by finite element analysis

Stability evaluations based on the limit equilibrium methods have an artificial separation from the stress distribution-consolidation analysis. The latter is used to predict subsequent changes in shear strength at certain stages during construction, after which the equilibrium analysis is used to calculate the safety factor at that stage. Numerical analyses such as the finite element method potentially offer a possibility to evaluate embankment stability as a coupled continuous effect of stressstrain-strength-time analysis. Information on the state of stresses obtained from the finite element analysis is then used to evaluate the stability conditions. Because the normal effective stresses 6"N and shear stresses ~N in each element are calculated, a local safety factor may be calculated as the ratio of the local shear strength to the local shear stress (Chen and Chameau, 1982): C"-F(Y"N tan(~"

FN =

(4.31)

The mean safe~y factor for the assumed failure surface may be obtained as the ratio of the total shear strength to the total shear stress as follows: +

tan ') a h

F=

(4.32) Z"[ N dA

where Z = summation over the whole failure surface dA = bottom area of a vertical column. There are other ways, based on the finite element analysis, in which the stability conditions may be estimated (Chowdhury, 1978). One of them is the evaluation of stability conditions on the basis of propagation of plastic zones (Lo and Lee, 1973; Lacasse et al., 1977; Almeida and Ramalho-Ortigao, 1982; Teunissen et al., 1986). Stability conditions have also been evaluated on the basis of the development of effective stress paths related to yield and failure surfaces (Tavenas and Leroueil, 1977; Tavenas et al., 1978; Folkes and Crooks, 1985; Crooks 1987; Jardme and Hight, 1987).

Stability Analysis

176

4.7

REFERENCES

Aas, G. (1976). Totalsp~inningsanalyser, prinsipp, gmnnlag. Norske Sivilingeniorers Forening. Kurs i Jordartsegenskaper- bestemmelse og anvendelse. Gol 2022/5. Aas, G., Lacasse, S., Lunne, T. and H6eg, K. (1986). Use of in situ tests for foundation design on clay. Proc. of In Situ '86, a Speciality Conference on Use of In Situ Tests in Geotechnical Engineering, Blacksburg, Virginia. ASCE Geotechnical Special Publication 6, New York, pp. 1-30. Almeida, M.S.S. and Ramalho-Ortigao, J.A. (1982). Performance and finite element analyses of a trial embankment on soft clay. Proc. International Symposium on Numerical Models in Geomechanics, Ziirich, pp. 548-558, A.A. Balkema. Azzouz, A.S., Baligh, M.M. and Ladd, C.C. (1981). Three-dimensional stability analyses of four embankment failures. Proc. 10th International Conference on Soil Mechanics and Foundation Engineering, Stockholm, Vol. 3, pp. 343-346. Azzouz, A.S., Baligh, M.M. and Ladd, C.C. (1983). Corrected field vane strength for embankment design. Journal of the Geotechnical Engineering Division. ASCE, Vol. 109, No. GT5, pp. 730-734. Baecher, G.B. and Ladd, C.C. (1985). Reliability analysis of the stability of embankments of soft clays. Proc. for MIT Special Summer Course 1.60S, Recent Developments in Measurement and Modeling of Clay Behavior for Foundation Design, Lecture 1. Baligh, M.M. and Azzouz, A.S. (1975). End effects on stability of cohesive slopes. Journal of the Geotechnical Engineering Division, ASCE, Vol. 101, No. GT11, pp. 1105-1117.

Becker, D.E., Crooks, J.H A. and Been, K. (1987). Interpretation ofthe field vane test in terms of in situ and yield stresses. International Symposium on Laboratory and Field Vane Shear Strength Testing, Tampa, Florida, ASTM STP 1014.

Bergdahl, J., Hartlen, J., Larsson, R., Lechowicz, Z., Szymanski, A. and Wolski W. (1987). Shear strength increase in organic soils due to embankment loading. 8th Krajowa Konferencja Mechaniki Gruntow i Fundamentowania, Wroclaw, Vol. 1, pp. 21-32. Bishop, A.W. (1955). The use of the slip circle in the stability analysis of slopes. Geotechnique, Vol. 5, No. 1, pp. 7-17. Bjerrum, L. (1972). Embankments on soft ground. Proc. ASCE Specialty Conference on Performance of Earth and Earth Supported Structures, Purdue University, Lafayette, Indiana, Vol. 2, pp. 1-54.

References

177

Chen, R.H. and Chameau, J.L. (1982). Three-dimensional slope stability analysis. Proc. 4th International Conference on Numerical Methods in Geomechanics, Edmonton, Canada, Vol. 2, pp. 671-677. Chowdhury, R.N. (1978). Slope analysis. Developments in Geoteclmical Engineering, 22, Elsevier, Amsterdam-Oxford-New York. Cousins, B. F. (1978). Stability charts for simple earth slopes. Journal of the Geotechnical Engineering Division, ASCE, Vol. 104, No. GT2, pp. 267-279. Crooks, J.H. (1987). Some observations on the stability of structures founded on soft clays. Proc. International Symposium on Prediction and Performance in Geotechnical Engineering, Calgary, pp. 27-38. A.A. Balkema. Folkes, D.J. and Crooks, J.H.A. (1985). Effective stress paths and yielding in soft clays below embankments. Canadian Geotechnical Journal, Vol. 22, No. 3,pp. 357- 374. Fredlund, D.G. and Krahn, J. (1977). Comparison of slope stability method of analysis. Canadian Geotechnical Journal, Vol. 14, No. 4, pp. 429-439. Fredlund, D.G., Krahn, J. Pufahl, D.E. (1981). The relationship between limit equilibrium slope stability methods.Proc. 10th International Conference on Soil Mechanics and Foundation Engineering, Stockholm, Vol. 3, pp. 409-416. Gens, A., Hutchinson, J.N. and Cavounidis, S. (1988). Three-dimensional analysis of slides in cohesive soils. Geotechnique, vol. 38, No. 1, pp 1-23. Golebiewska, A. (1976). An application of vane shear testing in organic soils. Ph. D. Thesis, Warsaw Agricultural University (in Polish) Golebiewska, A. (1983). Vane testing in peat. Proc. 7th Danube European Conference on Soil Mechanics and Foundation Engineering., Kishinev, Vol. 1, pp. 4953. Hungl, O. (1987). An extension of Bishop's simplified method of slope stability analysis to three dimensions. Geoteclmique, Vol. 37, No. 1, pp. 113- 117. Jamiolkowski, M., Ladd, C.C., Germaine, J.T. and Lancellotta, R. (1985). New developments in field and laboratory testing of soils. Theme lecture. Proc. 1l th International Conference on Soil Mechanics and Foundation Engineering, San Francisco, Vol. 1, pp. 57-153. Janbu, N., Bjerrum, L. and Kjaernsli, B. (1956). Veiledning vid losning av fundamenteringsoppgaver. Norwegian Geotectmical Institute, Publication 16, Oslo. Janbu, N. (1973). Slope stability computations. Embankment-Dam Engineering Casagrande volume. John Wiley and Sons, New York.

17 8

Stability Analysis

Jardine, R.J. and Hight, D.W. (1987). The behaviour and analysis of embankments on soft clay. Special Publication on Embankments on soft soils, Bulletin of the Public Works Research Center, Athens, pp. 33-158. Lacasse, S.M., Ladd, C.C. and Barsvary, A.K. (1977). Undrained behaviour of embankments on New Liskeard varved clay. Canadian Geotechnical Journal, Vol. 14, No. 3, pp. 367-388.

Ladd, C.C. (1985a). Overview of clay behavior. Proc. for MIT Special Summer Course 1.6OS, Recent Developments in Measurement and Modeling of Clay Behaviour for Foundation Design, Lecture 2. Ladd, C.C. (1985b). Stability evaluation for staged construction, Proc. for MIT Special Summer Course 1.6OS, Recent Developments in Measurement and Modeling of Clay Behaviour for Foundation Design, Lecture 15.

Ladd, C.C. (1991). Stability evaluation during staged construction. The 22nd. Karl Terzaghi Lecture. Journal of Geoteclmical Engineering, ASCE, Vol. 117, No. 4, pp. 540-615.

Landva, A.O. (1980). Vane testing in peat. Canadian Geotechnical Journal, Vol. 17, No. l, pp. 1-19.

Landva, A.O. (1986). In-situ testing of peat. Proc. of In Situ '86, a Speciality Conference on Use of In Situ Tests in Geoteclmical Engineering, Blacksburg, Virginia, ASCE Geotechnical Special Publication 6, pp. 191-205.

Landva,,A.O., Pheeney, EE., La Rochelle, P. and Briaud, J.L. (1986). Structures on peatland - geotechnical investigations, Proc. Advances in Peatlands Engineering, Ottawa, 31-52. Larsson, R. (1980). Undrained shear strength in stability calculation of embankments and foundations on soft clays. Canadian Geotechnical Journal, Vol. 17, No. 4, pp. 591-602. Larsson, R. (1986). Consolidation of soft soils. Swedish Geotechnical Institute, Report No.29, Link6ping. Larsson, R.(1990). Behaviour of organic clay and gyttja. Swedish Geotechnical Institute, Report No.38, Link6ping. Larsson, R. and S~illfors, G. (1981). Hypothetical yield envelope at stress rotation. Proc. 10th International Conference on Soil Mechanics and Foundation Engineering, Stockholm, Vol. 1, pp. 693-696. Larsson, R., Bergdahl, U. and Eriksson, L. (1984). Evaluation of shear strength in cohesive soils with special reference to Swedish practice and experience. Swedish Geotechnical Institute, Information No.3, 32 pp. Link6ping. Also in shorter version in ASTM Geotechnical Testing Journal, Vol. 10, No. 3, 1987.

References

179

Lechowicz, Z. (1986). Evaluation of the increase in shear strength in organic subsoil loaded by embankanent. Speciality Conference on Land Reclamation, Land Reclamation Facult~ Warsaw Agricultural University, 3, 77-84 (in Polish). Lechowicz, Z. (1994). All evaluation of the increase in shear strength of organic soils. Proc. International Workshop on Advances in Understanding and Modelling the Mechanical Behaviour of Peat, Delft, pp. 167-179, A.A. Balkema. Lechowicz, Z., Szymanski, A. and Wolski, W. (1984). Consolidation-strength analysis for soft soils. Proc. Sedimentation/Consolidation Models, ASCE, San Francisco, 107-120. Lo, K.Y. and Lee, C.F. (1973). Analysis of progressive failure in clay slopes. Proc. 8th International Conference on Soil Mechanics and Foundation Engineering, Moscow, Vol. 1.1, pp. 251-258. Lunne, T. and Kleven, A. (1981). Role of CPT in North Sea Foundation Engineering. Proc. Cone penetration testing and experience, St Louis, MO. Also in Norwegian Geotechnical Institute, Publication No. 139, Oslo. pp. 1- 14. Mayne, P. W. and Mitchell, J.K. (1988). Profiling of overconsolidation ratio in clays by field vane. Canadian Geotechnical Journal, Vol. 25, No.l, pp. 150- 157. Morgenstern, N.R. (1963). Stability charts for earth slopes during rapid draw down. Geoteclmique, Vol. 13, No. 2, pp. 121-131. Morgenstern, N.R. and Price, V.E. (1965). Analysis of the stability of general slip surfaces. Geotechnique, Vol. 15, No. 1, pp. 79-93. Runesson, K. (1978). On nonlinear consolidation of soft soils. Chalmers University of Tectmology, Department of Structural Mechanics, Publ. 78:1, Gothenburg. Spencel; E. (1967). A nlethod of analysis of the stability of elnbanklnents assuming parallel inter-slice forces. Geoteclmique, Vol. 17, No. 1, pp. 11-26. Tavenas, E and Leroueil, S. (1977). Effects of stresses and time on yielding of Clays. Proc. 9th International Conference on Soil Mechanics and Foundation Engineering, Tokyo, Vol. 1, pp. 319-326. Tavenas, E, Blanchet, R., Garneau, R. and Leroueil S. (1978). The stability of stage-constructed embankments on soft clays. Canadian Geoteclmical Journal, Vol. 15, No. 2, pp. 283-305. Tayloh D.W. (1956). Fundamentals of soil mechanics, Jolm Wiley and Sons, New York. Terzaghi, K. (1956). Theoretical soil mechanics, Jolm Wiley and Sons, New York. Teunissen, J.A.M., Bauduin, M.H. and Calle, E.O.E (1986). Analysis of failure of an embankment on soft soil: A case study. Proc. 2nd International Symposium on Numerical Models in Geomechanics, Ghent, pp. 617-626.

180

Stability Analysis

Whitman, R.V. and Bailey, W.A. (1967). Use of computers for slope stability analysis. Journal of the Soil Mechanics and Foundations Division, ASCE, Vol. 93, No. SM4, pp. 475-498. Wolski, W., Szymanski, A., Mirecki, J., Lechowicz, Z., Larsson, R., Hartlen, J., Garbulewski, K. and Bergdahl, U. (1988). Two stage-constructed embankments on organic soils. Swedish Geotechnical Institute, Report No. 32, Link6ping.

Wolski, W., Szymanski, A., Lechowicz, Z., Larsson, R., Hartlen, J. and Bergdahl,U. (1989). Full-scale failure test on stage-constructed test fill on organic soil. Swedish Geoteclmical Institute, Report No.36, Link6ping. Wright, S.G., Kulhawy, F.H. and Duncan, J.M. (1973). Accuracy of equilibrium slope stability analysis. Journal of the Soil Mechanics and Foundations Division, ASCE, Vol. 99, No. SM 1O, pp. 783-791. Wroth, C.P. and Houlsby, G.T. (1985). Soil mechanics-property characterization and analysis procedures: Theme Lecture. Proc. 1lth International Conference on Soil Mechanics and Foundation Engineering, San Francisco, Vol. 1, pp. 1-55.

181

Chapter 5

Analysis of Subsoil Deformations A. Szymanski, Department of Geotechnics, Warsaw Agricultural University

5.1

GENERAL

The design of embankments on organic soils involves estimation of the magnitude and rate of the subsoil deformations. In order to model the existing subsoil as well as the boundary conditions, two main cases are usually considered in these calculations (Fig. 5.1): case 1 - one-layer subsoil, and case 2 - multi-layer subsoil.

o)

J Xau=O Compressible layer &u=0 / ;" . . :" .".".".".".. ..-

i , _

., 120

.~.4. ~16~176

.;-'. -_.,1500

~

" l " ~ ""--''T . . . .

6

7

8

OF HUMIFICATION

H

Coefficient of secondary consolidation versus degree of humification in peat. (Larsson 1990). H = Degree of humification according to yon Post (1924). Permeability, 6 0.1

9

I

!

_

1

pllll

I

k (x 10 m/s)

!

1

95

l

~ ltrl'~

X

0

X

O" ..8..,

x

0

o L-

0

o

>

10

X X X

0

X X

o 1

100

ll'll

O

0

I

1

X

0

l

l

0

0

|

i

Xx

0

4

!

o e= f(k) I

I

1 I II

10

l

. I

l

I

I l[lJ

x e= f (6") 1_.

100

t

1

1

IllL

1000

Effective stress, 6" (kPa) Fig. 5.13.

Consolidation characteristics of peat from Antoniny (Yong et al., 1988).

Empirical relations for initial permeability and permeability change index for peat can be found in Chapter 3.7.

192

Analysis of subsoil deformations

5.3

ANALYSIS OF "FINAL" DEFORMATION

5.3.1

Type and scope of analysis

For embankments on a thin layer of soft subsoil, an evaluation of the final settlement is sufficient to design the embankment. In this case, the settlement is estimated by one-dimensional analysis or empirical formulae. When embankments are to be constructed on deep organic subsoil, a full description of the behavoiur of the soft subsoil is required to evaluate the performance of the embankment. In order to evaluate the vertical and horizontal displacements as well as the effective stresses in the subsoil, calculation methods accounting for two-dimensional deformations and changes in geometries during the consolidation process should be used (see Chapter 5.7). If the simplified solution of the consolidation theory is used, the one- dimensional deformation prediction should be supplemented by an estimation of the initial plastic movement occurring in the initial stage of loading because of shear and lateral deformations in the subsoil. The total settlements should be estimated as a sum of the initial movement Si and long-term displacement Sf created by primary consolidation S~ and secondary compression S~: S = S i q- Sf

(5.2)

or S = S i + Se+ Ss

(5.3)

The initial movement and the secondary deformation of the subsoil under an embankment play significant roles in the consolidation process in organic soils. The deformations can also be estimated by use of finite elements and theory of elasto-plasticity. The applied soil models may then for example be hyperbolic - strain hardening (elastic Duncan - Chang model, Duncan 1980) or Critical State models (Cam Clay model, Wroth and Houlsby 1980) for soils with isotropic properties or anisotropic models, (e.g. Runesson 1978), for soils with anisotropic properties. The selection of model also depends on the effective stress level and the expected soil displacements under the embankment, (range of elastic and plastic deformations). Elastic models cannot be used when the plastic deformations in the subsoil are large. In such cases elasto-plastic models should be applied.

Analysis of "Final" Deformation

193

Initial settlement and horizontal movement

5.3.2

If the applied construction load has a limited lateral extension, shear and lateral deformations occur in the subsoil, which results in initial vertical and horizontal displacements. Such deformations are mainly associated with undrained conditions. Therefore, the computation of initial settlement S i and horizontal movement Sh is often based on theory of elasticity using Poisson's ratio v = 0.5 and an undrained modulus of elasticity E u. The initial deformations calculated by the theory of elasticity become:

S~- Iv q B/E.

(5.4)

and Sh =I h q H/E u

(5.5)

where q = stress applied to the subsoil B = width of the loaded area I

= influence factor which depends on the geometry of the problem ( I and Ih refer to vertical and horizontal direction respectively)

E -

undrained modulus of elasticity,

H = thickness of the compressible layer. Application of this theory to the prediction of initial movements requires the estimation of modulus E from laboratory tests or from empirical correlations with the undrained shear strength. Larsson (1986) points out that the undrained modulus E values vary from 80 ~fu for organic soils to 2000 ~f~ for low-plastic clays. The results of investigations carried out by Foott and Ladd (1981) indicate that the undrained modulus for normally consolidated soils can be calculated from the empirical formula: E u = 1:f~215 lnF/Ip

(5.6)

where ~f~ = undrained shear strength from vane shear tests or direct simple shear tests, F = calculated factor of safety against shear failure, Iv -

plasticity index.

Analysis of subsoil deformations

194

When the soil conditions vary with depth, a harmonic mean value of E u is often used.

Calculation of the initial deformation in soft subsoil under an embankment with Eqn. 5.4 and 5.5 requires evaluation of the displacement influence factors Iv and Ih from the theory of elasticity. Diagrams for various conditions have been presented by Steinbrenner (1934), Janbu et a1.(1964) and Poulos (1972), Figs. 5.14, 5.15 and 5.16.

a

3.0 ..,, ......... .......... C,_,,~o,,~,, ~ L-length 2.0

H ~

,-

0.9

fo.

~ 21

Iv=)J1 )u 0

5

~ :o.s

1.o

o.%.~ 0

Fig. 5.14.

01

dL

1 l lllll]

1

J

I

lllllll

10 H/B

l

l

I IJJ|II

100

l

[

IJIIU

~o

~oo

Hf/B

~ooo

1000

Influence factor Iv for evaluating the vertical displacements in the centre of a loaded area (Janbu et al., 1964).

00

x/B

1

2

00

=

~,

Influence factor, I h 0.1

0.2

O.Z,

-0.1

N

0.6 0.8

0.2

.5

1.0~''_q._.

[ B , 9

,,-1

fil"~l ~liJ J I i l J

Sh:Ih

0.3

2

0.2

'E

qH/Eu }

I x_ H

/////////////i////////////////,4

Fig. 5.15.

~

Case Oistrib.

of

Eu

with depth

'

1

'

2

3

~] 1.5Eu

0.5Eu

Influence factor I h f o r evaluating the horizontal displacements under a uniform strip load (Poulos, 1972). (a) Horizontal surface displacements. (b) Horizontal displacements beneath edge.

195

Analysis of "Final" Deformation

]' 2

/ r--

I T-li~176176 I

!, I

!

.~ b "1" ]

I

61~

~ =1

I i

i,a

! !

10 .I 0

~l

i

I

I

0.2

0.4

0.6

0.8

fl,f2 Fig. 5.16. The coefficients fl and 1'2 for Steinbrenner's formula.

Using Janbu's diagram to estimate the influence factor Iv with good accuracy is difficult for small values of the H/B ratio. In such a case, therefore, the factor Iv can be calculated directly from the analytical solution of the elastic theory according to Steinbrenner (1934). The settlement under a comer of a uniform strip load is expressed as:

(5.7)

(q B/E u) ((1-v2)fl + (1- v - 2v 2) f2)

Sic =

where"

fl = ~

1 fL

-In

B

(1+~1)~C1+Ch-2 + In((L/B)+'~I)'~h (L/B) (I+~CI+Ch-I)

1

(5.8)

L/B + ~CI+Ch-I

and 1/B f2 ~

(5.9)

(H/B) arctan 2rt

(H/B)~C

1-I-Ch- 1

where C 1 = 1 d-(L/B)

2

(5.1o)

Analysis of subsoil deformations

196 and Ch =

1 + (H/B) 2

(5.11)

where L = length of the loaded area, B = width of the loaded area, H = thickness of compressible layer. The coefficients fl and fz can also be determined from the diagram presented in Fig. 5.16. For the special case ofv - 0.5 equation 5.7 is reduced to Si~ = (q" B . 0.75 9fl) / E,. To use the equation (5.7) for a point within a loaded area, the area should be divided into four rectangles with comers at the considered point. To obtain the settlement at the point, the comer settlements of the four rectangles should be calculated and added.

5.3.3

Empirical prediction of "final" settlement

The final settlement of the subsoil can be estimated with empirical formulae, based on results of compressibility tests or collected field observations. Such empirical methods are presented by Ostromecki (1956), Drozd-Zajac (1968), Flaate (1968), Niesche (1977) and Carlsten (1988). The Ostromecki formula gives settlements of peat with thickness less than 4.5 m and loaded to 10-50 kPa. The equation is as follows: S = 1.08. C t o

(5.12)

where S = settlement (m) C = coefficient (a function of H/t 0 and ~'d) t o = q/7.55 normalized depth of dewatering (m) H = peat thickness (m) q = stress applied to the subsoil (kPa) 7a = dry unit weight of peat A nomogram for evaluation of the coefficient C is presented in Fig. 5.17.

Analysis of "Final" Deformation

197

0.9

]

r

08 '

,

. . ~ ~i. ~ , ~ > ' [ |

/,"

'

j

0.7

....

o.6

~

,.,,=

~9 o.~

~

f

0.3

.

/

/--/"

-/

i1/"

..,,I

/,//,

I-" ~

.i

~ ,

~

~

l

.

,

~_~

,

'

t

!

~I i . , /

~

~~''"

o.1 oo

~

0

~

0.2 0./, 0.6 0.8

_.-~ ~ ~

,

~

u ~ ~ ~ .

~

~;;t'~

---1.2; ,,_. _ . - ~ r - t ~ 1.66

_~~.----.-.---1

"

0.9~

~ ~ ~ ~ 0.2

I

~ -

i

-'--2

.,,,,,,~ ~ r " '.~

~

"

,

,

_~"",

--~.~

, ,

,

, " ------.-----------" -

. . . . . . . ~ .

,

3

/*

Coefficient C for the Ostromecki (1956) formula.

Settlement observations of peats of different degrees of decomposition enabled Niesche (1977) to elaborate an empirical approach to evaluate the final deformation of peat at loading. On the basis of field observations, Niesche gives a nomogram for the evaluation of deformation when the load q and degree of decomposition are known. This nomogram is presented in Fig. 5.18. An estimation of final settlement in organic clay and gyttja can be made with the Drozd-Zajac method (1968). The authors give the following empirical equation: (e 0 -0.36) 1"1

S = 3.6H

lg(3.12q ~e0) l+e 0

where S = settlement (ram) H = thickness of compressible layer (m) e 0 = initial void ratio q = stress increase (bar)

,

,---r-

H/t,

Fig. 5.17.

,

(5.13)

Analysis of subsoil deformations

198 70

60

--

-

50

,....

E "6 E

z.O

30

3

0

10

20

30

40

50

60

70

80

90

100

Degree of decomposition, (%)

Fig. 5.18. Niesche's (1977) diagram for settlement calculation.

The settlement investigations in peats with different water contents performed by Flaate gave an empirical nomogram for evaluating the coefficient of volume change m v (Fig. 5.5). In this approach, the calculation of the total settlement is made with: Sf = m v q H

(5.14)

The final settlement in peat when the water content is in the range 700 - 1500 % can also be estimated on the basis of the empirical diagram (Fig. 5.19) presented by Carlsten (1988) to estimate the relative vertical deformation ~. In this case the final settlement is calculated as" Sf = e H

(5.15)

The diagram anticipates that the peat is normally consolidated. In the case of slightly overconsolidated peat, the following correction to evaluate the calculation load q" for estimation of settlement was suggested: q'=q-(g'p -or'0) where q = applied load ~ p= preconsolidation pressure ~'0 = in situ vertical stress before load application

(5.16)

Analysis of "Final" Deformation

199

100 Apptied l

toad (kPa)

~9. ~ - - . ~ - , , - ~

~- 80 60

~-- ~ ' - ~

W

E

_~

O .m

"6 5O

Applied

E L.

~. ~

toad (kPa)

100 /

O

I

~ 30 ~

.-~ ~.-

I,

20

10

123

0

9 560

.

.

.

.

1000

Water content Fig. 5 . 1 9 .

.

.

.

.

15100 .

.

.

.

(%)

C a r l s t e n ' s (1988) d i a g r a m f o r c a l c u l a t i o n t i o n ~ in p e a t .

2000

of vertical deforma-

Empirical methods are based on field settlement measurements of particular constructions. They can be used to predict settlements only in the specific kind of subsoil for which they were elaborated. The use of empirical methods for calculation of settlements in the soft soil therefore often gives low credibility for the results. The settlements for the final design should always be estimated from parameters obtained in laboratory or field tests and one- or two-dimensional calculations.

5.3.4

Prediction of settlement in one-dimensional consolidation

The estimation of settlement in one-dimensional consolidation requires knowledge of the increase in vertical stress developed in the subsoil as a result of the load from the embankment. The stress increase is calculated by the theory of elasticity. The state of stress in the subsoil under an embankment can be estimated by Gray's approach (1936) presented in Fig. 5.20. in which the stress components ( ~ - vertical, 6h - horizontal, I: - shear stress) are calculated as follows: ~v = q (13 + x c~/a-z (x-b)/R22)/~

(5.17)

cyh = q ([3 + x c~/a + z(x-b)/R22 +2zln(R1/R0)/a)ht

(5.18)

I: = -q (z cz/a-z2/R22 )/n

(5.19)

200

Analysis of subsoil deformations

~"

I

..J

~-!-~1

I

q/unit

J

QreQ

\

\\RI R~ ~ //

X\

/

/R2

Fig. 5.20. Vertical "embankment" loading (Gray, 1936).

On the basis of this solution, Osterberg (1957) published a diagram to estimate the influence factor I for evaluation of the vertical stress increase beneath an embankment with unlimited length (Fig.5.21). In the case of an embankment with limited length 1, the influence factor can be determined by using Fadum's (1948) approach and the chart for vertical stresses beneath the comer of a uniformly loaded rectangle (Fig 5.22). If a point within a loaded area is considered, the area should be divided into four rectangles with comers at the considered point. To obtain the vertical stress at the point, the stresses beneath the comer of the four rectangles should be calculated and added. The settlement calculations for one-dimensional consolidation can be made for one or more subsoil strata. In the case of many strata, the total settlement is defined as the sum of all settlements. It is assumed that the whole subsoil stratum is taken into account in the calculations (see Chapter 9). The final time-dependent deformation Sf is calculated as the sum of the initial settlement Si, consolidation settlement S and settlements due to secondary compression S. The simplest way to determine the consolidation settlement is by means of: So = A~' v H / M

(5.20)

or

So=a~H

(5.21)

201

Analysis of "Final" Deformation b/z =oo 0.5 c~:::==:==~:: 3

0./-,

O.B

u.0

u 0.3 ,2 o u C 0 .._.,

bl b 2 0 2

0.2

C

0.1

'

Gv=I-q 1 -

I=II+I2 6v 0 i

0.01

,--7

I i iIIli

0.1

a/z

10

1

Fig. 5.21. Influence factor I for vertical stress due to embankment loading (Osterberg, 1957).

0.261

I

0.22

m=t/z n=b/z

0.18

~v =ql

I

I I IIII"

rn= o o ~

2.5 1.8

0.6 I

0.14 0.10

0.5

n are/i///f"--~

Note" m and interchangeable

0.~ 0.3 0.2

0.06

Fig. 5.22. Vertical stress beneath the corner of a uniformly loaded rectangle (Fadum 1948).

0.1

0.02 ~.~

~

0.1

1

10

Analysis of subsoil deformations

202 where M = S~ = H = ~ = Act"V =

modulus of compressibility consolidation settlement thickness of soft subsoil consolidation strain effective vertical stress increment

The above formula can be used for homogeneous subsoils with small moduli variation within the actual stress interval, e.g. heavily overconsolidated soils, and with relatively small thickness H compared to the loaded area, (i. e. relatively small change in A~' v with depth). In other cases taking moduli variation into account, the consolidation strain e~ is determined according to the formula: Ae eo =

Cr

1+e 0

where e0 = Ae = c" 0 = cr p = 6",r = Cr = C~ =

(Y'

~

log

1 +e 0

Co

P +~ cr 0

1 +e 0

O'vf

log ~

(5.22) cr p

initial void ratio decrease in void ratio initial vertical effective stress preconsolidation pressure final effective vertical stress recompression index compression index

The final relative compression at varying moduli can also be determined by means of Larsson's (1981) proposition with the estimation of three different moduli. (~ p -(~'0

Eo = ~ + M0

0" L -(3" p

~

+ My

1 M'(~',r ~'L ) +1) ~ ln( M' ME

(5.23)

The necessary explanations are presented in Fig. 3.20. The method can be used for single layer or multi layer subsoils. Based on extensive investigations, performed by the Swedish Geotechnical Institute on organic soils, an empirical nomogram was established for estimation of the

Consolidation Analysis

203

moduli values for some types of peat depending on the degree of decomposition and water content (see Chapter 5.2) A simple calculation of secondary settlement S~ describing the time-dependent movements due to secondary compression after pore pressure equalization can be based on a coefficient of secondary consolidation C~ or Co~, which is relevant for the actual void ratio or water content after primary consolidation. The settlement S~ is then expressed as: S~ = Ca log(tf/tp ) H/(1 +e0) or

(5.24)

S~ - C~e log(tf/tp ) H where tf = time of the end of the period covered by the prognosis tp = time for the end of primary consolidation The coefficient of secondary consolidation is determined in incremental oedometer tests or it may be estimated from empirical relations such as those presented in Chapter 5.2.

5.4

CONSOLIDATION ANALYSIS

5.4.1

Type and scope of analysis

When constructing embankments on organic soils, it is often necessary to utilize the shear strength increase in the subsoil during construction by stages. Such cases involve an accurate estimation of the subsoil deformation rate and the excess pore pressure dissipation in the consecutive stages. The position and thickness of each subsoil layer and the prediction of the effective stresses with time makes it possible to estimate the shear strength increase and to select a preliminary construction time schedule with consideration to stability. The same requirements for an accurate prediction of the consolidation process apply when preloading and surcharging is used in order to stop subsequent settlements. The consolidation analysis of soft soils can be made with methods based on onedimensional theory, in which the settlement St at time t is calculated as: St = U Sc

(5.25)

Analysis of subsoil deformations

204 where U = degree of consolidation So = total consolidation settlement

Due to a significant variability of the soil parameters during the deformation process, the consolidation prediction should be based on a method which accounts for the variation of the parameters with time and preferably also for creep effects. In the case of construction of embankments on deep soft subsoil, especially when the subsoil consists of several layers, the consolidation prediction should be made with methods which take into consideration the interaction of several consolidating layers with different characteristics, as well as the changing geometry and stresses due to large strains in the soil. In construction of embankments on soft soils, the very slow consolidation process causes great difficulties in designing a practical construction schedule. In order to accelerate the consolidation process, vertical drains of different types are often used. The prediction of the deformation process can then be carried out, assuming one-dimensional state of strain and vertical and axi-symmetrical pore water flow.

5.4.2

Empirical prediction of the consolidation course

In practice, the estimation of the settlements is often sufficient to design the embankment when the organic soil is a surface layer with limited thickness. In such a case, the preliminary calculation of the degree of consolidation U can be made with empirical formulae for the specific kind of subsoil for which they were elaborated. The preliminary estimation of consolidation course in peaty subsoil can be made by using Carlsten's diagrams prepared on the basis of experience from Swedish peats (Carlsten, 1988). To obtain simple diagrams for estimation ofthe rate of settlements in peat, the consolidation equation was simplified to the following form: 0.52 (wN)~ -( U = 1 - 0.6 e

qO.5 -)t (5.26)

The equation (5.26) is valid for the case where there is free drainage at both the top and the bottom of the peat layer. Limits for the investigated peats in the empirical data base are: thickness of peat 2 - 6m; water content 900 - 1500 %; applied load q i_

s

-- ~ $4 $3 $2

~'~'

-

Sc

$I

u~ r~ C)

0 So

$I $2S3

Si-1

Fig. 5.24. Estimation of consolidation parameters according to the Asaoka (1978) method.

Using the observed settlements at various times, the consolidation curve is determined as S i = f(t i ), where t i = t0+iAt (Fig. 5.24). The points on the consolidation curve are then used in the linear relationship Si= f(Si.1). The intersection point ofthe line thus obtained and the bisector of the coordinates is the total settlement S Gand the curve inclination gives the consolidation parameter ~1" Settlement within the time t can be calculated from the formula: S t = So (1- exp((lnl~ 1/At)O )

(5.28)

Asaoka's method is based on Terzaghi's consolidation theory, where all parameters are constant and independent of time and deformation. Therefore, the parameters in Asaoka's method change during the consolidation process owing to creep effects (~1' > 131)"This is illustrated by the observation of the settlements of a stage constructed embankment at the Antoniny site in Fig. 5.25.

Consolidation Analysis 1.2 Si (m) ~. o

207

s]E o.__g_~

/

~LIst2 g~

tg p =o.9~

O.8 SI I =0. 0.6

0.2 0.0 0.0

I

I

I

I

0.2

0.4

0.6

0.8

1.0 1.2 Si-1 (m}

Fig. 5.25. Change in consolidation parameters in Asaoka's method during consolidation o. organic subsoil at the Antoniny site (Wolski et al, 1988).

5.4.3

Prediction of one-dimensional consolidation at small strains

Terzaghi's linear, one-dimensional consolidation theory (1924) may be used for relatively thin compressible layers and homogeneous soil. Due to the assumptions adopted in this theory, such as constant relationship between void ratio and effective stress, constant permeability and small strains, its applicability is limited to relatively stiff thin layers with small changes in parameter values. The usual form of Terzaghi's equation is:

~U 8t

~2U Cv

(5.29) 3x 2

where u = excess pore water pressure % = coefficient of consolidation = ( k M) / ( g Pw) t = time; x = vertical space coordinate

208

Analysis of subsoil deformations

The degree of vertical consolidation U v c a n be calculated from the relationship between U v and T v represented by curves in Fig. 5.26. For soft subsoils in which the coefficient of volume change m v and permeability k are changeable and are known functions of depth, the consolidation prediction can be made with Schiffman and Gibson's (1964 ) equation:

a~u

1 dk au

%mv(X) au =

~-

ax 2

k

(5.3o)

dx c3x

k(x)

31:

where k = permeability m v - coefficient of volume change Yw = unit weight of water. I

I

I

I

I

I

II

i

I

I

: .'.'.':.'.

";::

. : :'.;

:"

:.'.'.'.-.':

i 0 I/)

c 0.60o u

l'~

~ .' ...' : .': ..~

Curve I

Curve 2 Curve 3

: : ; ''"."i:

: : ". ".'.'.'."

::

'.

:.'..::

I

I

I 1

::

. ".'." :.'.'.'"

Half-closed tayers

i z

T M / / / ,,///z,

1.00 0.001

I

Open layers

: : : .'.'.'..."

,0.80 O~

I

Zv- ;

> 020-

Z)

I

l

l

l

i

i

i Jl

0.01

I

I

Time

J

I

t I

factor,

II

0.1

I

I

I

1

11 I II 1

Tv

Fig. 5.26. Relationship between time factor T v and degree of consolidation U.

209

Consolidation Analysis

5.4.4

Prediction of one-dimensional consolidation at large strains When the course of one-dimensional consolidation of thick and soft soil layers is to be predicted, the change in soil parameters, as well as the change in load and geometry because of the large strains and total deformations, has to be accounted for. In this case, the consolidation prediction among other things involves the application of a reduced or convective coordinate system (Fig.5.27). Reference plane (x=O .,,

a)

Reference

Reference plane (z=O)

plane (~':O) b)

l"(x t)

No

initial depth subsoil

Fig. 5.27.

B

Zo

volume of solids

~(xo t) present depth subsoil

Coordinate systems used in consolidation analysis. (a) Lagrange coordinate system at time t=0. (b) Convective coordinate system at time t. (c) Reduced coordinate system at time t.

Approaches to the consolidation prediction with large strain analysis for onedimensional state of strain and pore water flow were presented by Gibson and Schiffman (1981) and by Yong and Ludwig (1984). These methods were originally designed for the extreme cases of consolidation of slurries.Gibson and Schiffman proposed the following equation in a reduced coordinate system:

_(y~ _1) d~ Yw

k

~ + ~ b e~ f

de (1--~-e) ~z

where y~ - unit weight of solids e = void ratio z - reduced coordinate.

~z

k

de" ~et + D e

yw(l+e) de

~z

Dt

=0 x

(5.31)

210

Analysis of subsoil deformations

Yong and Ludwig (1984) proposed a consolidation equation with the application of a convective coordinate system (Fig.5.28) as"

1

de Du

=0 l+e dcr' Dt

(5.32)

X

where = convective coordinate u = excess pore water pressure Du = material derivative Dt x

Unit

area

>,,

.....

IZ

V/,,~'S 0 LIDS//,//,,~"'-.~

,.,-..

0 Z

=

Vs

4-, .,.._, + r--

N~,

ii -i-,, N ,,._.,

0

VOIDS Soil

skeleton

Fig. 5.28. Relationship between convective position ~ and reduced coordinate z.

An accurate description of the consolidation process, taking time effects on the compressibility of the soil into account, leads to the following equation in a convective coordinate system (Szymanski 1991):

Consolidation Analysis

.

~

.

%

.

.

211

1 1 0~J

l+e

deP Du dcr" Dt V

+~ x

d~1

=0

(5.33)

dt

where deP = change in void ratio due to primary consolidation de ~ = change in void ratio due to secondary consolidation deP

= C r

log(t~'p/G'vo) + C o log(t~' v / t~'p )

de ~ = eo - C~ log(t / tp)

(5.34)

(5.35)

e ~ = initial void ratio tp

= time w h e n e - eo

C~ = recompressionindex Co = compression index C~ = coefficient of secondary consolidation CY'vo= initial effective vertical stress c'p = preconsolidation pressure ~'v = effective vertical stress The application of differential equations to predict settlements and excess pore pressure dissipation requires these equations to be solved by means of numerical procedures because of the non-linear nature of their coefficients. Numerical solutions can be based on a finite difference scheme with the application of the fmitestrain consolidation analysis or a piece-wise linear approach. The finite consolidation analysis developed by Gibson et al. (1981) is based on the reduced coordinate system. The governing equation requires recalculation at each step for the current void ratio value and the boundary conditions defined in terms of void ratio. In the piece-wise linear iterative analysis developed by Yong and Ludvig (1984), the derivation for finite difference consolidation is performed with respect to a convective coordinate system. With the piece-wise linear iterative approach, non-linear soil properties and non-homogeneous material can be accommodated in the analysis. This approach is based on updating the excess pore pressure explicitly.

Analysis of subsoil deformations

212

Calculation of the consolidation process with these types of analysis requires the specific gravity of solids, the relationship between void ratio and effective stress and the coefficient of permeability estimated from oedometer tests. The application of the large-strain consolidation analysis to layered subsoil requires taking into account the boundary problem which appears at the level between compressible layers, because of the differences in pore pressure and pore pressure gradient values obtained from numerical calculations for each different layer at the boundary level. One way of solving this problem is to use common imagined boundaries with imaginary mesh points (Cargill, 1982). This iterative numerical procedure is complex when the soil consists of several thin layers. A more sophisticated method without the necessity of solving boundary problems in the numerical calculation for layered subsoil can be applied by using the implicit scheme in fimte consolidation analysis. This procedure, based on equation 5.33, has been used e.g. to predict the deformation process in the subsoil under a test embankment constructed at the Antoniny site. Results obtained from numerical computations based on the original Terzaghi procedure and Yong's equations for this embankment are shown in Fig.5.29. "

I

i

!

9\ x

~ ~

....

"~-.~,. ~ ~..~.~

E

I

- - - - - CONMULT without creep

0.4

.....,

i

- - - - Terzaghi's met,

~ ~176

0.8

i

.---9 observed value

CONMULT

with creep ~\

~ .....

.

c 1.2 E l.n

1.6

2.0 - ~tja 0

:: ..-.. ~.-..: ..-.;..:..:.,~,

120

240

360

480

,

,

600 720 840 Time. (days)

Fig. 5.29. Measured and calculated settlement of organic subsoil at the Antoniny site.

Consolidation Analysis 5.4.5

Layered

213

soils

Terzaghi's equation can be used in layered soils with a modified calculation procedure. In this case, it is not possible to solve the equations analytically and numerical methods have to be used, Helenelund (1951) proposed the following graphical method: The soil is divided into layers in such way that the coefficient of consolidation c v within each layer may be assumed constant. For each layer the boundary conditions must be fulfilled. Each layer is assumed to have a thickness Ax. The pore pressure is assumed to be u at time t and u" at time (t+At). With designations according to Fig.5.30, the consolidation equation can be written:

(U'i-U i )/At = c v ( 1 / A x ) [ ( U i + l - U i ) / A x - (u i -ui. 1 ) / A x ] =

(5.36)

Cv [(Ui+l + ui.1 - 2u i )/(Ax) 2]

,=

QI t_

tn (b i_

13.

I I I

I

i-1

0 I..

, t!

L

I

i.'~l I t*txt

I I I I I

ui-1 !

0(,J x LU

i I &X I

&X

i

I

&X

Vertical coordinate, x Fig. 5.30.

i*1

I

,,.._ v

Designations for graphical solution of the pore pressure dissipation in one-dimensional consolidation, ( H e l e n e l u n d 1951)

By selecting c v At/(Ax) 2 = AT v it is thus possible to proceed from the pore pressure at time t = 0 and step-wise calculate the remaining excess pore pressure after a certain time for consolidation and thereby estimate the consolidation process for the entire soil strata. With consideration to required accuracy and simplicity, a value of AT v = 1/4 is usually selected for calculation by hand.

Analysis of subsoil deformations

214

Then u" i = 1/4 (Ui+l+Ui.1) + 1/2 u i =1/2 [(l/2)(Ui+l+Ui ) + (l/2)(Ui+Ui_ 1 )]

(5.37)

Hence u i arithmetic mean of the pore pressures at the boundaries towards the surrounding layers. Example: A homogeneous clay layer. Load increase = q. Select t = 1 year = 3.15 9107 s and c v = 10 8 mVs AT v = 1/4 gives Ax=~]10 "8 x 3.15 x 107/(1/4) = 1.12 m The graphical construction is made according to Fig. 5.31.

At :0 U=Uo

I I I o a,X--t12m~ Fig. 5.31.

a•

AX

ere

Example of a graphical solution according to Helenelund's method when c v is constant, (After Hansbo 1984).

When both the % and the k-values vary, the thickness zXxi of a layer with coefficient of consolidation %i is determined by the selected time At according to Ax i = ~/cViAt/AT v . The thickness of the layer is thus directly proportional to the square root of the coefficient of consolidation. Furthermore, at both sides and in the vicinity of the boundary between two layers, the demand for continuity of the velocity of the water flow must be met,

ki (~U/~x)i = ki+l (~U/~x)i+l

(5.38)

Consolidation Analysis

215

How this can be solved graphically is shown in Fig. 5.32, where Cvl/Cv2 = 1.8, kl/k2 = 2, Cv2/Cv3 = 2.25 and k2/k3 = 1.5 etc. The graphical construction requires extra construction lines to determine the inclination of the pore pressure isochrone between the midpoints of the layers. At =0. U=Uo I I I I

(J

I

I I I I I

.c_ 0 i..

I I

r-~

az l:2ax 9

klCvl

az2=l.5ax ,

az3=ax.,

k2cv2

, k3cv3 ,

Fig. 5.32. E x a m p l e

of a numerical

solution

when

ax

. k3cv3

,,,~r

etc

B

c v and k vary (Hansbo

1984).

From the variation of c v and k we obtain Ax2=~/2.25. Ax3= 1.5 Ax3 and mx 1~ 118"Axe= 2.0-Ax3 etc. In the construction, tan c~ denotes (3u/3x)l and tan ~=(3u/3x)2 at time At. By using the extra construction k 1 tan c~ = k 2 tan 13 is obtained and hence the boundary conditions between the layers are fulfilled. In this graphical method, it is also possible to account for the variation ofk and c v during the consolidation process by letting Ax vary so that at all times AT v = [At/(Ax) 21 (kM/g Pw)=1/4

(5.39)

After each step, the increase in effective stress and the relative compression for each layer is calculated. New values of moduli, permeabilities and coefficients of consolidation are estimated and the whole situation may be reevaluated with new geometries of the problem, taking into account changes in the loading situation and associated changes in pore pressures that have occurred during the load step. The calculations in the next step then proceed from this new situation. Performing this kind of step by step calculation by hand is very time-consuming and therefore the number of steps that can be made and the number of variations in

216

Analysis of subsoil deformations

parameters and boundary conditions that can be taken into account is limited. Therefore, the calculations should preferably be carried out in a computer, where any number of steps and variations can be included. Such calculation programmes have been developed by Magnan etal. (1979) and Mesri and Choi (1985). The CONMULT-programme developed by Magnan et al. (1979) has been revised at SGI to take new models of soil compressibility and empirical observations into account (Larsson 1986). In these types ofprogrammes, also creep effects occurring during and after the dissipation of excess pore pressure can be taken into account. In the SGI model, this is done by modifying the Terzaghi equation to M

~)u

C)

gPw ~x

~t

C)Uet

~U

( k - - ) - ~ ~ x ~t

(5.40)

where uet is the additional excess pore pressure that develops due to the creep effects. The programmes can also be used to predict swelling at unloading and secondary swelling or compression thereafter. The effect of partially unsaturated soil can also be taken into account. The CONMULT-programme and the SGI model has also been used to predict the deformation process under the test embankment constructed at the Antoniny site. Results from these calculations are shown in Fig. 5.33.

ui

3

STAGE STAGE

Q 100

2

V

2O0

~00

3oo

~

E o Ii

LI.I ~ 1,5m

f

D 3.5, there is a need for backfill material with a higher friction angle

Replacement

278

B. Inclination of slope 1:3 30 = 350 n=3.0

E

20

e.

o

~

co ~

/

/

LU

..

0

1

2

3

4 5 6 7 Height of embankment, m

8

9

10

Fig. 8.4. Excavation width vs height of embankment. (D= depth of soft soil)

of thumb,

Rule

no.

2

Blasted rock (~ = 42 ~ in refill and embankment. 20" = 42 ~

f

E x~ .c: "O . m

~ o

~-~ 9 } n = 2.0 .I

~

n-1.5

10"

UJ

0

0

1

2

3

4 5 6 7 Height of embankment, m

8

9

10

Fig. 8.5. Excavation width vs height of embankment. (D= depth of soft soil)

279

Excavation and Backfill

If the most dangerous slip surface does not pass through the crest of the embankment (Cf Fig. 8.2) and the criteria described above (Eqv. 8.1) are fulfilled there will be no problems with lateral displacement. E m p i r i c a l rules employed in various countries:

The OECD report on "Construction of roads on compressible soils" (1979) gives a scheme of empirical rules from various countries, developed for determining the excavated cross-section beneath road embankments. The rules have been developed according to the local conditions in each country and do not normally consider the properties of the backfill or subsoil. For this reason, the OECD report states that the more general application in some cases can lead to problems. A check on the stability during excavation is of course necessary. If this calculation gives unsatisfactory values of the safety factor, the excavation can be performed in terraces (Fig. 8.6).

.N_

x O ' / / / / /

\ \ \ \ \ \\

/ / / / / / / @~r

///////////

\ \,~x,\~;

/// // ~\ \

Fig. 8.6. Excavation with terraces.

An alternative is to keep the pit open to a minimum during excavation. This means that excavation and backfill are performed in parallel as shown in Fig. 8.7.

x

x

X___ __,,

x _\ __

__~_._

Fig. 8.7. Excavation and backfill. 8.2.3

Limitations

The excavation cannot normally be made deeper than 5-6 m, depending on the capacity of the excavator. The method is also costly in comparison to preloading, for instance. When the excavation is made below water, compaction of backfill may be

Replacement

280

1. ORGANIC SOILS

W,"R,~/,,7,,"~,,~,,'7",,'-Rx2~,~

"~...~...:,.:

1.5 m ~j~ Org. soil ////I//////////////////11

~:~

.

.

.

.

/ ~< 1.5m \~JJ Org. soil J._" / / / / / / / / / / / / / / / / / / / / / / . ~ I

2. COMPRESSIBLE SOILS

Shoulder

Soft soil

y///,,///////////,~;,# ~-Wb

",,-,, 9 ~

-t-7, ~ 7 ,~w,,7, e ",, ~T., - ~R,,~

.3 J -

Wo

]_//////~//////////,~/

ID13 Soft so it

Depending on the shoutder width line a or line b witt be determinant for the excavation width

3. ALL COMPRESSIBLE SOILS ,

7..~

/..~

.-..~.-~=_-" Any soft soil

Any soft soil

Minimum width 4. PEAT 7\ I D -~ 1.5rn

Fig. 8.8.

\\

~

Peat

Empirical rules employed in various countries for determining the width of excavation of compressible soil beneath road embankments.

ProgressiveDisplacement

281

difficult and consideration should be given to using material which will not require compaction. This material shall be non moisture-susceptible.

8.2.4

Construction aspects

Depending on the conditions, it is possible to choose between a normal excavator and a dragline. The dragline has the advantage that it can be placed at a distance from the pit and the vehicles removing the soil from the site need not get close to the excavation. An important factor in minimising costs and maintaining production is the availability of conveniently located spoil areas for depositing the unsuitable material. Close control of the excavation operation is necessary to ensure that pockets of soft material are not overlooked. These may lead to post-constructional problems such as differential settlement, lateral displacement or even instability.

8.3

PROGRESSIVE DISPLACEMENT

8.3.1

Description of the method

Progressive displacement is a method of constructing an embankment by placing an overload on the ground surface. A continuous shear failure is produced in the soft soil and as a consequence the soft soil is squeezed away and replaced by material from the overload. Usually, a surcharge is applied to increase the force and to compensate for loss in embankment height due to settlement and lateral displacement. This method can be applied when the subsoil has a low shear strength, which is normally the case in organic soil (with the exception of peat with a low degree of decomposition). If the subsoil is stiffer there is an altemative method, where the embankment is constructed in its full length. The displacement is then created from bursting in the subsoil. In assessing the feasibility of the displacement method for a particular situation, consideration must be given to the influence of the displaced soil on the adjacent structures. Relatively large areas are likely to be affected and structures at distances from the edge of the embankment of five times the depth of treated soil are known to have been influenced. The method has proved most successful with high embankments and when the subsoil is very soft and too deep for normal excavation. With well-planned and carefully executed schemes, the method has produced a satisfactory solution even

282

Replacement

for roads with high standard. Plan

Longitudinal section

!, !, !,

Lever of future road

:_i_..$.ur.c.h.a.r.ge" :...i...i.._...,~ :.'..'..............-.........

....

/.

/

/

/

/

/

/

/

/

Fig. 8.9. P r o g r e s s i v e displacement.

Displacement methods are currently used mainly in the United States, Canada and Scandinavia. Good results have been reported from these countries, particularly when suitable rock fill has been employed and where the construction period is sufficiently prolonged to allow the movements to be largely completed. Where a stronger layer, or a dessicated crust, overlies the very soft compressible material, it may prove necessary to remove this by excavation methods. Alternatively, a blasting technique may be employed to achieve the same effect (cf. Fig. 8.10). Cross section

... -z/

tl

2 ~1~_~ " \ ~- --- --.J_"2.__~~C

....... ~"" : ' " ":,:': "'":." "':": ":" I},'D'.".:'-':.'.':';:

9,.-.,.-7..:-/7

a (m)

Plan

t

R ~

J

I ~ J,

'l'l'Jllll I "

k_) ,,_2

"~ ~

, II

I! u ~ ~ ~

~~176

~,_) ,.-) Fig. 8.10. Blasting technique.

,, ...-._7

Progressive Displacement

8.3.2

283

Design considerations

1 Initial data

Density The weight of the embankment is calculated from an assumed value of the density of the fill. If no exact values are available, the following values can be chosen: 9 Multi-gradedfill

- 1.9t/m 3

9 Fill consisting only of boulders and coarse blasted rock

- 1.7 t/m 3

Friction angle When calculating stability and earth pressure, the friction angle is chosen with respect to the material in the embankment fill. The friction angle can be estimated as: 9 Gravel, cobbles, boulders and coarse-grained moraines

- 35 ~

9 Blasted rock with large boulders

- 42 ~

Shear strength When checking stability, earth pressures and displacement depth, the shear strength is chosen from measured values. Using bursting, the shear strength of the soil will normally be reduced. The reduction is judged for each case. As a rule, the reduction can be assumed to be about 20%.

I Dimensioning Displacement is created by causing a failure in the soil through a surcharge on the ground. The necessary height of the surcharge can be estimated from H >_0.4.~f~ where H is the embankment height in metres and q:fu is the shear strength in kPa. This corresponds to an embankment load exceeding the failure load by at least 20 %. To make the displacement easier, it is possible to use bursting or excavation in front of the fill and consequently reduce the resistance against displacement. Bursting and excavation are also used to control the direction of the displaced masses. When bursting is used, the shear strength is normally decreased. The decrease is judged for each case, but as a rule a 20 % reduction in shear strength can be assumed.

Replacement

284

H1H0 s 9. / . . . / . .

/.../.../

S- surcharge Hproj-height of future road H- total height of embankment and surcharge Fig. 8.11.

The necessary embankment height H, including surcharge, as a function of the shear strength t fu"

The displacement depth is influenced by soil stratigraphy and properties of the fill. If the subsoil consists of a homogeneous cohesive soil, especially clay, the displacement depth can be derived from a trial calculation. If the ground water table is assumed to be at the original ground surface, the depth can roughly be estimated from: Of" hi + P'f" h2 - 0.55"qTfu h

Z

~..

Ps where h z = Displacement depth, given as an assumed distance from the level of the ground surface after displacement, m ~f~ = Undrained shear strength, kPa P'~ - The density of the original subsoil beneath the ground water table, t/m 3 P'f = The density of the fill below the ground water table, t/m 3 Pf = The density of the fill above the ground water table, t/m 3 If there are layers of non-cohesive material in the cohesive soil, the fill may stop at these layers. The displacement depth may also be influenced by measures taken during construction, such as bursting and excavation. The lateral displacement is normally limited to the vertical plane through the foot of the embankment. If the top layer is very soft, this vertical plane passes through the intersection between the slope of the embankment and the lower level of the soR material.

Progressive Displacement

285

Original ground surface

: PiPi :

~j'

7 . . .

/...

i ~.

. . "

jo! h 2 . ~ displacement !1 ~ t h

/.

..

/.

..

/.

../.

.

.

/.

9 . / .

.

./..

./..

Fig.8.12.Calculationof displacementdepth.

~. I

.../.../.../.../.../.../.../.../.../.../.../.../...

Fig.8.13.Lateraldisplacement.

9 /.

.

-/

Replacement

286

8.3.3

Limitations

As mentioned before, it is necessary to consider the effect on adjacent structures from the progressive displacement. Other factors that affect the design considerations are the availability of a suitable fill (such as crushed rock), the construction time required, and the maintenance costs resulting from long-term deformations. The shear strength in the subsoil cannot be too high since this creates a large resistance in the soil against failure. For this reason, the method is very difficult to use in peat with a low degree of decomposition. The entanglement between the fibres in the peat results in a high shear strength and the shear strength also increases with effective vertical pressure. The method is preferably used for high embankments. For low embankments, other methods may be better from both economic and functional aspects. Normally, the fill does not reach firm bottom layers, but leaves a layer of compressible soil beneath the fill. Though this layer may be very thin, it may take a long time to reach full consolidation in this layer. The progressive displacement shall be combined with preloading, i.e. the embankment with surcharge shall be in place for at least 4-6 months. During this preloading, the settlements should be measured continuously.

8.3.4

Construction aspects

The following factors affect the result of progressive displacement: 9 The higher the embankment, the higher will be the probability of failure in the subsoil. 9 A narrow embankment is more likely to reach firm bottom layers, due to smaller movements in the soil. 9 When the embankment is very wide, the displacement is mostly longitudinal. Sometimes the volume of the displaced masses may be so large that there is a need for bursting or excavation in front of the fill. 9 Ensure sufficient preloading time. 9 The settlements shall be measured during the preloading period. Sometimes extra time for preloading will be needed due to adjustments of the surcharge. 9 Materials forced upwards will change the level of the ground surface in the vicinity and cause changes in dewatering conditions. 9 Progressive displacement in water causes the bottom of the lake or river to heave so that the water becomes muddy. This will have an effect on currents, river channels, bird life and fish life.

Progressive Displacement

Damaged tel ephone cables

287

/

Accumua l ted water~

Damage totrees

Tn '

L " 5D

Fig. 8.14. Effects on the environment close to the construction area.

9 It is advantageous for the progressive displacement method from an economic viewpoint if there are excess masses in the vicinity of the treated area. It is also valuable if the volume of the masses is so large that extra surcharge masses are available at low cost. It is often difficult to calculate the exact volume of the masses required. 9 Ensure that there is a sufficient, safe distance to adjacent structures. The claims due to settlement damage may be considerable. Note that structures within a distance of five times the depth of the treated soil may be affected. 9 Culverts must be placed outside the area affected by the displacement. 9 Excavations or piling close to the displacement area may be unsuitable or even impossible. For this reason, progressive displacement will affect future contruction works in the area. 9 In materials with high sensitivity, shear failures may become uncontrolled unless care is taken. 9 Displacement to depths greater than 15 rn should be avoided. The following demands should be placed on the filling material: 9 The masses shall be coarse-grained to create good binding in the fill. Preferably the fill shall consist of coarse, blasted rock. Moraine with a large content of stones and boulders, stones and boulders from boulder ridges, or gravel containing stones and boulders can be used. 9 If the alternative solution is chosen, i.e. if the embankment is constructed to the full length and the displacement is created through bursting, the demands on the fill are not as stringent. One condition, though, is that the material shall be freedrained.

288

Replacement

Transitions: For transitions to areas with better beating capacity, progressive displacement is normally complemented with excavation. If and when there is a transition between two different embankments built by progressive displacement, it is very difficult to replace all soft soil down to firm bottom. Normally, some material will be left and there is a need for extra surcharge in this area. It is advisable to avoid this problem by making the displacement take place in only one direction and supplementing it with excavation where the soil is shallower. Transitions from progressive displacement to other construction methods, such as embankment piling, are not advisable.

Construction documents: The following aspects shall be detailed in the documents presented to the contractor. 9 Demands on the fill material. Demands on placing the fill. 9 Temporary transport roads to the site. Temporary diversions for traffic on existing roads. Restrictions with respect to existing buildings. 9 The site where the contractor is to obtain the fill material. The direction in which the displacement is to be managed. 9 Location of a place to store excavated material. 9 Preparatory works, such as bursting or excavation. 9 Performance, filling, surcharge and necessary preloading time. 9 Necessary measures for controlling and facilitating the progressive displacement. Since areas in the vicinity ofthe site for displacement will be affected, it is important to observe effects on buildings and other structures in the neighbourhood. 9 Excavation of material that has been lifted up on the side of the embankment. This is normally not necessary unless the embankment is located m an urban district. 9 Transitions 9 Time schedule 9 Inspection programme. Includes follow-up of quantity of fill material, sounding to check the effect of the displacement and levelling of the embankment with surcharge.

Progressive Displacement 8.3.5

289

Case history

The example concerns a progressive displacement performed in Sweden for road 50 between J6nk6ping and Orebro, approximately 200 km west of Stockholm. At this site, the soil stratigraphy was as follows: On top there was a 2-3 m thick layer of peat with a high degree of decomposition and 1-3 m of gyttja. Beneath this organic material was 2-6 m soft clay. The total depth to firm bottom layers was about 8-10 m.

+25

Level of future road

7

i

'

SurchQrge. ,

Exc~176l TI I x

7

/

* IO

0 I

0/100 I

0/200 I

0/300 Length

I

Fig. 8.15. Longitudinal section of the progressive displacement.

For stability and settlement reasons, it was decided that the embankment should be pressed down to firm bottom layers using the progressive displacement method in combination with excavation. The replacement was made through excavation between the sections 0/010 and 0/030, where the excavation depth was gradually increased to 4 m. In the section between 0/030 and 0/300, the excavation was to 4 m depth and the rest of the replacement performed through progressive displacement.

Replacement

290

Between sections 0/300 and 0/310, the excavation depth was decreased gradually from 4 to 1 m. For reasons of stability, masses from the excavation had to be transported away. Refilling was performed in direct connection with the excavation. The refill material consisted of blasted rock and the fill was placed with a surcharge of at least 3 m over the level of the future road. The necessary height of the embankment including overload can be calculated from the following formula: H>0.4.%

Length or

'

"

9~

i.~

9 N.

9

10 -15 m

~> H t o t a I

"

~"

9 \/

Level of f u t u r e r o a d / ; \ " i . z, a

"

~

~x \

"

"~:x

~

"

~x

X

9

1' " ,~. " I

9

Fill " ~. ~

.

o"

',,

Soft ]:

Fig. 8.16.

""

"

"

""

"

"'"

H pr~

~

. \ " . . \\ Originalground surfaceV \ I

\ "

~

soil

"Firm'friction'rnateriai

.

~

--T

I

................T,~,,,,,~ ~

.

*

-

. ... 'x-~, . I~ v-

Surcharge J

~

.

.) / . . ,,/

Htotal ' -I

i

__.. "

"

"

"'""

Safety margin for placing the fill. The fill is subsequently transported to the front by caterpillar.

The shear strength in the clay was about 9 kPa and consequently the necessary load was 3.6 m. The surcharge was placed with the same crest width as the future road (7 m) and its length was 30 m. This temporary surcharge was moved forward gradually as the progressive displacement was achieved. After removing this temporary surcharge, the terrace was covered with fine-grained rock material and a new surcharge was made from basecourse material. This surcharge was to have a crest width of 6 m and the level was to be at least 2.5 m above the level of the future road.

Progressive Displacement

291

Surcharge of blasted, rock to c r e a t e failure in the \soft soil

. Temporary surcherge

\

from base course material

-:::.. : . . :9.......... -.N.

\

II..- . . . ._. .~. . . -

9 f C o v e r of f i n e - g r a i n e d

,

-- rock material ,.,

w

3i ~ _~._.

" ,a

"

"~

,,

~"

~

~,"

3m

_" t~

" v

Blasted

:".'.':

:'.'.'.".'.-.':': 9

friction

material.':-'.':

\ -."~- - ~

rock

~ ~

i:

9

t~.

_

".'.'.'".'-.'.

9~ .

.

t - - ' - -S ' oftsoil ,6 m

.....

J

:..'..'.-.':

.'.-~'.':'.-:

Fig. 8.17. Diagram of progressive displacement. The documents also contained detailed instructions conceming the direction in which the organic material was to be squeezed out. Instructions regarding suitable machines (caterpillars) to place the surcharge were given. To make the displacement easier a dragline was used. The dragline removed loose pressed-out masses close to the front of the fill. One condition was that the excavation pit in front of the fill had to be at least 5-6 rn long and at least 4 m deep. After finishing the progressive displacement, the surcharge functioned as a preloading for 3 months. In the instructions to the contractor, it was stated that this preloading period could be prolonged by up to 6 months and that the surcharge was not to be removed before the Road Administration had given their consent. The following control measurements were made: 9 The consumption of fill material was measured and compared to the (theoretically) calculated value. If the difference was too large, construction was to be stopped and an investigation made to clarify the reason. One possible cause of differences could occur in connection with the fill stopping at a layer of non-cohesive material. In this case, measures such as bursting or extra excavation were to be taken. 9 Levelling was to be carried out in sections, every 20 m, and was to start as soon as the large movements close to the front had stopped. The results were communicated to the Road Administration in order to make a decision on the need for an extra surcharge.

292

Replacement

8.4 REFERENCES Carlsten,P (1995). Construction methods for roads in peatland areas. European conference on soil mechanics and foundation engineering, 11, Copenhagen, MayJune 1995. Proceedings, vol. 8. The interplay between geotechnical and engineering geology, pp. 8.13-8.18 Carlsten, P. (1989). V~igbyggnad ph torv. Handbok. V~igverket. Publ. 1989:53. Bod~,nge. 35 p. Hartl6n, J. (1985). Pressure berms, soil replacement and lightweight fills. Soil improvement methods. International geotechnical seminar, 3, Singapore, Nov., 1985. Proceedings, pp. 101-111. Holtz, R.D. (1989). Treatment of problem foundations for highway embankments. National Cooperative Highway Research Program. Synthesis of Highway Practice 147. Washington, DC. 72 p. Oppbygging av fyllinger (1994). Statens Vegvesen. HS,ndbok 176. Oslo. 124 p. Swedish National Road Administration (1991). U r g r / i ~ g f6r v~igbank. Allm/in teknisk beskrivning. V~igverket, V/ig-och Brokonstruktion. Geoteknik. Publikation 1991:06. Bodange. 53 p. Swedish National Road Administration (1979). Nedpressning av v/igbank. Statens V~igverk; TU 139.39 p.

293

Chapter 9

Staged Construction W. Wolski, Department of Geotechnics, Warsaw Agricultural University

9.1

GENERAL

Staged construction consists in the filling of an embankment at a controlled rate, so as not to cause failure but to permit an increase in shear strength due to consolidation. In such a way, the obtained strengthening of the foundation soil should be sufficient to support safely the required load. Thus, in staged embankments, the precompression technique is used, which according to Johnson (1970) is defined as: "compressing the soil under an applied stress prior to placing or completing the structure load". In the case of the staged embankment, the first stage of embankment (preloading with first stage) compresses the subsoil prior to the filling of the second stage. In the case of roads or dykes, two or three stages are often used. When settlements after the end of embankment construction have to be minimised, surcharging is used. This is a temporary preloading with load in excess of the permanent fill. Where there is a high ground water level the settling embankment is gradually submerged. Because of uplift the effective load is then decreased. A schematic explanation of the staged construction is shown in Fig. 9.1. The staged construction technique is frequently used in conjunction with the installation of vertical drains in order to accelerate the consolidation process and to reduce the period before the next stage of the embankment construction can be commenced. The staged embankment is considered a very useful technique of construction on most organic soils. But, in spite of an apparent simplicity, it needs: 9 a good quality soil investigation 9 a design well-adjusted to the foundation behaviour, taking into account the relationship between the stress state developed in the subsoil and stability 9 careful monitoring of the subsoil behaviour during the construction period.

294

Staged Construction

"0 0 0 _A

Possible load decrease due to uplift / ~, f

I

...... i

I

........I.......-

nd stage

I Surcharge

Permanent fill Time

r

C

E @

Fig. 9.1. Staged construction scheme.

9.2

PRECOMPRESSION TECHNIQUE

9.2.1

Introduction

Staged embankments are constructed by the use of precompression technique in which each stage is utilised as preloading before the next stage is placed. Each stage should be constructed according to earthwork standards, using layers. Unlike temporary preloading fills, used for improving soft, soils before a structure is built, each stage of the embankment should be made of an appropriate soil, with proper compaction. Depending on the site conditions, it may be necessary to construct several layers of the first stage as a working mat - without compaction, to support the hauling vehicles and the equipment for compaction. When the top layer is composed of peat it is important to keep the surface intact, which can be done by removing vegetation with very light equipment to avoid disturbance of the surface mat. In some cases, e.g. where there is a top layer of decomposed peat, it would be advisable to use a geotextile as a separating layer (see Chapter 11) to prevent local failure. The first stage of the embankment is usually wider than the next one, thus providing loading berms which improve stability condition in the subsoil (Fig. 9.2).

Precompression Technique

295

/

1II,,'~I//

j~

F

~"

/

Surchorge

2nd sta.qe Ist

stage

\

x

~'~

Lomdingberm

l 1~~_

I//,~,WlX,.~ O'vf throughout the entire soft layer. To fulfil this requirement, the degree of consolidation at the midpoint of the compressible stratum at the time of surcharge removal can be computed according to the following proposition by Johnson (1970):

Precompression Technique

297

CI

n wilt settle ,//~ after surcharge f l ' / / A movot

c-

I |

-7~176176

13.

n

\N .4--"

ffvs at

o

~

x

Sand

Fig. 9.4. Stresses at the time of surcharge removal.

CYvf log (1 + ~ ) CYv o Uff+~,:) = , (~ vf (T'vs log (1 + + ) (~' G' VO

(9.2)

VO

where CY'vs = the vertical effective stress under the permanent load with surcharge after excess pore pressure dissipation ~'vo = initial vertical effective stress Cy'vf = final vertical effective stress under the permanent load (see notations given in Figs. 9.3 and 9.4.). A solution of this formula, that is the value for U(f§ is given in Fig. 9.5. From this value the moment for surcharge removal can be indicated. The above requirement leads to the conservative design in which the actual precompression settlement will be greater than designed Sol. It is important to emphasise that the design of preload should be considered as preliminary and the decision to remove the surcharge should be based on field observations.

Staged Construction

298 ~,

100 I

.f, g o s0F

~

~.o

r~m

30

L

~176t

"5 ~

0

i

0.2

i

0.4

l

I

I

,

0.6 0.8 1.0 1.2 1.4 1.6 1.8 2.0 Surchorge to embankment Iood rcttio. ~vs/6'vf

Fig. 9.5. Diagram for estimating the surcharge to eliminate primary settlement (after Johnson, 1970).

Surcharging to compensate for secondary compression: This is particularly important for embankments on organic soils. Effects of the secondary compression may result in a significant settlement during the economic life of the embankment. This is especially evident when vertical drams are used, which drastically reduce the time required for primary consolidation. For a preliminary estimation of the reduction in secondary settlement due to surcharge, in normally consolidated soils, the diagram given in Fig. 9.6 can be utilised (Ladd, 1976). The notations used in the diagram are as follows: G'v~ = vertical effective stress after excess pore pressure dissipation under the load with surcharge ~'vf = final vertical effective stress under the permanent load only aider pore pressure dissipation C ae = modified coefficient of secondary compression tp = time at the end of primary consolidation t~r = time at which the surcharge is removed R = surcharge ratio According to Ladd (1976) for the first cycle of secondary compression the reliable surcharge ratio is obtained when the "maximum reduction" line is used. The time at which the surcharge is removed t~r should be close to the tp value, i.e. tfftp> 1. A preliminary assumption of the surcharge value can be based on the secondary behaviour of the Swedish clays described by Larsson (1981). It was found that

Precompression Technique 100

299 ~\,

,

80

\\

,

,

Minimum

.c_ .x2_ 60 c-

C 0

20 ~o

~

~

~

" -... /Maximum ~

0

0

10

20

Surcharge

Fig.9.6.

Averooe

[Reduction k, ----7='-"~, 30

/.,13

50

ratio R s = ( % s - % f ) / S v f , (%)

Reduction in rate of secondary compression due to surcharging (after Ladd, 1976).

secondary settlements will take place when the effective stress in the soil exceeds 0.8 ~'p (where C'p is preconsolidation pressure). Hence surcharging ought to result in the stress G'v~= ~' f/0.8, where C'vf = effective stress in subsoil under final load. This is valid only in limited depth, where change of additional load distribution is small by depth. For a more detailed analysis, the method given by Johnson (1970) can be used. The general layout of the surcharge design for partial compensation to secondary compression is given in Fig. 9.7. The degree of consolidation required under surcharge fill loading to produce primary consolidation plus the desired amount of seeondary compression can be expressed as follows:

U(f+sr)-Up (1 +

log

tso

)

(9.3)

where Up = degree of consolidation required under surcharge loading to produce settlement equal to primary consolidation C~ = coefficient of secondary compression = strain at the centre of the compressible stratum, caused by primary consolidation under permanent loading t~o = time determined by the useful life of the embankment or by the amount of secondary compression for which compensation is desired tp = time corresponding to primary consolidation under permanent load.

300

Staged Construction

/due touplifi Possible lo

[qf

decrease

,,,

t,tog time

=.._ v

tsc \/o.. \ v~;,

\R

~ S c f

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