Developments in the Formulation and Reinforcement of Concrete (Woodhead Publishing in Materials)

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Developments in the formulation and reinforcement of concrete

Related titles: Durability of concrete and cement composites (ISBN 978-1-85573-940-6) Concrete and other cement-based composites are by far the most widely used manmade construction materials in the world. However, major problems of infrastructure deterioration have been caused by unanticipated premature degradation of these materials. This book provides an up-to-date review of several of the main forms of degradation, examining what is known about their causes and control. Trends in modelling and prediction of service lives are also examined. Advanced civil infrastructure materials (ISBN 978-1-85573-943-7) In recent decades, developments in materials in response to the call for more durable infrastructures have led to many exciting advancements. The purpose of this book is to provide an up-to-date review of several emerging infrastructure materials that may have a significant impact on repair and/or new construction. An area that is given high priority in this book is the materials design concept that has led to the creation of advanced composites; such design methodology has been made possible by several key advancements in materials science and mechanics. Also, each chapter concludes with selective examples of real world applications using such advanced materials. This book will be an invaluable reference for undergraduate and graduate students of civil engineering, materials, architecture and building engineering, and will serve as a guide for researchers and professional engineers in the construction industry. Durability of composites for civil structural applications (ISBN 978-1-84569-035-9) This comprehensive book on durability of FRP composites will make it easier for the practising civil engineer and designer to use these materials on a routine basis. It addresses the current lack, or inaccessibility, of data related to the durability of these materials, which is proving to be one of the major challenges to the widespread acceptance and implementation of FRP composites in civil infrastructure. The book should help further the acceptance of composites for civil structural applications, by providing this source for practising engineers, decision makers, and students involved in architectural engineering, construction and materials, disaster reduction, environmental engineering, maritime structural technology, transportation engineering and urban planning. Details of these and other Woodhead Publishing materials books, as well as materials books from Maney Publishing, can be obtained by: • •

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Developments in the formulation and reinforcement of concrete Edited by Sidney Mindess

Woodhead Publishing and Maney Publishing on behalf of The Institute of Materials, Minerals & Mining CRC Press Boca Raton Boston New York Washington, DC

Woodhead publishing limited Cambridge, England

Woodhead Publishing Limited and Maney Publishing Limited on behalf of The Institute of Materials, Minerals & Mining Woodhead Publishing Limited, Abington Hall, Granta Park, Great Abington, Cambridge CB21 6AH, England www.woodheadpublishing.com Published in North America by CRC Press LLC, 6000 Broken Sound Parkway, NW, Suite 300, Boca Raton, FL 33487, USA First published 2008, Woodhead Publishing Limited and CRC Press LLC © 2008, Woodhead Publishing Limited The authors have asserted their moral rights. This book contains information obtained from authentic and highly regarded sources. Reprinted material is quoted with permission, and sources are indicated. Reasonable efforts have been made to publish reliable data and information, but the author and the publishers cannot assume responsibility for the validity of all materials. Neither the author nor the publishers, nor anyone else associated with this publication, shall be liable for any loss, damage or liability directly or indirectly caused or alleged to be caused by this book. Neither this book nor any part may be reproduced or transmitted in any form or by any means, electronic or mechanical, including photocopying, microfilming and recording, or by any information storage or retrieval system, without permission in writing from Woodhead Publishing Limited. The consent of Woodhead Publishing Limited does not extend to copying for general distribution, for promotion, for creating new works, or for resale. Specific permission must be obtained in writing from Woodhead Publishing Limited for such copying. Trademark notice: Product or corporate names may be trademarks or registered trademarks, and are used only for identification and explanation, without intent to infringe. British Library Cataloguing in Publication Data A catalogue record for this book is available from the British Library. Library of Congress Cataloging in Publication Data A catalog record for this book is available from the Library of Congress. Woodhead Publishing ISBN 978-1-84569-263-6 (book) Woodhead Publishing ISBN 978-1-84569-468-5 (e-book) CRC Press ISBN 978-1-4200-7609-7 CRC Press order number WP7609 The publishers’ policy is to use permanent paper from mills that operate a sustainable forestry policy, and which has been manufactured from pulp which is processed using acid-free and elementary chlorine-free practices. Furthermore, the publishers ensure that the text paper and cover board used have met acceptable environmental accreditation standards. Project managed by Macfarlane Production Services, Dunstable, Bedfordshire, England (e-mail: [email protected]) Typeset by SNP Best-set Typesetter Ltd., Hong Kong Printed by TJ International Limited, Padstow, Cornwall, England

Contents

Contributor contact details Introduction S Mindess, University of British Columbia, Canada

1 1.1 1.2 1.3 1.4 1.5 1.6 1.7 1.8 1.9 2 2.1 2.2 2.3 2.4 2.5 2.6

Autoclaved aerated concrete R Klingner, University of Texas at Austin, USA Introduction to autoclaved aerated concrete Applications of autoclaved aerated concrete Structural design of autoclaved aerated concrete elements Seismic design of autoclaved aerated concrete structures Design example: three-story autoclaved aerated concrete shear-wall hotel Further background on earthquake performance of autoclaved aerated concrete shear-wall structures Development of seismic design factors (R and Cd) for ductile autoclaved aerated concrete shear-wall structures Acknowledgements References High-density and radiation-shielding concrete P Lessing, Idaho National Laboratory, USA Introduction Applications/case studies The case of DUAGG® and DUCRETE® Future trends Sources of further information and advice References

ix

xiii

1 1 4 6 11 14 34 41 42 42 44 44 47 47 66 73 76 v

vi 3

3.1 3.2 3.3 3.4 3.5 4 4.1 4.2 4.3 4.4 4.5 4.6 4.7 4.8 5 5.1 5.2 5.3 5.4 5.5 6

6.1 6.2 6.3 6.4 6.5 7 7.1 7.2 7.3 7.4

Contents High-strength concrete O E Gjørv, Norwegian University of Science and Technology, Norway Introduction Applications Future trends Sources of further information and advice References

79

Sprayed concrete (shotcrete) N Banthia, University of British Columbia, Canada Introduction Mix proportioning and process implications Strength and stiffness Kinematics and rebound Toughness, impact resistance and fiber reinforcement Concluding remarks Acknowledgements References

98

79 80 90 93 95

98 99 101 102 108 111 111 111

Hot weather concreting C Ishee, Florida Department of Transportation, USA Introduction Applications/case studies Future trends Sources of further information and advice References

114

Underwater concrete A K Al-Tamimi, College of Engineering, The American University of Sharjah, United Arab Emirates Introduction Development of underwater concrete Quality control of underwater concrete Application/case study References

136

Fibrous concrete reinforcement S Mindess, University of British Columbia, Canada Introduction How do fibres work? Types of fibres Mix proportioning, fabrication and placement

154

114 114 131 133 134

136 137 138 144 151

154 155 156 157

Contents

vii

7.5 7.6 7.7 7.8 7.9 7.10

What do fibres do? High performance fibre reinforced concrete Hybrid fibre systems Applications of fibre reinforced concrete Concluding remarks References

158 161 163 164 165 166

8

Lightweight concrete T W Bremner, University of New Brunswick, Canada Introduction Applications/case studies Production of lightweight concrete Future trends Sources of further information and advice References

167

Self-compacting concrete (SCC) M Geiker, Technical University of Denmark, Denmark Significance of self-compacting concrete Selected properties of self-compacting concrete Applications/case studies Future trends Sources of further information and advice References

187

8.1 8.2 8.3 8.4 8.5 8.6 9 9.1 9.2 9.3 9.4 9.5 9.6 10

167 176 178 181 183 184

187 188 195 200 202 204

Recycled materials in concrete C Meyer, Columbia University, USA 10.1 Introduction 10.2 Fly ash 10.3 Ground granulated blast furnace slag (GGBFS) 10.4 Recycled concrete 10.5 Recycled waste glass 10.6 Recycled tires 10.7 Recycled plastics 10.8 Other recycled materials 10.9 Future trends 10.10 References

208

11

231

Foamed concrete V Bindiganavile and M Hoseini, University of Alberta, Canada 11.1 Introduction 11.2 Definitions and classifications

208 210 211 213 215 220 222 223 225 227

231 232

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Contents

11.3 11.4 11.5 11.6 11.7 11.8 11.9 11.10 11.11

Materials Mix design Production of foamed concrete Properties of foamed concrete Fiber reinforced foamed concrete Applications Research needs Acknowledgements References

232 235 236 237 248 250 251 252 252

12

Polymer concrete Y Ohama, Nihon University, Japan Introduction Production techniques for polymer concrete Practical applications, recycling and quality standards Future trends Sources of further information and advice References

256

Future developments in concrete L Czarnecki, Warsaw University of Technology, Poland, W Kurdowski, Institute of Mineral Building Materials, Poland and S Mindess, University of British Columbia, Canada Introduction Does concrete have a future? General factors influencing the development of concrete Functional concrete Nanocement and nanoconcrete Concluding remarks References

270

12.1 12.2 12.3 12.4 12.5 12.6 13

13.1 13.2 13.3 13.4 13.5 13.6 13.7

Index

256 257 259 263 268 268

270 271 272 276 279 282 282

285

Contributor contact details

(* = main contact)

Chapter 3

Introduction

Professor Odd E. Gjørv Norwegian University of Science and Technology, NTNU Department of Structural Engineering Rich. Birkelandsv. 1a NO-7491 Trondheim Norway E-mail: [email protected]

Dr Sidney Mindess Department of Civil Engineering University of British Columbia Vancouver BC Canada E-mail: [email protected]

Chapter 1 Professor Richard E. Klingner The University of Texas at Austin Engineering 1 University Station C2100 Austin, TX 78712-0284 USA E-mail: [email protected]

Chapter 2 Dr Paul A. Lessing Idaho National Laboratory Box 1625 Idaho Falls, ID 83415-2218 USA E-mail: [email protected]

Chapter 4 Dr Nemy Banthia Department of Civil Engineering University of British Columbia Vancouver BC Canada E-mail: [email protected]

Chapter 5 Dr Charles Ishee Florida Department of Transportation 5007 Northeast 39th Avenue Gainesville, FL 32609 USA E-mail: [email protected]

ix

x


Chapter 6

Chapter 9

Dr Adil K. Al-Tamimi Department of Civil Engineering College of Engineering The American University of Sharjah P.O. Box 26666 Sharjah United Arab Emirates E-mail: [email protected]

Associate Professor Mette R. Geiker Department of Civil Engineering Technical University of Denmark Brovej, Building 118 DK-2800 Kgs. Lyngby Denmark E-mail: [email protected]

Chapter 10 Chapter 7 Dr Sidney Mindess Department of Civil Engineering University of British Columbia Vancouver BC Canada E-mail: [email protected]

Dr Christian Meyer 622 SW MUDD Mail Code 4709 500 West 120th Street New York, NY 10027-6699 USA E-mail: [email protected]

Chapter 8

Chapter 11

Professor Emeritus Theodore W. Bremner Honorary Research Professor Department of Civil Engineering 17 Dineen Drive University of New Brunswick Fredericton NB Canada E3B 5A3 E-mail: [email protected]

Dr Vivek Bindiganavile* and M. Hoseini Structural Engineering University of Alberta 3-020 Markin/CNRL Natural Resources Engineering Facility Edmonton Alberta Canada T6G 2W2 E-mail: [email protected]


xi

Chapter 12

Chapter 13

Professor Yoshihiko Ohama 14-10-402 Hiyoshi 2-chome Kohoku-ku Yokohama 223-0061 Japan E-mail: [email protected]

Dr Lech Czarnecki* Warsaw University of Technology Warsaw Poland Email: [email protected] Dr Wieslaw Kurdowski Institute of Mineral Building Materials Cracow Poland Dr Sidney Mindess University of British Columbia Vancouver BC Canada E-mail: [email protected]

Introduction

S MINDESS, University of British Columbia, Canada

Concrete is the most widely used man-made material in the world; indeed, of all materials, only water is used in greater quantities (Table I.1). In many ways, concrete literally forms the basis of our modern society. Almost every aspect of our daily lives depends, directly or indirectly, on concrete. We need only consider the obvious examples: roads, bridges, runways, dams, water conduits, buildings of all types, and so on, to realize its importance. However, because it is so ubiquitous, we tend to take it for granted. The general view of concrete can be expressed by the often-quoted remark, “You mix together cement, gravel and water, and it gets hard. What else is there to know about concrete?” Concrete production is now, of course, a mature technology. “Modern” concretes have been in use since the middle of the 19th century, and the ancient Greeks and Romans both produced mortars which may be considered to be forerunners of today’s concretes. It is thus not surprising that concrete is often considered to be a rather “boring” topic, compared to the more dramatic modern discoveries in such areas as information technology, communications, or medical technology. However, in this case, appearances are deceiving; in fact, modern concretes constitute a sophisticated family of materials. Portland cement itself is a complex material, manufactured by first burning an intimate mixture of limestone and clay or shale in a kiln at temperatures in the range of 1400°C to 1500°C, and then intergrinding the resulting clinker with gypsum. In producing concrete, this cement is now commonly combined with one or more supplementary cementing materials, such as fly ash, silica fume, blast furnace slag, or limestone powder. In addition, modern concretes usually contain one or more chemical admixtures to modify the properties of either the fresh or the hardened concrete. Concrete is thus a very complex, though not completely understood, system. Today, exciting developments in cement and concrete are underway. There are several forces driving this resurgence in cement and concrete research. Perhaps the most significant is the drive to make the concrete industry much more sustainable. The production of Portland cement is highly energy intensive. As well, the cement industry is a significant conxiii

xiv

Introduction Table I.1 Annual production of selected materials Concrete Steel Salt Sugar Oil

∼3.8 billion m3 ∼8.7 billion tonnes ∼1 billion tonnes ∼200 million tonnes ∼135 million tonnes ∼5.2 billion tonnes

tributor of greenhouse gases: the production of one tonne of cement leads to almost one tonne of CO2 being released into the atmosphere. This is estimated to account for about 6% of the worldwide production of greenhouse gases. Thus, driven by environmental (and hence governmental) pressures, there is a real incentive for the concrete industry to become much “smarter” about concrete production. This involves, amongst other things, an increasing use of supplementary cementing materials to replace Portland cement, and an increased emphasis on durability rather than strength. There is a movement in the industry to performance specifications rather than the current prescriptive specifications for concrete; this should permit modern concrete producers to make much more efficient use of their materials. Related to this is the development of a large number of special concretes that are optimized for particular applications. These concretes are now available only because of the advances in our understanding of the chemistry and physics of concrete, and the development of the admixtures mentioned above. Indeed, most of the chapters in the present volume deal with these new types of concretes. The authors have described both the underlying science and the production applications of these new materials. A careful reading of these chapters should lead to a better understanding of the possibilities inherent in concrete technology – we are limited only by our imaginations in the types of concretes that will be available in the future.

1 Autoclaved aerated concrete R KLINGNER, University of Texas at Austin, USA

1.1

Introduction to autoclaved aerated concrete

Autoclaved aerated concrete (AAC) is a concrete-like material with very light weight, obtained by uniformly distributed, closed air bubbles. Material specifications for this product are prescribed in ASTM C1386. Because AAC typically has one-sixth to one-third the density of conventional concrete, and about the same ratio of compressive strength, it is useful for cladding and infills, and for bearing-wall components of lowto medium-rise structures. Because its thermal conductivity is one-sixth or less that of concrete, it is energy-efficient. Because its fire rating is slightly longer than that of conventional concrete of the same thickness, it is very fire-resistant. It is not susceptible to mold. Because of its internal porosity, it has very low sound transmission, and is acoustically very effective.

1.1.1 Historical background of AAC AAC was first produced commercially in Sweden, in 1923. Since that time, its production and use have spread to more than 40 countries on all continents, including North America, Central and South America, Europe, the Middle East, the Far East, and Australia. This wide experience has produced many case studies of use in different climates, and under different building codes. Background material on experience with AAC in Europe is given in RILEM (1993). In the US, modern uses of AAC began in 1990, for residential and commercial projects in the southeastern states. US production of plain and reinforced AAC started in 1995 in the southeast, and has since spread to other parts of the country. A nationwide group of AAC manufacturers was formed in 1998 as the Autoclaved Aerated Concrete Products Association (http://www.aacpa.org/). Design provisions for AAC are provided in the 1

2


Code and Specification of the Masonry Standards Joint Committee (MSJC), and in the technical manuals available on the web site of the AACPA. The AACPA includes one manufacturer in Monterrey, Mexico, and many technical materials are available in Spanish as well as English.

1.1.2 AAC elements AAC can be used to make unreinforced, masonry-type units, and also factory-reinforced floor panels, roof panels, wall panels, lintels, beams, and other special shapes. These elements can be used in a variety of applications including residential, commercial, and industrial construction. Reinforced wall panels can be used as cladding systems as well as load-bearing and non load-bearing exterior and interior wall systems. Reinforced floor and roof panels can be efficiently used to provide the horizontal diaphragm system while supporting the necessary gravity loads.

1.1.3 Materials used in AAC Materials for AAC vary with manufacture and location, and are specified in ASTM C1386. They include some or all of the following: fine silica sand; Class F fly ash; hydraulic cements; calcined lime; gypsum; expansive agents such as finely ground aluminum powder or paste; and mixing water. Details of the mixture designs used by each producer depend on the available materials and the precise manufacturing process, and are not publicly available. The finely ground aluminum powder or paste produces expansion by combining with the alkaline slurry to produce hydrogen gas. AAC can be reinforced internally in the manufacturing process with welded wire cages, and also at the job site with conventional reinforcement.

1.1.4 How AAC is made Overall steps in the manufacture of AAC are shown in Fig. 1.1, and described below. Sand is ground to the required fineness in a ball mill, if necessary, and is stored along with other raw materials. The raw materials are then batched by weight and delivered to the mixer. Measured amounts of water and expansive agent are added to the mixer, and the cementitious slurry is mixed. Steel molds are prepared to receive the fresh AAC. If reinforced AAC panels are to be produced, steel reinforcing cages are secured within the molds. After mixing, the slurry is poured into the molds. The expansive agent creates small, finely dispersed voids in the fresh mixture, which increases the volume by approximately fifty percent in the molds

Autoclaved aerated concrete

3

Cement Lime Water Ball mill

Sand/fly ash

Expanding agent

Mix

Reinforcement wires Cast

Longitudinal cut Expansion Autoclave

Anticorrosive treatment

Assembling

To job site

AAC products

1.1 Overall steps in manufacture of AAC.

within three hours. Within a few hours after casting, the initial hydration of cementitious compounds in the AAC gives it sufficient strength to hold its shape and support its own weight. After cutting, the aerated concrete product is transported to a large autoclave, where the curing process is completed. Autoclaving is required to achieve the desired structural properties and dimensional stability. The process takes about 8–12 hours under a pressure of about 174 psi (12 bars) and a temperature of about 360°F (180°C) depending on the grade of material produced. During autoclaving, the wire-cut units remain in their original positions in the AAC block. After autoclaving, they are separated for packaging. AAC units are normally placed on pallets for shipping. Unreinforced units are typically shrink-wrapped, while reinforced elements are banded only, using corner guards to minimize potential localized damage that might be caused by the banding.

1.1.5 AAC strength classes AAC is produced in different densities and corresponding compressive strengths, in accordance with ASTM C1386 (Precast Autoclaved Aerated Concrete Wall Construction Units). Densities and corresponding strengths are described in terms of “strength classes” (Table 1.1).

4


Table 1.1 Typical material characteristics of AAC in different strength classes Strength class

Specified compressive strength lb/in2 (MPa)

Nominal dry bulk density lb/ft3 (kg/m3)

Density limits lb/ft3 (kg/m3)

AAC 2.0

290 (2.0)

AAC 4.0

580 (4.0)

AAC 6.0

870 (6.0)

25 31 31 37 44 50 44 50

22 28 28 34 41 47 41 47

(400) (500) (500) (600) (700) (800) (700) (800)

(350)–28 (450)–34 (450)–34 (550)–41 (650)–47 (750)–53 (650)–47 (750)–53

(450) (550) (550) (650) (750) (850) (750) (850)

Other strength classes within these ranges and densities may be produced depending on specific design requirements.

Table 1.2 Dimensions of plan AAC wall units AAC unit type

Width, in. (mm)

Height, in. (mm)

Length, in. (mm)

Standard block Jumbo block

2–15 (50–375) 4–15 (100–375)

8 (200) 16–24 (400–610)

24 (610) 24–40 (610–1050)

Table 1.3 Dimensions of reinforced AAC wall units Product type

Thickness, in. (mm)

Height or width, in. (mm)

Typical length, ft (mm)

Wall panel Floor panel Lintel/beam

2–15 (50–375) 4–15 (100–375) 4–15 (100–375)

24 (610) 24 (610) 8–24 (200–610)

20 (6090) 20 (6090) 20 (6090)

1.1.6 Typical dimensions of AAC units Typical dimensions for plain AAC wall units (masonry-type units) are shown in Table 1.2. Typical dimensions for reinforced AAC wall units (panels) are shown in Table 1.3.

1.2

Applications of autoclaved aerated concrete

Autoclaved aerated concrete (AAC) can be used in a wide variety of structural and non-structural applications (Barnett et al. 2005), examples of


5

which are shown in the following figures. Figure 1.2 shows an AAC residence in Monterrey, Mexico, in which the AAC is used as structure and envelope. Figure 1.3 shows an AAC hotel in Tampico, Mexico, in which the AAC is again used as structure and envelope. Figure 1.4 shows an AAC cladding application on a high-rise building in Monterrey, Mexico. In each of the above applications, the thermal and acoustical efficiency of the AAC makes it an attractive choice for building envelope.

1.2 AAC residence in Monterrey, Mexico (courtesy Xella Mexicana).

1.3 AAC hotel in Tampico, Mexico (courtesy Xella Mexicana).

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1.4 AAC cladding in Monterrey, Mexico (courtesy Xella Mexicana).

1.3

Structural design of autoclaved aerated concrete elements

1.3.1 Integrated US design context for AAC elements and structures Prior to October 2003, proposed AAC masonry buildings in the US had to be approved on a case-by-case basis. Since that date, project approvals can be obtained under the general evaluation-service reports ICC ES 215 (2003) and ICC ESR-1371 (2004). Since early 2005, project approvals for AAC masonry structures can be obtained through the inclusion of design provisions for AAC masonry in the mandatory-language Appendix A of the 2005 MSJC Code and Specification. It is also expected that reinforced AAC panels will be analogously addressed through ACI 318. This design context is shown schematically in Fig. 1.5 and is discussed in more detail in the rest of this chapter. Loads for structural design of AAC should be taken from appropriate load codes, such as ASCE 7. AAC masonry elements are designed using the provisions of Appendix A of the 2005 MSJC Code and Specification. Reinforced AAC panels are designed using manufacturers’ recommendations.

1.3.2 US design and construction provisions for elements and structures of AAC masonry In the US, development of masonry design provisions by an ANSI consensus process is the responsibility of the Masonry Standards Joint Committee (MSJC), sponsored by the American Concrete Institute (ACI), the American


7

Model codes

AAC masonry design appendix in MSJC Code and Specification

ASTM specifications unique to AAC masonry

R, Cd

ASTM specifications for AAC material

Reinforced AAC panel appendix in ACI 318

ASTM specifications unique to reinforced AAC panels

1.5 Integrated US design background for AAC elements and structures.

Society of Civil Engineers (ASCE), and The Masonry Society (TMS). The MSJC Code and Specification is essentially referenced directly by US model codes (International Building Code and NFPA Code). The MSJC design provisions cover a wide variety of design approaches (strength, allowable-stress, empirical) and materials (clay, concrete, glass block). Based on the combination of test results from the University of Texas at Austin, the University of Alabama at Birmingham, and elsewhere, a strength design approach was developed for AAC masonry, with provisions that are generally similar to current strength-design provisions for other types of masonry, and for reinforced concrete. The proposed design provisions, commentary, and “super-commentary” were introduced, refined by, and approved by MSJC in 2004, in the form of a mandatory-language Appendix to the 2005 MSJC Code and Specification. They produce final designs similar to those produced by the proposed ACI provisions for reinforced AAC elements, described below. Flexural resistance of AAC masonry elements is computed assuming yielded flexural reinforcement and an appropriate equivalent rectangular stress block. Maximum reinforcement is limited to ensure tensioncontrolled behavior. Deformed reinforcement must be used, and must be surrounded by grout. Development and splice requirements are the same as for conventional masonry; only the grout is considered, and bond failure and splitting are addressed. In-plane shear resistance of AAC masonry elements is computed as the sum of resistance from masonry plus deformed reinforcement in intermediate bond beams only. In-plane shear resistance from AAC masonry is checked with respect to web shear, crushing of the diagonal strut, and

8


sliding shear. Out-of-plane resistance of AAC masonry elements is computed using beam shear equations similar to those used for conventional masonry. Capacity design for shear is required. These design requirements are accompanied by corresponding construction requirements in the MSJC Specification, which is mandated by the MSJC Code. Construction requirements address quality assurance, materials and execution.

1.3.3 US design provisions for reinforced AAC panels In the US, development of design provisions for reinforced concrete under the ANSI consensus process is the responsibility of ACI Committee 318. The latest version of that committee’s document, ACI 318-05, is essentially referenced directly by US model codes. The design provisions of ACI 318-05 address the strength design of a wide variety of conventional reinforced concrete elements similar to AAC applications, including prefabricated wall panels. Based on the combination of test results from the University of Texas at Austin, the University of Alabama at Birmingham, and elsewhere, a strength design approach has been developed for reinforced AAC elements, that is consistent with ACI 318-05, whose provisions are generally similar to ACI 318-05 strengthdesign provisions for reinforced concrete elements, and that produce final designs similar to those produced by the proposed MSJC provisions for AAC masonry. The first set of proposed design provisions, commentary, and “supercommentary” was introduced to ACI Subcommittee 523A (Autoclaved Aerated Cellular Concrete) in the Fall of 2002. Because ACI 523A is a relatively new subcommittee, the design provisions, commentary, and “supercommentary” were introduced as appendices to a non-mandatory design guide on AAC. After the guide has been approved by Committee 523, it will be offered to ACI 318 as a basis for a mandatory-language appendix to that document. Because these provisions must be discussed and refined within ACI Committee 318 as well as ACI 523, their timetable for approval will probably be extended longer than for their counterpart provisions for AAC masonry. It is hoped that this will be approved for the 2014 edition of the ACI 318 document.

1.3.4 Handling, erection and construction with AAC elements AAC masonry units are laid with a polymer-modified, thin-bed mortar. AAC panels are lifted and placed using specially designed clamps, and are aligned using alignment bars.


9

When AAC elements are used as a load-bearing wall system, the floor and roof systems are usually designed and detailed as horizontal diaphragms to transfer lateral loads to shear walls. The tops of the panels are connected to the floor or roof diaphragms using a cast-in-place reinforced concrete ring beam. AAC floor and floor panels can be erected on concrete, steel or masonry construction. All bearing surfaces should be level and minimum required bearing areas (to prevent local crushing) should be maintained. Most floor and roof panels are connected by keyed joints that are reinforced and filled with grout to lock the panels together and provide diaphragm action to resist lateral loads. A cast-in-place reinforced concrete ring beam is normally placed along the perimeter of the diaphragm, completing the system.

1.3.5 Electrical and plumbing installations in AAC Electrical and plumbing installations in AAC are placed in routed chases. Care should be taken when laying out chases to ensure that the structural integrity of the AAC elements is maintained. Do not cut reinforcing steel or reduce the structural thickness of the AAC elements in critical areas. When analyzing the vertically spanning AAC element, horizontal routing should be permitted only in areas with low flexural and compressive stresses. In contrast, when the AAC element is intended to span horizontally, vertical routing should be minimized. When possible, it may be advantageous to provide designated chases for large quantities of conduit or plumbing.

1.3.6 Exterior finishes for AAC Unprotected exterior AAC deteriorates when exposed to cycles of freezing and thawing while saturated. To prevent such freeze-thaw deterioration, and to enhance the aesthetics and abrasion resistance of AAC, exterior finishes should be used. They should be compatible with the underlying AAC in terms of thermal expansion and modulus of elasticity, and should be vapor permeable. Many different types of exterior finishes are available, and the most common are discussed here. Polymer-modified stuccos, paints or finish systems are the most common exterior finish for AAC. They increase the AAC’s water-penetration resistance while allowing the passage of water vapor. Heavy acrylic-based paints containing aggregates are also used to increase abrasion resistance. There is generally no need to level the surface, and horizontal and vertical joints may be chamfered as an architectural feature, or may be filled.

10


Masonry veneer may be used over AAC panels in much the same way that it is used over other materials. The veneer is attached to the AAC wall using masonry ties. The space between the AAC and the masonry can be left open (forming a drainage wall), or can be filled with mortar. When AAC panels are used in contact with moist or saturated soil (for example, in basement walls) the surface in contact with the soil should be coated with a waterproof material or membrane. The interior surface should either remain uncoated, or be coated with a vapor-permeable interior finish.

1.3.7 Interior finishes for AAC Interior finishes are used to enhance the aesthetics and durability of AAC. They should be compatible with the underlying AAC in terms of thermal expansion and modulus of elasticity, and should be vapor permeable. Many different types of interior finishes are available, and the most common are discussed here. Interior AAC wall panels may have a thin coat of a mineral-based plaster to achieve a smooth finished surface. Lightweight interior gypsum-based plaster may provide a thicker coating to level and straighten walls, and to provide a base for decorative interior paints or wall finishes. Interior plasters have bonding agents to enhance their adhesion and flexibility, and are commonly installed by either spraying or troweling. When applied to the interior surface of exterior AAC walls, gypsum board should be attached using pressure-treated furring strips. When applied to interior walls, moisture-resistant gypsum board can be applied directly to the AAC surface. For commercial applications requiring high durability and low maintenance, acrylic-based coatings are often used. Some contain aggregates to enhance abrasion resistance. When ceramic wall tile is to be applied over AAC, surface preparation is normally necessary only when the AAC surface requires leveling. In such cases, a Portland cement- or gypsum-based parge coat is applied to the AAC surface before setting the ceramic tile. The ceramic tile should then be adhered to the parged wall using either a cement-based thin-set mortar or an organic adhesive. In moist areas such as showers, only a Portland cement-based parge coat should be used, and the ceramic tile should be set with cement-based thin-set mortar only.

1.3.8 Typical construction details for AAC elements A wide range of construction details for AAC elements is available on the web sites of individual manufacturers, accessible through the web site of the AACPA. An example is given in Fig. 1.6.


11

Load-bearing vertical wall system

Interior bearing wall

1/2″

Exterior bearing wall

1/2″

1.6 Construction details for load-bearing wall panels (courtesy Aercon Florida).

1.4

Seismic design of autoclaved aerated concrete structures

Because it has been used extensively in Europe for more than 70 years, AAC has been extensively researched there (RILEM 1993). Outside of the US, seismic qualification of AAC components and structures is based on experience in the Middle East and Japan. In the US, it is based indirectly

12


on that experience, and directly on an extensive experimental and analytical research program conducted at the University of Texas at Austin, and described further here and in Tanner et al. (2005a,b), Varela et al. (2006) and Klingner et al. (2005a,b). That research program developed design models, draft design provisions, and seismic design factors (R and Cd). In the rest of this chapter, the US approach to seismic design of AAC structures is summarized; a design example is presented; and the research background for the design procedure is reviewed.

1.4.1 Basic earthquake resistance mechanism of AAC structures Structures whose basic earthquake resistance depends on AAC elements are generally shear-wall structures. Lateral earthquake loads are carried by horizontal diaphragms to AAC shear walls, which transfer those loads to the ground. General response of shear-wall structures to lateral loads is discussed in the Masonry Designers’ Guide (MDG 2006), and is not repeated here. Earthquake design of AAC shear-wall structures is similar to earthquake design of conventional masonry shear-wall structures. A complete design example is given later in this document. The technical justification for the design steps is given in the Commentary to Appendix A of the 2005 MSJC Code and Specification, and is also discussed at the end of this chapter.

1.4.2 Seismic design factors (R and Cd) for ductile AAC shear-wall structures in the US Because AAC structures (whether of masonry units or reinforced panels) in practically all parts of the US must be designed for earthquake loads, it is necessary to develop seismic design factors (R and Cd) for use with ASCE 7, the seismic load document referenced by model codes such as the 2003 IBC. The seismic force-reduction factor (R) is intended to account for ductility, and for structural overstrength. It is based on observation of the performance of different structural systems in previous strong earthquakes, on technical justification, and on tradition. Because AAC is a new material in the US, its seismic design factors (R and Cd) must be based on laboratory test results and numerical simulation of the response of AAC structures to earthquake ground motions. The proposed factors must then be verified against the observed response of AAC structures in strong earthquakes. Values of R and Cd for ductile AAC shear-wall structures have been proposed in two code-development forums.

Autoclaved aerated concrete •

•

13

In October 2002, seismic design factors were proposed to and approved by ICC ES (a model-code evaluation service), as part of a proposed IC ES listing for AAC structural components and systems produced by members of the Autoclaved Aerated Concrete Products Association (AACPA). That listing is intended to make it easier to use such systems throughout the US, until the consensus design provisions proposed above are incorporated in MSJC and ACI documents, and are referenced by model codes. In 2005 and 2006, the same seismic design factors were considered by the Building Seismic Safety Council and by the International Code Council. In September 2006, the ICC Structural Committee approved the R and Cd values, shown in Table 1.4, for reinforced AAC masonry, for inclusion in the 2007 IBC Supplement. The values will become final if they are sustained in ICC public comment hearings in May 2007.

1.4.3 ASTM specifications for AAC construction ASTM traditionally deals with specifications for materials and methods of test. For the past several years, standards-development work regarding AAC has been going on in two ASTM committees: •

In 1998, ASTM Subcommittee C-27.60 (Precast Concrete Elements of AAC) developed a material standard for AAC: C1386-98 (Standard Specification for Precast Autoclaved Aerated Concrete Wall Units). Subcommittee C27-60 has also developed a standard for reinforced AAC panels: C1452-00 (Standard Specification for Reinforced Autoclaved Aerated Concrete Units). That subcommittee has also developed a standard method of test for determining the modulus of AAC: C1591: (2004) (Standard Test Method for Determining the Modular of Elasticity of AAC).

Table 1.4 R and Cd values for reinforced AAC masonry Response modification coefficient, R

2

System overstrength factor, Ω0

2.5

Deflection amplification factor, Cd

2

System limitations and building height limitations (feet) by seismic design category as determined in Section 1616.3 A or B

C

D

E

F

NL

35

NP

NP

NP

14 •

Developments in the formulation and reinforcement of concrete In 2003, ASTM Subcommittee C-15.10 (Autoclaved Aerated Concrete Masonry) developed a standard for AAC masonry: C1555-03a (Standard Practice for Autoclaved Aerated Concrete Masonry). That standard references the AAC material provisions of ASTM C1386-98, and also contains construction provisions. It has been incorporated into the 2005 MSJC Specification.

1.5

Design example: three-story autoclaved aerated concrete shear-wall hotel

This example illustrates the preliminary design of a three-story AAC shearwall hotel in Asheville, North Carolina, a zone of moderate seismic risk, using the loading provisions of the 2003 IBC. The principal lateral forceresisting elements of the structure are transverse shear walls. This example problem is carried out using the AAC masonry design and detailing provisions of the 2005 MSJC Code and Specification.

1.5.1 Design steps 1)

Choose design criteria: • propose plan, elevation, materials, fAAC′ • calculate D, L, W, E loads • propose structural systems for gravity and lateral load 2) Design transverse shear walls for gravity and earthquake loads 3) Design exterior walls for gravity and wind loads • earthquake loads will be carried by longitudinal walls in-plane • out-of-plane wind loads will be carried by longitudinal walls out-ofplane using vertical and horizontal strips

1.5.2 Step 1: Choose design criteria The plan and elevation of the building are shown in Figs 1.7 and 1.8. Architectural constraints Water-penetration resistance:

Movement joints:

A single-width AAC masonry wall will be used. Exterior protection will be provided by low-modulus acrylic stucco. To control crack widths from shrinkage of AAC walls, use vertical control joints every bay.


15

North

Elevator

8-in. AAC planks, untopped

20 ft

Stairs

10 ft 20 ft

20 ft typ 20 ft typ

7 @ 20 ft = 140 ft

1.7 Design criteria: plan of the building. Roof, parapet R 3 3 @ 11 ft 2 1 Typical facade

1.8 Design criteria: elevation of the building.

Design for fire Use and occupancy: Group B Use Type I or Type II construction (noncombustible material) No area or height restrictions 2- or 3-hour rating required Must meet separation requirements of Table 602 of the 2003 IBC Bearing walls: 4-hr rating (8-in. nominal AAC masonry OK) Shafts: 2-hr rating (8-in. nominal AAC masonry OK) Floors: 2-hr rating (planks and topping OK) Specify materials 12-in. AAC masonry units (ASTM C1555), fully mortared Thin-bed mortar (ASTM C1555)

16


Class 6 AAC (fAAC′ = 6 MPa or 870 psi), assumed unit weight 45 pcf Deformed reinforcement meeting ASTM A615, Gr. 60 Floors and roof of untopped AAC planks with diaphragm reinforcement Structural systems Gravity load:

Lateral load:

Gravity load on roof and floors will be transferred to transverse walls. Gravity load on corridor will be transferred to spine walls. Lateral load (earthquake will govern) will be transferred by floor and roof diaphragms to the transverse shear walls, which will act as statically determinate cantilevers. (See Tables 1.5 and 1.6.)

1.5.3 Summary of design procedure Calculate design lateral load from earthquake (All section and table references are to the 2003 International Building Code) 1)

Determine the structure’s seismic use group (related to nature of occupancy), and select the corresponding seismic importance factor, IE, in accordance with Section 1616.2 and Table 1604.5. 2) Determine the site class (A through F) in accordance with Section 1615.1.1 and Table 1615.1.1. Table 1.5 Calculate design roof load due to gravity Dead load

Planks EPDM membrane, gravel HVAC, roofing

Live load

Ignore reduction of live load based on tributary area

30 lb/ft2 20 lb/ft2 30 lb/ft2 80 lb/ft2 total 20 lb/ft2

Table 1.6 Calculate design floor load due to gravity Dead load

Planks HVAC, floor finish, partitions

Live load

Use weighted average of corridor and guest rooms. Ignore reduction of live load based on tributary area

30 lb/ft2 20 lb/ft2 50 lb/ft2 total 60 lb/ft2

Autoclaved aerated concrete 3)

4)

5) 6)

7)

8)

17

Determine the ordinates of the maximum considered response acceleration for short periods, SMS, and for a 1-second period, SM1, depending on the geographical location of the structure (Figure 1615), and adjusted for site class effects, in accordance with Section 1615.1.2 (Equations 16-38 and 16-39). Determine the key ordinates of the design response spectrum for short periods, SDS, and at 1 second, SD1, as two-thirds the values determined in Step 3 above (Section 1615.1.3, Equations 16-40 and 16-41). Determine the design response spectrum using the key ordinates from Step 4 and Section 1615.1.4 (Equations 16-42 and 16-43). Determine the structure’s seismic design category (A through F) based on its seismic use group (Step 1), the key ordinate SDS, and the key ordinate SD1, using Section 1616.3, Table 1616.3(1), and Table 1616.3(2). Determine the required design approach for each seismic design category in accordance with Section 1616.6, including, for higher seismic design categories, the effects of plan structural irregularities (Table 1616.5.1.1) and vertical structural irregularities (Table 1616.5.1.2). Determine the seismic load effect, E and Em, for use in the load combinations of Section 1605, including the effects of redundancy (r), system overstrength (Ω0), in accordance with Section 1617.

Now we will discuss each step in more detail, with our example for Asheville, North Carolina. Step 1: Determine the structure’s seismic use group (related to nature of occupancy), and select the corresponding seismic importance factor, IE, in accordance with Section 1616.2 and Table 1604.5 In accordance with Section 1616.2, • •

•

Seismic Use Group II structures are those whose failure would result in substantial public hazard, or so designated by the building official; Seismic Use Group III structures are those containing essential facilities required for post-earthquake recovery, or so designated by the building official; and Seismic Use Group I structures are those not assigned to Seismic Use Group II or III.

Step 2: Determine the site class (A through F) in accordance with Section 1615.1.1 and Table 1615.1.1 In accordance with Table 1615.1.1, site classes are assigned as shown in Table 1.7.

18


Table 1.7 Site class definitions Site class

Soil profile name

A B C D E E

Hard rock Rock Very dense soil and soft rock Stiff soil profile Soft soil profile –

F

–

Average properties in top 100 ft

其

Described in terms of soil shear wave velocity, standard penetration resistance, and undrained shear strength

Described in terms of plasticity index, moisture content, and undrained shear strength Described in terms of vulnerability to liquefaction or collapse, high organic content, very high plasticity, or very high flexibility

Step 3: Determine the ordinates of the maximum considered response acceleration for short periods, SMS, and for a 1-second period, SM1, depending on the geographical location of the structure (Figure 1615), and adjusted for site class effects, in accordance with Section 1615.1.2 (Equations 16-38 and 16-39) Determine the maximum considered earthquake acceleration response in % g for short periods (Figure 1615(1)) and for 1-second periods (Figure 1615(2)) as a function of geographical location. For Asheville, North Carolina, for example, SS = 0.40 g, and S1 = 0.13 g. For illustration, assume Site Class D (stiff soil profile). Then the acceleration-dependent site coefficient, Fa, is 1.48 (Table 1615.1.2(1)), and the velocity-dependent site coefficient, Fv, is 2.28 (interpolating in Table 1615.1.2(2)). Then the maximum considered short-period response acceleration is: SMS = Fa ⋅ SS = 1.48 ⋅ 0.40 g = 0.59 g And the maximum considered 1-second response acceleration is: SM 1 = Fv ⋅ S1 = 2.28 ⋅ 0.13 g = 0.30 g Step 4: Determine the key ordinates of the design response spectrum for short periods, SDS, and at 1 second, SD1, as two-thirds the values determined in Step 3 above (Section 1615.1.3, Equations 16-40 and 16-41) These design spectral ordinates correspond to values with a 10% probability of exceedance within a 50-year period.


19

Continuing with our example for Asheville, North Carolina, the design response acceleration for short periods is: SDS =

2 2 ⋅ SMS = ⋅ 0.59 g = 0.39 g 3 3

and the design response acceleration for a 1-second period is: SD1 =

2 2 ⋅ SM 1 = ⋅ 0.30 g = 0.20 g 3 3

Step 5: Determine the design response spectrum using the key ordinates from Step 4 and Section 1615.1.4 (Equations 16-42 and 16-43) Define T0 ≡ •

SD1 S and TS = D1 . Then: SDS SDS

For periods less than or equal to T0, the design spectral response acceleration, Sa, is given by Equation 16-20: Sa = 0.6

• •

SDS T + 0.4SDS T0

(Equation 16-20)

For periods greater than T0 and less than or equal to TS, the design spectral response acceleration, Sa, is equal to SDS. For periods greater than TS, the design spectral response acceleration, Sa, is given by Equation 16-21: S=

SD1 T

(Equation 16-21)

The resulting design acceleration response spectrum is given in Fig. 1.9.

Response acceleration, g

0.45 0.4 0.35 0.3 0.25 0.2 0.15 0.1 0.05 0 0

0.2

0.4

0.6

1 0.8 Period, sec

1.2

1.4

1.6

1.8

1.9 Design response spectrum for Asheville, NC site class D (stiff soil).

20


Step 6: Determine the structure’s seismic design category (A through F) based on its seismic use group (Step 1), the key ordinate SDS, and the key ordinate SD1, using Section 1616.3, Table 1616.3(1), and Table 1616.3(2) According to Table 1616.3(1), a structure in Seismic Use Group I (the default case), and with a short-period design spectral ordinate, SDS, of 0.39 g, is assigned to seismic design category C (Table 1.8). The asterisk refers to footnotes that do not apply in this case. According to Table 1616.3(2), a structure in Seismic Use Group I (the default case), and with a 1-sec period design spectral ordinate, SD1, of 0.48 g, is assigned to seismic design category C (Table 1.9). The asterisk refers to footnotes that do not apply in this case. The two tables assign the structure to the same seismic design category, C. Had they assigned the structure to different categories, the more severe classification would have governed. Because the structure is less than or equal to 35 ft in height, and is assigned to SDC C, it can be designed as ordinary reinforced AAC masonry according to the draft provisions of the 2007 Supplement to the 2006 IBC. Continue with the design from Fig 1.9:

Table 1.8 Seismic design category based on short-period response accelerations Value of SDS

SDS < 0.167 g 0.167 g ≤ SDS < 0.33 g 0.33 g ≤ SDS < 0.50 g 0.50 g ≤ SDS

Seismic use group I

II

III

A B C D*

A B C D*

A C D D*

Table 1.9 Seismic design category based on 1-second period response accelerations Value of SD1

SDS < 0.167 g 0.067 g ≤ SDS < 0.133 g 0.133 g ≤ SDS < 0.20 g 0.20 g ≤ SDS

Seismic use group I

II

III

A B C D*

A B C D*

A C D D*


21

Step 7: Determine the required design approach for each seismic design category in accordance with Section 1616.6, including, for higher seismic design categories, the effects of plan structural irregularities (Table 1616.5.1.1) and vertical structural irregularities (Table 1616.5.1.2) Plan structural irregularities include: • • • •

plan eccentricities between the center of mass and the center of stiffness; re-entrant corners; out-of-plane offsets; non-parallel systems.

These can increase seismic response. Vertical structural irregularities include: • • • • •

stiffness irregularity; mass irregularity; vertical geometric irregularity; in-plane discontinuity in vertical lateral-force-resisting elements; discontinuity in capacity – weak story.

These can also increase seismic response. The building under consideration here has no plan or vertical structural irregularities. Step 8: Determine the seismic load effect, E and Em, for use in the load combinations of Section 1605, including the effects of redundancy (ρ) and system overstrength (Ω0), in accordance with Section 1617 Overstrength need not be addressed, because the lateral-force-resisting system is statically determinate. Redundancy must be addressed. In our case, redundancy is the same on every floor:

ρ = 2−

20 rmax A

(Equation 16-54)

For shear walls, r is the maximum value of the product of the shear in the wall or wall pier and 10/lw, divided by the story shear. Because there are 16 uncoupled transverse shear walls, the shear per wall is 1/16 times the total story shear, and r=

( 161 ) ⋅ ( 1020 ) = 0.0313

22


The area A refers to the diaphragm level above the story in question:

ρ = 2− r = 2−

20 rmax A

20

0.0313 3 × 50 × 140

r = 2 − 4.41 r = −2.41 However, ρ must not be less than 1.0, and need not exceed 1.5 (2003 IBC, Section 1617.2.1). The former value governs in this case. Finally, in accordance with the 2003 IBC, Section 1617.1.1, the design seismic load effect E is E = rQE Now compute the seismic base shear. In accordance with the 2003 IBC, Section 1617.4, seismic base shear is to be computed using the procedures of ASCE 7-02, Section 9.5.5.: V = CsW where Cs is the design seismic coefficient (defined below), and W is the building weight. In accordance with ASCE 7-02, Section 9.5.5.2.1: Cs = In our case:

SDS R I

( )

(Equations 16-36, 16-37)

SDS = 0.39 g R =2 (ordinary AAC masonry shear wall systems) I = 1.00 (ASCE 7-02, Table 9.1.4) Therefore SDS 0.39 = = 0.20 R 2 I 1 This is multiplied by the redundancy factor of 1.0, giving a product of 0.40. In other words, the building must be designed for 40% of its weight, applied as a lateral force. This force is distributed triangularly over the height of the building. The weight of a typical floor is its area, times the dead load per square foot, plus the interior transverse wall weight, plus the spine wall weight, plus the weight of the exterior walls. For simplicity, assume that the roof weighs the same as a typical floor, and ignore the parapet. Cs =

( ) ()

Floor weight: 50 lb/ft2 × 50 × 140 ft2 = 350 kips Transverse wall weight: 7 × 20 × 11 ft2 × 45 lb/ft2 = 69.3 kips


23

Spine wall weight: 2 × 130 × 11 ft2 × 45 lb/ft2 = 128.7 kips Perimeter wall weight: 2 × (140 + 50) × 12 ft2 × 45 lb/ft2 = 188.1 kips Total weight of a typical floor is 736.1 kips. The design base shear (Table 1.10) is calculated assuming a linear distribution of forces over the height of the structure. Total design base shear is 2208 kips × 0.20 = 441.7 kips. At the roof level, the factored design lateral force is the factored design base shear (441.7 kips), multiplied by 0.50 (the quotient of WH/SUM) for the triangular distribution, or 220.9 kips. At the next level down, the factored design lateral force is 441.7 kips, multiplied by 0.333, and so forth. At each level, the factored design moment is the summation of the products of the factored design lateral forces above that level, each multiplied by its respective height above that level. The load factor for seismic loads is 1.0. Factored design shear and moment diagrams for the building are shown in Table 1.11 and Fig. 1.10. Table 1.10 Design base shear Level

Weight

Height

W×H

WH/SUM

R 3 2

736.1 736.1 736.1 2208.3

33 22 11

24,291 16,194 8,097 48,582

0.50 0.333 0.167

Table 1.11 Factored design shear Level

Fu, k

H, ft

Vu, k

Mu, k-ft

R 3 2

220.9 147.2 73.6

33 22 11

220.9 368.1 441.7

0 5923 29,616 44,424

0

220.9 k 147.2 k 73.6 k

2,430

220.9 368.1

6,479

441.7

Vu, kips

1.10 Moment diagrams for the building.

11,338 Mu, kip-ft

24


20 ft

4 ft

1.11 Transverse wall I beam.

1.5.4 Step 2: Design building for gravity plus earthquake loads (All references are to the 2005 MSJC Code and Specification) The transverse direction is critical for this building. The 16 transverse walls are conservatively assumed to be uncoupled, so that each functions as an independent cantilever. Design each transverse wall as I beam, assuming flange widths of 4 ft (Fig. 1.11). Shear design of a typical transverse wall for earthquake loads From the 2005 MSJC Code, Appendix 3.4.1.2.1: fVAAC = f 0.95 w t f AAC 1+ ′

Pu 2.4 f AAC ′ wt

Include the effects of axial load, assuming that a typical transverse wall carries its self-weight plus the distributed floor weight on a tributary width of 20 ft: Self-weight of wall: 9.9 kips/floor Floor weight: 60 lb/ft2 × 20 × 20 = 24 kips/floor Total unfactored axial load at base, P, is 3 × (9.9 + 24) = 101.7 kips. Assume that the critical load case is Pu = 0.9D. Then fVAAC = 0.8 × 0.95 × 240 × 11.9 in.2 870 1 +

0.9 × 101, 700 2.4 870 × 240 × 11.9

fVAAC = 0.8 × 0.95 × 240 × 11.9 in.2 870 1 + 0.45 fVAAC = 77, 165 lb


25

This exceeds (1/16 walls) times the factored design base shear (1/16 × 441.7 kips = 27.6 kips), and the transverse walls are satisfactory for shear thus far. While floor-level bond beams are required, no shear reinforcement is required. Check sliding at the base of the wall. By A.1.8.5 and A.3.4.1.2.3: VAAC = m ACC Pu Conservatively assume that Pu comes from 0.9D only: VAAC = 1.0 × 0.9 × 101.7 kips = 91.5 kips fVAAC = 0.8 × 91.5 = 73.2 kips Sliding shear resistance governs. In-plane flexural design of transverse shear walls for earthquake loads Each transverse shear wall has a plan length of 20 ft. The factored base moment per wall is (1/16) × 11,338 ft-kips, or 708.63 ft-kips. The critical load case is 0.9D + 1.0E. The factored axial load (see above) is 0.9 × 101.7 kips, or 91.5 kips. Using a spreadsheet, the interaction diagram for the wall is shown in Fig. 1.12. Flexural reinforcement consisting of 1 #4 bar at each end is required. The bars should be placed in grouted cores at least 12-in. square (at intersections of web and flanges). Check splice requirements and percent area requirements. Assume a 2000-psi grout strength by the proportion specification of ASTM C476. By A.3.3.3.1: 1600 1400

φ Pn, kips

1200 1000 800 600 400 200 0 0

500

1000

1500

2000 2500 φ Mn, ft-kips

3000

3500

4000

1.12 Strength interaction diagram by spreadsheet AAC transverse shear wall f ′AAC = 870 psi, 20 ft long, 12 in. thick, #4 bars at ends.

4500

26

Developments in the formulation and reinforcement of concrete ld =

0.13db2 f yg K AAC fg′

=

0.13 × 0.52 × 60, 000 × 1.0 = 7.58 in. 12 − 0.5 2, 000 2

(

)

12 inches governs. By Code A.3.3.1, the maximum percent area in a plastic hinge zone is 3%. For a 12-in. square core, a #4 bar easily satisfies this requirement. Because the wall is symmetrically reinforced, maximum reinforcement limitations (Code A.3.3.5) are satisfied. Now check capacity design for shear (Code Section A.1.3). First try to meet the capacity design provisions of that section. At an axial load of 91.5 kips, the nominal flexural capacity of this wall is 750 ft-kips, divided by the strength reduction factor of 0.9, or 833 ft-kips. The ratio of this nominal flexural capacity to the factored design moment is 833 divided by 708.6, or 1.18. Including the additional factor of 1.25, that gives a ratio of 1.47:

φVn ≥ 1.47Vu Vn ≥

1.47 1.47 Vu = Vu = 1.84Vu = 1.84 × 27.6 = 50.7 kips φ 0.8

Vn (governed by sliding shear) is 91.5 kips, considerably greater than this. The wall is satisfactory without shear reinforcement. A nominal 8-in. wall could probably be used instead of 12 in. Comments • The most laborious part of this design is calculation of the design lateral force for earthquake loads. Once that calculation is done, design of the lateral-force-resisting system is straightforward, even for a region of moderate seismic risk such as Knoxville. • This structural system could be designed for increased capacity. Increased flexural capacity would be quite easy to achieve, but increased shear capacity (to meet capacity design requirements) would probably require intermediate bond beams.

1.5.5 Step 3: Design exterior walls for gravity plus out-of-plane wind (All references are to the 2005 MSJC Code and Specification) The critical panel will be at the top of the building, where the wind load is highest. The panel must be designed for out-of-plane wind. Load effects in vertical jamb strips will be increased by the ratio of the plan length of openings to the total plan length.


27

Use factored wind load (components and cladding) on wall ρu = 50 lbs/ft2 Reinforcement in a 3-in. grouted cell at 4 ft. on center (Fig. 1.13): h = 11ft t = 12 in. The wall is considered simply supported at top and bottom (Fig. 1.14). Flexural capacity a)

Determine design moment wu = p ⋅ width = 50 lb/ft 2 ⋅ 4 ft = 200 lb/ft Mu =

(Section A.3.2)

wl 2 200 ⋅ (11)2 = = 3, 025lb / ft = 36, 300 lb / in. 8 8

48 in.

1.13 Plan view of 4 ft section of wall.

1.14 Wall supported at top and bottom.

28 b)

Developments in the formulation and reinforcement of concrete Try #4 bar T = As f y = 0.20 ⋅ 60, 000 = 12, 000 lb a=

T 12, 000 = = 0.34 in. 0.85 f ACC ′ b 0.85 ⋅ 870 ⋅ 48

( a2 ) = 12.0 ⋅ (6 − 0.234 ) = 70, 000 lb/in.

Mn = As ⋅ f y ⋅ d −

ΦMn = 0.9 ∗ 70, 000 lb/in. = 63, 000 lb/in. > Mu OK Use a #4 bar. Outside of a plastic hinge zone, Code A.3.3.1 imposes a maximum bar area of 4.5% of the cell. Using a 3-in. grouted core, the area ratio is (0.5/3)2, or 0.028, easily satisfying the requirement. This bar size will easily satisfy the maximum reinforcement limitations of A.3.3.5 for out-ofplane flexure, and the design is satisfactory for flexure.

Shear capacity (a)

Determine factored loads and maximum shear force for a single panel. wu = 50 psf, and the panel is 4 ft wide: Vu =

(b)

200 ⋅ 11 = 1, 100 lb 2

Determine shear capacity of panel: VAAC = 0.9 f AAC ′ An + 0.05Pu = 0.9 870 ⋅ 48 ⋅ 6 = 7645 lb ΦVAAC = 0.8 ∗ (7645) = 6, 116 lb > Vu = 1, 100 lb

OK

1.5.6 Step 4: Design floor diaphragms for in-plane actions Design requirements for AAC floor diaphragms are not given in the 2005 MSJC Code and Specification, because that can be applied to many different types of floor systems. The design procedure given here is based on the requirements of the ICC ES 215 evaluation report, which was developed based on research at the University of Texas at Austin. The procedure is also given at the AACPA website (www.aacpa.org). f AAC ′ = 870 psi fgrout ′ = 2000 psi


29

f y = 60, 000 psi Ring beam reinforcement 2 #5 Grouted key reinforcement 1 #5 Factored transverse lateral load in each bay, Fu = 220.9 kips/16 bays = 13.81 kips (a)

Design diaphragm for flexure, assuming that load is uniformly distributed along span (Fig. 1.15)

Plan view of diaphragm

Fu

Grouted keys

Ring beam

240 in.

(a) Elevation

b = 240 in.

(b)

1.15 (a) Design diaphragm for flexure, assuming that load is uniformly distributed along span, plan view; (b) Design diaphragm for flexure, assuming that load is uniformly distributed along span, elevation view.

30

Developments in the formulation and reinforcement of concrete M=

wu l 2 Fu × l 13, 810 ⋅ 240 = = = 414, 188 lb/in. 8 8 8

T = As f y = 2 ⋅ 0.31 ⋅ 60, 000 = 27, 200 lb a=

C 37, 200 = = 0.09 in. 0.85 fgrout ′ b 0.85(2000)(240)

d = length of key − ring beam/2 − 2 ∗ U-block thickness = 240 − 4 − 4 = 2387 in.

( a2 ) = 0.9 ⋅ 37, 200 ⋅ (238 − 0.09)

ΦMn = ΦAs f y ⋅ d −

= −7, 970, 000 lb/in. ≥ Mu (b)

OK

Design diaphragm for shear based on adhesion (i) Panel-to-panel joint (Fig. 1.16(a))

The total resistance is the adhesion of the grouted area plus the adhesion of the thin-bed mortar area. bgrout = 5in. bthin−bed = 3in. Vgrout = t grout ⋅ bgrout ⋅ l = 36 ⋅ 5 ⋅ 240 = 43, 200 lb Vthin − bed = t thin − bed ⋅ bthin − bed ⋅ l = 18 ⋅ 3 ⋅ 240 = 13, 000 lb Vtotal = Vgrout + Vthin − bed = 55, 200 lb

φVtotal = 0.67 ⋅ 55, 200 lb = 36, 980 lb > Vu = (ii)

Fu = 6, 900 lb OK 2

OK

Panel bond beam joint (Fig. 1.16(b))

bgrout = 8 in. Vgrout = t grout ⋅ bgrout ⋅ l = 36 ⋅ 8 ⋅ 240 = 69, 100 lb Fu = 6, 900 lb 2 Design diaphragm for shear based on truss model fVtotal = 0.67 ⋅ 69, 100 = 46, 300 lb > Vn

(c)

OK

One #5 bar in each grouted key. Each plank is 2 ft wide, so there are 10 planks. The load applied to each node is 1/10 of the total load, or 1.38 kips.


31

Section D - D AAC floor panel

Grouted key AAC joint

Thin bed mortar at AAC joint (a)

Bond beam

Bond beam AAC joint

AAC floor panel

(b)

1.16 (a) Design diaphragm for shear based on adhesion: panel-topanel joint; (b) Design diaphragm for shear based on adhesion: panel bond beam joint.

In this model the compression chords act as diagonal compression members. There are two types of nodes: loaded nodes (on the upper side of Fig. 1.17) and unloaded nodes (on the lower side) (Fig. 1.18). The critical diagonal compression occurs in the panels next to the support. The component of that compression parallel to the transverse walls is onehalf the total factored load on the panel, or one-half of 13.81 kips, or 6.91 kips. The total compressive force in the diagonal, and also the tension force

32

Developments in the formulation and reinforcement of concrete Fiu Node 1

Node 3

Tension reinforcement

Compression strut

Node 2 Node 4

1.17 Design diaphragm for shear based on truss model.

Node 1

Node 2

Cpanel T grouted

Fiu T ring3

T ring3

Cpanel Cpanel Node 3

T ring1

Node 4

Cpanel Tring beam

T ring3 Cpanel T grouted

T ring2

T ring2

1.18 Loaded and unloaded nodes.


33

in the associated tension tie, is essentially that shear, because of the aspect ratio of the panels: C panel =

6.91 kips 6.91 kips × 20 2 + 2 2 = = 6.97 kips = Tgrouted 2 20 cos tan −1 20

(

)

Check capacity of compression strut: Wstrut = 6 in. Tpanel = 8 in. Fstrut = 6.94 kips/48 = 145 psi < 0.75 (0.85 f AAC ′ ) = 0.75(0.85)(870) = 555 psi OK Check capacity of tension tie in grouted key: Tgrouted key = 6.94 kips < ΦAs f y = 0.75 ∗ 0.31 ∗ 60, 000 = 14, 000 lb

OK

Tension ties in ring beams have already been checked, and are satisfactory.

1.5.7 Overall comments •

•

• •

Although it is located in a region of moderate seismic risk, this building needs comparatively little reinforcement, because of the large plan area of its bearing walls. Considerable simplicity in design and analysis was achieved by letting transverse shear walls resist lateral loads as statically determinate cantilevers. Design of AAC bearing walls is inexpensive and straightforward for this type of building. Many types of floor and roof system are possible with AAC. To adapt this design to other types of floor or roof elements, the unit weight would have to be changed appropriately; the connection details would have to be changed appropriately; and the diaphragm actions would have to be checked appropriately. For example, if hollow-core prestressed concrete planks were used, the unit weight would increase, and so would the seismic base shear and overturning moment. Shear design of the transverse shear walls would still govern. Details of the connections between walls and floor or roof would be similar to those used with the AAC planks. Shears in the horizontal diaphragms would be transferred in topping only.

34


1.6

Further background on earthquake performance of autoclaved aerated concrete shear-wall structures

1.6.1 Introduction to earthquake performance of AAC shear-wall structures Because the performance of AAC shear walls is fundamental to the earthquake resistance of AAC structures, the University of Texas studies first examined the performance of AAC shear walls under reversed cyclic loads like those imposed by strong earthquakes.

1.6.2 Brief description of AAC shear-wall tests at the University of Texas at Austin The Texas tests involved a group of 17 AAC shear-wall specimens, with aspect ratios (height divided by plan length) from 0.6 to 3. Some of the specimens were of AAC masonry; others were of reinforced panels, oriented horizontally; and still others were of reinforced panels, oriented vertically. To develop design equations for shear capacity, some of the specimens were designed to fail in shear, with low aspect ratios and heavy flexural reinforcement. Others were designed to fail in flexure, with high aspect ratios and light flexural reinforcement. AAC material strengths ranged from the lowest to the highest commercially available. The test setup is shown in Fig. 1.19.

Lateral load

External reinforcement Tie down boxes for rods

Axial load system

External reinforcement Tie down boxes for rods

1.19 Test setup for ACC shear-wall specimens (University of Texas at Austin).


35

1.6.3 Overall performance of AAC shear walls tested at UT Austin AAC shear walls tested at UT Austin behaved reliably. The sheardominated walls failed in a variety of mechanisms, discussed further below. The flexure-dominated walls failed, as expected, by yielding of the longitudinal reinforcement. As shown in Fig. 1.20, a typical flexure-dominated AAC shear wall reaches displacement ductilities of about 4, and story drifts of about 1% with little degradation in strength or stiffness. Shear-dominated walls failed by web-shear cracking, flexure-shear cracking, or sliding shear. Based on observed performance of AAC shear walls, design formulas were developed for each of these behaviors, for walls laid with unreinforced AAC masonry units, and also for walls made of reinforced AAC panels. The design formulas for AAC masonry have been incorporated as described above into the 2005 MSJC Code and Specification; the design formulas for reinforced AAC panels are being incorporated into the draft ACI 523A Design Guide.

1.6.4 Two-story AAC assemblage tested at UT Austin

30

133

20

88

10

44

0 –1.2

–0.8

–0.4

0.0

0.4

0.8

1.2

0

–10

–45

–20

–89

–30

–134 Drift ratio (%)

1.20 Typical load-displacement curves for flexure-dominated AAC shear wall.

Base shear (kN)

Base shear (kips)

To confirm the validity of the design procedures that had been developed based on tests of AAC shear walls, a complete two-story AAC assemblage, consisting of AAC shear walls and AAC floor planks, was built at UT Austin and tested under reversed cyclic loading.

36


Description of assemblage The two-story assemblage specimen consisted of two flanged walls connected by floor slabs (Fig. 1.21). The walls were constructed of vertically oriented AAC panels with internal reinforcement and additional fieldplaced longitudinal reinforcement, and the floor slabs were constructed of internally reinforced AAC panels. Vertically oriented panels were selected instead of AAC masonry units. On the upper level, the floor panels were oriented longitudinally; on the lower level, transversely. A complete description of the specimen is provided in Tanner (2005a). Vertically oriented, reinforced panels were selected because those had shown the most critical behavior in the shear wall tests described above. The reinforcement in the shear walls of the two-story assemblage Specimen consisted of flexural (longitudinal) reinforcement and foundation dowels (Fig. 1.22). The flexural reinforcement continued up the height of the specimen with a splice just above the first-story slab (first elevated slab). The dowels extended 24 in. (61 mm) above the foundation, and were included to increase the sliding shear capacity of the specimen. They were also placed at the level of the first elevated slab to prevent sliding at the bed of leveling mortar placed between the vertical panels and the slab.

North

1.21 Isometric view of two-story AAC assemblage specimen.


37

#4 Vertical reinforcement #5 Dowels

North

50

222

40

178

30

133

20

89

10

44

0

0

4

8

12

16

20

24

28

32

36

40

0

–10

–45

–20

–89

–30

–134

–40

–178

Force per story (kN)

Force per story (kips)

1.22 Plan view of horizontal section at base, showing flexural reinforcement and dowels in two-story AAC assemblage specimen.

–222

–50 Time

1.23 Loading history for two-story AAC assemblage specimen.

Instrumentation was used to measure global and local behavior of the two-story assemblage specimen. Measurements included applied loads, horizontal displacements, and various local deformations. Before the assemblage was tested, the AAC material was verified to have a compressive strength of 495 psi (3.4 MPa) and a splitting tensile strength of 45 psi (0.31 MPa). The assemblage was subjected to the reversed cyclic loading history shown in Fig. 1.23. After the maximum load was reached, the specimen was loaded under displacement control to tip displacements of 0.5 in. (13 mm), 0.8 in. (20 mm) and 1.5 in. (38 mm). A constant axial load was applied through the self-weight of the specimen and the loading equipment. The total axial load on the specimen is 60 kips (267 kN).

38


1.6.5 Summary of assemblage behavior Overall assemblage behavior is summarized in Fig. 1.24. Total base shear is the summation of the equal shears applied to each floor level. Positive displacements are to the south; negative, to the north. The displacements shown in the plot include base sliding. The following behaviors were observed in the two-story AAC assemblage: • • • • • •

flexural cracking at the base; minor vertical cracking on the north end of the east wall; web-shear cracking in the webs of both walls in the lower story; yielding of the flexural reinforcement at all four corners of the assemblage; separation of the flanges from the webs in the lower story; and separation of the vertical joint at the location of reinforcement at the northeast corner.

160

712

120

534

80

356

40

178

0 –1.0 –0.8 –0.6 –0.4 –0.2 –40

0.0

0.2

0.4

0.6

0.8

0 1.0 –178

–80

–356

–120

–534

–160

–712 Average drift ratio in % (south positive)

1.24 Overall hysteretic behavior of two-story AAC assemblage specimen.

Total base shear (kN)

Total base shear (kips)

The loads at which these behaviors were observed were consistent with the predictions based on previous tests of AAC shear walls, and also consistent with the design procedures and equations. After the flexural reinforcement yielded, both walls exhibited flexural behavior, consisting largely of rigid-body rocking. Vertical displacements were observed at the wall bases on the tension side, due to yielding and bond deterioration of the tensile reinforcement. Crushing of the compression toe was avoided, due to lateral support by the flanges. The tested flexural capacity of each wall fell within predicted limits corresponding to


39

neglecting the contribution of flexural dowels and including the contribution. After flexural yielding, distributed web-shear cracks continued to form in the walls at the lower level. Diagonal cracks formed around the dowels, separating those dowels from the webs of the AAC shear walls. This reduced the effectiveness of the dowel action, which in turn reduced the sliding-shear capacity of each wall. Degradation of dowel action is also identified by spalling of AAC around the diagonal cracks. Maximum slip between the AAC shear walls and the foundation exceeded 0.5 in. (13 mm). The wall displacement, corrected for this slip, is shown in Fig. 1.25. After three cycles of flexural rocking, to displacement drift ratios of 0.32% (loading south) and 0.24% (loading north), vertical cracks began to form at the interface between the web and the flanges. As the displacements increased, the flange panel did not slide with the web in the direction of loading, resulting in local damage to the flange and finally instability of the flange at both the north and south ends of the specimen. At the north end, the flange damage was accompanied by a large vertical crack in the east web. Testing of the two-story assemblage specimen was halted due to this damage. Final cracking patterns for each exterior face of the specimen are shown in Figs 1.26–1.28. The cracks shown in grey formed at the time of yielding of the flexural reinforcement; subsequent cracks are shown in black.

1.6.6 Summary of response of two-story AAC assemblage

80

356

60

267

40

178

20

89

0 –0.6 –20

–0.4

–0.2

0.0

0.2

0.4

0 0.6 –89 –178

–40 E Top W Top

–60

Force per story (kN)

Force per story (kips)

After removing the measured base slip, the two-story assemblage specimen reached drift ratios between 0.24% and 0.42%, and final displacement ductilities ranged from 2.8 to 5.8 (Table 1.12).

–267 –356

–80 Drift ratio in % (south positive)

1.25 Hysteretic behavior of two-story AAC assemblage specimen with slip removed.

40


1.26 Cracks in the east wall at the end of the test.

1.27 Cracks in the west wall at the end of the test.

1.28 Cracks in the south and north wall at the end of the test.


41

Table 1.12 Drift ratios and displacement ductilities for each wall, with base slip removal Shear wall/Direction of loading

Displacement ductility

Drift ratio (%)

East wall/south East wall/north West wall/south West wall/north

5.8 2.8 4.4 2.8

0.42 0.24 0.37 0.24

For design purposes, these results justify an assumption of an available flexural ductility of at least 3.0, reasonably consistent with that observed in previous tests of flexure-dominated AAC shear walls at UT Austin.

1.7

Development of seismic design factors (R and Cd) for ductile autoclaved aerated concrete shear-wall structures

The seismic force-reduction factor R specified in seismic design codes is intended to account for energy dissipation through inelastic deformation (ductility), and for structural overstrength. The factor R is based on observation of the performance of different structural systems in previous strong earthquakes, on technical justification, and on tradition. For innovative structures such as AAC structures, the force-reduction factor R and the corresponding displacement-amplification factor Cd must be based on laboratory test results and numerical simulation of the response of those structures subjected to earthquake ground motions. The proposed factors must then be verified against the observed response of those structures in strong earthquakes. Using the observed experimental response of the ductile (flexuredominated) AAC shear walls and the two-story assemblage tested at UT Austin, inelastic analytical models were developed and subjected to ground motions representing design earthquakes for highly seismic regions of the US. For flexure-dominated AAC structures, values of R and Cd were established that would result in a suitably low probability of failure under those design earthquakes, and under maximum considered earthquakes. For use with the 2000 IBC, the R value is 3, and the Cd value is also 3. Those values are consistent with the values for moderately reinforced shear walls of conventional masonry. ICC hearings have approved the lower values of R and Cd used in the example problem.

42


1.8

Acknowledgements

Some of the material in this chapter is adapted from a draft guide to AAC construction, under preparation by Committee 523A of the American Concrete Institute. The author acknowledges the contributions of colleagues to that material. The University of Texas research whose results are described here was supported financially and technically by the Autoclaved Aerated Concrete Products Association.

1.9

References

ACI 318-05 (2005): ACI 318-05 Building Code Requirements for Structural Concrete and Commentary, American Concrete Institute, Farmington Hills, Michigan, 2005. ACI 523A (2002) Autoclaved Aerated Cellular Concrete, American Concrete Institute, Farmington Hills, Michigan, 2002. ASCE 7-05 (2005): Minimum Design Loads for Buildings and Other Structures (ASCE 7-02), American Society of Civil Engineers, Reston, Virginia, 2005 (with Supplement). ASTM C1386 (1998): ASTM C1385 (Standard Specification for Precast Autoclaved Aerated Concrete (PAAC) Wall Construction Units), ASTM International, West Conshohocken, Pennsylvania. ASTM C1452 (2000): ASTM C1452 (Standard Specification for Reinforced Autoclaved Aerated Concrete Units), ASTM International, West Conshohocken, Pennsylvania. ASTM C476 (2002): Standard Specification for Grout for Masonry, ASTM International, West Conshohocken, Pennsylvania, 2002. ASTM C1555 (2003): ASTM C 1555 (2003a) (Standard Practice for Autoclaved Aerated Concrete Masonry), ASTM International, West Conshohocken, Pennsylvania. ASTM C1591 (2004): ASTM C 1591 (Standard Test Method for Determining the Modulus of Elasticity of AAC), ASTM International, West Conshohocken, Pennsylvania. ASTM A615 (2006): Standard Specification for Deformed and Plain Carbon-Steel Bars for Concrete, ASTM International, West Conshohocken, Pennsylvania, 2006. Barnett et al. (2005): Barnett, R. E., Tanner, J. E., Klingner, R. E. and Fouad, F. H. “Guide for Using Autoclaved Aerated Concrete Panels: I – Structural Design,” ACI Special Publication SP 226, Caijun Shi and Fouad H. Fouad, Editors, American Concrete Institute, Farmington Hills, Michigan, April 2005, pp. 17–28. IBC 2000: International Building Code, 2000 Edition, International Code Council, Washington, DC, 2000. IBC 2003: International Building Code, 2003 Edition, International Code Council, Falls Church, Virginia, 2003. IBC 2006: International Building Code, 2006 Edition, International Code Council, Washington, DC. ICC ES 215 (2003): “Acceptance Criteria for Seismic Design Factors and Coefficients for Seismic-Force-Resisting Systems of Autoclaved Aerated Concrete (AAC),”


43

Evaluation Report AC215, ICC Evaluation Service, Inc., Whittier, California, November 1, 2003. ICC ESR-1371 (2004): “Autoclaved Aerated Concrete (AAC) Block Masonry Units,” Evaluation Report ESR-1371, ICC Evaluation Service, Inc., Whittier, California, October 1, 2004. Klingner et al. (2005a): Klingner, R. E., Tanner, J. E., Varela, J. L., Brightman, M., Argudo, J. and Cancino, U., “Technical Justification for Proposed Design Provisions for AAC Structures: Introduction and Shear Wall Tests,” ACI Special Publication SP 226, Caijun Shi and Fouad H. Fouad, Editors, American Concrete Institute, Farmington Hills, Michigan, April 2005, pp. 45–66. Klingner et al. (2005b): Klingner, R. E., Tanner, J. E. and Varela, J. L., “Technical Justification for Proposed Design Provisions for AAC Structures: Assemblage Test and Development of R and Cd Factors,” ACI Special Publication SP 226, Caijun Shi and Fouad H. Fouad, Editors, American Concrete Institute, Farmington Hills, Michigan, April 2005, pp. 67–90. MDG (2006): Masonry Designers’ Guide, 5th edition, Phillip J. Samblanet, ed., The Masonry Society, Boulder Colorado, 2006. MSJC Code and Specification (2005): ACI 530-05 / ASCE 5-05 / TMS 402-05 (Building Code Requirements for Masonry Structures) and ACI 530.1-05 / ASCE 6-05 / TMS 602-05 (Specifications for Masonry Structures), American Concrete Institute, Farmington Hills, Michigan; American Society of Civil Engineers, Reston, Virginia; and The Masonry Society, Boulder, Colorado. RILEM (1993): Autoclaved Aerated Concrete: Properties, Testing and Design, RILEM Recommended Practice, RILEM Technical Committees 78-MCA and 51-ALC, E & FN Spon, London. Tanner et al. (2005a): Tanner, J. E., Varela, J. L., Klingner, R. E., “Design and Seismic Testing of a Two-story Full-scale Autoclaved Aerated Concrete (AAC) Assemblage Specimen,” Structures Journal, American Concrete Institute, Farmington Hills, Michigan, vol. 102, no. 1, January–February 2005, pp. 114–119. Tanner et al. (2005b): Tanner, J. E., Varela, J. L., Klingner, R. E., Brightman M. J. and Cancino, U., “Seismic Testing of Autoclaved Aerated Concrete (AAC) Shear Walls: A Comprehensive Review,” Structures Journal, American Concrete Institute, Farmington Hills, Michigan, vol. 102, no. 3, May–June 2005, pp. 374– 382. Varela et al. (2006): Varela, J. L., Tanner, J. E. and Klingner, R. E., “Development of Seismic Force-Reduction and Displacement Amplification Factors for AAC Structures,” EERI Spectra, vol. 22, no. 1, February 2006, pp. 267–286.

2 High-density and radiation-shielding concrete P LESSING, Idaho National Laboratory, USA

2.1

Introduction

2.1.1 Definition of high-density concrete High-density concrete is produced using special heavy aggregates and can have a density of up to 400 lb/ft3. Why would anyone be interested in a heavyweight concrete when low-density concrete has so many beneficial construction and insulation applications? The answer is that heavyweight concrete is primarily useful for nuclear radiation shielding but can also be useful for counterweights and blast shielding (applications where physical space is limited). Gamma-rays and X-rays can be shielded by a mass of material containing heavy atoms. To shield against neutrons it is necessary to have a mass of material that contains atoms that can both “thermalize” and capture neutrons. The hydrogen atom in water that is chemically bonded in concrete “thermalizes” the fast neutrons which may then be captured by other atoms such as boron which have high neutron-capture cross-sections. Reference books are available that cover aggregate characteristics, mix proportioning, and standards for conventional high-density concrete.1,2

2.1.2 Characteristics of various heavy aggregates Historically inexpensive high-density mineral aggregates such as barite, ferro-phosphorus, goethite, hematite, ilemnite, limonite, magnetite, and steel punching and steel shot have been used to produce high-density concrete. Table 2.1 shows typical densities of heavy aggregates and the concrete manufactured using these aggregates. Boron-containing additions such as colemanite, boron frits, and borocalcite have been used to improve the neutron-shielding properties of heavy concrete. However, they may negatively affect the setting and strength of the concrete. 44

High-density and radiation-shielding concrete

45

Table 2.1 Physical properties of traditional heavyweight aggregates and concrete Aggregate

Water % (retained or chemically bound)

Aggregate specific gravity

Aggregate bulk density lb/ft3

Concrete density lb/ft3

Geothite (hydrous iron ore) Barite (barium sulfate) Hematite (iron ores) Magnetite (iron mineral) Ferro-phosphorus (slag) Steel punchings or shot

10–11

3.4–3.7

130–140

180–200

0

4.0–4.6

145–160

210–230

–

4.9–45.3

180–200

240–260

–

4.2–5.2

150–190

210–260

0

5.8–6.8

200–260

255–330

0

6.2–7.8

230–290

290–380

2.1.3 Requirements for radiation-shielding concrete High level functional requirements for spent nuclear fuel storage applications were gathered into a report by Haelsig.3 Haelsig tabulates the requirements that he determined to be applicable to the conceptual design of a multi-purpose cask (MPC) and allocates them to MPC subsystems. In several appendices, Haelsig lists the design criteria of several vendor supplied concrete casks that were then licensed by the Nuclear Regulatory Commission (NRC). The design criteria/requirements vary by manufacturer, but are in the general categories of: service life, criticality safety limit, surface contact dose, dose for storage, storage facility dose limits, ambient environment, fuel cladding temperature limits, maximum decay heat power, concrete temperature limits, canister internal pressure limits and leak tightness, seismic ground accelerations, tornado loads, flooding loads basis, snow and ice loads, cask drops and fuel impact acceleration, and overall shipping package width on transport vehicle. Obviously, the mechanical and physical properties of a heavy concrete (e.g., attenuation of radiation, density, strength, thermal conductivity, etc.) have an impact on how a cask design meets most of these functional requirements. For instance, the Haelsig report calculates how the wall thickness will be greatly reduced using a heavy versus ordinary concrete while still meeting the radiation criteria.4 However, no physical properties are specified for the concrete (other than in-service temperature requirements). The physical properties of a specific grade of concrete need to be

46


Table 2.2 Maximum temperatures of concrete in storage cask Source

References cited

Normal “bulk” temp.

“Local” longterm temp.

Short-term (accident) temp.

Depleted Uranium Concrete Container Feasibility Study5

ACI-349 Appendix A and NRC Guidance ACI-349 Appendix A

66°C (150°F)

93–149°C (200–300°F)

177–343°C (350–650°F)

≤66°C (≤150°F)

≤93°C (≤200°F)

177(surface)– 343°C (local) (350–650°F)

66°C (150°F)

149°C (local) (300°F)

Pacific Nuclear “Nuhoms” System Sierra Nuclear “VSC” System6

Babcock & Wilcox Fuel Company7

ACI-349 Appendix A and NRC Guidance

121°C (250°F)

coupled with a specific cask design in order to meet the overall functional requirements. Table 2.2 shows the important aspect of temperature requirements for concrete spent fuel storage casks listed by various manufacturers. Pacific Nuclear and Sierra Nuclear appear to interpret NRC guidance somewhat differently. Pacific Nuclear’s criteria distinguish between duration and location (short versus long term, surface versus bulk, and local). Examining Table 2.2, it appears that the long-term maximum exposure (local) temperature for concrete in spent nuclear fuel storage casks should be less than 149°C (300°F). This is fairly consistent with a performance test sponsored by the Electric Power Research Institute (EPRI).8 A relevant standard from the American Cement Institute (ACI) standard 349 Appendix A.4 states: A.4.1- The following temperature limitations are for normal operation or any other long-term period. The temperatures shall not exceed 150°F except for local areas, such as around penetrations, which are allowed to have increased temperatures not to exceed 200°F . . .”

ACI-349 Appendix A.4 was obviously written for conventional concrete using rock aggregate such as quartzite and not for synthetic aggregate. Therefore, to qualify under ACI-349, long-term exposures should be kept under 200°F (93.3°C). Using this standard, concrete fabricated with synthetic aggregate would qualify if the strength did not deteriorate at temperatures from 90°C to 125°C.


2.2

47

Applications/case studies

In the United States, depleted uranium (uranium having 235U content less than natural uranium’s 0.711 wt%) has been generated as tails from uranium enrichment and spent fuel reprocessing. This comprises approximately 500,000 metric tons (uranium content). About 470,000 metric tons are stored as pure DUF6 in steel cylinders at US gaseous diffusion enrichment sites. There is approximately 225,000 metric tons of elemental fluorine associated with this stored DUF6. Intact cylinders normally contain DUF6 with purity exceeding 99.9%. In addition, the UDOE-EM currently owns about 19,500 metric tons of elemental uranium (MTU) in the form of high purity DUO3 resulting from historical weapons production programs at US defense complexes. In 2002, the DOE awarded Uranium Disposition Services, LLC (UDS) a contract to design, build and operate two DUF6 conversion facilities at Paducah, Kentucky and Portsmouth, Ohio. The facilities were designed to convert DUF6 into uranium oxide for disposal (or an alternate use) and aqueous hydrogen fluoride which is to be sold. The facilities were expected to be in full operation by June 2008.9 When the conversion facilities are operational, large quantities of high purity depleted uranium oxide powder will be available. This material could be used as a feedstock to manufacture aggregate suitable for inclusion into radiation shielding heavy concrete.

2.3

The case of DUAGG® and DUCRETE®

In the mid-1990s the Idaho National Laboratory (INL) developed new methods to produce high-density aggregate (synthetic rock) primarily consisting of depleted uranium oxide.10 The objective was to develop a low-cost method whereby depleted uranium oxide powder (UO2, U3O8, or UO3) could be processed to produce high-density aggregate pieces (DUAGG) having physical properties suitable for disposal in low-level radioactive disposal facilities or for use as a component of high-density concrete used as shielding for radioactive materials.11

2.3.1 Fabrication of DUAGG High purity, dense, sintered UO2 and U3O8 pellets were fractured to form aggregate. This aggregate was used with Portland cement and sand to fabricate a heavy concrete. In addition, INL conducted experiments where UO2 and U3O8 powder was sintered using liquid phase sintering techniques

48


to produce DUAGGTM. The DUAGG was also used as a heavy aggregate in concrete. Figure 2.1 shows a process flow diagram for the production of synthetic DUAGGTM heavy aggregate. The DUAGG consisted of approximately 80% uranium oxide powder pressed and then densified via liquid phase sintering where the liquid was provided by about 20% synthetic basalt. Physical properties of materials: UO3: crystal density – 7.3 g/cm3; bulk density – 3 g/cm3 UO2: crystal density – 10.9 g/cm3; bulk density – 4 g/cm3 Additives: average bulk density – 2.5 g/cm3 DUAGG: approx. green density – 5.4 g/cm3; approx. sintered density – 8.6 g/cm3 Crushed DUAGG: approx. bulk density – 5.2 g/cm3 DUAGG compositions The compositions of several synthetic basalts are listed in Table 2.3. Historically, Composition Number 5 of Table 2.3 has been the formula most frequently used in various waste encapsulation studies at INL. The composi-

UO3 + dirt (weigh)

Ducrete process Clay & additives (weigh)

Waste heat to dryer

Heat

Storage bin Combine?

1.) Rotary calciner gas fired

2.) Grind - attrition mill (wet)

3.) Mixer / Blender

Waste heat to dryer

Size: (comparable to jellybeans)

5.) Agglomerator

Exhaust

Exhaust

Briquettes

Storage bin

Pre heat

Storage bin

6.) Briquettor

8.) Mix concrete intensive shear? pump from bottom?

4.) Dryer

Storage bin

7.) Furnace - 1300 °C dry H2 atmosphere gas or electric? continuous or batch? 9.) Cast concrete forms, etc.

Cement + Sand + Water

2.1 Process flow diagram for production of DUAGGTM heavy aggregate (illustrative).


49

Table 2.3 Chemical analyses of five synthetic basalt melt compositions (wt %) Composition SiO2 No.

Al2O3 FeO

Fe2O3 CaO

MgO Na2O K2O

TiO2

ZrO2

1 2 3 4 5

8.34 8.09 7.87 6.93 5.20

6.62 5.30 7.09 2.76 0.79

3.85 3.76 3.31 3.38 2.77

2.15 2.00 1.34 1.74 1.46

0.59 2.94 6.62 11.4 21.4

0.14 1.24 3.06 4.96 9.91

48.7 46.8 45.9 41.6 32.6

17.1 17.3 15.5 17.0 16.4

9.21 9.74 7.08 7.61 6.70

3.29 2.79 2.20 2.64 2.86

Table 2.4 Average analysis of 224 North American basalts Melt

SiO2

Al2O3

FeO

Fe2O3

CaO

MgO

Na2O

K2O

TiO2

ZrO2

Avg.

49.9

16.0

8.4

2.7

9.7

6.5

2.7

0.8

1.6

1.5

tions can be compared to the average of 224 analyses of natural North American basalts, as shown in Table 2.4. Comparison of Table 2.3 and Table 2.4 shows that some of the INL synthetic basalts are enriched in iron oxides at the expense of the alumina and silica content. In addition, Compositions 4 and 5 have been highly enriched in TiO2 and ZrO2 to promote the formation of the crystalline phases (e.g. zirconolite (CaZrTi2O7), which can incorporate fairly large amounts of uranium oxide into its crystalline structure). Compositions 4 and 5 form liquid phases that can result in a higher degree of crystallinity upon cooling to room temperature than some of the other basalt compositions. Synthetic basalt is made using several precursor chemicals. For initial experiments, precursor chemicals were mixed in given ratios, melted, and cooled. The resultant mass was ground into a fine powder. Later experiments demonstrated lower cost methods to replace the mixing, melting, and grinding steps of producing synthetic basalt. It was discovered that uranium oxide powder could be mixed directly with calcined soil, additive chemicals, and clay and then the DUAGG could be green-pressed and sintered at quite moderate temperatures to good densities. The local INL soil used for the production process is designated Subsurface Disposal Area (SDA) Lake Bed Soil and has the following chemical analysis (in weight percentages): 69.9 SiO2, 13.2 Al2O3, 4.7 FeO, 4.1 CaO, 1.9 MgO, 1.5 Na2O, 3.3 K2O, 0.8 TiO2, 0.11 MnO2, 0.10 BaO, 0.06 ZrO2, 0.06 B2O3, 0.04 NiO, 0.02 SrO, and 0.02 Cr2O3. A typical batch formulation using soil and additives for melt # 5 is given in Table 2.5. Batch formulas can be adjusted for soil of different compositions. Soils available at the INL make excellent starting precursor material because they have a

50


Table 2.5 Batch formula of “Melt No. 5” basalt using “SDA” soil Chemical component

Number of moles

Batch wt%

SDA soil Al2O3 Fe2O3 CaO MgO NaHCO3 K2CO3·1.5 H2O TiO2 ZrO2

0.5917 0.0003 0.0541 0.0274 0.0192 0.0316 0.0022 0.1733 0.0570

51.36 0.88 11.84 2.11 1.06 3.64 0.50 18.98 9.63

Table 2.6 Favored mill forms for Melt No. 5 basalt using “SDA” soil and “Allen” clay Chemical component

Number of moles

Batch wt%

SDA soil Allen Kaolin clay Fe2O3 CaO MgO NaHCO3 K2CO3·1.5 H2O TiO2 ZrO2 Al2O3

0.4061 0.1889 0.0808 0.0588 0.0353 0.0181 0.0230 0.2440 0.0812 –

36.183 16.901 11.548 2.952 1.274 1.357 3.402 17.434 8.950 –

very fine particle size, are of volcanic origin (a composition similar to basalt), and contain very little organic matter. It was also discovered during DUAGG production that part of the soil can be replaced with clay. The benefits are that clay acts as (1) a lubricant that sticks less to metal pressing dies and (2) a powder binder. Clay greatly aids in the green forming of the aggregate when using mass production methods such as briquetting. A variety of different generic types of clays were investigated for use in the DUAGG production process, including Kaolin, Ball, and Bentonite clays. An example was H-56 clay which was supplied by K-T Clay Company. A similar batch formulation for Melt No. 5 basalt uses “Allen” type Kaolin clay from K-T Clay Company as shown in Table 2.6. The “Allen” clay exhibits a great deal of plasticity and was determined to be a satisfactory choice for inclusion in the mill batches. DUAGG that was pressed and sintered (reducing atmosphere), using the batch formula of Table 2.6, appeared to have good strength and showed


51

nondetectable amounts of uranium during leaching studies. However, during the leaching tests, trace amounts of iron were found to be leaching from the DUAGG. The leaching has no effect on the DUAGG shielding performance. However, if the DUAGG is buried in a landfill, such leaching may eventually lower DUAGG strength. Therefore, a new composition was devised in which extra CaO was substituted for the iron oxide, and boron oxide was added for increased neutron attenuation. This formulation, designated as SRP-5, is shown in Table 2.7. The molar formula is shown in Table 2.8. Sufficient moles of Ti were provided so that it would be theoretically possible to form a stoichiometric amount of CaZrTi2O7 (zirconolite) if the Ti atoms reacted with all of the Zr and a portion of the Ca with sufficient Ti remaining to react with the remaining Ca to form CaTiO3. Boron was added in an amount equivalent to about 1 wt%. Table 2.7 Boron-added DUAGGTM compositiona Sample no.

UO2 content vol. %

Type of clay additive

Description of mill additions

Anticipated bulk density (g/cm3)

Anticipated apparent density (g/cm3)

Anticipated open porosity %

SRP-5

80

Allen

Boron-no iron

>8.07

>8.46

24 months of exposure the release rate of uranium in a cement pore solution was low and no deleterious products from the alkaliaggregate reaction were seen. Evidence was later presented by ORNL24 that a protective layer of re-crystallization products from the basalt phase of the DUAGGTM covers the surface and this slows the release of uranium. These precipitated phases are due to the high silica content of the basalt, and are thought to include: Schoepite (UO3 . 2H2O), Soddyite (U5Si2O19 . 6H2O), and Uranophane (Ca[UO2]2[SiO3OH]2 . 5H2O).

2.4

Future trends

The largest applications for heavy concrete shielding are for spent nuclear fuel (SNF) and high level waste (HLW) containers. Most current shielding systems for SNF or vitrified HLW use either steel or concretec because of their relatively low cost, wide availability, known fabrication characteristics, and radiation-shielding effectiveness. For steel cask systems, separate neutron shields typically containing hydrogenous material are added since the thickness of steel required to attenuate neutrons to c

All concrete systems use some steel for structural and/or thermal purposes but the predominant material for shielding is concrete.


67

acceptable external doses is impractical. While ordinary concrete is not a particularly good gamma shield (per unit thickness), it is cheap enough that the practical approach is to use copious volumes. In sufficient thickness, it shields both the neutron and gamma radiation source terms effectively. On the straight basis of aggregate cost it is impossible to compare DUAGGTM with gravel, since gravel is inexpensive having been mined and washed rather than fabricated at a high temperature like DUAGGTM. However, there are overall economic factors that need to be considered. These include some environmental advantages to using the huge stockpiles of depleted uranium oxide (converted from corrosive UF6) for some useful purpose rather than burial.25 More importantly, it would seem that the key to gaining the full economic benefit and acceptance of DUCRETETM is to devise a combination storage and transportation cask which uses DUCRETETM as the shielding.26 An empty DUCRETETM cask can be transported via rail-freight from the manufacturing facility to a storage site and ultimately (after filling) to a repository or disposal location. This feature also provides an economic advantage as most concrete storage systems are not transportable. New concrete shielding must be built at each location where they are used. This transportable feature assures that if the United States opens an interim spent fuel storage facility at Yucca Mountain or at private sites, that the DUCRETETM shield can be transported to that site and re-used. There is a precedent for such a system in the GNB CONSTOR Cask27 (see Fig. 2.13) which uses high density concrete (4.1 g/cm3) combined with steel to provide a low cost system for transportation and storage of spent nuclear fuel.

2.4.1 Casks for spent nuclear fuel Shipping casks Transportability is a necessary design feature for the use of DUCRETETM casks at storage sites and then for their reuse as part of the repository waste package or at LLW disposal sites. The high density of DUCRETE allows the cask to be considerably smaller in diameter and remarkably lighter than conventional concrete casks. This is illustrated in Fig. 2.14. The DUCRETETM cask is 35 tons lighter and 100 cm smaller in diameter than the Sierra Nuclear Cask (SNC Model VSC) ventilated storage cask manufactured using ordinary concrete.28

2.4.2 Repository overpacks Figure 2.15 shows one particular concept developed by Duke Power for a DUCRETETM high level waste (HLW) storage container or “overpack”.

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Secondary lid (welded) Sealing plate (welded) Primary lid bolted Trunnion Fuel assembly Outer liner Inner liner Reinforcement Basket Heavy concrete

2.13 GNB CONSTOR combination storage and shipping cask (uses a heavy-concrete liner).

X-section of concrete cask (92 ft2)

Air outlet

SNF fuel assemblies

X-section of DUCRETE cask (44 ft2)

Cask lid

Multipurpose sealed basket (shell wall) Concrete cask liner Concrete

Air inlet duct Air entrance

2.14 Sierra nuclear ventilated spent-fuel dry storage cask made with concrete compared to one made with DUCRETETM.


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DUCRETETM Overpack System MPC ( 3 8 in.)

24 in.

13 in.

61 in. MPC ( 3 8 in.) canister

147 in.

120 in.

½ in. Steel shell

DUCRETE TM concrete (12 in.)

90 in.

DUCRETETM thickness (in.)

Surface dose rate (mrem/h)

Dose rate @ 2 (mrem/h)

10 11 12 13 14 15 16

41 27 19 15 12 9 8

10 6 4 3 2 2 1

2.15 Drawing of a high level waste (HLW) container made from DUCRETETM – also, shielding effectiveness vs. wall thickness.

This container holds four each of sealed HLW canisters. High level waste is much different than spent nuclear fuel and must be packaged differently. The surface dose rate (mrem/hr) drops off rapidly with DUCRETE wall thickness. Figure 2.16 shows a schematic of waste container stored in an underground repository. Provisions are made for access roads and gantry cranes for moving and positioning the waste containers.

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Gantry crane Concrete liner Access road

Rock

SG98 0051a

2.16 Schematic of waste containers stored in an underground repository.

2.4.3 Russian RBMK heavy concrete spent fuel casks Russia has approximately the same amount of stored depleted uranium as does the United States. Russia also has the need to transport and store spent nuclear fuel from various reactors. They have already adapted a metal/ heavy-concrete dry storage and transportation cask (GNB CONSTOR type – see Fig. 2.13) for spent fuel from their RBMK-type reactors. This cask contains heavy concrete utilizing steel shot and barium sulfate. Several years ago Oak Ridge National Laboratory (ORNL), Teton Technologies, General Nuclear Services, Inc. (GNSI) and VNIIEF of Russia joined together to collaborate on DUCRETETM cask design and fabrication.29 Both ORNL (Y-12) and VNIIEF were to produce 100–200 kg of DUAGGTM for testing. GNSI was to perform chemical testing and optimization of DUAGGTM and GNSI was to handle cask design and modeling tasks. Some additional work was to be performed by a Russian team30 on micro-reinforcement (metal and/or polymeric fibers) of concrete to increase fracture toughness and therefore the ability to absorb energy (from a rocket attack or airplane crash). The lead institution was the Russian Federal Nuclear Center – All Russia Science and Research Institute of Experimental Physics (RFNC-VNIIEF). The RFNC-VNIIEF is located in Sarov, Nizhniy Novgorod Oblast, Russia. This city was originally established in 1946 as Arzamas-16 as the location of the Soviet Union’s nuclear weapons research and development facilities. In 1994, Arzamas-16 was renamed Kremlev and in 1995 President Yeltsin officially changed the city’s name to Sarov (its original historical name). Sarov is sometimes called the “Russian Los Alamos”, but the RFNC-


71

VNIIEF also performs research in many other areas including materials science and mechanical engineering. Future production of DUAGGTM A report31 as of July 2005 indicated that RFNC-VNIIEF had made significant progress in improving and simplifying a DUAGGTM formulation and its production and had made DU concrete with high compressive strength and high density (6.4 g/cm3). Work was also ongoing on UO2 – stainless steel cermets with large (1 mm dia.) UO2 particulates mixed with stainless steel in about a 50 vol% ratio. For the last couple of years ORNL has not been funded for participation in the Russian collaboration. It is thought that the RFNC-VNIIEF is going forward with plans to produce casks. However, it is not clear if they will produce the UO2 – stainless steel cermet or large quantities of DUAGGTM to produce their version of DUCRETETM. Also, it is not clear what might be the source of funding for such developments.

2.4.4 Shields The use of heavy concrete (probably “toughened”) for shielding may have a future due to the proliferation of terrorist attacks on various structures. Toughened DUCRETETM A toughened version of DUCRETETM is thought to be necessary for applications where enhanced mechanical performance is desired like thin-walled (2 inch thick) waste containers, the corners/edges of large casks, for casks with thin (or no) metal walls, casks “hardened” against terrorist attacks, or for blast shields. Spent nuclear fuel sitting in certain dry-storage casks in the yards of nuclear power plants are thought to be vulnerable to terrorist attacks. The US Army Engineer Waterways Experiment Station (WES) supports the US Army Corps of Engineers mission in Survivability and Protective Structure using the Geodynamics Research Facility and the Projectile Penetration Research Facility. They have worked on high density and toughened concrete,32 sometimes in conjunction with Sandia National Labs on terminal effects of projectiles.33,34 Many additives and fabrication techniques are known to increase the compressive strength of concrete; however, for spent fuel and high level waste containers the critical design criteria are related to impact and dropping. In these cases, the relevant property is not compressive strength,

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but: (1) flexural strength (maximum stress at failure under bending), (2) impact resistance, (3) energy absorption, and (4) fracture toughness. These properties of concrete can be increased through the use of fibrous reinforcement. For shielding applications, steel fibers have the advantages (over plastic fibers) of holding stresses while under load and greatly increasing the tensile/bend (flexural) fracture strength of the concrete and moderately increasing the compression strength (likely due to shear effects). The level of property improvement gained in fibrous concrete is dependent on a number of factors including: (a) concrete mix, (b) fiber aspect ratio (length/diameter), (c) fiber strength, (d) fiber volume fraction, (e) concrete age, (f) fiber/matrix bond, (g) fiber modulus of elasticity (stiffness). Steel fibers include the categories of spun fibers and wires. The stainless steel versions of the wires are very corrosion resistant. Larger diameter steel fibers (wires) can be added to concrete with large aggregate (e.g., no larger than 3/8″ diameter) while smaller aggregate is necessary when using chopped fibers. The larger aggregate sizes will usually accommodate less fiber per cubic yard. Metal mats (loosely woven long continuous fibers) can only be filled with cement grout (small aggregate/sand or cement only) since small particles are necessary to penetrate between the continuous fibers in the mat. An example of the use of metal mats is the SIMCON (slurry infiltrated mat concrete) fabricated using MmatTEC promoted by Ribbon Technology Corp. In tests comparing the mat with normal fiber reinforcement, the mat materials offer a substantial improvement in flexural strength (increasing from 300 psi for ordinary concrete to 3000–6000 psi for SIMCON) and also large increases in fracture toughness (measured as area under load vs. deflection curve). Ribbon Technology Corp. has accumulated a large amount of fracture strength and toughening data using ordinary Portland cement mixes with various metal fibers and metal mats.35 The advantage of using metal mats is that a lower volume fraction of metal is needed when compared to chopped metal fibers in order to achieve the same level of toughening, plus the mat can be placed in specific volumes of a cast shape (e.g., edges or outside) in order to give the desired effect at a lower cost. Of course, polymeric fibers can also be added to concrete or DUCRETETM to provide toughening.36 All types of polymers have been utilized in concrete: polypropylene, polyethylene, polyester, acrylic, aramid, etc. Since the polymers have hydrogen atoms as part of their long chains, they also could help attenuate neutrons in DUCRETETM used as shielding. However, some concern has been expressed over polymeric fibers’ longterm stability in spent nuclear fuel casks due to temperature and radiation damage.


2.5

73

Sources of further information and advice

Below is a list of laboratories or companies that have been involved with research and development of DUAGGTM and DUCRETETM and companies that manufacture dry storage casks. Some cask manufacturing companies utilize concrete-liners and others do not. An example of a dry cask “system” manufactured by Holtec International is shown in Fig. 2.17. Figures 2.18 and 2.19 show an Energy Solutions Model VSC-24 (concrete lined cask) being loaded, in-service, and being shipped via rail inside a special Model TS125 transportation cask.

2.17 A “Holtec” dry storage case being set in place on a “pad”.

2.18 Energy Solutions VSC-24 cask being loaded (left) and in-service on pad at a nuclear plant (right).

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2.19 Energy Solutions VSC-24 cask will be shipped inside an Excellox6 transportation cask Model TS125 via rail car.

2.5.1 US national laboratories Idaho National Laboratory Paul A. Lessing, PhD Idaho National Laboratory Materials Department MS-2218 Box 1625 Idaho Falls, Idaho 83415-2218, USA Tel: 208-26-8776 E-mail: [email protected] Oak Ridge National Laboratory Leslie R. Dole, PhD Oak Ridge National Laboratory PO Box 2008 MS-6166 Oak Ridge, TN 37831-6166, USA Tel: 865-576-0382 E-mail: [email protected] Web page: http:/www.dole.nu.lesdole/

2.5.2 US technology development and worldwide cask fabrication companies The Idaho National Laboratory licensed their DUAGGTM and DUCRETETM technology to Teton Technologies who sublicensed to Starmet Corporation of Barnwell, SC. Starmet assembled and briefly ran a DUAGGTM fabrication pilot plant, but Starmet is no longer in business.


75

Dry cask technology advice Teton Technologies William Quapp, MS, PE 860 W. Riverview Dr. Idaho Falls, Idaho 83401, USA Tel: 208-535-9001 Email: [email protected] Pacific Development Services Inc. John C. Ritchie, BSc, MBA 1802 N Carson St., Suite 212 Carson City, NV, 89701, USA Tel: 702-940-7832 Email: [email protected] Dry cask fabrication companies There have been many changes in licenses and company ownership within the dry cask fabrication community within recent years. Below is a partial list of companies that currently manufacture dry spent fuel casks. AREVA (France) 2 Rue Paul Dautier, F-78141 Velizy-Villacoublay Cedex, France, Tel: +33 1 39 26 30 00 Fax: +33 1 39 26 27 00 Vectra Technologies sold the NUHOMS® license to AREVA, then Framatome, in the early 1990s. AREVA currently fabricates several different types of storage casks. Their Transnuclear TN series of metal casks have self-contained steel and borated resin shielding; no concrete is used. Energy Solutions, LLC [USA] 423 West 300 South, Suite 200 Salt Lake City, Utah 84101 Tel: 801-649-2000 E-mail: [email protected] Energy Solutions sells the VSC-24 cask system (see Fig. 2.18). Equipos Nucleares SA (ENSA) [Spain] C/ Jose Ortega y Gasset 20, 5 Floor, E-20006 Madrid, Spain, Tel: +34 91 5 553 617 Fax: +34 91 5 563 149 ENSA’s main focus is on spent fuel transport as opposed to dry cask storage; however, they do manufacture canisters for dry storage systems. Holtec International [USA] Holtec Centre, 555 Lincoln Drive West, Marlton, NJ 08053, USA, Tel: +1 856 797 0900

76


Holtec is a major manufacturer of casks at the present time. They have complete system solutions (wet storage, dry storage, and shipping). (See Fig. 2.17.) Siempelkampstraße 45, 47803 Krefeld, Germany, Tel: +49 2151 894 481 Fax: +49 2151 894 488 Holtec Manufacturing Division (HMD) Keystone Commons, 200 Braddock Avenue, Pittsburgh, PA 15145, USA, Tel: +1 412 823 3773 Fax: +1 412 823 6669 [See Above comments] GNS Gesellschaft für Nuklear-Service mbH [Germany] Hollestraße 7A, D-45127 Essen, Germany, Tel: +49 201 109 1828 Fax: +49 201 109 1125 GNS is the major European supplier for dry spent fuel cask storage, their technology is cast iron. NAC International [USA] Atlanta Corporate Headquarters 3930 East Jones Bridge Road, Norcross, GA 30092, USA, Tel: +1 678 382 1229 Fax: +1 678 382 1429 USEC purchased NAC International in late 2004.

2.5.3 Russian institutes Serge G. Ermichev (e-mail: [email protected]) V.I. Shapovalov N.V. Svirodov RFNC-VNIIEF 37 Mir Avenue, Sarov, 607190 Nizhniy Novgorod region Russia Tel: (83130) 4-24-14 or 4-57-92 V.K. Orlov VNIINM, Moscow, Russia

2.6

References

1 Waddell J.J. and Dobrowolski J.A., Concrete Construction Handbook, Third Edition, McGraw-Hill, 1993. 2 Kosmatka S.H. and Panarese W.C., Design and Control of Concrete Mixtures, Thirteenth Edition, Portland Cement Association, 5420 Old Orchard Road, Skokie, Ill 60077-1083.


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3 Haelsig R.T., “Depleted Uranium Concrete Container Feasibility Study”, EGG Report MS-11400, September 1994. 4 MPC transport 10 mrem/h at 2 m and 200 mrem/h on contact; MPC storage 20 mrem/h contact; DHLW storage 20 mrem/h on contact. 5 Haelsig R.T., “Depleted Uranium Concrete Container Feasibility Study”, Prepared under INEL contract to Scientech, Inc., 1585 N. Skyline Drive, Idaho Falls, ID 83402 by Packaging Technology, Inc., 4401 A Industry Drive East, Tacoma WA 98424-1800, May 16, 1994. 6 “Safety Analysis Report for the Ventilated Storage Cask System” PSN-91001 Rev.0 Pacific Sierra Nuclear Associates, Scotts Valley, CA, October 1991. 7 “BR-100 Shipping Cask Preliminary Design Report” 51-1177082-01, page II-311, Babcock & Wilcox Company, Lynchburg, VA. 8 ”Performance Testing and Analyses of the VSC-17 Ventilated Concrete Cask”, EPRI Report TR-100305, Project 3073-1, PNL-7831, UC-85, Final Report 1992, Prepared by Pacific Northwest Laboratories and Idaho National Engineering Laboratory (EG&G). 9 Chillicothe Gazette, Report, June 28, 2006. 10 Lessing P.A., “Development of DU-AGG (Depleted Uranium Aggregate)”, INEL-95/0315, Idaho National Laboratory, Idaho Falls, Idaho, September 1995. 11 Quapp W.J. and Lessing P.A., “Radiation Shielding Composition”, U.S. Patent No. 5,786,611; Granted July 28, 1998. 12 Lessing P.A., “Process Description for the Production of Depleted Uranium Aggregate”, INEL/INT-97-00662, Idaho National Engineering Laboratory, June 1997. 13 Lessing P.A. and Gillman H., “DU-AGG Pilot Plant Design Study”, INEL96/0266, Idaho National Engineering Laboratory, July 1996. 14 Quapp W.J., Miller W.H., Taylor J., Hundley C., and Levoy N., “DUCRETE: A Cost Effective Radiation Shielding Material”, Spectrum 2000, Sept. 24–18, Chattanooga, TN. 15 Lessing P.A., “Development of ‘DUCRETE’ ”, INEL-94/0029, Idaho National Engineering Laboratory, Box 1625, Idaho Falls, ID 83415, October 1994. 16 Lessing P.A., “Development of DU-AGG (Depleted Uranium Aggregate)”, INEL-95/0315, Appendix B, Idaho National Engineering Laboratory, Idaho Falls, Idaho, September 1995. 17 Quapp W.J. and Lessing P.A., “Radiation Shielding Composition”, U.S. Patent 6,166,390; Granted December 26, 2000. 18 “Heavyweight Concrete: Measuring, Mixing, Transporting, and Placing”, ACI 304.3R-89, American Concrete Institute, Detroit, MI. 19 Quapp W.J., “An Advanced Solution for the Storage, Transportation and Disposal of Spent Fuel and Vitrified High Level Waste”, Global ’99: International conference on future nuclear systems, summary article in Nuclear News; 42 (12), Nov. 1999. 20 Sterbentz J.W., “Shielding Evaluation of a Depleted Uranium Heavy Aggregate Concrete for Spent Fuel Storage Casks”, INEL-94/0092, Idaho National Engineering Laboratory, October 1994. 21 Tang J., “Shielding Characteristics of Various materials on PWR Waste Packages”, BBAC00000-001717-0210-00008 Rev 00, DOE Yucca Mountain Project Office, M&O Contractor, February 1998.

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22 “American National Standard Measurement of the Leachability of Solidified Low-Level Radioactive Wastes by a Short-Term Test Procedure”, ANSI/ANS16.1–1986 or “Static Leach Test Method” in the Nuclear Waste Materials Handbook of the Materials Characterization Center, MCC-1P, Pacific Northwest National Laboratory, Richland, WA. 23 Dole L.R., Ferrada J.J., and Mattus C.H., “Radiation Shielding for Storage and Transportation Cask Using Depleted Uranium Oxide in Cementitious Matrices”, U.S./Russian Depleted Uranium Workshop: Review of ISTC Projects, Oak Ridge National Laboratory, May 17–21, 2004. 24 Dole L.R., Ferrada J.J., and Mattus C.H., “Cask Size and Weight Reductions Through the Use of DUCRETE”, Russian-American Workshop on Use of Depleted Uranium and Review of International Science and Technology Center (ISTC) Projects, Moscow/Serov, Russia, June 19–23, 2005. 25 Quapp W.J. and Cooley C.R., “New Options for Managing Depleted Uranium Materials”, International Conference on Nuclear and Hazardous Waste Management (SPECTRUM ’98), Denver, CO, Sept. 13–17, 1998, American Nuclear Society – ISBN: 0-89448-635-7. 26 Hopf J.E., “Conceptual Design Report for a Transportable DUCRETE Spent Fuel Storage Cask System”, INEL-95/0167, Idaho National Engineering Laboratory, April 1995. 27 Diersch R. and Jack A., “GNB, The CONSTOR Steel-Concrete Cask for Transport and Storage of Spent Nuclear Fuel from RBMK”, INMM, Washington, DC, January 1998. 28 Powell F.P., “Comparative Economics for DUCRETE Spent Fuel Storage Cask Handling, Transportation, and Capital Requirements” INEL-95/0166, Idaho National Engineering Laboratory, April 1995. 29 Haire M.J., Doe L.R., Arrowsmith H.W., Denton M.S., Shapovalov Vi.I., and Matveev V.Z., “A Collaboration to Develop the Next-Generation SNF/HLW Cask”, 2003 International High-Level Radioactive Waste Management Conference, March 30–April 2, 2003, Las Vegas, Nevada. 30 Dole L.R., Ferrada J.J., and Mattus C.H., “Radiation Shielding for Storage and Transportation Cask Using Depleted Uranium Oxide in Cementitious Matrices”, J.S./Russian Depleted Uranium Workshop: Review of ISTC Projects, May 17–21, 2004, Oak Ridge National Laboratory, Oak Ridge, TN. 31 “ORNL Foreign Trip Report #226719, Report of Foreign Travel to Russia for M.J. Haire, C.W. Forsberg, L.R. Dole, and R.G. Wymer”, Oak Ridge National Laboratory, July 13, 2005. 32 Moxley R.E., Adley M.D., and Rohani R., “Impact of Thin-Walled Projectiles with Concrete Targets”, Shock and Vibration 2, 1995 5 355–364. 33 Forrestal M.J., et al., “An Empirical-Equation for Penetration Depth of OgiveNose Projectiles into Concrete Targets”, International Journal of Impact Engineering, 1994 15 (4) 395–405. 34 Frew D.J., et al., “The Effect of Concrete Target Diameter on Projectile Deceleration and Penetration Depth”, International Journal of Impact Engineering, 2006 32 (10) 1584–1594. 35 Hackman L.E., et al., “Slurry Infiltrate Mat Concrete (SIMCON)”, Concrete International, 1992 (12) 53–56. 36 Fiber Reinforced Concrete, Portland Cement Association, 5420 Old Orchard Road, Skokie, Illinois 60077-1083, ISBN 0-89312-091-X, 1991.

3 High-strength concrete O E GJØRV, Norwegian University of Science and Technology, Norway

3.1

Introduction

The theoretical basis for producing high-strength concrete was originally developed in the field of ceramic materials in the late 1950s and early 1960s. Based on single-phase polycrystalline ceramic materials, it was shown that reduced particle dimension increased the strength. The dependence of the particle size on the strength was explained on the basis of Griffith’s theory for the rupture of brittle materials with internal cracks. In accordance with this theory, the strength of the material should increase with decreasing pore and particle size by a square root law. Later on it was shown that there was a similar relationship between microstructure and strength for cement pastes with densely packed cement particles at a very low water/cement ratio. However, it was not until the early 1970s that new and very effective agents for dispersing the fine cement particles in water became available, and then, a tremendous advance in the production of high-quality concrete was achieved. At the same time, large quantities of ultra-fine condensed silica fume particles also became available. Therefore, a commercial basis for production of concrete with very high density and strength was established, and a rapid development of high-strength concrete took place. Since a low porosity concrete with high density also will enhance the overall performance of the material, the term “high-performance concrete” was also soon introduced, which is inclusive of the term “high-strength concrete”. More and more, however, the term “high-performance concrete” was mostly used and specified for concrete durability rather than for concrete strength. In the literature, there are a number of definitions of both “high-strength concrete” (HSC) and “high-performance concrete” (HPC), but as properly discussed by Aïtcin in his book on high-performance concrete (Aïtcin 1998), there is no clear consensus about the meaning of either of these terms. In the literature, some people try to define high-strength concrete as different from “normal strength concrete”, “ordinary concrete” or “usual 79

80


concrete”, but what is “normal”, “ordinary” or “usual” is rapidly changing. Also, some definitions are based on a maximum water/cement ratio, but nor is the term “water/cement ratio” easy to define any longer. For many years, when concrete was mostly based on pure Portland cements and simple procedures for concrete production, the concept of water/cement ratio was the fundamental basis for characterizing concrete quality. Since a number of different cementitious materials and reactive fillers are now being used for concrete production, the concrete properties are more and more being controlled by various combinations of such materials. In addition, the concrete properties are also increasingly being controlled by the use of various types of processed concrete aggregate, new concrete admixtures and sophisticated production equipment. Therefore, the old and very simple terms “water/cement” or “water/binder ratio” for characterizing concrete quality successively lost their meaning. As a consequence, performance definition and specification for concrete quality based on performance criteria are more commonly being applied. In order to stimulate the use of high-performance concrete for highway applications in the USA, the Federal Highway Administration in the early 1990s defined high-performance concrete (HPC) by four durability and four strength parameters, which included (Goodspeed et al. 1996): Durability • freeze/thaw durability • scaling resistance • abrasion resistance • chloride permeability. Strength • compressive strength • elasticity • shrinkage • creep. Based on requirements for each parameter, four different performance grades were defined, and details of test methods for determining the performance grades given. Then, applications of the various HPC grades for various exposure conditions were recommended.

3.2

Applications

3.2.1 General At an early stage, high-strength or high-performance concrete was mostly applied to high-rise buildings, bridges and offshore structures, but it was successively applied to a variety of other applications such as:

High-strength concrete • • • • • • •

81

harbor and coastal structures hydraulic structures underground construction pavements and industrial floors water treatment plants storage facilities for aggressive waste and chemicals concrete products.

For some of the applications, the mechanical properties were decisive, while for others, the durability properties were the most important; but very often, it was the combined enhanced mechanical and durability properties which were the basis for the various applications. For the various partners in new projects, the importance and benefits of applying high-strength or high-performance concrete were somewhat different. For the owner, increased service life, reduced concrete volume and costs, reduced construction time and increased space and comfort in highrise buildings by reduced swaying were the most important. For the designer, improved properties such as compressive strength, E-modulus, durability and rapid attainment of final creep as well as reduced dead load were important. For the contractor, fast track construction and cheaper alternatives were important. For the concrete producer, the use of high-tech production increased the profit and the market, as well as giving beneficial effects on conventional production. Also, from an environmental point of view, savings in cement and aggregate and increased service life contributed to a more sustainable development. Over the years, a number of technical committees and working groups have summarized the extensive literature and experience from the various areas of application of high-strength or high-performance concrete, a brief outline of which is given in the following.

3.2.2 High-rise buildings For many years, the columns of high-rise buildings were the largest application of high-strength concrete in buildings. In 1965 and in the early 1970s, a number of buildings with high-strength concrete were constructed in the Chicago area with columns typically having a design compressive strength of 62 MPa, but successively, much higher design strengths were achieved (Table 3.1 and Fig. 3.1). Very often, the high E-modulus for improved stiffness of the columns was the motivation for the selection of high-strength concrete.

3.2.3 Bridges For long-span bridges, both high strength and low weight are important. For the bridges listed in Table 3.2, design strengths of up to 79 MPa based on

82


Table 3.1 Buildings with high-strength concrete (FIP/CEB 1990) Building

Location

Year+

Total stories

Lake Point Tower Midcontinental Plaza Frontier Towers Water Tower Plaza Royal Bank Plaza River Plaza Helmsley Palace Hotel Richmond – Adelaide Toronto Larimer Place Condominiums Texas Commerce Tower City Center Project Trump Tower 499 Park Avenue Petrocanada Building S.E. Financial Center Chicago Mercantile Exchange 1130 A.Michigan Ave. Pacific Park Plaza

Chicago Chicago Chicago Chicago Toronto Chicago New York Centre

1965 1972 1973 1975 1975 1976 1978 1978

70 50 55 79 43 56 53 33

52 62 62 62 61 62++ 55 61

Denver

1980

31

55

Houston Minneapolis New York New York Calgory Miami Chicago

1981 1981 – – 1982 1982 1982

75 52 68 27 34 53 40

52 55 55 59 50 48 62+++

Chicago Emeryville, CA Melbourne Seattle Dallas Chicago Paris

– 1983

– 30

52 45

– 1983 1983 1986 1988

44 76 72 15 –

55 66 69 97 65

Chicago Seattle Seattle Seattle

1989 1989 1989 1989

79 58 44 62

83 115 115 94

Collins Place Columbia Center Interfirst Plaza 900 N. Mich. Annex Grande Arche de la Défence South Wacker Tower Two Union Square Pacific First Center Gateway Tower

Max design strength (MPa)

+ Year in which high strength concrete was cast. ++ Two experimental columns of 76 MPa strength were included. +++ Two experimental columns of 97 MPa strength were included.

normal weight and up to 69 MPa based on lightweight concrete were applied. For cantilever bridges, the use of normal weight or lightweight concrete was sometimes varied from one span to another. In Norway, two floating bridges for strait crossings were also built, where a 65 MPa type of lightweight concrete was typically used for the floating pontoons (Meaas et al. 1994; Hasselø 2001). Many bridges are also exposed to very severe environments (Fig. 3.2), and in particular, chloride penetration and steel corrosion have proved to

High-strength concrete Building height, ft 12,000 12,000 psi 1000 (82.7 MPa) (82.7) 12,000 (82.7)

900 800

Building height, m 300 12,000 (82.7)

19,000 (131.0)

17,000 (117.2) 19,000 (131.0)

12,500 (82.2)

700

83

14,000 (96.5)

600 500

14,000 (96.5)**

400

250 200 150 100

300 200

50

100

So ut h

Pr ud en t

31 1

Tw o

0

ia l C Pla hi z c a 1 ag * W 98 o ac 9 ke r C Dr hi iv c e So 1 ag * ci 98 o et 9 y C Ce O le n ne ve te Pe la r† ac 19 nd ht 91 re e Ce At nte la r* n Tr 199 ta u N mp 1 ew Yo Pal Da rk ac in C e* Bo 19 ity sw 91 or M th 22 in T 5 ne ow W ap er es t W 19 olis * ac 91 ke r C Dr hi iv c e G 1 ag * at 9 o ew 88 ay To Se we a r† Pa 19 ttle ci 90 fic Fi rs tC e Se ntr a e† Tw 1 ttle o U 989 ni on Sq Se uar a e† 19 ttle 88

0

* Reinforced concrete frame † Composite concrete/steel frame ** Also includes one experimental column of 17,000 psi

3.1 High-strength concrete in buildings (FIP/CEB 1990).

represent a special challenge for their durability and long-term performance (Gjørv 2002). For many bridges, experience has shown that specification of a concrete with high strength is not necessarily enough to ensure proper durability, and in some cases, deep chloride penetration has been observed even during concrete construction, before the concrete has gained sufficient maturity and density (Fig. 3.3).

3.2.4 Offshore structures When the first concept for fixed offshore concrete structures in the North Sea was introduced in the late 1960s, the offshore technical community showed much skepticism. At the same time, however, the results of a comprehensive field investigation of more than 200 conventional concrete sea structures along the Norwegian coastline were published, demonstrating that the general condition of these structures was quite good, even after service periods of up to 50–60 years (Gjørv 1968, 1994). These results contributed, therefore, to convincing the most skeptical operators that concrete could also be a possible and reliable construction material for offshore installations in the North Sea. However, the appearance of corrosion on embedded steel that typically took place in all of these conventional concrete structures after a service period of only 5–10 years was not acceptable

Table 3.2 Bridges with high-strength concrete (FIP/CEB 1990) Bridge

Location

Year

Max span Max design (m) strength (MPa)

Willows Bridge Nitta Highway Bridge San Diego to Coronado Kaminoshima Highway Bridge Ootanabe Railway Bridge Fukamitsu Highway Bridge Akkagawa Railway Bridge Kylesku Bridge Selbjørn Bridge Deutzer Bridge Pasco-Kennewick Intercity Coweman River Bridge Linn Cove Viaduct N Parrot Ferry Bridge Ottmarsheim Pont de Tricastin Tower Road Bridge Pont du Pertuiset Pont de Joigny Arc sur la Rance Giske Sandhornøya Boknasundet Helgelandsbrua

Toronto Japan California Japan

1967 1968 1969 1970

48 30 43 86

41 59 41+ 59

Japan Japan Japan Scotland Norway Germany Washington Washington Carolina California France France Washington France France France Norway Norway Norway Norway

1973 1974 1976 – 1977 1978 1978 – 1979 1979 1979 – 1981 1988 1988 1988 1989 1989 1990 1990

24 26 46 79 212 185 299 45 54 195 172 142 49 110 – – 52 154 190 425

79 69 79 53 40 69+ 41 48 41 43+ 30+ 30+ 62 65 60 60 55 55+ 60+ 65

+ Lightweight concrete.

3.2 Storseisund Bridge (1989), which is one of the many concrete bridges severely exposed along the Norwegian coastline (courtesy of J Brun).

High-strength concrete

85

3.3 Helgelandsbrua (1990) in Norway, in which a high chloride penetration already took place during concrete construction.

3.4 Development of offshore concrete structures in the North Sea (courtesy of Aker-Kvæner).

to the offshore technical community. Therefore, in order to gain acceptance for the first offshore concrete platform, both increased concrete quality and concrete cover beyond that required by current concrete codes and much stricter programs for quality assurance and quality control had to be introduced. During the construction of the Ekofisk Tank (1973), the first edition of Recommendations for Design and Construction of Concrete Sea Structures was published by the international organization for prestressed concrete (FIP 1973). Shortly after, Det Norske Veritas (DNV 1974) in their Rules, the Norwegian Petroleum Directorate in their Regulations (OD 1976) and ACI in their Recommendations (ACI 1978) adopted the new and stricter durability requirements for fixed offshore concrete structures. After the first breakthrough for use of concrete for offshore installations in the North Sea in the early 1970s, rapid development took place (Fig. 3.4).

86


During the period from 1973 to 1995, altogether 28 major concrete platforms containing more than 2.5 million cubic meters of high-performance concrete were installed, and by 2007, there were 34 concrete structures in the North Sea, most of which were produced in Norway (Table 3.3). In other parts of the world, a number of offshore concrete structures have also been produced, and so far, a total of 50 various types of offshore concrete structures have been installed (Moksnes 2007). For the first North Sea concrete structures in the early 1970s, it was not so easy to produce a high-strength concrete which also contained entrained air for ensuring proper frost resistance. Extensive research programs were carried out, however, and the quality of concrete and the specified design strength increased from project to project. Thus, from the Ekofisk Tank which was installed in 1973, to the Troll A Platform installed in 1995, the design strength successively increased from 45 to 80 MPa. Also, the water depths for the various installations successively increased; in 1995, the Troll A Platform was installed at a water depth of 303 m. From the tip of the skirts to the top of the shafts of this gravity base structure, the total height was 472 m, which is taller than the Empire State Building. After production in one of the deep Norwegian fjords, the Troll A Platform containing 245,000 m3 of high-strength concrete, 100,000 t of reinforcing steel and 11,000 t of prestressing steel, was moved out to its final offshore destination, and this operation was the biggest movement of a man-made structure ever (Fig. 3.5). In 1995, the Heidrun platform was also installed in deep water of 350 m, but this structure was a tension leg floating platform consisting of lightweight concrete with a design strength of 65 MPa.

3.5 The Troll A Platform (1995) on its way out to the final destination in the North Sea (courtesy of Aker-Kværner).


87

Table 3.3 Offshore concrete structures with high-strength concrete in the North Sea (Moksnes 2007) Year

Field

Operator

Platform type

1973

Ekofisk

Phillips

1975 1975 1975

Beryl A Brent B Frigg CDP1

Mobil Shell Elf

1976 1976 1976

Brent D Frigg TP1 Frigg MCP-01

Shell Elf Elf

1977 1977 1977 1978 1978

Dunlin A Frigg TCP2 Statfjord A Cormorant A Ninian Central

Shell Elf Mobil Shell Chevron

1978 1981 1982

Brent C Statfjord B Maureen ALC

Shell Mobil Phillips

1981

Schwedeneck A

Texaco

1981

Schwedeneck B

Texaco

1984 1986 1987 1988

Statfjord C Gullfaks A Gullfaks B Oseberg A

1989 1989

Gullfaks C North Ravenspurn

1989 1992 1992

Ekofisk Barriere NAM F3 Snorre CFT

Mobil Statoil Statoil Norsk Hydro Statoil Hamilton Bros. Phillips Shell Saga

Caison, Jarlan Wall GBS 3 shafts GBS 3 shafts GBS 1 shaft, Jarlan Wall GBS 3 shafts GBS 2 shafts GBS 1 shaft, Jarlan Wall GBS 4 shafts GBS 3 shafts GBS 3 shafts GBS 4 shafts GBS 1 shaft, Jarlan wall GBS 4 shafts GBS 4 shafts GBS Art. column GBS Monotower GBS Monotower GBS 4 shafts GBS 4 shafts GBS 3 shafts GBS 4 shafts

1993 1993

Sleipner A Draugen

Statoil Shell

1994

Heidrun Found

Conoco

1995 1995 1995 1995

Harding Troll A Heidrun TLP Troll B

1999

South Arne

BP Shell Conoco Norsk Hydro Amerada Hess

Depth (m)

Concrete volume (m3)

70

80,000

118 140 98

52,000 64,000 60,000

140 104 94

68,000 49,999 60,000

153 104 145 149 136

90,000 50,000 87,000 120,000 140,000

141 145 92

105,000 140,000 –

25

3,620

16

3,060

145 135 141 109

130,000 125,000 101,000 116,000

GBS 4 shafts GBS 3 shafts

216 42

244,000 9,800

Protection Ring GBS Suction anchors, 3 cells GBS 4 shafts GBS Monotower Suction anchors, 19 cells GBS foundation GBS 4 shafts Concrete TLP Concrete Semi

75 43 310

105,000 23,300 7,800

82 251

77,000 85,000

350

28,000

106 303 350 340

37,000 245,000 63,000 41,000

60

35,000

GBS 1 shaft

88


3.2.5 Special applications Along with the extensive programs for production of offshore concrete structures, extensive research for further development of high-strength and high-performance concrete took place. Thus, in the early 1990s, concretes based on high-quality natural rock aggregate and lightweight aggregate with compressive strengths of up to 198.6 MPa (Gjørv and Rønning 1992) and 102.4 MPa (Zhang and Gjørv 1991), respectively, were obtained, the latter having a fresh concrete density of 1865 kg/m3. As a spin-off of this research, a number of other special applications such as high-strength concretes for highway pavements, industrial floors and hydraulic structures were also developed. In the Scandinavian countries, extensive maintenance and rehabilitation of highway pavements due to heavy traffic from studded tires had been a big problem for many years. In 1985, therefore, an accelerated load facility for full-scale testing of the abrasion resistance of highway concrete pavements exposed to heavy traffic from studded tires was built in Norway (Gjørv et al. 1990). By increasing the concrete strength from 50 to 100 MPa, the abrasion of the concrete was reduced by approximately 50%, and at 150 MPa, the abrasion was comparable to that of high-quality massive granite blocks (Fig. 3.6). Compared to a high-quality asphalt highway pavement, this represented an increased service life of the highway pavement by a factor of approximately ten. Later, various types of concrete with high abrasion resistance for a variety of other applications were developed.

Partial abrasion (mm/104 rev.)

3.0

Wet

Dry Wet Type of aggregate

Syeniteporphyr Hornfels Quartzdiorite Jasper

2.5 2.0 Dry 1.5 1.0 Massive granite (wet) 0.5

Massive granite (dry) 50

150 100 28 days compressive strength (MPa)

3.6 Relationship between compressive strength and abrasion resistance of concrete (Gjørv et al. 1990).


89

Compressive strength (MPa)

500

100

50

10 0.3

0.4

0.5

0.6

0.7

Porosity

3.7 Relationship between compressive strength and porosity of cement paste (Bache 1981).

Gradually, a new generation of cementitious materials with extremely low porosities and ultra high-strength properties were also developed (Fig. 3.7). Different techniques such as DSP, MDF and RPC were applied in order to reduce the porosity, but all these materials had one thing in common by optimizing the packing of all particles involved. As a result, water/binder ratios of 0.10 to 0.20 were achieved, which were much lower than those of traditional high-performance concrete. The DSP materials (densified with small particles) took advantage of the combined action of condensed silica fume and superplastcizers. By also replacing the rockbased aggregate by small particles of high-quality ceramic aggregate such as calcined bauxite, new materials with compressive strengths of 140 to 500 MPa, flexural strengths of 20 to 70 MPa and E-moduli of 50 to 100 GPa became available for a variety of new industrial applications and new concrete products. The MDF materials (macro defect free) also took advantage by additional use of a polyvinyl alcohol polymer (PVA), which first acts as a powerful dispersant of the particle system and then acts as reactive binder. As a result, impressive flexural strengths of up to 200 MPa were achieved. The RPC materials (reactive powdered concrete), which were the latest type of ultra high-strength materials developed, went one step further than the DSP materials in down-scaling the maximum size of the coarse aggregate to 300 μm. By such a small size aggregate in combination with pressure on the material during hardening, impressive compressive strengths of up to 800 MPa were achieved.

90


3.3

Future trends

3.3.1 General In many countries there has been a rapidly increasing deterioration of many important concrete infrastructures (Gjørv 2002). This is not only a technical and economic problem, but also has a great impact on available resources, environment and human safety (Gjørv and Sakai 2000). For new concrete infrastructures, therefore, there is a great need for proper application of high-performance concrete. In many countries, there is also a rapid development of high-rise buildings, and increasing population and shortage of land area also tend to move more activities and constantly new types of activities into more severe marine environments. All of these developments will require more high-performance concrete in the years to come. For many of the bridges and offshore installations which have been exposed to a severe environment for some time (Figs 3.2 and 3.8), experience has shown that the use of high-strength concrete alone does not necessarily ensure proper durability and long-term performance. Although the overall condition of the concrete structures in the North Sea appears to be quite good (Gjørv 1994, Moksnes and Sandvik 1996, FIP 1996), several of these structures already have some extent of steel corrosion, and for some of them, very costly repairs have been carried out and costly protective measures applied. Many of the observed corrosion problems can be related to lack of proper quality control or special problems during concrete construction, but experience also clearly demonstrates that a high compressive

3.8 The offshore concrete structures in the North Sea are exposed to a very harsh and hostile environment.


91

Chloride concentration (% Cl as weight of concrete)

0.35

elevation: + 14.4 m elevation: + 7.8 m elevation: – 11.5 m

0.3 0.25 0.2 0.15 0.1 0.05 0 0

10

20

30

40

50

60

70

Distance from the exposed surface (mm)

3.9 Chloride penetration into a 20-year-old North Sea concrete platform (Sengul and Gjørv 2007).

strength alone does not necessarily give a high resistance against chloride penetration. For one of the concrete platforms in the North Sea which was subjected to a very thorough investigation after a service period of approximately 20 years, Fig. 3.9 reveals that a deep chloride penetration both above and below water had taken place (Sengul and Gjørv 2007). Although the scatter of test results was relatively high, the chloride penetration was deepest in the upper part of the splash zone (+14.4 m) and lowest in the constantly submerged part of the shaft (−11.5 m). In the upper part of the shaft, a chloride front of approximately 0.07% by weight of concrete at a depth of approximately 60 mm was observed. For the nominal concrete cover of 75 mm specified, this indicated that an early stage of steel corrosion had already been reached. A specified water/cement ratio of less than 0.40 and a minimum cement content of 400 kg/m3 in combination with a nominal concrete cover of 75 mm, had not been sufficient to prevent chloride penetration from reaching embedded steel within a service period of approximately 20 years. While it is relatively easy to control the resistance of concrete, both against freezing and thawing and expansive alkali reactions, by following established precautions and procedures, extensive experience demonstrates that electrochemical corrosion of embedded steel represents the most critical and greatest threat to the durability and long-term performance of concrete structures in chloride exposed environments. Therefore, in order to gain a more controlled durability and service life of new important concrete structures in severe environments, a rapid international development on both probability-based durability design and performancebased concrete quality control has taken place (Gjørv 1993, 2002). A

92


probability-based durability design has already been applied to several new major projects (Gehlen and Schiessl 1999, McGee 1999). In order to meet the very strict requirements based on such a durability design, a proper type of cement or binder system has proved to be very important (Årskog et al. 2007). Thus as a result of a proper durability design, the Rion-Antirion Bridge (2001) in Greece was produced with a high-performance concrete based on blast furnace slag cement (Fig. 3.10). A probability-based durability design has already been shown to provide a very good basis for a better utilization of high-performance concrete for new concrete structures in severe environments.

3.3.2 Probability-based durability design A probabilistic approach to the durability design was developed in the European research project “DuraCrete” in the late 1990s (DuraCrete 1999). Further development and simplification of the results from this project have provided the basis for new specifications and guidelines both for durability design and performance-based concrete quality control of new important concrete structures in Norwegian harbors (Norwegian Association for Harbor Engineers 2007a, b). Over the last few years, the above Norwegian specifications and guidelines have been successfully applied to a number of projects with important concrete structures in severe environments. One of these projects includes a new large city development with a service life requirement of 300 years (Fig. 3.11), which is currently under construction in Oslo harbor (Årskog and Gjørv 2007). Although a probability-based durability design does not guarantee a given service life of a new concrete structure, such a design provides the basis for engineering judgment of all of the factors which are considered the most relevant for the durability, including the scatter and variability of all these factors. Hence, a good engineering basis for comparing and select-

3.10 The Rion-Antirion Bridge (2001) in Greece was produced with a high-performance concrete based on blast furnace slag cement.


93

3.11 Utilization of high-performance concrete in a new city development project in Oslo harbor, where all the concrete substructures have a service life requirement of 300 years (courtesy of Tjuvholmen KS).

ing one of several technical solutions for a given environment is obtained, and durability requirements, which are possible to verify and control during concrete construction, are specified. Extensive experience has shown that much of the durability problems, which occur after some time of service, can be related to lack of proper quality control and special problems during concrete construction. Upon completion of a new structure, therefore, it is very important to provide documentation of the construction quality and durability achieved before the structure is handed over to the owner. As part of the durability design, the owner must also be provided with a service manual for a regular condition assessment and preventive maintenance of the structure. For concrete structures in chloride-containing environments, it is the regular monitoring of the real chloride penetration and assessment of the future corrosion probability in combination with protective measures which provide the ultimate basis for achieving a more controlled durability and service life.

3.4


International conferences “Very High Strength Cement-Based Materials” (1984), Boston, USA, Proc. of an Int. Symp., ed. by J F Young, Matrs. Res. Soc., Pennsylvania, ISBN 0-931837-07-3. “Utilization of High Strength Concrete” (1987), Stavanger, Norway, Proc. of 1st Int. Symp., ed. by I Holand, D Holland, B Jakobsen and R Lenchow, Tapir, NTNU, Trondheim, ISBN 82-519-0797-7.

94


“High Strength Concrete” (1990), Proc. of 2nd Int. Symp., Berkeley, USA, ed. by W T Hester, ACI SP-121. “High Strength Concrete” (1993), Lillehammer, Norway, Proc. of 3rd Int. Symp., ed. by I Holand and E Sellevold, Norwegian Concrete Association, Oslo, ISBN 82-91341-00-1. “Utilization of High Strength/High Performance Concrete” (1996), Paris, France, Proc. of 4th Int. Symp., Paris, ed. by F de Larrard and R Lacroix, École Nationale des Ponts et Chaussées, Paris, ISBN 2-85978-257-(1-3-5). “Utilization of High Strength/High Performance Concrete” (1999), Sandefjord, Norway, Proc. of 5th Int. Symp., ed. by I Holand and E. Sellevold, Norwegian Concrete Association, Oslo, ISBN 82-91341-25-7. “Utilization of High Strength/High Performance Concrete” (2002), Leipzig, Germany, Proc. of 6th Int. Symp., ed. by G Kønig, F Dehn and T Faust, Leipzig University, Leipzig, ISBN 3-934178-18-9. “Utilization of High Strength/High Performance Concrete” (2005), Washington, USA, Proc. of 7th Int. Symp., ed. by H G Russel, ACI SP-228. “Concrete Under Severe Conditions – Environment and Loading” (1995), Sapporo, Japan, Proc. of 1st Int. Symp., ed. by K Sakai, N Banthia and O E Gjørv, E & FN Spon, London and New York, ISBN 0-419-19870-9. “Concrete Under Severe Conditions – Environment and Loading” (1998), Tromsø, Norway, Proc. of 2nd Int. Symp., ed. by O E Gjørv, K Sakai and N Banthia, E & FN Spon, London and New York, ISBN 0-419-23850-6. “Concrete Under Severe Conditions – Environment and Loading” (2001), Vancouver, Canada, Proc. of 3rd Int. Symp., ed. by N Banthia, K Sakai and O E Gjørv, Univ. of British Columbia, Vancouver, ISBN 0-88865-782. “Concrete Under Severe Conditions – Environment and Loading” (2004), Seoul, Korea, Proc. of 4th Int. Symp., ed. by B H Oh, K Sakai, O E Gjørv and N Banthia, Seoul National University and Korea Concrete Institute, Seoul, ISBN 89-8949902-X 93530. “Concrete Under Severe Conditions – Environment and Loading” (2007), Tours, France, Proc. of 5th Int. Symp., ed. by F Toutlemonde, K Sakai, O E Gjørv and N Banthia, Laboratoire Central des Ponts et Chaussées, Paris, ISSN 2-7208-2495-X.

State-Of-The-Art Reports “State-Of-The-Art-Report on High Strength Concrete” (1984), Report by ACI Committee 363–84, ACI Journal. “Guide for the Design and Construction of Fixed Offshore Concrete Structures” (1984), Report by ACI Committee 357R-84, ACI. “High Strength Concrete” (1987), ACI Report SCM-15, ACI. “Literature Review of High Strength Concrete Properties” (1988), by L J Parrott. Review carried out by C and CA Services, January. “High Strength Concrete” (1990), State-Of-The-Art-Report, FIP SR 90/1/CEB Bulletin d’Information No. 197, London. “High Performance Concrete” (1991), by P Zia, M L Leming and S M Ahmed, A State-Of-The-Art-Report by the Strategic Highway Research Programme, National Research Council, Washington, DC, USA.


95

“State-Of-The-Art-Report on Offshore Concrete Structures for the Arctic” (1991), Report by ACI Committee 357, 1R-91, ACI. “High Strength Concrete” (1992), ACI Compilation 17, ACI. “State-Of-The-Art-Report on High Strength Concrete” (1992), Report by ACI Committee 363R-92, ACI. “Application of High Performance Concrete” (1994), FIP/CEB Bulletin d’Information No. 222, Lausanne, ISBN 2-88394-025-8. “High-Strength Concrete” (1994), by J A Farny and C Panarese, Portland Cement Association, Engineering Bulletin, Skokie, USA. “High Performance Concrete” (1995), Recommended Extensions to the Model Code 90. Research Needs, CEB Bulletin d’Information No. 228, Lausanne, ISBN 2-88394-031-2. “Durability of Concrete Structures in the North Sea” (1996), State-Of-The-ArtReport, FIP, London, ISBN 1-874266-30-1. “Curing of High-Performance Concrete: Report of the State-of-the-Art” (1999), by K W Meeks and N J Carino, NISTR 6295, Gaithersburg, USA.

Books “High Strength Concrete” (1985), ed. by H Russel, ACI SP-87. “Production of High Strength Concrete” (1986), by M B Peterman and R L Carrasquillo, Noyes Publications, ISBN 0-8155-1057-8. “High Performance Concrete: From Material to Structure” (1992), ed. by Y Malier, E & FN Spon, London, ISBN 0-419-17600-4. “High Performance Concrete in Severe Environments” (1993), ed. by P Zia, ACI SP-140. “High Performance Concrete: Properties and Applications” (1994) ed. by S P Shah and S H Ahmad, Edward Arnold, London, ISBN 0-340-059922-1. “High Performance Concrete” (1998), by P-C Aïtcin, E & FN Spon, London and New York, ISBN 0-419-19270-0.

3.5

References

ACI Committee 357 (1978), “Design and Construction of Fixed Offshore Concrete Structures”, ACI. ACI Committee 363R-84 (1984), “State-Of-The-Art-Report on High Strength Concrete”, ACI Journal. Aïtcin P-C (1998), “High Performance Concrete”, E & FN Spon, London, ISBN 0-419-19270-0. Årskog V, Ferreira, M, Liu, G and Gjørv O E (2007), “Effect of Cement Type on the Resistance against Chloride Penetration”, Proc. V. 1 of 5th Int. Conf. on Concrete Under Severe Conditions: Environment and Loading, ed. by F Toutlemonde, K Sakai, O E Gjørv and N Banthia, Laboratoire Central des Ponts et Chaussées, Paris, ISSN 2-7208-2495-X, 367–374. Årskog V and Gjørv O E (2007), “A New City Development Project in Oslo Harbor With 300 Years Service Life Requirement”, Proc. V. 1 of 5th Int. Conf. on Concrete Under Severe Conditions: Environment and Loading, ed. by F Toutlemonde, K Sakai, O E Gjørv and N Banthia, Laboratoire Central des Ponts et Chaussées, Paris, ISSN 2-7208-2495-X, 851–862.

96


Bache H H (1981), “Densified Cement/Ultra-Fine Particle-Based Materials”, Aalborg Portland, Aalborg, Denmark. DNV – Det Norske Veritas (1974), “Rules for Design, Construction and Inspection of Fixed Offshore Structures”, Oslo, Norway. DuraCrete (1999), “General guidelines for durability design and redesign, The European Union – Brite EuRam III, Project No. BE95–1347: Probabilistic performance based durability design of concrete structures”, Report No. T7-01-1. FIP (1973), “Recommendations for the Design and Construction of Concrete Sea Structures”, London. FIP (1996), “Durability of Concrete Structures in the North Sea”, State-Of-The-ArtReport, FIP, London, ISBN 1-874266-30-1. FIP/CEB (1990), “High Strength Concrete”, State-Of-The-Art-Report, FIP SR 90/1/ CEB Bulletin d’Information No. 197, London, UDC 624-012-46. Gehlen C and Schiessl P (1999), “Probability-Based Durability Design for the Western Scheldt Tunnel”, Structural Concrete (2) 1–7. Gjørv O E (1968), “Durability of Reinforced Concrete Wharves in Norwegian Harbours”, Ingeniørforlaget, Oslo, 208 p. Gjørv O E (1993), “Durability of Concrete Structures and Performance-Based Quality Control”, Proc. Int. Conf. on Performance of Construction Materials in the New Millenium, ed. by A S El-Dieb, M M R Taha and S L Lissel, Shams University, Cairo, ISBN 977-237-191, 10 pp. Gjørv O E (1994), “Steel Corrosion in Concrete Structures Exposed to Norwegian Marine Environment”, Concrete International, 16, (4) 35–39. Gjørv O E (2002), “Durability and Service Life of Concrete Structures”, Proc. of 1st FIB Congress 2002, Session 8, 6, Japan Prestressed Concrete Engineering Association, Tokyo, 1–16. Gjørv O E and Rønning H R (1992), 1st Prize Award in a Compressive Strength of High Strength Concrete Contest, Norwegian Ready Mix Concrete Association, Oslo, Norway. Gjørv O E and Sakai K (2000), “Concrete Technology for a Sustainable Development in the 21st Century”, Proc. of Int. Workshop in Lofoten, Norway, ed. by O E Gjørv and K Sakai, E & FN Spon, London and New York, ISBN 0-419-25060-3, 386 p. Gjørv O E, Bærland T and Rønning H R (1990), “Abrasion Resistance of HighStrength Concrete Pavements”, Concrete International, 12, (1) 45–48. Goodspeed C H, Vanikar S and Cook R A (1996), “High Performance Concrete Defined for Highway Structures”, Concrete International, 18, (2) 62–67. Hasselø J A (2001), “Experiences with Floating Bridges”, Proc. of 4th Int. Symp. on Strait Crossings, Bergen Norway, ed. by J Krokeborg, Swets & Zeitlinger Publ., Lisse, ISBN 90-2651-845-5, 333–337. McGee R (1999), “Modelling of Durability Performance of Tasmanian Bridges”, Proc. of 8th International Conference on the Application of Statistics and Probability, Sydney, Australia. Meaas P, Landet E and Vindøy V (1994), “Design of Sahlhus Floating Bridge (Nordhordland Bridge)”, Proc. of 3rd Int. Symp. on Strait Crossings, Ålesund, Norway, ed. by J Krokeborg, Balkema Publ., Rotterdam, ISBN 90-5410-388-4, 729–734. Moksnes J (2007), Private communication.


97

Moksnes J and Sandvik M (1996), “Offshore Concrete in the North Sea – A Review of 25 Years Continuous Development and Practice in Concrete Technology”, Proc. of Odd E Gjørv Symp. on Concrete for Marine Structures, ed. by P K Mehta, CANMET/ACI Ottawa, Canada, 1–22. Norwegian Association for Harbor Engineers (2007a), “Durable Concrete Harbour Structures-Recommended Specifications for New Concrete Harbor Structures”, TEKNA, Oslo, 2. edit., (In Norwegian). Norwegian Association for Harbor Engineers (2007b), “Durable Concrete Harbour Structures Practical Guidelines for Design and Durability, Quality Control”, TEKNA, Oslo, 2. edit., 48 p. (In Norwegian). OD – Norwegian Petroleum Directorate (1976), “Regulations for the Structural Design of Fixed Offshore Structures”, Stavanger, Norway. Sengul O and Gjørv O E (2007), “Chloride Penetration into a 20-year-old North Sea Concrete Platform”, Proc. V. 1 of 5th Int. Conf. on Concrete Under Severe Conditions: Environment and Loading, ed. by F Toutlemonde, K Sakai, O E Gjørv and N Banthia, Laboratoire Central des Ponts et Chaussées, Paris, ISSN 2-7208-2495-X, 107–116. Zhang M H and Gjørv O E (1991), “Mechanical Properties of High-Strength Lightweight Concrete”, ACI Materials Journal, 88, 240–247.

4 Sprayed concrete (shotcrete) N BANTHIA, University of British Columbia, Canada

4.1

Introduction

Shotcrete is an all-encompassing term used to describe pneumatically projected concrete or mortar using either the dry-mix process or the wet-mix process. “Gunite”, an old term, refers only to the dry-mix process. In the dry-process (Fig. 4.1), a bone-dry cementitious mixture is blown to the nozzle through a hose, where water is added by the nozzleman. Since an intimate mixing of the water and dry materials does not occur at the nozzle, dry-process shotcrete relies heavily on the skills of the nozzleman who manipulates the nozzle in order to produce an effective mixing on the application surface. The amount of water added at the nozzle is critical, as insufficient water will increase both the material rebound and dust, while excessive water will cause the mix to slough off. Only small aggregates are used in dry-process shotcrete, since large aggregates tend to rebound from the application surface. Cement contents are generally high as effective cohesiveness is required to reduce rebound and produce a certain shotcrete “build-up”. In the wet-process, a concrete mixture (typically ready-mix concrete) including all of the mix water is fed to the hopper of the machine, and the mixture is then pumped to the nozzle where compressed air is added to accelerate the mix onto the application surface (Fig. 4.1). Due to the reduced possibility of rebound, large aggregates (12.5 mm or larger) are often used in the mix. Since the nozzleman has no control over the mixture proportioning of the final product, the quality of wet-mix shotcrete is far less dependent upon the skills of the nozzleman. With less dust and less rebound, wet-mix shotcrete is becoming the preferred process of producing shotcrete. The choice of the process – dry or wet – however, depends upon the prevailing conditions at the site and the engineering culture of the region. One way to understand the differences between the two types of shotcrete and concrete is to compare their penetration resistances in the fresh state (Fig. 4.2; Refs. 4 and 5). Such curves are obtained by inserting an 98

Sprayed concrete (shotcrete)

99

A B

Dry-process A: Bone-dry cementitious materials B: Water Wet-process A: Premixed materials with water B: Air

Penetration resistance (MPa)

4.1 Dry-process and wet-process shotcrete (adapted from Ref. 1). 2.5 Dry-mix 2 1.5 1 Cast or wet-mix 0.5 0 0

2

4 6 Deflection (mm)

8

10

4.2 Consistency of shotcrete as measured by instrumented penetrometer.

instrumented penetrometer in the fresh state, and are often used in the case of dry-process shotcrete to determine its water content and ascertain its acceptability. As seen in Fig. 4.2, right after its placement, the in-situ penetration resistance of dry-process shotcrete is significantly higher than both cast concrete and wet-process shotcrete. It is generally believed that, in the case of shotcrete, the early products of cement hydration never get the time to rearrange themselves within the mix, resulting in a much greater penetration resistance.

4.2

Mix proportioning and process implications

The matrix in shotcrete is different from that in cast concrete, and has its own distinct features. Typical mix designs are given in Table 4.1. Placement using pneumatic compaction and lack of forms in shotcrete requires that the material be more cohesive, adhere well to the surface and resist sloughing off. Aggregates preferred in shotcrete are therefore generally rounded. Highly angular aggregates are known to cause problems in pumping, and

100


Table 4.1 Typical mix designs for dry- and wet-process fiber reinforced shotcrete Components

Dry-process 3

kg/m Cement Silica fume additives Blended aggregate Steel fibers Accelerators Superplasticizer Water reducer Air entraining agent Water Total

420 50 1670 60 13 – – – 170 2383

Wet-process % dry materials 19.0 2.2 75.5 2.7 0.6 – – – – 100

kg/m3

% wet materials

420 40 1600 60 13 6L 2L If required 180 2321

18.1 1.7 68.9 2.6 0.6 0.3 0.1 If required 7.7 100

also increase the possibility of rebound. Further, aggregates used in shotcrete tend to be smaller than in cast concrete. High aggregate rebound and placement difficulties further dictate that shotcrete mixes contain far less total aggregate content (fine and coarse) than cast concrete (2,3). While cast concrete has anywhere from 50–60% coarse aggregate by mass, shotcrete mixes only have 30% coarse aggregate content, which is further reduced during shooting due to rebound. Supplementary cementing materials (such as fly ash, silica fume and metakaolin) are used more commonly in shotcrete than in conventional cast concrete in order to achieve improved pumpability, better adhesion and cohesion, greater build-up thickness, reduction in the required quantity of accelerator, and a reduction in rebound. Shotcrete mixes already have higher than normal cement contents (400–450 kg/m3), and a further increase in the in-situ cement content occurs due to the higher than proportional rebound of aggregates during shooting; in the dry-process, it is not uncommon to have in-situ cement contents approaching 600–700 kg/m3 (4,5). Once in place, shotcrete also has a different spatial distribution of its various components. Water in dry-mix shotcrete, given that it is introduced only at or near the nozzle, is far less uniformly distributed through the placement than in cast concrete. In the wet-process, although the water is uniformly distributed, the pneumatic compaction results in internal voids that are far different in size ranges and spatial distribution than in cast concrete. Furthermore, the lack of bleed channels in shotcrete produces a material with different internal structure and transport properties than traditional cast concrete. Shotcrete, therefore, owing to its distinct mix design and placement procedures, develops a number of properties different from cast concrete.


101

4.3 X-ray image showing a preferential fiber orientation in shotcrete (6).

Unlike ordinary fiber reinforced concrete, in which fiber orientation is known to be three-dimensional random, in shotcrete, fibers are distributed in a more or less two-dimensional random fashion (6) (Fig. 4.3), due to the pneumatic placement. This introduces anisotropy, and has a clear influence on the reinforcing efficiency of the fibers.

4.3

Strength and stiffness

Shotcrete displays a different rheology, strength gain mechanism, compressive/tensile strength ratio, and creep characteristics. A commonly used indicator of matrix quality in cast fiber reinforced concrete is the water/cement ratio, which is expected to have a unique relationship with strength. The application of this relationship, however, requires, among other things, that the basic condition of complete consolidation be met – a condition rarely met in shotcrete. Another difficulty in applying a water/cement ratio vs. strength relationship to dry-process shotcrete is that an exact determination of the in-situ water/cement ratio in the mix is never possible (5). These factors when combined result in a poor correlation between the water/ cement ratio and strength of dry-process shotcrete, as illustrated in Fig. 4.4 (7). In the case of shotcrete, therefore, a direct measurement of porosity is necessary for assessing its strength and quality. When the volume of permeable voids (ASTM C642-97) is plotted as a function of strength, a much better correlation is seen to exist (5,7).

102


f¢c (MPa)

90

30 0

Exponential curve

Armelin; 1997 Bindiganavile; 1998

60

fc = 132.6e–1.93w/c R = 0.32 0

0.1

0.2

0.3

0.4

0.5

0.6

4.4 Water/cement ratio vs. dry-mix shotcrete compressive strength (7). Note a poor correlation.

For wet-mix shotcrete, while the water/cement ratio is the same as the one in the initial mix, strengths can sometimes be as much as 30% lower than its cast counterpart (8). Shotcrete has richer mixes to begin with, and during the process of shooting the rebound of aggregate renders the mixes even richer. It is well known that for concrete of a given water/cement ratio, leaner mixes with a higher aggregate/cement ratio develop higher strengths (9) and wet-mix shotcrete appears to follow a similar trend. However, when the flexural strength is considered, comparisons indicate that richer shotcrete mixes acquire greater flexural strengths than their cast counterpart for an identical water/cement ratio (8). Another property of interest is the deformability of the matrix itself as defined by its elastic modulus. Dry-mix shotcrete with a greater loss of aggregate through rebound exhibits a consistently lower elastic modulus than both wet-mix shotcrete and cast concrete (10). Wet-mix shotcrete, on the other hand, develops a different internal structure and creep characteristics from those of cast concrete (10).

4.4

Kinematics and rebound

One primary concern with the dry-process shotcrete, as discussed previously, is the high rebound; nearly 20–40% of material and up to 75% of fibers may be lost through rebound (4). The use of various mineral admixtures in shotcrete has therefore increased dramatically in order to control rebound (3). The rebound performance of four mineral admixtures – silica fume (SF), carbon black (CB), high reactivity metakaolin (HRM) and fly ash (FA) – in dry-process shotcrete is compared in Fig. 4.5 at 10% cement replacement rate (5). Notice that steel fiber rebound is always greater than the material rebound and increases proportionately with it. Note also a definite increase in rebound with an increase in the mean particle size of the admixture. The shape of the particle also appears to be important; while fly ash, silica fume and carbon black are all spherical, HRM has an irregular “platelet” structure.


Rebound (%)

40 30

CB - 0.05 μm SF - 0.1 μm HRM - 1.0 μm Fly ash - 10 μm Cement - 50 μm

Fiber y = 0.1853x + 35.598 R2 = 0.0412

Material

20

103

y = 2.3022x + 22.343 R2 = 0.3371

10 0 –1.5

–1

–0.5

0

0.5

1

1.5

2

Log (a); a = Mean size (μm)

4.5 Material and fiber rebound as a function of admixture particle size. Table 4.2 Particle velocities in dry-mix shotcrete Study

Reported particle velocity (m/s)

Stewart (1933)11 Ryan (1973)12 Valencia (1974)13 Ward and Hill (1977)14 Parker (1976)15 Blume et al. (1978)16

90–50 90–120 135 35–56 30–60 27–35

An additional concern, as discussed previously, is that the rebound material consists primarily of aggregates. Lack of sufficient aggregate volume for volumetric stabilization and insufficient curing render shotcrete highly susceptible to cracking due to plastic shrinkage (17). Pneumatic compaction in shotcrete necessarily requires its placement at a high velocity. In Table 4.2 some typical measurements of particle velocity are given (1). Notice the wide scatter in measured velocity values. Further, due to differences in particle shape, size and specific gravities, various particles will acquire different velocities in the shotcrete stream – a fact not reflected in Table 4.2. Using high-speed photography, an extensive investigation was carried out at the University of British Columbia on understanding the kinematics of particle motion in shotcrete. Figure 4.6 shows some high-speed images of a particle as captured by a high-speed camera running at 1000 frames/sec. These images were used to develop a generalized kinematic model of particle motion in shotcrete and combined with plasticity theory to predict rebound (18–21). Essentials of this model as applied to aggregates particles are presented below. Based on high-speed imaging, aggregate velocity (V) was found to be inversely proportional to its size (f) at a given air volume and would be described by a general expression (A and B are constants): 1 V = A ⎡⎢ ⎤⎥ ⎣f ⎦

B

4.1

104


(1) t = 0 ms

(2) t = 16 ms

(3) t = 33 ms

(4) t = 53 ms

Rebound

4.6 High-speed images of a fiber in dry-process shotcrete stream.

In modeling rebound in shotcrete, the usual assumption of fresh concrete being a Bingham fluid is of limited usefulness since it fails to account for the elastically stored energy in the fresh substrate, which when transferred back to the impinging particles may cause it to “pop” out of the bed. In addition, it is necessary to quantify the resistance a particle would experience to its penetration into the substrate at a high speed (which brings into relevance the strain-rate sensitivity of the substrate) and the adhesion developed between the substrate and the particle which would then resist debonding of the latter. For an experimental assessment of this process, a small portable penetrometer with a 25.4 mm diameter hemispherical ball was developed which is inserted into the bed of fresh shotcrete and then pulled out. In actual modeling of rebound, the process could then be divided into two phases, the penetration phase and the reaction phase. During the penetration phase, one could make use of the theory of plasticity (22) to develop the parameters of the hydrostatic stress-field that developed around the impinging aggregate. The final depth

105

b


p

dh

da

p

R

c

p a

p

b

du(a)

dh

p a p p p

4.7 Contact stress field developed during shotcreting.

of penetration of the particles would be determined simply by balancing the work necessary to create the indentation (pd · Va) and the kinetic energy available in the fast moving particle (W1). In other words (Fig. 4.7): W1 = pd ⋅ Va

4.2

In the reaction phase, the rebound energy (W2) is given by (22) W2 =

3p 2 2 3 ⎡ 1 − n c2 1 − n c2 ⎤ + p a* ⎢ 10 Ei ⎥⎦ ⎣ Ec

4.3

The coefficient of restitution, e, which is the percentage of energy returned to the particle for rebound is then given by e2 =

V′ V ′2 W2 = = 2 V V W1

4.4

From Eqns 4.2, 4.3 and 4.4, e can be further written as e2 =

( ) ( )( )

3p 2 4 R 10 p

3/4

1 m E* 2

−1 / 4

⋅ p ⋅ ( pd )−3 / 8 ⋅ V −1 / 4

4.5

where m is the mass of the impacting particle, p is the static penetration resistance (peak contact stress as determined using the penetrometer), pd is the dynamic penetration resistance and V is the particle velocity. The values of pd are obtained experimentally as the slope of the particle kinetic

106


energy (W1) vs. volume displaced (Va) curve (Eqn 4.2). The values of p generally varied between 0.3 MPa (wet consistency) and 1.0 MPa (dry consistency) with a typical values of 0.5 MPa (Fig. 4.2). The dynamic contact stress, pd, on the other hand was typically greater than 3 MPa and depended strongly upon the mix design parameters. Equation 4.5 can be further simplified by using the expressions, Ec = Y/ecp, where Y is the yield strength of the substrate given by Y = p/3. e2 =

( )

3p 2 4 R 10 p

3/4

(2.25e cp )

( ) m 2

−1 / 4

⋅ p1 / 2 ⋅ pd −3 / 8 ⋅ V −1 / 4

4.6

If one can assume that ecp is a constant, Eqn (4.6) can be rewritten in terms of a constant K and an impact factor, Y: e = K .y where K =

4.7

( )

3p 2 4 R 10 p

3/4

(2.25e cp )

( ) m 2

−1 / 4

and y = p1 / 2 ⋅ pd −3 / 8 ⋅ V −1 / 4

Notice that in the final form (Eqn 4.7) the rebound energy of a particle can be entirely represented by the impact factor Y. With a known value of the coefficient of restitution e, the available energy for rebound W2 can be calculated. This can then be equated to the energy necessary for debonding (WD) given by WD = [s 0 (pa *2 )]d *

4.8

And finally, the condition of rebound is given by W2 ≥ 1 ⇒ REBOUND WD

4.9

During the computations, the process of rebound was modeled as a stochastic process with particle velocity, static and dynamic contact stresses and the ultimate concrete strain taken as the stochastic variables. The model thus determined the probability of rebound for a given particle size and when all particle sizes in a given mix were considered, the overall rebound could be predicted. Some model predictions are compared with the experimental findings in Figs 4.8–4.11. In Fig. 4.8, aggregate rebound is plotted as a function of aggregate size. Notice that the model predicts a linear variation in rebound with the logarithm of aggregate size, which relates very well to the experimental observations. In Fig. 4.9, the influence of silica fume content is shown. Once again, the model accurately predicts a decrease in rebound with an increase in the silica fume content as observed in reality. In Fig. 4.10, rebound dependence on cement content is shown. Notice a drop in rebound with an increase in the cement content, and a good prediction by

Aggregate rebound (% by mass)


107

80 70 60 50

Experimental, R2 = 0.89

40

2

Analytical, R = 0.99

30 20

Experimental Analytical

10 0 0.01

0.1

1 10 Aggregate size (mm)

100

4.8 Rebound as predicted by model for various aggregate sizes.

Overall rebound (% by mass)

50 Experimental Analytical

45 40

Analytical, R2 = 1

35 30


25 20 –2

0

2

4

6

8

10

12

14

16

18

Silica fume content (% mass subst to cement)

Overall rate of rebound (% by mass)

4.9 Rebound as predicted by model for various silica fume contents. 55 Experimental Analytical

50 45 40


35 30

2 Analytical, R = 0.99

25 20 300

350 400 450 500 550 Cement content of the mix (kg/m3)

600

4.10 Rebound as predicted by model for various cement contents.

108


Overall rebound (% by mass)

70 60

Analytical, R2 = 0.94

50 Experimental, R2 = 0.84

40 Experimental Analytical

30 20

0

1

2

3

4

5

6

7

8

9

Shooting consistency to 9 mm cylindrical needle (MPa)

4.11 Rebound as predicted by model for various shooting consistencies.

the model. Finally, the influence of shooting consistency on rebound is plotted in Fig. 4.11. Notice an increase in the rebound when shotcrete with a higher shooting consistency (i.e., a greater resistance to penetration needle) is shot. Note also a good match between the experimental findings and the model predictions. In the case of fibers, the fiber rebound is shown to be proportional to a specific fiber parameter called the modified aspect ratio (l/√d) (23). High fiber rebound in shotcrete remains a critical issue and significant further research is needed.

4.5

Toughness, impact resistance and fiber reinforcement

Due to the nature of the applications, requirements of material deformability, toughness and energy absorption are often greater in shotcrete than in cast concrete. In repair applications, a greater toughness and cracking resistance is required for durability. When used as ground covering or support, two distinct situations may arise: in the first, slow quasi-static ground movements may occur and in the second, rapid or dynamic ground deformations may occur. The latter, often called “rock-bursts”, impose very severe toughness requirements on shotcrete as discussed below. For an enhancement of quasi-static toughness, energy absorption and impact resistance fibers are generally used in shotcrete. Fibers control cracking, and act as stress-transfer bridges thus enhancing resistance to crack growth and fracture (24). Some typical load deflection responses for fiber reinforced shotcrete beams are shown in Fig. 4.12. While plain shot-


Load (kN)

20 18 16 14

109

DD fiber (0.5%) hybridized with 0.5% of secondary micro fiber

DD fiber at 0.5%

12 10 8 6 4 2

DD fiber at 0.3%

DD fiber at 0.1%

0 0

0.2

0.4

0.6

0.8 1 1.2 Displacement (mm)

1.4

1.6

1.8

2

4.12 Toughness of shotcrete (ASTM C1018) for various fiber systems. 35

Material Fiber

Rebound (%)

30 25 20 15 10 5 0 Plain-Dry Plain-Wet 30 kg/m3

60

30

60

kg/m3

kg/m3

kg/m3

Dry-mix

Wet-mix

4.13 Comparison of two processes for fiber reinforced shotcrete: rebound.

crete beams would fail in a brittle manner, fiber reinforced shotcrete beams would carry loads far beyond the peak load resulting in a curve with a long descending branch. The area under this curve is a measure of the absorbed energy and it is often called “toughness”. Note in Fig. 4.12 that the improvements in “toughness” are proportional to fiber volume fraction, and that fiber hybridization appears to be highly effective. The process used for shotcrete, dry or wet, is expected to influence not only the rebound values but also the strength, toughness, and long-term durability of fiber reinforced shotcrete. In Fig. 4.13, the two processes are compared on the basis of rebound (25). Notice that the rebound in the dryprocess – for both material and fiber – far exceeds the rebound in the wetprocess. Note also that the percentage rebound is inversely proportional to the initial fiber content in the design mix.

110


Load (kN)

20 16 Batched: 60 kg/m3 In-situ: 50 kg/m3 Batched: 30 kg/m3 In-situ: 24 kg/m3

12 8 4 0

Plain dry-mix shotcrete

0

1

0.5

1.5

2

2.5

3

3.5

4

Deflection (mm) (a)

Load (kN)

20 16

Batched: 60 kg/m3 In-situ: 57 kg/m3

12 8 4 0

Batched: 30 kg/m3 In-situ: 28 kg/m3

Plain wet-mix shotcrete

0

0.5

1

1.5

2

2.5

3

3.5

4

Deflection (mm) (b)

4.14 Comparison of flexural toughness for the two processes: (a) dryprocess and (b) wet-process.

In Fig. 4.14, “flexural toughness” curves of fiber reinforced shotcrete produced by the two processes are compared. These curves were obtained as per the ASTM C1018 procedure.* Notice that the “toughness” of wetprocess shotcrete for a given initial design mix fiber content is greater than that of the dry-process shotcrete. This is expected, as the fiber rebound in the dry-process is far greater than that in the wet-process. However, a closer observation of Fig. 4.14 reveals that even for a given effective in-place fiber content, the wet-process produces a greater toughness. The reasons for this are not clear, and in fact the observation is a bit puzzling as the fibers in the dry-process shotcrete are expected to be more preferentially aligned (see Fig. 4.3) than in the wet-process shotcrete and this preferential orientation would be expected to produce a better toughness. This, however, does not appear to be the case. Rock-bursts occur in tunnels in hard-rock mines due to high in-situ, mining and seismically induced stresses. Three main mechanisms (26) are * ASTM C1018 has recently been withdrawn, and replaced by ASTM C1609. However, the procedure for obtaining the load vs. deflection curves is essentially the same in both standards.


111

recognized: seismically induced rock falls where a seismic wave accelerates a volume of rock, rock mass fracturing and bulking due to strain build-up at an opening, and finally, rock ejection due to momentum transfer from a remote disturbance. When ejection of a rock occurs, it is not uncommon to have rocks almost 1 m size ejected with typical ejection velocities of about 6 m/s (27) and as high as 50 m/s (28). Typically, support systems are expected to withstand a dynamic energy release of about 100 kJ/m2 (29,30), and although some full-scale impact tests have been reported (31), there has hardly been any systematic study of the influence of such dynamic impact events on shotcrete linings of different thickness, with different types of reinforcement including fibers, and rock bolting patterns. Particularly lacking is a thorough understanding of the constitutive response of shotcrete materials under variable strain-rates, and the interaction between the retaining elements and the containment elements and its influence on the structural response of the entire assembly.

4.6

Concluding remarks

This chapter presents a brief state-of-the-art on shotcrete. It demonstrates that due to its unique mix designs, placement techniques, compaction dynamics, strength gain mechanisms, and internal structure, shotcrete is distinctly different from cast concrete, and hence our conventional understanding of cast concrete should be applied to shotcrete only with caution. Significant further efforts are necessary for a fundamental understanding of the nature of shotcrete and to develop materials suitable for various applications. This chapter places particular emphasis on fiber reinforced shotcrete and highlights the various benefits of fiber inclusions in shotcrete.

4.7

Acknowledgements

The author would like to thankfully acknowledge the continued financial support of the Natural Sciences and Engineering Research Council of Canada.

4.8

References

1 Austin, S.A. and Robins, P.J. (Eds), Sprayed Concrete: Properties, Design and Applications, Whittles Publishing, 1995. 2 Gilkey, H.J., “Water-Cement Ratio vs. Strength: Another Look”, J. of American Concrete Institute, 57(10), 1961, 1287–1312. 3 Morgan, D.R., McAskill, N., Neill, J. and Duke, N. F., “Evaluation of silica fume shotcretes”, Proceedings of CANMET/ACI workshop on condensed silica fume in concretes, Montreal, 1987.

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4 Armelin, H.S., “Rebound and Toughening Mechanisms in Steel Fiber Reinforced Dry-mix Shotcrete”, Ph.D. Thesis, University of British Columbia, 1997. 5 Bindiganavile, V. and Banthia, N., “Rebound in Dry-Mix Shotcrete: Influence of Type of Mineral Admixture”, ACI Materials Journal, 97(2), 2000, 1–5. 6 Ramakrishnan, V., Coyle, W.V., Dahl, L.F. and Schrader, E.K., “A Comparative Evaluation of Fiber Shotcretes”, Concrete International, Jan. 1981. 7 Banthia, N. and Bindiganavile, V., Proc. International Conference on Infrastructure Regeneration and Rehabilitation, Sheffield, 1999. 8 Banthia, N., Trottier, J.-F. and Beaupré, D., “Steel Fiber Reinforced Shotcrete: Comparisons with Cast Concrete”, ASCE, J. of Materials in Civil Eng., 6(3), 1994, 430–437. 9 Neville, A., Properties of Concrete, Fourth edn. John Wiley and Sons, Inc., 1996. 10 Chan, C., Banthia, N., and Sakai, K., Proc. Int. Workshop on Concrete Technology for a Sustainable Development in the 21st Century, Svolvaer, Norway, June 1998. 11 Stewart, E.P., New Test Data Aid Quality Control of Gunite. Engineering NewsRecord, Nov. 9, 1933, 4. 12 Ryan, T.F., Gunite: A Handbook for Engineers. Cement and Concrete Association, Wexham Springs, 1973, 63. 13 Valencia, F.E., Practical Aspects of Shotcrete Application. Use of Shotcrete for Underground Structural Support, SP-45, ACI/ASCE, Detroit, 1974, 114–129. 14 Ward, W.H. and Hills, D.L., Sprayed Concrete – Tunnel Support Requirements and the Dry-mix Process, Shotcrete for Ground Support, SP-54, ACI, Detroit, 1977, 475–532. 15 Parker, H.W., Field Oriented Investigation of Conventional and Experimental Shotcrete for Tunnels, PhD. Thesis, University of Illinois at Urbana-Champaign, USA, 1976, 630. 16 Blumel, O.W., Lutsch, H. and Stehno, G., State-of-the-Art Shotcrete Technology. Shotcrete for Underground Support III, Engineering Foundation, New York, 1978, 15–26. 17 Banthia, N. and Campbell, K. “Restrained Shrinkage Cracking in Bonded Fiber Reinforced Shotcrete”, RILEM – Proc. 35, The Interfacial Transition Zone in Cementitious Composites, Eds. Katz, Bentur, Alexander and Arligui, E and F N. Spon, 1998, 216–223. 18 Armelin, H. S. and Banthia, N., ‘Mechanics of Aggregate Rebound in Shotcrete – (Part 1), RILEM, Materials and Structures, 31, March 1998, 91–98. 19 Armelin, H.S. and Banthia, N., “Development of a General Model of Aggregate Rebound in Dry-mix Shotcrete – (Part 2)”, RILEM Materials and Structures, 31, April, 195–202. 20 Armelin, H.S., Banthia, N., Morgan, D.R. and Steeves, C., “Rebound in Dry-Mix Shotcrete”, Concrete International, 19(9), 1997, 54–60. 21 Armelin, H.S., Banthia, N. and Mindess, S., “Kinematics of Dry-Mix Shotcrete”, ACI Materials J., 96(3), 1999, 283–290. 22 Hill R., The Mathematical Theory of Plasticity, Oxford University Press, London, 1950. 23 Armelin, H.S. and Banthia, N., “Steel Fiber Rebound in Dry Mix Shotcrete: Influence of Fiber Geometry”, ACI Concrete International, 20(9), 1998, 74–79.


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24 Banthia, N., Trottier, J.-F., Wood, D. and Beaupré, D., “Steel Fiber Dry-Mix Shotcrete: Influence of Fiber Geometry”, ACI Concrete International, 14(5), 1992, 24–28, and “Influence of Fiber Geometry in Wet-Mix Steel Fiber Reinforced Shotcrete”, ACI Concrete International, 16(6), 1994, 27–32. 25 Bindiganavile, N. and Banthia, N., “Process Dependence of Shotcrete for Repair”, International Journal of Materials & Product Technology, UK, 2005, 23(3), 240–256. 26 Kaiser, P.K., Proc. 3rd Int. Symp. on Rock Burst and Seismicity in Mines, Balkema, Rotterdam, 1993, 13–27. 27 Jager, A.J., et al., Proc. Int. Deep Mining Conference: Technical Challenges in Deep Level Mining, South African Institute of Mining and Metallurgy, 1990, 1155–1177. 28 Ortlepp, W.D., 3rd Int. Symp. on Rock Burst and Seismicity in Mines, Balkema, Rotterdam, 1993, 101–106. 29 Stacey T.R., Ortlepp, W.D. and Kirsten H.A.D., J of the South African Institute of Mines and Metallurgy, May–June 1995, 137–140. 30 Kirsten, H.A.D., “Fiber Reinforced Shotcrete”, World Tunneling, Nov. 1997, 411–414. 31 Tannant, D.D., Kaiser, P.K. and McCreath, D.R., Large Scale Impact Tests on Shotcrete, Laurentian University, March 1995, 45.

5 Hot weather concreting C ISHEE, Florida Department of Transportation, USA

5.1

Introduction

Hot weather concreting is generally referred to as the placement of concrete in weather conditions which require attention to avoid excessive heating or drying of the placed concrete. Building codes and specifications generally limit the maximum placing temperature of concrete to 30°C [85°F] to ensure that the concrete does not set early or show signs of plastic shrinkage cracking upon hardening. When the concrete is placed at temperatures above this, it is commonly termed hot weather concreting. Hot weather concrete does not only refer to placing at higher temperature, but can include other situations that would cause similar effects such as high air temperature, low relative humidity, wind velocity and intensity of solar radiation. Any of these situations or combination thereof can adversely affect the quality of fresh and hardened concrete properties (ACI 305R-99). Most of the time these situations cannot be avoided and a high quality concrete is desirable to produce a structure that will have a long service life. Certain precautionary measures are required to ensure the desired service life is obtained.

5.2

Applications/case studies

Difficulties in placing concrete in high temperatures include the likelihood of problems with transporting, finishing and consolidation, which typically results in a hardened concrete product with lower strength and subject to cracking. These effects can be overcome to produce a high performing concrete if material selection and evaluation is performed carefully, concrete placing temperatures are controlled, plastic properties are taken into consideration, and anticipated hardened properties are used in the design of the structure. 114

Hot weather concreting

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5.2.1 Material selection and mix design evaluation One of the best practices to ensure adequate performance from a concrete mixture is to select a concrete mix design that is going to be appropriate for the placement and environment that the structure will be exposed to. Many of the typical mix design calculations can be used in choosing the proportion of coarse aggregates, fine aggregates, cement, mineral admixtures, chemical admixtures and water. One of the more difficult issues with hot weather concreting is to address the selection of cement type, mineral admixture addition and chemical admixtures used for the concrete mix. Another aspect of concrete mix design that is often overlooked is creating a trial batch and exposing it to conditions that simulate hot weather concrete conditions to ensure a successful placement.

Selection of cement Selection of the most appropriate cement for an application can be difficult. ASTM C150 specifies the use of five different types of Portland cements. The two types of cements applicable to hot weather concrete are a slowersetting Type II or Type IV. The Type II cement is generally selected because of the lower heat of hydration and lower content of tricalcium aluminate (C3A). The Type IV cement is best for massive and hot weather concrete applications but is not readily available in the United States or other parts of the world. Concretes produced with the slower-setting cements will be less likely to exhibit plastic-shrinkage cracking. Selecting a Portland cement with a slower rate of heat development and simultaneously producing a lower dissipation of heat from the concrete will result in lower peak temperatures. This will result in less thermal expansion and a lower risk of thermal cracking (ACI 305R-99). Most recently, the specification for cement (ASTM C150 or AASHTO M 85) has included a maximum heat of hydration to reduce the amount of heat generated by the typical Type II cement. The Florida Department of Transportation allows a maximum on 335 kJ/kg [80 cal/g] for concrete placed in severe environmental conditions. If the Type II cement exceeds this limit, but is less than 370 kJ/kg [88 cal/g], then the concrete mix design must have a mineral admixture to replace a portion of the cement to reduce the total amount of heat generated (FDOT 2007). If available, blended hydraulic cement (ASTM C1157 or ENV197-1) is better for most hot weather applications. Blended hydraulic cements are obtained by creating a composite of reactive materials such as: Portland cement clinker, granulated blast furnace slag, pozzolanic materials, fly ash, burnt shale or silica fume. Selection of an appropriate hydraulic cement would be based on four requirements:

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(1)

If the hydraulic cement contains lime as calcium as CaO and magnesia as MgO, then their presence must be in the form of free lime and not periclase. This can be tested by measuring the autoclave expansion and contraction (ASTM C595). (2) Investigating the demand for water of the hydraulic cement. As the amount of the secondary material increases the lowering of water demand is a considerable attribute. (3) Have hydraulic cement that has a heat of hydration between 200 kJ/kg [48 cal/g] and 250 kJ/kg [60 cal/g] at 72 hours. This type of hydraulic cement should produce enough heat to set the concrete in a workable time, but not so much heat as to induce cracking. (4) Determine if there is potential for any delayed expansion problems in the concrete. Duggan and Scott (1989) proposed a test which states that if more than a 0.05% expansion occurs within 20 days of casting then there is a high risk of deleterious expansion and potential of delayed expansion problems (Owens 1992). One of the trends in current construction practices is to obtain access to in-place concrete elements as quickly as possible. As such, there is an increased demand for early strength requirements. Therefore, there has been a general trend to increase the Portland cement content in a typical mix design. However, for hot weather concrete applications, such practice is contradictory to getting a better performing element. Hot weather concreting results in an increase in the hydration rate and thereby increases the early-age properties. When additional Portland cement is added to the mix, additional production of heat will be generated within the element at early ages. This excessive heat production will result in an increase in the thermal expansion of the concrete and increase the potential for thermal cracking. To reduce the risk of excessive early age thermal expansion, the reduction of Portland cement within the mixture and the replacement of portions of the Portland cement with mineral admixtures are beneficial. Additionally, lower water content in the concrete below 150 liters/m3 [252 lbs/yd3] is desirable to reduce the plastic shrinkage cracking (Owens 1992).

Use of mineral admixtures Mineral admixtures have been used in concrete for as long as concrete has been made. However, some of the benefits of mineral admixtures are just recently being understood. Some typical mineral admixtures used in concrete are fly ash, ground granulated blast furnace slag and silica fume. Each of these mineral admixtures affects the properties of the concrete differently in hot weather conditions. Fly ash is the byproduct of coal burning power plants and is used to replace portions of the Portland cement in mixture design. ASTM C618


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outlines the requirements of fly ash for use in concrete as a mineral admixture. Fly ash typically is used at a replacement rate of 20–40% of ordinary Portland cement. Fly ash can reduce the rate of slump loss of concrete under hot weather conditions and this reduction in slump loss is inversely proportional to the percentage of cement that was replaced. Fly ash typically reduces the early age rate of strength gain of the concrete, but this is recovered at later ages. Due to the reduction in early age strength, fly ash has a greater potential for plastic shrinkage cracking and requires adequate curing to protect the concrete from cracking (Soroka and Ravina 1998). Ground granulated blast furnace slag (GGBFS) is a byproduct of the steel mill industry and is also used as a mineral admixture in concrete production. ASTM C989 outlines the requirements of GGBFS. GGBFS typically is used at a replacement rate of 30–70% of ordinary Portland cement. If mixed and cured properly, concrete made with GGBFS has better performance characteristics in hot weather conditions than ordinary Portland cement with respect to strength and pore structure. Concrete with GGBFS should be moist cured as quickly as possible and for a minimum of 7 days to ensure that the material achieves its benefits (Austin and Robins 1992). Silica fume is a byproduct of producing silicon metal or ferrosilicon alloys and needs to meet the requirements of ASTM C1240. Silica fume typically is used at a replacement rate of 5–10% of ordinary Portland cement. Silica fume can be used to increase the strength of concrete and to reduce the permeability. The increase in strength causes an inverse in plastic shrinkage of the concrete. Increasing the percentage of silica fume in a concrete mix will also increase the plastic shrinkage. However, the plastic shrinkage, if controlled properly, can be permitted with no cracking in the concrete due to the early increase in strength. The fineness of the silica, as represented by its specific surface area and bulk density, is a strong indicator of its potential to increase plastic shrinkage cracking in hot weather (Al-Amoudi et al. 2006). In Florida, the typical concrete mix design for hot weather conditions will use a fly ash or GGBFS. These materials typically cost less than Portland cement and both can help reduce the amount of heat generated in the concrete mix design. When the concrete is going to be placed in a severely aggressive environment, exposed to sea water and in hot weather conditions, the typical mix design will also have silica fume as a ternary blend to provide early strength gain and reduce the amount of heat generated. Use of chemical admixtures There are many types of chemical admixtures that can provide better performance characteristics of concrete. Some of these benefits include lower

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water demand and extended periods of workability. Chemical admixtures can provide these benefits without any reduction in strengths comparable to concrete without chemical admixtures placed at lower temperatures. The effectiveness of the chemical admixtures is dependent upon the reactions with the cement they are to be used with. Any chemical admixture without a history of better performance in hot weather conditions should be evaluated prior to use (ACI 305R-99). Chemical admixtures should be added in accordance with the manufacturer’s technical data sheets. The dosage of most chemical admixtures is mix design specific and should be evaluated prior to use. Some chemical admixtures are designed to allow for an extended setcontrol of the freshly mixed concrete. Most extended set-control admixtures comply with the requirements of ASTM C494 as a Type B, retarding admixture, or Type D, water-reducing and retarding admixture. These admixtures are often referred to as hydration control admixtures and benefit the concrete in that they can temporarily stop the hydration process of both the silicate and aluminate phases in the Portland cement. These extended set-control admixtures are designed to allow for longer haul times or for additional finishing times when needed. Water-reducing admixtures are chemical admixtures designed to reduce the water:cement ratio of concrete without adversely affecting the rheological properties. Most of these water-reducing admixtures comply with the requirements of ASTM C494 as Type A, water-reducing, or Type F, highrange water-reducing mixtures. One major benefit of these materials is that they provide up to 15% of the water in a concrete mix design. Typically, water-reducing admixtures do not affect the setting time of the concrete at lower dosages, but at higher dosages can increase the setting time. Other chemical admixtures can provide high-range, water-reducing and retarding effects on the freshly mixed concrete. Most high-range, waterreducing and retarding admixtures will comply with the requirements of ASTM C494 as a Type G and ASTM C1017 as a Type II for plasticizing and retarding admixtures. These admixtures are often referred to as superplasticizers and can provide significant benefits for producing flowing concrete in hot weather concrete conditions (ACI 305R-99). Most superplasticizers are synthetic water-soluble polymers such as sulfonated naphthalene formaldehyde (SNF), sulfonated naphthalene polymer (SNP), modified sugar-free lignosulfonate polymer (MLP), and most recently polycarboxylic ether polymers (PCE). Research has shown that the type of superplasticizer affects the plastic shrinkage strain in the concrete (AlAmoudi et al. 2006). The interaction between the cement and the superplasticizer is crucial because there have been cases where the wrong combination resulted in faster slump loss and additional plastic shrinkage (Ravina and Soroka 2002).


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Most of the concrete produced in the United States will have some form of chemical admixture added to the mix design. With hot weather conditions, admixtures are typically used to control the plastic properties of the mix without any long-term strength reductions. It is not uncommon to see multiple chemical admixtures in a mix design designed for hot weather conditions. In Florida, a combination of a high-range, water-reducing and retarding admixtures will be used with a mid-range, water-reducing admixture. The high-range admixture will be kept constant in all of the production loads of concrete and the mid-range will be adjusted for changes in temperature, humidity, and absorption of the aggregates. Because of the amounts of chemical admixtures being used, a majority of the concrete facilities have concrete mix designs for the hot weather summer conditions and a different set of mix designs for winter conditions. Mix design verification process Prior to large-scale production, mix designs should be verified for use in hot weather concreting. A trial mix should be batched in accordance with ASTM C192 with a few exceptions. The batch size should be large enough to reproduce typical issues regarding the heat produced by the cementitious materials, at least 0.1 m3 [3.5 ft3]. The slump of the trial batch should be determined after the initial mixing procedure (three minutes of mixing, three minutes of rest, and two minutes of remixing). After mixing, the mixer should be stopped and covered with wet burlap or an impermeable cover material. The trial batch should remain in the mixer for at least 90 minutes or the anticipated transit time after the completion of the initial mixing procedure. During the extended mixing period, the concrete should be remixed intermittently for 30 seconds every five minutes after which it should be re-covered. At the end of the 90 minute period, the concrete should be remixed for a minimum of one minute and a slump reading should be obtained to verify that the concrete slump is within the desired range. If the slump is below the target range, water may be added to adjust the slump to achieve the desired results. If additional water is added, the concrete should be remixed for a minimum of two minutes. Other desired plastic properties (such as temperature, air content, and unit weight) should be acquired and samples should be cast to obtain hardened properties. It is important to ensure that the mix temperature is not less than 35°C [95°F] at any time during the mixing process. Plastic properties of the concrete should be within the allowable ranges or close to the desired target after water additions and final mixing. The procedure listed above is the process that is used to approve any concrete mix design for hot weather use on a Florida Department of Transportation construction project (FDOT 2007).

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The total amount of water added to the mix (including any additional water added to adjust the final slump) constitutes the design mix water content. Ensure that the total water to binder ratio does not exceed the maximum water to binder ratio desired or specified. If a procedure for cooling of the concrete is to be evaluated, it should follow the above stated mixing procedure. As an alternative to the trial mixing procedure, a full-size production batch may be used for verification of the mixture proportions, provided the required temperature levels of the concrete are obtained (ACI 305R-99).

5.2.2 Cooling of concrete Once the concrete has been proportioned to reduce the amount of heat produced, and if additional mix design adjustments do not reduce the maximum temperature, then cooling of the concrete prior to placement is necessary to achieve the desired concrete properties. One of the easiest and most cost effective ways to limit the placing temperature of concrete is to properly engineer the process for producing the concrete. There are several ways to limit the placement temperature of the concrete: selection of time and temperature of placement, use of chilled water replacement for mixing water, use of ice water for replacement of mixing water, cooling of coarse aggregates, and use of liquid nitrogen. Time and temperature of placement One of the cheapest ways to ensure cool concrete is to schedule the placement at the coolest part of the day, which is typically at night or early in the morning. The relative humidity and wind velocity may also need to be taken into account to prevent the concrete from dehydrating. The distance the concrete will be hauled should be kept as short as possible so as not to expose the concrete to additional heat and the truck staging time should be minimized to keep the concrete temperature low. Additionally, the concrete should be placed and finished as quickly as possible to minimize the prolonged exposure of concrete to the higher temperatures. Formulas given in ACI 305R-99 can be used to predict the concrete temperatures as expressed in equation 5.1 T=

0.22(TaWa + Tc Wc ) + Tw Ww + TaWwa 0.22(Wa + Wc ) + Ww + Wwa

where Ta Tc

= temperature of aggregate = temperature of cement

[5.1]


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Tw = temperature of batched mixing water from normal supply excluding ice Wa = dry mass of aggregate Wc = mass of cement Ww = mass of batched mixing water Wwa = mass of free and absorbed moisture in aggregate at Ta (Note: All temperatures are in ºC or ºF and all masses are in kg or lb) (ACI 305R-99). Concrete can expect to gain approximately 8°C [15°F] for each hundredweight of cement being used in a 0.75 m3 [1 yd3] if heat dissipation does not take place. If a concrete mix design were to contain 227 kg [500 lb] of cement and is placed at 32°C [90°F], then it can be anticipated that the concrete will increase in temperature to 82°C [180°F] within a day. With normal dissipation, the same 32°C [90°F] placed concrete could reach 60°C [140°F] before night. In the event the night time temperature dropped drastically, then it is likely that the exposed surface of the concrete could crack due to the differential temperature gradients. By reducing the initial placing temperature, the likelihood of cracking due to thermal effects is reduced (Scanlon 1997). Any large concrete placements in Florida will be placed in either the evening hours or early morning hours to reduce the concrete temperature as much as possible. This method can be very effective if the placement is not thicker than 1 m [3 ft], the materials are carefully selected, and the placement occurs just before sunrise.

Chilled water replacement for mixing water Concrete mixed with chilled water is often used to lower the temperature. The use of chilled batch water alone will lower the concrete temperature approximately 5°C [9°F]. Recently, manufacturers of heat pumps have been advertising equipment which can be used to cool the concrete as well. There are two types of heat pumps that are typically used to cool concrete: water to water and water to air. Water to water heat pumps are the most common and the most economical method used. In the event no groundwater is available, the water to air method can be used. Most heat pump modules can cool 75 m3 [100 yd3] of concrete in a twelve-hour period. If a higher rate is necessary, then additional modules are typically employed. Heat pumps which are used to cool the concrete for hot weather applications can also be used to heat concrete for cold weather applications. In Florida, this method of cooling concrete has not been used very often because of the initial capital costs and because typically the concrete temperature needs to be lowered at a higher rate than can be obtained with heat pumps.

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Ice water replacement of mixing water The most commonly used method for reducing the temperature of the concrete is to replace a portion of the mixing water with ice. This has two benefits: first, it lowers the mix water temperature and secondly, it lowers the mix temperature of the concrete by extracting heat during the phase change from ice to water. Ice can be used to substitute up to 80% of the batch water. Typically the amount of cooling that is achieved is limited to about 11°C [20°F]. Ice can be added directly into a ready-mix truck or premixed at the plant in large tanks with the mixing water (Sumodjo 2005). In order to ensure correct proportioning, the ice must be weighed and introduced into the mixing operation in finely graded sizes. This process typically requires a large capital investment due to the cost of ice, transportation, refrigerated storage, and handling. If blocks of ice are used, then a crusher/ slinger unit will also be required to finely crush a block of ice and blow it into the mixer. An alternative is to establish an ice plant near the concrete facility. Typical ice plants can produce 40 metric tons [44 tons] of ice in a 24-hour period. Formulas given in ACI 305R-99 can be used to predict the concrete temperatures as expressed in equations 5.2 and 5.3. With ice (SI units): T=

0.22(TaWa + Tc Wc ) + Tw Ww 0.22(Wa + Wc ) + Ww + Wi + Wwa TaWwa − Wi (79.6 − 0.5Ti ) + 0.22(Wa + Wc ) + Ww + Wi + Wwa

[5.2]

With ice (in.-lb units): T=

0.22(TaWa + Tc Wc ) + Tw Ww 0.22(Wa + Wc ) + Ww + Wi + Wwa 79.6Ww TaWwa − Wi (128 − 0.5Ti ) + 0.22(Wa + Wc ) + Ww + Wi + Wwa

[5.3]

where Ti = temperature of ice Wi = mass of ice (Note: The temperature of free and absorbed water on the aggregate is assumed to be the same temperature as the aggregate; all temperatures are in ºC or ºF) (ACI 305R-99). The use of ice to replace part of the mixing water is rather frequent. Some of the benefits of using ice to lower the placing temperature are: it can be purchased when needed, large quantities are available, and the amount of ice can be increased or decreased depending on the temperatures through-


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out the day. The ice can be added at a second staging area in the concrete production facility so as to not slow down the amount of concrete being delivered to a project site. Cooling of coarse aggregates Coarse aggregates typically make up 60 to 80% of a concrete mix design. Thus, the use of aggregates with a decreased temperature can have significant effects on the concrete placing temperature. Reducing the coarse aggregate temperature by 1°C [2°F] typically lowers the placing temperature of the concrete by 0.5°C [1°F]. Typical processes used to lower aggregate temperatures include: sprinkling, air blast or water chilled soaking. When smaller amounts of temperature reduction are desired, sprinkling the coarse aggregate with water can lower the concrete placing temperatures by 5°C [9°F]. This process relies on the evaporative cooling of the coarse aggregate utilizing only enough water to keep the stockpile wet, not saturated. Use of chilled water is unnecessary, as the heat loss is a result of evaporation. This is one of the more economical methods to cool concrete but has a limited amount of usage (Sumodjo 2005). Another method used for cooling the aggregates is by blowing air through the moist aggregates. The air flow can bring the aggregate to within 1°C [2°F] of the wet bulb temperature and will enhance evaporative cooling. The effectiveness of this method depends upon the ambient temperatures, relative humidity, and velocity of the air flow through the coarse aggregates. By adding a chilled air unit instead of air at ambient temperatures, the air cooling system can reduce the concrete placing temperature by 13.5°C [25°F]. The typical amount of coarse aggregates that can be air cooled is about 150 m3 [200 yd3] per hour. The main drawback is a relatively high installation cost compared to other methods (ACI 305R-99). If large temperature reductions are needed, then soaking the coarse aggregates in chilled water is required. The effectiveness of this method relies on the concrete facility being able to contain the coarse aggregate in a bin or silo so that the material is cooled in a short period of time. Care must be taken to ensure that all of the material is evenly inundated to provide a consistent slump of concrete from load to load. To obtain this consistency, modifications to the aggregate scooping device might be needed to allow the aggregate to drain properly. This process can produce concrete with as high as a 20°C [35°F] reduction in concrete placing temperatures. The process requires a large capital investment because dedicated bins or silos are required, as are installation of a chiller, and modifications to the scooping device. For a large project that will require a continuous amount of cooled concrete for placing in a hot weather environment, this procedure will typically be the most cost effective. This process was recently used at a concrete

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production facility supplying concrete to the Escambia Bay Bridges that were destroyed by Hurricane Ivan. The replacement bridges will take three years to complete and will mainly require hot weather concrete to be placed. Examples of the coarse aggregate soaking tanks and modified frontend loader are given in Figs 5.1 and 5.2. Other types of liquids and gases have been used on certain projects such as dry ice (solid CO2) to cool the aggregates. Dry ice is solid at temperatures of approximately −78.5°C [−110°F] and has a latent heat of 137 kcal/kg. At −20°C [−4°F] and with a pressure of 20 bars [2.0 MPa], dry ice is in a liquid state called liquefied carbonic acid gas (LCAG). LCAG is used to cool

5.1 Modified front-end loader.

5.2 Coarse aggregate soaking tanks.


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aggregates as it is sprayed onto the aggregate piles at a rate of 32 kg [70.5 lb] per minute. Dry ice is limited to cooling aggregate piles because adding the dry ice to the concrete would result in carbonation and loss of workability. After the dry ice has successfully cooled the aggregate stock piles, exhausting the carbonic acid gas from the mixer becomes critical prior to the addition of any other materials. Research has shown that the use of dry ice can result in decreasing the temperature of the concrete by up to 12°C [20°F] (Takeuchi et al. 1993). Liquid nitrogen for cooling concrete Liquid nitrogen has also been used as a method for cooling concrete for over twenty years. Liquid nitrogen (LN) is an inert cryogenic fluid with a temperature of −196°C [−320°F]. LN is injected directly into the batch water storage tank, aggregate, or mixer via lances to lower the temperature of the concrete as much as practical without freezing. LN can be stored at the batch plant or on the project site and if used on the project site itself, then repeated cooling of the concrete and greater control of the concrete temperature is possible. LN can be set up at a project or plant within a few days and can supplement other cooling methods to achieve a reduction in concrete temperature when necessary (Beaver 2004). LN is produced by compressing and cooling nitrogen gas to a point below its evaporation point of about −196°C [−320°F]. Typically, LN is injected directly into the mixer without making changes to the mixer itself. LN is injected into the mixer with a lance, which can move in and out of the ready-mix truck using a pneumatic cylinder. Typically, the lance can move vertically and horizontally, to allow the operator to position it correctly. The flow of LN through the lance is controlled by a pneumatically operated ball valve. The entire sequence of inserting the LN into the mixer is computer controlled to assist the operator in positioning the truck properly and when the cooling is complete. Typically, when the lance is inserted into the mixer it appears as though there is a release of toxic fumes, but in reality, this is just inert nitrogen and water vapor boiling off. The rotation of the mixing drum is essential when the LN is injected to the concrete to prevent nitrogen pools from accumulating within the mixer. In the event that too much LN is injected into the mixing drum, the mixing drum can become damaged or even ruptured. LN can cool concrete by more than 12°C [20°F] and has been used to obtain concrete with a placing temperature as low as 2°C [35°F]. The LN cooling process is flexible enough that it can be used for readjustment of the concrete in the field if certain processes are in place and the LN is being used at the project site. This method requires additional mixing time and the mixer is partially sealed to minimize the loss of coolant. The primary

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drawback to the use of LN for cooling concrete is the initial cost set up compared to other cooling methods. All material safety data sheets (MSDS) need to be read and understood by all personnel using LN (Sumodjo 2005). LN is not used very often and is typically reserved for use when all other options of cooling concrete have been exhausted. The Texas Department of Transportation has started using this technology with varying degrees of success.

5.2.3 Effects on plastic properties Once the concrete has been placed into the forms and is still in a plastic state, the high ambient temperature at the surface of the concrete can still cause problems. There are four main areas that have to be addressed to ensure that the concrete has adequate quality at early ages. These four areas include rate of slump loss, setting time of concrete, plastic shrinkage cracking, and proper curing. The stiffening of fresh concrete and the associated slump loss are mainly caused by the hydration of the cement. Evaporation or absorption of the mixing water into the aggregate may also reduce the amount of free water in the concrete, hindering the proper hydration of the cement. The hydration rate of cement increases with rise of temperature and generally follows the Arrhenius equation. When mixing concrete in hot weather, cement will hydrate at an accelerated rate, have a shorter setting time, and a higher rate of slump loss. The rate of slump loss is increased with a reduced amount of water available (Soroka and Ravina 1998). The increased rate of slump loss is typically resolved by adding mineral or chemical admixtures to the mix, and properly evaluating the mix design prior to placing it in the field with a trial batch. Decrease in time of set results in greater difficulty with transporting, handling, compacting, finishing, and additional need for cold joints. The initial setting time of the concrete is reduced due to increase of ambient temperature, decrease of ambient humidity, and increase of air/wind velocity. The same conditions also decrease the final setting time of the concrete (Ahmadi 2000). Most adverse field conditions can be overcome by making the following provisions: • • •

Placing concrete in the cooler part of the day and extending the time of setting in a condition that is workable. The use of a water fog spray nozzle to keep the air cool and the field humidity as high as possible (ACI 305R-99). Shading the concrete surface during exposure to solar radiation and high wind velocity can reduce the rate of evaporation on the concrete surface by 50% or more (Hasanain et al. 1989).


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Plastic shrinkage cracking typically occurs on the surface of freshly placed concrete when the surface stresses exceed the tensile strength of the concrete, and is caused by the evaporation of water from the surface of the concrete. Plastic shrinkage cracks can be identified as having a spacing of 0.3 to 1.0 m [1 to 3 ft] apart and normally do not extend to the free edge of the concrete. When the bleed water on the concrete surface evaporates, loss of water from beneath the concrete surface will occur and plastic shrinkage cracking takes place (Hasanain et al. 1989). ACI Committee 305 suggests an evaporation rate of 1.0 kg/m2/h [0.2 lb/ft2/h] or less so as not to exceed the amount of bleed water produced and thus induce plastic shrinkage cracking: h ER = ⎡t 2.5 × × a2.5 ⎤ × (1 + 0.4 × v) × 10 −6 [5.4] ⎣⎢ ⎦⎥ 100 where

(

ER t h a v

= = = = =

)

evaporation rate (lb/ft2/h) temperature of concrete (°F) ambient humidity (%) ambient air temperature (°F) wind velocity (mph)

Concrete mixtures incorporating slag, fly ash, silica fume, high cement contents, finely ground cements, high air contents, or super-plasticizers typically bleed at a rate within the limits recommended by ACI Committee 305. Therefore, some agencies have lowered the allowable evaporation rate for these types of mixes or require additional fogging systems to create additional humidity near the surface of the concrete to reduce the evaporation rate (Hover 2006). In some cases, revibrating before floating can close the plastic shrinkage cracking in large placements. When floating the surface, plastic shrinkage cracks can also be closed by striking the surface on each side of the crack before the concrete has reached final set (ACI 305R-99). The purpose of curing concrete is to maintain a proper amount of moisture within concrete during the early ages to develop the desirable properties at later ages. From the strength gain curve for concrete, it may be seen that the early age strength gains are much more rapid than at later ages (Fig. 5.3). As such, the concrete benefits from curing at early ages. Research has shown that concrete can lose 43% of its strength if not cured properly when placed in hot weather with a low humidity. The most effective curing operations are applied to the concrete as quickly as possible after initial set. Delaying the application of the curing material by as little as one day in hot weather can drastically reduce the later age properties of the concrete. For hot weather applications, it is recommended that the concrete be cured for at least seven days.

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Developments in the formulation and reinforcement of concrete Compressive strength gain curve compared to 360-day results 10000 Compressive strength (psi)

13% 7%

8000 22%

6000 4000 58%

2000 0 0

28

56

84 112 140 168 196 224 252 280 308 336 364 Time (days) Raw data

Predicted

5.3 Strength gain curve for concrete.

There are two types of curing process available for use in hot weather. The first is to apply water to the surface of the concrete through ponding, sprinkling, spraying, or saturated materials (burlap, rugs, sand, straw, or cotton mats). The second is to prevent excessive loss of water through impervious paper or plastic sheets, membrane-forming curing compounds or evaporation reducers (Al-Ani and Al-Zaiway 1988). A typical practice in Florida is to use a combination of these two methods such that membrane-forming curing compound is applied to the concrete until it has reached initial set. Once the initial set has been obtained, then a layer of burlap mats can be applied to insure no surface water is lost. Recently, there was a bridge deck in Florida where a contractor did not take the weather conditions into account. The concrete was designed with too short a set time and was not cured properly (see Figs 5.4 and 5.5). The bridge deck suffered severe cracking and had to be completely removed and replaced at the contractor’s cost.

5.2.4 Effects on hardened properties Despite the use of chemical and mineral admixtures, cooling the concrete and ensuring proper curing has taken place; concrete exposed to hot weather at early ages may still be prone to reduced performance. In the event that the long-term performance of hardened concrete properties is not taken into account, the structure may be subject to reduced life cycle or decreased load capacity. There are four main areas to be addressed with respect to hardened concrete at later ages: compressive strength, flexural/tensile strength, modulus of elasticity and durability.


129

5.4 Concrete bridge deck in Florida.

5.5 Severe cracking in the concrete bridge deck in Florida.

Owing to the environmental conditions associated with hot weather concreting, the amount of evaporation of water from the fresh concrete results in a lower effective water content. As previously stated, the lower effective water content usually results in a concrete that is less workable, and additional water may be needed on site for additional workability and to reduce slump loss. Upon the introduction of water to concrete on site, there is an increase in the water to binder ratio and capillary porosity. An estimate for the compressive strength of the concrete placed and cured in hot weather conditions, (fct)T, can be calculated from the following equation:

130

Developments in the formulation and reinforcement of concrete (fct)T = −14.15 + 7.06 lnT + (9.80 − 0.125T)lnT

[5.5]

where (fct)T = compressive strength of the concrete in MPa t = age of compressive strength testing in days T = temperature of the concrete in °C (Zivkovic 1992) This phenomenon, in which the increase in concrete placing temperature decreases the compressive strength of the concrete, has been confirmed in the field. The Tandy Center in downtown Fort Worth, Texas, showed a strength reduction of about 2.1 MPa [300 psi] for every 5.6°C [10°F] increase in placing temperature. The study involved samples which were cured in laboratory conditions with fairly uniform temperature and humidity after placement compared to samples cured in the field (Dodson and Rajagopalan 1979). Not only is compressive strength reduced with a rise in concrete temperature, but the modulus of rupture and splitting tensile properties are also reduced due to an increase in placing temperature. The reduction in modulus of rupture and splitting tensile strength is typically attributed to lower effective water, higher water to binder ratio, and improper curing. The modulus of rupture of concrete placed and cured in hot weather conditions, (fr)T, can be calculated from the following equation: (fr)T = (1.24 − 0.01Tc fr ≤ 0.935fr

24˚C ≤ Tc ≤ 45˚C

[5.6]

where fr = modulus of rupture of concrete placed and cured at approximately 24°C [75°F] Tc = temperature of concrete placed and cured in hot weather conditions. The splitting tensile strength of concrete placed and cured in hot weather conditions, (fst)T, can be calculated from the following equation: (fst)T = (1.132 − 0.00552Tc)fst ≤ 0.955fst

24˚C ≤ Tc ≤ 45˚C

[5.7]

where fst = splitting tensile of concrete placed and cured at approximately 24°C [75°F] Tc = temperature of concrete placed and cured in hot weather conditions (Abbasi and Al-Tayyib 1990). The modulus of elasticity can be reduced by as much as 17.5% when the concrete is prepared and cured at 45°C [115°F]. The modulus of elasticity of concrete placed and cured in hot weather conditions, (Ec)T, can be calculated from the following equation:

(Ec)T = (1.4775 − 0.0145Tc)Ec


131

35˚C ≤ Tc ≤ 45˚C

[5.8]

where Ec = modulus of elasticity of concrete placed and cured at approximately 26°C [78°F] Tc = temperature of concrete placed and cured in hot weather conditions (Abbasi and Al-Tayyib 1990). Durability of hardened concrete is affected by hot weather conditions and can drastically reduce the service life of a structure. One of the main causes of deterioration of a structure is corrosion of steel reinforcement especially in aggressive environments. The corrosion of steel due to salt exposure is found to increase drastically with higher humidity. Areas that are exposed to salt water environments are at a higher risk for corrosion than dry areas, even with both exposed to hot weather conditions. The diffusion coefficient of reinforced concrete exposed to salt water is increased as compared to airborne salts. This increase in diffusion coefficient is mainly due to the increase in capillary porosity, plastic shrinkage cracking, and higher water to binder ratio. The service life of a concrete structure exposed to hot weather can be extended with lower water to binder ratio and increase in the cover of reinforcing steel (Morinaga 1992).

5.3

Future trends

5.3.1 Initiation of cracking One of the largest problems that affects the durability of concrete is cracking of the concrete. Currently, considerable research is being devoted to the prediction of early age cracking in concrete. Some of these techniques use completely saturated samples, other testing techniques use sealed samples, and still other tests are being developed for the surface tension of the concrete. Regardless of the test type, the environmental exposure conditions of the in-situ concrete should be accounted for prior to development. Currently, there is a lot of promising ongoing research focusing on the prediction of plastic shrinkage in concrete materials (Sant et al. 2006, Schindler 2004).

5.3.2 Concrete modeling Modeling of concrete properties is becoming a key area for future research. Some of the better known models are Stadium® developed by Materials Service Life, HYPERCON developed by the National Institute of

132


Standards and Technology (NIST) and LIFE-365 developed by E.C. Bentz and M.D.A. Thomas. There are several advantages and disadvantages with any type of modeling software package. The main advantage of a modeling software package is that it can give the user predicted performance characteristics in a short amount of time and without extensive laboratory work. The major disadvantage is that they are often misused and based on assumptions not always associated with real world events. As technology advances and more data is inputted into models in a holistic approach, these modeling software packages should become more valuable and could be used for predicting the service life of a structure. One future initiative of predictive modeling could be used to evaluate a mix design for the appropriate materials and the best time to place the concrete.

5.3.3 Nondestructive evaluation techniques Recent advancements in evaluating the data collected from nondestructive techniques is starting to show some real promise for examining an existing structure. Some of the more advanced research has been coming out of the Federal Institute for Materials Research and Testing (BAM) in Berlin, Germany. One of the most promising current research initiatives has been the development of data fusion techniques based on information from different types of nondestructive tests applied to concrete structures. Perhaps a future initiative might incorporate the use of nondestructive techniques on freshly placed concrete to predict consolidation, potential for plastic shrinkage cracking, cover of steel reinforcement and density of concrete.

5.3.4 Materials specialty engineers In recent years, there has been an initiative taken by the concrete industry to incorporate the relatively large amount of information regarding cement and concrete material properties being generated on a daily basis. New tests are being developed which aid end users in determining concrete properties, new models are being developed to predict service life, and new techniques are being evaluated to determine in-situ properties. What is also becoming apparent is that without proper guidance from an individual who knows the limitations of different tests, models and evaluation techniques, a poor quality concrete can still be placed without anyone’s knowledge until the structure starts to fail. There has been more demand for individuals who specialize in the understanding of local materials to produce a structure that will meet the owner’s performance requirements. Without these


133

individuals, it will become increasingly difficult to obtain the performance requirements desired by the owner of the structure.

5.4


5.4.1 American Concrete Institute One of the preferred resources for information in North America is the American Concrete Institute. ACI produces different types of documents for mix design calculations, placing, curing, and testing of concrete. One group in particular that specializes in hot weather concreting is Committee 305 (Hot Weather Concrete), and is currently chaired by James Cornell.

5.4.2 Japan Concrete Institute One of the preferred resources for information in Asia is the Japan Concrete Institute. JCI has several research committees investigating issues pertaining to setting of concrete, material durability, and cracking of concrete. One group in particular that looked at concrete under hot weather conditions is Committee JCI-TC-033A (Technical Committee on Performance Evaluation of Concrete under Natural Weathering Conditions) and was chaired by Noboru Saeki. Another group that is looking at the time-dependent behavior of concrete is Committee JJCI-TC061A (Technical Committee on timedependent behavior of cement-based materials) and is chaired by Tada-aki Tanabe.

5.4.3 RILEM One of the preferred resources for information in Europe is the International Union of Laboratories and Experts in Construction Materials, Systems and Structures (RILEM). RILEM has several different technical committees researching current issues to advance the materials used for construction. One group that is looking at early age cracking of concrete is Committee 195-DTD (Recommendation for test methods for autogenous deformation and thermal dilation of early age concrete) and is chaired by Tor Ame Hammer. Another group that is looking at concrete at high temperatures is Committee 200-HTC (Mechanical concrete properties at high temperature – Modelling and applications) and is chaired by Ulrich Schneider. One last group that is also looking at the durability of concrete cracking is Committee CCD (Concrete cracking and its relation to durability: Integrating material properties with structural performance) and is chaired by Jason Weiss.

134


5.5

References

AASHTO M 85: Standard Specification for Portland Cement, American Association of State Highway and Transportation Officials (2007). Abbasi A F, Al-Tayyib A J (1990), ‘Effect of hot weather on pulse velocity and modulus of elasticity of concrete.’ Materials and Structures, Vol. 23, Number 5, 334–340. ACI Committee 305 (1999), ‘Hot Weather Concreting (ACI 305R-99),’ American Concrete Institute, Farmington Hills, Michigan, 20 pages. Ahmadi B H (2000), ‘Initial and final setting time of concrete in hot weather.’ Materials and Structures, Vol. 33, Number 8, 511–514. Al-Amoudi O S B, Maslehuddin M, Abiola T O (2006), ‘Effect of type and dosage of silica fume on plastic shrinkage in concrete to hot weather.’ Construction and Building Materials, Vol. 18, Number 10, 737–743. Al-Ani S H, Al-Zaiway M A K (1988), ‘The effect of curing period and curing delay on concrete in hot weather.’ Materials and Structures, Vol. 21, Number 3, May, 205–212. ASTM C1157-03: Standard Performance Specification for Hydraulic Cement, ASTM International (2003). ASTM C494/C494M-05a: Standard Specification for Chemical Admixtures for Concrete, ASTM International (2005). ASTM C618-05: Standard Specification for Coal Fly Ash and Raw or Calcined Natural Pozzolan for Use in Concrete, ASTM International (2005). ASTM C1240-05: Standard Specification for Silica Fume Used in Cementitious Mixtures, ASTM International (2005). ASTM C989-06: Standard Specification for Ground Granulated Blast-Furnace Slag for Use in Concrete and Mortars, ASTM International (2006). ASTM C150-07: Standard Specification for Portland Cement, ASTM International (2007). ASTM C192/C192M-07: Standard Practice for Making and Curing Concrete Test Specimens in the Laboratory, ASTM International (2007). ASTM C595-07: Standard Specification for Blended Hydraulic Cements, ASTM International (2007). ASTM C1017/C1017M-07: Standard Specification for Chemical Admixtures for Use in Producing Flowing Concrete, ASTM International (2007). Austin S A, Robins P J (1992), ‘Performance of Slag Concrete in Hot Climates’, in Walker M J, Concrete in Hot Climates, London, E & F N Spon, 129– 139. Beaver W (2004), ‘Liquid Nitrogen for Concrete Cooling.’ Concrete International, Vol. 26, Issue 9, 93–95. Dodson C J, Rajagopalan K S (1979), ‘Field Tests Verify Temperature Effects on Concrete Strength.’ Concrete International, Vol. 1, Issue 12, 26–30. Duggan C R, Scott J F (1989), ‘Alternative cement for hot climates.’ Concrete. Journal of the Concrete Society, London, Vol. 20, Number 2, 18–20. ENV197-1: Cement. Composition, specifications and conformity criteria for low heat common cements, European Standards (2000). FDOT – Florida Department of Transportation (2007), Standard Specification for Road and Bridge Construction.Tallahassee, Florida, 926 pages.


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Hasanain G S, Khallaf T A, Mahmood K (1989), ‘Water Evaporation from Freshly Placed Concrete Surfaces in Hot Weather.’ Cement and Concrete Research, Vol. 19, Number 3, 465–475. Hover K C (2006), ‘Evaporation of Water from Concrete Surfaces.’ ACI Materials Journal, Vol. 103, Number 5, September–October, 384–389. Morinaga S (1992), ‘Life prediction of reinforced concrete structures in hot and salt-laden environments’, in Walker M J, Concrete in Hot Climates, London, E & F N Spon, 129–139. Owens P L (1992), ‘The Selection of Hydraulic Cements to Satisfy the Requirements for Concrete Construction in Hot Climates’, in Walker M J, Concrete in Hot Climates, London, E & F N Spon, 187–197. Ravina D, Soroka I (2002), ‘Admixture Effects on Hot-Weather Concrete’ Concrete International, Vol. 24, Issue 5, 34–38. Sant G, Lura P, Weiss J (2006), ‘Measurement of Volume Change in Cementitious Materials at Early Ages: Review of Testing Protocols and Interpretation of Results.’ 85th Annual Transportation Research Board Meeting, January 22–26, Washington, D.C., 19 pages. Scanlon J (1997), ‘Controlling Concrete During Hot and Cold Weather’ Concrete International, Vol. 19, Issue 6, June, 52–58. Schindler, A K (2004), ‘Effect of Temperature on the Hydration of Cementitious Materials.’ ACI Materials Journal, Vol. 101, Number 1, 72–81. Soroka I, Ravina D (1998), ‘Hot Weather Concreting with Admixtures.’ Cement and Concrete Composites, Vol. 20, Issue 2–3, 129–136 Sumodjo F (2005), ‘Precooling Mass Concrete: Liquid Nitrogen Proved to be the Best Alternative on the San Francisco-Oakland Bay Bridge.’ Concrete Construction, Vol. 50, Number 8, Aug., 36–40. Takeuchi H, Tsuji Y, Nanni A (1993), ‘Concrete Precooling Method by Means of Dry Ice’ Concrete International, Vol. 15, Issue 11, Nov., 52–56. Zivkovic S D (1992), ‘The Effect of Increase Temperature on Fresh and Hardened Concrete’, in Walker M J, Concrete in Hot Climates, London, E & F N Spon, 129–139.

6 Underwater concrete A K AL-TAMIMI, College of Engineering, The American University of Sharjah, United Arab Emirates

6.1

Introduction

Underwater concrete “UWC” is one special type of high performance concrete used in the past, present, and in the foreseeable future as long as there is need to construct bridges, with foundations in soil with high water levels, and almost all off- and on-shore structures. The term high performance concrete refers to concrete that performs particularly well in at least three key performance indicators: strength, workability, and service life. Therefore, underwater concrete should meet these performance criteria and it remains a viable and economic choice for consultants and contractors. UWC requires special and careful monitoring during all stages of construction; i.e. special considerations for selecting the right materials, specialized apparatus for the quality control, design, and methods of construction. Underwater concrete was specially designed to enhance constructability and performance in water environments. Using the underwater concrete technique may avoid engineers using the “old-style” of construction by isolating the water, and therefore, minimize interruption to plant operation and result in high savings. “UWC” is a highly flowable concrete that can spread into place under its own weight and achieve good compaction in the absence of vibration, without exhibiting defects due to segregation and bleeding. Underwater concrete technology has developed dramatically in recent years, so that the mix can be proportioned to ensure high fluidity as well as high resistance of washout and segregation. The construction of a wide range of structures including bridge piers, harbors, sea and river defences over many decades, and the development of offshore oilfields, has required placement of concrete underwater. This process can be successfully carried out and sound, high-quality concrete can be produced if sufficient attention is paid to the concrete mix design and the production method applied. 136

Underwater concrete

137

The stability of fresh concrete depends on the rheological properties and placement conditions. It can be characterized by the concrete resistance to washout, segregation and bleeding and is affected by the mix proportioning, aggregate shape and gradation, admixtures, vibration and placement conditions. The differential velocity at the interface between the freshly cast concrete and surrounding water can erode some cement and other fines. Such erosion can increase the turbidity and contamination of the surrounding water, and impair strength and durability, as well as bond to reinforcement steel and existing surfaces. The improvement of the in-situ properties of underwater concrete is related to the enhancement in washout resistance [1–3]. A superplasticizer (SP) is used to ensure high fluidity and reduce the water/powder ratio (W/ P). An anti-washout admixture (AWA) is incorporated to enhance the yield value and viscosity of the mix and hence the washout resistance and segregation resistance [1, 4]. The majority of AWAs are water-soluble polymers that increase the yield value and viscosity of cement paste and concrete [4, 5]. A statistical design approach was used to establish statistical models and to provide an efficient means of evaluating the influence of key mix variables on the fresh and hardened concrete characteristics that affect the performance of underwater concrete [6, 7]. The derived models include mixes with 380 to 600 kg/m3 of powder, W/P ratios of 0.34 to 0.46, sand/ aggregate ratios of 0.42 to 0.50, as well as AWA and SP dosages varying between 0.005% and 0.265% and 0.05% and 2.65%, respectively. The slump flow and the washout resistance are influenced, in order of importance, by the concentrations of AWA and cement, then by the water/cement ratio and dosage of SP, and various combinations of these parameters. The sand/ aggregate ratio had a secondary effect on these properties [6, 7].

6.2

Development of underwater concrete

Underwater concrete (UWC) continues to flow with time under its own weight before it starts to harden, unlike ordinary fresh concrete, which usually assumes its stable shape very rapidly. The commonly used standard tests for workability of ordinary fresh concrete, such as the slump test or the flow (spread) test, are inadequate. Flowing concrete usually drops to one-third or less of the original height in the slump test, and the result is a collapsed heap, the height of which is likely to be determined by the angle of repose of the largest particles. Therefore, visual measurement of collapsed concrete, or simply the height of the slumped concrete, does not differentiate the characteristics of two cohesive concretes. Moreover, even if a measurable slump is obtained, the sample will continue to settle and show increasing slump with time. The DIN flow table, which had been developed in Germany as a workability test for ordinary concretes and

138


adopted as the British Standard test (BS 1881:1984), is no more satisfactory than the slump test. The test is intended to measure a bulk property of the concrete, but the end-point condition for the flowing concrete (510 mm spread) can only be achieved by assuming that the concrete spreads into a disc of 21 mm thickness, equal to the size of the largest particles, and clearly not representative of the bulk. The test is also operator-sensitive (manual jolting of the base plate, perfectly level position). Many other tests have thus been developed to assess UWC, and they will be described briefly in the following section.

6.3

Quality control of underwater concrete

6.3.1 Flow/spread test There have been several tests proposed based on this principle. The version of the test described here was originally developed in Germany by Graf [8] in the 1930s (Fig. 6.1). The test measures the spread of a sample of fresh concrete after it has been molded into the shape of a truncated cone and allowed to slump following the removal of the mold. The slumped concrete is then subjected to a controlled amount of jolting. The term “spread” test appears to be more appropriate than “flow” in order to avoid confusion with other “flow” tests. The “spread” describes much better the principle of the test in which the sample spreads in all directions. Flow of concrete tends to imply moving or “flowing” in one direction, restrained within a container or pipe. The test had originally been aimed at the assessment of workability of medium range concrete mixes and remains in use for such purposes in

130

700

6.1 Spread/flow table test apparatus [8].

200

40

700

200

Underwater concrete

139

several European countries. It is widely used in Germany, its country of origin. The test can also be used for fresh mixes of high and very high workability, where collapsed slumps are recorded. This capability has increased the use of the test for assessment of superplasticized and other special flowing fresh mixes. The apparatus consists of a flat, square (700 mm × 700 mm) plywood top plate, which has its upper surface lined with a metal sheet at least 1.5 mm thick. Center-lines at 90 degrees are engraved on to the surface of the metal lining together with a concentric circle of 200 mm in diameter. The mass of the top plate should be within 16 kg ± 1 kg. The top plate is attached to a bottom plate by hinges along one side. The top plate is fitted with a handle at the center of the edge opposite to the hinged side. The handle is used for lifting the top plate; however, the height of the lift is restricted to 40 mm ± 1 mm by metal retainers. The bottom plate extends forward by at least 120 mm along the side with the handle to provide a foothold. Spread values in the range of 450 to 600 mm (18 to 24 in.) were recommended for underwater concrete used in drilled shaft construction [9] and values of 550 to 650 mm (21 to 26 in.) were reported in underwater concrete repair [10].

6.3.2 The Orimet test The Orimet was developed by Bartos [11, 12] specifically as a method for a rapid assessment of very highly workable, flowing fresh concrete mixes on construction sites. The test is based on the principle of an orifice rheometer which is applied to fresh concrete. The Orimet test is applicable to fresh concrete mixes of very high workability, preferably mixes for which the result of the slump test is greater than 150 mm or which record a collapse slump. The test is used for specifications of workability (mobility) of fresh concrete mixes, for the compliance with specifications and for a rapid check of adjustments of mix proportions/admixtures on construction sites where very high workability of a fresh mix has to be maintained. It is particularly suitable for superplasticized and other flowing mixes. The Orimet consists of a vertical casting pipe fitted with an interchangeable orifice at its lower end. A quick-release trap door is used to close the orifice. The basic Orimet is provided with an orifice having an 80 mm internal diameter which is appropriate for assessment of concrete mixes of aggregate size not exceeding 20 mm. Depending on the composition of the mix and the workability required, orifices of other sizes, usually from 70 mm to 90 mm in diameter can be fitted instead. The casting pipe, the orifice and the trap door mechanism are supported by an integral tripod which folds back to facilitate transport (Fig. 6.2). A sample of at least 7.5 liters of fresh mix is required.


60

1090

600

980

140

6.2 Orimet tester: basic dimensions [11].

The same sample can be re-tested rapidly. Normal assessment requires at least two, preferably three samples to be tested. The Orimet test includes two stages: 1) Go or not Go for underwater concrete mixes, and 2) Recommended value of 3–5 seconds for good underwater concrete [12].

6.3.3 The washout-resistance test The washout-resistance test was developed at the University of Paisley [13, 14], particularly as a method for the assessment of the non-dispersability of fresh concrete placed underwater. The test is based on the principle of evaluating non-dispersability by direct contact of fresh concrete with water (Fig. 6.3). The washout-resistance test is applicable to fresh concrete mixes of any level of workability conditions, to evaluate their suitability for an underwater application. The test assembly consists of a barrel containing water (30 l) with a pipe and a spray head connected at the bottom. The test sample is placed on a

Underwater concrete

141

6.3 Washout test [15].

frame, which is freely suspended on an electronic balance. The balance is supported on a bench. The test is simple. A sample of a concrete mix is put into a mold on the plate. The mold is removed and the plate with the sample is placed on the frame suspended from the balance. The tap on the pipe connected to water tank is turned on. Water from the spray head washes out the sample until the tank is empty. A computer connected to the electronic balance records the whole washout process. The test produces diagrams showing the loss of mass during the test. The measurement recorded directly from the balance at any moment during the test is the mass of the sample and of the plate resting on it plus the mass and pressure of water which is poured on the sample. The net amount of lost material is the direct measurement from the balance minus the effect of the pressure and weight of the poured water. The end result is the loss of material expressed as the percentage loss of the original sample, accompanied by graphs of the washout mass during the test, and visual assessment of the sample after test. The variability of the washout results inherently increases when a greater mass has been washed out from the original sample. However, the test is able to recognize clearly the dosage of underwater admixtures and the suitability of fresh concrete to be placed underwater. The advantages of this test are: • • • •

good simulation of concrete-water interaction in practice a simple test procedure highly sensitive to washout resistance acceptable mass of the test samples (1 kg).

However, the apparatus is not designed for use on site. The apparatus is expected to be used by concrete laboratories and companies producing underwater concrete admixtures. Washout values of 1% to 6.6% are recommended for underwater concrete [15].

142


6.3.4 The plunge test The plunge test originates from Belgium where it was used by the University of Ghent for an indication of how the fresh concrete mix will perform when either dropped through water or subjected to flow of water over it. The plunge test is applicable to fresh concrete mixes from low to medium workability. A high workability mix will tend to flow out from the wire basket. The plunge test (Fig. 6.4) consists of a vertical transparent plastic pipe embedded at the bottom into a concrete-filled container to provide a seal around the pipe and to form a stable base (Fig. 6.4(a)). A test sample is placed in a wire basket (Fig. 6.4(b)), which is plunged during the test through water in the plastic pipe (Fig. 6.4(c)). The actual test arrangement and the size of the equipment has varied from one research center to another [16]. A sample of 4–30 kg is required. The sample cannot be retested. The test requires at least two samples to be tested. Plunge test case study at the American University of Sharjah – 2007 The loss in mass of the concrete in the receiving container is equal to M1 − M2 (M1 – mass of concrete before test; M2 – mass of concrete after test). Washout or loss of mass of sample, expressed as a percentage of the initial mass of the sample is given by the following formula: D=

M1 − M 2 × 100 M1

(where D = washout, %)

Major problems with this test are: • •

•

Variation of the rate of fall of the basket full of concrete through the water. Variation of the size of the basket and the diameter of the holes. Also the wire basket retains the sample, therefore the test can show a lesser “washout” than in practice. Fresh concrete placed underwater is expected to have very high workability. In this case parts of the mix will often flow out from the basket without any washout and the measurement of the washout becomes unreliable.

Using the plunge test, washout results in the range between 2% and 3% were found acceptable to produce underwater concrete [17].

6.3.5 The filling ability test The test has been developed by the Technology Research Center of the Taisei Corporation in Japan [18]. The test is used to measure the filling ability of a new type of concrete called Self Compacting Concrete (SCC).

Underwater concrete Plastic pipe 200 mm OD × 3 mm wall Supporting wire

20

Steel reinforcing collar 20 mm wide × 3 mm thick

120 150 dia. Expanded metal wire mesh ‘Expamet mesh hex 2’

1.7 m

315 dia. 20 L

Plastic container

320

(a)

280 dia.

Dense impervious concrete to provide seal around pipe and to form stable base

(b)

(c)

6.4 Plunge test examples [16].

143

144

Developments in the formulation and reinforcement of concrete Middle wall

2400Pa

Sliding door

45 cm Reinforcing bars (D13 mm)

Concrete 59 cm

R1 R2

14 cm 14 cm

Filling height

14 cm

28 cm

4@5 cm=20 cm

6.5 Apparatus for evaluating filling ability [18].

SCC has excellent deformability and high resistance to segregation, meaning that it can be placed in heavily reinforced formwork without the need for vibrators. The test is suitable for high workability mixes such as SCC. The apparatus consists of a vessel, which is divided by a middle wall into two rooms, shown by R1 and R2 in Fig. 6.5. At the bottom of the wall is an opening, which is controlled by means of a sliding door. Deformed reinforcing bars with nominal diameters of 13 mm are installed at the opening with center-to-center spacing of 50 mm – creating a clear spacing of 35 mm between the bars. To conduct the test, the concrete sample is placed in R1. A small amount of pressure – about 2400 Pa – is applied to the concrete to assist its flowing ability. The sliding door is then moved upwards to allow the concrete sample to flow through the opening and then through and around the reinforcing bars and fill section R2 of the vessel. The filling height of the concrete in R2 is then measured and recorded as the filling ability of the mix. A difference of 0–30 mm between R1 and R2 is recommended to produce typical good quality underwater concrete [19].

6.4

Application/case study

6.4.1 Design of UWC mixtures [20] UWC mixtures were designed using a Portland cement, and coarse aggregate consisting of round natural quartz and sandstone particles with a nominal aggregate size of 20 mm. Well-graded quartzite sand with a finesse modulus of 2.74 was also employed. The relative density values of the coarse aggregate and sand were 2.50 and 2.56, respectively, and their absorption rates were 1.7% and 1%, respectively.

Underwater concrete

145

A new generation copolymer-based SP was used which has a solid content and specific gravity of 30% and 1.11, respectively. This SP was developed for self-compacting concrete. The SP was used at dosages varying from 0.2 to 2.1%, by mass of cement. Welan gum was selected as the AWA. Welan gum is a high molecular-weight, water-soluble polysaccharide obtained through a controlled microbial fermentation [5]. It is used to increase the viscosity of mixing water, and hence that of the cement paste. The powderbased welan gum was mixed with part of the mixing water, 10% solution, using a high-shear mixer. This was done to prevent the AWA from continuing its hydration during mixing and agitation. All mixes were prepared in 25-liter batches and mixed in a drum mixer. The mixing sequence consisted of homogenizing the sand and coarse aggregate for 30 s, then adding 50% of the mixing water in 15 s. After mixing for 2–3 min, the mixer was stopped for 5 min while the contents were covered. The cement was then added along with the remaining solution of water and SP. The AWA was added last. The concrete was then mixed for a further 3 min. The workability of the concrete was evaluated using the slump test. Because of the viscous nature of concrete containing an AWA, the readings of the measurement were delayed for one minute following the removal of the slump cone. The test consisted of determining the mass loss of a fresh concrete sample weighing 2.0 ± 0.2 kg which was placed in a perforated basket and allowed to freely fall three times through a 1.7 m-high column of water [21]. The Orimet test described earlier was used for determination of the flow time of a fresh concrete mix [14]. The Orimet was provided with an orifice which reduced the internal diameter from 120 mm within the casting pipe to 90 mm at the end of the orifice. A sample of at least 7.5 l of fresh mix was used. The flow time was measured from the time at which the trap door was opened until the flow of the concrete from the orifice was finished. Three measurements of the time flow were determined. The study was concentrated on determining the effects of SP, AWA dosage and cement content on the slump, flow time and the washout loss and also the degree of variation between washout loss and slump. As summarized in Table 6.1, the investigated mixtures were prepared with a fixed W/CM of 0.43 corresponding to high-quality underwater concrete. These mixtures were made with 100% Portland cement which varied from 420 to 520 kg/m3. The concentrations of AWA were 0.02% and 0.13%, by mass of cementitious materials, corresponding to relatively low and medium dosages used in underwater concrete. The dosage of SP used varied from 0.40 to 1.8%. Results of the fresh properties of the 16 mixes used in this investigation are given in Table 6.1. Mixes 1-A, 1-B, 1-C and 1-D are replicates to determine the variability of slump, flow time and washout loss obtained by the plunge test. This mix was made with 0.43 W/CM, 470 kg/m3 of cement and

0.075

0.075

190 9.3 3.3

1.10

1.10

190 9.5 3.0

955

955

Slump (mm) Flow time (s) Washout after 3 drops (%)

0.43 470 202 695

0.43 470 202 695

W/C Cement (kg/m3) Water (kg/m3) Fine aggregate (kg/m3) Coarse aggregate (kg/m3) Superplasticizer (%) Welan gum (% of cement mass)

1-B

1-A

Mix

190 9.5 3.0

0.075

1.10

955

0.43 470 202 695

1-C

180 9.0 3.1

0.075

1.10

955

0.43 470 202 695

1-D

40 none 5.6

0.02

0.40

1012

0.43 420 181 736

2

Table 6.1 Mix proportioning and test results of mixes

0 none 3.6

0.13

0.40

1012

0.43 420 181 736

3

0.02

0.2

898

0.43 520 224 653

5

180 110 1.47 6.0 10.5 8.7

0.02

0.40

898

0.43 520 224 653

4

90 10 2.0

0.13

0.40

898

0.43 520 224 653

6

190 14 6.6

0.02

1.80

1012

0.43 420 181 736

7

240 10 7.8

0.02

2.1

1012

0.43 420 181 736

8

80 none 4.2

0.13

1.80

1012

0.43 420 181 736

9

190 8 5.6

0.43 420 181 736 1012 2.1 0.13

10

265 2.9 11.2

0.43 520 224 653 898 1.80 0.02

11

220 2.4 4.7

0.43 520 224 653 898 1.80 0.13

12

260 2.0 6.5

0.43 520 224 653 898 2.1 0.13

13

Underwater concrete

147

Table 6.2 Repeatability of test parameters

Mean (N = 4) Coefficient of variation Estimate error (95% confidence limit) Relative error

300

Slump

Flow time

Washout loss

188 mm 2.7% 5 mm

9.3 s 2.5% 0.24 s

3.1% 4.6% 0.14%

2.7%

2.6%

4.5%

SP = 1.8%

420 kg/m3

Slump (mm)

250 200

520 kg/m3 SP = 1.8%

SP = 0.4%

150 SP = 0.4%

100 50

0 slump 0

AWA = 0.02%

AWA = 0.13%

6.6 Variations of slump with concentrations of SP and AWA, and cement content.

contained 1.1% and 0.075% of SP and AWA, respectively. Table 6.2 shows the mean measured results of the four replicate mixes, coefficients of variations, as well as the standard errors with 95% confidence limit for each of the three measured properties. The relative errors for slump and flow time are lower and shown to be limited to approximately 2.7%. On the other hand, the relative error for washout loss was 4.5% indicating the greater degree of experimental error for the washout resistance test. Figure 6.6 illustrates the effect of increasing SP and AWA concentrations and the dosage of cement on the slump. For any given dosage of AWA, an increase in the dosage of SP or cement content resulted in a substantial increase in workability. For example, with 0.02% of welan gum, concrete made with 420 kg/m3 of cement exhibited an increase of slump from 40 mm to 190 mm when the SP dosage was increased from 0.4 to 1.8%. The increase of AWA for a fixed dosage of SP and cement content resulted in a drop in slump. For example, for the mixture made with 520 kg/m3 of cement and containing 0.4% of SP, the increase in AWA dosage from 0.02% to 0.13% led to a reduction of slump from 180 mm to 90 mm. For all mixtures, an increase in cement content resulted in an improvement in workability. For a constant dosage of 0.02% of AWA and 1.8% of SP, the slump increased

148


from 190 mm to 265 mm when the cement content increased from 420 kg/m3 to 520 kg/m3. The washout loss is affected by the concentration of AWA and SP, and the cement content. Figure 6.7 presents the variation of washout loss with different dosages of AWA, SP and different cement contents. For any given concrete, the increase in SP dosage and cement content increased the washout mass loss. In fact, the increase in fluidity due to higher additions of SP resulted in an increase in washout mass loss regardless of the AWA content. However, an increase in AWA concentration resulted in a reduction of washout mass loss. The improved resistance to water dilution of concrete containing AWA is due in part to the ability of the polymer to retain some of the mixing water. The AWA polymers also become adsorbed onto cement grains along with imbibing and fixing part of the mixing water, resulting in further retention of suspended cement particles. For example, for mixtures made with 0.02% welan gum and 420 and 520 kg/m3 of cement, the increase of SP from 0.4% to 1.8% led to increases in washout mass loss of 18% and 7%, respectively. For fixed dosages of SP of 0.4 and 1.8%, the increase of cement content from 420 to 520 kg/m3 resulted in increases in washout mass loss of 88% and 70%, respectively. The effect of increasing SP, AWA and cement content on flow time is shown in Fig. 6.8. An increase in SP resulted in a reduction of flow time for fixed contents of cement and SP. However, an increase in AWA concentration led to an increase of flow time for mixes made with constant SP and cement contents. In fact, for a given concentration of SP and cement content, the increase in AWA content increases the viscosity of concrete and therefore the flow time increased. For example, for the concrete made with 520 kg/m3 and containing 0.13% of AWA, the increase of SP from 0.4% to 1.8% resulted in significant reduction of flow time (from 10 s to 2.4 s). For the mixture containing 0.02% of AWA and 1.8% of SP, increas-

Washout mass loss (%)

14 12

420 kg/m3 520 kg/m3

SP = 1.8% SP = 0.4%

10 8 6

SP = 1.8% SP = 0.4%

4 2 0 AWA = 0.02%

AWA = 0.13%

6.7 Variations of washout loss with concentrations of SP and AWA and cement content.

Underwater concrete

149

20

Flow time (s)

420 kg/m3 15

520 kg/m3

SP = 1.8% SP = 0.4%

10 5

SP = 1.8%

SP = 1.8%

SP = 0.4% 0 AWA = 0.02%

AWA = 0.13%

6.8 Variations of flow time with concentrations of SP and AWA and cement.

Washout loss (%)

12

C = 420 kg/m3 AWA = 0.02% C = 520 kg/m3 8 Increasing AWA = 0.02% SP 6

10

4 AWA = 0.13%

2

AWA = 0.13%

0 0

50

100

150 200 Slump (mm)

250

300

6.9 Changes in washout loss vs. slump for concretes made with different AWA concentrations and cement content.

ing the cement content from 420 to 520 kg/m3 resulted in a substantial reduction in flow time (14 s to 2.9 s). Figure 6.9 presents the relationship between washout loss and slump of various mixtures containing 0.02% and 0.13% welan gum and 420 and 520 kg/m3 of cement. For a given concrete, an increase in slump is shown to increase washout loss. With 180 mm of slump, the lower washout mass loss (3.8%) occurred with concrete made with a high dosage of AWA (0.13%) and 520 kg/m3 of cement. For similar slump and 0.13% of AWA, the concrete made with 420 kg/m3 cement exhibited more washout loss (5.6%). However, the decrease of AWA content from 0.13% to 0.02% of concrete made with 520 kg/m3 resulted in increase in washout loss from 3.8% to 10.5% for the same slump. The washout mass loss can decrease with an increase of AWA concentration, despite the additional SP content required to maintain a given fluidity. The relationship between slump and flow time is presented in Fig. 6.10. The increase in slump resulted in a reduction in flow time. Concrete made

150

Developments in the formulation and reinforcement of concrete 70

C = 420 kg/m3 C = 520 kg/m3

Flow time (s)

60 50

Increasing SP

40

AWA = 0.13%

30 20

AWA = 0.13%

AWA = 0.02%

10 AWA = 0.02%

0 0

50

100

150 200 Slump (mm)

250

300

6.10 Changes in flow time vs. slump for concretes made with different AWA concentrations and cement content.

with 520 kg/m3 of cement content exhibited lower flow time. For a fixed slump, the increase in AWA concentration increased flow time, which directly affected the viscosity of the concrete. The degree of water retention, and therefore the remaining free water needed to lubricate the concrete, increases with the dosage of AWA, which acts on the aqueous phase. The experimental error in slump and flow time is shown to be limited to 2.6%. On the other hand, the relative error of washout loss was higher than 4.5%. This showed that the slump and flow time are influenced by the dosage of superplasticizer (SP), the cement content and the concentration of anti-washout admixture (AWA). The washout loss is affected by the concentration of AWA, cement content and the dosage of SP. The washout resistance is enhanced by the increase in AWA concentration and reduction of SP dosage. The increase in the SP dosage for a given AWA and cement increases the slump and mass loss by washout. However, for a fixed dosage of SP, the increase in AWA dosage reduces the slump and mass loss by washout. It also showed that for any given concentration of AWA, the addition of SP enhanced fluidity which was reflected by a reduction in flow time and an increase in slump.

6.4.2 Placement methods of UWC Several methods have been applied to place fresh concrete in underwater environments as shown in the following list. The Tremi and Pump methods are the most effective and practical used in recent construction: • • •

Tremi Pump Bottom dump buckets

Underwater concrete • •

151

Toggle bags Bag work.

In order to make the most of underwater concrete construction, the concrete should be produced properly, choosing the right composition, tested well in the production plant and before placing using the appropriate quality control mentioned above. The fresh concrete should be placed continuously to avoid inferior quality produced by an interrupted casting. The site condition, cost, and experience of workmanship will dictate the suitable methods. The Tremi and Pump methods are very common practices, however their procedures are different. The Tremi method operates by means of gravity flow while a pump applies pressure to cast concrete in its final location. The Tremi consists of a rigid pipe with hopper fixed on top of it to collect the concrete and force it down by gravity pressure. Placing enough continuous fresh concrete on the hopper overcomes the friction developed between the internal wall of the Tremi and the concrete to ensure uninterrupted flow of concrete in the pipe. The Pump method is more effective when speedy process is required in severe environments; however, it has also some drawbacks which should be taken into consideration. Both the Tremi and Pump methods have produced good quality underwater concrete; however, it was reported that the Tremi method is more practical in construction for two reasons [10, 22]: 1.

The Tremi produces a more uniform rate of flow than the Pump where the concrete exists at uncontrolled high speed causing great disturbance to the concrete that has already been poured. 2. Owing to the difference between the pump pressure and the fast speed of concrete flow, a vacuum will be created in the pump line. This vacuum will affect the composition of the mix and produce segregation. Concrete consultants and contractors should be familiar with the applications of these two methods and the consequences of their use when deciding which is the appropriate method to adopt.

6.5

References

1 Sonebi, M. “Development of high-performance, self-compacting concrete for underwater repair applications”, Ph.D. Thesis, Université de Sherbrooke, Canada (Sep. 1997) 420 p. 2 Yamaguchi, M., Tsuchida, T. and Toyoizumi, H. “Development of high-viscosity underwater concrete for marine structures”, Marine Concrete, International Conference on Concrete in the Marine Environment, Concrete Society (Sep. 1986) 235–245. 3 Khayat, K. H., Gerwick, B. C. and Hester, W. T. “Self-levelling and stiff consolidated concretes for casting high-performance flat slabs in water”, ACI Concrete International: Design and Construction (15) 8 (1993) 36–43.

152


4 Khayat, K. H. “Effects of anti-washout admixtures on fresh concrete properties”, ACI Materials Journal (92) 2 (March 1995) 164–171. 5 Ghio, V. A., Monteirio, P. J. M. and Gjørv, O. E. “Effect of polysaccharide gums on fresh concrete properties”, ACI Materials Journal (91) 6 (Nov. 1994) 602–606. 6 Khayat, K. H., Sonebi, M., Yahia, A. and Skaggs, C. B. “Statistical models to predict flowability, washout resistance and strength of underwater concrete”, in Production Methods and Workability of Concrete, Glasgow, June 1996 (E & FN Spon, London, 1996) 463–481. 7 Khayat, K. H., Yahia, A. and Sonebi, M. “Applications of statistical models for proportioning underwater concrete”, Fourth International Conference on Recent Advances in Concrete Technology, Supplementary Papers, Japan, (June 1998) 95–113. 8 Graf, O. “Experiments on the behaviour of reinforcement in concrete of various compositions”, Deutscher Ausschuss für Eisenbeton, 71 (1933), 37–60. 9 Brown, D., Bailey, J. and Schindler, A. “The use of self-consolidating concrete for drilled shaft construction”, Proc. Geo Construction QA/QC Conf., 2005, pp. 437–448. 10 Yao, S. X. and Gerwick, B. C. “Underwater Concrete Part II: Proper mixture proportioning and Underwater concrete Part III”, Concrete International, February and March 2004. 11 Bartos, P. “Workability of flowing concrete: assessment by a free orifice rheometer”, Concrete, 1978 (12) 28–30. 12 Sonebi, M. and Bartos, P. J. “Filling ability and plastic settlement of selfcompacting concrete”, Materials and Structures, (35) 8 (2002) 462–469. 13 Ceza, M. and Bartos, P. J. M. “Development of an Apparatus for Testing the Washout Resistance of Underwater Concrete Mixtures”. ACI Concrete in Marine Environment, Proceedings Third CANMET/ACI International Conference, Canada, 1996, pp. 111–126. 14 Bartos, P. Orifice rheometer as a test for flowing concrete, in developments in the use of superplasticizers (W.M.Malhortra, Ed.) ACI SP-68, USA, June 1982, 467–682. 15 Ceza, M. and Bartos, P. J. M. “Assessment of washout resistance of a fresh concrete by the MC-1 test, Production methods and workability of concrete, P.J.M. Bartos et al., 1996, E & FN Spon, pp. 399–413. 16 Neeley, B. D. “Technical report of USA Army Engineering-Waterway Experiment Station”. Evaluation of Concrete Mixtures for use in Underwater Repairs, 1988. 17 Sonebi, M., Tamimi, A. K. and Bartos, P. J. M. “Application of factorial models to predict the effect of anti-washout admixture, superplasticizer and cement on slump, flow time and washout resistance of underwater concrete”, Materials and Structures, (33) 5 (2000) 317–323. 18 Haykawa, M. “Development and Application of Super Workable Concrete”, Proc of Intl. RILEM Workshop on Special Concretes Workability and Mixing, Paisley, 1993, edited by Prof. Bartos, pp. 183–190. 19 EFNARC, Specification and guidelines for self-compacting concrete, February, 2002. 20 Sonebi, M., Tamimi, A. and Bell, D. “Analysis of the performance of fresh underwater concrete produced with polysaccharide gum and superplasticizers using

Underwater concrete

153

plunge and the Orimet tests”, Proceedings of 14th International Conference on Building Materials, Ibausil, Vol. 1, Weimar, Germany, pp. 147–156, Sept. 2000. 21 CRD C61: “Test Method for Determining the Resistance of Freshly-Mixed Concrete to Washing Out in Water”, US Army Experiment Station, Handbook for Concrete, Vicksburg, Mississippi, Dec. 1989, 3 p. 22 Netherlands Committee for Concrete Research, Underwater concrete, HERON, 1973.

7 Fibrous concrete reinforcement S MINDESS, University of British Columbia, Canada

7.1

Introduction

Plain concrete is a brittle material, with low tensile strength and strain capacities. The use of short, discontinuous fibres to strengthen and toughen such materials, which are much weaker in tension than in compression, goes back to ancient times. Probably the oldest written account of such a composite material (clay bricks reinforced with straw), occurs in Exodus 5:6–7: “And Pharaoh commanded the same day the task-masters of the people, and their officers, saying: ‘Ye shall no more give the people straw to make bricks, as heretofore: let them go and gather straw for themselves.’ ”

In modern times, the use of fibres to reinforce cementitious materials goes back to about 1900, when the invention of the Hatschek process enabled the production of asbestos cement. Over the past forty years, there has been a steady increase in the use of fibres in cement and concrete. Today, about 100 million cubic metres of fibre reinforced concrete are produced annually: • • • •

60% – slabs on grade 25% – fibre shotcrete 5% – precast members 10% – miscellaneous other applications.

However, the lack of truly structural applications should be noted; the reasons for this will be discussed later. Fibres are not added to concrete to increase its strength, though some modest increases in strength may occur. The main role of the fibres is to bridge across the matrix cracks that develop as concrete is loaded, and thus to provide some post-cracking ductility. Fibres should not be considered as a replacement for conventional reinforcing bars, even though in some applications this may be the case. They are, in fact, complementary methods of 154

Fibrous concrete reinforcement

155

reinforcing concrete, and there are many applications in which they should be used together.

7.2

How do fibres work?

The mechanical behaviour of fibre reinforced concrete (FRC) depends largely on the interactions between the fibres and the brittle concrete matrix: physical and chemical adhesion; friction; and mechanical anchorage induced by complex fibre geometry or by deformations or other treatments on the fibre surface. The “first generation” steel fibres, produced by shearing thin sheets of steel, were not very efficient, because they were too smooth to bond well with the matrix. Subsequently, many different fibre geometries (Fig. 7.1) were developed to improve the mechanical anchorage, which is the most important of the bonding mechanisms. Surface treatments of the synthetic (mostly polypropylene) fibres have been similarly employed to improve the fibre-matrix bond. As FRC is stressed (either by external loads or by shrinkage or thermal stresses), there is initially elastic stress transfer between the fibres and the matrix. Because the fibres and the matrix have very different elastic moduli, shear stresses develop at the fibre/matrix interface. When the shear stress at the interface is exceeded, debonding gradually begins to occur, and frictional shear stresses become the dominant stress transfer mechanism. At some point during this gradual transition from elastic to frictional stress transfer, some cracking of the matrix occurs, and some frictional slip occurs in the debonded areas. Of course, we are primarily interested in how the fibres in FRC inhibit crack extension once the matrix has cracked, i.e., how they behave in the post-cracking zone. This is governed primarily by the nature of the pull-

Crimped or non-straight

Hooked Button end

Indented Twisted polygonal

7.1 Some types of available steel fibres.

156


out of the fibres from the matrix. It must be emphasized that failure by fibre pull-out is much the preferred mode of failure of FRC; much more energy is consumed in pulling the fibres out of the matrix than in breaking them. It is possible to define a critical length, lc, at which the fibres break rather than pulling out. This must be taken into account when designing or choosing fibres for a particular application. In a properly designed FRC, following the appearance of the first crack, a process of multiple cracking begins, in which the brittle matrix cracks into successively smaller segments (held together by the fibres bridging these cracks). This leads to toughening of the composite. The crack width and crack spacing during this process can be controlled by proper selection of the fibres and the matrix. As stated above, straight, smooth fibres cannot develop sufficient adhesional or frictional bond to be efficient; thus, essentially all of the fibres presently used in practice are deformed in some way (Fig. 7.1), or are surface treated, to increase the bonding with the matrix. This turns out to be much more significant than the fibre length in controlling the degree of bonding.

7.3

Types of fibres

A number of different types of fibres have been developed for use specifically with concrete, though it must be remembered that within each fibre type there are a number of different producers and fibre geometries, leading to different fibre properties. They may be classified as follows: Steel fibres may be produced by cutting wires, shearing sheets, or from a hot-melt extract, and are still the most commonly used fibres. As shown in Fig. 7.1, they are almost always deformed in some way to enhance the fibre-matrix bond. They have been found to be extremely durable in concrete, even though they may rust visibly when exposed at the concrete surface. In some cases, where surface rusting is unacceptable, or in very aggressive environments (e.g., refractory applications) stainless steel fibres may be used. Glass fibres are produced by drawing molten glass in the form of fine filaments through a special bushing. Typically, 204 filaments are drawn simultaneously, and after solidification these are formed into a single strand. Ordinary E-glass (soda-lime glass) fibres and A-glass fibres (borosilicate glass) are not stable in the highly alkaline concrete environment. For use in concrete, alkali resistant glass fibres, typically containing about 16–20% zirconia must be used. Asbestos fibres have been used since about 1900 in the manufacture of asbestos cement pipes, roofing materials, and other building components.

Fibrous concrete reinforcement

157

They have a particular affinity for the cement matrix, and are very effective as reinforcement. Unfortunately, there are significant health risks associated with the production of the asbestos fibres themselves, and so they have largely been replaced with other types of fibres, primarily cellulose fibres. Synthetic fibres have become very common in the last few years. Unlike the fibres mentioned above, they have a significantly lower elastic modulus than does the concrete matrix. At relatively low fibre volumes (

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