Challenges, Opportunities and Solutions in Structural Engineering and Construction

CHALLENGES, OPPORTUNITIES AND SOLUTIONS IN STRUCTURAL ENGINEERING AND CONSTRUCTION

PROCEEDINGS OF THE FIFTH INTERNATIONAL STRUCTURAL ENGINEERING CONSTRUCTION CONFERENCE (ISEC-5), LAS VEGAS, USA, 22–25 SEPTEMBER 2009

Challenges, Opportunities and Solutions in Structural Engineering and Construction

Editor

Nader Ghafoori University of Nevada, Las Vegas, USA

AND

CRC Press/Balkema is an imprint of the Taylor & Francis Group, an informa business ©2010 Taylor & Francis Group, London, UK Typeset by Vikatan Publishing Solutions (P) Ltd., Chennai, India Printed and bound in the USA by Edwards Brothers, Inc, Lillington, NC All rights reserved. No part of this publication or the information contained herein may be reproduced, stored in a retrieval system, or transmitted in any form or by any means, electronic, mechanical, by photocopying, recording or otherwise, without written prior permission from the publisher. Although all care is taken to ensure integrity and the quality of this publication and the information herein, no responsibility is assumed by the publishers nor the author for any damage to the property or persons as a result of operation or use of this publication and/or the information contained herein. Published by: CRC Press/Balkema P.O. Box 447, 2300 AK Leiden, The Netherlands e-mail: [email protected] www.crcpress.com – www.taylorandfrancis.co.uk – www.balkema.nl ISBN: 978-0-415-56809-8 (Hbk) ISBN: 978-0-203-85992-6 (eBook)

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Table of contents

Preface

XV

Acknowledgements

XVII

Reviewers

XIX

Committees

XXI

Keynote papers High-performance materials in earthquake-resistant concrete bridges M. Saiidi, C. Cruz & D. Hillis

3

Implications of air pollution on future electricity generation A. Singh

7

‘‘Root of all evils’’ misunderstanding of construction industry structure D.T. Kashiwagi

15

Concrete and masonry structures Application of nonlinear damper in reinforced concrete structure control F. Hejazi, J. Noorzaei & M.S. Jaafar

25

Behavior of cylindrical R/C panel under combined axial and lateral load T. Hara

31

Calculation method research on the flexural capacity of PSRC beam S. Qin, Y. Wang, F. Li & Z. Ding

39

Cyclic loading deterioration effect in RC moment frames in pushover analysis G. Ghodrati Amiri, B. Mohebi & S.A. Razavian Amrei

45

Evaluating shear capacity of RC joints subjected to cyclic loading using ANN A. Said & E. Khalifa

51

Evaluation of drift distribution in performance-based retrofitting of RC frames G. Ghodrati Amiri & A. Gholamrezatabar

57

Experimental research on high-frequency fatigue behavior of concrete Y. Chen, W. Shao, H. Han, Z. Yin & R. Azzam

63

Experimental study of self-centering RC frames with column yielding mechanism K. Sakino & H. Nakahara

69

Experimental study on high-strength R/C member in tension and shear T. Tamura

75

Improving the behavior of reinforced concrete beams with lap splice reinforcement A.M. Tarabia, M.S. Shoukry & M.A. Diab

81

Load testing a historic monument F.D. Heidbrink

87

V

Modeling of concrete beams prestressed with AFRP tendons Y.J. Kim

93

Nonlinear finite element analysis of unbonded post-tensioned concrete beams U. Kim, P.R. Chakrabarti & J.H. Choi

99

Predicting shear strength of cyclically loaded interior beam-column joints using GAs A. Said & E. Khalifa

105

Punching shear strength of RC slabs using lightweight concrete H. Higashiyama, M. Mizukoshi & S. Matsui

111

Shear strength and deformation prediction in steel fiber RC beams T. Nyomboi & H. Matsuda

119

Torsional resistance of confined brick masonry panel P.K. Singh

127

Steel structures Effect of shear lug on anchor bolt tension in a column base plate P.K. Khan

135

Elastic-plastic bending load-carrying capacity of steel members P. Juhás

141

Local stability and carrying capacity of thin-walled compressed members P. Juhás, M. Al Ali & Z. Kokorud’ová

149

Finite element analysis of wind induced buckling of steel tank S. Borgersen & S. Yazdani

157

Performance based analysis of RBS steel frames P. Alexa & I. Ladar

161

Practical non-prismatic stiffness matrix for haunched-rafter pitched-roof steel portal frames H.K. Issa & F.A. Mohammad

167

Relationship between strength of scaffolds and shear rigidity of frames H. Takahashi, K. Ohdo & S. Takanashi

173

Sequential failure analysis of tension braced MRFs M. Lotfollahi & M.M. Alinia

181

The optimization of the industrial steel building T. Žula & S. Kravanja

189

Composite structures A finite element model for double composite beam S. Duan, R. Niu, J. Xu & H. Zheng

197

A new composite element for FRP-reinforced concrete slabs Y.X. Zhang & Y. Zhu

203

An experimental study on double steel-concrete composite beam specimens S.J. Duan, J.W. Wang, Q.D. Zhou & H.L. Wang

209

Behaviour of FRP wrapped circular reinforced concrete columns M.N.S. Hadi & V. Yazici

215

Contribution of NSM CFRP bars in shear strengthening of concrete members A.K.M.A. Islam

221

VI

Effect of transverse reinforcing on circular columns confined with FRP G. Ghodrati Amiri, A. Jaberi Jahromi & B. Mohebi

229

Experimental investigation of FRP wrapped RC circular and square hollow columns M.N.S. Hadi & Y. Kusumawardaningsih

235

Numerical study on strengthening composite bridges K. Narmashiri & M.Z. Jumaat

241

Repair systems for unbonded post-tensioned 1-way & 2-way slabs with CFRP P.R. Chakrabarti, U. Kim, M. Busciano & V. Dao

247

Strengthening a concrete slab bridge using CFRP composites S.H. Petro, J.T. Peaslee & T.G. Leech

253

Strengthening effect of CCFP for RC member under negative bending I. Yoshitake, S. Hamada, K. Yumikura & Y. Mimura

259

Structural behaviour of reinforced palm kernel shell foamed concrete beams U.J. Alengaram, M.Z. Jumaat & H. Mahmud

265

Dynamic impact and earthquake engineering A note on the model based on the constant Q damping assumption and its corrected models A.S. Takahashi

273

A neural-oscillator model for human-induced lateral vibration on footbridges M. Yoneda

279

Analysis of large dynamic structures in the entertainment industry D.P. Cook & R.T. Robinson

285

Comparison of different standards for progressive collapse evaluation procedures A. Saad, A. Said & Y. Tian

291

Correlation between minimum building strength and the response modification factor L.G. Daza

297

Effect of infill walls in structural response of RC buildings I. Idrizi, N. Idrizi, Z. Idrizi, S. Idrizi & I. Idrizi

303

Estimation of statically equivalent seismic forces of single layer reticular domes H. Abdolpour, Z. Zamanzadeh & A. Behravesh

309

Experimental study of RC slab-CFT column connections under seismic deformations Y. Su & Y. Tian

315

Identification of frequency dependency of quality factor in subsurface ground O. Tsujihara

321

Integrated design and construction to mitigate wind-induced motions of tall buildings K. Moon

327

Mitigation of high acceleration shock waves in hybrid structures S.G. Ladkany & S. Sueki

333

Prefabricated multi-story structure exposed to engineering seismicity ˇ J. Witzany, T. Cejka & R. Zigler

339

Satisfying drift and acceleration criteria with double FP bearing M. Malekzadeh & T. Taghikhany

345

Seismic analysis of interlocking block in wall–foundation–soil system M.S. Jaafar, F. Hejazi, A.A. Abang Ali & J. Noorzaei

351

VII

Shear crack width of RC column with cut-off rebar under cyclic loading T. Tsubaki, M. Dragoi & J. Onishi

357

Structural behavior of steel frame connections subjected to blast G.S. Urgessa & T. Arciszewski

363

Bridges and special structures Applicability of AASHTO LRFD live load distribution factors for nonstandard truck load Y.J. Kim, R. Tanovic & R.G. Wight

371

Full scale test on a bridge PC box girder C. Mircea, A. Ioani & Z. Kiss

377

Long-term deflections of long-span bridges J. Navrátil & M. Zich

385

Structural optimization and computation A revised BESO method for structures with design-dependent gravity loads X. Huang & Y.M. Xie

393

Investigating the buckling behaviour of single layer dome form of space structures Z. Zamanzadeh, H. Abdolpour & A. Behravesh

399

Reasoning on structural timber design for target reliability L. Ozola

405

Shape optimization of shell structures with variable thickness M. Kegl, D. Dinevski & B. Brank

411

Structural damage detection in plates using wavelet transform G. Ghodrati Amiri, A. Bagheri, S.A. Seyed Razzaghi & A. Asadi

415

The non-local theories based on different types of weighted functions and its application Z. Wang & Q. Yang

421

The MINLP approach to structural synthesis S. Kravanja & T. Žula

427

Construction materials Basic study on physical property for calcium-solidification material and on Ca-based concrete A. Shimabukuro & K. Hashimoto

435

Correlations between filler type and the self compacted concrete properties M. Gheorghe, N. Saca & L. Radu

441

Delphi study on Portland cement concrete specifications of ITD H. Sadid, V. Miyyapuram & R. Wabrek

447

Effects of bamboo material on strength characteristics of calcium-based mortar H. Kawamura, K. Hashimoto & A. Shimabukuro

451

Effects of remediation and hauling on the air void stability of self-consolidating concrete N. Ghafoori & M. Barfield

457

Estimation of marine salt behavior around the bridge section E. Iwasaki & M. Nagai

463

Evaluation of alkali-silica reactivity using aggregate mineralogy and expansion tests N. Ghafoori & M.S. Islam

467

VIII

Evolution of Portland cement pervious concrete construction J.T. Kevern

473

Experimental research on regional confined concrete columns under compression X.M. Cao, J.C. Xiao, Z.H. Huang & T.J. Ren

479

Experimental study on dry-shrinkage of lightweight cement mortar T. Watanabe & A. Mori

485

Flexural behavior of high strength stone dust concrete V. Bhikshma, R. Kishore & N.H.M. Raju

491

Hemp: Rediscovered raw building material F. Khestl

495

Influence of admixtures on performance of roller compacted concrete P. Hafiz & A.R. Khaloo

501

Investigation of the effect of aggregate on the performance of permeable concrete C. Lian & Y. Zhuge

505

Investigations on flexural behavior of high strength manufactured sand concrete V. Bhikshma, R. Kishore & C.V. Raghu Pathi

511

Moisture permeability and sorption-desorption isotherms of some porous building materials R. Miniotaite

515

Nested ANOVA model applied to evaluate variability of ready-mixed concrete production C. Videla & C. Imbarack

521

On characteristics of bamboo as structural materials T. Tada, K. Hashimoto & A. Shimabukuro

527

Optimization of fly ash content in suppressing alkali-silica reactivity N. Ghafoori & M.S. Islam

533

Overdosing remediation of plastic SCC exposed to combined hauling time and temperature N. Ghafoori & H. Diawara

539

Research into the optimum level of rock-derived micro-fine particles in sand for concrete T. Kaya, K. Hashimoto & H. Yamamoto

545

Retempering remediation of transported SCC under extreme temperatures H. Diawara & N. Ghafoori

551

Strength property of concrete using recycled aggregate and high-volume fly ash T. Ishiyama, K. Takasu & Y. Matsufuji

557

Strength, sorptivity and carbonation of geopolymer concrete A.A. Adam, T.C.K. Molyneaux, I. Patnaikuni & D.W. Law

563

Suitability of some Ghanaian mineral admixtures for masonry mortar formulation M. Bediako, E. Atiemo, S.K.Y. Gawu & A.A. Adjaottor

569

Ultra light-weight self consolidating concrete M. Hubertova & R. Hela

575

Wood use in Type I and II (noncombustible) construction D.G. Bueche

581

Composite materials Computational models for textile reinforced concrete structures W. Graf, M. Kaliske, A. Hoffmann, J.-U. Sickert & F. Steinigen

IX

589

Properties of natural fiber cement boards for building partitions Y.W. Liu & H.H. Pan

595

Studies on glass fiber reinforced concrete composites – strength and behavior B.L.P. Swami, A.K. Asthana & U. Masood

601

Use of bamboo composites as structural members in building construction T.H. Nguyen, T. Shehab & A. Nowroozi

605

Young’s modulus of newly mixed cementitious extrusion-molded materials T. Watanabe & A. Mori

609

Construction methods Active pier underpinning of Jin-bin light rail bridge in Tianjin J. Bu, N. Sun & S. Huang

617

CFRP liner quality control for repair of prestressed concrete cylinder pipe A. Allan & H. Carr

623

Configuration, evaluation and selection tool (CET) for tunnel construction methods B. Schaiter & G. Girmscheid

629

Formwork specific, process orientated geometrical-path-velocity-time-model (GPVT-model) M. Kersting & G. Girmscheid

635

Open building manufacture systems: A new era for collaboration? M.D. Sharp & J.S. Goulding

641

Precast ferrocement barrel shell planks as low cost roof S.F. Ahmad

645

Tall building boom – now bust? I.R. Skelton, D. Bouchlaghem, P. Demian & C. Anumba

651

The state-of-the-art of building tall I.R. Skelton, D. Bouchlaghem, P. Demian & C. Anumba

657

Construction management An anatomy of speculative claims in construction H.Y. Pang & S.O. Cheung

665

Builders’ perceptions of the impact of procurement method on project quality S. Saha & M. Hardie

671

Business model of the prefab concrete industry – a two-dimensional cooperation network T. Rinas & G. Girmscheid

677

Conservation project management by the architectural digital photogrammetry F. Navarro, A.L. Rodríguez, V. Ávila & C. Loch

683

Construction productivity and production rates: Developing countries C.R. Guntuk & E. Koehn

687

Contractors’ influence within the design process of design-build projects H. Haroglu, J. Glass, T. Thorpe & C. Goodchild

693

Delays in the Iranian construction projects: Stakeholders and economy E. Asnaashari, A. Knight & A. Hurst

699

Designing the relationship between contractor and client to partnership K. Spang

705

X

Determining schedule delay causes under the Build-Operate-Transfer model in Taiwan J.B. Yang & C.C. Yang

711

Dey Street Tunnel: The challenges of a design build project in a congested urban setting M. Trabold

717

Developing a document management model for resolving contract disputes for contractor J.B. Yang & K.M. Huang

723

Development of a decision-making model for requirements management N. Krönert & G. Girmscheid

729

Improving the MEP coordination process through information sharing and establishing trust T.M. Korman

735

Key competences of design-build clients in the People’s Republic of China B. Xia & A.P.C. Chan

739

PPP-risk identification and allocation model: The crucial success factor for PPPs T. Pohle & G. Girmscheid

745

Privileges and attractions for private sector involvement in PPP projects A.P.C. Chan, P.T.I. Lam, D.W.M. Chan, E. Cheung & Y. Ke

751

Training of skills and thinking in structural timber design L. Ozola

757

Use of alternative dispute resolution in construction: A comparative study S.O. Cheung, P.S.P. Wong & P. Kennedy

763

Construction maintenance and infrastructure Construction safety in the repair and maintenance sector A.P.C. Chan, F.K.W. Wong, M.C.H. Yam, D.W.M. Chan, C.K.H. Hon, D. Dingsdag & H. Biggs

771

Challenges of a substation and infrastructure upgrade in an urban downtown setting M.L. Cochrane & C.D. Wagner

777

Construction of concrete embedded, direct fixation, ballasted, LVT and special trackwork K.H. Dunne, N. Slama & K. Wong

783

Design issues of the Palmdale Water Reclamation Plant expansion K. Monroe, J. Stanton, P. Wong & S. Maguin

789

Fuzzy logic based diagnostic tool for management of timber bridges S. Ranjith, S. Setunge, R. Gravina & S. Venkatesan

795

Organizational behavior Behaviors of leadership in architectural offices E. Kasapoglu ˘

805

Gendered behavior: Cultures in UK engineering and construction organizations B. Bagilhole

811

Knowledge management (KM): ‘Integrating past experiences’ A. Weippert & S. Kajewski

817

Managing innovative change within organisations and project team environments A. Weippert & S. Kajewski

823

Personality types of civil engineers and their roles in team performance K. Gautam & A. Singh

829

XI

System service oriented cooperation – lessons for the construction industry D. Lunze & G. Girmscheid

835

Sustainability and energy conservation Building environmental assessment tool S. Vilˇceková, E.K. Burdová & I. Šenitková

843

Building passive design and hotel energy efficiency B. Su & Q. Wang

851

Climatic effects on building facades R. Miniotaite

857

Energy consumption related to winter housing thermal performance B. Su

863

Green energy and indoor technologies for smart buildings F. Vranay, Z. Vranayova, D. Ocipova & D. Lukasik

869

Indoor air quality, distribution systems and energy simulations R. Nagy & I. Šenitková

873

Structural sustainability of high performance buildings M.M. Ali & P.G. Dimick

879

The use of green materials in the construction of buildings’ structure B.O. Russell

885

Engineering economics An interactive model for reduction of failure costs: A process management approach J.E. Avendano Castillo, S.H. Al-jibouri & J.I.M. Halman

893

Forecasting low-cost housing demand in urban area in Malaysia using ANN N. Yasmin Zainun & M. Eftekhari

899

Influence of construction costs on schedule performance J.A. Kuprenas

903

LC maintenance strategy development and decision-making model for street maintenance A. Fastrich & G. Girmscheid

907

Macroeconomic costs within the life-cycle of bridges T. Zinke, T. Wachholz & T. Ummenhofer

915

Maintenance life cycle cost model for drainage systems of infrastructures T. Gamisch & G. Girmscheid

921

Trend analysis of cost performance for public work projects P.P. Shrestha, D.R. Shields & L. Burns

927

Smoothing methodology for time series data F. Khosrowshahi

933

Underlying mechanisms of failure costs in construction J.E. Avendano Castillo, S.H. Al-jibouri & J.I.M. Halman

939

Information technology A survey of the current i-Build practices in the Taiwanese construction industry H.J. Chien, H.W. Chien & J.R. Chang

XII

947

Construction information technology and a new age of enlightenment P.S. Brandon

953

How building information modeling has changed the MEP coordination process T.M. Korman, L. Simonian & E. Speidel

959

Organisational e-readiness in the built environment: People, process, technology E.C.W. Lou & J.S. Goulding

965

Technology projects and their impact on the engineering and construction process M.M. Shoura

969

Geotechnical engineering, foundation and tunneling Dynamic effects of machines on foundations buildings J. Vondrich & E. Thöndel

977

Investigations of the dynamic state of turbo sets foundations A.O. Kolesnikov, V.N. Popov & O.M. Kolesnikov

983

Model tests on bearing capacity of soil-bags H. Yamamoto & S. Jin

987

The construction pre-control of a foundation pit in Shanghai Y. Chen, Z. Yin, J. Wu, M. Wang, Y. Chen & R. Azzam

993

Using BESO method to optimize the shape and reinforcement of underground openings K. Ghabraie, Y.M. Xie & X. Huang

1001

Author index

1007

XIII

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Preface

The mission of the International Structural Engineering and Construction (ISEC) conference is to promote innovative and integrative approaches in life cycle systems in civil and building engineering that include constructability, specifications, design, bidding and construction. The previous ISEC conferences were held in Honolulu, USA (2001); Rome, Italy (2003); Shunan, Japan (2005); and Melbourne, Australia (2007). The purpose of the Fifth International Structural Engineering and Construction Conference (ISEC-5), held in Las Vegas, USA, from September 22 to 25, 2009, was to present and publish recent developments and innovations on these subjects and to continue the transfer of advanced knowledge and technologies as wildly as possible. The exchange of information included all branches of structural and civil engineering, construction engineering and management, contracting and claims, architecture, quality control, housing, materials, education and ethics. These proceedings contain 163 technical articles that were presented during ISEC-5 conference. Each manuscript was peer-reviewed and selected from over 325 abstracts and full papers submitted from 40 countries. My sincerest gratitude is expressed to many reviewers, who are hereby gratefully acknowledged, for their generous efforts. Thanks are also extended to members of the international and local scientific, advisory, and organizing committees; sponsors and cooperating institutions; for their tremendous support towards a successful ISEC-5. Nader Ghafoori Editor Las Vegas, September 2009

XV

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Acknowledgements

Sponsor University of Nevada, Las Vegas Co-sponsors American Society of Civil Engineers American Concrete Institute Association for the Advancement of Cost Engineering Canadian Society of Civil Engineers Chartered Institute of Building, UK Chinese Society of Civil Engineers Concrete Reinforced Steel Institute Japan Concrete Institute Japan Society of Civil Engineers Structural Engineering Institute

XVII

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Reviewers

The Editor gratefully acknowledges the contributions made by the following reviewers who provided valuable comments and recommendations. Sayed Ahmad David Akers Iyad Alattar Chimay J. Anuba Mary Barfield Sai-On Cheung Paul Chynoweth Hamidou Diawara Francis Edum-fotwe Dongping Fang Roger Flanagan Nader Ghafoori Ian Gilbert Gerhard Girmscheid Jack Goulding Dean T. Kashiwagi Mohan Kumaraswamy Kamran Nemati Indubhushan Patnaikuni Habib Sadid Swapan Saha Aly Said Ingrid Senitkova Pramen Shresta Amarjit Singh Ying Tian Ali Touran Frank Yazdani

XIX

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Committees

ISEC Executive Committee Amarjit Singh, Chair, University of Hawaii, USA Frank Yazdani, Secretary, North Dakota State University, USA Nader Ghafoori, University of Nevada, Las Vegas, USA Takashi Hara, Tokuyama College of Technology, Japan Indubhushan Patnaikuni, RMIT University, Australia International Scientific and Technical Committee Nader Ghafoori, Chair, University of Nevada, Las Vegas, USA Chimay J. Anumba, The Pennsylvania State University, USA Nemkumar Banthia, University of British Columbia, Canada Franco Bontempi, University of Rome "La Sapienza", Italy Fabio Casciati, University of Pavia, Italy Sai-On Cheung, City University of Hong Kong, China John Christian, University of New Brunswick, Canada Paul Chynoweth, University of Salford, UK Dejan Dinevski, University of Maribor, Slovenia Francis Edum-fotwe, Loughborough University, UK Dongping Fang, Tsinghua University, China Roger Flanagan, University of Reading, UK Mike Forde, University of Edinburgh, UK Ian Gilbert, University of New South Wales, Australia Gerhard Girmscheid, IBB, ETH Zurich, Switzerland Jack Goulding, University of Salford, UK Takashi Hara, Tokuyama College of Technology, Japan Makarand Hastak, Purdue University, USA Roozbeh Kangari, Georgia Tech, USA Dean T. Kashiwagi, Arizona State University, USA Mohan Kumaraswamy, University of Hong Kong, China Barzin Mobasher, Arizona State University, USA Antonio Nanni, University of Washington, USA Kamran Nemati, University of Washington, USA Indubhushan Patnaikuni, RMIT University, Melbourne, Australia Janaka Y. Ruwanpura, University of Calgary, Canada Swapan Saha, University of Western Sydney, Australia Saiid Saiidi, University of Nevada, USA Ingrid Senitkova, Technical University of Kosice, Slovakia Ahmad Shuaib, American Concrete Institue, USA Amarjit Singh, University of Hawaii, USA Takahiro Tamura, Tokuyama College of Technology, Japan Ali Touran, Northeastern University, USA Francois Toutlemonde, Laboratoire Central des Ponts et Chaussees, France Mumtaz A. Usmen, Wayne State University, USA Ramakrishnan Venkataswamy, University of South Dakota, USA

XXI

Mike Xie, RMIT University, Melbourne, Australia Frank Yazdani, North Dakota State University, USA Local Scientific and Technical Committee Nader Ghafoori, Chair, University of Nevada, Las Vegas, USA David Akers, California Nevada Cement Association, San Diego, USA Iyad Alattar, FHWA, Carson City, USA Reed Gibby, Nevada Department of Transportation, Carson City, USA Tie He, Nevada Department of Transportation, Carson City, USA Samaan Ladkany, University of Nevada, Las Vegas, USA Neil Opfer, University of Nevada, Las Vegas, USA Habib Sadid, Idaho State University, USA Aly Said, University of Nevada, Las Vegas, USA Harry Teng, University of Nevada, Las Vegas, USA Ying Tian, University of Nevada, Las Vegas, USA Local Organizing Committee Nader Ghafoori, Chair, University of Nevada, Las Vegas, USA Mary Barfield, University of Nevada, Las Vegas, USA Greg Desart, GES Inc., USA Hamidou Diawara, University of Nevada, Las Vegas, USA David Goldstein, Geo Tek, Las Vegas, USA Mohammad Shahidul Islam, University of Nevada, Las Vegas, USA Dara Nyknahad, University of Nevada, Las Vegas, USA Samuel Palmer, Terracon, Las Vegas, USA Jonna Sansom, City of Henderson, USA

XXII

Keynote papers

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

High-performance materials in earthquake-resistant concrete bridges M. Saiidi, C. Cruz & D. Hillis University of Nevada, Reno, Nevada, USA

ABSTRACT: Three unconventional details for plastic hinges of bridge columns subjected to seismic loads were developed, designed, and implemented in a large-scale, four-span reinforced concrete bridge. Shape memory alloys (SMA), special engineered cementitious composites (ECC), elastomeric pads embedded into columns, and post-tensioning were used in three different piers. The bridge model was subjected to two-horizontal components of simulated earthquake records of the 1994 Northridge earthquake in California. The multiple shake table system at the University of Nevada, Reno was used for testing. Over 300 channels of data were collected. Test results showed the effectiveness of post-tensioning and the innovative materials in reducing damage and permanent displacements. The damage was minimal in plastic hinges with SMA/ECC and those with built in elastomeric pads. Conventional reinforced concrete plastic hinges were severely damaged due to spalling of concrete and rupture of the longitudinal and transverse reinforcement. 1

2.1

INTRODUCTION

Shape memory alloys are special metallic materials that combine two or more alloys to accomplish certain features. Several types of these alloys are available and have been used in many products but their application in civil/structural engineering has been limited. The particular feature that was of interest in the current study was the superelastic memory effect that allows the SMA yield and dissipate energy but return to its original length upon stress removal under a range of temperature representing those encountered in bridges. The most common type of SMA with this feature is a combination of Nickel and Titanium of approximately equal proportions know as Nitinol (NiTi). The ECC is a grout composed of cement, sand, water, and a patented poly vinyl fiber. It might include fly ash as a substitute for a part of the cement. What makes ECC unique is its ability to undergo large tensile strains of up to 5%. Micro-cracks develop but are spanned by fibers that through a special coating allow for partial slip and relatively large deformations. By using SMA/ECC in column plastic hinge zones it is possible to substantially reduce concrete spalling and damage and to minimize residual lateral displacements (Saiidi & Wang 2006, Saiidi et al. 2007, Saiidi et al. 2009).

Except for bridges that are categorized as critical structures, bridges structures are designed to undergo substantial nonlinear deformations during strong earthquakes and experience serious damage and permanent drift. The design objective for non-critical bridges is to prevent collapse. A new approach to earthquakeresistant concrete bridge design is emerging in which the ‘‘no-collapse’’ target performance is considered to be inadequate. Based on this new approach, even non-critical bridges are to remain functional or nearly so after strong earthquakes. For bridges to continue to be functional the column damage (damage indicator 1) should be none or minimal and permanent lateral displacements (damage indictor 2) should be very small. One approach to accomplish the new emerging performance objective is to make use of highperformance materials and unconventional details that would address one or both damage indicators. This paper describes three high-performance column details and their application in a large-scale bridge model tested on multiple shake tables.

2

SMA combined with ECC

HIGH-PERFORMANCE MATERIALS AND DETAILS

2.2

Built in elastomeric pads

Elastomeric materials are capable of undergoing large tensile strains without failing. They also have a relatively high damping characteristic. Because of their relatively low stiffness, elastomeric pads have been used in civil engineering structures as base isolators to lengthen structural vibration period and take

Three types of high-performance columns were studied: columns with shape memory alloy (SMA) combined with engineered cementitious composites (ECC), post-tensioned columns with built in elastomeric pads, and post-tensioned columns with conventional reinforced concrete.

3

advantage of reduced seismic forces and higher damping. In 2002 a different application of elastomeric pads was explored by incorporating them into the plastic hinge region of concrete bridge columns (Kawashima and Nagai, 2002). The objective in that study was to eliminate spalling of concrete and to reduce damage in the plastic hinge region. The performance was satisfactory up to moderate levels of lateral displacements. Under large displacements, the rubber did not provide sufficient restraint for the column longitudinal bars and the bars buckled and failed due to low cycle fatigue. A modified version of elastomeric pads was developed by the author and his research team to prevent bar buckling. Figure 1 shows a view of the new pad. The fundamental difference between this version and an earlier version of the pad was that the new pad incorporated steel shims to prevent bar buckling and

the pad was relatively thick to allow for large rotation. The pad was vulcanized to steel plates at the top and bottom, and a central steel pipe was used to prevent shear deformations and to serve as a duct for a post-tensioning rod. The outer holes in Fig. 1 were predrilled to allow for the passage of column longitudinal bars. The other holes where only in the end steel plates for the attachment of headed steel dowels to anchor the pad in concrete. 2.3

Post-tensioned columns

One effective approach to reduce residual displacements in columns subjected to earthquake loading is post-tensioning (Sakai & Mahin, 2004). As the column is displaced laterally an axial post-tensioned force tends to return the column to its original position and recenter the column. The post-tensioning approach addresses one of the two damage indicators, the reduction of permanent lateral displacement. However, the column is still susceptible to damage due to spalling of concrete and penetration of damage to the column core under large displacements. Another drawback with this detail is the relatively small amount of energy dissipation these columns offer. To address this problem, mild steel is used in the plastic hinges and is anchored in the footing and the superstructure.

3

IMPLEMENTATION IN LARGE-SCALE BRIDGE MODEL

The aforementioned details were implemented in the piers of a 4-span bridge model. A similar 4-span bridge model incorporating conventional reinforced concrete piers had been tested in a previous study directed by the first author (Nelson et al. 2006). The Figure 1.

New UNR elastomeric pad.

Figure 2.

Bridge geometric details.

4

superstructure was a continuous prestressed reinforced concrete slab, 107 ft (32.6 m) long and 7.5 ft (2.30 m) wide with circular columns having a diameter of 12 in (304.8 mm) and clear height of 72 in (1830 mm). The superstructure consisted of three solid rectangular section beams in each span that were transversely and longitudinally post-tensioned, making the entire superstructure behave as a unit (Fig. 2). An additional dead load of 180 kips (800 kN) was placed on the bridge model deck in order to ensure realistic representation of stresses in the columns. The upper column plastic hinges were made with conventional concrete and steel. Detailed information for the bent design can be found in Hillis and Saiidi (2009). The dimensions of the three bents are illustrated in Fig. 3. Essential details of the columns

Figure 3.

Geometric details of bents.

Figure 4.

Side-view column details.

Figure 5.

Upper plastic hinge after final motion: left to right, SMA pier, PT pier, and ISO pier.

Figure 6.

Lower plastic hinge after final motion: left to right, SMA pier, PT pier, and ISO pier.

5

the earthquake damage compared with conventional reinforced concrete construction. The new built-in shimmed elastomeric pads used in ISO pier proved to be effective in minimizing the earthquake damage even under relatively large displacement ductilities. In the ISO pier the presence of post-tensioning force reduced residual displacements. The post-tensioned pier (PT bent) incorporating conventional reinforced concrete plastic hinges was successful in minimizing residual displacements, but suffered from severe damage due to spalling of concrete and rupture of the transverse steel.

used in SMA, PT and ISO bents are shown in Fig. 4. Over 300 channels of transducers were attached to the model to measure displacements, strains, accelerations, and rotations. The bridge model was subjected to seven coherent earthquake runs simulating the 1994 Northridge southern California earthquake record. The amplitude of the motions increased in successive runs to determine the response under different levels of earthquakes. The first five motions were applied in the two orthogonal horizontal directions. The latter two were applied only in the transverse directions because of the limitations of the abutment loading system. During test 7, several steel bars ruptured in conventional reinforced concrete plastic hinges in two of the piers and hence the model was considered to have failed. 4

ACKNOWLEDGEMENTS The study presented in this article was funded by the US National Science Foundation under the NEESR Grant No. CMS-0420347 and CMS-0402490. The support of program directors Dr. Steve McCabe and Dr. Joy Pauschke is appreciated. The authors are also indebted to the staff of the structures laboratory at the University of Nevada, Reno particularly Dr. Patrick Laplace and Dr. Sherif Elfass.

SUMMARY RESULTS

The measured displacement histories for all three piers indicated that the residual displacement was insignificant in all three piers, and that the recentering technique of post-tensioning used in the PT and ISO piers and the superelastic feature of SMA incorporated in the SMA/ECC pier were effective. As mentioned in the previous section during the last motion the upper plastic hinges in the ISO pier failed due to rupture of the longitudinal and transverse bars. The upper plastic hinge in the PT and SMA piers underwent severe concrete damage exposing the longitudinal and transverse bars. Figure 5 shows the upper plastic hinges after the final test. Except for the post-tensioning forces, the details were conventional reinforced concrete plastic hinges meeting the current code requirements. In the same piers, there was little damage in the lower plastic hinges in the SMA pier and ISO pier, with the damage being limited to minor cracking (Fig. 6). In the PT pier, however, the damage was severe and the spiral reinforcement fractured in one of the columns. Two conclusions are drawn: (1) that both SMA/ECC combination and the built in elastomeric pad drastically reduced the damage making the bridge potentially serviceable even after a strong earthquake that led to the failure of several plastic hinge, and (2) the plastic hinge in the PT pier was severely damaged. Post-tensioning alone was not sufficient to keep the PT pier serviceable. 5

REFERENCES Hillis, D. and Saiidi, M. (2009), ‘‘Design, Construction, and Preliminary Analysis of an Innovative Bridge Model,’’ Center of Civil Engineering Earthquake Research, Department of Civil Engineering, University of Nevada, Reno, Nevada, Report No. CCEER-09-xx (in preparation). Kawashima, K. and Nagai, M.: Development of a reinforced concrete pier with a rubber layer in the plastic hinge region, Structural and E. Engineering, Proc. JSCE, 703/I-59, 2002, pp. 113–128. Nelson, R., Saiidi, M. and Zadeh, S. 2007, ‘‘Experimental Evaluation of Performance of Conventional Bridge Systems,’’ Center of Civil Engineering Earthquake Research, Department of Civil Engineering, University of Nevada, Reno, Nevada, Report No. CCEER-07-4. Saiidi, M. and Wang, H. ‘‘An Exploratory Study of Seismic Response of Concrete Columns with Shape Memory Alloys Reinforcement,’’ American Concrete Institute, ACI Structural Journal, Vol. 103, No. 3, May–June 2006, pp. 436–443. Saiidi, M., Zadeh, M. Ayoub, C. and Itani, A. ‘‘A Pilot Study of Behavior of Concrete Beams Reinforced with Shape Memory Alloys,’’ Journal of Materials in Civil Engineering, ASCE, Vol. 19, No. 6, June 2007, pp. 454–461. Saiidi, M., O’Brien, M. and Zadeh, M. ‘‘Cyclic Response of Concrete Bridge Columns Using Superelastic Nitinol and Bendable Concrete,’’ American Concrete Institute, ACI Structural Journal, Vol. 106, No. 1, January–February 2009, pp. 69–77. Sakai, J. and Mahin, S. ‘‘Mitigation of residual displacements of circular reinforced concrete bridge columns,’’ 13th World Conference on Earthquake Engineering: Conference Proceedings, Vancouver, British Columbia, Canada, August 1–6, 2004, Paper No. 1622.

CONCLUSIONS

The material presented in this article showed that the superelastic characteristic of SMA bars observed in individual bar tests may also be observed in SMAreinforced concrete columns. The combination of ECC and SMA was found to significantly reduce

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Implications of air pollution on future electricity generation A. Singh University of Hawaii at Manoa, Honolulu, USA

ABSTRACT: Scientists are unable to agree on the real causes of global warming. Whatever the cause, one plausible ingredient—carbon emissions—stands to damage the air environment and health. This paper argues that any source of energy that produces carbon must be heavily curtailed. Considering that most of the world’s energy comes from coal, other substitutes must be legislated, and the energy sources selected must deliver in gigawatt quantities day and night and be ‘‘clean’’. Only nuclear energy fits that bill. This paper discovers that modern day nuclear energy is ‘‘renewable’’. Nuclear energy cost is competitive against energy from fossil fuels, and safety concerns are easily managed while people’s perceptions are becoming more favorable towards nuclear energy. Various concerns of nuclear energy are dispelled. 1

gases such as chlorofluorocarbons (CFCs). Greenhouse gases—produced by natural and industrial processes, result in CO2 levels of 380 parts per million per volume (ppmv) in the atmosphere. The levels in 1900 were about 300 ppm. (Wikipedia: Greenhouse Gas 2008; Patterson 2005). But, water vapor has been shown to be the largest contributor to greenhouse gases by far (Lindzen 1992; Hieb (2008) reports that 95 percent of all greenhouse gases are water vapor: ‘‘Of the 186 billion tons of CO2 that enter Earth’s atmosphere each year from all sources, only six billion tons are from human activity. Approximately ninety billion tons come from biologic activity in Earth’s oceans and another ninety billion tons from such sources as volcanoes and decaying land plants’’ Apparently, only 3.2 percent of atmospheric CO2 is generated from human activities such as coal plants and fossil fuel burning, whereas the plant kingdom and natural volcanic activity contribute to natural CO2 . In contrast, 99.99 percent of water vapor is natural and comes from oceans and clouds. And, eighteen percent of methane and 65 percent of CFCs are from human activity (Hieb 2008). Even if all human induced methane and CFCs increased ten times, which is realistically impossible, they would likely have a miniscule effect on global warming. In addition, Essenhigh (2008), a professor of energy conversion, believes that CO2 is simply unable to drive global warming, but that global warming may drive CO2 increases. And, if humankind wishes to reduce the greenhouse effect of water vapor, it is absolutely beyond our control. An increase of 2◦ C can occur if the CO2 concentration increases to 450 ppm, which may take a century or two (How much 2006). But, this estimate is based on models that make too many assumptions and can therefore not stand up to scientific rigor. At most, the

INTRODUCTION

Hansen et al. (2006) from the National Aeronautics and Space Administration (NASA) put the rate at 0.2◦ C per decade over the past thirty years. Maybe that However, Robinson et al. (2007) establish that sea-surface temperatures in the Sargasso Sea were higher in 1,000 AD by 1◦ C and in 1,000 BC by 2◦ C. Antarctica is reported to have warmed only 0.2◦ C from 1850 to 2000, and actually cooled markedly during the 1990s while the Southern Hemisphere rose by 1.4◦ C over the past century (New Evidence 2006). According to data provided by Robinson et al. (2007) the average ice core temperatures were this high 110,000 years ago. Of one thing there is no doubt: global warming is a fact and has been an old story for the past 15,000 years, helping us emerge from the ice age into a beautiful garden. But, not all parts of the Earth are heating uniformly. The current alarm about global warming is important, but must be placed in perspective and not exaggerated (Hieb 2003; Michaels 1998). Therefore, it is meaningful to note that the cause of global warming is disputed vehemently by scientists. The differences of opinion and interpretation between science writers and climatologists, astronomers and solar-terrestrial physicists, atmospheric scientists and astrophysicists, chemists and paleoclimatologists, and so many other scientists, are substantial, such that it is hard to believe in any one of them with any certainty (Kaplincki 2006).

1.1

Greenhouse gases

Important greenhouse gases are water vapor, methane, nitrous oxide, carbon dioxide, and miscellaneous

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human contribution to the greenhouse effect is 0.28 percent, which is too small to matter (Hieb 2003). It can thus quite safely be deduced that CO2 is probably not the big culprit of global warming that the media has made it out to be (Chandler 2007; Beck 2006), though it is not entirely causeless.

2

as SO2 causes constriction of the finer air tubes of the lungs, thus making it difficult to breathe naturally.’’ (Air Pollution 2008) 2.3

Ozone

Although CO2 is not a lung irritant, Jacobsen (2008) found that CO2 serves to increase ground-level ozone. This ozone is a lung irritant. Cases of asthma have gone up, such that by 2020, there could be twentynine million Americans suffering from asthma (Pew 1998).

HEALTH EFFECTS OF AIR POLLUTION

2.1 Fog and smog The fact is that while the relatively small amounts of CO2 and nitrous oxides put into the atmosphere probably do not have a significant impact on global warming, they have a very significant impact on air pollution and air quality. Nitrogen oxide gases are produced from fossilburning power plants and the transportation sector. They irritate the lungs of humans, birds, and animal species, causing bronchitis. Owing to reduced resistance to respiratory infections, they increase the incidence of pneumonia among humans. Several pollutants are produced by burning fossil fuels and contribute to the noxious fumes that cause smog; these include carbon monoxide, nitrogen oxides, sulfur oxides, and hydrocarbons. Hydrocarbons combine in the atmosphere to form tropospheric ozone that descends to surface levels and becomes a major component of smog. ‘‘Human exposure to ozone can produce shortness of breath and, over time, permanent lung damage. Research shows that ozone may be harmful at levels even lower than the current federal air standard. In addition, it can reduce crop yields’’ (The Hidden 2008). Our entire industrial activity, which is supposed to alleviate the human condition, is creating conditions that harm us.

2.4

Health risks

Children are more quickly affected by air pollution and more easily develop bronchitis and earaches (Outdoor 2008). It is well known that the incidence of respiratory diseases is on the rise all around the world. In one of the longest, largest studies on the effects of air pollution on lung cancer and heart diseases, 500,000 adults were surveyed in more than one hundred cities from 1982 to 1998, Pope et al. (2002) found air pollution as a convincing cause of increased lung cancer and cardiopulmonary diseases. ‘‘More than 220 million Americans breathe air that is one hundred times more toxic than the goal set by Congress ten years ago, according to figures calculated by the Environmental Defense Fund (EDF). And for eleven million people, the cancer risk from their neighborhood air is more than one thousand times higher than Congress’s goal, the group says’’ (Most Americans 1999). 2.5

Global spread of air pollution

A report last year identified China as the worst air polluter in the world; 656,000 Chinese per year die from diseases caused by air pollution alone. The corresponding numbers for India are 527,000 (Platt 2007). We also know that air pollution from China is traveling across the Pacific. NASA satellite data have confirmed that nearly 10 billion pounds of aerosol pollution reached North America from East Asia (NASA 2008), the largest contributor of which was China. Cliff (2006) reports that we are already breathing Chinese pollution in North America.

2.2 Acid rain The phenomenon of acid rain occurs in industrialized areas that emit nitrogen oxides and sulfur oxides into the atmosphere. These gases—and smog—combine with water vapor in clouds to form sulfuric and nitric acids, which become part of rain. Carbon dioxide combines with water in the clouds to form carbonic acid. As the acids accumulate on the surface after acid rain, lakes and rivers become too acidic for plant and animal life (The Hidden 2008). Humanity is simply hurting itself. Acid rain falls on a third of China’s territory and 70 percent of Chinese rivers and lakes are toxic, unfit for drinking. Moreover, the sulfur dioxide (SO2 ) produced in coal combustion, in addition to causing acid rain, causes about 400,000 premature deaths a year. Most of these deaths are from lung and heart-related diseases

3

THE KYOTO AGREEMENT

The Kyoto Agreement accepts that global warming is a result of burning fossil fuels, which we have shown is possibly unconvincing. The Kyoto Agreement goes a step further in their false premise to require that gaseous emissions be cut by 2012 by 5.2 percent below the emission levels of 1990 (Bloch 2008). At best, the effect of the Kyoto Agreement would be a

8

Germany produces, and nine times the world average (Solar/Wind, 2008 (1997 data)). The World Energy Outlook of the International Energy Agency (2006) says ‘‘the current pattern of energy supply carries the threat of severe and irreversible environmental damage—including changes in global climate.’’ Therefore, it is imperative to reverse the trend of energy production in the world. Importantly, the world is fast running out of oil, which will bring fossil fuel electricity generation and transportation closer to a standstill. Some other form of energy generation will have to be substituted at a fast pace. World demand is quickly depleting oil reserves. The oil company, Royal Dutch Shell, estimates that ‘‘ . . . after 2015 supplies of easy-to-access oil and gas will no longer keep up with demand’’ (Shell 2008). The same article consequently concludes that there will be a need for nuclear power and alternate sources. It is possible to produce a maximum of 20 percent of the United States’ energy needs through wind power (Solar Energy International 2008), another 5 percent by solar power, and about 7 percent by hydroelectric power. However, these will not replace or close down existing fossil fuel power plants, unless repealed by legislation, which is definitely recommended. The total world installed capacity for solar power is a mere 0.8 GW (Solar/Wind 2008). Solar power for consumption on a mass scale is currently impeded by technical difficulties in storing energy during cloudy periods and night. The technology is simply undeveloped for a reliable, continuous supply of electric power. The feasibility of wind and sun sources of energy producing in gigawatt quantities day and night is not established. Geothermal energy is not available everywhere. Thus, we have to think of ‘‘clean’’ alternates other than solar and wind energy to meet world energy demands. Surely, there must be a better way to generate electricity that does not damage health. Luckily, there is.

reduction in global temperature by one-twentieth of a degree Fahrenheit by 2050 (Global Warming 2003). Thus, even in its own argument, the Kyoto Agreement doesn’t go far enough. A deeper and bolder global agreement is needed, one that has real teeth, like eliminating coal generation by 2050. 4

FEASIBLE SOLUTION

If we accept the likely premise that humans will not forego their electricity use and modes of transportation, and will thus resist reverting to the Middle and Dark Ages, what is the solution if we don’t want to damage our health? It is evident that we must target energy production, for which two areas stand out most prominently: energy production and transportation fuel. This paper is focusing only on energy production. Technologists propose renewable energies such as solar, wind, and geothermal. However, like any energy engineer will tell you, the electricity that can be potentially harnessed from these sources is not more than 25 percent of our needs. Night production of solar and wind is a problem, and storage mechanisms are in an immature stage of development. What’s more, hydroelectricity causes severe ecological damages of its own, not to mention that the potential is limited. Many novel sources of energy, such as tidal and wave power, are still being researched for safe and reliable implementation, since hostile ocean conditions pose challenges for wave structures (Wave and tidal 2000; Wave power 2008). Ocean thermal energy conversion is a new possibility for renewable energy, but one that lacks a track record. So what’s next, if we want to steer away from fossil fuel energy, but still want a decent standard of life? Practical engineers and politicians cannot live on hope that new technology will arrive one day. Nations could be steeped deep in crises by that time. It’s not reasonable to wait till new inventions arrive to secure our future. 5

6

NUCLEAR ENERGY

The only logical and available answer for the world is nuclear energy, at least for the foreseeable future or until some other technology proves to be effective. Let’s look at nuclear energy in greater detail:

ENERGY PRODUCTION

The aim now is to produce ‘‘clean’’ energy, in ‘‘large quantities,’’ with no ‘‘environmental effects.’’ Well, the total electric installed capacity in the United States was 1,000 gigawatts (GW) in 2005 (Industry 2008; Wind power 2008). The distribution was approximately as follows: Coal, 49%; Nuclear, 19.3%; Natural gas 18.7%; Hydropower 6.5%; Fuel oil 3.0%; Biomass 1.6%; Wind 1.2%; Geothermal and Solar 0.6%. Thus, 73% of the electricity comes from fossil sources. The USA consumes 12,000 kWh of electricity per person per year. This is twice the amount that

6.1

Production capability and trends

Of the relatively large countries, 70 percent of France’s electricity comes from nuclear energy, while for the United States the number is 20 percent. China and Russia are extremely busy cornering the world’s uranium market and India is also on the path of nuclear renaissance. The United States is also apparently heading towards a nuclear revival.

9

be recycled into new nuclear fuel and does not need to be stored in repositories, such as the Yucca Mountain repository in Nevada. Hence, neither radioactive decay nor storage are engineering problems for the future, at all.

There are 442 nuclear power plants in the world, of which 104 are in the United States. Worldwide, thirtyfour reactors are under construction, and 280 are proposed. In China alone, 116 new reactors are planned or proposed (Uranium 2006; Freeman 2005). As many as twenty-nine new reactors may receive licenses for construction in the United States (Biello 2007). General Electric, the world’s largest utility company, plans to enter into partnerships for nuclear construction around the world (General 2007; New Nuclear 2005). These statistics illustrate that the world is moving head on towards nuclear power. One ton of nuclear fuel delivers as much energy as 20,000 tons of coal. Consequently, the advantages in the logistic management of uranium compared to coal are evident. Apparently, there is no shortage of uranium on Earth. Uranium can even be extracted from sea water. There are about 4.5 billion metric tons of uranium available in the world’s oceans (Uranium from Sea water 2006). This is enough to last humankind approximately 36,000 years, compared to 130 to 300 years from coal reserves, for electricity production (World Coal Institute 2008; Elert 2005).

6.4

Safety of nuclear power plants

The safety of nuclear power plants centers on two main issues: (1) maintaining public safety in the event of radioactivity leaks, and (2) eliminating damage through malfunctions or accidents. Radioactivity leakage has been a concern for many decades. In 2000, four of the eight reactors in Pickering, Canada were finally shut down as a result of tritium leaks, which is a cancer causing substance (Sierra Club 2001). The Oyster Creek Plant in New Jersey was reported to have elevated levels of cesium-137 near the plant. Cesium-137 is another carcinogenic substance (Cacchioli and Larsen 2006). There are many more such cases around the world. Thus, when people are concerned about radioactivity leaks from nuclear power plants, it is not altogether without reason (Leak forces 2000). Nevertheless, it is not well known that the Three Mile Island (Pennsylvania) accident was contained without immediate harm to anyone. The world can also be confident that a Chernobyl type of poor design will never be repeated again in IAEA supervised countries, though human errors cannot be ruled out (for anything). In over 12,700 cumulative reactor years of commercial operation in thirty-two countries, there has never once been a death outside of Chernobyl. Moreover, current reactor design emphasis has shifted in the last eight years from reliance on containment structures to safety through improved design of the reactor plant itself (History of Nuclear 2008). Safety (2007) reports: ‘‘The U.S. Nuclear Regulatory Commission (NRC) specifies that reactor designs must meet a 1 in 10,000year core damage frequency, but modern designs exceed this. U.S. utility requirements are 1 in 100,000 years, the best currently operating plants are about 1 in 1 million and those likely to be built in the next decade are almost 1 in 10 million.’’ Advanced nuclear reactors, known as nextgeneration reactors, such as the ones going up in Japan (the first of which was constructed in 1996), contain safety improvements based on operational experience. Beyond the safety engineering already standard in Western reactors, they have passive safety systems, which require no operator intervention in the event of a major malfunction. All modern reactors are designed to automatically shut down in the event of earthquakes. Owing to an emphasis on safety, safety systems account for about one-fourth of the capital costs of modern reactors (Safety 2007). Safe (2004) reports that Generation-IV reactors, which will be in

6.2 Carbon saving In an MIT study that aimed to expand current worldwide nuclear generating capacity almost threefold by the year 2050, it was found that 1.8 billion metric tons of carbon emissions would be saved annually (MIT 2003). This much is equal to about one-third of the total current carbon emissions. Nuclear energy production is not ‘‘carbon-free,’’ but it does minimize carbon emissions in its process cycle (Nuclear Comeback, 2007). The world has no choice but to resort to the only electricity generating technique that can deliver the goods, once fossil fuels are depleted, which is expected to start around 2012 (Shell 2008). Needless to say, the period between 2015 and 2020 might be tumultuous for the world owing to energy shortages and skyrocketing prices of oil. These will further affect food distribution, plausibly causing widespread famines (Edwards 2000). The disruption to society from mining to manufacturing to transportation might be tremendous. 6.3 Nuclear waste and fuel recycling That nuclear waste has a disposal problem is a myth from the days of old technology when conventional thermal reactors operated in a ‘‘once-through’’ mode. Today, it is possible to recycle spent nuclear waste from thermal reactors by reprocessing in a ‘‘closed’’ fuel cycle, or from fast reactors by reprocessing in a balanced ‘‘closed’’ fuel cycle (MIT 2003). Robinson and colleagues (2007) affirm that spent nuclear fuel can

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service by 2030, will provide dramatic improvements in reactor design. Generation-III+ reactors are already markedly improved over the Generation-I. These assure that radioactivity leakage is minimized below harmful levels. 6.5

gas, 1200; hydroelectricity, 4000; nuclear, 31 (all in Chernobyl). This data is self-explanatory, in that fatalities from other energy sources are much more inspite of the hype of nuclear catastrophe. Moreover, because all deep-earth minerals contain radioisotopes, they generate radioactivity when burning (McBride et al. 1978). Aubrecht (2003) reported that coal has uranium and thorium radioisotopes ranging representatively from 1 ppm to 2 ppm. Their conclusion was ‘‘that Americans living near coal-fired power plants are exposed to higher radiation doses, particularly bone doses, than those living near nuclear power plants that meet government regulations.’’ Francis (2001) discovered that ‘‘a coal plant releases about 74 pounds of uranium-235 each year, enough for two or more nuclear bombs.’’

Pilferage of nuclear material

There has been general concern that a multitude of nuclear power plants in the world will make them more susceptible to pilferage of nuclear material for terrorist operations. There are many cost-effective practicable steps that can be taken to prevent this, as well as circumstances that already prevent it: 1. Place all new nuclear power plants under International Atomic Energy Agency (IAEA) safeguards 2. Increase security, much of which is already enhanced at nuclear plants around the world, 3. Weapons-grade plutonium and uranium require purification up to 80 percent levels, which nuclear plants don’t do. 4. Generation-IV reactors are more resistant to attempts to divert material for illegal weapons manufacturing (Safe 2004).

6.8

Studies have shown that nuclear energy is cheaper than energy from coal or gas, while other studies find nuclear energy comparable or slightly more expensive than coal (‘‘The Economics of Nuclear Power’’ 2007). Overall, nuclear power is competitive with coal from a cost perspective. A report from the Organization for Economic Cooperation and Development (OECD 2005) found that nuclear power was cheaper than fossil fuels among 80 percent of the countries in a represesntative sample (Nuclear Energy Agency 2005). Data was projected to the year 2010. However, the World Nuclear Association (Ritch 2006) claims that the OECD (2005) report underestimates the nuclear advantage and so claims that the generating costs for the year 2010, projected at a 5 percent discount rate, are 2.1 to 3.1 cents/kWh for nuclear energy; 2.5 to 5.0 cents/kWh for coal; and 3.7 to 6.0 cents/kWh for natural gas. Additionally, nuclear energy production costs in the United States have dropped from a total of 2.47 cents/kWh in 1981 to 1.72 cents/kWh in 2003. Uranium prices have risen dramatically, but coal prices rose by 42 percent between 2000 and 2006 (U.S. Price 2008). Further, operating costs of nuclear plants in the United States dropped by 44 percent between 1990 and 2003 (Ritch 2006).

Moreover, existing pilferage around the world is unknown in IAEA supervised reactors. What happens in Russia, China, and Pakistan by way of missing uranium fuel can continue to happen with or without increased nuclear energy. 6.6

Safety from terrorism

Shopping malls, football games, and conventional industrial facilities have traditionally lesser security than nuclear power plants. Thus, nuclear plants are inherently safe. With regard to aerial or missile attacks, the following studies were done: 1. A Boeing 767 hitting at 560 kph would cause no penetration of the containment. 2. Sandia labs demonstrated that an F4 Phantom jet hitting a 3.7-m concrete slab at 765 kph would destroy the plane. The concrete penetration would only be 6 cm (Safety 2007). Because containment structures are huge, an attack inside a plant that causes loss of cooling, core melting, and breach of containment, would not result in significant radioactive releases (Safety 2007). The death and destruction from terrorist attacks at other installations can have worse outcomes. 6.7

Costs of nuclear power

6.9

Construction costs of nuclear plants

In OECD (2005), nuclear power construction costs are believed to be in the order of $2.3 billion for a 1.2-GW nuclear plant. Add to this the economies of scale that can bring about savings of 15 to 20 percent. Holt and Behrens (2003) report that costs range anywhere on average from $3 to 6 billion. It is not clear what size the plants are in these two latter reports. Many nuclear plants are constructed with two and three reactors together; the CANDU reactor in Pickering, Canada, had eight reactors. In this regard, the final cost

Safety comparison with coal

How safe is nuclear energy compared to its rivals coal and others? For the period 1970–1992, immediate fatalities were as follows: coal, 6400; natural

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that increased nuclear energy will come to be a reality in the United States in the next few years.

data of OECD as verified in Robinson et al. (2007) is taken as representative: the current estimated cost of a 1.2-GW plant is $2.6 billion. Thus, estimates of electricity use to 2050 in the USA indicate that if new clean energy plants must be constructed and the old, dirty coal plants be replaced, the total cost for nearly 1,000 additional gigawatts will be $2.95 trillion. Spread over 41 years—a rough economic estimate—brings the total annual investment to $72 billion, which is easily financed in the current economic environment of the United States. How much do coal-fired plants cost to construct? More than 130 new coal-fired plants have been proposed over the next ten to twenty years. However, Odell (2008) reports, ‘‘The costs of constructing and operating these plants are highly uncertain due to multiple factors in the industry, and the owners will face significant financial, economic, and environmental risks.’’ Given that coal plants might simply be shut down due to air pollution concerns, implies that investors will be unable to recover their invested sums. This is making bankers and lenders balk. Nevertheless, coal plants are typically cheaper than nuclear plants. A 1-GW coal plant can be expected to cost between $1.3 to 2.77 billion depending on whether it is a traditional plant, an integrated gasification combined cycle plant, or a fluidized bed plant (What is a coal fired plant 2008; Energy Information Administration 2008; Wald 2007). So, coal plants could be half as expensive to construct, or else nearly as expensive as nuclear plants, but definitely more expensive to operate.

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8

SUMMARY AND CONCLUSIONS

Coal and oil still contribute up to 50 percent of the world’s electricity, but we don’t have time on our side to continue using them. Oil will be depleted in about three decades, while the accompanying gasses and particulate matter that are emitted by burning coal and fossils fuels are major air pollutants that have serious health risks for world citizens and animal life. In addition, CO2 causes surface ozone to form, which is a lung irritant. It is appropriate to ask what type of air we are bequeathing to future generations. From an engineering and medical perspective, we have to reverse direction now. Nuclear energy is a clean available alternative to the hazards of burning coal, biomass, and fossil fuels for electricity. I argue that neither radioactive waste, nor the safety of nuclear plants, nor the threats of terrorism are significant concerns in relation to nuclear energy. In addition, all spent fuel can be recycled with modern nuclear technology. Moreover, nuclear plants emit less radioactivity than coal plants, since all deep-earth materials, such as coal, have some uranium and thorium. More ominously, uranium-235 can be extracted from coal emissions. Sooner or later, every country in the world, rogue or not, will be able to do so. While the capital costs for coal plants to install 1 to 1.2 GW of electricity range from half the cost of nuclear plants to equal the cost, the increasing health hazards of coal plants are beginning to make coal plants a risky venture that is turning away financiers. Operation costs of coal plants are significantly higher than nuclear plants, not to mention the enormous logistics of transporting huge quantities of coal from mines and coal extraction factories to coal power plants, given that one ton of uranium produces as much energy as 20,000 tons of coal. Public opinion of nuclear energy is now favorable by a 2:1 ratio, making it likely that the future of electricity in the USA and the world is likely to be nuclear energy in the years to come. It takes four years to construct a 1-GW nuclear plant, and ten years to develop uranium mines. Thus, if we are serious about maintaining our quality of life, and breathing clean air, we must make a conscious policy agreement now to switch to nuclear energy. The world has no other sensible alternative at the current moment. The future of electricity generation is staring us in the face and the technology is sitting there for us to adopt, unless we wish to revert to the middle ages.

PUBLIC OPINION ON NUCLEAR POWER

In the eyes of the public and numerous bureaucrats and legislators, nuclear power is still a dirty word. The damage to the good effects of nuclear power was considerable after the Three Mile Island episode. In many parts of the United States, however, public opinion is now becoming favorable toward nuclear power. In a survey, Bisconti (2007) discovered that ‘‘Using a mix of low-carbon sources, including nuclear energy and renewables, makes sense to the public for producing the electricity we need while limiting greenhouse gases. There is near consensus (85 percent) on this concept, and this consensus encompasses the range of demographic groups.’’ Moreover, it was seen that fifty-six percent of the public would ‘‘definitely build more nuclear power plants in the future,’’ while ‘‘72 percent agree that we should keep the option to build more nuclear power plants in the future.’’ Overall, about 63 percent favor nuclear energy, while 31 percent oppose it. It can thus be interpreted that there is a more than good chance

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ACKNOWLEDGMENTS

Hieb, Monte. 2008. ‘‘Global Warming: A Chilling Perspective.’’ . Hieb, Monte. 2003. ‘‘Global Warming: A Closer Look at the Numbers: Water Vapor Rules the Greenhouse System.’’ January 2003. . Hansen, J., Sato, M., Ruedy, R., Lo, K., Lea, D.W., and Medina-Elizade, M. 2006. ‘‘Global Temperature Change.’’ Proc. Natl. Acad. Sci., 103, 14288–14293, doi: 10.1073/ pnas. 0606291103. Holt, M., and Behrens, C.E. 2003. ‘‘Nuclear Energy in the United States.’’ Congressional Research Service. . ‘‘How Much CO2 Emission is Too Much?’’ 2006. . November 6, 2006. ‘‘Industry statistics.’’ 2008. Edison Electricity Institute. . International Energy Agency (IEA). 2006. World Energy Outlook. IEA, Paris. Jacobson, M.Z. 2008. ‘‘On the Causal Link Between Carbon Dioxide and Air Pollution Mortality.’’ Geophysical Research Letters (GRL), doi. 10.1029/2007GL031101. Kaplincki, J. 2006. ‘‘A Cosmic Connection.’’ Nature, September 14, 2006. ‘‘Leak Forces Pickering Shutdown.’’ 2000. CBC News, November 10. . Lindzen, Richard S. 1992. ‘‘Global Warming: The Origin and Nature of the Alleged Scientific Consensus’’ Cato Institute, Vol 15 No. 2, . Martin, Robert 2008. ‘‘History of Nuclear Power Safety.’’ The Nuclear Info Ring. . McBride, J.P., Moore, R.E., Witherspoon, J.P., and Blanco, R.E. 1978. ‘‘Radiological Impact of Airborne Effluents of Coal and Nuclear Plants.’’ Science: 202, 1045. Michaels, P. 1998. ‘‘Global Deception: The Exaggeration of the Global Warming Threat.’’ Center for the Study of American Business, Policy Study No. 146. MIT. 2003. ‘‘The Future of Nuclear Power.’’ An Interdisciplinary MIT Study, Massachusetts Institute of Technology, Cambridge, Mass. ‘‘Most Americans Face Cancer Risk from Toxic Air Pollution.’’ 1999, April 23. . NASA. 2008. ‘‘Satellite Measures Pollution from East Asia to North America.’’ Goddard Flight Space Center. , March 17, 2008. ‘‘New Evidence Shows Antarctica Has Warmed in Last 150 Years.’’ 2006. September 6. Science Daily. ‘‘New Nuclear Power Plants Consortia.’’ 2005. http:// www.aaenvironment.com/NewNukes4.htm. ‘‘Nuclear Comeback.’’ 2007. [Video recording], New York: The TV Set: First Run/Icarus Films. Nuclear Energy Agency and International Energy Agency. 2005. ‘‘Projected Costs of Generating Electricity.’’ Paris: OECD Publishing.

This paper is extracted, condensed, and rewritten ‘‘with permission from ASCE’’ from an article by the same author titled ‘‘The Future of Energy’’, Leadership and Management in Engineering, Vol. 9 (1), pp. 9–25, Jan 2009. REFERENCES ‘‘Air Pollution, the Environmental Imperative for Renewable Energy: An update.’’ 2008. . Aubrecht, G.J. 2003. ‘‘Nuclear Proliferation Through Coal Burning.’’ Physics Education Research Group, Department of Physics, Ohio State University. Beck, C. 2006. ‘‘Water Vapor Accounts for Almost All of the Greenhouse Effect.’’ . Biello, D. 2007. ‘‘Nuclear Power Reborn.’’ Scientific American. . Bisconti, A.S. 2007. ‘‘Public Supports Climate Change Action, but is Unclear on Nuclear Energy’s Role in Preventing Greenhouse Gases.’’ Perspectives on Public Opinion, Nuclear Energy Institute. . Bloch, M. 2008. ‘‘What is the Kyoto Protocol?’’ . Cacchioli, J., and Larsen, E. 2006. ‘‘Radioactive Isotope Found Near Oyster Creek Nuclear Power Plant.’’ Alliance for Nuclear Responsibility. . Chandler, D.L. 2007. ‘‘Climate Myths: CO2 Isn’t the Most Important Greenhouse Gas.’’ New Scientist Environment, May 16, 2007. Cliff, S. 2006. ‘‘We Are Breathing Chinese Pollution.’’ New Perspectives Qrtrly. 23(4), 78–79. Edwards, R. 2000. ‘‘July 2030: Going Back to the Future.’’ The Ecologist. . Elert, G. (ed.). 2005. ‘‘Price of Coal.’’ The Physics Factbook. . Energy Information Administration. 2008. . Electric Power Monthly, March 2008. Essenhigh, R. 2008. ‘‘Does CO2 Really Drive Global Warming?’’ . Francis, D.R. 2001. ‘‘Energy Study Gives Black Marks to Coal, Boost to Nukes.’’ The Christian Science Monitor, 29 May 2001. Freeman, M. 2005, ‘‘China’s 21st Century Nuclear Energy Plan.’’. Executive Intelligence Review, http://www. larouchepub.com/other/2005/3208china_htr.html. ‘‘General Electric Interested in Lithuania Nuclear Power Plant Project,’’ 2007, Space Daily. Jan 23, 2007. http:// www.spacedaily.com/reports/General_Electric_ Interested_ In_Lithuania_Nuclear_Power_Plant_Project_ 999.html.

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Odell, AA.M. 2008. ‘‘New Coal Plants: At What Cost?’’ Sustainability Investment News, March 20. OECD. 2005. ‘‘Projected Costs of Generating Electricity.’’ Nuclear Energy Agency. Paris: OECD Publication. ‘‘Outdoor Air Pollution: Possible Health Effects.’’ 2008. March 2008. Patterson, T. 2005. ‘‘The Geologic Record & Climate Change’’ . Pew Environmental Health Commission. 1998. ‘‘Attack Asthma: Why America Needs a Public Health Defense System to Battle Environmental Threats.’’ . Platt, K.H. 2007. ‘‘Chinese Air Pollution Deadliest in World.’’ National Geographic News, July 9, 2007. Pope III, C.A., Burnett, R.T., Thun, M.J., Calle, E.E., Krewski, D., Ito, K., and Thurston, G.D. 2002. ‘‘Lung Cancer, Cardiopulmonary Mortality, and Long-Term Exposure to Fine Particulate Air Pollution,’’ JAMA, 287:1132–1141. Radiation Leak. 2008. Web Currents . Robinson, A.B., Robinson, N.E., and Soon, W. 2007. ‘‘Environmental Effects of Increased Atmospheric Carbon Dioxide.’’ Journal of American Physicians and Surgeons, 12, 79–90. Ritch, John. 2006. ‘‘The New Economics of Nuclear Power.’’ World Nuclear Association. London: Carlton House. Safe, Secure and Inexpensive Power from Latest Generation of Nuclear Reactors.’’ 2004. Inside Science News Service. . Safety of Nuclear Power Reactors.’’ 2007. Nuclear Issues Briefing Paper 14, September 2007. . Shell Predicts Energy Shortage by 2015.’’ 2008. Calgary Herald, January 26, 2008.

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Sierra Club of Canada. 2001. ‘‘The Canadian Nuclear Lesson.’’ http://www.sierraclub.ca/national/programs/atmosphere-energy/nuclear-free/reactors/nuclear-and-climchg-6-01.html. Solar Energy International. 2008. ‘‘Energy Facts.’’ Solar Energy International. . March 2008. Solar/Wind.’’ 2008. . The Economics of Nuclear Power.’’ 2007. Uranium Information Centre Briefing Paper 8, December 2007. http://www. uic.com.au/ nip08.htm. The Hidden Cost of Fossil Fuels.’’ 2008. Union of Concerned Scientists. . Uranium from Seawater (Part 1).’’ 2006. Peak Oil Debunked. . [Blog] January 7, 2006. Uranium from Seawater (Part 2)’’ 2006. Peak Oil Debunked, . [Blog] January 8. U.S. Price of Coal: Annual Data.’’ 2008. . Wald, M.L. 2007. ‘‘Costs Surge for Building Power Plants.’’ New York Times, July 10. Wave Power. 2008. MIT Energy Initiative. http://web.mit.edu/ mitei/research/transformations/ocean.html. Wave and Tidal Power. 2000. Fujita Research Reports. http://www.fujitaresearch.com/reports/tidalpower.html. What is a Coal-Fired Plant? ( 2008) Appalachian Voices 2008. . Wind Power: U.S. Installed Capacity (Megawatts), 1981– 2007.’’ 2008. American Wind Energy Association. . World Coal Institute. 2008. ‘‘Where is Coal Found?’’ http://www.worldcoal.org/pages/content/index.asp?Page ID=100.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

‘‘Root of all evils’’ misunderstanding of construction industry structure D.T. Kashiwagi Arizona State University, Tempe, Arizona, USA

ABSTRACT: In tough economic times, the construction industry participants seek for efficiency, value, ‘‘winwin,’’ low cost, maximized profit, and best value. In tough economic times, the problems of the industry are magnified, as construction services are commoditized. Analysis of the Construction Industry Structure (CIS) shows that a stable construction industry should maximize value, vendor profit, quality and performance, while minimizing cost and risk. PBSRG has been testing out the CIS concepts for 15 years on over 600 tests ($650 M of construction services) with dominant results: 98% on time, on budget, and meeting client expectations, minimizing up to 90% of client risk/project management functions. Movement from the low bid award environment to a best value environment will lower costs and increase profits. The answer lies in the transfer of risk and accountability to the industry participants who have the expertise to minimize risk (and not pass risk.) 1

time despite a continual effort to improve, the system may be stable, and the only way to increase production is to overhaul the entire system structure. Deming (1982, p. 317) states that: ‘We rely on our experience.’ The answer is selfincriminating: it is a guarantee that this company will continue to pile up about the same amount of trouble as in the past. Why should it change? Experience without theory teaches nothing. In fact, experience cannot even be recorded unless there is some theory, however crude, that leads to a hypothesis and a system by which to catalog observations. Sometimes, only a hunch, right or wrong, is sufficient theory to lead to useful observation.’’ The industry has tried to improve the construction performance by using different delivery systems (Gransberg & Windel 2008; Lam et al. 2004; Kashiwagi et al. 2002):

INTRODUCTION

Construction industry performance has suffered issues with performance and capability for the past ten years (Lepatner 2007; CFMA 2006; Simonson 2006; Flores & Chase 2005; Adrian 2001). Academic research has had minimal impact in assisting the construction industry (Kashiwagi et al. 2008; Adeyemi et al. 2009): 1. Academic research is funded by government agencies that are bureaucratic and not accountable for results. 2. The cycle of research development is too long for results to be tested, modified, and implemented. 3. Industry does not run hypothesis testing with academic researchers. 4. Research is done by graduate students who continually change. There is no corporate knowledge that stops the reproduction of the same concepts. Many of the studies propose solutions that are never tested, are too complicated to be tested, and regenerated five years later. 5. Research effort durations and funding levels are not enough to have repeated hypothesis testing.

1. 2. 3. 4.

Design-build (DB.) Construction Management @ Risk (CM@Risk) Indefinite Delivery Indefinite Quantity (IDIQ.) Public Private Partnerships (PPP.)

Deductive logic or common sense proposes the construction industry nonperformance as a systems problem instead of an individual’s or participant’s nonperformance (Sullivan et al. 2009). Directed solutions toward industry participants will not solve a systems issue. For example, if the contractor’s cannot manage themselves, putting a construction manager over the contractor is not a solution. If the contractor is forced to submit the lowest possible bid using the lowest priced subcontractors and materials, tighter inspection will not raise the level of performance. Management, direction, control, and inspection will not increase a vendor’s performance (Deming 1982; Goldratt 1980).

The construction industry and the academic research community have been reactive in nature. The method of research depends heavily on literature search and survey of industry participants. This mode of research forces the continuation of the existing status quo. Without actual tests and test results (customer satisfaction/expectation, on time, and change order rates), decision making by industry experts will maintain the status quo. Deming (1982) stated that two important ingredients of progressive thought were ‘‘theory’’ and deductive logic concepts. If problems continue over

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A stable system’s production cannot be increased by management and control. The entire system must be changed to increase production.

moved to the right hand side (Quadrant I—Price Based Award and Quadrant II—Value Based or Best Value). The price based system has the following characteristics (Goodridge et al. 2007):

2

1. The client’s representative directs through specifications and controls and inspects the contractor’s work for compliance to minimum standards. 2. There is no transfer of control and accountability to vendors/contractors. 3. The client’s specification’s minimum standards are turned to maximums due to the price based environment and driven downward by vendors (Fig. 2). 4. Contractors who are short on experience, reactive, and only do what they are directed, be-come more competitive because they can give a lower initial price, but increase the prices when the deviations are identified. Low performing contractors are far more likely to submit deviations due to a lack of perception, incomplete directions from the owner, unforeseen events which are normally identified and handled by the high performance contractors, but unknown to the inexperienced contractor. 5. Contractors who manage and minimize risk and who are the better value when considering total or actual project cost, become noncompetitive because they do not fully utilize the change order or deviation system (Fig. 3). The logic model here forces the vendors to not think of the client, to degrade performance, and to create profit from the lowering of quality and increasing of risk.

PROBLEM

The construction industry and construction management academic research groups have not been able to improve construction industry performance, nor assisted the industry to increase their technical expertise/craftsperson skill level, nor explain why the construction industry performance is continually low and problematic. The construction industry is one of the few industries that have the same problem as twenty years ago, and whose industry productivity has fallen in the last few years (Adrian 2001). 3

HYPOTHESIS

The construction industry performance is a system’s problem. Deductive logic and dominant concepts will replace literature search and industry perceptions and be used to confirm the system’s problem, and a systems solution will be created to correct the problem. The academic research community has actually increased the problem by using concepts such as management, technical solutions, and delivery systems that are unenforceable and are not consistent with the goals of efficiency, minimization of risk, dominant measurements and accountability. 4

Due to the worldwide competitive environment, buyers of design and construction services have moved to the right hand side (Quadrant I—Price Based Award and Quadrant IIP—Value Based or Best Value). The price based system has the following characteristics (Goodridge et al. 2007):

CONSTRUCTION INDUSTRY STRUCTURE

The Construction Industry Structure (CIS) (Fig. 1) (Kashiwagi 2009) defines the construction or any industry based on performance and competition. Due to the worldwide competitive environment, buyers of design and construction services have

1. The client’s representative directs through specifications and controls and inspects the contractor’s work for compliance to minimum standards.

Figure 1. Construction industry structure. Source: (Kashiwagi 2009).

Figure 2. Minimum/Maximum dilemma. Source (Goodridge et al. 2007).

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13. 14. 15. 16.

Increased inspections. Detailed specifications. Increased flow of information. Contract and procurement officer become more important. 17. Reactive as contractors submit bids at the last minute. In the best value environment, risk is transferred to the high performance vendor (who minimizes the risk.) The only vendor personnel who can reasonably accept the risk are the experienced, high performing individual(s) who have expertise and can minimize the risk. Low performing vendors and individuals bring both the vendor and the buyer risk and need to be managed, directed, and controlled by the client (Sullivan et al. 2005). They are never the best value.

Figure 3. Price based award. Source: (Goodridge et al. 2007).

2. There is no transfer of control and accountability to vendors/contractors. 3. The client’s specification’s minimum standards are turned to maximums due to the price based environment and driven downward by vendors (Fig. 2). 4. Contractors who are short on experience, reactive, and only do what they are directed, become more competitive because they can give a lower initial price, but increase the prices when the deviations are identified. Low performing contractors are far more likely to submit deviations due to a lack of perception, incomplete directions from the owner, unforeseen events which are normally identified and handled by the high performance contractors, but unknown to the inexperienced contractor. 5. Contractors who manage and minimize risk and who are the better value when considering total or actual project cost, become noncompetitive because they do not fully utilize the change order or deviation system (Fig. 3). The logic model here forces the vendors to not think of the client, to degrade performance, and to create profit from the lowering of quality and increasing of risk.

5

In the best value or value based environment, designers and contractors must do the following to be awarded work (Fig. 4) (Kashiwagi et al. 2006): 1. Show that they are the best value. Compete based on past performance of the company and key contractor components including project manager and site superintendent and critical sub-contractors, or lead architect, project manager, and critical subconsultants/engineers. 2. Know what has to be done. Quantify the scope of work and the risk that they do not control that is not in the scope of the project and have a plan to manage and minimize the risk through preplanning and clear communications with the client. 3. Be able to communicate. Interview of their key personnel to identify if they can predict and create a baseline of the project from beginning to end, if they can manage and minimize the risk that they do not control, if they can preplan and be accountable, and

The characteristics of the price based, specification driven, management, direction, and controlled environment include (Kashiwagi et al. 2004; Kashiwagi et al. 2002): 1. 2. 3. 4. 5. 6. 7. 8. 9. 10. 11. 12.

BEST VALUE ENVIRONMENT

Subcontractors’ prices are shopped. Manufacturers continually lower quality. Standards are subjective. No performance measurements. Social relationships are important. Prices are meaningless due to no connection to performance. Inefficient. Incentives used. No accountability. No transfer of risk and control to vendor. Increased transactions. Increased management positions.

Figure 4. Best value requirements for selection. Source: (Kashiwagi 2009).

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to identify if they have the expertise to manage the project. 4. Fair price. The price is determined by the professional/vendor, and it must be competitive. If the value of the different proposers cannot be identified, the low price proposer will be awarded the project. There is not a client party to double check and assist the proposer, on their price and baseline plan. Professionals/Vendors must accept accountability and liability for submitting a correct price.

Figure 6. Construction industry structure. Source: (Kashiwagi 2009).

The best value professional/vendor is then required to have a baseline plan based on time and cost, to have a weekly risk report (WRR) and risk management plan (RMP) to manage and minimize all concerns and risks that they do not control before the project begins and track deviation of time and cost during execution of the contract, justifying and validating every deviation (Pauli et al. 2007). The designer and/or contractor and all their key components will then be rated on their performance, and that rating will become 50% of their future performance rating that will be a large factor on their ability to get future work with the client. Figure 5 shows a vendor’s business perspective of why best value environments motivate vendors to use highly trained personnel who can manage and minimize risk. Vendors respond to three types of owners: owners who transfer risk and control of projects to expert vendors, owners who partner with the vendors and share the risk, and the price based owner who directs, manages, and hires only the vendor who is the lowest price (assuming that all vendors are the same) (Sullivan et al. 2005). Vendors usually have high performers (who they must pay the highest salaries), medium performers, and inexperienced personnel (who earn the lowest salaries, need constant management and direction and are the most inefficient). The most efficient relationship is between the best value client and the high performing personnel. The vendor can maximize their profit even though the high performance personnel have a higher salary cost.

Figure 7. PIPS/PIRMS filters. Source: (Kashiwagi 2009).

The efficiency and the ability of the high performing personnel to manage and minimize risk to finish on time, minimizing change orders, and doing quality work creates a win-win between the client and vendor (high value, lowest possible cost and maximized profit for the vendor). However, many clients do not understand this principle as they do not transfer risk and accountability, they maximize their decision making, and they buy based on low price and use the management, direct, control, and inspect approach. The Vendor Business Model (Fig. 7) identifies the client who uses the management, direction, and control approach as the source of the risk of nonperformance. By selecting the price based award with its management, direction, control, and inspection approach, the client has forced the vendor (whose objective is the maximization of profit) to send the most inexperienced personnel to the client. Another drawback of the price based system approach is that there is no motivation of the vendor’s personnel to become highly trained, proactive, and quality oriented. This describes the dilemma of the design and construction industry as they attempt to maintain the professionalism of their practices. Their two largest problems are the inability to increase the number of highly trained craftspeople and the turnover of contractors without regard to experience (Angelo 2003; Krizan & Winston 1998; Post 2000; Simonson 2006). 6

PRICE BASED ENVIRONMENT IS A SYSTEM ISSUE

The description of the price based environment matches the design and construction industry performance

Figure 5. Vendor business model. Source: (Kashiwagi 2009).

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by other participants, mainly the client in the form of over-expectations, items outside of the scope, decision making at the wrong time during the process, and the changing of expectations) (Fig. 8). High performers/experts see the project from beginning to end, before they compete for a project, and know the risk that they do not control before they accept the project. High performance firms (Fig. 9):

previously discussed. The Construction Industry Structure (CIS) (Fig. 1), the Minimum Requirement/ Maximum Value (Fig. 2), the Price Based Contractor Reaction (Fig. 3) and the Vendor Business (Fig. 5), all identify the design and construction price based delivery system as the major source of nonperformance, no accountability, and diminishing industry skill levels. The price based environment is setup and controlled by the client/buyer. The deductive logic shows that the potential solution lies in changing the system from a price based system to a best value system. 7

1. Minimize risk before they start a project by putting the right expertise on the project who knows how to do the project based on experience. 2. Identify the scope of the project, a baseline schedule, what the project will cost, and the solution of the project before project award. 3. Can also identify what risk that may affect the project due to client over-expectations, client nonperformance, problems caused by other participants (permitting, review bodies, client related individuals) potential unforeseen conditions (defined by the scope and baseline schedule.) 4. High performance vendors maximize their profit by finishing ahead of schedule. 5. High performance vendors are motivated by profit (finishing ahead of schedule and meeting client expectations of time, cost, and quality.)

PIPS/PIRMS

The Performance Information Procurement System/ Performance Information Risk Management System (PIPS/PIRMS) creates a best value environment and is a system solution to the design performance issue. PIPS/PIRMS has the following major phases (Fig. 6): 1. Phase I: Selection of the best value. 2. Phase II: Pre-award/pre-planning, and creation of weekly risk report (WRR) and risk management plan (RMP.) 3. Phase III: Project delivery by risk management. The process has six major filters if awarded based on performance and price (Fig. 7): 1. Past performance of the vendor, critical personnel (project manager/lead designer), and critical sub-vendors (engineering and professional consultants). 2. Identification of the general scope of the project described in two pages. 3. Risk Assessment/Value Added submittal that identifies the risk that the vendor does not control and that is not in the scope, and how they will minimize that risk, and add value (that is not in the scope as a requirement, and it will a dominant difference between vendors) that makes them different from their competitors. 4. Interview of the critical personnel of the vendor. 5. Prioritization of the best value based on performance (the past performance information, scope rating, RAVA rating, and interview rating and price.) 6. Creation of the WRR and RMP. 7.1

Figure 8. Inexperienced vs. experienced vendor risk model. Source: (Kashiwagi 2009).

High performers have minimal technical risk

A major departure from the traditional practices is the handling of risk. Information Measurement Theory (IMT) identifies that by definition, high performance/expert design firms and their personnel have minimal or no technical risk. The only risk they have is risk that they do not control (risk that is brought

Figure 9. New risk model. Source: (Kashiwagi 2009).

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The authors propose that the client’s misunderstanding of risk minimization is the biggest risk, and their management, direction, inspection, and decision making approach allows risk to be maximized because it allows non-expert vendors who do not have the expertise in minimizing risk to ‘‘seem competitive’’ and ‘‘qualified.’’ As decisions and expectations are made by the client and the designer, they are actually creating risk that will occur as the project unfolds (Fig. 10). Traditionally the clients believe that the difference between their expectations and the actual state is risk and it is the vendor who is creating the risk.

schedule, and force all participating parties who integrate with them, to be accountable for their actions by documenting all deviations to the scope. 2. Preplan and create a baseline plan, and thus have the technical skills to do the design. 3. Quality Control. Review and correct their own work. 8

PIPS/PIRMS TEST RESULTS

The best value Performance Information Procurement System (PIPS) and follow-on Performance Information Risk Management System (PIRMS) has been undergoing tests for the past 15 years. The results include (Kashiwagi 2009):

7.2 Transfer risk and accountability to vendor Inexperienced vendors cannot see from beginning to end, are reactive to the client’s behavior and needs, and are concerned with technically being qualified to do the project. They do not consider the non-technical risk that they do not control. Procurement processes that use contracts that concentrate on the technical requirements of a project attract vendors and personnel who are inexperienced and reactive (Figs. 7, 10). The PIPS/PIRMS process forces the vendors to identify their relative level of expertise by:

1. Over 600 tests delivering $650 M of design and construction services. 2. 98% performance (on time, no contractor generated cost change orders, and client satisfaction.) 3. Client project/risk management minimized up to 90%. 4. Vendors make 5–15% greater profit. 5. Process has been used successfully by Arizona State University to deliver $1.5 B of non-construction services (IT Networking, Food Services, and other professional services.) 6. Processes used by the U.S. Army Medical Command and has minimized change orders by 68% and time delays by 52%.

1. Identifying what they think the scope of the project is. 2. Identifying what may be expected that is not in the scope, or what may stop them from completing their project that is not identified in the scope (the risk that they do not control) and propose how they will manage and minimize the uncontrolled risk. 3. Identify what makes them dominantly better than their competition. 4. Having their key individuals interviewed to identify their level of perception, vision, and capability to preplan and create a conservative baseline plan.

The former procurement director and now Associated Vice-President of University Business Services Ray Jensen stated that: ‘‘I have been successful in the business of procurement and services delivery for the past 30 years. I saw in PIPS/PIRMS, improved solutions of performance/contract administration issues that are so dominant, that I am willing to change my approach to the business after 30 years’’ (Jensen 2009).

The selected best value vendor must then: 1. Propose a milestone schedule, manage and minimize the risk that will deviate from the miles-tone

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CONCLUSIONS AND RECOMMENDATIONS

The construction industry and the academic research community have not had a significant impact on improving construction performance because they may not have perceived it to be a systems/environmental problem. One of the reasons that the systems solution was not perceived or considered is the traditional research methodology of construction management of using literature search and survey of experts which maintains the status quo approach of using management techniques to solve the industry problems. The authors replace the traditional approach by using Deming and Goldratt’s analysis of stable systems and

Figure 10. Traditional risk model. Source: (Kashiwagi 2009).

20

theory of constraints to hypothesize that a continually broken industry may actually be a ‘‘stable system.’’ The traditional client management, direction, and control approach to increasing construction performance and minimizing construction risk was identified by deductive logic as a system that cannot increase construction performance. Two more critical analyses were done:

Jensen, R. 2009. Interview performed by Dean Kashiwagi on February 11, 2009. Arizona State University, Tempe, Arizona. Kashiwagi, D.T. 2009. Best Value. Tempe: Performance Based Studies Research Group (PBSRG). Kashiwagi, D., Sullivan, K., Badger, W. & Egbu, C. 2008. Business Approach to Construction Research. COBRA 2008 The construction and building research conference of the Royal Institution of Chartered Surveyors, Dublin Institute of Technology, London, UK, (September, 2008). Kashiwagi, D., Koebergen, H., Zenko, D. & Sullivan, K. 2006. Bridging the Gap: Performance and Efficiency in Design Build Delivery. ASC International Proceeding of the 42nd Annual Conference, Colorado State University, Fort Collins, CO, Track 20 (April 19–22, 2006) Kashiwagi, D., Koebergen, H. & Egbu, C. 2004. Construction Nonperformance is a Process Problem. COBRA 2004—Construction and Building Research Conference, Leeds Metropolitan University, Thorpe Park, Leeds, UK, Track 2 (Sept. 7, 2004). Kashiwagi, D. & Mohammed, M. 2002. A Study Comparing US Construction Delivery Systems Based on Business and Management Principles. COBRA 2002—Construction and Building Research Conference, Nottingham Trent University, UK, pp. 37–53 (2002) Kashiwagi, D. & Savicky, J. 2002. The Relationship between the Specification, Low-Bid Process and Construction Nonperformance. First International Conference on Construction in the 21st Century, Miami, FL, pp. 371–377 (2002). Krizan, W.G. & Winston, S. 1998. Scarcity of Skilled Workers Will Put Brakes On Growth. Engineering News Record, 240 [4], pp. 95, 98, 101. Lam, E., Chan, A. & Chan, D. 2004. Benchmarking designbuild procurement systems in construction. Benchmarking. Bradford: 2004. Vol. 11, Iss. 3; p. 287. Lepatner, B.B. 2007. Broken Buildings, Busted Budgets. The University of Chicago Press. United States of America: Chicago. Pauli, M., Sullivan, K. & Kashiwagi, D. 2007. Utilization of Risk Management to Show Value and Increase Competitiveness. COBRA 2007 Construction and Building Research Conference, Georgia Institute of Technology, Atlanta, GA, USA, CD Track 59 (September 6–7, 2007). Post, N.M. 2000. No Stamp of Approval On Building Plans: Contractors sound off over difficulties with bid documents. Engineering News Record, 244 [17], pp. 34–37, 39, 42, 45–46. Simonson, K. 2006. Quick Facts. Association of General Contractors, Chief Economist Report. Sullivan, K., Kashiwagi, J. & Kashiwagi, D. 2009. The Optimizing of Design Delivery Services for Facility Owners. Unpublished Manuscript. Sullivan, K., Egbu, C. & Kashiwagi, D. 2005. Forcing Contractors to Improve with Minimized Management Effort. CIB W92 Construction Procurement: The Impact of Cultural Differences and Systems on Construction Performance, University of Nevada—Las Vegas (UNLV), Las Vegas, NV, 2, pp. 683–691 (February 8–10, 2005).

1. A 15 year research test of over 600 deliveries resulted in 98% performance, up to 90% less management and control function and increased the vendors profit by as much as 15% (100% increase.) 2. The process was taken outside of the construction industry and successfully run on $1.5 B of services at Arizona State University, resulting in over $100 M of efficiency. This research has tremendous impact on future construction management research because it proposes that the management, direction, and control approach is actually the source of the construction performance risk. It proposes that the systems problem requires the use of leadership/alignment techniques, quality control and quality assurance (dominantly understood in the manufacturing industries), and will affect the future roles of all the participants in the construction industry.

REFERENCES Adeyemi, A., Kashiwagi, J., Kashiwagi, D. & Sullivan, K. 2009. New Procurement Approach in Graduate Education. Manuscript submitted for publication. Association of Schools of Construction of Southern Africa. Livingstone: Zambia. Adrian, J.J. 2001. Improving Construction Productivity. Construction Productivity Newsletter, Vol. 12, No. 6. Angelo, W. 2003. Project Management: Keeping Costs Under Control. Engineering News Record. 250 [1], p. 45. CFMA’s 2006 Construction Industry Annual Financial Survey. Moss-Adams, LLP, Eighteenth edition. Deming, E.W. 1982. Out of the Crisis. Mass.: Mass. Institute of Technology. Flores, V. & Chase, G. 2005, Project Controls from the Front End. Cost Engineering. April 2005, Vol. 47, No. 4; pp. 22–24. Goldratt, E. & Cox, J. 1992. The Goal: A Process of Ongoing Improvement. North River Press. (May 1992). Goodridge, S., Kashiwagi, D., Sullivan, K. & Kashiwagi, J. 2007. The Theoretical Evolution of Best Value Procurement Research. Symposium on Sustainability and Value through Construction Procurement 2006, CIB W092— Procurement Systems, Digital World Center, Salford, United Kingdom, pp. 310–321 (November 29–December 2, 2006). Gransberg, D. & Windel, E. 2008. Communicating Design Quality Requirements for Public Sector Design/Build Projects. Journal of Management in Engineering. April 2008, Vol. 24, Iss. 24; pp. 105–110.

21

Concrete and masonry structures

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Application of nonlinear damper in reinforced concrete structure control F. Hejazi, J. Noorzaei & M.S. Jaafar University Putra Malaysia, Selangor, Malaysia

ABSTRACT: The application of modern control techniques to diminish the effects of seismic loads on building structures offers an appealing alternative to traditional earthquake resistant design approaches. Over the past decade there has been significant research conducted on the use of damper devices for dissipating seismic energy. This paper describes the development of a numerical finite element algorithm used for analysis of reinforced concrete structure equipped with shakes energy absorbing device subjected to dynamic load such as earthquake. For this purpose a new nonlinear viscose damper is proposed and a finite element program code for analysis of reinforced concrete frame buildings is developed. The effect of proposed damper is evaluated by implementation in a simple model of reinforced concrete frame building. The results show that using new proposed damper as seismic energy dissipation system effectively reduced structural response in earthquake excitation.

1

of structures (Hybrid systems) are conducted in many investigations to obtain a good compromise of energy dissipation systems combination (Yang & Agrawal 2002). It is clear from critical literatures review that most of the investigators employed the finite element techniques for the purpose of idealization under plane strain idealization (Bouaanani & Lu 2008; MartinezRodrigo & Romero 2003; Sekulovic et al. 2002). So far little or no information is available about three dimensional formulation framed structures, equipped with damper element. Hence there is need to improving upon physical and constitutive modeling of reinforced framed building equipped with viscous dampers. Hence in the present work a new three dimensional nonlinear viscous damper is formulated and it is applicable in reinforced concrete structures. Then the proposed damper is applied to one story example of reinforced concrete framed buildings subjected to earthquake and seismic response of model is investigated.

INTRODUCTION

The traditional approach to design earthquake resistance building is providing adequate strength and stiffness against lateral forces. Alternatively, latest concepts of earthquake energy dissipation system and damper device have been devoted via advance technology and techniques to reduce earthquake effect and preclude seismic damage of buildings. Recently many investigations have been conducted to evaluate and analyze the seismic response of structures equipped different types of damper. Viscous dampers are known as effective energy dissipation devices improving structural response to earthquakes. The damping force developed by the viscous damper depends on the physical properties of the fluid used in the device (Constantinou & Symans 1992). So energy dissipation due to sloshing liquid in toms shaped dampers is studied analytically accounting for nonlinearities and viscous effects so as to be applicable at resonance (Modi et al. 1990). Performance of viscous dampers based on the experimental study carried out on steel frame model, subjected to horizontal and vertical earthquake vibrations with varying intensities (Zhang & Soong 1992). In the other hands, the role of viscous damping in preventing buildings from collapse during intense earthquake ground motion was extensively investigated by using numerical modeling (Soda 1996). The result of application of viscous dampers to structures are shown by analyzing of few structures equipped energy dissipation devices and demonstrated their advantages and disadvantages (Tezcan & Uluca 2003). Examining the performance of application viscous dampers with other various type of dampers for control

2

PROPOSED FINITE ELEMENT MODELLING OF FRAMED STRUCTURE EQUIPPED WITH ENERGY DISSIPATION SYSTEM

The following elements have been used for the purpose of finite element idealization of reinforced concrete framed buildings equipped with viscous dampers: 2.1

Beam column element

In this study, a two node, three dimensional beamcolumn element having two rigid ends of different

25

lengths for simulating the finite widths of the beamcolumn connection as shown in Figure 1 (Thanoon et al. 2004). So the stiffness matrix of this element that derived using bending theory for small transverse displacements is and coefficients of this matrix are calculated using the following equation: 

L

krs =

EI (x)ψr (x)ψs (x)dx

Frame element Rigid ends

(1)

0

where L is the member length, EI is the flexural rigidity of the member, ψr (x) is the displacement function (shape function) along the member resulting from unit displacement along the rth degree of freedom (ur = 1), and ψs (x) is the shape function when (us = 1) is impressed on the member. For the member with two rigid ends, the stiffness coefficients are obtained in a similar way after modifying the shape function ψ for the member with uniform cross section. These functions are written choosing the origins different for different segments of the element for convenience. Stiffness coefficient is now obtained by re-writing equation (1) in the following form:  krs =



a

EI ψr (x)ψs (x)dx +

0

 +

0

0

Figure 1.

Mathematical model of building structure.

c

a

b

Elastic frame Element

c

EI ψr (x)ψs (x)dx

Rigid Block

Hinge zone

Figure 2.

Inelastic modeling of frame member.

Rigid Block

b

EI ψr (x)ψs (x)dx

(2)

where a and b is length of two rigid block in both end of element and c is length of elastic part as showed in Figure 2. The stiffness and damping properties of frame element are time dependent and a function of its deformation status and deformation histories, therefore member of stiffness matrix are changed in each time steps of imposing external load. The beam-column element with two unequal rigid ends in three dimensional is shown in Figure 2. In this study rigid end blocks and plastic zone have zero length.

Rigid Block

a

i

Hinge Zone

d

Damper

c

Hinge Zone

Rigid Block

d

b

j

Figure 3. Three dimensional damper element with two plastic hinges at ends.

2.2 Damper element as energy dissipation device Figure 3 shows the viscous damper element proposed in the present study (Hejazi 2008). The first zone is the rigid block zone located at each end of the member. This element also consists of three different zones. The second zone is the 3D plastic hinge zone at each end assumed to have zero length. The remaining intermediate part of the member represents the third zone which is function of the viscous damper properties. These zones represent the finite

width of the damper joints, the inelastic and the elastic properties of the member. The central part of the member (located between two plastic hinges) is assumed to reflect the elastic behaviour of the member (elastic element), while the plastic hinge zones reflect the inelastic behaviour of the member (inelastic element). Damping coefficient (Cd ) for proposed 3-D damper element was calculated using the following relation:

26



L

Cd =

C¯ d (˙x)ψr (x)ψs (x)dx

3

(3)

0

where L is the member length, C¯ d is the damper damping constant coefficient, ψr (x) and ψs (x) is the shape functions as defined before. So, by expanding of equation (3) for three different zones of damper element, damping coefficient is now obtained in the following form: 

a

Cd = 0

In the present study, the stiffness method for structural analysis has been integrated with the finite element method to analyze a building system equipped with damper. Newmark’s direct step-by step integration is used for dynamic analysis. Predictor-corrector method for the solution of the resulting equation of motions in the time domain (Newmark 1959). The equation of motion for an elasto-plastic system equipped control system subjected earthquake load obtained from the consideration of equilibrium of forces is given by:

 c C¯ d (˙x)ψr (x)ψs (x)dx+ C¯ d (˙x)ψr (x)ψs (x)dx



+

0 b

C¯ d (˙x)ψr (x)ψs (x)dx

CONSTITUTIVE MODELING TIME MARCHING SCHEME FOR FRAMED STRUCTURES EQUIPPED DAMPER

(4)

0

M u¨ + q(u, u˙ ) = Fc + Fe

Similar to the beam-column element, a and b is length of two rigid block in both end of element and c is length of damper part as showed in Figure 3. If the relative velocity between two floors where damper installed intermediate of them in time (t) of earthquake excitation is denoted as x˙ (t), then a linear viscous damper force in corresponding time is calculated with this equation (Chart & Wong 2000): Fd (t) = Cd x˙ (t)

where q is the vector of internal resisting forces which depends upon the displacement u and velocity u˙ , M is the mass matrix of the system, u¨ is the acceleration vector, Fc is imposed control force and Fe is the applied earthquake load vector. The internal resisting forces are defined by the stiffness matrix K and damping matrix C and the control force due to viscous damper elements defined in pervious section.

(5)

where Fd is linear viscous damper force and it is used in equation of motion. With the availability of high-technology manufactured dampers, the structural engineer has the freedom of imposing additional damping in the structure by introducing manufactured dampers. Manufactured dampers that are used in buildings can produce forces that vary linearly with the relative velocities between the ends of the dampers. So the most common damper force design for structural engineering purposes is (Chart & Wong 2000): Fd (t) = Cd ˙x(t)η

(7)

4

DEVELOPMENT OF COMPUTER PROGRAM CODE

The existing finite element code developed by Thanoon (1993) has been extensively modified in view of the proposed physical and material models and adopted computational procedures for carrying out the 2D and 3D analysis of reinforce concrete framed buildings equipped viscous damper devices subjected to static and seismic/dynamic loads. The computer program has been written in Fortran language compatible with power station environment.

(6)

where η is velocity exponential coefficient or power law coefficient of damper and generally it is between 0 to 2. The first generation of manufactured viscous dampers used a power law coefficient equal to one. Structural engineers selected this value of η for design because for η equal to one the manufactured damping force, like the natural damping force, is a linear function of velocity. In the case of η =1 damper is called nonlinear damper and the force is not linear function of velocity. Therefore damper damping coefficient (C¯ d ) and power law coefficient (η) are two foremost effective parameters of damper force. So in the application of proposed damper in some model, effects of these parameters on structures response are evaluated.

5

CALIBRATION OF THE DEVELOPED PROGRAM CODE

In order to verify the accuracy of the developed finite element program code, an example has been analyzed. The verification proceed has been done by employing the commercial package SAP2000 software to analyze the same example. Figure 4 shows the structural and dynamic model for single storey reinforced concrete frame building, the material and section properties are also shown in same figure. The structure was subjected to an actual earthquake that occurred in Zanjiran-Iran (1985), as depicted in Figure 5.

27

3m Viscous Damper

5m

Columns Section

400 mm

400 mm

As1 = 1200 mm2 As2 = 800 mm2 As3 = 1200 mm2

Figure 6. Comparing the results of developed finite element program code and SAP2000 software (displacements of top node in X direction).

fc = 30 MPa Ec = 30000 MPa fy = 400 MPa Es = 200000 MPa

With out damper With damper

Beams Section

400 mm

300 mm

Figure 4. erties.

As1 = 1200 mm2 As2 = 1200 mm2

fc = 30 MPa Ec = 30000 MPa fy = 400 MPa Es = 200000 MPa

Single storey frame example and sections propFigure 7. Displacement of top node of single story model in X direction (C¯ d = 1300).

It is obvious from this plot that the displacements predicted by proposed model follows a similar trend obtained via commercial package SAP-2000. 6.1

The single story reinforced concrete structure shown in Figure 4 further has been analyzed for the cases of with and without supplemental viscous damper acting under seismic excitation. In this model damper damping coefficient (C¯ d ) and power law coefficient (η) are assumed equal to 1300 and 1 respectively. Figure 7 shows the typical time history response of tip deflection of the frame. It is clear from these plots that the magnitude of the displacement response of the structure to earthquake excitation has been decreased significantly, when the building structure is equipped with damper device compare to those without any energy dissipation system (uncontrolled structure). Considerable reduction in values of displacements is about 75% for the structure analyzed. Also time history displacement of same node in Y direction is shown in Figure 8. Similar to previous response, using damper in the modeled framed, is effectively reduced structure response in vertical direction though the displacement is very low. In article, it was seen that damping coefficient (C¯ d ) and power law coefficient (η) are two major parameters on damper force. So effects of

Figure 5. Zanjiran-IRAN (1985) earthquake acceleration record (m/Sec2 ).

6

Example: Single story framed building

PARAMETRIC STUDY ON EFFECT OF DAMPING

To evaluation, effect of application proposed damper on the structure and analysis of frame buildings by demonstrated computational strategies, a numerical examples have been selected and attempt has been made to analyze the models through developed finite element program code in various viscous damper properties. A damper element is used as a diagonal member as frame bracing system shape. Figure 6 shows the comparison of the horizontal displacement at the tip of the frame.

28

Displacement (mm)

0.0012

Without damper With damper

0.0008 0.0004 0 -0.0004 -0.0008 -0.0012 0

Figure 8.

1

2

3 Time (Sec)

4

5

6

Displacement of top node of single story model in Y direction (C¯ d = 1300).

these parameters on response of single story model are evaluated and results are showed in Figure 9. In this figure maximum displacement of top node in X direction during earthquake excitation for various damping coefficient (c = 0, c = 50, c = 100, . . ., c = 1300) and power law coefficient (η = 0, η = 0.2, η = 0.4, . . ., η = 2) is plotted. Increasing of damper damping coefficient is lead to enhance of damper damping matrix member’s values and finally increase the damper force. Then by increasing the damper force, more values of earthquake force is diminished by this force and therefore the structure response will reduce. As seen in plot, by increasing the damper damping coefficients, response of structure is reduced. When this coefficient is equal to zero, the damper force be come zero and response of structure is similar of model without damper element. The range of damping coefficient is different in various structures and it is depends on structures properties, imposed earthquake load density and buildings target performance levels. As seen in Figure 9 in this model, damping coefficient equal to 1300 is most effective on reducing of structures response against earthquake, therefore this coefficient value is suggest for this model. In the other hand, power law coefficient or damper velocity exponential coefficient, is define the effect of relative velocity between the ends of the dampers with damper force. If this coefficient is equal to one, the damper operation be come linear function of velocity and in other case, damper has nonlinear performance. Whatever this exponential coefficient is increased, effect of velocity on damper force will enhanced and, more part of earthquake energy will dissipate by damper device. As observed in figure, by increasing of power law, displacement is decreased. It is clear that in case of velocity exponential coefficient equal to 1.4 in range of damping coefficient equal to 500, damper performance is maximum in order to reduce displacement of structure. So these values are optimum values of

Figure 9. Comparison top node displacement of single story example in X direction with various damper velocity exponential coefficients (η) and damping coefficient (C¯ d ).

damper parameters to design effective passive control system to dissipate the earthquake energy and reduced structures seismic response. This type of damper devices can be designed and installed into both new buildings and existing structures. As they are relatively small and inconspicuous, they can be incorporated into a structure without compromising its appearance. So by notice to high performance of dampers device to reducing structural seismic response, this system is suitable technique for rehabilitation and retrofitting of exist buildings. Hence, as seen through explained example, by evaluation of damper parameters effect on the structures response, the structural design engineers are able to choose suitable damper properties for desire design of structure base on the request performance demand level of building and maximum effect of damper devices to diminish the seismic load.

7

CONCLUSIONS

This paper has reported development of special energy dissipation device for use as the seismic response control dampers for structures innovative from finite

29

element technique. A three dimensional viscous damper element applicable in reinforced concrete structures is proposed and formulated. For this propose a finite element program code for analysis of reinforced concrete structures equipped proposed damper elements is developed. The application of finite element code has been shown by analysis of one story reinforced concrete framed structures with supplemental viscous damper devices. Validation of developed program code is carry out by compare of results with SAP2000 software and it showed the accuracy and admissibility of program. By comparing seismic responses of structures without energy dissipation system, and structures with proposed viscous damper elements shows that using damper devices effectively reduced structural response subjected to earthquakes. (75% reduction for single story example). Also the optimum design of damper is eligible by evaluation of damper parameters effect on the structures response, and chooses suitable damper properties for desire design of structure base on the request performance demand level of building and maximum effect of damper devices to diminish the seismic load.

Chart, G.C. and Wong, K. 2002 ‘‘Structural Dynamic for Structureal Engineers’’. John Wiley & Sons. Jann N. Yang and Anil K. Agrawal. 2002. ‘‘Semi-active hybrid control systems for nonlinear buildings against near-field earthquakes’’. Journal of Engineering Structures.Vol. 24. p271–280. Martinez-Rodrigo, M. and Romero, M.L. 2003.‘‘An optimum retrofit strategy for moment resisting frames with nonlinear viscous dampers for seismic applications’’. Journal of Engineering Structures.Vol. 25. p913–925. Miodrag Sekulovic, Ratko Salatic, and Marija Nefovska. 2002. ‘‘Dynamic analysis of steel frames with flexible connections’’. Journal of Computers and Structures. Vol. 80. p935–955. Modi, V.J. Welt, F. and Irani, M.B. 1990. ‘‘On the suppression of vibrations using nutation dampers’’. Journal of Wind Engineering and Industrial Aerodynamics.Vol. 33. p273–282. Najib Bouaanani and Fei Ying Lu. 2008. ‘‘Assessment of potential-based fluid finite elements for seismic analysis of dam-reservoir systems’’. Journal of Computers and Structures. Article in press. Newmark, N. 1959. ‘‘A method of computation for structural dynamics’’. Journal of Eng Mech Division ASCE.Vol. 85. p67–94. Ri-Hui Zhang and Soong, T.T. 1992. ‘‘Seismic design of viscoelastic dampers for structural applications’’. Journal of Structural Engineering.Vol. 118. p1375–1392. Soda, S. 1996. ‘‘Role of viscous damping in nonlinear vibration of buildings exposed to intense ground motion’’. Journal of Wind Engineering and Industrial Aerodynamics.Vol. 59. p247–264. Semih S.Tezcan and Ozan Uluca. ‘‘Reduction of earthquake response of plane frame buildings by viscoelastic dampers’’. Journal of Engineering Structures.Vol. 25. 2003. p1755–1761. Waleed, A. Thanoon, D.K. Paul, M.S. Jaafar, and D.N. Trikha. 2004. ‘‘Influence of torsion on the inelastic response of three-dimensional r.c. frames’’. Journal of Finite Elements in Analysis and Design.Vol. 40. p611–628. Waleed, A.M. Thanoon. 1993. ‘‘Inelastic Dynamic Analysis Of Concrete Frames Under Non-Nuclear Blast Loading’’. PhD Thesis, University of Roorkee. India.

ACKNOWLEDGMENT The research was financially supported by Ministry of Science, Technology and Innovation of Malaysia under Research Project No. 5450366 and gratefully acknowledged. REFERENCES Constantinou, M.C. and Symans, M.D. 1992. ‘‘Experimental and Analytical Investigation of Seismic Response of Structures with Supplemental Fluid Viscous Dampers’’. Tech. Rep. NCEER-92-0027, National Center of Earthquake State Univ. of New York (SUNY) at Buffalo N.Y. Farzad Hejazi. 2009. ‘‘Optimization of Earthquake Energy Dissipation in Framed Buildings’’. PhD Thesis, University Putra Malaysia, Malaysia. (under process).

30

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Behavior of cylindrical R/C panel under combined axial and lateral load T. Hara Tokuyama College of Technology, Shunan, Japan

ABSTRACT: R/C cylindrical panels have been often used for the storage reservoir, tank, stack and cooling tower structures. They usually possess the high ultimate strength under lateral uniformly distribute loading. Therefore, they are used for the wall structure. However, when R/C cylindrical panels have heavy roof structure such as the underground LNG tanks or they are huge structures such as tall stacks or the huge cooling towers, they are subjected to combined axial compressive and lateral flexural loadings. In this paper, the deformation characteristics and the ultimate strength of R/C cylindrical panel under combined axial and flexural loadings are investigated numerically. In numerical analyses, FEM procedures are used based on the degenerate shell formulation. The numerical parameters are obtained from the experimental results and the numerical models are constructed based on the experimental specimens. The distribute or the concentrate loadings are considered as the flexural loading. INTRODUCTION

960

10

X

Y

Y Z

950 Figure 1.

X

R/C cylindrical panels have been often used for the storage reservoir, tank, stack and cooling tower structures. They usually possess the high ultimate strength under lateral uniformly distribute loading. Therefore, they are used for the wall structure in underground or the roof structure. However, when R/C cylindrical panels have heavy roof structure such as the underground LNG tanks or they are huge structures such as tall stacks or the huge cooling towers, they are subjected to combined axial compressive and lateral flexural loadings. From the previous paper (Hara 2008), the comparisons between the numerical and the experimental results concerning R/C cylindrical shell panel under the concentrate or the distribute lateral loading were presented. In the analyses, R/C cylindrical shell were pin-supported on both meridional and hoop edges. Both numerical and experimental results were well agreed. In this paper, the deformation characteristics and the ultimate strength of R/C cylindrical panel subjected to a combined axial and a flexural loadings are investigated numerically under the same geometric and supporting conditions. R/C panels are made by micro concrete and steel wire meshes. The specimens are 960 mm × 950 mm plan and the thickness of the panel is 10 mm. The span to depth ratio is 5 (see Figure 1). The panels contain a reinforcement mesh sheet in the middle of the shell panel. The reinforcement wire is 0.75 mm diameter and placed with equidistant spacing 5 mm in both meridional and hoop directions. The specimens are supported on both meridional and hoop edges by roller hinges. The deformations are evaluated at each

R=688.75 960

0.75@5

Z

1

190

Geometric dimensions of R/C shell (unit: mm).

loading step. The ultimate load is obtained by the peak loading on the load deflection curve. In numerical analyses, FEM schemes are used based on the degenerate shell formulation. The numerical parameters are obtained from the experimental results and the numerical models are constructed based on the experimental specimens. Load deflection characteristics and the ultimate strength are computed under several combinations of loadings. 2 2.1

DEFINITION OF THE SPECIMEN Geometric dimensions

R/C panel has the cylindrical shape with 960 mm × 960 mm plan and has 688.75 mm radius and 10 mm

31

thickness (see Figure 1). φ0.75 mm stainless wires are used as the reinforcements and are placed in the middle of the shell thickness in both meridional and hoop direction. They are placed in equi-distance 5 mm. The specimen is made by use of the steel mold to avoid the geometric imperfections. The micro concrete with aggregate size 2.5 mm is used. The material properties are shown in Table 1 and 2.

Two types of lateral loading conditions are considered. One is the concentrate load at the center of the shell plane. The other is the quasi-distribute load on the shell plane. The quasi-distribute load is applied as 64 concentrate loads, that are summarized to the one concentrate load via whiffle tree loading system (see Figure 2).

3 2.2

NUMERICAL PROCEDURE

Supporting and loading conditions 3.1

Specimens are pin-supported by steel ball-hinges of 11 mm diameter arranged at 20 mm spaces on both meridians. Also, both hoop edges are pin-supported by the same apparatus. Material properties of concrete.

Compressive strength (MPa) Tensile strength (MPa) Young’s modulus (GPa) Poisson’s ratio

Table 2.

38.2 3.8 23.6 0.20

240

240

120

240

240

Loading Point (Uniformly )

115

240

Loading Point (Concetrate)

Finite element

In numerical analysis, the geometric and material nonlinearlities are considered. 9 nodes Heterosis element is used and 2 × 2 reduced integration is adopted to avoid the numerical problems. The numerical simulation is performed under the displacement incremental scheme. The yield condition of concrete is defined as the Dracker-Prager type, which is assumed that concrete yields when the equivalent stress based on mean stress and second deviatoric stress invariants reaches uniaxial compressive strength (Hinton 1984). The crushing condition is contorolled by strain. The ultimate compressive strain of concrete is assumed as 0.003 by Kupfer’s experiment (Kupfer

235 449 206 21

115

240

3.2

Material properties of steel.

Yield stress (MPa) Tensile stress (MPa) Young’s modulus (GPa) Tangential modulus (GPa)

12 0

In numerical analyses, the finite element procedure is applied. Figure 3 shows the FE mesh of this analysis. The full model is adopted. The model is divided into 32 elements in meridional and hoop directions, respectively. Each element is divided into 8 concrete layers and 2 steel layers. Boundary conditions are pinsupported along both meridional and hoop edges. All rotations are free on all edges.

55 120 120 120 120 120 120 120 55 950

Table 1.

Numerical model

190

60 120 120 120 120 120 120 120 60 960 10

R=688.75 960 (Unit:mm)

:Loading Points Figure 2.

Quasi-uniformly distributed load.

Figure 3.

32

FE Mesh.

1969). Also, after cracking of concrete, the tension stiffenig parameters accounting for the tensile strength of concrete are introduced. The material nonlinearlities of steel are assumed to be bilinear stress-strain relation for the reinforcement.

4

NUMERICAL RESULTS

Two types of lateral loading conditions are considered. One is the concentrate load at shell center. The other is the quasi-uniformly distribute load of the shell plane. After applying the axial force in meridional direction, these two types of lateral loadings are applied, respectively. 4.1

Point A Figure 5.

Fundamental load bearing characteristics

4.1.1 R/C shell under axial compression Figure 4 represents the relation between an axial load and deformation in the direction perpendicular to the shell plane at the center of the shell. Figure 5 shows the deformations at loading stages A and B. Deformations appear around both hoop edges. Up to loading stage A, deformations appear only around both hoop edges as shown in Figure 5. After loading stage A, deformation mode of the shell changes. The deformation at loading stage B is also presented in Figure 5.

Point B

Deformation under axial compression.

Load(kN) 4

3

4.1.2 R/C shell under lateral concentrate load Figure 6 shows the load deformation relation under lateral concentrate load applied at the center of the shell.

2

1

Load(kN) 400

0

B 300

0.1

0.2

0.3

0.4

Displacement(cm) Figure 6.

Lateral concentrate load–displacement relation.

200

100

0

R/C shell is pin-supported at both hoop and meridional edges. Targeted displacement point is the same as the loading point. The deformation is in the direction perpendicular to the shell plane. Figure 7 represents the deformation mode under concentrate lateral load at the center. Large deformations appear around at the center of the shell. The deformation is the local phenomena along the meridional direction. R/C cylindrical shell surface deforms one and a half wave deformation in the hoop direction.

A

0.01

0.02

0.03

0.04

0.05

Displacement(cm) Figure 4.

Axial load–displacement relation.

33

Figure 7.

Deformation under lateral concentrate load. Figure 9. load.

Load(kN)

Deformation under lateral uniformly distribute

100

80

around the both hoop edges, there are steep deformations and the lateral load plays the predominant role for R/C shell deformation under the distribute loading.

60

4.3

To investigate the deformation characteristics and the ultimate strength of R/C shell under combined axial and lateral loadings, parametric study is performed. Figure 10 shows the relation between lateral concentrate load and deformation in the direction perpendicular to the shell plane at the center of the shell after applying the axial loadings. The applied axial load is the 20% of the ultimate strength represented in Figure 4. Figure 11 shows the total deformation. The deformation pattern is the same as that only under concentrate loading (see Figure 7). Therefore, the effect of the axial loading concerning the combined loading effect is small. To define the influences of the axial loadings into the ultimate strength under combined axial and lateral concentrate loadings, the ultimate strength is computed by use of the several axial loading levels as the parameter. Figure 12 shows the load-deformation relation under combined loading condition. Applied axial loadings are 10%, 30% and 50% of the ultimate axial loading. From Figure 12, the larger the applied axial loading level is, the smaller the ultimate lateral loadings. However, the ultimate deformation is larger with increasing the axial loading level. Figure 13 shows the relation between the ultimate strength and the applied axial compressive loading level under combined axial and lateral concentrate loadings. Figure shows the same tendencies as shown

40

20

0

0.1

0.2

0.3

0.4

0.5

Displacement(cm) Figure 8. relation.

4.2

R/C shell under combined axial force and lateral concentrate load

Lateral uniformly distribute load–displacement

R/C shell under lateral distribute load

Figure 8 shows the load-deformation relation under the lateral uniformly distribute load. Uniformly distribute load is applied as the series of concentrate load mentioned in the previous chapter. R/C shell is pinsupported at both hoop and meridional edges. The displacement in the direction perpendicular to the shell plane is evaluated at the center of the shell surface. Figure 9 represents the deformation mode under lateral uniformly distribute load. Large deformations appear along the central meridian of the shell. R/C cylindrical shell surface also deforms one and a half wave deformation in the hoop direction. Total deformation is the same as that under concentrate loading. However,

34

Load(kN)

Load(kN) 4

4

3

3

2

2

0.1P 0.3P 0.5P

1

1

0

0

0.1

0.2

0.3

0.1

0.2

0.4

0.3

0.4

Displacement(cm)

Displacement(cm) Figure 12. Load deformation relation of R/C shell under combined axial and lateral concentrate loadings.

Figure 10. Relation between combined 20% of axial ultimate load and lateral concentrate load–displacement at center.

Load(kN) 4

3

2

1

0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

(Axial Load)/(Compressive Strength)

Figure 11. Deformation under combined 20% axial ultimate load and concentrate load.

Figure 13. Ultimate load and the applied axial compressive load level under combined axial and lateral concentrate loadings.

in Figure 12. The applied axial force plays an important role for the ultimate strength of the shell under combined axial and lateral concentrate loadings.

after applying the axial loadings. The ordinate shows the total applied loading intensity to R/C shell surface. The applied axial load is also the 20% of the ultimate load represented in Figure 4. The ultimate strength under combined loading is larger than that under only lateral distribute loading.

4.3.1 R/C shell under combined axial force and lateral uniformly distribute load Figure 14 shows the relation between lateral uniformly distribute load and deformation in the direction perpendicular to the shell plane at the center of the shell

35

Load(kN)

Load(kN) 150

150

100

100

50

50

0.1P 0.3P 0.5P

0

0

0.1

0.2

0.3

0.1

0.2

0.4

0.3

0.4

Displacement(cm)

Displacement(cm)

Figure 16. Load deformation relation of R/C shell under combined axial and lateral distribute loadings.

Figure 14. Relation between combined 20% of axial ultimate load and lateral uniformly distribute load– displacement at center.

Load(kN) 150

100

50

0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

(Axial Load)/(Compressive Strength)

Figure 15. Deformation under combined 20% axial ultimate load and uniformly distribute load.

Figure 17. Ultimate load and the applied axial compressive load level under combined axial and lateral distribute loadings.

Figure 14 shows the total deformation. The deformation pattern is also the same as that only under uniformly distribute loading (see Figure 9). From the deformation analysis shown in Figure 5 under axial loading, R/C shell shows the convex deformation in upward. Consequently, R/C shell deforms to the bidirectional convex configuration. In such case, R/C shell represents the high ultimate strength under lateral uniformly distribute load. Therefore, the ultimate strength

increases comparing with that only under laterally distribute load. The effect of the axial load concerning the combined loading effect is large under lateral uniformly distribute load. To define the influences of the axial loadings into the ultimate strength under combined axial and lateral uniformly distribute loadings, the ultimate strength is

36

From the parametric investigations, following conclusions are obtained.

also computed under the same conditions as Figure 12. Figure 16 shows the load-deformation relation under combined axial and uniformly distribute loadings. Applied axial loadings are 10%, 30% and 50% of the ultimate axial loadings. The ultimate strength grows with the axial compressive load level. From Figure 16, the different results from that under combined axial and concentrate loadings are obtained. The larger the applied axial loading level is, the larger the ultimate lateral loading is. Also, the ultimate deformation is smaller with increasing the axial loading level. Figure 17 shows the relation between the ultimate strength and the applied axial compressive loading level under combined axial and lateral uniformly distribute loadings. From Figure 17, the increasing rate of the ultimate strength is small if the applied axial compression level exceeds 20% of axial ultimate load. Consequently, the applied axial force plays a predominate role for the ultimate strength of the shell under combined axial and lateral distribute loadings. 5

1. The deformation of R/C shell under concentrate loading contributes to the local deformation around the applied load. 2. The deformation of R/C shell under distribute loading contributes to the global deformation on the shell surface. 3. The larger the axial compressive load is, the smaller the ultimate strength of R/C shell is when the combined axial and concentrate lateral loadings are applied. 4. The larger the axial compressive load is, he larger the ultimate strength of R/C shell is when the combined axial and lateral distribute loadings are applied. However, the effect of axial load is almost the same if the axial load level exceeds 20% of the ultimate compressive strength. REFERENCES Hara, T. 2008. Numerical and experimental evaluation of R/C shell. Proceedings of International Conference on Advances in Structural Engineering and Mechanics, 133–144. Hinton, E. and Owen, D.J.R., 1984, Finite element software for plates and shells, Prineridge Press, Swansea, UK. Kupfer, H. and Hilsdorf, K.H., 1969, Behavior of concrete under biaxial stress, ACI Journal 66(8) 656–666.

CONCLUSIONS

In this paper, the deformation characteristics and the ultimate strength of R/C cylindrical panel subjected to a combined axial and a flexural loadings are investigated numerically under the same geometric and supporting conditions as the previous researches.

37

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Calculation method research on the flexural capacity of PSRC beam S. Qin, Y. Wang & F. Li College of Civil Engineering, Chongqing University, Chongqing, China

Z. Ding ZheJiang Academy of Building Research, Hang Zhou, China

ABSTRACT: This experiment is carried out for the purpose of finding out a proper method that can be used to calculate the flexural capacity of normal section of PSRC beam. According to the test results of 6 experimental beams and two of China’s prevailing occupation standard on this field, a new method is put forward, which is specially used to calculate the flexural capacity of this beam. By using comparison analyses, the following is clear: (a) Simple superposition method is only suited for special circumstance that is the steel must be located symmetry along the height of the cross-section. However, the error is notable while be used to calculate SRC (steel reinforced concrete) beam. (b) Coordinate analytical method based on plane cross-section assumption has better applicability and calculation accuracy on both PSRC beam and SRC beam. 1

prestress, the simple additive method and coordinate analytical method based on the current two regulations of China were put forward to calculate the flexural capacity of such beam. In addition, evaluations upon rationality and applicability of these two methods were completed after comparing the results of experiment and theoretical calculation.

PREFACE

Numerous researches on Steel Reinforced Concrete (SRC) structures have been done world widely, among which some of them were completed in China. As for the calculation theory of SRC structure, it is now basically a well-developed system, with three major calculation theories in use world widely, as follows: a. The stiffness discounted method based on steel structure design, with a considering of rigidity reducing caused by outer concrete, is mainly used in the United States. b. The ultimate state method based on concrete structure design is mainly used by the former Soviet Union. According to this method, the flexural capacity of SRC members is calculated analogous to reinforced concrete, but considering the influence of stress distribution of steel ribs. This method is also adopted in the Steel Reinforced Concrete Structure Technical Regulations of China. c. The additive method, mainly used in Japan, is to overlap the flexural capacities of concrete and steel together as the final flexural capacities of SRC members. This method can also be seen in Steel Reinforced Concrete Design Codes of China.

2

SIMPLE ADDITIVE METHOD

Now borrowing ideas from the ninth reference, regarding that the PSRC beam which is consist of steel ribs (S) and prestressed reinforced concrete (PRC), thus the flexural capacity of PSRC beam can be acquired by adding the two part’s flexural capacities together. The calculation formula is given by prc

ss M ≤ Mbu + Mby

where M = design value of bending moment at prc ss section; Mby = flexural capacity of steel ribs; Mbu = flexural capacity of PRC department. If the earthquake action dose not counts in, the flexural capacity of the steel is calculated by ss Mby = γs fss Wss

As for flexural capacity of Prestressed Steel Reinforced Concrete (PSRC) beams, few researches have been done world widely; additionally, the current two regulations of China have different viewpoints on the collaboration of concrete and steel. For reasons above, experiment is carried out. From the test results of 6 PSRC beams, and considering the influence of

(1)

(2)

where Wss = elastic resistance moment of cross section of steel; γs = plastic coefficient of cross section, which is used to lower the calculation error caused by the difference between pure bending model and real eccentrically tension situation of steel section

39

(for I-shape steel γs = 1.05); fss = design strength of steel ribs under tension, compression, or bending The flexural capacity of PRC department is represented by the equation prc

Mbu = γ hb0 · (As fsy + Ap fp )

d. (3)

where As = area of re-bar under tension; AP = area of prestressed reinforcement; fsy = design value of tensile strength of re-bar; fP = design value of tensile strength of prestressed reinforcement; hb0 = the distance from the action spot of resultant tensile force of reinforcement (including re-bar and prestressed reinforcement) to the outer boundary of compressive region; γ hb0 = the distance from the action spot of resultant tensile force to the action spot of resultant compressive force; γ = coefficient of internal force arm. According to document 7, γ = 0.875. SRC beams can be considered as PSRC beams with a prestressed degree equal 0; hence the simple additive method is suit for SRC beams also. That is to say, with a universal applicability. 3

e. f. g. h.

COORDINATE ANALYTICAL METHOD

the depth of compression zone obtained by the plane cross-section assumption. The corresponding maximum compressive stress is equal to the compressive strength fc . For rebar and steel rib, ideal elastoplastic constitutive relation of stress-strain (σ –ε) is adopted. According to full plastic assumption, the stress distribution of the web of steel rib can be simplified to two rectangular stress distribution models, one in tensile, the other in compressive. When at ultimate state, because of the thin thickness and the short force arm of the web, the contribution of it to the total flexural capability is small, thus the tensile region of the web can be considered as a full yield region. Tensile strength (fpy ) of prestressed reinforcement being arrived while at ultimate state. The tensile strength of concrete is ignored. Local buckling of steel rib does not occur. In terms of the assumption above, the distribution of strains, actual stresses, and simplified stresses in concrete and steel rib over the depth of the section at ultimate state is as shown in Fig. 1.

3.3

3.1 Failure morphology

The flexural capacity of normal section of solid-web PSRC beams

3.3.1 Judgment of stress state at ultimate state At ultimate state the relative depth of compression zone ξ is given by

Numerous researches shows that, no matter SRC beams or PSRC beams as long as the cross-section well designed, and enough shear keys being putted between the steel’s compressive flange and concrete, the over reinforced failure of PSRC beam seldom occurs while at ultimate state of failure. The failure morphology is similar to a well-designed RC beam that is, the failure occurs by yielding of tensile reinforcement (including re-bar, prestressed reinforcement, flange and partial web of steel ribs), not by crushing of concrete.

ξ = x/h0

(4)

where x = height of equivalent rectangular stress diagram of compressive region of concrete; h0 = the distance from the outer boundary of compressive region to the action spot of resultant tensile force. These tensile forces including the force in re-bar, prestressed reinforcement, and tensile flange of steel rib, however the force in tensile web is ignored for little contribution to the total flexural capacity. To determine the exact stress status of normal section at ultimate state, the critical relative depth ξi of compression zone of the following several situations need to be understood first:

3.2 Fundamental assumption For rectangular PSRC beam with solid-web steel encased, the following hypotheses must be obeyed when calculate the flexural capability of this beam: a. Each material on the cross section obeys the plane cross-section assumption. Data in the sixth document demonstrate that, before the bottom flange of steel yielding, this assumption applies well. Even if the bottom flange of steel yielded, for little growth of flexural capability generated after that, hence the plane cross-section assumption can also be used in analysis, and without a significant error. b. The ultimate compression strain of concrete at compressive region is given by εcu = 0.003. c. At ultimate state, the distribution of compressive stress at compressive region is assumed as a rectangular distribution. The equivalent height of this compressive region can be expressed by β1 times

a. At this condition, the strain of top steel flange is exactly equal to the tensile yielding strain when the concrete strain in outer compression fiber simultaneously reaches the crushing strain εcu . Thus, the critical relative depth ξ1 of compression zone can be determined, as follow: ξ1 βh01 aa − ξ1 βh01

=

(aa − ξ1 βh01 ) · εcu εcu ⇒ ξ1 = · β1 εay h0 · εay (5)

40

Strain Figure 1.

Actual stress state

Simplified stress state

Stress and strain of normal section at ultimate state.

where β1 = the ratio of height of equivalent rectangular stress block to the depth of neutral axis. If the strength grade of concrete is lower than C50, in this case β1 = 0.8; If the strength grade of concrete equal C80, then β1 = 0.74; Else if the strength grade is just between C50 and C80, the value of β1 can be calculated by linear interpolation method. b. The stress of top steel flange is exactly equal to zero when the concrete strain in outer compression fiber simultaneously reaches the crushing strain εcu . According to plane cross-section assumption, the critical relative depth ξ2 of compression zone can be derived from ξ2 h0 /β1 = aa ⇒ ξ2 = aa /h0 · β1

strain εcu . Thus, according to plane cross-section assumption, the critical relative depth ξ5 of compression zone can be derived from  ξ 5 h0 β 1 εcu εcu · (h − as) · β1 = ⇒ ξ5 = h − as εcu + εsy h0 · (εcu + εsy) (9) f. At this condition, the strain of prestressed reinforcement is exactly equal to the tensile yielding strain when the concrete strain in outer compression fiber simultaneously reaches the crushing strain εcu . Thus, according to plane cross-section assumption, the critical relative depth ξ6 of compression zone can be derived from  ξ 6 h0 β 1 εcu · (h − ap ) εcu = ⇒ ξ6 = · β1 h − ap εcu + εp1 h0 · (εcu + εp1 ) (10)

(6)

c. At this condition, the strain of top steel flange is exactly equal to the compressive yielding strain when the concrete strain in outer compression fiber simultaneously reaches the crushing strain εcu . Thus, according to plane cross-section assumption, the critical relative depth ξ3 of compression zone can be derived from  ξ3 h0 β1 εcu aa εcu  = ⇒ ξ = · β1 3   ) εay h0 · (εcu − εay ξ3 h0 β1 0 − aa (7)

 , ε , ε = they are strains of the top comwhere εay ay sy pressive flange of steel, the bottom tensile flange of steel, and the bottom longitudinal tensile re-bar. εp1 = strain increment of prestressed reinforcement from the state of none stress exist in concrete near the action spot of prestressing tendon, to the state of yielding of the this tendon. By sorting these ξi above, and comparing them with the actual ξ , the real stress distribution of each component at ultimate state can be known clearly.

d. At this condition, the strain of bottom steel flange is exactly equal to the tensile yielding strain when the concrete strain in outer compression fiber simultaneously reaches the crushing strain εcu . Thus, according to plane cross-section assumption, the critical relative depth ξ4 of compression zone can be derived from  ξ4 h0 β1 εcu εcu · (h − aa) · β1 = ⇒ ξ4 = h − aa εcu + εay h0 · (εcu + εay) (8)

3.3.2 Critical depth of relative compression zone (ξb ) The constituents of PSRC members is more complex than usual RC members; also the position and relative stress value of tensile re-bar, steel rib, and prestressed reinforcement are uncertain. It is concluded that the way to ascertain ξb only by one material is inapplicable. As known to all, the concrete in compressive region reaches its ultimate compressive strain εcu dose not means all of the prestressed reinforcement, tensile re-bar, and the bottom flange of steel yield at the same time. By analysis each ξi above, it is cleared that, under the ξ1 and ξ2 situation, all of the prestressed reinforcement, tensile re-bar, and the bottom flange of

e. At this condition, the strain of bottom longitudinal steel bar is exactly equal to the tensile yielding strain when the concrete strain in outer compression fiber simultaneously reaches the crushing

41

steel yield while the concrete in compressive region reaches its ultimate compressive strain εcu . Based on the discussion above, the value of ξ3 , ξ4 , ξ5 or ξ6 is relative lager than ξ1 or ξ2. Therefore, to assure all the tensile reinforcements yielding while the beam is at the critical failure state, ξb is given by ξb = Min(ξ3, ξ4 , ξ5 , ξ6 )

3.3.3 Depth of compression zone of concrete At ultimate state, three conditions about the location of neutral axis in cross section are sorted. They are the neutral axis dose not go through the steel, the neutral axis just through the top flange of the steel, and the neutral axis through the web of steel. The condition that the neutral axis through the top flange of the steel is thought as a criterion to distinguish the other two as mentioned above, and the simplified state of stress and strain under this condition is shown in Figure 2. The actual depth (c) of compression zone of concrete can be determined by the following trial method. First the assumption that the neutral axis just going through the top flange of steel is adopted, that is to say ξaf = ξ2 . Then compare of the total tensile force Taf and the total compressive force Paf at the same cross-section, as follows:

(11)

The above method to ascertain ξb is precise and conservative enough, but the process is more complex. Hence the simplified method based on current specifications and codes of China is put forward. that is to divide the PSRC beam into two parts, the PRC department and the SRC department, and calculate the ξb of each part separately, then take the smaller one as the PSRC beam’s ξb at last. The ξb calculated in this method is relatively small, so the calculated flexural capability is more or less smaller than actual flexural capability. According to the third reference, the ξb of the PRC department is given by ξpb =

β1 1+

0.002 εcu

+

1. Taf = Paf , then x = β1 c. 2. Taf > Paf , it is means the actual depth of compression zone of concrete is deeper than the assumed one, then another assumption ξ = ξ3 is adopted, and once again to compare the total tensile force T and the total compressive force P across the section.

(12)

fpy −σpo Ep εcu

– if T = P, in this case x = β1 c. – if T < P, it is means the actual depth of compression zone of concrete is shallower than the assumed one, thus x can be calculated according to plane cross-section assumption and equilibrium condition of the section. – else if T > P, it is means the actual depth of compression zone of concrete is deeper than the assumed one, or the top flange of steel have already yielded, thus x can be calculated by equilibrium condition of the section.

where σpo = stress of prestressed reinforcement at the moment of zero normal stress state of concrete next to the prestressed reinforcement; EP = elasticity modulus of prestressed reinforcement. The value of ξpb can also be taken referring to the Highway Reinforced Concrete and Prestressed Concrete Bridges Design Specification of China, as seen in Table1. If different kinds of re-bar are located at the tensile region, the value of ξpb equals to the minimum one in Table 1 below. Referring to the ninth reference, and considering the compatibility of it to the third reference, β1 is introduced. Then

3. Taf < Paf , it is means the actual depth of compression zone of concrete is too shallow that the actual neutral axis dose not reach the top flange of steel rib, then the assumption ξ = ξ1 is adopted.

(13)

Correspondingly, compare the total tensile force T and the total compressive force P across the section.

where fay = design value of tensile strength of steel rib; ES = elasticity modulus of rebars.

– if T = P, in this case x = β1c. – if T < P, it is means the actual depth of compression zone of concrete is shallower than the assumed one, the total cross section of steel have already

ξsb =

β1 1+

Table 1.

fsy + fay 2 × Es εcu

Relative compressive area depth at limit state.

Types of steel reinforcement

Concrete grade

C50 and below

C55/C60

C65/C70

C75/C80

HPB235 HRB335 HRB400, RRB400 Strand, steel wire Planished concrete reinforcing bar

0.62 0.56 0.53 0.40 0.40

0.60 0.54 0.51 0.38 0.38

0.58 0.52 0.49 0.36 0.36

– – – 0.35 –

42

Strain

Stress

Stress in web

Figure 2. Stress and strain of cross section with neutral axis through top flange of steel.

Figure 3. Stress and strain of cross section when neutral axis through the web.

yielded, and x can be calculated by equilibrium condition of the section. – else if T > P, it is means the actual depth of compression zone of concrete is deeper than the assumed one, the top flange of steel is in tensile state without yielding, and x can be calculated according to plane cross-section assumption and equilibrium condition of the section.

Strain

For situation III (see Fig. 4), taking moment about the top flange of steel gives Mu = fc bx · (aa − x/2) + fs As · (aa − as ) + fa Aaw hw /2 + fs As · (h − aa − as )

3.3.4 Flexural capacity Mu For convenience, the height of steel web hw is replaced by the distance between the center lines of top flange and bottom flange, and the compression area of concrete is calculated without deduction of the area of steel rib. The stress of the top flange can be expressed by σaf which may be tensile or compressive, yielding or not. For situation I, the neutral axis just goes through the top flange of steel (see Fig. 2), taking moment about the top flange of steel gives

+ fa Aaf hw + fpy Ap · (h −

(16)

+ fa Aaf hw + fpy Ap · (h − aa − ap )

4

COMPARISON BETWEEN THEORETICAL CALCULATION AND TEST RESULTS

In order to understand the force performance of PSRC beams and find out the design method of flexural capability of it, three groups of 6 simple beams were researched. The 6 specimens are consisting of 2 SRC beams and 4 PSRC beams. The prestressed reinforcement is composed of two bunches of prestressed tendon with six 5 mm high-strength steel wire in each bunch. Steel rib adopted is Q235 hot rolled steel H-beams (HN200 ×100 × 5.5 × 8), as shown in Figure 5. Comparison of theoretical calculation and test results is shown in Table 2. According to the data in Table 2 the following is clear: (a) Simple superposition method is only suit for special circumstance that is the steel must be located symmetry along the height of the cross-section. However, the error is notable while be used to calculate SRC (steel reinforced concrete) beam. (b) Coordinate analytical method based on plane cross-section assumption has better applicability and calculation accuracy on both PSRC beam and SRC beam. It is found that the calculated result using coordination analytical method not only close to the measured value, but also conservative enough.

Mu = fc bx · (aa − x/2) + fs As · (aa − as ) aa

Stress in web

Figure 4. Stress and strain of cross section when neutral axis dose not pass the steel skeleton.

In addition, x must satisfy the ξ < ξb condition so as to make sure the ductile fracture appearing while at ultimate state, that is the failure initiates by yielding of rebar, prestressed reinforcement, and the bottom flange of steel yielding at ultimate state. If this condition can not be satisfied, adjustment of the sectional dimension or concrete grade is suggested to meet this requirement.

+ fa Aaw hw /2 + fs As · (h − aa − as )

Stress

(14)

− ap )

For situation II (see Fig. 3), taking moment about the neutral axis gives Mu = fc bx · (x/β1 − x/2) + fs As · (x/β1 − as ) + σaf Aaf · (x/β1 − aa ) + 0.5σaf tw · (x/β1 − aa )2 + 0.5fa tw · (hw − x/β1 + aa )2 + fa Aaf · (hw − x/β1 + aa ) + fs As · (h − as − x/β1 ) + fpy Ap · (h − ap − x/β1 ) (15)

43

Figure 5. Table 2.

Dimensions and reinforcements of specimens. Test values and theory values of the carrying capacity for the specimens (kn·m).

Beam number

SB-1

PSB-1a

PSB-1b

SB-2

PSB-2a

PSB-4b

Type of test beam Degree of prestress Measured value of Mu (Simple superposition method) Mu1 (Coordination analytical method) Mu2

SRC1 0 151.90 112.90 (25.67%) 148.54 (2.21%)

PSRC1 0.160 200.90 176.22 (12.28%) 198.57 (1.16%)

PSRC1 0.173 199.90 175.58 (12.17%) 198.70 (0.6%)

SRC2 0 240.10 125.98 (47.53%) 239.20 (0.37%)

PSRC2 0.160 313.60 212.88 (32.12%) 300.32 (4.23%)

PSRC2 0.173 312.60 212.02 (32.18%) 298.54 (4.50%)

Note: 1) Superscript 1 means that the steel is symmetrical in layout while subscript 2 means unsymmetrical layout of steel. 2) Relative error between computed value and measured value of bending moment is shown in the parentheses.

5

CONCLUSIONS

University of Architecture & Technology, vol. 31, no. 3, 1999, pp. 211–214. CN3-78 Reinforced Concrete Structural Design Guidelines of the Soviet Union, Peking: China Architecture & Building Press, 2000. Code for Design of Concrete Structures, National standards of People’s Republic of China, GB50010-2002, Peking: China Architecture & Building Press, 2002. Explanation of Steel reinforced concrete structure and the standard, Architectural Institute of Japan, 1998. Highway Reinforced Concrete and Prestressed Concrete Bridge Design Specifications, Industry Standards of China, JTG D62-2004, Peking: China Communications Press, 2005. Li Feng, Qin Shihong, and Ding Zhichao, ‘‘Calculation Method Research on the Flexural Performance of Prestressed Steel Reinforced Concrete Beam,’’ J. Journal of harbin institute of technology, Harbin, 2007. Li Shaoquan and Sha Zhenping, ‘‘Superposition Method for Calculating Bearing Capacity on Normal Section of Beam with Steel Skeleton and Bars Reinforced Concrete,’’ J. China Civil Engineering Journal, vol. 37, no. 7, 2004, pp. 1–5. Steel Reinforced Concrete Design Codes, Industry Standards of China, YB9082-97, Peking: Science Press (in Chinese). 1998. Steel Reinforced Concrete Structure Technical Regulations, Industry Standards of China, JGJ138-2001, Peking: China Architecture & Building Press, 2001. Ye Lieping and Fang Ehua, ‘‘State-of-the-art of Study on the Behaviors of Steel Reinforced Concrete Structure,’’ J. China Civil Engineering Journal, vol. 33, no. 5, 2000, pp. 1–11. Zhao Hongtie, Steel and Concrete combinatorial structures, Peking: Science Press, 2001.

a. Simple superposition method making use of the concept in China’s Steel Reinforced Concrete Design Codes is a simple calculation, suit for manual computation with a higher accuracy for PSRC beam than SRC beam. However, it is only suit for special circumstance that is the steel must be located symmetry along the height of the crosssection; furthermore, the result is conservative excessively. Because of these reasons, this method can be used for scheme selection and preliminary design just in terms of its characters. b. Coordinate analytical method with tight logical derivation, combines China’s current specifications and codes well. Hence, not only this method can be used for all kinds of arrangement of steel rib, but also the calculated value is close to the experimental result. However the computing process is complex. Therefore, this method is quite suit for computer calculation. c. Both of the two methods have universal applicability for SRC and PSRC structure. Not only rectangular section but also I-shaped and T-shaped sections can be calculated by the two methods REFERENCES Bai Guoliang and Zhao Hongtie, ‘‘Calculating Method of Ultimate Moment Capacities on Full Web Type Steel Reinforced Concrete Beams,’’ J. Journal of Xi’an

44

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Cyclic loading deterioration effect in RC moment frames in pushover analysis G. Ghodrati Amiri Center of Excellence for Fundamental Studies in Structural Engineering, School of Civil Engineering, Iran University of Science & Technology, Tehran, Iran

B. Mohebi & S.A. Razavian Amrei School of Civil Engineering, Iran University of Science & Technology, Tehran, Iran

ABSTRACT: The aim of this research is to investigate effect of hysteresis loops in static nonlinear analysis. One of the inefficiencies of static nonlinear analysis is that nonlinear behavior of structural elements due to cyclic deformations is approximately considered in the analysis, and only one quarter of a full hysteretic loop is considered. For investigating the effect of this inefficiency in analysis results, three intermediate concrete moment frames are selected. These models are selected from regular RC structures. The notified procedures in FEMA-356 and proposed plastic hinges in this guideline are utilized for performing static nonlinear analysis. A coefficient for consideration of stiffness degradation and strength deterioration is proposed by FEMA-356 in nonlinear static analysis. This coefficient for intermediate RC moment frames is equal to unity. For calculation of this coefficient, in this paper, the nonlinear dynamic analysis is used. Clough and Takeda Hysteretic loops and a hysteretic loop that considers effects of severe stiffness degradation, strength deterioration and pinching are assumed in nonlinear dynamic analysis. By comparison of results the value of this coefficient is obtained 25% more than the value proposed by FEMA-356.

1

demands. In fact, some seismic codes have begun to include them to aid in performance assessment of structural systems (IIEES 2002). Although seismic demands are best estimated using nonlinear timehistory (NTH) analyses, NSPs are frequently used in ordinary engineering applications to avoid the intrinsic complexity and additional computational effort required by the former. As a result, simplified NSPs recommended in ATC-40 (ATC 1996) and FEMA-356 (FEMA 2000) have become popular. These procedures are based on monotonically increasing predefined load patterns until some target displacement is achieved. In the implementation of pushover analysis, modeling is one of the important steps. The model must consider nonlinear behavior of structural elements. Such a model requires the determination of the nonlinear properties of each component in the structure that are quantified by strength and deformation capacities. Lumped plasticity idealization is a commonly used approach in models for deformation capacity estimates. The real plastic behavior of a member during an earthquake is a cyclic behavior and after one cycle of loading and unloading, properties of the curve such as strength and stiffness is changed. Because in the nonlinear static analysis the applied load is not cyclic and is an incremental load and it is in the specified direction, therefore some kind of modeling

INTRODUCTION

Since inelastic behavior is intended in most structures subjected to infrequent earthquake loading, the use of nonlinear analyses is essential to capture behavior of structures under seismic effects. A set of earthquake records which have been selected accurately, can give a detailed evaluation of expected behavior and seismic performance of structures. With accepting this fact that the suitability and accuracy of calculation instruments are being increased, but there are some restriction for using nonlinear dynamic analysis. Some of these restrictions are the difficulty of interpreting the results and sensitivity of results to selection of earthquake records so that this problems, make using this kind of analysis very difficult for practical purposes. Therefore due to its simplicity, the structural engineering profession has been using the nonlinear static procedure (NSP) or pushover analysis, described in FEMA-356 (FEMA 2000) and ATC-40 (ATC 1996). It is widely accepted that, when pushover analysis is used carefully, it provides useful information that cannot be obtained by linear static or dynamic analysis procedures. In performance assessment and design verification of building structures, approximate nonlinear static procedures (NSPs) are becoming commonplace in engineering practice to estimate seismic

45

is needed, that can give the appropriate estimation of cyclic behavior of member. Also it should consider the effects of strength deterioration and stiffness degradation. For achieving this purpose, as shown in Figure 1, the backbone curve of actual hysteretic behavior of a member is used and it has been idealized to curve shown in Figure 2. The specifications of these idealized curves have been explained in some standards and pre-standards like FEMA356 (FEMA 2000). After performing the analysis, it is needed to evaluate the demand of structure. This evaluation can be done by using capacity spectrum method or constant coefficients method as described in ATC40 (ATC 1996) and FEMA356 (FEMA 2000) respectively. In the second method (constant coefficients) the demand of structure is defined by target displacement parameter. For calculating this parameter some coefficients are used and one of these coefficients is for taking into account the amount of strength deterioration and stiffness degradation. This coefficient in FEMA356 (FEMA 2000) for intermediate RC frames is equal to one. But this coefficient can be different depend on the members specifications and their hysteretic behavior. In this paper, it has been tried to calculate amount of effect of hysteresis loops and strength deterioration and stiffness degradation in estimating target displacement by using nonlinear dynamic analysis with different hysteresis loops for members and nonlinear static method with standard plastic hinges.

For performing nonlinear dynamic analyses computer program, IDARC (Kunnath et al. 1992) has been used. Also for performing nonlinear static analyses computer program ETABS (Computers & Structures 2005) has been used.

2

Nonlinear static analysis is based on a principle which says structure’s response can be simulated with a system which has one degree of freedom with equivalent characteristics. Based on this theory, structure’s response is just related to first vibration mode and its shape doesn’t change during analysis period. Although both of these theories seems to be wrong, but accurate estimation from system’s maximum reflection can be obtained for those structures which their first vibration mode is dominant. The aim of nonlinear static analysis is to evaluate expected behaviors of structural system by estimating the resistance and displacement demand under designed earthquake and comparing the requirements with existing capacities in selected performance level. Nonlinear static analysis (pushover) is a method to estimate the force and displacement demand which it simply does load redistribution for internal forces in members which tolerate more than their elastic limited forces. Specifying force-displacement curve (capacity curve) is one of its most important results. Amount of base shear against lateral displacement of the reference point which is in the roof, are used to draw the curve. It’s used for specifying the target displacement. 2.1

Figure 1.

Backbone curve of hysteresis behavior.

Figure 2.

Idealized capacity curve.

NONLINEAR STATIC METHOD

Lateral load distribution

Using adequate load distribution figure in evaluation of behavior of a building is one of important steps. In fact, the shape of lateral loading presents the distribution of inertia forces in designed earthquake. It is clear that lateral forces change because of intensity of earthquake and duration of earth’s excitation time. Contribution of inertia forces during earthquake is constant and maximum displacement can be compared with what is occurred in designed earthquake as if one loading shape is used. These assumptions are sometimes accurate or not. They are accurate when structure’s responses are not mostly affected by upper vibration modes and it has only one kind of yielding mechanism. Some case of lateral loading shapes gives accurate approximates for displacement demands. Using two kind of lateral loading shapes are recommended by researchers in some situations. Applying uniform loads which floor’s forces are proportionate with its mass and the shape of load contribution similar to different modes is widely accepted by them.

46

Figure 4.

Figure 3.

3

Trilinear curve (Park et al. 1987).

Building plan and selected frame for analyze.

DESCRIPTION OF SAMPLES

Three RC frames with 4, 8 and 12 floors are used for this research. These frames have been designed based on intermediate ductility according ACI318-05. (ACI 2005) They are in high seismic risk areas with PGA = 0.35 g (BHRC 2005) and soil is in type II, similar to group C in FEMA-356. Structures’ plan and the separated frame are shown in Figure 3. It is supposed that concrete compressive strength is 24 Mpa and steel’s yield stress is 400 Mpa. Floor’s dead load is 700 Kg/m2 and live load is 200 Kg/m2 . Live load participating percentage in earthquake force calculation is 20% for residential using (BHRC 2005).

4

HYSTERETIC BEHAVIOR MODELING

Hysteresis model which is used for showing nonlinear behavior of members is in company with three parametric models for mentioning stiffness degradation, strength deterioration and pinching (Park et al. 1987). Push curve of force-displacement is shown as a trifurcation curve in Figure 4. As it is shown, points consist of cracking point, yielding point and ultimate strength point. For showing the effect of stiffness degradation, strength deterioration and pinching behavior, α, β and γ parameters are used. Parameter’s concepts and their effects have been shown in Figure 5. Stiffness degradation which is represented by α is introduced by selecting a common point on protraction of first skeletal curve which is aiming the unloading lines to the horizontal axis. This parameter specified the stiffness degradation degree and mentioning the reduction of confined area of hysteresis loops. Strength deterioration is represented by β, which shows reduction rate of strength. This parameter is defined in a ratio of

Figure 5. Definition of parameters in 3 parameters hysteresis loops (Park et al. 1987).

addition of maximum response damage, dδm /δu , and addition of normalized hysteresis energy, dE/δu Py . β=

dδm /δu dδm = dE/(δu Py ) dE/Py

(1)

pinching behavior is represented by γ . According to Figure 5, by lowering the maximum point to the py

47

level, γ is defined in protraction of previous unloading line. Renewed loading lines aim the new target point to reach the displacement of crack closure. Pinching behavior causes the reduction of hysteresis loop’s area and reduction of energy loss indirectly. By using three parametric models, hysteresis models can be recycled such as, Clough (Clough et al. 1965) which doesn’t contain strength and stiffness reduction or Takeda (Takeda et al. 1970) that contains the effects of stiffness and strength reduction. Three hysteresis loops have been used in this research. Clough, Takeda and a model which its stiffness degradation, strength deterioration and pinching behavior parameters are maximum amount of normal time, have been used. 5

Figure 6.

have been multiplied to ratio of two mentioned areas. Finally all records multiplied to 0.35 as a PGA of site. The records specifications are presented in Table 1.

INPUT RECORDS

Seven earthquake records have been used for nonlinear dynamic analysis which the distance between them and the fault is 10–24 km. They are selected from soil type group (II) (BHRC 2005), which the effect of closeness to the fault and soft soil don’t exist. For using these records, their specification should be compatible with site. So they have to be scaled in a way with their responses are mostly compatible with real earthquakes which is occurred in the site. 5.1

Newmark method (Bathel 1996).

6 6.1

U˙ t+t = U˙ t + [(1 − δ)U¨ t + δ U¨ t+t ]t

For comparing effect of deferent earthquake records on structure, they should be scaled by a unique criterion. Earthquake records which used in this research have been scaled to 0.35 g. The method of scaling record is describing in this section. All records have been scaled to their maximum PGA. Therefore their PGA will be equal to 1 g. The response spectrum of each scaled record has been specified with considering 5% damping ratio. The area under response spectrum curve between periods 0.1 and 3 second has been calculated and compared with the same area in the standard reflection curve (Sa /PGA) in Iranian code. The records

Ut+t = Ut + U˙ t t +

Real Scaled Year PGA (g) PGA (g)

1 2 3 4 5 6 7

1995 1999 1989 1994 1985 1992 1987

0.821 0.376 0.479 0.514 0.594 0.385 0.377

  1 − α U¨ t + α U¨ t+t t 2 2 (3)

Selected method for dynamic analysis

Inelastic nonlinear dynamic analysis of RC moment frames has been done by computer program, IDARC (Kunnath et al. 1992). Moment-curvature curves have been used for presenting the nonlinear behavior of members which made by using concrete and steel stress-strain curves, member’s dimension, and reinforcement and confinement limitation. Newmark method is also used for analyzing, and hysteresis models, which mentioned before, are used as the members’ hysteresis behavior. Nonlinear dynamic analysis and nonlinear static analysis results are shown in Table 2.

Specifications of used records.

Number Records’ name



(2)

where α and δ are parameters that can be determined to obtain integration accuracy and stability. Newmark originally proposed as an unconditionally stable scheme the constant average acceleration method (also called trapezoidal rule), in which case δ = 1/2 and α = 1/4 (Figure 6). 6.2

KOBE KOCAELI, TURKEY LOMA PRIETA NORTHRIDGE N. PALM SPRINGS CAPEMENDOCINO SUPERSTITI HILLS

Newmark method for dynamic analysis

The Newmark integration scheme can also be understood to be an extension of the linear acceleration method. The following assumptions are used (Bathel 1996):

Scaling earthquake records

Table 1.

NONLINEAR DYNAMIC ANALYSIS

0.446 0.57 0.517 0.45 0.578 0.709 0.45

48

Table 2.

Nonlinear static and dynamic results. 4 Story Frame

8 Story Frame

12 Story Frame

Records’ Name

Clough

Takeda

Severe

Clough

Takeda

Severe

Clough

Takeda

Severe

Kobe Kocaeli Lomap Northridge Palm Springs Capemendocino Superestiti hills Average of Dynamic Analyses Static Analyses with First Mode Loading Shape Static Analyses with Uniform Loading Shape C2

201.1 167.9 173.4 185.3 244.7 143.3 200.4 188.0 210.58

258.3 262.1 224.3 272.7 342.1 157.1 277.6 256.3

227.9 336.7 163.7 276.4 361.9 218.0 Fail 264.1

173.2 400.4 231.4 296.9 298.7 180.5 288.1 267.0 460.53

284.4 462.9 256.1 450.5 352.9 243.9 358.0 344.1

516.5 528.3 223.2 443.5 373.6 259.3 402.8 392.5

262.6 554.0 178.6 333.0 355.2 220.5 355.6 322.8 659.17

323.1 538.4 208.1 353.3 364.0 324.8 392.5 357.7

736.8 843.8 247.9 376.2 439.0 435.6 721.4 543.0

1.108

1.682

7

202.16 1

399.77 1.363

1.405

1.289

1.47

1

NONLINEAR STATIC ANALYSIS

Nonlinear static analysis has been used based on displacement which is mentioned in FEMA-356. (FEMA 2000) In this method structure’s nonlinear model is pushing by a specified loading pattern until reaching specified displacement called target displacement. Deformations and internal forces are calculated and compared with allowable amounts, after achieving to the target displacement. Target displacement which is the maximum displacement that structures can experience in earthquake loading is calculated by spectrum value in effective period of structure and applying special factors. 7.1

1

627.53

Figure 7. 2000).

Calculating major effective period

period with specified damping. C0 is modification factor to relate spectral displacement of an equivalent SDOF system to the roof displacement of the building MDOF system. In this research it is considered to be equal to 1.2. C1 is used as a modification factor to relate expected maximum inelastic displacements to displacements calculated for linear elastic response and it is equal to 1 when Te > T0 . Therefore in this research it is considered to be equal to 1. C2 is a modification factor to represent the effect of pinched hysteretic shape, stiffness degradation and strength deterioration on maximum displacement response. Value of C2 is different for different framing systems and structural performance levels. But it is equal to one for intermediate moment frames in all performance levels. For some structures which they don’t have a stable and complete hysteresis loops, cyclic motions can cause to extension of damage and not only lead to strength degradation but also it cause to increase deformations and decrease strength. Because this factor is assumed to be 1 in FEMA-356 (FEMA 2000), in this research with different hysteresis loops

The major effective period can be obtained by bilinear model:  Ki Te = Ti (4) Ke where Ti is the base period of structure with linear behavior assumption and Ki is lateral elastic stiffness as shown in Figure 7. 7.2 Calculating target displacement In this method target displacement is calculated based on Equation 5. δt = C0 C1 C2 C3 Sa

Te2 g 4π 2

Simplified force-displacement curve (FEMA

(5)

where, Te is the effective period in the considered direction. Sa is the amount of site spectrum in effective

49

and it is 25.3% more than 1 in Takeda model. If the structure has incomplete hysteresis loops, high stiffness degradation and strength deterioration and pinching, this factor is 51.9% more. 8

CONCLUSIONS

Nonlinear static analysis is an efficient method for seismic prediction of structure which is widely used based on displacement. It doesn’t consider member’s nonlinear behavior changes which are the results of cyclic behaviors, so by considering these effects in a factor this problem can be approximately solved. C2 is the factor for calculating target displacement. So for the structures which their deformation is medium and the stiffness degradation and strength deterioration is normal, the factor for Takeda model accrued and it is 25.3% more than 1. If it’s deformation, stiffness degradation, strength deterioration and pinching be so high, the factor will be 51.9% more than 1. REFERENCES

Figure 8. analysis.

ACI Committee 318, ‘‘Building Code Requirements for Structural Concrete and Commentary (ACI 318M-05)’’, American Concrete Institute. Applied Technology Council; Seismic Evaluation and Retrofit of Concrete Buildings, Volume 1, ATC-40 Report, Redwood City, California, 1996. Bathe, K.J, ‘‘Finite Element Procedure’’, Prentice-Hall, 1996. Building and Housing Research Center, Iranian Code of Practice for Seismic Resistant Design of Buildings, Standard No. 2800-05, Third Edition, BHRC Publication No. S-253, 2005. Clough, R.W., Benuska, K.L. and Wilson, E.L., ‘‘Inelastic Earthquake Response of Tall Buildings’’, Proceedings, Third World Conference on Earthquake Engineering, New Zealand, Vol. 11, New Zealand National Committee on Earthquake Engineering, 1965. Computer & Structures Inc., ‘‘Linear and Nonlinear Static and Dynamic Analysis and Design of Building Systems Users Guide, ETABS Ver. 9’’, Berkeley, California, 2005. Federal Emergency Mnagement Agency; Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA356, Washington, D.C., 2000. International Institute of Earthquake Engineering and Seismology. ‘‘Seismic Retrofitting Provisions for Existing Buildings’’. First edition, Iran, May–June 2002. Kunnath, S.K., Reinhorn, A.M. and Lobo, R.F., ‘‘IDARC: A Program for the Inelastic Damage Analysis of Reinforced Concrete Structures’’, Report No. NCEER-920022, National Center for Earthquake Engineering Research, State University of New York at Buffalo, 1992. Park, Y.J., Reinhorn, A.M. and Kunnath, S.K., ‘‘IDARC: Inelastic Damage Analysis of Reinforced Concrete Framed-Shear Wall Structures’’, NCEER-87-0008, 1987. Takeda, T., Sozen, M.A. and Nielsen, N.N., ‘‘Reinforced Concrete Response to Simulated Earthquakes’’, Journal of the Structural Division, ASCE, Vol. 96, ST12, Dec. 1970.

Capacity curves obtained from nonlinear static

and by using nonlinear dynamic analysis and comparing it with nonlinear static analysis, this factor is calculated. C3 is a modification factor to represent increased displacement due to dynamic P- effects. Because studied structures have a positive post-yield stiffness, this factor is assumed to be equal to 1. Two kind of loading pattern are used for doing nonlinear static analysis based on FEMA-356 (FEMA 2000): 1- according to the shape of first vibration mode 2- loading pattern which is in proportion with floors mass. After performing analyses and by using capacity curves and idealizing them to bilinear curves, the target displacement has been calculated. Capacity curves of structures are shown in Figure 8. By calculating target displacement for each structure, it is possible to equivalent the nonlinear static analysis results with nonlinear dynamic analysis results using hysteresis model without stiffness degradation and strength deterioration (C2 is 1), so C2 can be defined by calculating the ratio of nonlinear dynamic analysis of the models which have stiffness degradation and strength deterioration to the first model (which doesn’t have any degradation and deterioration). Results are shown in Table 1. Results are separately shown for Clough, Takeda, a model with high stiffness degradation and strength deterioration, and nonlinear static analysis. The results have been calculated and averaging has been done for calculating C2 . C2 is more than 1 in all the conditions

50

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Evaluating shear capacity of RC joints subjected to cyclic loading using ANN A. Said University of Nevada, Las Vegas, USA

E. Khalifa Shair and Partners, Cairo, Egypt

ABSTRACT: For the last few decades, many RC structures collapsed during earthquakes leading to severe losses in lives and properties. Observation of damages indicated that in many cases the main reason was the lack of the shear capacity of beam-column joints. This is usually caused by inadequate reinforcement and detailing of the joint reinforcement. The capacity of beam-column joints is influenced by various key parameters. The effect of each of these parameters has some limit of uncertainty due to the complexity of the joint behavior. Consequently, existing shear design formulae for joints produce varying results depending on the parameters accounted for in each respective formula. This paper investigates the shear behavior of interior beam-column joints subjected to cyclic loading using artificial neural networks (ANNs) based on experimental testing results collected from the literature. The paper aims to clarify the effect of some of the key parameters affecting the shear capacity of the cyclically loaded interior joints including joint shear reinforcement, concrete compressive strength, column axial stress, and joint aspect ratio. The study also evaluates the accuracy of current code formulae of the ACI-ASCE Committee 352 (2002) and Architectural Institute of Japan (1998) using experimental testing results. 1

these studies in accurately predicting the shear capacity of the joints and the influence of each of the design parameters on the shear strength. For the same beamcolumn joint specimen, each of the existing formulae predicts a different value for the shear strength. This study aims to investigate the feasibility of using artificial neural networks (ANNs) to predict the shear capacity of the monotonically loaded exterior beamcolumn joints. The study will also compare ANN predictions to those obtained from the following equations: ACI-ASCE Committee 352 (2002) and the Architectural Institute of Japan (1998). The parameters investigated in this study are joint volumetric reinforcement ratio, concrete compressive strength, joint aspect ratio, and column axial stress.

INTRODUCTION

Beam-column joints are critical regions in reinforced concrete structures. An extensive number of studies were conducted to investigate both the behavior and design parameters of this area. Design parameters such as joint shear reinforcement, concrete compressive strength, joint aspect ratio, and column axial stress are all critical factors in the behavior of beam-column joint. In a planar frame, joint failure results in multiple reductions of redundancy of the structure, whereas a failure in beam or column results in a single reduction of redundancy (Said & Nehdi 2004). Accordingly, strength hierarchy ensures that joint failure is avoided and that a strong column weak beam will limit failure to the beam. Through the last four decades, several studies were conducted to investigate the behavior of beam-column joints and its failure mechanisms, most of these studies focused on cyclic loading. These studies utilized both analytical techniques (Will et al. 1972; Noguchi 1981; Pantazopoulou & Bonacci 1994; Elmorsi et al. 2000) and experimental techniques (Higashi & Ohwada 1969; Durrani & Wight 1982; Otani et al. 1984; Kitayama et al. 1987; Endoh et al. 1991; Joh et al. 1991; Oka & Shiohara 1992; Teraoka et al. 1994) to investigate the shear behavior of the joint. However, despite the extensive analytical and experimental studies conducted, discrepancy still exists between

2

BEHAVIOR OF INTERIOR BEAM-COLUMN JOINT

Interior beam-column joints have a great importance in reinforced concrete structures. The effect of cyclic loading conditions on interior joints is much higher than the effect of monotonic loading. The reasons behind this are: 1. Larger forces can be generated on the joint for the case of cyclic loading depending on the direction of

51

and bases on the type of joint the factors of the formula vary. The general formula is as follows:  Vn = 0.083 γ fc bj hc

where Vn is the nominal shear strength of Type 2 joints, fc is the concrete cylinder strength (MPa), hc is the depth of the column in the direction of joint shear being considered (mm), bj is the effective width of the joint (mm), it is defined as the smaller value of:

Figure 1. Strut and truss models proposed by Park and Paulay (1975) for interior beam-column joints.

forces (the ground motion) rather than the monotonic loading case. 2. According to Chopra (2007), the amount of lateral displacement of a RC structure when subjected to cyclic loading is almost twice the amount of the displacement generated by the same force value when applied monotonically to the joint.

3.1

bb + b c 2  bb + (mhc + 2)

(2b)

bc

(2c)

(2a)

where m = 0.50 for the case of no eccentricity between the beam and column centerlines, γ = 15 for Type 2 interior planar joints (database case). Accordingly the formula becomes:  Vn = 1.245 fc bj hc (3)

In any reinforced concrete frame subjected to seismic loading, beams and columns experience flexure and shear forces. These forces are transformed into higher shear values acting on the joint and they might cause a shear failure in the joint. This type of failure has severe damaging results on the structure. The strut and truss models proposed by Park and Paulay (1975) can be used for the cyclically loaded interior beam-column joints. As shown in Figure 1, two mechanisms are used for the transfer of loads through the joint. The first one is the strut mechanism which accounts for the concrete contribution to the shear strength of the joint. In this mechanism, a single concrete compression strut is used to transfer the shear forces through the joint. The second one is the truss mechanism which accounts for the contribution of joint shear reinforcement in transferring the shear forces through the joint. In this mechanism, the load is transferred through a steel tie represented by the joint shear stirrups. To ensure the presence of the tie mechanism, a strong and uniform bond stress distribution along the beam and column reinforcement should exist.

3

(1)

3.2

Design equation of the Architectural Institute of Japan (1998)

Most of the recommendations provided in the Japanese design guidelines for the cyclically loaded beamcolumn joints are based on studies conducted by Aoyama (1993) on the behavior of cyclically loaded beam-column joints. According to his study, it is stated that there are two earthquake design methods. The first is the strength design, in this method the structure is designed to sustain large lateral load resistance capacity. The second method is the ductility design method, where the structure is designed to have a large inelastic deformation capacity. It is very important for any structure not to suffer brittle failure by dissipating the energy of the earthquake through plastic hinges formed in the beams. This actually represents the strong column weak beam theory. This theory states that the structure should be designed to have a stronger column than the beam to increase the dissipation of energy, and to ensure the simultaneous formation of plastic hinges in the beams. Based on his study, the Architectural Institute of Japan (1998) provides the following formula for calculating of the shear capacity of cyclically loaded beam-column joints.

PREVIOUSLY PROPOSED FORMULAE AND EQUATIONS ACI-ASCE Committee 352 Formula (2002)

Vu = k × φ × Fj × bj × D

According to the ACI-ASCE Committee 352 (2002), the cyclically loaded joints are categorized as Type 2. Type 2 joints are the ones designed to have sustained strength under deformation reversals into the plastic range (seismic loading case). The ACI-ASCE Committee 352 (2002) proposes a general formula for the design of beam-column joints

(4)

where k = 1, φ = 0.85, Fj = 0.80 * ( fc )0.70 (MPa), D is the column depth, bj = effective column width. This leads the formula to be: Vu = 0.68 × (fc )0.70 × bj × D

52

(5)

3.3

Artificial neural network approach

groups of input vectors, each vector representing one of the investigated parameters in the study and the output vector represents the shear capacity of the joint. Table 1 represents the database range of the parameters investigated in the study.

Artificial Neural Networks are one of the most popular artificial intelligence techniques used in engineering applications. Multi-layer perceptron networks (MLP) have been widely used in engineering applications. They are able to map a given input (s) into desired output (s), and accordingly detect hidden and complex behavioral trends of such engineering problems by learning through the database used to train the system (Haykin 1994). The architecture of the MLP networks consists of an input layer which represents the investigated parameters in the network, an output layer which represents the final result of the network or the behavior under investigation, and a number of hidden layers. Each layer contains a number of processing elements that are fully or partially connected to the elements in successive layers. The strength of the bond between processing elements is a numerical value called the weight of the connection. The simulation process in ANN can be expressed as the operation of detecting the optimum weights such that the network can predict an accurate value for the output within the database range. 4

5

ANN MODEL

To predict the shear strength of cyclically loaded beam-column joints, an ANN was constructed with the following components: an input layer, an output layer and two hidden layers. The input layer contains four variables representing the common shear design parameters of reinforced concrete beam-column joint (volumetric reinforcement ratio, concrete compressive strength, joint aspect ratio, and column axial stress). The output layer includes one unit representing the shear capacity, Vn and the hidden layers consisted of eight and four processing units consecutively. Full bonding connections were used between the processing elements and the elements in other consecutive layers. The software used in this model is MATLAB (2007). This software is commonly used for the simulation process of engineering problems. This software divides the given database into training and testing groups to increase the accuracy of the model and give a better understanding of the effect of each parameter in the output layer. Figure 2 represents the architecture of the proposed model.

EXPERIMENTAL DATABASE

The most important factor contributing to the performance of ANNs is the learning material used in the training process. Accordingly, it is imperative to train a network model on a comprehensive database to capture the actual embedded relationships between the parameters of the input and output layers. In this study, the aim is to detect the relationships between the different parameters being considered and their effect on the shear capacity of interior beam-column joints under cyclic loadings. In this study, shear capacity of this joint type is investigated using a database consisting of 58 concrete beam-column connections collected from published literature and listed in Khalifa (2008). The accuracy of the network was improved by imposing several limitations on specimens in the database used by the ANN model. Specimens failing due to joint shear were strictly used, with no beams in the transverse direction. Specimens with high strength concrete, and reinforcement welding into the joint were omitted. The database was formatted into

6

RESULTS AND DISCUSSIONS

To consider an ANN successful, it must be able to accurately predict output values for input values within the range of the database used in the training and the testing process. To evaluate the accuracy of the proposed network, a comparison was held between the

Table 1. The parameters range for the investigated database. Parameter

Minimum Maximum

Joint aspect ratio 1 Concrete compressive strength MPa 21.2 Volumetric reinforcement ratio (%) 0 Column axial stress (MPa) 0

1.3 70 3.15 17.8

Figure 2.

53

Architecture of artificial neural network model.

network predicted outputs which represent the shear capacity and those calculated using the formulae by ACI-ASCE 352 (2002) Architectural Institute of Japan (1998) The performance of each model was evaluated based on both the ratio of measured to predicted (or calculated) shear strength (Vm /Vp ), and the average absolute error (AAE) calculated using the following equation: AAE =

  1  Vm − Vp  × 100 n Vm

(6)

The average value, the standard deviation (STDV ), and coefficient of variation (COV ) for Vm /Vp , and the average absolute error (AAE) of the ANN model and ACI-ASCE 352 (2002) are listed in Table 2. The shear strength of beam-column joints calculated using current shear design provisions are plotted against the experimentally measured values in Figure 3. Figure 3a indicates that the ACI shear design guidelines for reinforced concrete beam-column joints are highly inaccurate even without application of reduction factors as shown. This formula neglects the influence of the joint aspect ratio and the column axial stress, and the contribution of the joint reinforcement to the shear capacity of the joint. Using the selected data for this study and knowing the actual capacity of the specimens obtained from the experimental programs results, the average absolute error AAE for this formula is 63%, which is significantly high, and the STDV for Vm /Vp of this formula is 0.29. It is recommended that this formula should not be used to estimate the shear capacity of beam-column joints due to its lack of accuracy. It should rather be used to estimate the minimum shear strength of the joint based on concrete properties and joint dimensions. Design equations proposed by the Architectural Institute of Japan (1998) resulted in highly inaccurate prediction of the shear strength of the cyclically loaded interior bream-column joints. Figure 3a represents a plot of the actual experimental shear strength values versus the calculated ones using this formula. This formula neglects the influence of the joint aspect ratio, the column axial stress, and the contribution of joint reinforcements to the shear capacity of the joint.

(a)ACI-ASCE 352 (2002)

(b) Architectural Institute of Japan (1998)

(c) ANNs Figure 3. Measured versus calculated shear capacity of beam-column joints.

Table 2. Performance of different formulae for the calculation of shear strength of RC interior beam-column joints under cyclic loading.

Using the selected data for this study and knowing the actual capacity of the specimens obtained from the experimental programs results, the average absolute error AAE for this formula is 90%, which is extremely high, and the STDV for Vm /Vp of this formula is 0.297. Neglecting several major factors governing the behavior of the joint refute the accuracy and the validity of this formula.

Vmeasured /Vpredicted Method

AAE (%) Average STDV COV

ACI-ASCE 352 (2002) 63 AIJ (1998) 90 ANN 8.15

0.77 0.651 0.99

0.29 38.7 0.297 48.00 0.0988 10

54

The proposed model for the ANNs produced much more accurate outputs for predicting the shear capacity of joints than the formula proposed by ACI-ASCE 352. Figure 4c shows that this model reduced the AAE between the actual and the predicted values to a very small value (8.15%). The model also resulted in a smaller scatter for the data with STDV of 0.0988. The small value of AAE ensures the accuracy of selecting the investigated parameters as the key factors governing the shear behavior of joints. 7

Joh, O., Goto, Y. and Shibata, T. 1991. Influence of Transverse Joint, Beam Reinforcement and Relocation of Plastic Hinge Region on Beam-Column Joint Stiffness Determination. In ACI Special Publications SP 123-12: Design of Beam-Column Joints for Seismic Resistance, Farmington Hills, Michigan, pp. 187–223. Joint ACI-ASCE Committee 352, 2002, ‘‘Recommendation for Design of Beam-Column Connections in Monolithic Reinforced Concrete Structures’’, American Concrete Institute, Farmington Hills, Mich, 40 p. Kitayama, K., Otani, S. and Aoyama, H. 1987. Earthquake Resistant Design Criteria for Reinforced Concrete Interior Beam-Column Joints. In Pacific Conference on Earthquake Engineering, Wairakei, New Zealand, pp. 315–326. Noguchi, H. 1981. Nonlinear Finite Element Studies on Shear Performance of RC Interior Column-Beam Joints. In IABSE Colloquium, Delft, The Netherlands, pp. 639–653. Oka, K. and Shiohara, H. 1992. ‘‘Test on High -Strength Concrete Interior Beam-Column Sub-Assemblages’’. In 10th World Conference on Earthquake Engineering, Madrid, Spain, pp. 3211–3217. Otani, S., Kobayashi, Y. and Aoyama, H. 1984. Reinforced Concrete Interior Beam-Column Joints under Simulated Earthquake Loadings. In US-New Zealand- Japan Seminar on Design of Reinforced Concrete Beam-Column Joints, Monterey, CA. Pantazopoulou, S. and Bonacci, J. 1994. On Earthquake Resistant Reinforced Concrete Frame Connections. Canadian Journal of Civil Engineering 21: 307–328. Park, R. and Paulay, T. (1975). ‘‘Reinforced Concrete Structures’’, John Wiley and Sons, United States of America, 769 p. Teraoka, M., Kanoh, Y., Hayashi, K. and Sasaki, S. 1997. Behavior of Interior Beam-and-Column Sub Assemblages in RC Frame. First International Conference on High Strength Concrete, Kona, Hawaii, pp. 93–108. Teraoka, M., Kanoh, Y., Tanaka, K. and Hayashi, K. 1994. Strength and Deformation Behavior of RC Interior BeamColumn Joint Using High Strength Concrete. In Proceedings, Second US-Japan-New Zealand- Canada Multilateral Meeting on Structural Performance of High Strength Concrete In Seismic Regions, Honolulu, Hawaii, pp. 1–14. The Math Works., 2007, ‘‘MATLAB (2007)’’, Orchard Hill, Michigan, United States. Will, G.T., Uzumeri, S.M. and Sinha, S.K. 1972. Application of Finite Element Method to Analysis of Reinforced Concrete Beam-Column Joints. In Proceeding of Specialty Conference on Finite Element Method in Civil Engineering, CSCE, EIC, Canada, pp. 745–766.

CONCLUSIONS

The purpose of this study was to study the feasibility of using artificial neural networks to predict the shear strength of monotonically loaded exterior beam-column joints. The proposed technique outperformed existing equations in the ACI code and the literature. The study also shows that ANNs are very useful tool for complex engineering problems. Further refinement to the proposed technique can be provided through incorporating new experimental research results. REFERENCES Aoyama, H. Empirical versus Rational Approach in Structural Engineering—What We Learned from New Zealand in the Trilateral Co-operative Research on Beam-Column Joint, ACI Special Publication SP-I Detroit, September 1993, pp. 31–57. Architectural Institute of Japan, 1998. Recommendations of RC Structural Design after Hanshin-Awaji Earthquake Disaster-Cause of Particularly Noticed Damages and Corresponding RC Structural Details. Chopra, A.K. 2007. Dynamics of Structures, Prentice Hall, Englewood Cliffs, New Jersey. Durrani, A.J. and Wight, J.K. 1982. Experimental and Analytical Study of Beam to Column Connections Subjected to Reserve Cyclic Loading. Technical Report UMEE82 R3, Department of Civil Engineering, University of Michigan, 295p. Elmorsi, M., Kianoush, M, R. and Tso, W.K. 2000. Modeling Bond-Slip Deformations in Reinforced Concrete Beam-Column. Canadian Journal of Civil Engineering 27: 490–505. Endoh, Y., Kamura, T., Otani, S. and Aoyama, H. 1991. Behavior of RC Beam-Column Connections Using LightWeight Concrete. Transactions of Japan Concrete Institute, 319–326. Haykin, S. (1994). ‘‘Neural Networks: A Comprehensive Foundation’’, Macmillan, New York, 842 p. Higashi, Y. and Ohwada, Y. 1969. Failing Behavior of Reinforced Concrete Beam-Column Connections Subjected to Lateral Load. Memories of Faculty of Technology Tokyo Metropolitan University, Tokyo, Japan, pp. 91–101.

55

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Evaluation of drift distribution in performance-based retrofitting of RC frames G. Ghodrati Amiri Center of Excellence for Fundamental Studies in Structural Engineering, School of Civil Engineering, Iran University of Science & Technology, Tehran, Iran

A. Gholamrezatabar Department of Civil Engineering, Shomal University, Amol, Iran

ABSTRACT: As an experienced retrofit scheme, concentric structural brace elements are installed in reinforced concrete (RC) frames to upgrade original existing buildings. In the current research installed brace elements are designed based on force and satisfying seismic performance levels of structural elements under lateral loading. The drift distributions of stories for both forces-based design (FBD) and performance-based design (PBD) of steel braces are evaluated in retrofitted RC frames. On the basis of pushover analysis damage distributions and corresponding performance levels of formed plastic hinges are investigated as well. Results show, performance base design causes different drift distributions pattern from that of force base design. It is recognized that designing brace elements on the base of force concentrates structural damages, while performance base design spreads damages and therefore prepares higher seismic performance levels and satisfies retrofit goals. 1

may be selected in accordance with required performance. However the codes do not clearly prepare a guideline to design braces elements of RC frames based on performance. The first attempt was standardized the performancebased design approaches by Federal Emergency Management Agency (FEMA) that sponsored the development of national guidelines for the seismic retrofit of buildings; ATC-33 project (1992). Displacementbased design (DBD) as a simplified approach of performance based-design (PBD) was proposed initially for bridges (Kowalsky et al. 1995). These two approaches have been used interchangeably, because performance objects can be related to the level of damages to the structure which can be expressed in term of displacement or drift. The approach was adopted by Structures Engineering Association of California (SEAOC) to design new buildings (2000) on the basis of pushover analysis. In current paper brace elements of RC frames design based on strength and performance level. The static nonlinear pushover analysis was performed because it may provide much of the needed information. According to pushover analysis method predetermined load applied and gradually increased until the displacement of roof reaches to target displacement. The drift distributions of retrofitted frames were evaluated. Formation of plastic hinges and their performance levels were investigated.

INTRODUCTION

Installing steel brace elements to the reinforced concrete (RC) frames is a feasible effective approach that provides retrofit purposes. Several researchers conducted experimentally and analytically investigations to study seismic behavior of rehabilitated RC frames. A. Ghobarah and H. Abou Elfath (2001) retrofitted a nonlinear three stories building using concentric an eccentric inverted-V steel brace elements. T.ELAmoury and A. Ghobarah modeled nine and eighteen stories RC frames analytically to study seismic behavior under various scaled motions. M.R. Maheri, R. Kousari and M. Razazan developed experimental testes on the ductile RC frames with X-brace elements. Seismic evaluation of RC frames with inverted V-brace elements, formation of plastic hinges and their local and global performance levels shows a considerable improvement in rehabilitated RC frames. Generally the seismic rehabilitation is achieved to upgrade the original performance to satisfy higher seismic demands of current code. Over the last couple of decades a lot of endeavors put on developing a new method of analysis and explicit designing regarding to various level of hazard and multiple criteria for levels of local and global performance. Now after establishing wide research on conventional method and using advances of earthquake engineering; retrofit technique

57

2

MODELS GEOMETRY AND MECHANICAL PROPERTIES

Models selected for analytical investigation consist of four reinforced concrete frames with three, five, seven and ten stories. The frames have three spans of 4.0 m and the same height of 3.1 m. Frames are symmetric and tensional effects are minimal. Frames were considered being moment resisting with intermediate ductility. Dimensions of beams, columns and details of reinforcement calculate to provide enough capability for frames to withstand seismic loads estimated by previous Iranian code. Due to that square column sections from 300 mm to 600 mm were used and, with reinforced ratio not exceeding about 2%. Beams sections varied from 300 × 300 to 450 × 600 (mm2 ). Beams and columns cross-sections keep the same in every two stories. It reassures that strong-column and week-beam principal is reserved. Concrete comprehensive of 21 MPa and steel yield strength of 350 MPa were used for materials. The initial elastic module for concrete material is 80 GPa and for steel material is 200 GPa. 3

(a)

(b) Figure 1.

modeling (a, b, c, IO, CP, LS) extracted from tables which was prepared by FEMA273. Introduced parameters obtain based on geometry, concrete properties, corresponding transitive and longitude steel ratio. In other words bending and shear characteristics of hinges for the beams are defined according to the parameters in Equations 1 and 2.

RETROFIT SCHEME

Many approaches and techniques have been studied and practiced for recent 20 years to upgrade and improve seismic performance of existing structures. The aims of rehabilitation are: (a) to recover original structure performance (b) to upgrade original performance; and (c) to reduce seismic response; so as to building earthquake vulnerability. As an effective scheme concentric inverted V-brace elements are added to existing RC frames to upgrade original performance of them. Structural steel brace elements install using three connections. Plates and bolds connects brace element to column and beam at one end at their joint and fixed it to the middle of mid-span beam at the other end. Based on conventional approach new added structural elements reduce large response displacement, so higher performance for retrofitted RC frames are expected. 4

Generalized load-deformation relationship.

3.77

V  b w d fc

ρ − ρ ρbal

(1) (2)

For columns axial force-bending moment curve for plastic hinges were defined and assign to linear elastic columns. The same procedure conducted for beam plastic hinges, was used to find parameters of nonlinear modeling for column plastic hinges using computer macro. Corresponding reinforcement in the column are calculated using Equations 1 and 3.

NONLINEAR DEFINITIONS AND MODELING

P A g fc

Structural elements were modeled using elastic elements with two nonlinear hinges at their ends. In fact hinges represent material nonlinearity during gradually loading. Due to that beams modeled using linear elastic element with two point moment hinges. Yield moment introduce as a bending moment. Nonlinear behavior of plastic hinges defined according to generalized load-deformation relation diagram; Figure 1a and 1b. In current research parameters of nonlinear

(3)

Load-deformation relation that represents nonlinear behavior of brace elements in tension and compression specified for each element. Parameters of nonlinear modeling or properties of brace elements hinges depend on yield stress and buckling limits in compression. Performance levels of hinges were extracted from tables which are recommended in chapter 5 of FEMA273.

58

5

DESIGN OF BRACE ELEMENT

were calculated according to modified Iranian Seismic Code. Linear dynamic analysis (LDA) performed and the amount of brace forces calculated. Brace elements were design for tension and compression according to Iranian steel structures.

In current research two different methods were carried out to design brace elements for each story. First traditional method is based on force. According to force-based design (FBD) philosophy the design criteria are defined by limits on stress from prescribe lateral loading. As second method, advanced methodology that is based on performance based-design set up for designing of brace elements. PBD procedure proposed a more general design philosophy regarding to multiple design criteria expressed in term of stated performance objects when the structure subjected to level of seismic hazard. 5.1

5.2

Second upgraded method for designing braces is based on performance objects which expressed acceptable level of damage. According to performance baseddesign concept the aim is providing one of categorized performance level, such as immediate occupancy (IO), collapse prevention (CP) and life safety (LS). To achieve the aims, nonlinear static analysis (NSA) was performed to determine (1) capacity (2) demand (3) performance. The capacity spectrums were obtained via pushover analysis. To achieve demand spectrum we used standard spectrum which was reduced due to 5% damping. According to definition the interaction of obtained spectrums determines performance point. Responses of building can be estimated based on assessing performance and hazard levels at target displacement. The responses can be checked against acceptability limits on either global system levels in term of stability and interstory drift or local element performance levels such as the element strength and sectional plastic rotation. In current investigation the brace elements design on the base of local element performance level observation at target displacement is developed. According to definition, target displacement is maximum

Force-based design

RC frames were braced using various tube sections, and increased lateral load were applied. Seismic loads

Figure 2.

Performance-based design

Performance base design procedure.

5

3

2

Story

Story

4

Force Base Design

3 Force Base Design

2

Performance Base Design

Performance Base Design

1

1

0

0.005

0.01

0.015

0.02

0

0.01

0.02

(b)

(a) 7

10 9 8 7 6

6

Story

Story

5 4 Force Base Design

3

Performance Base Design

2 1 0

0.01

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0.03

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5 4 3 2 1

Force Base Design Performance Base Design

0

0.05

drift(%h)

0.005

0.01

0.015

0.02

0.025

0.03

drift(%h)

(c)

Figure 3.

0.03

drift(%h)

drift(%h)

(d)

Drift distribution of braced RC concrete frames (a) Three stories, (b) Five stories, (c) Seven stories, (d) Ten stories.

59

IO

IO

IO B

IO

E

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B

D B E CP

IO

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B B

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B D CP

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(a)

Figure 4.

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(a)

(b)

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(b)

Formation of plastic hinges and local performance levels of structural elements (a) push1 (b) push2—Three Stories. B

CP

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LS

B B

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E

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C

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E

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CP B

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(a)

Figure 5.

IO

CP

B

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B

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B

IO

B B

B

B

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IO

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B

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B B

(a)

(b)

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(b)

Formation of plastic hinges and local performance levels of structural elements (a) push1 (b) push2—Five Stories. E

B

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(a)

Figure 6.

B

(b)

(a)

(a)

Formation of plastic hinges and local performance levels of structural elements (a) push1 (b) push2—Seven Stories.

probable displacement that building may experience during special earthquake. This drift is useful for engineers for the purpose of evaluation or retrofitting. Introduced displacement is estimated by Equation 4 due to the selected performance level. In this equation which is recommended by FEMA. δt = C0 C1 C2 C3 Sa

Te2 g 4π 2

maximum inelastic displacements, C2 modification factor to represent the effect of hysteretic shape and C3 coefficient factor for post-yield stiffness. Braced RC frames pushed to target displacement to evaluate formed plastic hinges performance. The drift of each frame obtain using PBS approach were compared with force base design approach.

(4) 6

parameters of nonlinear analyzing are C0 modification factor to relate spectral displacement and likely building roof displacement, C1 modification factor for

COMPARISON OF DRIFT DISTRIBUTIONS

Figure 3 shows interstory drift distribution of 3-story retrofitted frame carrying out both force based-design

60

to 1.1(QD + QL ) and 0.9(QD + QL )(QD ); dead load and QL ; live load), are pushed primarily according to FEMA273 recommendation. Then lateral load is gradually applied to achieve the target displacement. Consequently two types of nonlinear analysis could be defined such that for each of them lateral load follows one of defined gravity loads. Retrofitted RC frames using both performance based-design and force based-design were evaluated in term of formation of plastic hinges and their prepared local performance levels at target displacement. Figure 7 shows formation of plastic hinges at target displacement for frames retrofitted based on performance. As it is expected, almost all of formed plastic hinges remain in collapse prevention level. Figure 8 presents formation of plastic hinges in retrofitted frame based on force while subjected to lateral load. Comparison of them states FBD procedure concentrates damages on second floor structural elements. In other words columns damaged at target drift. Also chord plastic rotation of formed hinges exceeds CP level. However PBD distributes the damages along the height of frame. Seismic performance of rehabilitated five, seven, and ten RC frames are evaluated by conducting the same nonlinear analyzing procedure. Chord rotation in these RC frames retrofitted based design ranged from B to CP level while undergoing lateral loading. RC frames retrofitted using force based-design approaches could not provide expected seismic performance. Consequently many structural elements are significantly damaged. Formed plastic hinges with grater plastic rotation occurred in third floor column which is coded by D states a considerable concentrated damage in columns of retrofitted RC frames based on FBD approach. It is worth noting that reduction in brace area based on PBD procedure lead to an obvious difference in formation of plastic hinges and prevent

and performance based-design approach. Comparison of drift distribution of three story frame reveals that RC frame retrofitted setting up performance base design method predict smaller drift for second floor than force base design method. It means performance based-design maintain brace elements with bigger dimensions for second floor. However this approach allows third floor has bigger drift. It can be recognized that PBD method estimate an over strength for third floor elements. Due to that PBD procedure achieves smaller dimension than FBD procedure for third floor braces used section. Figures 1, 2 and 3 illustrate interstory drift distribution of five, seven and 10-story case study retrofitted based on two introduced design approaches. It is pointing out all three of them which are retrofitted using PBD method predict greater story drift for lower floors in compare with FBD approach. But as observed, frames retrofitted performing conventional approach base design show smaller drift for higher floors. An interesting point is that the values of calculated story drift based on PBD and FBD approaches coincide almost at 2/5 height of frames. As described before it can be obviously seen that the value of last story drift for PBD base retrofitting is smaller than one for FBD base retrofitting. It indicates an over strength presented in term of story drift.

7

EVALUATION OF EFFECT OF DESIGN APPROACHES ON DAMAGE

To evaluate the seismic capacity of retrofitted RC frames with more accurate and to obtain a precise understanding of differences in results of two applied design approaches, a complete set of nonlinear static (pushover) analysis were carried out. To achieve the aims, two types of gravity load which are restricted

B

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Figure 7.

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Formation of plastic hinges and local performance levels of structural elements (a) push1 (b) push2—Ten Stories.

61

from damages concentration. Due to that higher performance levels are achieved. Analyzing of seven and ten story RC frames with steel brace elements are clear testimony that proves inabilities of force base design retrofitting approach. 8

Ganzerli, S. Pantelides, C.P. & Reaveley, L.D. 2002. Performance-based design using structural optimization. Earthquake Engineering and Structural Dynamic; 29: 1677–1690. Ghobarah, A. & Abou Elfath, H. 2001. H. Rehabilitation of a reinforced concrete frame using eccentric steel bracing. Engineering Structure; 23:745–755. Ghodrati Amiri, G. & Gholamrezatabar, A. 2008. Evaluation of Performance of Reinforced Concrete Frame Retrofitted Using Concentric Steel Bracing; Structure and Steel (in Persian) 4:17–25. Higashi, Y. et al. 1981. Experimental studies on retrofitting of reinforced concrete structural members. In: Proceedings of the second seminar on repair and retrofit of structures. Ann Arbor (MI): National Science Foundation. pp. 126–155. IIEES (International Institute of Earthquake Engineering and seismology), 2002. Seismic Rehabilitation Code for Existing Building in Iran. Maheri, M.R. et al. 2003. Pushover tests on steel X-braced and knee-braced RC frames. Engineering Structure; 25(13): 1697–1705. Masri, A.C. & Goel, S.C.1996. Seismic design and testing of an RC slab-column frame strengthened by steel bracing. Earthquake Spectra; 12(4): 645–666. Ministry of Housing and Urban Development, 2004. Iranian National building code for structural loading (Standard No.519, Part6). Iran. Ministry of Housing and Urban Development, 2004. Iranian National building code for steel structures (Part10). Iran. Nateghi-Alahi, F. 1995. Seismic strengthening of eight-storey RC apartment building using steel braces. Engineering Structure; 17(6): 455–461. Ohishi, H. et al. 1988. A seismic strengthening design and practice of an existing reinforced concrete school building in Shizuoka city. In. Proceedings of the ninth world conference on earthquake engineering. Vol. VII. pp. 415–420. Raul D. Berter & Vitelmo V. Berter. 2002 Performance-based seismic engineering: the need for a reliable conceptual comprehensive approach. Earthquake Engineering and Structural Dynamic; 31: 627–652. Rodriguez, M. & Park, R. 1991. Repair and strengthening of reinforced concrete buildings for seismic resistance. Earthquake Spectra; 7(3): 439–459. SEAOC Vision 2000.1995. Performance Based Seismic Engineering of Buildings. Structural Engineers Association of California. Canada. Stewart, J.P. et al. 2002 Ground motion evaluation procedures for performance-based design. Soil Dynamics and Earthquake Engineering; 22: 765–772. Zou, X.K. & Chan, C.M. 2005. Optimal seismic performancebased design of reinforced concrete buildings using nonlinear pushover analysis. Engineering Structures. 27: 1289–1302.

CONCLUSIONS

Comparing drift distributions for both applied retrofitting and applied retrofitting approach states results of PBD method more tend to a straight line pattern than FBD method. It can be distinguished PBD approach proposed for achieving higher seismic assessment; cross sectional areas of brace elements should distribute across the height of frame in manner of uniform drift distributions for interstory. Presented plastic hinges and consequently corresponding local performance levels of structural elements PBD approach prevent damages concentration due to miner plastic deformations. As a novel note, FBD method with strength limits and restrictions do not have enough ability to distribute braces optimal in other to use all seismic capability of week existing buildings as a retrofitting goal. However PBD procedure is more effective for retrofitting aims. REFERENCES American Institute of Steel Construction Inc (AISC), 2005. Seismic provisions for structural steel buildings, Standard ANSI/AISC 341-05. Chicago (IL, USA). ATC. Applied Technology Council, 1996. Seismic evaluation and retrofit of concrete buildings-volume 1 (ATC40). Report No. SSC 96-01.Canada. Badoux, M. Jirsa, JO. 1990. Steel bracing of RC frames for seismic retrofitting. Structural Engineering ASCE; 116(1): 55–74. Building and Housing Research Center, 2005. Iranian Code of Practice for Seismic Resistant Design of Building-3rd Revision (Standard No. 2800). Iran. Building and Housing Research Center. 2005. Iranian Code of Practice for Seismic Resistant Design of Building-2nd Revision (Standard No. 2800). Iran. Chopra, A. Dynamics of Structures. 1995. Theory and Applications to Earthquake Engineering. Prentice-Hall: Englewood CliDs, NJ. FEMA 273, 1997. NEHRP Guideline for Seismic Rehabilitation of Buildings. Federal Emergency Management Agency. America. FEMA 356, 2000. NEHRP Guideline for Seismic Rehabilitation of Buildings. Federal Emergency Management Agency. America.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Experimental research on high-frequency fatigue behavior of concrete Y. Chen, W. Shao, H. Han & Z. Yin University of Shanghai for Science and Technology, Shanghai, P.R. China

R. Azzam RWTH Aachen University, Aachen, Germany

ABSTRACT: The fatigue strength of concrete under cyclic loads of vehicles is an important problem in calculation and design of bridge engineering. Based on this viewpoint, the fatigue tests of plain concrete under constant-amplitude and stepping-amplitude cyclic loads were conducted. The damage mechanism of plain concrete specimens under high-frequency fatigue loads was analyzed and the nonlinear accumulative fatigue damage formula was proposed. The fatigue equation P–S–N considering the failure probability p was given. The abovementioned research results made a good preparation for further study on high-frequency fatigue tests of concrete cylinders reinforced with carbon fibers. 1

2.1

INTRODUCTION

Bridge structures of highway and railway suffer cyclic loads from vehicles. The failure characteristics of these structures and their components under cyclic loads show that the fatigue failure loads are far lower than their original strength. The fatigue failure is abrupt and the consequences are very serious. In this paper, fatigue tests of plain concrete under constant-amplitude and stepping-amplitude cyclic loads were carried out, the damage mechanism was analyzed, the nonlinear fatigue accumulative damage model was established and loading frequency correction coefficient considering the effect of loading frequency on fatigue life was put forward. In addition, probability statistics analysis was conducted on the tests data and a fatigue equation considering the failure probability was proposed.

2

High frequency fatigue tests under constant-amplitude and stepping-amplitude cyclic load

2.1.1 Static loading tests 33 cylindrical C30 concrete specimens with 70 mm in diameter, 100 mm in height were prefabricated for the tests. The specimens were manufactured in plastic moulds and were tested after having been maintained for 28 days at the temperature 20 ± 3◦ C. Through the static loading tests it was obtained that the ultimate loading capacity of cylindrical specimens is Fu = 112.11 kN, and so the axial compressive strength is fco = 29.1 MPa. The compressive strength of cubic samples is fcu = 36.6 MPa, and the elastic modulus amounts to Ec = 4.73 × 104 MPa.

HIGH FREQUENCY FATIGUE TESTS OF PLAIN CONCRETE

The loading frequency of common fatigue tests is usually below 30 Hz, while the loading frequency of high frequency fatigue tests can reach 100 Hz. With high frequency, time and costs for reaching the fatigue limit and strength of the material are dramatically reduced. Here high frequency fatigue tests with constant amplitude and stepping amplitude are conducted.

Figure 1.

63

Relationship between fatigue load parameters.

Table 1.

The loading condition of high-frequency fatigue tests. Group

σm

σa

σmax

σmin

Ratio of σmin to σmax R

Number of specimens

Constant-amplitude cyclic loading tests

Group1 Group2 Group3 Group4 Group5

0.45 0.45 0.45 0.45 0.45

0.4 0.35 0.32 0.3 0.28

0.85 0.8 0.77 0.75 0.73

0.05 0.1 0.13 0.15 0.17

0.059 0.125 0.169 0.200 0.233

4 3 3 4 3

Stepping-amplitude cyclic loading tests

Group1

0.45

0.45

0.73 0.75 0.77 0.78 0.85 0.75 0.8

0.17 0.15 0.13 0.11 0.05 0.15 0.1

0.233 0.200 0.169 0.141 0.059 0.200 0.125

4

Group2

0.28 0.3 0.32 0.34 0.4 0.3 0.35

4

2.1.2 The introduction of high frequency fatigue testing machine The high frequency fatigue tests were conducted in AMSLER HFP (High Frequency Pulsator)—100 HFP 5100 testing machine, which was produced by ZWICK/ ROELL company. The basic technical parameters of testing machine include the maximum test load (100 kN), maximum vibration frequency (150 Hz), maximum static loading capacity (100 kN) and maximum dynamic loading capacity (100 kN ± 50 kN). Figure 2. The loading form of stepping-amplitude cyclic loading tests.

2.1.3 Test schemes 1. Constant-amplitude cyclic loading tests: The test parameters include upper limit fatigue f f load Fmax , lower limit fatigue load Fmin , mean load f Fm , the biggest stress σmax , the smallest stress σmin , mean stress σm , the stress amplitude σa and the difference between σmax and σmin , σ . The relationship among the parameters is shown in Figure 1. Table 1 indicates the loading condition for the high frequency fatigue tests under constantamplitude cyclic loads. 2. Stepping-amplitude cyclic loading tests: Figure 2 shows the loading form of steppingamplitude cyclic loading tests and Table 1 shows the loading condition. 2.2

Table 2. Fatigue life results of high-frequency fatigue tests with constant amplitude. Specimen σmax

Test results

Constant-amplitude cyclic loading tests were carried out at 5 stress levels using 17 specimens while stepping-amplitude cyclic loading tests were divided into 2 groups using 8 specimens in total. The fatigue life of specimens is shown in Tables 2 and 3. Among the data, two are invalid (tests were not successful) and indicated with delete line, three exceed the limit fatigue life of the national norm 2 × 106 and are indicated with underline.

64

σmin

Fatigue life

Mean fatigue life

Group 1 PCF15 PCF16 PCF19 PCF24

0.85

0.059

10331 944 14708 55090

26710

Group 2 PCF13 PCF17 PCF20

0.8

0.125

61063 41773 75718

59518

Group 3 PCF5 PCF21 PCF22

0.77

0.169

244165 252934 3870

248550

Group 4 PCF6 PCF12 PCF18 PCF23 Group 5 PCF7 PCF8 PCF9

0.75

0.200

0.73

804676 1515514 2100011 2100000 1057368 0.233 2100003 1405245 940873 1174859

Table 3. Fatigue life results of high-frequency fatigue tests with stepping amplitude. Fatigue life σm

Specimen PCF11

Group 1 σa 0.45 0.28 0.3 0.32 0.34 0.4

2000000 2000000 2000000 2000000 60589

PCF12

PCF14

PCF25

2000000 2000000 2000000 1032433 –

2000000 2000000 2000000 1551877 –

2000000 2000000 2000000 2000000 202896

(a)

Group 2 Specimen PCF3 PCF4 PCF18 PCF10 0.45 0.3 2000000 2000000 2000000 2000000 0.35 814144 943842 595244 485954

2.3

(b)

High-frequency fatigue failure mechanism of C30 concrete specimens

According to the results of high-frequency fatigue tests, there are 2 main modes of fatigue failure: 1. Vertical failure. In the loading process, the loading plates of fatigue testing machine fully coincide with the end planes of concrete specimen. stress, the vertical micro-cracks will arise and develop into vertical macro-cracks, which will result in the rupture of specimen. 2. Conical failure. In the process of loading, the cracks of concrete specimens don’t develop along the loading direction. The final rupture macro-cracks look like cones.

(c)

From the fatigue testing results, it is found that the vertical failure is the main failure mode. For the stepping-amplitude cyclic loading tests, due to long time vibration, inner and surface cracks fully developed, some specimens even crushed. 3

(d)

THE NONLINEAR ACCUMULATIVE HIGH-FREQUENCY FATIGUE DAMAGE FORMULA OF C30 PLAIN CONCRETE

Based on the axial fatigue testing results of high strength concrete, Wu modified the linear accumulative fatigue damage formula with consideration of the discreteness of fatigue tests and the effect of loading order. He put forward the following formula for computing the accumulative fatigue damage and predicting the residual fatigue life: D = γ1 γ2 ·

 ni =1 Nfi i

(e)

Figure 3. si (ni )∼ni curves under constant-amplitude cyclic loads at five stress levels.

of loading order respectively. Nfi is the fatigue life and ni is the cyclic loading number, both correspond to the stress level i. Cao (2004) replaced γ1 γ2 with the function si (ni ), and gave the accumulative damage formula at the

(1)

where, γ1 and γ2 are correction coefficients in consideration of the discreteness of fatigue tests and the effect

65

stress level i after ni cycles of loading: Dt = si (ni )

ni Nfi

(2)

Figure 3 shows si (ni ) ∼ ni curves for concrete specimens in the constant-amplitude cyclic loading tests at 5 stress levels. The following equations are obtained through curve fitting. Smax Smax Smax Smax Smax

= 0.73 : s(n) = 2.08524n−0.99139 , R = 1; = 0.75 : s(n) = 2.41503n−0.99854 , R = 1; = 0.77 : s(n) = 1.48873n−0.41478 , R = 0.95248; = 0.80 : s(n) = 6.28489n−0.98726 , R = 0.99974; = 0.85 : s(n) = 1.51394n−1.01304 , R = 0.99994;

In the above-listed equations, Smax is maximum stress level, R is correlation coefficient. Five above-listed equations could be unified as: si (n) = ai nbi

index, and σf is the coefficient of fatigue strength. Basquin equation is also suitable for the description of σa ∼Nf relation in high-frequency fatigue tests. By making logarithmic transformation in both sides of Equation 5,

(3)

where ai and bi are material properties corresponding to the stress level i, which can be obtained by constantamplitude cyclic loading tests. Hence, the following accumulative damage formula is suggested.  b ni  ni si (ni ) = ai ni i · (4) D= Nfi Nfi i i

lg σa = lg σf + b lg(2Nf )

High frequency: σa = 0.774587 × (2Nf )−0.06597 (7) Low frequency: σa = 0.426177 × (2Nf )−0.03455

(8)

The fatigue limit σa corresponding to different cyclic loading number can be calculated from Equations 7 and 8.

5

BASQUIN FUNCTIONS

Figure 4 shows lg S ∼ lg Nf curves obtained from normal low-frequency fatigue tests and high frequency fatigue tests with constant-amplitude cyclic loads respectively. The data of low frequency fatigue tests are obtained from the reference (Cao 2004) and the compressive strength of concrete is 20.47 MPa. Compared with the data of low-frequency fatigue tests, it could be found that at the same stress level, fatigue life in high-frequency fatigue tests is longer, which reflects the effect of loading frequency on fatigue performance of concrete. In 1910, Basquin established the famous Basquin equation for constant-amplitude cyclic loading tests, which shows the relationship between stress amplitude and fatigue life. σa = σf (2Nf )b

(6)

The Basquin equations for the curves in Fig. 7 are:

Equation 4 is not only suitable for computing the fatigue damage under constant-amplitude cyclic loads, but also suitable for analyzing the development of fatigue damage under stepping-amplitude cyclic loads. 4

Figure 4. lgS∼lgNf curves under high-frequency and low-frequency fatigue loads.

CORRECTION COEFFICIENT FOR HIGH FREQUENCY FATIGUE TESTS

According to Chinese national norm, N = 2 × 106 is adopted as the limit low-frequency fatigue life of concrete for designing highway bridge. From the comparison between the low-frequency and high-frequency fatigue test results in Fig. 7, it can be drawn that high frequency can improve the fatigue characteristics of concrete. When the results of high-frequency fatigue test are applied to solve the practical low-frequency fatigue problems, the effect of high loading frequency should be considered. 5.1

The definition of correction coefficient

High frequency correction coefficient is defined as the ratio of fatigue limit σaL under low frequency condition to fatigue limit σaH under high frequency condition:

(5)

where σa is the stress amplitude, Nf is the fatigue life corresponding to the stress amplitude σa , b is Basquin

φ=

66

σaL σaH

(9)

Normally, with the increase of the loading frequency, the anti-fatigue characteristics of materials are improved, so φ is less than 1. The smaller φ is, the more positive is the reaction of the materials to the high loading frequency. Conversely, the closer φ to 1 is, the smaller is the influence of loading frequency to fatigue characteristics. When φ is equal to 1, the high loading frequency and the low loading frequency have the same fatigue characteristics. From Equations 5 and 9, the following equation is obtained: φ = φ0 (2Nf )

b

5.2

If the high-frequency correction coefficient φ is known, then the fatigue characteristics of concrete under lowfrequency loading condition could be obtained from the results of high-frequency fatigue test through the following equation:  σaL = φσfH (2Nf )bH

σfL  σfH

(10)

b = bL − bH

φ0 is the ratio of strength coefficient of low frequency fatigue test to that of high frequency fatigue test, while b is the difference between strength index of low frequency fatigue test and that of high frequency fatigue test.

6 6.1

Cycle 1× 1 × 105 1 × 106 2 × 106

0.403 0.346 0.297 0.284

104

0.303 0.280 0.258 0.252

FATIGUE CURVES AND FATIGUE EQUATION Fatigue curves

S – lg N data points from the high-frequency fatigue tests and the corresponding fitting curves are shown in Figure 6. S is the stress level and N is the corresponding fatigue life.

Table 4. High-frequency and low-frequency fatigue limits along with high-frequency correction coefficient corresponding to various fatigue cycles. High frequency High frequency Low frequency correction fatigue limit fatigue limit coefficient φ

(11)

Table 4 shows the high-frequency correction coefficient φ corresponding to different fatigue life from 1 × 104 to 2 × 106 , which is obtained from high and low frequency fatigue tests. Figure 5 shows two lg σa ∼ lg(2Nf ) curves obtained from modified results of high-frequency fatigue test and results of low-frequency fatigue test, respectively. Two curves are very good congruent. So it is feasible to calculate the results of concrete fatigue under low-frequency condition from that of high-frequency fatigue tests.

where φ0 =

The correction of high-frequency fatigue test results

6.2

0.868 0.887 0.807 0.751

P–S–N fatigue equation considering failure possibility

For the given failure possibility p , the equivalent  1 fatigue life N = η ln(1 − p ) b can be calculated. The results are listed in Table 5. Table 6 shows the regression coefficients c, d and correlation coefficient r that

Figure 5. lg σa ∼ lg(2Nf ) curves obtained from modified curves of high-frequency fatigue tests and low-frequency fatigue tests.

Figure 6. S− lg N data points from the high frequency fatigue tests and the corresponding fitting curves.

67

The fatigue equation with failure possibility 50% is:

Table 5. Equivalent fatigue life corresponding to failure probability, p.

lg S = 0.08502 − 0.0038 (1 − R2 ) lg N

Stress level S Failure probability p

0.85

0.80

0.77

0.75

0.73

0.05 0.1 0.2 0.3 0.4 0.5

348 1148 3986 8677 15744 26119

18562 24481 32669 39124 44918 50511

161234 164712 168418 170776 172605 174174

179158 273516 425147 560093 691754 827678

152066 224914 338242 436497 530647 626441

7

Regression coefficient d

Regression coefficient c

Correlation coefficient r

0.05 0.1 0.2 0.3 0.4 0.5

0.9645 1.0158 1.0882 1.1425 1.1849 1.2162

0.0212 0.0252 0.0304 0.0339 0.0364 0.0380

−0.9331 −0.9557 −0.9734 −0.9749 −0.9656 −0.9467

CONCLUSIONS

The following conclusions can be drawn: 1. There are two main failure modes of plain concrete in the high frequency fatigue tests. They are vertical failure and conical failure. 2. The results of low-frequency fatigue tests can be calculated from the results of high-frequency fatigue tests using high-frequency correction coefficient φ. 3. It is more reasonable to establish the P–S–N fatigue equation corresponding to certain failure possibility. 4. The research work contributes to the discussion of the characteristics of fibre reinforced concrete column in high-frequency fatigue tests.

Table 6. Regression coefficients c and d along with correlation coefficient r, corresponding to various fatigue failure probability, p. Failure probability p

(13)

Figure 7 shows the lg S∼ lg N fatigue curves with different failure probability. REFERENCES Aas-Jakobsen K. Fatigue of concrete beams and columns. NTH Institute for Betonkonstruksjoner Bulletin, 1970, 70(1): 148. Alliche A. Damage model for fatigue loading of concrete. International Journal of Fatigue, 2004, 26(9): 915–921. Cao W. Experimental and theoretical research on fatigue properties of plain concrete under triaxial cyclic loading with constant lateral pressure. Doctoral Dissertation for Dalian University of Technology, 2004. Hsu T.C. Thomas. Fatigue of plain concrete [J]. ACI Materials Journal, 1981, 78(8): 292–305. Liang Y. Robert & Zhou J. Prediction of fatigue life of asphalt concrete beams. International Journal of Fatigue, 1997, 19(2): 117–124. Matsushita H. & Tokumitsu Y. A study on compressive fatigue strength of concrete considered survival probability. Proceeding of JSCE, 1972, 198: 127–138. Mihashi H. Stochastic approach to study fatigue of concrete. Engineering Fracture Mechanics, 1987, 28(5–6): 785–793. Sain T. & Chandra Kishen J.M. Energy-based equivalence between damage and fracture in concrete under fatigue. Engineering Fracture Mechanics, 2007, 74(15): 2320–2333. Wu G. & Lv Z.T. Study on the stress-strain relationship of FRP-confined concrete circular column without a strainsoftening response. Journal of Building Structures, 2003, 24 (5): 1–9. Wu P.G. & Zhao G.Y. & Bai M.L. Fatigue behavior of high strength concrete under compressive cyclic loading. China Civil Engineering Journal, 1994, 27 (3): 33–4.

Figure 7. lg S∼ lg N fatigue curves corresponding to different failure probabilities.

correspond to various fatigue failure probability p , which could be obtained from the linear regression of the data in Table 5 using lg S = lg d − c(1 − R2 ) lg N . p could be determined from the reliable possibility requirement, and then c, d could be received from Table 6. P–S–N equation can be obtained by substituting c, d into: lg S = lg d − c (1 − R2 ) lg N The fatigue equation with failure possibility 5% is: lg S = −0.01568 − 0.0212 (1 − R2 ) lg N

(12)

68

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Experimental study of self-centering RC frames with column yielding mechanism K. Sakino & H. Nakahara Kyushu University, Fukuoka, Japan

ABSTRACT: This paper discusses the hysteresis behavior of a precast concrete structural system. The system is proposed based on innovative concepts for precast concrete frames using precast concrete units assembled by unbonded post-tensioned (PT) high strength steel bars. The precast concrete units are classified into three types according to their shapes. Each concrete unit is composed of one or two reinforced concrete beams and one or two plain concrete columns confined by steel tube. PT steel bars run through conduits embedded in the column of the precast concrete units. Each unit is connected to other unit with a connecting device and PT steel bars at a column-to-column joint at the mid-height portion of the columns in frames. After that, the reinforced concrete beams of the precast concrete units are connected by cast-in-sight concrete. The most significant characteristic of these unbonded PT precast concrete systems is their self-centering capability that results in essentially no residual drift after seismic events. The structural behavior under cyclic lateral load of a proposed precast concrete frame was examined by experimental study. 1

INTRODUCTION

In many reinforced concrete buildings, reinforced concrete frames and structural walls appear together. When lateral force resistance is provided by the combined contribution of frames and structural walls, it is customary to refer to them as a wall fame system, a dual system or a hybrid system. The Japanese term related to this type of system can be translated literally as wall frame system. A goal in our research project is to establish a reliable performance-based seismic design (P-BSD) method for dual system buildings in which following abilities are required.

(a) 1. To control the largest story drift angle induced in the buildings during major earthquake ground motions (EQGMs) within limited value such as 0.01 rad 2. To control damages of non-structural elements as less as possible 3. To decrease permanent (residual) drift as small as possible 4. To avoid an expensive post-earthquake repair process 5. To afford large column-free spaces which could be easily remodeled 6. To be easily demolished when necessary A proposed example of the prototype model for dual system buildings which might satisfy the demands described above is shown in Figure 1. The prototype building, which is so called double tube system, are composed of following structural elements.

(b) Figure 1.

69

Prototype building: (a) plan; and (b) section.

capability that results in essentially no residual drift after EQGMs. The three-bay and two-story posttensioned concrete frame system reported in this paper is shown in Figure 2. Figure 3 shows the deformed configuration of an unbonded post tensioned precast concrete beam and column sub-assemblage under lateral load due to EQGM. The behavior under lateral load is governed by the behavior of the interfaces between beams and columns which are fabricated monolithically. Opening of these interfaces at a selected level of lateral load provides a softening of the lateral load-drift behavior, in other words, a kind of hinges are formed at the top and bottom of the columns. Upon unloading, the post-tensioning force and the axial force due to gravity load tend to restore the structure to its original position (self-centering) by closing the gaps at the open interfaces. The characteristics and merits of the unbonded post-tensioned precast concrete frame system are itemized as follows.

Composite Structural Wall: The rolls of this element are to avoid weak story, to afford lateral stiffness and to behave as hysteretic dampers. Frame System for Peripheral Tube: The roles of this frame system are to afford torsional rigidity of buildings, to sustain gravity load and to behave as a self-centering system which brings essentially no residual drift after EQGMs. Truss Girder with Hysteretic Damper: The roles of this element are to afford an outrigger action to structural wall and to absorb earthquake energy by a bottom cord, which is attached to the structural wall, designed as hysteretic damper. Post Column: The rolls of this element are to sustain gravity load and to resist overturning moment induced by earthquakes together with truss and structural wall. An elastic-plastic behavior of the composite structural wall systems has been experimentally investigated and reported elsewhere (Sakino et al. 2004, Sakino & Hitaka 2006). The objectives of this paper are to propose a self-centering RC structural frame for the frame system for peripheral tube and to experimentally investigate a seismic performance of the new self-centering frame. The proposed self-centering frame system is different from a self-centering system proposed in the USA (Seo & Sause 2005) from the view points of a construction method and a collapse mechanism under lateral loading.

2

1. Cracking in slab system and elongation of beams are not occurred, since the collapse mechanism of the frame is formed by openings of the interfaces between beams and columns as shown in Figure 3. This ensures a rigid-slab hypothesis. 2. As the position of the story in the frame is lower, the lateral load carrying capacity of the story increases due to the gravity load affect. 3. From the view point of a post-earthquake repair process, no hinging actions in beams are preferable. 4. On the other hand, the unbonded post-tensioned precast frame system has following defects. 5. An energy absorption capacity is hardly expected, because nonlinear behavior of the concrete in compression (refer to the deformed configuration shown in Figure 3) results in a narrow hysteresis loop.

SELF-CENTERING FRAME SYSTEM

The frame system proposed in this paper is composed of precast concrete beam and column units assembled by using connecting devices and unbonded PT bars. The most significant characteristic of these unbonded PT precast concrete members is their self-centering

Figure 2.

Self-centering frame system studied.

70

The column-to-column connecting device shown in Figure 4 is fabricated by using H shape steel so called as H-200 × 200 × 8 × 12 with stiffeners. These connecting devices were placed at the mid-height portions of the columns in frame. The eight connecting devices and the precast concrete units, which were eight T shape units and four cruciform units, were assembled by using PT bars at the site where the test was conducted. After that, the reinforced concrete beams of the precast concrete units are connected by cast-insite RC beams. Eight deformed bars with nominal diameter of 19 mm (D19 bars) were used as longitudinal bars of beams, and D 6 bar hoops were placed at 50 mm space for transverse reinforcements. At the ends of three-span continuous beams, steel plates with 19 mm thickness were placed to be used as anchor devices for beam longitudinal bars which were welded to the steel plate. The fabrication of the frame specimen showed that the construction procedure described above was successfully verified to be feasible. The properties of steel materials are shown in Table 1. The cylinder strengths of concrete used for precast concrete units and cast-in-site concrete for connection RC beams were 38.4 MPa and 32.1 MPa, respectively.

Figure 3. Schematic deformed configuration of unbonded PT precast column.

6. A story collapses mechanism, which should be avoided during EQGMs, could be introduced to the frame system. 7. Serious damages of column concrete could accompany opening of the interfaces between beams and columns. These defects in the self-centering frame system can be overcome by introducing the composite structural walls with hysteretic dampers or by confining the column concrete by a steel tube which has a role of a formwork as well.

Column

H steel shape Load cell

3 3.1

PT bar

EXPERIMENTAL PROGRAMS Stiffener

Test specimen

The matters of major interest in the experimental study were to examine feasibility of the construction method and the structural behavior of the proposed self-centering frame system. A specimen was threebay, two-story model, and was scaled to 1/3 in order to utilize the available test facility. The dimensions and detailing of the specimen are shown in Figure 2, where an exterior view of the specimen is shown in left half, an arrangement of steel bars and PT bars is in right half. The sections of beams and columns are also shown in Figure 2. The specimen is composed of precast concrete units and column-to-column connecting devices. The precast concrete units are classified into two types according their shapes, i.e. T shape type and cruciform type. Each concrete unit is composed of two reinforced concrete beams and one or two plain concrete columns confined by square steel tube. A space of 10 mm width is provided between the beam surface and the steel tube which is used as a formwork for column concrete. Four conduits to thread the PT bars are embedded in each column.

Figure 4. Table 1.

Column-to-column connecting device. The properties of steel materials.

Type of steel  200 × 200 × 6 D19 Steel bar D6 Steel bar H200 × 200 × 8 × 12 13φ Post tensioned bar

71

Yong’s Elongaσu modulus tion σy (MPa) (MPa) (GPa) (%) 397 394 333 325 306 1243

474 548 503 456 464 1288

205 173 177 207 204 192

34.3 16.8 15.1 41.8 37.3 11.0

5MN Testing machine Wire

2400 Roller Loading beam

Counter weight

1440

640

1600

H-300×300×9×14

Wire

Hydraulic jack(500kN)

Roller Load cell(500kN)

Pin

Retaining wall

Counter weight Twist preventer

Twist preventer

2350

PC rod

Mechanism to maintain the same story drift angle for the first and second stories

Round steel bar PC rod

FL

800

800

800

800

800

800

800

5600

Figure 5.

Loading set-up. 3

2

1

0

-1

-2

Figure 6.

3.2

-3

Axial force in each column.

0

2

4

6

8

10

12

Experimental apparatus and procedure Figure 7.

A loading method is schematically shown in Figure 5. The vertical load corresponding to gravity load in the columns was applied at first to each column by using testing machine and three steel loading beams acting as vertical load distributor. The vertical load applied by testing machine was 600 kN and was kept constant during test. After applying the vertical load through testing machine, additional axial loads in columns were introduced by PT bars. The loads in PT bars were measured by specially made load cells shown in Figure 4. The axial load in each column introduced by testing machine, weight of loading beams and post tensioned bars is shown in Figure 6. The horizontal loads were applied to the specimen in a manner of pushing in both directions as shown in Figure 5, which introduced the compression axial force into the reinforce concrete beam. The loading pattern was a cyclic type with alternating drift reversals. The peak drifts were increased stepwise from 0.005 h, where h was the total height (2200 mm) of the two-story specimen, until 0.02 h with incremental drift of 0.005 h after

Loading program.

three successive cycles at each drift level as shown in Figure 7. It is noteworthy that the story drift angles of the first and second story were kept to be the same during the test by the special mechanism attached to the specimen as shown in Figure 5.

4 4.1

EXPERIMENTAL RESULTS Load-deformation relationships

Figure 8 shows a relation between the lateral load and story drift angle. The maximum lateral load at each amplitude increases, as the drift amplitude becomes larger. The larger drift amplitude results in larger opening at the interfaces between columns and beams as shown in Figure 3, hence larger forces in PT bars. This is verified by Figure 9 which shows a relation between story drift angle and total elongation of

72

columns divided by total column height. The values of total elongation are taken as average values measured in four two-story columns. The increase of maximum lateral load can be attributed to an increase of axial loads in columns due to the increase of forces in PT bars. This phenomenon suggests that an introducing of larger initial forces in PT bars could bring larger lateral load capacity at relatively small story drift amplitude. As shown in Figure 8, the proposed frame system has the self-centering capability as expected. A detailed and quantitative estimation of self-centering capability is shown in Figure 10 which describes a relationship between story drift amplitude and residual story drift. As shown in Figure 10, the average residual story drift shown by dotted line can be estimated as small as 10% of the maximum story drift undergone. An observation of the specimen after the test revealed that there was substantially no damage in appearance. Figure 11 shows relations between tension forces in four PT bars and story drift angles. The four PT bars are named as A, B, C, D, and they are referred as PT-A or PT-B and so on, hereafter. The places of these four PT bars are shown in Figure 6 under the same symbols A, B, C, D. As shown in Figure 6, PT-A and PT-C are placed in bottom half of left side column, and

Figure 10. Relationship between residual story drift and story drift undergone. 100

50

0

-50

-4

-3

-2

-1

0

1

3

2

4

(a) 100

50

Figure 8.

0

Experimental cyclic lateral load-drift behavior.

-50

-4

-3

-2

-1

0

1

2

3

4

(b)

Figure 11. Tension force-drift behavior of PT bars: (a) PT-A, B; and (b) PT-C, D.

PT-B and PT-D are in middle part of the same column. The tension forces in PT-A and PT-C are shown by solid lines in Figures 11(a) and (b), and those in PT-B and PT-D are by dotted lines. As shown by solid lines in Figure 11, tension forces in PT bars in bottom half of the column are not fluctuated when they are placed in compression side due to bending moment, and are

Figure 9. Relationship between axial elongation of columns and story drift.

73

where, s M is a bending moment component of four PT bars in the column, c M is a ultimate moment capacity of plain concrete section of the column under an axial force Nm , c σB is a compressive strength of concrete confined by square steel tube (Sakino & Sun 1994), c D is a depth of concrete column section and c b is an effective width of concrete column section which is defined by neglecting an ineffective width due to the conduits as shown in Figure 13. It is noteworthy that the axial force Nm is calculated by considering shearing forces in the beams and a tension axial force component of the resultant force of the four PT bars in the column measured by load cells. As shown in Figure 12, the calculated lateral load capacities of the first and second stories, which are 312 kN and 313 kN, respectively, are almost same as the experimental lateral load capacity of 320 kN.

Figure 12. External load and internal force in the frame at the ultimate state.

5

CONCLUSIONS

The following conclusions are reached on bases of the experimental study on the self-centering RC structural frame compose of precast concrete units, column-tocolumn connecting devices and unbonded post tensioned bars. Figure 13.

Ineffective width due to conduits.

1. It is verified to easily fabricate the three-bay, twostory self-centering frame by construction procedure proposed in this paper. 2. The hysteresis behavior of the frame under cyclic lateral loading was stable, and showed self-centering capability. 3. The value of maximum residual story drift angle was less than 0.002 rad as expected. 4. The lateral load capacity of the frame specimen can be predicted with a very reasonable accuracy by the calculation method used in this paper.

increased when in tension side. This means that resultant force of four PT bars in the bottom half (and/or top half) of the column has a bending moment component as well as tension force component. On the other hand, the dotted lines in Figure 11 show that tension forces in PT-B and PT-D behave in a similar manner. This means that resultant force of four PT bars in the middle part of the column has only axial-tensionforce component. The hysteresis behavior of PT bars in the column described above is consistent with the deformed configuration of unbonded post tensioned precast concrete column shown in Figure 3. 4.2

REFERENCES Sakino, K. & Sun. Y. 1994. Stress-strain curve of concrete confined by rectilinear hoop. Journal. Structural and Construction, Engineering. AIJ, 461, 95–104. Sakino, K. et al. 2004. Cyclic test on composite wall-frame subassmblage. 13th World conference on earthquake engineering, Vancouver, B.C., Canada, August 1–6, 2004 Paper No. 3172. Sakino, K. & Hitaka, T. 2006. Experimental study on overturning moment-resisting structural walls with steel hysteretic dampers. Prod. of 8th National conference of earthquake engineering, DVD-ROM, 2006, April 18–23. Seo, C.Y. & Sause, R. 2005. Ductility demands on self-centering systems under earthquake loading. ACI Structural Journal, Title No. 102-S28, pp. 275–285, March–April 2005.

Ultimate lateral load capacity

Figure 12 shows external vertical and lateral loads applied by loading system and shearing forces in columns at the loading point when the story drift reached the largest value of 0.02 h at first. The lateral load of 320 kN is the maximum lateral load measured in this test. The shearing force, Qc , of each column is a value calculated from the bending moment capacities, Mc , at the top and bottom of each column. The bending moment of columns is obtained by Equation (1). Mc =s M  + cM = N2m c D −

cM

Nm c σcB ·c b



(1)

74

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Experimental study on high-strength R/C member in tension and shear T. Tamura Tokuyama College of Technology, Tokuyama, Japan

ABSTRACT: The relations between the axial tensile force and the shear strength of the high-strength R/C member are discussed from experimental data. In the experiment, 16 specimens were provided that have various compressive strengths of concrete and have two types of tensile strengths of rebar. Test results showed that the members subjected to axial tension had decreased load-carrying capacity. Furthermore, it became clear that in the case of a higher strength concrete beam subjected to axial tension, the shear strength of the beam declined rapidly. On the other hand, it was confirmed that the axial tensile force shifts the destruction mode from shearing to bending. In this case, the axial tensile force contributes to a reduction of the compressive stress in the concrete compression region, when the ultimate loading capacity of the member increased finally.

1

INTRODUCTION Mu = (As · fyd − As · σs ) · [d − (a/2)]

Recently in cities, many superstructures are built on large and tall scales. High-strength concrete has been used as a primary building material, because has about 3–5 times the compressive strength and higher durability in comparison with normal concrete. These characteristics are used to make many skyscrapers and to reduce the cross-section of the members and so on. The effective use of the land and the expansion of the residential space due to the span expansion became possible as a result. The fracture property of a reinforced concrete member has been studied by many researchers. From those results, although the shear fracture is complicated, the fracture mechanism was clarified and the design equation was sufficiently accurate. However, in the case of the member subject to the axial force and shear, there are only a few studies. In a previous paper, the author reported on a reinforced concrete beam that underwent axial force and bending. In the case of high-strength concrete, the investigation of a similar situation is very important. This research aims to clarify the fracture mechanism of a high-strength concrete member subjected to axial force and shear. 2 2.1

+ As · σs · (d − d  ) ± N · I /(A · y)

(1)

where Mu = ultimate moment (kN · cm); As = area of the tensile rebar, As = area of the compressive rebar; fyd = yield stress of rebar; σs = compressive stress of rebar; d = effective depth; d  = depth of the compressive rebar; a = equivalent depth of concrete compressive area; N = axial force (positive for compression); I = moment of inertia; A = area of concrete; y = distance to the tensile rebar from the center of gravity. 2.2

Nominal shear strength

Nominal shear strength of a reinforced concrete (R/C) member is generally determined by concrete strength, reinforcement ratio, effective depth of a cross-section, shear span to depth ratio, and applied axial force. For example, in the JSCE design equation, the shear strength of an R/C beam with shear reinforcement (Vyd ) is calculated as the following Equation 2. However, in JSCE, an upper limit is set on the term of the concrete design strength if the concrete compressive strength exceeds 0.72 N/mm2 .

ULTIMATE STRENGTH

Vyd = Vcd + Vsd

Ultimate bending strength

(2)

where,

The ultimate bending strength of the member considering the axial force is calculated by the following Equation 1. The influence (i.e., stress intensity) by axial tension is taken into consideration by conversion into the bending stress intensity.

Vcd = βd · βp · βn · fvcd · bw · d/γb  fvcd = 0.20 3 fcd (N/mm2 )

75

(3) (4)

βd =

where Nd is the design axial load, which is taken as positive for compression and negative for tension. Md is the ultimate moment, and Mo , named the decompression moment, is the bending moment when the axial stress is calculated in bending. If Mo has the same signs to Md , it is taken as positive. Also, bw = web width; d = effective depth; pv = As /(bw · d); As = area of tensile rebars; fcd = concrete design compressive strength; γb = 1.3. Vsd is the shear strength, which is covered by the shear reinforcing bar.

fvcd ≤ (N/mm 2 )

Here,

  4 1 d (d : m)

βd is 1.5 when βd > 1.5  βp = 3 100pv βp is 1.5 when βp > 1.5 βn = 1 + Mo /Md (Nd ≥ 0) βn is 2 when βn > 2 βn = 1 + 2Mo /Md (Nd < 0)

Vsd = [Aw fwyd (sin αs + cos αs )/ss ]z/γb

βn is 0 when βn > 0.

where Aw = total cross-sectional area of shear rebar within the ss ; fwyd = design yield strength of shear rebar, less than 400 N/mm 2 ; however, in case of the characteristic compressive strength of concrete σck over 60 N/mm2 it is less than 800 N/mm2 ; αs = an angle toward the axis of the center of gravity; ss = Interval of the arrangement of the shear rebars; z = distance from center of the tensile rebar to center of the compressive stress; γb = 1.1

16@100=1600 D13

180 100

150

Figure 1. Table 1.

25

1500 1800

200

6

150

(mm)

Specimen with stirrup. Material properties of rebar.

3

Type of rebar

D13(L)

D13(H)

φ6

Yield stress (σy : N/mm2 ) Tensile strength (ft : N/mm2 ) Elastic modulus (Es : kN/mm2 )

334 457 202

401 643 208

226 282 193

Table 2.

(5)

EXPERIMENTAL PROGRAM

3.1

Specimens

Figure 1 shows the test beam subjected to axial force. Sixteen specimens were provided. All of the specimens for the experiment have the same dimensions. Three

Bending strength. Experimental results

Specimen

N

L-A1-0 L-A1-T40 L-B1-0 L-B1-T40 L-C1-0 L-C1-T40 L-D1-0 L-D1-T40

0 40 0 40 0 40 0 40

H-A2-0 H-A2-T40 H-B2-0 H-B2-T40 H-C2-0 H-C2-T40 H-D2-0 H-D2-T40

0 40 0 40 0 40 0 40

Pv

Mv

Pb

Mb

Mode

Mu

My /Mu (Mb /Mu )

32.9 32.9 43.5 44.4 63.7 66.6 82.5 84.9

– 67.1 – 68.4 79.2 67.6 82.0 72.0

– 1933 – 1970 2282 1946 2361 2073

69.3 69.5 73.4 80.6 107.1 99.2 83.7 91.9

1995 2002 2106 2321 3085 2856 2411 2647

S M S M M M M M

2149 1944 2177 1974 2224 2026 2263 2063

(0.93) 1.00 (0.97) 1.00 1.03 0.96 1.04 1.00

27.0 27.0 33.4 33.4 68.9 68.9 104 104

– – – 74.6 87.1 73.8 – 80.0

– – – 2147 2508 2125 – 2303

66.9 62.8 72.1 84.8 96.0 107 80.6 79.5

1927 1809 2078 2442 2765 3084 2321 2290

S S S M M M S S

2550 2345 2569 2364 2658 2453 2727 2522

(0.76) (0.77) (0.81) 0.91 0.94 0.87 (0.85) (0.91)

fck1

N : Expected axial force (tension) (kN) fck1 : Concrete compressive strength (N/mm2 ) Pv : Yielding load (kN) Mv : Yielding moment (kNcm) Pb : Breaking load (kN) Mb : Breaking moment (kNcm) Mu : Ultimate bending moment calculated by equation (1) (kNcm) S: Shear failure M: Bending failure.

76

deformed bars (D13) are placed as the tensile reinforcements and the compressive reinforcement, respectively. Then, there are two types of steel strength: fy = 295 N/mm2 and 390 N/mm2 . Thirteen stirrups (6) are placed at 15 cm intervals. There is a hole at the both ends of the beam to introduce the axial tension and the points are the supporting points of the all beams. The material properties of both the main and shear rebars are shown in Table 1. Also, Table 2 shows the material properties of concrete. There are four types concrete compressive strengths: approximately 24, 40, 70, and 100 N/mm2 .

and is held constant after reaching the expected tension. Next, the transverse loads are provided by the transverse actuator that introduces the load onto two points by the loading beam. The transverse load increases continuously until the beam fails under the displacement controlled system. To determine the shear strength of the member subjected to axial tension, a non-stressed test carried out. During the loading test, new cracks are marked on the face of the beam at each loading stage. Dial gauges are placed at the loading point and the center of the span to measure the deflection of the beam. Then the bending strain is measured by wire strain gauges at the canter of the tensile reinforcement and the top of the beam.

3.2 Test apparatus and procedure The testing apparatus for the test with axial tensile force is shown in Figure 2. It is composed of two oil pres-sure actuators controlled by the electro-hydraulic servomechanism. Also both supporting points are hinged by bearing joints. In the test of the member subjected to axial tension, the axial tension is introduced onto both ends of the beam via a longitudinal actuator

4 4.1

Longitudinal actuator Specimen

Dial gauge

Figure 2. Table 3.

Ultimate shear strength

Table 2 shows the experimental conditions, experimental results and the ultimate bending strength, Mu , calculated by the Equation 1. The ultimate state of the fracture mode of the member was divided roughly into bending failure (M) and the shear failure (S) in the Table 2. Table 3 shows the ultimate shear strength, Vb , and nominal shear strength, Vyd , calculated in Equation 2 and the terms of the equation. From these tables, it is observed that the beam of type L is in the transition area where a beam fails in bend mode or in shear mode. In Table 2 and Figure 3, My /Mu shows the precision of Mu calculated by Equation 1 against the experimental yield strength My . When the member failed in shear,

Transverse actuator Hinged support

EXPERIMENTAL RESULTS

Test apparatus for bending test with axial tension. Shear strength.

Specimen

fvcd (N /mm2 )

βd

βp

βn

Vcd (kN )

Vsd (kN )

Vyd (kN )

Vb (kN )

Vb /Vyd

L-A1-0 L-A1-T40 L-B1-0 L-B1-T40 L-C1-0 L-C1-T40 L-D1-0 L-D1-T40

0.60 0.60 0.68 0.68 0.72 0.72 0.72 0.72

1.5 1.5 1.5 1.5 1.5 1.5 1.5 1.5

1.28 1.28 1.28 1.28 1.28 1.28 1.28 1.28

1.00 0.89 1.00 0.92 1.00 0.94 1.00 0.95

20.7 18.7 23.5 21.6 24.9 23.4 24.9 23.7

13.4 13.4 13.4 13.4 13.4 13.4 13.4 13.4

34.1 32.1 36.9 35.0 38.3 36.8 38.3 37.1

34.6 34.8 36.6 39.1 53.6 49.6 45.9 41.9

1.01 (1.08) 0.99 (1.12) (1.40) (1.35) (1.20) (1.13)

H-A2-0 H-A2-T40 H-B2-0 H-B2-T40 H-C2-0 H-C2-T40 H-D2-0 H-D2-T40

0.60 0.60 0.68 0.68 0.72 0.72 0.72 0.72

1.5 1.5 1.5 1.5 1.5 1.5 1.5 1.5

1.28 1.28 1.28 1.28 1.28 1.28 1.28 1.28

1.00 0.90 1.00 0.92 1.00 0.94 1.00 0.95

20.7 18.7 23.5 21.6 24.9 23.4 24.9 23.7

13.4 13.4 13.4 13.4 13.4 13.4 13.4 13.4

34.1 32.1 36.9 35.0 38.3 36.8 38.3 37.1

33.5 31.4 36.1 42.4 48.0 53.6 40.3 39.8

0.98 0.98 0.98 (1.21) (1.25) (1.46) 1.05 1.07

fvcd , βd , βp , βn : Term of equation (3) Vcd : Shear strength charged by concrete, calculated by equation (3) Vsd : Shear strength charged by shear reinforcing bar, Vyd : Nominal shear strength calculated by equation (2) Vb : Shear strength (experimental results).

77

Figure 3.

My /Mu fc relationships (in bending failure). Figure 4. Vb /Vyd fc relationships (in shear failure).

Figure 5.

Figure 6.

L-A1-0

L-A1-T40

L-B1-0

L-B1-T40

L-C1-0

L-C1-T40

L-D1-0

D2-T40

Ultimate crack state (Member subjected to axial tension, SD295).

H-A2-0

H-A2-T40

H-B2-0

H-B2-T40

H-C2-0

H-C2-T40

H-D2-0

H-D2-T40

Ultimate crack state (Member subjected to axial tension, SD390).

78

80

80 Load (kN)

100

Load (kN)

100

60 40 20 10

20 30 40 Deflection (mm)

50

40

0 0

60

100

100

80

80

60 40 L-B1-0 L-B1-T40

20 0 0

10

20 30 40 Deflection (mm)

50

Load (kN)

40 L-C1-0 L-C1-T40

20

60

H-B2-0 H-B2-T40

40

10

20 30 40 Deflection (mm)

50

60

10

20 30 40 Deflection (mm)

50

80 60 40 20 0 0

60

100

80

80 Load (kN)

100

Load (kN)

50

100

60

60 40 L-D1-0 L-D1-T40

20 0 0

20 30 40 Deflection (mm)

60

0 0

60

80

0 0

10

20

100

Load (kN)

60

20

L-A1-0 L-A1-T40

Load (kN)

Load (kN)

0 0

H-A2-0 H-A2-T40

10

20 30 40 Deflection (mm)

50

H-C2-0 H-C2-T40 10

20 30 40 Deflection (mm)

60

60 40 20 0 0

60

50

H-D2-0 H-D2-T40 10

20 30 40 Deflection (mm)

50

60

Figure 8. Load deflection relationships (Member subjected to axial tension, SD390).

Figure 7. Load deflection relationships (Member subjected to axial tension, SD295).

79

the Mb /Mu shows the precision of Mu calculated by Equation 1 against the experimental ultimate strength Mb . From this table and these figures, it is clear that Equation 1 is expressing the experimental results with sufficient accuracy in the case of the member that failed at the bending moment. In type L, a normal concrete beam failed in the shear mode. Also, a highstrength concrete beam failed in bend mode. However, all of the type L members failed in shear mode when the members were subjected to axial tension. In the type H SD390 rebar used, almost all of the beams failed in shear mode. However, (H-B2-T40), (H-B3-0) and (H-C3-T40) failed in bending mode. The accuracy of Equation 1 is declining in these cases. In Table 3 and Figure 4, Vb /Vyd shows the precision of the nominal shear strength Vyd calculated by Equation 2 against the experimental shear strength Vb . Here, fvcd of the member with high-strength concrete becomes the upper limit value of 0.72 in accordance with the condition of Equation 3. The safety factor γb of concrete material is set to 1.3. From Table 3 and Figure 4, it is observed that, even if the member is subjected to axial tension, Equation 2 expresses the experimental result with sufficient accuracy in the case of the members that failed in shear. However, when high-strength concrete is used, the accuracy of the equation drops compared with the case of the member made of normal concrete. 4.2

drops due to the action of axial tension. Then, the value of the load of the occurrence of diagonal shear cracks and the load of the completion of diagonal shear cracks both decrease. The grade of concrete strength influences the stiffness of the member and is related with the inclination of the initial slope of the load deflection relationship. These figures show that the initial inclination of the member is raised by using high-strength concrete. In all of the series, it is confirmed that the displacement at the maximum load of a beam subjected to axial tension is larger than a beam not subjected to axial tension. 5

To investigate the influence of axial tension on the shear strength of a member using high-strength concrete, an experimental study was conducted. Based on the results, these conclusions can be drawn: 1. Equation 1 expresses the experimental results with sufficient accuracy in the case of members failing at the bending moment. 2. Both Equations 1 and 2 express the bending strength and shear strength with sufficient accuracy for a member using high-strength materials and subjected to axial tension. 3. However, the accuracy of Equation 2 drops in the case of a member using high-strength concrete compared a member using normal concrete.

Ultimate crack state

To clarify the shear strength of a member using high-strength material, a further detailed examination will be required.

Figures 5 and 6 show the ultimate crack state of all members. It is observed that the diagonal shear cracks are completed in all members. Also, it is observed that many cracks increased when the beams were subjected to axial tension. In type L beams not subjected to axial tension, cracks spread in wide area as the concrete strength becomes larger. However, in L-D1-0, localized cracks are occurring despite the bending failure. In type L beams subjected to axial tension, cracks spread in a wide area irrespective of the concrete strength. Also, in type H beams not subjected to axial tension, cracks spread in a wide area as the concrete strength became larger. In type H beams subjected to axial tension, cracks spread in a wide area, except for the shear failed beam. Then the beam with the smallest concrete strength and the beam with the biggest strength failed in shear. 4.3

CONCLUSIONS

REFERENCES JSCE (2002). Standard Specifications for Design and Construction of Concrete Structures. Tamura, T., Shigematsu, T., Hara, T. and Maruyama, K. (1995). A Study of Proposed Design Equation for the Shear Strength of R/C Beams Subjected to Axial Tension, Proc. of JSCE, No.520/V-28, 225–234. Tamura, T., Shigematsu, T, Kadonaga, T. Tokuda, M. (2005). Experimental study on high strength concrete beam subjected to axial force and shear, Proc. of ISEC04, 51–56.

Load deflection relationships

Figures 7 and 8 show the load deflection relationship at the center of the beams. These figures show that the value of the load of the bending crack occurrence

80

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Improving the behavior of reinforced concrete beams with lap splice reinforcement A.M. Tarabia, M.S. Shoukry & M.A. Diab Structural Engineering Department, Alexandria University, Egypt

ABSTRACT: The main objective of this paper is to study the behavior of lap splice of steel reinforcement in tension zones in reinforced concrete beams. An experimental program is conducted on twelve simply supported concrete beams. The main studied variables are: cut-off ratio, lap splices length, the type, spacing and shape of transverse reinforcement in the splice zone. It is concluded that for the same value of the lap splice length as that recommended by Egyptian Code, without the use of transverse reinforcement at the spliced zone, the change of the cut off ratio from 25% to 100% resulted in a reduction in ductility. On the other hand, there is a drastic increase in ductility of beams when transverse reinforcement was used. 1

strength. From these tests, they concluded that stirrups increased splice strength, minimum stirrups as much as 20%, and heavy stirrups up to 50%. The splitting prior to failure gradually developed over the full splice lengths seemed almost to stabilize with a substantial center length remaining un-split until a final catastrophic failure occurred. Jeanty et al. (1988) tested thirteen specimens to study the effect of transverse reinforcement on the bond performance among other variables. The main conclusions of this research were that for beams with and without transverse reinforcement crossing the plane of splitting, the top bar factor was found to be 1.22, which means that the required lap splice length must be increased by 22% for spliced top tension bars. The presence of transverse reinforcement across the plane of potential splitting reduced significantly the required development length for both bottom-cast and top-cast bars. Hamad et al. (2006) investigated eighteen full-scale beam specimens. In this study, the amount of transverse reinforcement, bar size, and the bar type (black or galvanized) were considered. They concluded that in beams without transverse reinforcement in the splice region, surfaces of black and galvanized bars were relatively clean with limited signs of concrete crushing in the vicinity of very few bar lugs. In beams with transverse reinforcement in the splice region, however, there were relatively more signs of concrete crushing adjacent to the bar lugs indicating the positive role of confinement by transverse reinforcement in mobilizing more bar lugs in the stress transfer mechanism between the steel bars and the surrounding concrete.

INTRODUCTION

When reinforcement is spliced together within a concrete beam, it is necessary to overlap the bars long enough for tensile stresses in one bar to be fully transferred to other bars without inducing a pullout failure in the concrete. Most design codes allow the use of bars with lap splice and specify minimum length of the lap as well as the required transverse reinforcement. The Egyptian Code 203-2007 requires the lap splice length; Lo, for flexural members to be taken as the development length; Ld, when the area of steel in the section; As applied, is greater or equal to the required steel area; As required, and the percentage of lapped steel is less than 25% of the total steel area at that section. Otherwise, the lap splice length is taken as 1.3 Ld. For flexural bar terminated in a tension zone, excess stirrups are provided over a distance along each terminated bar from the point of cut off to 0.75 the depth of the element. It should be mentioned that all the previous versions of the Egyptian Code prohibited splicing more than one quarter of the total number of tension bars at the same section. According to ACI 318-05, the minimum length of lap for tension lap splices for Class A = 1.0 Ld and =1.3 Ld for class B. Stirrup area in excess of that required for shear and torsion is provided along each terminated bar or wire over a distance from the termination point equal to three-fourths the effective depth of member. Most of design codes do not specify a specific shape of transverse reinforcement required for spliced bars. Ferguson and Breen (1965) studied thirty-five beams, focusing on bar diameter, stirrups, and concrete

81

2

OBJECTIVES

Table 1.

The main objectives of this study were: 1. To study the behavior of reinforced concrete simply-supported beams with lap splice of tension steel reinforcement zones with different lap splice lengths and arrangements. 2. To obtain a spliced beam that can achieve at least the same strength and ductility of the same beam without any splices using transverse reinforcement with different shapes. 3. To investigate the old condition of the Egyptian code, which was removed from the last version of the code (2007), not to splice more than one quarter of tension steel at the same section.

3

EXPERIMENTAL PROGRAM

Twelve simply supported reinforced concrete beams of dimensions 150 mm × 260 mm × 2600 mm were tested in the reinforced concrete lab, Alexandria University (2008). All the specimens had the same concrete strength, and the same longitudinal reinforcement. Four 10 mm-diameter 400/600 high strength steel were used as tension reinforcement. Plain bars of 6 and 8 mm diameter agree with grade 280/450 were used for stirrups outside the splice zone and top reinforcement respectively. The test setup of the studied beams is shown in figure 1. The studied parameters are given in Table 1 and are discussed in the following sections. Additionally, the required lap splice length as well as transverse reinforcement required by several design codes were obtained in Table 2. Figure 2 shows reinforcement details of some of the test beams.

Details of tested beam specimens.

Average cube strength Splice Beam (N/mm2 ) length

Transverse reinforcement Cutoff shape in the ratio splice zone

D1 D2 D3 D4 D5

36.4 36.3 37.3 38.8 38.4

no splice 54 db∗ 54 db 27 db 27 db

0.00% 25% 100% 100% 100%

D6 D7

35.4 36.4

27 db 27 db

100% 100%

D8 D9

34.7 38.7

27 db 27 db

100% 100%

D10 D11 D12

37.6 36.6 38.0

75 db 27 db 27 db

100% 100% 100%

None None None None Stirrups with additional legs Separate stirrups Rectangular stirrups around spliced bars two interlocking Spirals Continuous rectangular stirrups. None Separate stirrups Separate stirrups

*db : Bar diameter. Table 2.

Calculated lap splice length by several codes. Required Calculated transverse lap splice reinforcement

Code

Egyptian Code 203-2007 540 mm ACI 318-05 300 mm Eurocode 2-1996 455 m

10O6/m /

950

P/2

900

P/2

ϕ6 mm @ 30 mm ϕ6 mm @ 30 mm ϕ6 mm @ 35 mm

950

2O /8 No Lap splice

D1

4 10

3.1 Test groups

2600 10O6/m /

The tested beams are divided into four groups. The main studied parameters were:

2O /8 D4

1. Cut off ratio; Group 1 Cut off ratio is ratio of the spliced area to the total

4 10 10O6/m / 2O /8

270

4 10

2 interlocking spirals O / 6 @ 60mm D8

4 10

Figure 2.

270

4 10

Details of some tested specimens.

area of tension bars of the beams. Three values of cut-off ratio were investigated: 0% (Beam D1), 25% (Beam D2), and 100% (Beam D3). No transverse reinforcement was used in the lap splice zone.

Figure 1.

2. Length of lap splice; Group 2 Three values of lap splice length were investigated: 54 db (Beam D3), 27 db (Beam D4), and 75 db

Test setup.

82

Table 3.

Figure 3.

Details of transverse reinforcement.

(Beam D10). The cut off ratio for beams in Group 2 was 100% with no transverse reinforcement in the lap splice zone. 3. Types of transverse reinforcement; Group 3 Beam D6 included vertical stirrups in the lap splice zone, while vertical stirrups with additional legs in the splice zone were used in Beam D5. Rectangular hoops around the spliced bars were used in Beam D7. Two interlocking spirals were provided in the splice zone in Beam D8. Continuous rectangular stirrups in the lap splice zone were used in beam D9. Beam D4 had no stirrups in the splice zone. All beams of this group had the same lap splice length (27 db) and 100% cut off ratio. The diameter and spacing of transverse reinforcement were 6 mm and 60 mm respectively. The shapes of the used stirrups are demonstrated in figure 3. 4 Stirrups spacing; Group 4 Three different spacing values of vertical separate stirrups were studied: S = 60 mm (Beam D6), S = 90 mm (Beam D12), and S = 120 mm (Beam D11). Beam D4 had no stirrups.

3.2

Beam

Average cube strength; fcu (N/mm2 )

Ultimate load Pu ,KN

Deflection at yield load y ,mm

Deflection at ultimate load u ,mm

D1 D2 D3 D4 D5 D6 D7 D8 D9 D10 D11 D12

36.4 36.3 37.3 38.8 38.4 35.4 36.4 34.7 38.7 37.6 36.6 38.0

80.0 77.5 84.3 52.5 82.5 87.5 85.0 88.0 87.5 87.5 80.0 83.0

11.42 10.00 14.00 *NY 12.58 9.42 26.00 11.61 11.08 9.32 12.00 11.80

16.40 18.74 18.00 7.20 17.75 34.00 36.00 47.50 43.00 27.00 20.00 21.00

Figure 4.

Crack pattern of beam D4 at failure.

Figure 5. failure.

Crack pattern at the bottom of beam D5 at

Figure 6.

Cracks at bottom and side of beam D8 at failure.

Test procedure and instrumentation

Figure 1 shows the details of the test rig. The load was applied using a calibrated hydraulic jack of 200 kN capacity. A strong spreader beam was used to transfer the vertical load to the tested beam through two concentrated loads 900 mm apart. Three dial gauges of 0.01 mm accuracy were used to record deflection at the center of the beams as well as under positions of the two loads. For each beam, at least two electrical strain gauges of 5 mm length and were used to measure steel strain. The load is applied in increments equal to 2.5 kN. 4

Main test results.

strain except for beam D4 where a sudden bond failure occurred. The ratio between the steel strain at the middle to that at the start of the lap splice was almost 0.50. This means that the lap splice worked efficiently along its all length. Figures 4–6 show the crack patterns of some of the tested beams. All beams in Group 1 had no stirrups in the lap zone. For beam D2, with 25% cut off ratio, longitudinal splitting cracks under the longitudinal lap spliced bar only began to form between flexural cracks at the bottom of the beam. These cracks became wide with the increase of load until a splitting of the concrete cover from bars

TEST RESULTS AND DISCUSSIONS

The main obtained results are given in Table 3. The longitudinal steel in all the beams reached the yield

83

cracks at the splice zone, and changed the mode of failure from bond failure to ductile flexural failure. In Group 4, the effect of the spacing of the vertical stirrups was studied. For beam D11, with 120 mm stirrups spacing, yield of longitudinal steel occurred at a load of 70.0 kN. Then splitting cracks appeared just prior to failure and after yield. After yielding, cracks extended and became wider, then, a ductile failure took place at a load of 80.0 kN. In beam D12, with 90 mm stirrups spacing, yielding of longitudinal steel occurred at a load of 77.5 kN. Prior to failure, the splitting cracks became wider, and the flexural crack at the start of the splice severely opened just before failure. At a load of 83 kN, the beam failed due to crushing of concrete in compression at the start of the lap splice. The splitting cracks were sufficiently wide but the use of stirrups prevented the occurrence of splitting failure. For beam D6 with 60 mm stirrups spacing, yield of longitudinal steel occurred at a load of 70 kN. At a load of 87.5 kN, a ductile failure took place. Splitting cracks did not appear at all.

occurred at a load of 77.5 kN accompanied by crushing of concrete in compression. For beam D3, with 100% cut off ratio, longitudinal splitting cracks began to form rapidly at the bottom of the beam between flexural cracks. Just after yield of longitudinal spliced bars, splitting cracks became wider, and failure occurred by splitting and loss of bond between spliced bars and concrete at a load of 84.3 kN. In Group 2, all beams had no stirrups in the lap zone. For beam D4 with a splice length = 27 db, flexural cracks propagated upward to the compression zone. At a load of 50 kN, a horizontal splitting crack along the splice length appeared. and a sudden bond failure occurred at a load of 52.5 kN. For beam D10, with lap splice length = 75 db, the extension of the flexural cracks at lap splice zone in beam D10 was about 0.6 the height of the beam. After the load reached 80 kN, the flexural and flexural shear cracks began to extend upwards with a slow rate. Just before failure, very narrow longitudinal splitting cracks occurred under the two ends of the splice, without a splitting failure. A ductile flexural failure occurred by crushing of concrete in compression nearby the concentrated load at a load of 87.5 kN. In Group 3, all beams had 100% cut off ratio and length of lap splice = 27db. For beam D5, yield of longitudinal steel occurred at a load of 80 kN. Compared with beam D1, and D4 (without transverse reinforcement), splitting cracks did not appear. After yielding of tension steel, the crack extended upwards, cracks became wider. Failure occurred at a load of 82.5 kN due to crushing of concrete in compression. In beam D6, with vertical stirrups, with load increase, flexural cracks propagated toward the compression zone. Yield of longitudinal steel occurred at a load of 70 kN. At a load of 87.5 kN, a ductile flexural failure took place due to crushing of concrete in compression with no splitting cracks. In beam D7, with rectangular stirrups in the tension splice zone, yield of longitudinal steel occurred at a load of 82.5 kN. Before failure, all cracks extended to be very close to the top surface of the beam and were concentrated around the applied load positions. At a load of 85.0 kN, a ductile failure took place without the occurrence of any longitudinal splitting cracks parallel to the bars. In beam D8, with two interlocking spirals in the splice zone, yield of longitudinal steel occurred at a load of 72.5 kN. Away from the splice zone, almost all the cracks extended to be very close to the top surface of the beam, and the vertical extension of the cracks was slow. No splitting cracks were observed. At a load of 88.0 kN, a ductile failure took place. For beam D9, with rectangular continuous stirrups in the splice zone, yield of longitudinal steel occurred at a load of 72.5 kN. No splitting cracks at the splice zone took place. A ductile flexural failure took place. It is clear that the use of transverse reinforcement eliminated the occurrence of splitting

4.1

Effect of cut off ratio (Group 1)

Figure 7 shows the relationship between load and midspan deflection for beams D1, D2, and D3. The figure shows that at any level of loading the mid-span deflection of beam D3 (100% cut off ratio) was less than that of beam D1. This may due to the doubling of reinforcement at mid-span section of beam D3. However, the deflection at mid-span of beam D2 (25% cut off ratio) was higher than that of beam D1at any level of load. Beam D2 achieved deflection at ultimate load; u, of 114% of that of beam D1, and for beam D3 this ratio was 110%. The area under the load-deflection curve was calculated to obtain the strain energy achieved by tested beams and was 0.886, 1.003, and 1.062 kN.m for beams D1, D2, and D3 respectively. 4.2

Effect of lap splice length (Group 2)

Figure 8 shows the relationship between load and mid-span deflection for the tested beams in this group. 100

80

Load (kN)

60

40

D1

D2

D3

20

0 0

Figure 7.

84

5

10 Deflection (mm)

15

Load deflection curve for Group 1.

20

100

100

80

80

60 Load (kN)

Load (kN)

60

GROUP 2

40 D1

D4

D3

GROUP 3 40

D1 D6 D9

D10

D4 D7

D5 D8

20 20

0 0

0 0

Figure 8.

10

Deflection (mm)

20

10

20

30

40

50 Deflection (mm)

30

Figure 9.

Load deflection curve for Group 2.

Load deflection curve for Group 3.

stirrups (beam D6) or vertical stirrups with additional leg (beam D5) in the splice zone, resulted in small values of mid-span deflections. Beams D7, D8, D9 with special transverse reinforcement in splice zone showed deformation higher than that of the reference beam D1 as well as that of the beams with separate stirrups. Beam D4 suddenly failed at 52.5 kN due to loss of bond strength. The calculated strain energy for the tested beams of D1, D4, D5, D6, D7, D8, and D9 was 0.886, 0.218, 1.017, 2.393, 2.506, 3.338, and 3.154 kN · m respectively. Spliced beams with transverse reinforcement D5, D6, D7, D8, and D9 achieved 115%, 270%, 283%, 377%, and 356% respectively of that of unspliced beam D1. From the previous results, it is clear that the use of transverse reinforcement increases the energy absorbed by the beams up to failure. The ratio between the maximum deflection at ultimate load and the deflection at yield load; u /y was 1.44, 1.41, 3.61, 1.38, 4.09, and 3.88 for beams D1, D5, D6, D7, D8, and D9 respectively. It is clear that beam D8 had the highest ratio, which indicates that beam D8 was the most ductile beam compared with other tested beams in this group.

Beams D4, D3, and D10 achieved an ultimate load of 66%, 105%, and 109% respectively of that of the reference beam D1. It is clear that beam D4, with a lap splice length of 27 db, did not reach the expected ultimate load. Figure 8 shows that after cracking, beams D1 and D4 showed almost the same behavior up to the sudden failure of beam D4, while beams D3 and D10 showed lower deformation than that of beam D1 at the same loads. After yielding, excessive deformations took place. Beam D1, D4, D3, and D10 achieved maximum deflection at ultimate load of 16.4, 7.2, 18, and 27 mm respectively. These results indicate that the use of lap splice length equals to that recommended by the Egyptian code (54 db), or greater (75 db) increased the maximum deflection at ultimate load. The calculated strain energy for beams D1, D4, D3, and D10 was 0.886, 0.218, 1.062 and 1.796 kN · m respectively. The ratio between the maximum deflection at ultimate load and the deflection at yield load; u/y, was 1.44, 1.29, and 2.90 for beams D1, D3, and D10 respectively. It is clear that beam D10 was the most ductile beam. 4.3 Effect of transverse reinforcement (Group 3) All beams in this group had transverse reinforcement except beam D4. Figure 9 shows the relationship between ultimate load and mid-span deflection for this group. The ultimate load of beams D4, D5, D6, D7, D8, and D9 was 52.5, 82.5, 87.5, 85, 88, and 87.5 kN respectively. It is clear that spliced beams with transverse reinforcement showed higher ultimate loads comparing with both un-spliced (beam D1), and spliced beam without confinement (beam D4). Beams D5, D6, D7, D8, and D9 achieved 103.1%, 109.4%, 106.3%, 110%, and 109.4% of the ultimate load of the un-spliced beam D1 respectively. This increase in the ultimate load values is mainly due to the use of transverse reinforcement, which eliminated the formation of splitting cracks at the tension splice zone and minimized the width of crack. The use of vertical

4.4

Group 4: Effect of spacing of stirrups (Group 4)

Figure 10 shows the relation between load and midspan deflection for beams D1, D4, D11, D12, and D6. The ultimate load of there beams was 80.0, 52.5, 80.0, 83.0, and 87.5 kN, respectively. The use of stirrups controlled the splitting cracks width at the tension splice zone in beams D11, and D12, and prevented the splitting cracks at the tension splice zone in beam D6. Beam D6 with 60 mm spacing showed the lowest deformation, before cracking. After cracking and before yield, figure 10 emphasizes that the decrease of stirrups spacing decreased the deflection at the working loads. After yielding, the presence of stirrups resulted in an increase in the mid-span deformation at

85

1. The behavior of a beam without any splice can be achieved in a spliced beam with 100% cut off ratio when: Lo = 27 db using transverse reinforcement with spacing ≤ 120 mm, or if the lap length ≥ 54 db without transverse reinforcement. 2. The use of a lap splice with 100% cut off ratio, with length of 27 db and without transverse reinforcement resulted in a brittle bond failure. 3. Using transverse reinforcement with lap splice length = 27 db, and 100% cut off ratio showed a better behavior than that of the un-spliced beam. Higher ultimate loads and an increase in the ductility were achieved comparing with the un-spliced beam. 4. Although the spacing of stirrups in the splice zone was twice, three times, and four times that recommended by several building codes, an increase in the ultimate strength and ductility were observed. 5. All beams with transverse reinforcement showed large values of deflection at ultimate load. These values ranged from two to three times that of the un-spliced beam.

100

Load

( kN )

80

60

GROUP 4 40

D1

D4

D6

D11

D12

D9

20

0 0

Figure 10.

10

20

30

40

50 Deflection (mm)

Load deflection curve for Group 4.

the ultimate load. Beams D11, D12, and D6 showed maximum mid-span deflections of 20 mm, 21 mm, and 34 mm, respectively, while beam D1 reached 16.4 mm ultimate deflection. The use of 120 mm, 90 mm, and 60 mm stirrups spacing resulted in an increase in the maximum deflection by 21%, 28%, and 107% of that of the un-spliced beam D1 respectively. The strain energy achieved by tested beams D1, D4, D11, D12, and D6 was 0.886, 0.218, 1.02, 2.39, 2.51 kN · m, respectively. One can conclude that the use of stirrups enhances the energy observed by the beams up to failure, because the use of stirrups controls the formation of splitting crack in the splice zone. The ratio of the maximum deflection at ultimate load to the deflection at yield load; u/y, was 1.44, 1.67, 1.78, and 3.61 for beams D1, D11, D12, and D6 respectively. As mentioned before, beam D4 did not reach the yield strength. It is clear that beam D6 with the lowest stirrups spacing had the highest ratio, and consequently the highest ductility. This means that the decrease of stirrups spacing, increases the ability of the beam to absorb energy, and to sustain excessive deformations after yielding. 5

REFERENCES ACI 318-05, 2005. Building Code Requirements for Structural Concrete and Commentary. American Concrete Institute, Michigan, 2005. Diab A.M. 2008. Lap Splices In Reinforced Concrete Beams Subjected To Bending, Master thesis. University of Alexandria, Egypt, December 2008. ECP 203-2007. Egyptian Building Code for Structural Concrete Design and Construction. Ministry of Housing, 2007. Eurocode 2 1992-1. Design Of Concrete Structures-Part 1: General Rules And Rules For Buildings, European Standard, European Committee for Standardization, October 2001. Ferguson, P.M. and Breen, J.E. 1965. Lapped Spliced For High Strength Reinforcing Bars. ACI Structural Journal, Proceedings Vol. 62, No. 9, Detroit, Michigan, September 1965. Hamad, B.S. and Fakhran, M.F. 2008. Effect of Confinement on Bond Strength of Hot-Dip Galvanized Lap Splices in High-Strength Concrete. ACI Structural Journal, Proceedings Vol. 103, No. 1, Detroit, Michigan, January 2006. Jeanty, P.R. et al. 1988. Investigation of ‘‘Top Bar’’ Effects in Beams. ACI Structural Journal, Proceedings Vol. 85, No. 3, Detroit, Michigan, February 1988.

SUMMARY AND CONCLUSIONS

Twelve concrete beams were tested to study the effect of lap splice of tension reinforcement with different splice lengths, cut off ratio, shape of the stirrups and its spacing on the behavior of these beams. From the results of the studied beams, the following conclusions were obtained:

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Load testing a historic monument F.D. Heidbrink Wiss, Janney, Elstner Associates, Inc., Northbrook, Illinois, USA

ABSTRACT: On June 22, 2006, a 230 kg (500 lb) granite fragment fell from the observation deck at the Perry’s Victory and International Peace Memorial. The circa 1915 memorial is located on South Bass Island, Ohio. As part of an emergency inspection, the underside of the reinforced concrete Upper Plaza level was surveyed. Extensive and severe freeze thaw deterioration was noted throughout. A load test was conducted to establish whether or not a severely deteriorated portion of the Upper Plaza can safely and reliably support the code prescribed loads with an appropriate factor of safety. At the conclusion of the test, one of the beams did not meet the acceptance criteria. Therefore, after long term repairs to the memorial are completed and public access to the plaza restored, it was recommended that the live load rating be reduced, that the Upper Plaza be shored or reinforced, or that the Upper Plaza be demolished and rebuilt. 1 1.1

INTRODUCTION The battle

On September 10, 1813, Commodore Oliver Hazard Perry defeated and captured six vessels of Great Britain’s Royal Navy at the Battle of Lake Erie. At the conclusion of the battle, Perry penned the now famous words to General William Henry Harrison: ‘‘Dear General: We have met the enemy and they are ours.’’ His success enabled Harrison to transport his army to Canada, defeating the British at the Battle of the Thames River. Many consider this victory to be the turning point of the War of 1812.

1.2

The memorial

As the centennial of the war approached in the early 1900s, it was decided that a memorial would be built to honor Commodore Perry and the men who fought with him, as well as the century of peace between the United States and Canada that followed the war. A design competition was held, and the architects chosen were Joseph Freedlander and Alexander Seymour, Jr. The Perry’s Victory and International Peace Memorial is located on an isthmus on South Bass Island in Lake Erie, adjacent to the city of Put-in-Bay, Ohio. The memorial was designed to appear as a Roman Doric column, considered the tallest in the world of its type (Fig. 1). The memorial stands 107 m (352 ft) above Lake Erie, with an observation deck at a height of 97 m (317 ft). The diameter at its base is 14 m (45 ft). The exterior of the memorial is pink granite from Milford, Massachusetts with a cast in place unreinforced concrete core lined with hard-fired face brick on the interior

Figure 1. Memorial.

Perry’s

Victory

and

International

Peace

(Fig. 2). The pink granite was chosen because from a distance it appears brighter than white. Coursing consists of seventy-eight rows of granite, with thirty blocks per course. An elevator runs through the interior of the memorial column, bringing visitors up to the observation deck level. Surrounding the memorial column is a raised Upper Plaza paved with granite and brick pavers. The Upper Plaza is raised above the natural grade level on a

87

the reinforced concrete structure. Three cores were removed for examination for freeze-thaw deterioration. One of these cores was tested for compressive strength. Notable conditions observed included concrete deterioration in the form of cracked, spalled, and delaminated concrete. Some areas of concrete sounded ‘‘dead’’ or hollow when tapped with a hammer, indicating delaminations, freeze-thaw deterioration, or other defects. In areas of spalled concrete, corroded reinforcing steel was observed. Areas of previously installed patches were noted. Sound testing of the previous repairs indicated that some areas may be delaminated and/or debonded. Evidence of water leakage through the deck structure in the form of water staining and efflorescence was observed on the underside of the structural slab (Fig. 3).

Figure 2. Memorial under construction (photo courtesy of National Park Service).

3.2

concrete structure, consisting of cast in place reinforced concrete columns supporting reinforced concrete beams and integral slab. Construction of the memorial began in 1912 and was completed in 1915. The memorial was declared a national monument on July 4, 1936, by President Franklin D. Roosevelt and is currently administered by the National Park Service. The NPS maintains an extensive archive at the site, including drawings, specifications, and photographs documenting the original construction and subsequent repair campaigns. Prior to the involvement of the author’s firm, the most recent extensive repairs had been completed in the early 1980s. The 1980s work included installation of a waterproofing membrane to protect the Upper Plaza structure from water infiltration. Prior to this work, the concrete structure was unprotected. 2

Structural analysis

The results of a structural analysis indicated that the Upper Plaza structure as originally designed would have adequate capacity to carry the design live load of 4.8 kPa (100 psf). There is no known method of structural analysis to reliably predict the behavior of the deteriorated concrete or determine whether any of the distress observed is a result of current structural inadequacy. However, observations of visible portions of the plaza structure as well as analysis of the concrete cores indicate that freeze-thaw deterioration of the concrete is extensive and severe. The distress is highly variable and the extent is difficult to quantify. Because the Upper Plaza could be easily shored at the basement level, it was possible to perform load testing to evaluate the behavior of the structure by monitoring deflections, and to determine whether it has sufficient strength and reliability in its current, partially deteriorated condition.

THE INCIDENT

On June 22, 2006, a 230 kg (500 lb) granite fragment fell from the observation deck at the top of the memorial and crashed into the Upper Plaza. An emergency investigation was conducted on the memorial in order to stabilize the area of the failure and inspect the remaining exposed areas of the observation deck. As part of a comprehensive follow-up investigation of the current condition of the memorial, the underside of the reinforced concrete Upper Plaza was surveyed. Extensive and severe freeze-thaw deterioration was noted throughout. 3 3.1

INVESTIGATION OF THE UPPER PLAZA Field survey

A visual survey and sounding were performed at the Upper Plaza to evaluate the current condition of

Figure 3. Extensive and severe freeze-thaw damage to underside of Upper Plaza.

88

4 4.1

LOAD TEST OF THE UPPER PLAZA Load medium

Once the decision to conduct the load test of the Upper Plaza was made, various methods for application of the test load were reviewed. Using dead weights or large test frames was ruled out since transportation to the island is limited to ferry boat or small airplane. With water plentiful from nearby Lake Erie, the use of a water tank or swimming pool was determined to be the best alternative. 4.2

Shoring

Prior to conducting the load test, steel frame shoring was installed under the structure with sufficient capacity to safely support the test loads and weight of the plaza in the event of a structural failure. The shoring was positioned approximately 50 mm (2 in) below the structure to allow for anticipated deflections, while also providing support should excessive deflections occur. 4.3

Figure 4.

Plan view of deflection instrument locations.

Crack mapping

The structure beneath the test area was inspected and sounded with a hammer for spalls, delaminations, cracks or other signs of distress. The distress conditions observed were recorded on an inspection data sheet. Significant existing cracks were marked and initial crack widths recorded directly on the structure. Changes in crack length and width were monitored and recorded during application of the test load. 4.4

Instrumentation Figure 5. View of load test in progress looking down from the observation deck (note repairs to observation deck facade).

Deflections of the beams and slab areas were monitored using cable extension transducers (CET) installed at twenty locations. Figure 4 is a plan view showing the locations of each installed instrument. Each CET has a measurement range of at least 50 mm (2 in) with a resolution of 0.025 mm (0.001 in). The CETs were wired into a panel board, which was interfaced to a data acquisition system. 4.5

based on measurements of the as-built dimensions of the beams, girders and slab in the test area. The required superimposed test load was determined to be 8.5 kPa (178.4 psf).

Test criteria 4.6

The load test was performed in accordance with the strength evaluation procedures prescribed in Chapter 20 of the American Concrete Institute Building Code Requirements for Structural Concrete (ACI 318-05). The design live load for this structure is 4.8 kPa (100 psf). As specified in Section 20.3.2 of ACI 318, the total test load (including dead load already in place) shall not be less than 0.85 (1.4D + 1.7L), where D is the dead load or self-weight and L is the design live load. The dead load was calculated

Load application

Three 5 m (16 ft) by 10 m (32 ft) by 1 m (40 in) deep lightweight swimming pools were installed adjacent to one another in the designated test area (Fig. 5). The 8.5 kPa (178.4 psf) test load was achieved with a water depth of 876 mm (34.5 in) of water over the entire test area. A large pump was used to bring lake water into the center pool. Two smaller pumps were used to pump the water out of the center pool and into each adjacent pool.

89

4.7

where max is the measured maximum deflection under the full test load, r max is the residual deflection, t is the span length, and h is the overall thickness of the member. The residual deflection r max is the difference between the initial deflection prior to loading and the final deflection after load removal. If the structure shows no evidence of failure, recovery of deflection after removal of the test load (Equation 2) is used to determine whether the strength of the structure is satisfactory.

Test procedure

At the onset of the test, an initial loading of ten percent of total test load was applied in order to assure that the structure was behaving as predicted and the loading mechanism and instrumentation were functioning as designed. Once it was determined that the structure was behaving as expected, the remainder of the test load was applied in four equal increments. It took approximately five hours to achieve the maximum load. At each load stage, the flow of water was stopped while readings were recorded on the deflection instruments and a crack survey of the test area was conducted. Once the total test load was achieved, it was kept in place for a period of 24 hours. During this hold period, deflection readings were taken periodically. After the test load had been in place for 24 hours, the water was drained from the pools in the reverse order of load stages. Deflection measurements were obtained after each load decrement and 24 hours after the test load was completely removed.

5 5.1

5.2

All of the elements in the area tested sustained the required test load without evidence of failure. Net deflections were determined by subtracting the average of the deflection at each support from the deflection reading obtained at midspan. The maximum net deflections measured at each location are summarized in Table 1. 5.2.1 Beams Of the six reinforced concrete beams tested, load testing showed that one of the beams did not meet the ACI criteria for both maximum deflection and residual deflection. The center of the beam at CET 13 had a maximum deflection of 5.4 mm (0.212 in), which exceeded the calculated deflection limit of 3.7 mm (0.147 in). In addition, CET 13 had a residual deflection of 1.6 mm (0.063 in), which exceeded 25 percent of the maximum deflection, or 1.3 mm (0.053 in). The beam at CET 13 was not observed to have distress conditions present based on the visual survey.

RESULTS Acceptance criteria

The ACI acceptance criteria require that the structural members tested show no evidence of failure, such as structurally significant cracking or spalling. ACI 318 also requires that the measured maximum deflections satisfy one of the following conditions: max ≤

2t 20, 000 h

(1)

5.2.2 Girders Of the two reinforced concrete girders tested, CET 4 had a maximum deflection of 2.8 mm (0.110 in), which exceeded the calculated deflection limit of 1.7 mm (0.068 in). However, CET 4 had a residual deflection of 0.7 mm (0.027 in), which did not exceed

or r max ≤ Table 1.

max 4

Test results

(2)

Summary of load test results.

CET Existing location Element condition

Calculated maximum deflection (mm)

Recorded maximum deflection (mm)

3 4 5 7 8 10 13 16 17 18

3.7 1.7 3.7 1.9 1.9 3.7 3.7 3.7 1.7 3.7

2.8 2.8 3.4 0.4 0.2 3.6 5.4 1.7 1.5 2.2

Beam Girder Beam Slab Slab Beam Beam Beam Girder Beam

Previous patch Hollow sounding Previous patch Hollow sounding Hollow sounding Previous patch No observed distress No observed distress Hollow sounding No observed distress

90

Recorded deflection exceeds calculated

25% of maximum deflection (mm)

Permanent deflection recorded (mm)

Permanent deflection recorded exceeds limit

Yes

0.7

0.7

No

Yes

1.3

1.6

Yes

25 percent of the maximum deflection and therefore met the acceptance criteria. The girder at CET 4 was observed to have been ‘‘dead’’ or hollow sounding based on the visual survey.

survey and sounding of the concrete in some areas. Visual observations, confirmed by the limited laboratory study, showed that extensive freeze-thaw deterioration has occurred. Similar conditions were observed in other areas of the Upper Plaza that were not a part of this load test. The load test results indicate that, in its current condition, the Upper Plaza structure does not meet ACI criteria to safely support the design live load. Therefore, as part of long term repairs to the memorial to restore public access to the Upper Plaza, it was recommended that the live load rating be reduced, that the Upper Plaza be shored or reinforced, or that the Upper Plaza be demolished and rebuilt.

5.2.3 Slabs Of the two slab areas tested, both met the ACI acceptance criteria. Both were considered ‘‘dead’’ or hollow sounding when surveyed, likely indicating early age freezing and severe freeze-thaw deterioration. Results of the petrographic examination of concrete cores taken from slab areas adjacent to the load test area correlate well with the results of the visual survey and sounding.

ACKNOWLEDGEMENT 6

CONCLUSION The author wishes to thank the National Park Service for their assistance with this test program.

Based on the overall findings, the load test results did not correlate well with the findings of the visual

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Modeling of concrete beams prestressed with AFRP tendons Y.J. Kim Department of Civil Engineering, North Dakota State University, Fargo, ND, USA

ABSTRACT: This → paper presents modeling of concrete beams prestressed with aramid fiber reinforced polymer (AFRP) tendons, based on the iterative sectional analysis and the finite element analysis (FEA) models. Two experimental beams were selected with different beam dimensions and different reinforcement ratios to validate the developed modeling approaches. Detailed comparisons between the experimental beams and the predictive models are made in terms of the flexural responses. The developed models adequately predict the experimental data with a maximum error of less than 5% at ultimate. 1

2

INTRODUCTION

Corrosion damage of steel tendons in a prestressed concrete member significantly influences the performance of the member. Various environmental factors contribute to the corrosion of steel tendons, for example, chloride, freezing-and-thawing, and temperature (Meta & Gerwick 1982; Sherman et al. 1996). Fiber reinforced polymer (FRP) tendons may provide superior durability to conventional steel tendons. FRP tendons consist of unidirectional fibers embedded in a matrix resin. The resins are usually thermosetting and epoxy is widely used. The non-metallic reinforcement provides a number of benefits, namely, favorable strength-to-weight ratio, good resistance to chemicals, low relaxation, and outstanding fatigue resistance (Grace & Abdel-Sayed 2000). Typically, two types of fibers are used for prestressed concrete applications, such as aramid fibers and carbon fibers. Carbon FRP (CFRP) provides higher strength and modulus in comparison to aramid FRP (AFRP), whereas the material cost of CFRP is more expensive than that of AFRP. Therefore, in order to use the structural advantages of FRPs with reasonable material costs, AFRP tendons may be recommended for prestressed concrete members. Although numerous research has been reported on prestressed concrete with FRP tendons, majority of the publications focused on experimental investigations (McKay & Erki 1993; Taerwe & Matthys 1995; Saafi & Toutanji 1998; Lees & Burgoyne 1999; Grace & Abdel-Sayed 2000). Limited efforts have been made to theoretically predict the behavior of FRP-prestressed members (Pisani 1998; Zou 2003b; Youakim & Karbhari 2007). This paper presents two theoretical modeling approaches to predict the flexural behavior of concrete beams prestressed with AFRP tendons, including the iterative sectional analysis and 3-dimensional finite element analysis models.

PRESTRESSED CONCRETE MEMBERS WITH FRP

As briefly mentioned in the previous section, FRPprestressed concrete members include AFRP or CFRP tendons. Glass FRP (GFRP) materials may not be recommended for prestressing applications because glass fibers are susceptible to alkali reaction in concrete which may accelerate creep-rupture of the prestressed GFRP tendons (Dolan 1999). McKay and Erki (1993) examined the load-carrying capacity of concrete beams prestressed with AFRP tendons, including time-dependent relaxation behavior of the prestressed AFRP tendons. Prior to cracking of the beams, there was no difference of the flexural stiffness between the beams prestressed with AFRP tendons and steel tendons. A significant drop of the stiffness was observed in the case of the beam with steel tendons when the applied load exceeded the yield load of the beam, whereas no such a change was found in the AFRPbeam, given that AFRP tendons did not demonstrate yield characteristics. The relaxation of the prestressed AFRP tendons was less than 8% in 10,000 hours after jacking. Niitani et al. (1997) reported the durable performance of AFRP tendons that did not show any changes in the mechanical properties after 6 years inside concrete beams. Saafi and Toutanji (1998) tested prestressed concrete beams with bonded or unbonded AFRP tendons. The load-carrying capacity of the bonded beams was greater than that of the unbonded beams. The beam prestressed with a combination of bonded and unbonded AFRP tendons showed higher ductility than other beams. Lees and Burgoyne (1999) tested concrete beams prestressed with AFRP tendons, including various bond parameters such as fully-bonded, unbonded, and partiallybonded tendons. The prestressed concrete beams with fully-bonded AFRP tendons showed the highest loadcarrying capacity; however, limited ductility was

93

Table 1.

Beam details and material properties. AFRP

Concrete

Prestress level (%)

Area (mm2 )

Ultimate strain (%)

Modulus (GPa)

Strength (MPa)

Modulus (GPa)

Experiment FEA Iterative

60

58

3.7

68

40.3

25.4

B1FEA40 B1Iter40

FEA Iterative

40

B1FEA20 B1Iter20

FEA Iterative

20

B1FEA0 B1Iter0 AR21 AR21FEA55 AR21Iter55 AR21FEA40 AR21Iter40

FEA Iterative Experiment FEA Iterative FEA Iterative

51

2.4

60

56.4

31.5

AR21FEA20 AR21Iter20

FEA Iterative

20

AR21FEA0 AR21Iter0

FEA Iterative

0

Beam

Description

B1 B1FEA60 B1Iter60

0 55

40

found. The beams with unbonded AFRP tendons exhibited improved ductility with a low load-carrying capacity (25%) in comparison to the fully bonded case. The beams with partially bonded AFRP tendons showed a high load-carrying capacity with high ductility. Grace and Abdel-Sayed (2000) tested double-tee concrete beams prestressed with a combination of bonded and unbonded CFRP tendons under various loading configurations, including vibration, impact, fatigue, and static loads. They found that there was no notable damage in the CFRP tendons when the beams failed in flexure and fatigue performance of the beams was satisfactory (i.e., survived after 7 million cycles). Zou (2003a) compared the performance of concrete beams prestressed with AFRP or CFRP tendons to that with conventional steel tendons. The AFRPprestressed beam showed reduced crack spacing in comparison to the steel-prestressed beams. The beam with CFRP tendons exhibited lower residual deflections compared to the beams with steel tendons that had yielded. The energy stored in the beams with prestressed FRP tendons was mainly elastic. ACI 440.4R-04 (ACI 2004) provides a design guide of the use of FRP tendons for prestressed concrete. The document includes anchorage issues, flexural and shear design, bond and serviceability, and design examples.

1992, Sen et al.1999), as shown in Table 1. Details of the beams are shown in figure 1. The AFRPreinforcement ratios were 0.15% and 0.36% for the B1 beam (McKay 1992) and AR21 beam (Sen et al. 1999), respectively. The identification code in Table 1 includes the base-beam (B1 or AR21), type of predictive model (FEA or Iter), and the level of prestress (e.g., 60 and 55 are the percentage of the ultimate strength). Typical material properties used for this study are also shown in Table 1. Both of the beams were simply supported and tested under four point bending. The load was monotonically applied until failure of the beams occurred. To study the effect of various prestressing levels in the AFRP tendons, additional predictive models were given (Table 1).

3

– Plane sections remain plane before and after bending. – Slippage between materials does not occur. – Premature anchorage failure does not happen.

4 4.1

PREDICTIVE MODELS Nonlinear iterative model

To predict the flexural behavior of the concrete beams prestressed with AFRP tendons, a nonlinear sectional analysis model was developed. The model was based on force equilibrium and strain compatibility. The following was assumed:

BEAM DETAILS

Two experimental concrete beams prestressed with AFRP tendons were selected for this study (McKay

94

P/2

P/2

150

2

c

Steel (100 mm each)

300

800

2000

AFRP

(a)

AFRP (28.8 mm2 each)

T AFRP

(d)

P/2 114

150

990

(c)

Figure 2. Iterative sectional analysis model: (a) typical cross section at midspan; (b) strain diagram; (c) stresses and forces before cracking; (d) stresses and forces after cracking.

(a) P/2

T AFRP

(b)

2250

AFRP (25.5 mm2 each)

AFRP

(b) Figure 1. AR21.

(b)

(c)

A typical sectional analysis model is shown in figure 2. The concrete was modeled using an equivalent stress block using Eq. 1 (Collins and Mitchell 1987), rather than a full integration of nonlinear stress profiles of the concrete.

α=

Concrete

Beam details (unit: mm): (a) Beam B1; (b) Beam

– The beam fails when the concrete strain reaches the crushing strain or the AFRP strain reaches the rupture strain (depending on the product, given in Table 1). – Tension stiffening after cracking does not influence the flexural behavior of the beam.



(a)

 εc 1 εc − εc0 3 εc0





2

/β

and β =

4 − εc /εc0 6 − 2εc /εc0

Figure 3. FEA model of B1FEA60 (unit: m): (a) mesh formulation (cut-away view to show the reinforcement); (b) initial camber; (c) deflection at ultimate.

3. Assume the concrete strain (εc ) and then establish the equivalent rectangular stress block using Equation 1. 4. Solve for the neutral axis depth (c) using force equilibrium (T = C) through a trial and error process. 5. Check whether the AFRP strain (εAFRP ) is less than its rupture strain (εfu ). If εAFRP > εfu , stop the iteration, otherwise go to the next step. Also check the concrete strain (εc ) as in the case of the AFRP strain. 6. Calculate the curvature of the section. 7. Calculate the internal moment. 8. Go to Step 3 and increase the concrete strain (εc ). Conduct the iteration until the AFRP tendon ruptures or the concrete crushes.



(1) where α and β are the stress block factors, εc is the concrete strain at an arbitrary load level, and εc0 is the concrete strain at the specified concrete strength of fc . Fully-cracked sectional properties were used when the tension stress at the bottom of the concrete  beam reached the modulus of rupture ( fr = 0.6 fc ), as shown in figure 2(d). The modeling procedure is briefly summarized as follows: 1. Establish an expression for the tension force (T ), including the AFRP tendons with an initial prestressing force, which is a function of the AFRP strain (εAFRP ). The tension force should be in equilibrium with the compression force (C) in concrete that is a function of the neutral axis depth (c). 2. Using similar triangles, determine the strains in the AFRP (εAFRP ) as a function of the concrete strain (εc ) and the neutral axis depth (c), then substitute them in Step 1.

4.2 Finite element analysis model The general-purpose finite element analysis (FEA) software package ANSYS was used to predict the flexural behavior of the concrete beams prestressed with AFRP tendons. A typical FEA model is shown in figure 3. The concrete was modeled using threedimensional composite elements having 8 nodes with 3 translational degrees of freedom per node. The concrete element was able to simulate crushing of

95

concrete in compression and cracking in tension. The behavior of the concrete was modeled based on the William and Warnke model (ANSYS 2008). A smeared crack analogy was adopted for the concrete element, rather than discrete cracking. The unidirectional AFRP tendon was modeled using threedimensional spar elements including 2 nodes with the same degrees of freedom as in the concrete element. The AFRP tendon included a linear elastic response until complete rupture occurred, with the material properties shown in Table 1. The prestressing effect was given by applying initial strains in the tendon element. A relatively fine mesh (i.e., maximum dimension length of 40 mm) was formulated. Nonlinear iterative solutions were conducted using the modified Newton-Raphson method with a constant stiffness predictor. Flexural loads were incrementally applied until the concrete failed or the AFRP ruptured. 5

mental beams and the FEA models. The flexural response of the beams with prestressed AFRP tendons was essentially bilinear with a transition at the cracking load. This is due to the linear characteristics of AFRP tendons that do not show any yield plateau. Beam B1 (experimental) was unloaded at about 80 kN (B-1-1) and reloaded until complete failure occurred (B-1-2), whereas the FEA model (B1FEA60) included only a single loading step (Fig. 4a). The fist loading cycle in the experiment (B-1-1) provided the cracking load and the second cycle (B-1-2) exhibited the ultimate load. Good prediction of the FEA model was made, including an error of 24.1% and 2.2% for the cracking load and the ultimate load, respectively. The relatively large error in the cracking load may be due to the fact that the modulus of rupture was input in the FEA model using the code provision that was mentioned in the previous section, whereas the actual cracking stress could be slightly different from the code equation. The prediction of Beam AR21 was also satisfactory (Fig. 4b), including an error of 4.6% for the ultimate load. It should be noted that the FEA predictions in

VALIDATION OF THE MODELS

Figure 4 shows a typical comparison of the loaddeflection response at midspan between the experi-

(a)

(a)

(b)

(b)

Figure 4. Comparison of the experimental and the FEA prediction: (a) B1 vs. B1FEA60; (b) AR21 vs. AR21FEA55.

Figure 5. Comparison of the FEA and the iterative sectional analysis model: (a) B1FEA60; (b) AR21FEA55.

96

6

0%

CONCLUDING REMARKS

This paper has presented two theoretical modeling approaches to predict the flexural behavior of concrete beams prestressed with AFRP tendons, including the nonlinear iterative sectional analysis and the FEA models. Good agreement between the experimental and theoretical beams was made including a maximum error of less than 5% at ultimate. According to the parametric study on various levels of initial prestressing forces in the AFRP tendons, there was no notable effect on the failure of the tendons, whereas the cracking load of the prestressed concrete beams was significantly influenced by the prestressing level. The on-going research includes an investigation of ductility issues for lightly reinforced prestressed concrete members with AFRP tendons.

20% 60% 40%

(a)

REFERENCES

20%

ACI 440.4R-04. 2004. Prestressing concrete structures with FRP tendons. ACI Committee 440, American Concrete Institute. ANSYS. 2008. Online manual, ANSYS Inc. Collins, M.P. & Mitchell, D. 1987. Prestressed concrete basics. Canadian Prestressed Concrete Institute. Dolan, C.W. 1999. FRP prestressing in the USA, Concrete International, 21(10): 21–24. Grace, N.F. & Abdel-Sayed, G. 2000. Behavior of carbon fiber-reinforced prestressed concrete skew bridges. Structural Journal, ACI, 97(1): 26–35. Lees, J.M. & Burgoyne, C.J. 1999. Experimental study of influence of bond on flexural behavior of concrete beams pretensioned with aramid fiber reinforced plastics. Structural Journal, ACI, 96(3): 377–385. McKay, K.S. 1992. Aramid fibre reinforced plastic tendons in pretensioned concrete applications, M.Eng. Thesis, Royal Military College of Canada, Kingston, ON, Canada McKay, K.S. & Erki, M.A. 1993. Flexural behaviour of concrete beams pretensioned with aramid fibre reinforced plastic tendons. Canadian Journal of Civil Engineering, 20: 688–695. Mehta, P.K. & Gerwick, B.C. 1982. Cracking-corrosion interaction in concrete exposed to marine environment. Concrete International, ACI, 4(10): 45–51. Niitani, K., Tezuka, M. & Tamura, T. 1997. Flexural behaviour of prestressed concrete beams using AFRP pre-tensioning tendons. Non-metallic (FRP) Reinforcement for Concrete Structures (FRPRCS-3): 663–670. Pisani, M.A. 1998. A numerical survey on the behaviour of beams pre-stressed with FRP cables. Construction and Building Materials, 12: 221–232. Saafi, M. & Toutanji, H. 1998. Flexural capacity of prestressed concrete beams reinforced with aramid fiber reinforced polymer (AFRP) rectangular tendons. Construction and Building Materials, 12: 245–249. Sen R., Shahawy, M., Rosas, J. & Sukumar, S. 1999. Durability of aramid fiber reinforced plastic pretensioned elements under tidal/thermal cycles. Structural Journal, ACI, 96(1): 95–104.

0% 60% 40%

(b) Figure 6. Variation of FRP strains depending on prestressing level: (a) B1-parametric; (b) AR21-parametric.

figure 4 were shifted to zero deflection when the beams were initially loaded in order to provide an appropriate comparison to the experimental data that had been reported without upward camber deflection. Figure 5 shows the moment-curvature relationship between the FEA and the iterative sectional analysis model. The curvature values of the FEA prediction were obtained based on the compression strains in concrete and the tension strains in the AFRP tendons. An abrupt jump of the curvatures immediately after the cracking of the beams was due to a sudden decrease in the moment of inertia, which is particularly important for lightly reinforced concrete members (ρ = 0.15% and 0.36% for Beams B1 and AR21, respectively). Figure 6 shows the strain variation in the AFRP tendons depending on the initial prestressing levels. The failure of the AFRP tendons was independent of the level of the prestress, whereas the cracking load of the prestressed beams was significantly influenced by the prestress level.

97

Sherman, M.R., McDonald, D.B. & Pfeifer, D.W. 1996. Durability aspects of precast prestressed concrete Part 2: chloride permeability study. PCI Journal, 41(4): 76–95. Taerwe, L. & Matthys, S. 1995. Structural behaviour of concrete slabs prestressed with AFRP bars. Non-metallic (FRP) Reinforcement for Concrete Structures (FRPRCS-2): 421–429. Youakim, S.A. & Karbhari, V.M. 2007. An approach to determine long-term behavior of concrete members prestressed with FRP tendons. Construction and Building Materials, 21: 1052–1060.

Zou, P.X.W. 2003a. Flexural behavior and deformability of fiber reinforced polymer prestressed concrete beams. Journal of Composites for Construction, ASCE, 7(4): 275–284. Zou, P.X.W. 2003b. Theoretical study on short-term and longterm deflections of fiber reinforced polymer prestressed concrete beams. Journal of Composites for Construction, ASCE, 7(4): 285–291.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Nonlinear finite element analysis of unbonded post-tensioned concrete beams U. Kim, P.R. Chakrabarti & J.H. Choi Department of Civil and Environmental Engineering, California State University, Fullerton, CA, USA

ABSTRACT: The main purposes of this study are to develop a sophisticated 3-D finite element model for simulating the nonlinear flexural behavior of unbonded post-tensioned beams, to compare analysis results with experimental results to verify the accuracy of the developed 3-D finite element model, and to investigate the effects of various prestressing forces on the flexural behavior of post-tensioned beams. To investigate the nonlinear flexural behavior of post-tensioned concrete beams, a 3-D finite element model was developed using ANSYS. ANSYS is a highly recognized and reliable commercial software that is used for finite element analysis. In order to validate the developed finite element model, four post-tensioned beams were tested at the structures laboratory of California State University, Fullerton and the test results were compared with the analysis results using ANSYS. 1

INTRODUCTION

of the beams. This study investigates the inelastic behavior of unbonded post-tensioned beams using the finite element method and experimental tests. The obtained comparison and analysis will be discussed at the end of this paper.

The inelastic flexural behavior of unbounded posttensioned concrete beams is inherently complicated, and reliable nonlinear behavior can usually be obtained through physical tests on actual beams (Chakrabarti 1995; Harajli 1991). However, tests are time consuming, expensive, and test results are generally limited to surface measurements. Thus, this study was conducted to compare and analyze the results between a finite element method and experimental tests to develop a reliable 3-D finite element model to simulate the flexural behavior of unbounded post-tensioned beams. The experimental tests were performed at the structures laboratory of CSUF (California State University, Fullerton). First, four post-tensioned test beams were constructed. Table 1 shows that two of the four experimental beams had an applied prestressing force of 31.1 kN (7000 lb). The remaining two beams had an applied prestressing force of 15.6 kN (3500 lb). Table 1 also illustrates that beams 41 and 42 had two #3 rebars in the upper top portion and two #3 rebars in the lower bottom portion of the beams, while beams 43 and 44 had two #4 rebars in the lower bottom portion

Table 1.

2

FLEXURAL TEST FOR POST-TENSIONED CONCRETE BEAM

Figure 1 shows the dimensions of the experimental post-tensioned concrete beams. The post-tensioned concrete beams were constructed with double-harped strands which can be seen

Post-tensioned concrete beam tests.

Beam number

Top reinforcement

Bottom reinforcement

Prestressing force (kN)

41 42 43 44

2–#3 Rebars 2–#3 Rebars 2–#3 Rebars 2–#3 Rebars

2–#3 Rebars 2–#3 Rebars 2–#4 Rebars 2–#4 Rebars

31.1 15.6 31.1 15.6

Figure 1.

99

Typical detail for post-tensioned concrete beams.

3

FINITE ELEMENT MODEL

This chapter specifically describes the finite element modeling and analysis techniques used for simulating the flexural behavior of post-tensioned concrete beams using ANSYS. ANSYS software is one of the most reliable and popular commercial finite element method programs (Lawrence 2007). 3.1 Figure 2.

Stirrup detail for post-tensioned concrete beams.

Figure 3.

Test setup for post-tensioned concrete beams.

Table 2.

Area of Strands (mm2 ) Ultimate Strength of Prestressing Strand (MPa) Area of Stirrups (mm2 ) Yield Strength of Stirrup, fy , (MPa) Compressive Strength of Concrete, f c , (MPa)

Table 3 shows the details of the element types which were utilized to construct the finite element model. Steel plates were placed at both ends of the beam in order to avoid unrealistic cracks due to stress concentrations. If the steel plates were not added to the ends of the beams in the finite element model, the concentrated prestressing forces would have been applied at very small areas, which would ultimately induce cracks that would initiate at the ends of the beams during the analysis procedure. However, this type of cracking mechanism would not occur during flexural tests for post-tensioned concrete beams. The concrete element type, Solid 65 was used because both cracking in tension and crushing in compression can be considered. 3.2

Properties of strands, stirrups and concrete. 23.2 1862 71 414 32.4

in Figure 1. Figure 1 also illustrates the two point loadings which were applied symmetrically on the tops of the beams. The beams were designed with a length of 3.7 m, a width of 152 mm, and a depth of 254 mm. Figure 2 illustrates the placement of the stirrups in the beams. Each stirrup is spaced 114 mm apart and a total of 31 stirrups were used. The supports of the beams were located 51 mm from the edges of the beams. Figure 3 shows the test setup utilized for the flexural test of the post-tensioned concrete beams, which was performed at CSUF. Strain gages were installed to measure the strain values on the top and bottom rebars and a LVDT was placed in the center of the test beam to measure the deflection. From this data, the stress level of the rebars can be calculated and the load-deflection curve can be obtained. Table 2 shows the material properties of the strands, stirrups, and concrete which were used in the construction of the post-tensioned concrete beams.

Element types

Real constants

The real constants of the post-tensioned concrete beams are described in Table 4 and Table 5. Table 4 Table 3.

Element types for post-tensioned concrete beams.

Material type ANSYS element type

Table 4.

Concrete

Steel plates

Stirrups

Strands

Solid 65

Solid 45

Link 8

Link 8

Real constants for PT beams (41 and 43).

Real constant set

Element type

Cross-sectional area (mm2 )

Initial strain

1 2 3

Solid 65 Link 8 Link 8

Blank 23.2 71.0

Blank 0.00682 Blank

Table 5.

Real constant for PT beams (42 and 44).

Real constant set

Element type

Cross-sectional area (mm2 )

Initial strain

1 2 3

Solid 65 Link 8 Link 8

Blank 23.2 71.0

Blank 0.00341 Blank

100

describes the real constants of beam 41 and beam 43. Table 5 demonstrates the real constants of beam 42 and 44. Tables 4 and 5 show the assigned initial strain values of the prestressing strands were 0.000682 for beams 41 and 43, and 0.00341 for beams 42 and 44. These strain values were calculated from the applied prestressing forces of the test beams. From Table 1, the applied prestressing forces were 31.1 kN for beams 41 and 43 and 15.6 kN for beams 42 and 44. 3.3

Material properties

Figure 6.

The stress-strain for the stirrups and rebars.

Figure 7.

Double-harped post-tensioned beam model.

Figure 8.

The meshed post-tensioned concrete beam model.

This section explains the material properties of the post-tensioned concrete beams. Figure 4 shows the stress-strain curve of the concrete. The value used for the uniaxial tensile cracking stress of concrete was 3.6 MPa (520 psi). During the analysis, if the tensile stress was over 3.6 MPa, cracking would begin to appear. Material properties of the strands were input as multi-linear isotropic material properties. Figure 5 illustrates the stress-strain curve of the strands. Figure 6 shows the stress-strain curve of the stirrups and rebars as bilinear isotropic material properties. 3.4

Modeling

A steel plate was attached at the end of the concrete beam. The stirrups, steel plates, and strands were also modeled as shown in Figure 7. Figure 7 illustrates how the double-harped shape of the strands was modeled as a finite element model. The strands are located

Figure 4.

Figure 5.

The stress-strain of the concrete.

76.2 mm (3.0 in.) from the bottom of the beam, in the middle of the span. The reason why the strands were not placed in the same places as the experimental tests is due to limitations in the size of the mesh.

The stress-strain curve of the strands.

101

If a mesh size less than 1.0 was used, the convergence would have a high tendency to fail in the analysis.

Table 8. Load increment for analysis of finite element model for beam 41.

3.5

Load step

Sub step

Loading on each node (N)

Total loading (N)

1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18

4 4 4 4 4 4 4 4 4 6 6 4 5 4 4 5 4 5

Prestressing Force Self Weight 445 890 1334 1779 2002 2224 2446 2669 3114 3558 4003 4448 4893 5338 5560 5649

0 3336 6005 8673 11342 14011 15345 16680 18014 19349 22017 24686 27355 30024 32693 35361 36696 37363

Meshing

Figure 8 shows the mesh generation of the posttensioned concrete beam model. A 1.0 mesh size was used for this model. Therefore, the concrete beam was meshed with cubes that have the dimensions of 25.4 mm (1.0 in) × 25.4 mm × 25.4 mm. 3.6

Boundary conditions and loading

The loads were applied as two point loadings which were distributed on 3 nodes to avoid stress concentration. The boundary conditions were modeled as a simply supported beam, which are the same as those of the test setup. 4

SOLUTION CONTROL

This chapter describes the solution controls used to analyze nonlinear materials. The values shown in Table 6 were used for simulating the post-tensioned concrete beam model.

Table 6.

Basics of the solution control.

Analysis options Calculate prestress effects Time at end of loadstep Automatic time stepping Number of substeps Max no. of substeps Min no. of substeps Write items to results file Frequency

Table 7.

The values in Table 7 were used for analyzing nonlinear material properties. In this particular case, the convergence criterion for force was discarded in order to avoid convergence problems and the reference value for the displacement criteria was changed to 1.6. Otherwise, if this had not been done, the convergence for the solution control would have had a high tendency to fail in the analysis.

Small displacement Off 0 On 5 30 2 All solution items Write every Substep

5

In order for the nonlinear analysis to be done accurately, the loads are required to have a gradual application, and the nonlinear analysis also requires handling of solution controls. Two point loadings were applied in small incremental loads on beam 41 using the load step and sub step as shown in Table 8.

Nonlinear convergence for solution control.

Line search DOF solution predictor Maximum number of iteration Cutback control Equiv. plastic strain Explicit creep ration Implicit creep ration Incremental displacement Points per cycle Set convergence criteria Label Reference Value Tolerance Norm Min. Ref

Off Program chosen 20 Cutback according to predicted number of iter. 0.15 0.1 0 10,000,000 13 U 1.6 0.05 L2 −1

ANALYSIS PROCESS

6

COMPARISON: TEST AND ANALYSIS

In this chapter, the comparison graphs of the loaddisplacement curves are illustrated in Figures 9 through 12 for beams 41 through 44. The percent differences between the actual tests and results of ANSYS are summarized in Table 9, Table 10, Table 11, and Table 12 for beams 41, 42, 43, and 44, respectively. These tables show the loads at specified deflections of 2.5 mm, 12.7 mm, and 22.9 mm for each beam specimen. For beams 41, 42, and, 43, the percentage differences at a deflection of 2.5 mm range from 27% to 35%, while the percentage differences at deflections of 12.7 mm and 22.9 mm range from 1.3% to 7.3%.

102

Figure 9.

Beam 41 load-deflection graph.

Figure 10.

Beam 42 load-deflection graph.

Figure 11.

Beam 43 load-deflection graph.

Figure 12.

Beam 44 load-deflection graph.

Table 9. Percent difference of load-displacement between actual test and ANSYS for beam 41. Displacement Load (ANSYS) Load (TEST) Difference

2.5 mm 18,014 N 13,122 N 27.16%

12.7 mm 26,243 N 26,688 N 1.67%

22.9 mm 34,250 N 33,805 N 1.3%

Table 10. Percent difference of load-displacement between actual test and ANSYS for beam 42. Displacement Load (ANSYS) Load (TEST) Difference

2.5 mm 15,568 N 10,008 N 35.71%

12.7 mm 23,130 N 22,018 N 4.81%

17.9 mm 28,956 N 27,527 N 4.94%

Table 11. Percent difference of load-displacement between actual test and ANSYS for beam 43. Displacement Load (ANSYS) Load (TEST) Difference

2.5 mm 18,682 N 12,454 N 33.33%

12.7 mm 32,693 N 30,691 N 6.12%

22.9 mm 48,483 N 44,925 N 7.34%

Compared to beams 41, 42 and 43, beam 44 had larger percentage differences but still had similar structural behavior. These relative large discrepancies may be explained by the idealized modeling related to material properties.

7

STRESS CONTOURS AND CRACKING

In this chapter, under the various levels of loadings according to the different load steps, the contours of the Z-component of stress and change in crack patterns can be seen for beam 41. The number of cracks increased and the region of cracking spread when the applied loads were augmented as shown in Figure 13. Figure 14 shows the crack pattern of the test beam. During the test, the sequence of crack development

103

Table 12. Percent difference of load-displacement between actual test and ANSYS for beam 44. Displacement Load (ANSYS) Load (TEST) Difference

2.5 mm 16,458 N 8674 N 47.30%

12.7 mm 30,246 N 22,907 N 24.26%

22.9 mm 44,925 N 37,141 N 17.33%

was marked with numbers and a total of 35 cracks developed at 26,688 N. Figure 15 shows the Z-component stress contour at total loading of 34,027 N.

actual beam test results than the partially prestressing case (15.6 kN). 2. The initial behavior shows more differences than the remaining behavior because the experimental post-tensioned concrete beams are not perfectly elastic within the initial stage. 3. From the comparison results, a modification factor of 0.75 is recommended to predict the loaddeflection behavior of unbonded post-tensioned beams using the proposed ANSYS model in this study conservatively.

Figure 13. Pattern of cracks at total loading of 14,678 N, 21,350 N, and 34,027 N.

If this study proves to be applicable on more experimental beam tests through analysis, more accurate results would be able to be investigated. More test results should be further investigated to more precisely evaluate the validity of the proposed FEM model. Furthermore, this FEM model can be used for simulating the nonlinear flexural behavior of a post-tensioned beam repaired with FRP sheets by adding elements of FRP to this model. REFERENCES

Figure 14. 26,688 N.

Crack pattern of the test beam at total loading of

Figure 15.

The stress contour at total loading of 34,027 N.

8

Chakrabarti, P.R. 1995. Ultimate stress for unbonded posttensioning tendons in partially prestressed beams. ACI Structural Journal 92(6): 689–697. Harajli, M.H. & Kanj, M.Y. 1991. Ultimate flexural strength of concrete members prestressed with unbounded tendons. ACI Structural Journal 88(6): 663–673. Lawrence, K.L. 2007. ANSYS Tutorial. Mission: SDC.

CONCLUSIONS AND FUTURE WORK

From this research, the following conclusions can be reached. 1. The results of the fully prestressing case (31.1 kN) for post-tensioned concrete beams are closer to

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Predicting shear strength of cyclically loaded interior beam-column joints using GAs A. Said University of Nevada, Las Vegas, USA

E. Khalifa Shair and Partners, Cairo, Egypt

ABSTRACT: One of the major problems in evaluating RC structures is estimating the strength of beam-column joints under cyclic loading. This is due to the lack of a clear formula to rely on. Behavior of the cyclically loaded beam-column joints is very complex and several mechanisms control it. Furthermore, several parameters are known to have significant effect on the shear capacity of the joint namely: joint shear reinforcement, concrete compressive strength, joint aspect ratio and column axial stress. The contribution of each of these parameters noticeably varies for each of the proposed formulae in the literature. This paper aims to evaluate some of the existing joint shear capacity formulae of cyclically loaded joints per ACI and AIJ, and subsequently optimize these formulae using genetic algorithms technique (GAs). The paper also is proposing a new formula for calculating the shear capacity of cyclically loaded beam-column joints. 1

INTRODUCTION

Interior beam-column joints have a great importance in reinforced concrete structures. The effect of cyclic loading conditions on interior joints is much higher than the effect of monotonic loading. The reasons behind this are: 1. Larger forces can be generated on the joint for the case of cyclic loading depending on the direction of forces (the ground motion) rather than the monotonic loading case. 2. According to Chopra (2007), the amount of lateral displacement of a RC structure when subjected to cyclic loading is almost twice the amount of the displacement generated by the same force value when applied monotonically to the joint. Through the last few decades, several studies were conducted to investigate the behavior of internal beamcolumn joints subjected to cyclic loading. These studies utilized both analytical techniques (Will et al. 1972; Noguchi 1981; Pantazopoulou & Bonacci 1994; Hwang & Lee 2000; Lowes & Altoontash 2003; Elmorsi et al. 2000) and experimental techniques (Higashi & Ohwada 1969; Durrani & Wight 1982; Otani et al. 1984; Kitayama et al. 1987; Endoh et al. 1991; Joh et al. 1991; Noguchi & Kashiwazaki 1992; Oka & Shiohara 1992; Hayashi et al. 1994; Teraoka et al. 1994; Walker 2001; Zaid 2001) to investigate the shear behavior of the joint. However, despite

105

the extensive analytical and experimental studies conducted, discrepancy still exists between these studies in accurately predicting the shear capacity of the joints and the influence of each of the design parameters on the shear strength. This study aims to investigate the feasibility of using genetic algorithms (GAs) to predict the shear capacity of the cyclically loaded interior beam-column joints. The study will also compare GAs predictions to those obtained from the following equations: ACI-ASCE Committee 352 (2002) and the Architectural Institute of Japan (1998). The parameters investigated in this study are joint volumetric reinforcement ratio, concrete compressive strength, joint aspect ratio, and column axial stress.

2

BEHAVIOR OF INTERIOR BEAM-COLUMN JOINT

In any reinforced concrete frame subjected to seismic loading, beams and columns experience flexure and shear forces. These forces are transformed into higher shear values acting on the joint and they might cause a shear failure in the joint. This type of failure has severe damaging results on the structure. The strut and truss model proposed by Park and Paulay (1975) can be used for the cyclically loaded interior beam-column joints. As shown in Figure 1, two mechanisms are used for the transfer of loads through the joint. The first one is the strut mechanism which accounts for the concrete contribution to

interior planar joints (database case). Accordingly the formula becomes:  (3) Vn = 1.245 fc bj hc 3.2

Figure 1. Strut and truss models proposed by Park and Paulay (1975) for interior beam-column joints.

the shear strength of the joint. In this mechanism, a single concrete compression strut is used to transfer the shear forces through the joint. The second one is the truss mechanism which accounts for the contribution of joint shear reinforcement in transferring the shear forces through the joint. In this mechanism, the load is transferred through a steel tie represented by the joint shear stirrups. To ensure the presence of the tie mechanism, a strong and uniform bond stress distribution along the beam and column reinforcement should exist. 3 3.1

Most of the recommendations provided in the Japanese design guidelines for the cyclically loaded beamcolumn joints are based on studies conducted by Aoyama (1993) on the behavior of cyclically loaded beam-column joints. Based on his study, the Architectural Institute of Japan (1998) provides the following formula for calculating of the shear capacity of cyclically loaded beam-column joints. Vu = k × φ × Fj × bj × D

Vu = 0.68 × (fc )0.70 × bj × D 3.3

According to the ACI-ASCE Committee 352 (2002), the cyclically loaded joints are categorized as Type 2. Type 2 joints are the ones designed to have sustained strength under deformation reversals into the plastic range (seismic loading case). The ACI-ASCE Committee 352 (2002) proposes a general formula for the design of beam-column joints and bases on the type of joint the factors of the formula vary. The general formula is as follows: (1)

where Vn is the nominal shear strength of Type 2 joints, fc is the concrete cylinder strength (MPa), hc is the depth of the column in the direction of joint shear being considered (mm), bj is the effective width of the joint (mm), it is defined as the smaller value of:

(2b)

bc

(2c)

(5)

Genetic algorithms approach

Genetic algorithms are search procedures that use the mechanics of natural selection and natural genetics. The genetic algorithm, first developed by John H. Holland in the 1960’s, allows computers to solve difficult problems. It uses evolutionary techniques, based on functional optimization and artificial intelligence to develop a solution. The sequences of operation of genetic algorithms are as follows: first a population of solutions to a problem is developed. Then, the better solutions are recombined with each other using some special procedures to form a new set of solutions. Finally the new sets of solutions are used to replace the unqualified original solutions and the process is repeated (El-Chabib 2006). A genetic algorithm is used in computing to find true or approximate solutions to optimization and search problems. Genetic algorithms are a particular class of evolutionary algorithms that use techniques inspired by evolutionary biology such as inheritance, mutation, selection, and crossover (Russell and Norvig 2003). Figure 2 presents the steps of typical genetic algorithm model.

ACI-ASCE Committee 352 formula (2002)

b b + bc 2  (mhc + 2) bb +

(4)

where k = 1, Ø = 0.85, Fj = 0.80 * ( fc )0.70 (MPa), D is the column depth, bj = effective column width. This leads the formula to be:

PREVIOUSLY PROPOSED FORMULAE AND EQUATIONS

 Vn = 0.083 γ fc bj hc

Design equation of the Architectural Institute of Japan (1998)

(2a)

4

EXPERIMENTAL DATABASE

The most important factor contributing to the performance of GAs is the input population representing the different initial solutions for the investigated problem.

where m = 0.50 for the case of no eccentricity between the beam and column centerlines, γ = 15 for Type 2

106

5

Start

To predict the shear strength of cyclically loaded beamcolumn joints, a genetic algorithms process was conducted for two main steps, the first one to optimize the existing design formulae mentioned before using the selected the database for improving it’s the performance of these formulae. The second step is to create a new design equation using the same selected database and the investigated parameters. The results of the two steps are compared to each other’s and to the results obtained from the pre optimized formulae. The software used in this model is MATLAB (2007). This software is commonly used for the simulation process of engineering problems. This software divides the given database into training and testing groups to increase the accuracy of the model and give a better understanding of the effect of each parameter in the output layer. Figure 2 represents the architecture of the proposed model.

Generate initial population

Evaluate objective function

Generate new population

Assess optimization criteria

GA MODEL

Yes Stop

No Selection Crossover Mutation

6

Figure 2. Steps of typical genetic algorithms presented by El-Chabib (2006).

To evaluate the accuracy of the proposed Genetic algorithms models, a comparison was held between the proposed new equation and those calculated using the formulae by ACI-ASCE 352 (2002) Architectural Institute of Japan (1998) in their original form and optimized ones. The performance of each model was evaluated based on both the ratio of measured to predicted (or calculated) shear strength (Vm /Vp ), and the average absolute error (AAE) calculated using the following equation:

Table 1. The parameters range for the investigated database. Parameter

Minimum

Maximum

Joint aspect ratio Concrete compressive strength MPa Volumetric reinforcement ratio (%) Column axial stress (MPa)

1

1.3

21.2

70

0 0

3.15 17.8

RESULTS AND DISCUSSIONS

  1  Vm − Vp  × 100 AAE = n Vm

Accordingly, it is imperative to supply the solving program with a comprehensive database to improve the quality of the algorithms procedures including selection, mutation, etc . . . . In this study, shear capacity of this joint type is investigated using a database consisting of 58 concrete beam-column connections collected from published literature listed in Khalifa (2008). The accuracy of the optimization process was improved by imposing several limitations on specimens in the selected database. Specimens failing due to joint shear were strictly used, with no beams in the transverse direction. Specimens with high strength concrete, and reinforcement welding into the joint were omitted. The database was formatted. Table 1 represents the database range of the parameters investigated in the study.

(6)

The average value, the standard deviation (STDV ), and coefficient of variation (COV ) for Vm /Vp , and the average absolute error (AAE) of the ANN model and ACI-ASCE 352 (2002) are listed in Table 2. The shear strength of beam-column joints calculated using current shear design provisions are plotted against the experimentally measured values in Figure 3. Figure 3a indicates that the ACI shear design guidelines for reinforced concrete beam-column joints are highly inaccurate even without application of reduction factors as shown. This formula neglects the influence of the joint aspect ratio and the column axial stress, and the contribution of the joint reinforcement to the shear capacity of the joint. Using the selected data for this study and knowing the actual capacity of the specimens obtained from the experimental programs results, the average absolute error AAE for this formula is 63%, which is significantly high, and the STDV for Vm /Vp of this formula is 0.29.

107

Table 2. Performance of different formulae for the calculation of shear strength of RC interior beam-column joints under cyclic loading. Pre-optimized

Post-optimized Vmeasured / Vpredicted

Vmeasured / Vpredicted

Method

AAE (%)

Ave.

STDV

COV

AAE (%)

Ave.

STDV

COV

ACI352 (2002) AIJ (1998) GA

63 90 –

0.77 0.651 –

0.29 0.297 –

38.80 48.00 –

36 36 21

1.223 1.223 1.07585

0.474 0.474 0.307

38.78 38.78 28.609

Figure 3a. Response of original and optimized formulae of ACI-ASCE 352 equations in calculating the shear capacity of the joint.

Figure 3b. Response of original and optimized formulae of Architectural Institute of Japan (1998) equation in calculating the shear capacity of the joint.

It is recommended that this formula should not be used to estimate the shear capacity of beam-column joints due to its lack of accuracy. It should rather be used to estimate the minimum shear strength of the joint based on concrete properties and joint dimensions. Optimization of this formula using the genetic algorithms managed to reduce the AAE of the formula to 36%. However this value is still high and the performance of the formula is yet questionable. Design equations proposed by the Architectural Institute of Japan (1998) resulted in highly inaccurate prediction of the shear strength of the cyclically loaded interior beam-column joints. Figure 3b represents a plot of the actual experimental shear strength values versus the calculated ones using this formula. This formula neglects the influence of the joint aspect ratio, the column axial stress, and the contribution of joint reinforcements to the shear capacity of the joint. Using the selected data for this study and knowing the actual capacity of the specimens obtained from the experimental programs results, the average absolute error AAE for this formula is 90%, which is extremely high, and the STDV for Vm /Vp of this formula is 0.297. Neglecting several major factors governing the

Figure 3c. Response of the proposed GA equation in calculating the shear capacity of the joint.

108

behavior of the joint refute the accuracy and the validity of this formula. The optimization process for the formula didn’t lead to an accurate form of the formula that could be designed in the design process. The proposed model for the GAs produced much more accurate outputs for predicting the shear capacity of joints than the formula proposed by the previous formulae. Figure 3c shows the performance of the new formula which could be formed as follows:  (7) Vud = 0.165 × hc bj × fc + 0.65Asj fy As noticed from this formula, the formula accounted for only 70% of the joint reinforcement. This result is justified because the actual lever arm between the compression and tension forces in the joint can never be the hall depth of the beam. This formula managed to reduce the error percentage to 18% which is significantly small. Among all the GAs optimization processes, the proposed formula resulted in the lowest AAE. The formula also resulted in a small scatter (0.165) which is less than other formulae. This formula can be used in the evaluation of shear strength of exterior beam-column joints subjected to monotonic loading.

7

CONCLUSIONS

The purpose of this study was to study the feasibility of using genetic algorithms to predict the shear strength of cyclically loaded interior beam-column joints. The proposed technique outperformed existing equations in the ACI code and the literature. The study also shows that GAs is very useful tool for complex engineering problems. Further refinement to the proposed technique can be provided through incorporating new experimental research results.

REFERENCES Aoyama, H. Empirical versus Rational Approach in Structural Engineering—What We Learned from New Zealand in the Trilateral Co-operative Research on Beam-Column Joint, ACI Special Publication SP-I Detroit, September 1993, pp. 31–57. Architectural Institute of Japan, 1998. Recommendations of RC Structural Design after Hanshin-Awaji Earthquake Disaster-Cause of Particularly Noticed Damages and Corresponding RC Structural Details. Chopra, A.K. 2007. Dynamics of Structures, Prentice Hall, Englewood Cliffs, New Jersey. Durrani, A.J. & Wight, J.K. 1982. Experimental and Analytical Study of Beam to Column Connections Subjected to Reserve Cyclic Loading. Technical Report UMEE82 R3, Department of Civil Engineering, University of Michigan, 295 p. Elmorsi, M., Kianoush, M.R. & Tso, W.K. 2000. Modeling Bond-Slip Deformations in Reinforced Concrete

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Beam-Column. Canadian Journal of Civil Engineering 27: 490–505. Endoh, Y., Kamura, T., Otani, S. & Aoyama, H. 1991. Behavior of RC Beam-Column Connections Using LightWeight Concrete. Transactions of Japan Concrete Institute, 319–326. Haykin, S. 1994. ‘‘Neural Networks: A Comprehensive Foundation’’, Macmillan, New York, 842 p. Hayashi, K., Teraoka, M., Mollick, A.A. & Kanoh, Y. 1994. Bond Properties of Main Reinforcing Bars and Restoring Force Characteristics in RC Interior Beam-Column Assemblages Using High Strength Materials. In Proc. 2nd US-Japan-New Zealand-Canada Multilateral Meeting on Structural Performance of High Strength Concrete In Seismic Regions, Honolulu, Hawaii, pp. 15–27. Higashi, Y. & Ohwada, Y. 1969. Failing Behavior of Reinforced Concrete Beam-Column Connections Subjected to Lateral Load. Memories of Faculty of Technology Tokyo Metropolitan University, Tokyo, Japan, pp. 91–101. Hwang, S.J. & Lee, H.J. 2000. Analytical Model for Predicting Shear Strength of Interior Reinforced Concrete BeamColumn Joints for Seismic Resistance. ACI Structural Journal 97 (1): 35–44. Joh, O., Goto, Y. & Shibata, T. 1991. Influence of Transverse Joint, Beam Reinforcement and Relocation of Plastic Hinge Region on Beam-Column Joint Stiffness Determination. In ACI Special Publications SP 123-12: Design of Beam-Column Joints for Seismic Resistance, Farmington Hills, Michigan, pp. 187–223. Joint ACI-ASCE Committee 352, 2002. ‘‘Recommendation for Design of Beam-Column Connections in Monolithic Reinforced Concrete Structures’’, American Concrete Institute, Farmington Hills, Mich, 40 p. Khalifa, E. 2008. Evaluation of Beam-Column Joint Strength Using Artificial Intelligence. M.Sc. Thesis, Department of Civil and Environmental Engineering, University of Nevada, Las Vegas, USA. Kitayama, K., Otani, S. & Aoyama, H. 1987. Earthquake Resistant Design Criteria for Reinforced Concrete Interior Beam-Column Joints. In Pacific Conference on Earthquake Engineering, Wairakei, New Zealand, pp. 315–326. Kurzweil, R. 1990. The Age of Intelligent Machines. MIT Press, Cambridge, Massachusetts. Lowes, L.M. & Altoontash, A. 2003. Modeling ReinforcedConcrete Beam-Column Joints Subjected to Cyclic Loading. Journal of Structural Engineering 129 (12): 1686–1697. Noguchi, H. & Kashiwazaki, T. 1992. Experimental Studies on Shear Performance of RC Interior Column-Beam Joints. In 10th World Conference on Earthquake Engineering, Madrid, Spain, pp. 3163–3168. Noguchi, H. 1981. Nonlinear Finite Element Studies on Shear Performance of RC Interior Column-Beam Joints. In IABSE Colloquium, Delft, The Netherlands, pp. 639–653. Oka, K. & Shiohara, H. 1992. ‘‘Test on High -Strength Concrete Interior Beam-Column Sub-Assemblages’’. In 10th World Conference on Earthquake Engineering, Madrid, Spain, pp. 3211–3217. Otani, S., Kobayashi, Y. & Aoyama, H. 1984. Reinforced Concrete Interior Beam-Column Joints under Simulated Earthquake Loadings. In US-New Zealand- Japan Seminar on Design of Reinforced Concrete Beam-Column Joints, Monterey, CA.

Pantazopoulou, S. & Bonacci, J. 1994. On Earthquake Resistant Reinforced Concrete Frame Connections. Canadian Journal of Civil Engineering 21: 307–328. Park, R. & Paulay, T. 1975. ‘‘Reinforced Concrete Structures’’, John Wiley and Sons, United States of America, 769 p. Teraoka, M., Kanoh, Y., Tanaka, K. & Hayashi, K. 1994. Strength and Deformation Behavior of RC Interior Beam-Column Joint Using High Strength Concrete. In Proceedings, Second US-Japan-New Zealand-Canada Multilateral Meeting on Structural Performance of High Strength Concrete In Seismic Regions, Honolulu, Hawaii, pp. 1–14. The Math Works., 2007. ‘‘MATLAB (2007)’’. Orchard Hill, Michigan, United States.

Walker, S.G. 2001. Seismic Performance of Existing Reinforced Concrete Beam Column Joints. MSCE Thesis, University of Washington, Seattle. 308 p. Will, G.T., Uzumeri, S.M. & Sinha, S.K. 1972. Application of Finite Element Method to Analysis of Reinforced Concrete Beam-Column Joints. In Proceeding of Specialty Conference on Finite Element Method in Civil Engineering, CSCE, EIC, Canada, pp. 745–766. Zaid, S. 2001. Behavior of Reinforced Concrete BeamColumn Connections under Earthquake Loading. PhD Dissertation, Department of Architecture, University of Tokyo, Japan.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Punching shear strength of RC slabs using lightweight concrete H. Higashiyama Kinki University, Osaka, Japan

M. Mizukoshi Takamatsu National College of Technology, Kagawa, Japan

S. Matsui Osaka Institute of Technology, Osaka, Japan

ABSTRACT: Lightweight aggregate concrete (LWAC) is useful to reduce the structural dead load. The punching shear strength of reinforced concrete (RC) slabs using LWAC decreases in comparison to that using normal-weight aggregate concrete (NWAC) due to the weakness of the aggregate in concrete. Therefore, a reduction factor for LWAC is specified in several international design codes/standards of JSCE, ACI, CEB-FIP, etc. In the present study, a reduction factor for the punching shear strength of RC slabs using LWAC was investigated using the experimental and analytical results. Consequently, the punching shear strength decreases with the density of LWAC and the reduction factor is remarkably related to the characteristic length of concrete proposed by Hillerborg et al. (1976).

1

The corresponding proposed reduction factor, which is derived from the experimental and analytical results, is compared with several design codes/standards.

INTRODUCTION

Density of lightweight aggregate concrete (LWAC) using expansive shale is 20 to 30% smaller than that of normal-weight aggregate concrete (NWAC). Therefore, LWAC is useful to reduce the dead load to the substructures and improve the earthquake resistance of concrete structures. The tensile and shear strengths of LWAC, however, decrease due to the lower strength of lightweight aggregate itself. Applying LWAC to highway bridge slabs, it is important to adopt a suitable design method taking into account the fracture properties of the material. In the practical design procedures, a reduction factor for LWAC is accepted in several design codes/ standards such as JSCE, ACI, and CEB-FIP. The JSCE standard specifications (2002) have a constant reduction factor of 0.7 for all lightweight aggregate concrete (ALWAC). The other LWAC is not specified in the JSCE standard specifications. The ACI code (2005) gives constant reduction factors considering the density of LWAC. On the other hand, the CEB-FIP Model Code 90 (2001) has an equation of reduction factor varying with the concrete density. From the previous test results (Higashiyama et al. 2006), however, the punching shear capacities of reinforced concrete (RC) slabs using LWAC decreased with the concrete densities. The purpose of this study is to propose experimentally and analytically the reduction factor for the punching shear capacity of RC slabs using LWAC.

2

DESIGN CODES/STANDARDS AND PREDICTING EQUATION

2.1

JSCE standard specifications

According to the JSCE standard specifications (2002), the punching shear capacity, Vc is: Vc = fpc βd βp βr up d

(1)

 fpc = 0.20 fc ≤ 1.2

(2)

βd =

 4 1/d ≤ 1.5

βp =

 3 100p ≤ 1.5

111

βr = 1 +

(d : m)

1 1 + 0.25u/d

(3) (4) (5)

where fc is the compressive strength of concrete, d is the average effective depth, p is the average reinforcement ratio, u is the perimeter of the concentrated load, and up is the perimeter of the critical section located at a distance of 0.5d from the periphery of the concentrated load (=u + π d).

The JSCE standard specifications (2002) give a constant reduction factor of 0.7 for ALWAC. All safety factors and material resistance factors in this calculation are taken as 1.0 for the objective of this research work. Hereinafter, the same procedure is followed.

2.4

Several punching shear predicting equations have been proposed so far worldwide. In the present study, the failure model proposed by Maeda & Matsui (1984) as shown in Figure 1, which has been recognized with applicability to the highway bridge RC slabs in Japan, is employed to evaluate the punching shear capacity. The equation can be expressed as follows:

2.2 ACI code According to ACI code 318-05 (2005), the punching shear capacity, Vc , is taken as the smallest of:

 Vc =

Vc =

   f αs d + 2 λ c b0 d b0 12

1   λ fc b0 d 3

Vc = fcv {2 (a + 2xm ) xd + 2 (b + 2xd ) xm } + ft {2 (a + 2dm ) Cd + 2 (b + 2dd + 4Cd ) Cm } (11)

(6)

ft = 0.269fc 2/3

(12)

fcv = 0.656fc 0.606

(13)

(7)

(8)

a f cv

ft

d

where fc is the compressive strength of concrete, d is the average effective depth, b0 is the perimeter of the critical section located at a distance of 0.5d from the periphery of the concentrated load, λ is a reduction factor to account for the concrete density, β is the ratio of the long side to short side of the concentrated loaded area, and αs is a factor to account for the location of the slab-column connection. The ACI code (2005) requires reduction factors of 0.85 for sand lightweight aggregate concrete (SLWAC) and 0.75 for ALWAC.

where a and b are side lengths of a loading plate in the main bar and distribution bar directions, respectively, xm and xd are depths of the neutral axis neglecting the tension side of concrete in the main bar and distribution bar directions, respectively, dm and dd are effective depths of tensile reinforcement for the main and distribution bars, respectively, Cm and Cd are depths of cover concrete for the tensile main

45˚ C

CEB-FIP Model Code 90

According to the CEB-FIP Model Code 90, Fib (2001), the punching shear capacity is given by:

ηt = 0.4 + 0.6

a

xm 2C d

(9)

dd

 ρ 2200

xm

(10)

where fc is the compressive strength of concrete, d is the average effective depth, b0 is the perimeter of the critical section located at a distance of 2d from the periphery of the concentrated load, p is the average reinforcement ratio, ηt is a reduction factor, and ρ is the oven-dry density of the LWAC. The CEB-FIP Model Code 90, Fib (2001) gives the equation of reduction factor taking into account the concrete density as outlined in Equation 10.

dd

Vc = 0.18ηt 1 +

 200 (100pfc )1/3 b0 d d

2C d



d

xd b xd



2C

b

2.3

x

    f 2 λ c b0 d Vc = 1 + β 6

Matsui’s predicting equation

2C m

dm

a

dm

2Cm

Figure 1. Punching shear failure model proposed by Maeda & Matsui (1984).

112

and distribution bars, respectively, ft and fcv are the tensile strength and shear strength of concrete (N/mm2 ), respectively. 3

EXPERIMENTAL APPROACH

3.1

(D10). Table 3 gives the details of slabs tested in the present study. Five types of slabs, which were varied in size, reinforcement ratio, and concrete material, were prepared and two same slabs in each type were made considering unexpected deviations of experimental results. All slabs were square, flat RC slabs. Figure 2 shows

Materials and test specimens

Table 1 gives the details of materials used herein. Ordinal Portland cement was employed. River sand and artificial lightweight aggregate were used for fine aggregates. And crushed stone and artificial lightweight aggregate were used for coarse aggregates. The fine and coarse artificial lightweight aggregates are composed of expansive shale. Mix proportions of concrete for specimens tested herein are listed in Table 2. Reinforcing bars with a deformed shape were used with a diameter of 6.35 mm (D6) and 9.53 mm Table 1.

Mixture proportions.

Unit content (kg/m3 ) Mix No. W/C W C S G 1 2

0.48 0.50

Ad

180 375 812 (RS) 555 (ES) 0.75 (Ad-1) 167 334 656 (ES) 913 (CS) 3.34 (Ad-2)

Notes: RS = river sand; ES = expansive shale; CS = crushed stone; Ad-1 = high performance air-entraining & water reducing admixture; Ad-2 = air-entraining & water reducing admixture.

Material properties.

Material

Type

Properties

Cement Fine aggregate

Portland River sand

Density: 3.15 g/cm3 Density: 2.59 g/cm3 Absorption: 1.39% Maximum size: 2.5 mm Density: 1.94 g/cm3 Absorption: 16.0% Maximum size: 2.5 mm Density: 2.70 g/cm3 Absorption: 0.71% Maximum size: 20 mm Density: 1.65 g/cm3 Absorption: 28.0% Maximum size: 15 mm

Expansive shale Coarse aggregate

Crushed stone

Expansive shale

Note: Density of aggregates is on saturated surface-dry condition. Table 3.

Table 2.

Figure 2.

View of test setup.

Details of test slabs. Main bar (mm)

Dist. bar (mm)

Specimen

Mix no.

Span length (mm)

Thickness (mm)

Upper bar Lower bar

Upper bar Lower bar

LC-1

1

1000

100

LC-2

1

1000

100

LC-3

1

1000

150

LC-4

1

1000

150

LC-5

2

600

80

D10-ctc. 160 D10-ctc. 80 D10-ctc. 240 D10-ctc. 120 D10-ctc. 160 D10-ctc. 80 D10-ctc. 240 D10-ctc. 120 non D6-ctc. 50

D10-ctc. 160 D10-ctc. 80 D10-ctc. 240 D10-ctc. 120 D10-ctc. 160 D10-ctc. 80 D10-ctc. 240 D10-ctc. 120 non D6-ctc. 50

113

Effective depth (mm)

Main bar

Dist. bar

Compressive strength (N/mm2 )

Concrete density (kg/m3 )

80.0

70.5

41.3

1810

80.0

70.5

41.3

1810

110.0

100.5

38.7

1840

110.0

100.5

38.7

1840

56.8

50.4

37.8

2084

2.0

1.5

1.5

Vexp / VACI

Vexp / VJSCE

2.0

1.0 Authors Hamada et al. JSCE Ito et al.

0.5

0.0 1400

1600

1800

2000

1.0

0.0 1400

2200

3

2200

1.0 0.8

Vexp / VMatsui

Vexp / VFIB

2000

(b) ACI code

1.5

1.0 Authors Hamada et al. JSCE Ito et al.

0.5

1600

1800

2000

0.6 0.4

Authors Hamada et al. JSCE Ito et al.

0.2 0.0 1400

2200

1600

1800

2000

2200

3

3

Concrete desnity (kg/m )

Concrete density (kg/m )

(c) CEB-FIP Model Code 90

(d) Matsui s equation

Comparison of design codes/standards and predicting equation with experimental results.

the view of test setup. All slabs were simply supported along four sides, which were free to rising of the slab corners. Concentrated load was applied at the center of slab by an oil jack through the square steel plate. Since the number of specimens tested herein was limited, test results conducted by Hamada et al. (2002), JSCE Concrete Library (2001), and Ito et al. (2005) were also included to evaluate the punching shear capacity. Hamada et al. (2002) tested using natural sand and expansive shale for the fine aggregate and expansive shale for the coarse aggregate. JSCE Concrete Library (2001) tested using natural sand for the fine aggregate and high strength lightweight aggregate made of fly ash for the coarse aggregate. Ito et al. (2005) tested using natural sand for the fine aggregate and high performance artificial lightweight aggregate made of sediments from the Yellow River in China for the coarse aggregate.

3.2

1800

Concrete density (kg/m )

2.0

Figure 3.

1600

3

Concrete density (kg/m ) (a) JSCE standard specifications

0.0 1400

Authors Hamada et al. JSCE Ito et al.

0.5

almost linear trend between the normalized punching shear capacities and the concrete densities. The normalized punching shear capacity of the smaller concrete density, however, is less than 1.0 as shown in Figure 3 (a) to (c). Matsui’s equation overestimates in all test results. From these results, it is noticed that the reduction factor for LWAC is significantly related to the concrete density.

4 4.1

ANALYTICAL APPROACH Finite element analysis

To propose a reduction factor for the punching shear capacity of RC slab using LWAC, the punching shear behavior was analyzed by means of a commercially available 3D finite element program, ATENA (Cervenka Consulting 2002). In this analysis, a fictitious crack model based on a crack opening law and fracture energy after cracking in tension was used. The stress-crack opening relation was used the formulation derived by Hordijk (1991) as follows:

Comparison of design codes/standards and predicting equation with experimental results

Normalized punching shear capacities, which mean the experimental values divided by the predicted values, versus the concrete densities are plotted in Figure 3 (a) to (d). It is noted that the normalized punching shear capacities decrease when lowering the concrete densities with some scattering. The design codes/standards, except for Matsui’s equation, have an

114

     σt w 3 w exp −c2 cr = 1 + c1 cr ft wt wt −

w 1 + c13 exp(−c2 ) wtcr

(14)

4.2

Figure 4. Stress-strain relation in compression on the ascending branch.

wtcr = 5.14

(15) k = 0.4 + 0.6

where σt is the tensile stress in the crack, ft is the tensile strength, w is the crack opening, wtcr is the crack opening at the complete release of stress, Gf is the fracture energy, and the constant values are c1 = 3.0 and c2 = 6.93. The stress-strain relation in compression on the ascending branch is linear up to fco and is given by the following formula from fco to fc (see Figure 4): 



p

p

εc − εeq p εc

fco =

2  f 3 c

(17)

εcp =

fc Ec

(18)

1−

(16)

where σc is the compressive stress, fc is the compresp sive strength, εeq is the equivalent plastic strain, and Ec is the Young’s modulus of concrete. Furthermore, the stress-displacement relation on the descending branch is given by the following formula derived by van Mier (1986):   wc σc = 1 − cr fc wc

(20)  ρ 2400

(21)

where, ft,L and ft,N are the tensile strengths of LWAC and NWAC, respectively, and ρ is the concrete density (kg/m3 ).

2

σc =

−

fco

fco

+

fc

Two types of RC slabs were determined as the analytical model as shown in Table 4. The thickness was 100 mm and 150 mm, the effective depth was 75 mm and 105 mm, and the reinforcement ratio was 1.19% and 1.17%. RC slabs were simply supported along four sides with a span length of 1000 mm. The analytical model is shown in Figure 5. The compressive strength of concrete was fixed at 40 N/mm2 . The concrete density was varied from 1400 kg/m3 to 2200 kg/m3 . Furthermore, the tensile strength (Walraven 2000), the Young’s modulus (AIJ 1991), and the fracture energy (Higashiyama et al. 2006) corresponding to the concrete density were determined by the following equations: ft, L = k · ft, N

Gf ft



Parameters of analytical model

 Ec = 2.1 × 10

5

ρ ρ0

1.5 

fc 200

where Ec is the Young’s modulus of concrete (kgf/cm2 ), fc is the compressive strength (kgf/cm2 ), and ρ is the concrete density (t/m3 ), ρ0 = 2.3 t/m3 .  Gf = 2.85 fc (ρ/1000)2.45 Table 4.

(23)

Analytical models.

Span Average Average length Thickness reinforcement effective Specimen (mm) (mm) ratio (%) depth (mm) Model-1 Model-2

1000 1000

100 150

1.19 1.17

Figure 5.

Finite element model.

(19)

where σc is the compressive stress, fc is the compressive strength, wc is the plastic displacement, and wccr is the plastic displacement at the zero compressive stress (=0.5 mm in this study). The details of the finite element program are referred to ‘‘ATENA Program Documentation’’ (Cervenka Consulting 2002).

115

(22)

75 105

From the analytical results mentioned above, the reduction factor for the punching shear capacity of RC slab using LWAC proportionally relates with the characteristic length of concrete as follows:

1.2

Reduction factor

1.0 0.8

 α = k 1 + k2

0.6

t = 150 mm Experiment

0.0 1200

(24)

1400

1600

1800

2000

2200

2400

where lch,L and lch,N are the characteristic lengths of LWAC and NWAC, respectively, and k1 and k2 are constant values. Equation 24 can be rewritten by substituting Equations 20 through 23 as follows:

3

Concrete density (kg/m )



Figure 6. Relation between the normalized punching shear capacity and the ratio of characteristic length.

α = k 1 + k2

ρ ρ0

2.95 (25)

where ρ is the concrete density (kg/m3 ), and ρ0 = 2300 kg/m3 (generally used value in Japan). When Matsui’s equation is used to predict the punching shear capacity, the values of the constant, k1 and k2 , are derived as 0.28 and 0.72, respectively, by using the least-squares curve fitting with the experimental data as shown in Figure 7. It should be pointed out that the reduction factor derived in Equation 25 is applicable only for the LWAC investigated herein.

1.2 Normalized punching shear capacity



t = 100 mm

0.4 0.2

1.0 0.8 0.6 0.4 t = 100 mm 0.2

t = 150 mm

5

0.0 0

0.2

0.4

0.6

0.8

1

CONCLUSIONS

1.2

Ratio of characteristic length

Figure 7. Relation between the reduction factor and the concrete density.

where Gf is the fracture energy (N/m), fc is the compressive strength (N/mm2 ), and ρ is the concrete density (kg/m3 ). 4.3

lch, L lch,N

Proposed reduction factor

Reduction of the punching shear capacity of RC slab using LWAC depends on a difference of the material fracture properties of concrete in tension. The characteristic length (Hillerborg et al. 1976) is a measure for the brittleness of the material and is defined by three material parameters, the Young’s modulus (Ec ), the tensile strength (ft ) and the fracture energy (Gf ). Figure 6 shows the relation between the normalized punching shear capacities, which mean the analytical values divided by Matsui’s equation, Equation (11), and the ratios of the characteristic lengths of LWAC and NWAC. It is noted that there is a proportional relationship between those values.

This paper presents the punching shear capacity of RC slabs using LWAC. The punching shear test results were compared with several design codes/standards. Several codes/standards give the reduction factor for LWAC. However, the predicting equations in design codes/standards for the punching shear capacity tend to be conservative. The normalized punching shear capacity linearly relates to the characteristic length of concrete. Predicting accuracy is the average of 0.99 and the standard deviation of 0.077. The reduction factor proposed herein is applicable for the punching shear capacity of RC slabs using LWAC.

REFERENCES ACI 318R-05. 2005. Building Code Requirements for Structural Concrete and Commentary, American Concrete Institute, Farmington Hills, MI. AIJ. 1991. Standard for Structural Calculation of ReinforcedConcrete Structures, Architectural Institute of Japan (in Japanese). Cervenka Consulting. 2002. ATENA Computer Program for Nonlinear Finite Element Analysis of Reinforced Concrete Structures, Program Documentation. Fib. 2001. Punching of Structural Concrete Slabs, Bulletin 12, Federation Internationale du Beton, Lausanne.

116

Hamada, S., Mao, M., Yoshitake, I. & Tanaka, H. 2002. Slabs with Lightweight Aggregate Concrete of High Strength, Session 9: High-performance concrete, Proc. of 1st fib congress, Osaka, pp. 267–272. Higashiyama, H., Mizukoshi, M. & Matsui, S. 2006. Prediction Method on Punching Shear Capacity of Reinforced Lightweight Concrete Slabs Considering Fracture Characteristics, Concrete Research Technology, JCI, Vol. 17, No. 2, pp. 23–31 (in Japanese). Hillerborg, A., Modeer, M. & Petersson, P.E. 1976. Analysis of Crack Formation and Crack Growth in Concrete by Means of Fracture Mechanics and Finite Elements, Cement and Concrete Research, Vol. 6, pp. 773–782. Hordijk, D.A. 1991. Local Approach to Fatigue of Concrete, Doctoral Dissertation, Delft University of Technology. Ito, H., Iwanami, M. & Yokota, H. 2005. Influence of PVA Short Fibers on Punching Shear Behavior of Lightweight Concrete Slab, Journal of Structural Engineering, JCSE, Vol. 54A, pp. 1321–1331 (in Japanese).

117

JSCE Concrete Library. 2001. Recommendations for Design and Construction of Concrete Structures Using High Strength Lightweight Aggregate Made of Fly Ash, 106 (in Japanese). JSCE Standard Specifications. 2002. Standard Specifications for Concrete Structures-2002, Structural Performance Verification, Japan Society and Civil Engineers (in Japanese). Maeda, Y. & Matsui, S. 1984. Punching Shear Equation of Reinforced Concrete Slabs, Journal of JSCE, No. 348/V-1, pp. 133–141 (in Japanese). van Mier, J.G.M. 1986. Multiaxial Strain-softening of Concrete, Part I, Fracture, Materials and Structures, RILEM, Vol. 19, pp. 179–190. Walraven, J. 2000. Design of Structures with Lightweight Concrete: Present Status of Revision EC2, Proc. of the Second International Symposium on Structural Lightweight Aggregate Concrete, Kristiansand, pp. 57–70.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Shear strength and deformation prediction in steel fiber RC beams T. Nyomboi & H. Matsuda Structural Engineering Department, Nagasaki University, Nagasaki, Japan

ABSTRACT: In this paper, an analytical shear strength predictive equation for steel fiber RC beams without stirrups is proposed. In the model, steel fibers are considered to resist tension through stress transfer after cracking, where a pullout mechanism is assumed to govern. Resistances from compressed concrete, crack/aggregate interlocking slip and flexural reinforcement dowel action are all considered in unified manner. By formulation and solution of force and moment equilibrium conditions as well as shear strain and curvature ratio analysis, a unified strength and deformation capacity predictive formula is proposed. The response of the proposed theoretical model is validated against FEM and experimental results obtained from a total of 8 steel fiber RC beams. Variably reinforced steel fiber (0 to 1.5% content by volume) RC beams were tested under a variable shear span to depth ratios. Results are showed to agree well with the FEM and experimental results.

1

INTRODUCTION

In reinforced concrete flexural members, premature shear failure, which is brittle and occurs suddenly, must be avoided, if maximum flexural bearing capacity in critical areas is to be attained. Steel fibers have been found to enhance the performance of RC beams through stress redistribution after cracking and by transforming the brittle failure to a ductile mode. In this regard and for purposes of development of design guidelines, prediction of the strength and deformation capacity is fundamental. Concrete reinforced with steel fibers have been applied in tunnels, wall cladding, industrial slabs, pipe repair, bridge decks and pavements (Balendar et al. 2001; Rossi 1994). A number of studies have also confirmed the potential of steel fibers as a supplement or replacement of stirrup reinforcements (Madhusudan et al. 1999; Yoon-keun et al. 1999; Pasacal et al. 1997; Colegero et al. 2004). Significant reduction in construction time and cost savings are noted as additional merits (Madhusudan et al. 1999; Greenough et al. 2007). Existing empirical and theoretical shear strength models (e.g Kyoung-kyu et al. 2007 and Lim at al. 1999) are mainly ultimate based. Evolution based theoretical models in which the strength behavior through yielding to ultimate failure can be predicted are unavailable. More over, the existing models ignore the dowel action of the reinforcements. The dowel action and the tensile capacity provided by the fibers significantly enhance the shear capacity of fiber reinforced concrete beams (Greenough et al. 2007). Lack of a universally accepted approach in the calculation of the strength for steel fiber reinforced concrete (SFRC) is one of the major obstacles in application of the material in design (Brandt 2008).

119

The objective of the present study is therefore to propose a unified theoretical model for the shear strength evolution response to failure in SFRC beams. Theoretical results from the proposed model are experimentally and numerically validated and are found to be in accord with these verification results. 2 2.1

SHEAR STRENGTH EVOLUTION MODEL Force relations

In the present study, a method that introduces a shear strain ratio is applied based on the interactions of the forces (Figs. 1, 2) and the sheared geometry of a beam (Gere & Timonshenko 1989) in which cracked and compressed regions are idealized (Fig. 1b). In the derivation of the model the following principle assumptions were made; (i) Plane section remain plane (ii) shear crack occurs at 45 degrees (iii) although concrete is brittle, it possesses some minimal tensile strength (iv) on yielding (concrete cracking), the fibers are assumed to initially elastically strain, de-bond and eventually pull out gradually from one side of the crack (v) dowel action of the flexural reinforcement bars contribute to the shear strength. 2.1.1 Compressive, tensile and aggregate/crack slip shear forces In a beam under bending-shear, tension and compression zones exist in which the compression zone is subjected to combined compressive and shear forces. The expressions for the concrete compressive and tensile forces shown in Figure 2 are given by:   w Fcc = σc b c − (1) ψ

K

c

h

N.A R.bar

Reinforcement bar J

M

Q /2

2.1.2 Fiber forces (Ft(1) , Ft(2) ) In fiber reinforced concrete beam, after tensile cracking occurs, tensile stresses are transmitted across the crack and a resistive mechanism is developed (Brandt 2008). The tensile stress transmitted is depended on; the effective number of fibers that bridge the crack orthogonally, bond strength and frictional forces. Practically, a complex interaction of these factors takes place. A simplified approach is adopted in the present study. For expressions of the fiber forces Ft(1) andFt(2) (Fig. 2a), an average effective fiber force is first established. The average number of effective fibers crossing a crack orthogonally in a plane (Stroeven & Hu 2006; Hannant 1975), is given by:

Steel fibers orientated along and across the crack path

Q kN L

b

Q /2

a l

(a) Basic beam details K‘

K

L

Compressed region

L‘

v

= w cos

h

= w sin

O

J‘

Cracked region

w

v

J‘

w

M

a

J

J

M‘

Neff = ζ Nf

h

(3)

Crack geometry at J

where, NF is the total number of fibers per unit area and ζ = 2/π is the average fiber projection (orthogonal) factor. A fiber subjected to a pull out force must first de-bond from the surrounding matrix before it pulls out (Yang et al. 2008). It is assumed in this study that the fibers behave elastically in the initial stage after which a pull out occurs (Fig. 2a). From equation (3) the general expression for the average effective tensile force carried by the fibers in a unit area is given by:

(b) Sheared profile of section JKLM

Figure 1.

Conceptual model beam details.

w

c

c

F cc

w

w

x

x1

Fa

Fv

F cv

o

x1

Fc

⎛ ⎜⎜c ⎝

o w

dx

sin

w

dF t (1)

F t (1) F ct F t (2)

Fs

Q/2

Fd w

w⎞ ⎟⎟ sin ⎠ o

Feff = Neff Ff = ζ Nf Ff

w

fp

εy ). According the critical conception limit strain εu = εcr . For the web subjected to bending with pin supporting along both longitudinal borders critical strain (Juhás 1984) εcr = 21.582/β 2

their bending webs and compressed flanges is complicated. Therefore, the representative experimental knowledge and results about the real elastic-plastic behavior and failure mechanisms of steel members in their decisive cross-sections and areas are very important for development and précising of the elasticplastic analysis and design.

EXPERIMENTAL RESEARCH AND SELECTED RESULTS

The experimental program included tests of 32 beams with usually welded I cross-sections of different geometrical dimensions, slenderness β and stiffening of the web. The designed geometrical dimensions are listed in Table 1 and static schemes of the tested beams are presented in Figure 2. The tested beams were made by usual productiontechnological process and conditions with some geometrical and material imperfections. The main aims of experimental program were investigate the elastic-plastic load-carrying capacity and actual behavior of steel members in the most loaded cross-sections and areas with accent on the development of plastic strains ε, local buckling of their webs w, global deflections v and mechanisms of member failure. The tested beams were loaded by successively increasing concentrated forces P, realized by hydraulic jacks. The tests continued till beginning of beam failure. All beams were horizontally supported to prevent their lateral-torsion buckling. The objectives of all realized beam tests were to investigate: – successive development of the elastic-plastic strains ε in the most loaded points, cross-sections and stiffened web fields by tensometers and measuring exchange, – local buckling w in the most loaded stiffened web fields by static and movable deflection pickups,

(12)

More theoretical works and some results about the elastic-plastic local stability and load-carrying capacity of the steel member webs already exist at present (Djubek et al. 1983, Salamon & Johnson 1995). But the real elastic-plastic behavior of steel members and

143

Table 1.

Geometrical dimensions of cross-sections. mm

Beams

h

b

tf

tw

N11, N21, N31, N41 N12, N22, N32, N42 N13, N23, N33, N34 N14, N24, N34, N44

380

160

10

3 4 5 6

C11, P11, R11, C12 R12, C13, P13, R13 C21, R21, C22, P22 R22, C23, P23, C23

520

200

10

6

820

β = (h − 2tf )/tw , βf = (b − tw )/2tf , γ = AW /A

Figure 3. Typical web and flange failure mechanism of tested beams.

Figure 2.

Static schemes of the tested beams.

beam loading P and web slenderness β. If the web slenderness β is not too large, then the buckling w is small up to some level of the loading. According to the classical stability conception the equivalent load has been accepted as the experimental critical load Pcr,exp . On the base of observed dependencies P − w the experimental critical load Pcr,exp was assigned for every relevant tested beam. The experimental critical load Pcr,exp can be generally less or higher then the theoretical elastic limit load Pel or even plastic limit load Ppl . Accordingly the critical load Pcr,exp can be elastic (Pcr,exp,el ), elastic-plastic (Pcr,exp,ep ) or plastic (Pcr,exp,pl ) experimental critical load, so

– global vertical deflections v in the place of end supports and in middle cross-sections of the beams by mechanical or electrical deflectometers.

• if Pcr,exp < Pel then Pcr,exp = Pcr,exp,el , • if Pel < Pcr,exp < Ppl then Pcr,exp = Pcr,exp,ep and • if Pcr,exp > Ppl then Pcr,exp = Pcr,exp,pl .

Accordance with the research aims all beams failure in consequence of the local web and flange buckling in their most loaded areas and stiffened fields. Generally, the web buckling was the reason for failure, but total failure of the tested beams formatted by induced buckling of their compassion flanges. The typical failure mechanism of tested beams in the decided web field subjected to pure bending is presented in Figure 3. The measured, evaluated and analyzed values of strains ε, buckling w and deflections v have offered very important information and principal knowledge about local and global failure mechanisms of tested beams in dependency on the level and process of their loading (Juhás 2005). The local stability of tested beams well characterized investigated dependencies of the load P and web buckling w in individual points of the most loaded areas and stiffened fields. The buckling w of the web subjected prevailing to bending depends first of all on its slenderness β. Generally, the web buckling w increased in accordance with increasing of the

The obtained experimental results confirmed that the elastic-plastic load-carrying capacity of the tested beams depends on local buckling of their webs (Juhás 2005). For illustration the dependences of load P and strains εx in the individual measured points of the most loaded cross-section of the beam N12 are presented in Figure 4, Pel and Ppl are theoretical elastic and plastic limit load, Pcr,exp , and Pu,exp are experimental critical and ultimate load. The obtained experimental knowledge about the real development of plastic strains ε in the most loaded cross-sections and areas of the tested beams are very important for the determination of the elastic-plastic and plastic load-carrying capacity of steel members from local stability aspects. The obtained experimental results confirmed that the load-carrying capacity of steel members and their webs depends on the development of the elastic-plastic strains ε in the most loaded cross-sections and their areas.

144

of the plastic hinges needed for development of the total plastic failure mechanism. It is sufficient for one plastic hinge in the most stressed cross-section with full plastic bending moment Mpl , if the maximal plastic strains εmax ≈ 4 εy (εy is the yield strain). Such plastic strains ε it can be assumed and allowed in the cross-sections of class 2, according to used standard classification. For statically undetermined structures, where the loading redistribution is also utilized, the maximal plastic strains should be higher, εmax > 4 εy . The real value of the maximal plastic strains εmax apparently depends on the measure of loading redistribution and on the effect of material hardening.

4

Figure 4. Dependences of the load P and strains εx in the individual measured points of the beam N12.

The investigated distributions of strains εx in the cross-sections subjected to bending or bending and shear with prevailing bending well answer to assumptions of the simple bending theory. From idealized distributions of the strains εx the experimental limit strains εcr,exp and εu,exp were assigned for every relevant tested web and beam. The experimental limit strains εcr,exp and εu,exp has rather large random variable, but they however depend on the web slenderness β. It means that the limit development of plastic strains at steel members can be defined from local stability aspects by relation εu – β, where εu is the maximal plastic strain in the compressed part of their most loaded cross-section. Therefore, according to the obtained experimental results and related theoretical results the empirical relation εu − β was specified (Galambos 1998, Juhás 2001, 2005). The relation εu − β allows the optional web classification of steel members from local stability aspects, in accordance with expected or select development of plastic strains ε in the most stressed cross-sections. The necessary development of plastic strains ε for achievement of full plastic bending moment Mpl in the most stressed cross-section of steel members is not to large. It is markedly outcome of the material hardening that in steel beams especially arises under bending loading. The necessary development of plastic strains ε depends on the static uncertainty of the member or structure and on the equivalent redistribution of the loading, that in some measure depends on the number

ELASTIC-PLASTIC BENDING LOAD-CARRYING CAPACITY

Following experimental and related theoretical results the author of paper worked out the original methodology for calculation of the elastic-plastic bending loadcarrying capacity of the steel members Mep with taking account local stability aspects (Juhás 2001, 2005). This methodology is based on the limit development of plastic strains in the most stressed cross-section. This development is defined by limit strains εu in the edge fibers of compressed web part subjected to bending. The limit strains result from proposed relation εu − β. The author methodology is already applied in the actual standard STN 73 1401:1998. The applied relation is slightly modified to respect the standard limit slenderness of the web β01 , β02 and β1 . The standard relation is shown in Figure 5. The presented methodology and standard relation allow to calculate the elastic-plastic bending moment Mep of the cross-section if the web slenderness β02 < β < β1 . For symmetrical I cross-section Mep = Mpl − Mel,w (εy /εu )2 ,

(13)

where Mel,w is the elastic-plastic bending moment of the web. The previous author linear relation Mep − β has had form (Juhás 1984): Mep = Mpl − (Mpl − Mel )[(β − β02 )/(β1 − β02 )]

145

(14) The similar linear relation is also applied in Specification AISC LRFD:1999. The limit slenderness β0 and β1 are, however, different. The particularity of this standard is that for open I cross-sections takes into consideration also residual stresses, differentially for rolled and welded cross-sections. The final EN 1993-1-1:2005 contains already some procedure for calculation of the elastic-plastic bending

–

–

–

– Figure 5. Relation of the limit strains εu to slenderness β applied in standard STN 73 1401:1998, steel S235.

–

– – –

Figure 6. Effective cross-section of class 2 according to new standard EN 1993-1-1:2005.

moment by the effective cross-section of class 2, which is presented in the Figure 6. It is clear from Figure 6 that assumed effective cross-section by new EN 1993-1-1:2005 is in conflict with assumptions simple bending theory and principal experimental knowledge.

5

CONCLUSIONS

The obtained and partially presented research knowledge and results have allowed following conclusions: – The elastic-plastic load-carrying capacity of the steel members and their cross-sections at large measure depends on local stability of their webs and flanges subjected to bending and compression. – The local stability of the steel member webs and flanges subjected to bending and compression

depends on the real development of the elasticplastic strains ε in the most stressed cross-sections and areas. According to the experimental and theoretical results the relation for limit plastic strains εu and web slenderness β from local stability aspects has been established. The proposed relation εu − β allows the common web classification of steel member cross-sections from local stability aspects in accordance with expected development of plastic strains ε. The proposed methodology enables to calculate the elastic-plastic bending load-carrying capacity of steel members depending on the development of plastic strains ε in the most stressed cross-sections and areas. The proposed methodology does need no classification of the cross-sections for the elastic-plastic and plastic calculation and design of steel members, if their needed rotation capacity or needed plastic strains ε are known or defined. The proposed methodology enables to calculate the elastic-plastic bending load-carrying capacity of the symmetrical and unsymmetrical cross-sections depending on the development of plastic strains εu The proposed methodology unifies the elastic, elastic-plastic and plastic calculation and design of steel members from local stability aspects. The proposed methodology enables also calculation and judgment of the elastic-plastic deflections of steel members and structures. At the end, the proposed methodology enables also calculation and judgment of the elastic-plastic deflections of steel members and structures.

According to present state it appears necessary for next improvement of the elastic-plastic calculation and design of steel members and structures predominantly: – To precise the limit web slenderness β1 for the cross-sections of steel members and structures, with taking account the real production-technological process and resulting material and geometrical imperfections. – To precise the relation εu − β for web slenderness β < β1 of steel members and structures, with taking account the material properties, geometrical dimensions and also interaction and stiffness of the web and flanges. – To work out simplified procedure for identification of the real plastic strains εu in the most stressed cross-sections and areas.

REFERENCES Beedle, L.S. et al. 1991. Stability of Metal Structures, A World View. Bethlehem: Headquarters.

146

Djubek, J., Kodnár, R. & Škaloud, M. 1983. Limit State of the Plate Elements of Steel Structures. Basel: Birkhäuser Verlag. Galambos, T.I. et al. 1998. Guide to Stability Design Criteria for Metal Structures. New York: Structural Stability Research Council, 4Ed. J.W. Juhás, P. 1984. Stability and the elasto-plastic flexural load-carrying capacity of steel structures. Building Research Journal 32(11): 833–863. Juhás, P. 1994. Classification of the cross-sections and elastic-plastic design of steel structures. Stavební obzor 3(3): 88–93.

147

Juhás, P. et al. 2000. Design of Steel Structures–Commentary to STN 73 1401:1998. Bratislava: Slovak Standards Institute (SÚTN). Juhás, P. 2005. Elastic-plastic load-carrying capacity of steel members and local stability aspects. Select Scientific Papers 1: 7–22. Mrázik, A., Škaloud, M. & Tocháˇcek, M. 1986. Plastic Design of Steel Structures. Chichester: Ellis Horvard Limited in co-edition with Prague: SNTL—Publishers of Technical Literature. Salamon, CH.G. & Johnson, J.E. 1995. Steel Structures– Design and Behavior. New Jersey: Prentice Hall.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Local stability and carrying capacity of thin-walled compressed members P. Juhás, M. Al Ali & Z. Kokorud’ová Technical University Košice, Slovakia

ABSTRACT: The paper presents fundamental information about realized experimental research of the local stability and load-carrying capacity of thin-walled compressed steel members with quasi-homogenous and hybrid cross-sections. The load-carrying capacity of such members is influenced by the local web buckling subjected in the elastic-plastic region. The research joins on previous research of the first author. The aim of this research has been oriented on the elastic-plastic post-critical behavior of thin web and its interaction with compact flanges. Some experimental results and their comparisons are presented in the paper, too.

1

INTRODUCTION

Table 1. The total research program and geometrical dimensions of tested members.

The continued effort for economic design of steel structures leads to decrease their weight by shape and material optimization through the using of thin webs, high-strength steels and their effective combination with usual structural steels. The efficiency of high-strength steels using and their combination with usual structural steels is evident in the case of beams mostly subjected to bending loading. But from the complex optimization analyses follow, that the using of high-strength steels and their combination with usual structural steels can be advantage also in the case of columns mostly subjected to compression loading, mainly in the case of thin-walled members. The paper presents basic information about realized experimental research of the elastic-plastic loadcarrying capacity of thin-walled compressed steel members with quasi-homogenous and combined crosssections. This research has been distinctively oriented on the investigation and analyses of local stability and post-critical behavior of the slender and ultraslender member webs and their interaction with compact flanges in the process of their loading and failure (Juhás 1997, 1999, 2006; Juhás et al. 2007, 2008).

2

THE EXPERIMENTAL RESEARCH PROGRAM AND TESTED MEMBERS

The experimental research program included the testing of 24 welded compression steel model members having quasi-homogenous and combined I crosssections with different dimensions, advisable elected to show, in decisive extent, the elastic-plastic postcritical effect of the slender webs and their interaction with flanges in process of their loading, transformation

149

Members

Geometrical dimensions [mm]

M.G

C.G

L

h

b

tf

d

tw

A

1 2 3 4

250 500 750 1000

112 212 312 412

60 90 120 150

6

100 200 300 400

2

B

1 2 3 4

250 500 750 1000

112 212 312 412

60 90 120 150

6

100 200 300 400

2

Figure 1.

Scheme of tested members.

and failure. Table 1 presents the total research program, designed geometrical dimensions and materials of the individual test members and groups. Scheme of the test members is illustrated in Figure 1. The basic geometrical and material characteristics of the individual test members groups are presented in Table 2. All test members are divided into 2 material groups (M.G.: A, B) and 4 cross-sectional groups (C.G.: 1, 2, 3 and 4). The materials group A is created by members with homogenous cross-section made from steel S235 and

Table 2.

specimens were taken from each of the used sheet to make normative shaped test specimens. The test specimens underwent a tension tests to find out the stressstrain diagrams and the actual material characteristics of applied steels. Characteristic stress-strain diagrams are illustrated in Figure 2, where the average values of determined yield stresses fy and limit tensile strength fu are presented. Mentioned yield stresses fy and ultimate tensile strength fu were assign to the relevant flanges and webs of the individual members. For the consistent evaluation and analyses of the experimental knowledge and results, it is also necessary to know the real geometrical dimensions of the test members. Therefore the detailed dimension measuring of the all members was done before test. The dimensions of cross-sections: height h, width b, thicknesses of flanges tf and webs tw was measured on the top, middle and bottom of each member. The average values of measured dimensions are considered as actual. Figure 2 shows the good quality of test members’ steels and required material characteristics. The determined yield stresses of flanges fyf and webs fyw are higher than the normative values. Also in the case of materials group A members with designed homogenous cross-sections, the determined flanges yield stress values were higher than the webs yield stresses, fyf > fyw . It means, that they are also material combined (my = 1.054, event. 1.143). In the case of materials group B it is categorical go about members with material combined cross-sections (my = 1.442, event. 1.563). For all of test specimens a good material ductility was also found, (A5 > 29%).

Geometrical characteristics of tested members.

Members

Geometrical characteristics

M.G

C.G

λy

λz

βf

βw

γ = Aw /A

A

1 2 3 4

5.03 5.01 4.99 4.98

16.02 20.80 23.02 24.28

4.83 7.33 9.83 12.33

50.0 100.0 150.0 200.0

0.217 0.270 0.294 0.308

B

1 2 3 4

5.03 5.01 4.99 4.98

16.02 20.80 23.02 24.28

4.83 7.33 9.83 12.33

50.0 100.0 150.0 200.0

0.217 0.270 0.294 0.308

Figure 2. Characteristic stress-strain diagrams and determined material characteristics.

3

group B that is created by members with combined cross-section made from steel S355 (flanges) and S235 (webs). The individual cross-sectional groups have different dimensions, but first of all they have different web slenderness βw . It is apparently, that the members are thin-walled at the compression loading. At the same time, according to local stability aspects, the flange dimensions are designed to be compact (slenderness βf ), when subjected to the elastic region of loading. At last, according to the global stability aspects and dimensions of the individual cross-sectional groups, the lengths of members L are designed to be quasi-compact (slenderness λy , λz ). The ratios γ give an evident characteristic of the economic efficiency of designed cross-sections. The flanges of all members were made out from 2 sheets, 6 mm thick (steel S235 and S355) and the webs from 2 sheets, 2 mm thick (steel S235). Three material

METHODOLOGY AND TESTS CONTENT

The tests of members were realized at the Bearing structures laboratory of the TU Košice—CEF by

Figure 3. General layout of the test, measurement of strains ε, deflections of the web w and buckling of the member v.

150

hydraulic loading machine. The general layout of the tests is illustrated in Figure 3. The tests had to bring out detailed investigation about transformation, failure and ultimate loadcarrying capacity of the test members, in consider of their geometrical and material parameters. In accordance with research target, the emphasis has been imposed on the elastic-plastic post-critical behavior of the slender webs and their interaction with flanges. In context with that, the initial shape deflections of members, mainly the initial buckling of slender webs are significant for the experimental results valuation and connected theoretical analyses. Therefore, the initial buckling of all the member webs on previously drawn raster by means of inductive sensors were finding out before testing start. During consecutive programmed overloading of test members, the strains ε in the middle cross-section were measured. Measurement was realized in 12 places, double-faced on the web in 6 places and also on the flanges in 6 places. The resistance tensiometers were used to measurement the strains ε by means of measuring apparatus Hottinger Balwin UPM 60 connected to computer for direct evaluation. According to member’s length, the deflections of web w were measured in 3 or more places elected in the characteristic positions. The web deflections w were measured using inductive sensors connected to

computer and also using mechanic gauges. In the case of members with ultra-slender webs (AS41∼AS43 and BS41∼BS43), the global buckling was also investigated in the middle cross-section on the edges of flanges. The member’s global buckling v was measured by means of mechanic gauges. The measurement of strains ε, lateral deflections of the web w and global buckling v of the members are illustrated in Figure 3. The members during the test were consecutively overloaded and released. The member overloading was regulated close to its behavior, measured values of strains ε and deflections of the web w. The test continued up to total failure, defined by the beginning of consecutive, continuous increasing of strains ε and deflections of the web w.

4

EXPERIMENTAL RESULTS

According to the mentioned methodology the experimental tests of all members were done. Table 3 explains the overall behavior of all tested members and also presents the experimental ultimate limit loads. In dependence on the sequential load increasing the deflections w and the strains ε of the tested members were recorded. Figure 4 illustrates the total failure of members AS22 and AS23 by local buckling of their

Table 3. Test results—transformation, failure and experimental ultimate loads of tested members. Transformation and failure

Member

N u,exp [kN]

 Original state: The initial web buckling in different amplitudes and shapes, depending on the web slenderness β w .  Transformation: The primary local web buckling and consecutive local buckling and warping of the flanges, in dependence of web slenderness β w and its initial buckling shapes. The consecutive changes and formation of new web’s buckling shapes. Consecutive impression of frontal (ending) plates into webs at the ends of members.  Failure: The webs and flanges are buckled along the members, buckling of the webs into multiple configurations of half-waves, different lengths of the several web’s half-waves buckling, reduction of web’s half-wave buckling lengths at the ending parts members. The local buckling and warping of the flanges, depending on web’s buckling. The impression of frontal (ending) plates into webs at the ends of members. The distinct impression of flanges into member’s cross-section was not manifested, neither at the most slender webs (β w,max = 200). The total buckling was not manifested, neither at the most slender members (λz,max = 24.28).

AS11 AS12 AS13 BS11 BS12 BS13 AS21 AS22 AS23 BS21 BS22 BS23 AS31 AS32 AS33 BS31 BS32 BS33 AS41 AS42 AS43 BS41 BS42 BS43

278.0 275.0 280.0 357.0 363.0 357.0 357.0 373.0 359.0 466.0 492.0 482.0 447.0 432.0 442.0 565.0 577.5 562.5 490.0 512.5 480.0 657.5 590.0 625.5

151

Figure 4.

20,0 15,0

The total failure of members AS22 and AS23.

Figure 7. Initial and final topographical form of web’s buckling shapes of member AS31.

w [mm]

10,0 5,0 L [mm]

0,0 -5,0 -10,0 -15,0

1000

-20,0

Figure 5. Initial and final webs buckling shapes along member AS41. —— initial shape, final shape.

20,0 15,0

w [mm]

10,0 5,0 L [mm]

0,0 -5,0 -10,0 -15,0

1000

-20,0

Figure 6. Initial and final webs buckling shapes along member BS41. —— initial shape, final shape.

Figure 8. BS23.

webs and flanges. The initial and final webs buckling shapes along of some selected tested members are illustrated in Figures 5 and 6. Figure 7 illustrates the topographical form of the initial and final web’s buckling shapes of member AS31. Figures 8 and 9 present the relations N –w and N –ε of the members BS23 and BS22.

The experimental knowledge and results presented in Table 3 and Figures 5, 6, 7, 8 and 9 mention very clearly on complicacy and variable elastic-plastic behavior of the individual test members during loading. But the experimental limit loads of the individual test members within the geometrical and material groups are relatively equivalent.

152

Middle and quarters web deflections, member

Table 4.

Figure 9.

5

Web and flanges strains ε of member BS22.

THEORETICAL LIMIT LOADS AND THEIR COMPARISON

According to the calculation procedures and formulae of the first author the theoretical limit loads of all test members were determined as following: Nel Npl Nul,el Nul,ep

Nu,y , Nu,z

Member

Nel

Np l

Nul,el

Nul,ep

Nu,z

Nu,y

AS11 AS12 AS13 AS21 AS22 AS23 AS31 AS32 AS33 AS41 AS42 AS43 BS11 BS12 BS13 BS21 BS22 BS23 BS31 BS32 BS33 BS41 BS42 BS43

249.9 251.3 249.9 377.6 377.1 376.0 540.2 542.8 549.7 651.5 653.9 654.6 253.6 261.6 257.0 376.4 376.0 370.1 544.6 549.0 546.1 643.0 654.7 650.9

260.2 261.8 260.3 416.7 415.6 414.4 560.3 563.0 570.0 713.8 716.6 717.4 339.8 350.6 345.1 527.8 527.3 517.9 711.3 716.6 713.5 888.2 902.1 896.2

249.9 251.3 249.9 322.8 321.9 320.9 424.5 426.8 432.2 491.0 493.9 494.6 253.6 261.6 257.0 321.2 320.9 314.9 429.2 432.6 430.7 486.7 494.4 490.5

260.2 261.8 260.3 360.9 359.2 358.2 444.8 447.2 452.7 552.0 555.4 556.2 339.8 350.6 345.1 469.2 468.7 459.2 592.8 597.2 595.0 691.8 709.2 695.7

260.2 261.8 260.3 356.1 354.3 353.3 435.6 438.1 443.4 539.1 542.0 542.8 335.5 346.5 341.0 454.9 454.7 445.7 569.9 574.3 572.2 659.5 676.1 663.0

260.2 261.8 260.3 360.9 359.2 358.2 444.8 447.2 452.7 552.0 555.4 556.2 339.8 350.6 345.1 469.2 468.7 459.2 592.8 597.2 595.0 691.8 709.2 695.7

Table 5.

limit elastic load of member definite by attaining the web yield stress fyw , limit plastic load of member definite by attaining the flanges yield stress fyf , limit elastic post-critical load of member definite by attaining the yield stress fyw in the outer fibers of cross-sectional web, limit elastic-plastic post-critical load of member definite by attaining the ultimate strain εu = εyf in the outer fibers of crosssectional web, limit buckling load of member according to axis y and z considering the elasticplastic post-critical behavior of the web.

Table 4 and 5 present the relevant theoretical values of the individual members’ limit loads and their comparison. All limit loads were calculated according to the real—measured dimensions and determined yield stresses of their flanges and webs. The ratios of elastic-plastic post-critical limit loads and the full plastic limit loads Nul,ep /Npl indicate at the negative effect of the web buckling on global loadcarrying capacity of the tested members, with increasing of the web slenderness β the ratios decrease. The ratios of elastic-plastic post-critical limit loads and the elastic post-critical limit loads Nul,ep /Nul,el

153

Calculated theoretical limit loads [kN].

Theoretical limit loads comparison.

Member

Nul,ep /Npl

Nul,ep /Nul,el

Nul,ep /Nu,z

AS11 AS12 AS13 AS21 AS22 AS23 AS31 AS32 AS33 AS41 AS42 AS43 BS11 BS12 BS13 BS21 BS22 BS23 BS31 BS32 BS33 BS41 BS42 BS43

1.000

1.042

1.000

0.865

1.117

1.014

0.794

1.048

1.021

0.775

1.124

1.024

1.000

1.341

1.012

0.888

1.460

1.031

0.834

1.381

1.040

0.780

1.425

1.049

indicate at the positive effect of the web plasticity behavior on local load-carrying capacity of the tested members, with increasing of the flange yield stress the ratios increase. The ratios of elastic-plastic post-critical limit loads and the global buckling loads Nul,ep /Nu,z indicate at the certain small possibility of the lateral torsional buckling of the tested members during loading, but it depends on their real support conditions. The average ratio values of calculated and compared theoretical limit loads for individual member cross-sectional groups are presented in Table 5. The maximum value of the ratio Nul,ep /Nu,z is only 1.049 (members BS41, BS42 and BS43).

6

COMPARISON OF THEORETICAL AND EXPERIMENTAL LIMIT LOADS

The numerical comparison of the crucial theoretical limit loads Nul,ep , Nu,z and investigated experimental limit loads Nu,exp is done in Table 6. The graphic evaluation and comparison of the theoretical limit loads Nu,z and experimental limit loads Nu,exp for the individual tested members and member groups is presented in Figures 10 and 11. In general, the buckling limit loads of the tested members Nu,z have the smallest values. However,

Figure 10. Theoretical and experimental capacities comparison, material group A.

Table 6. Theoretical and experimental limit loads comparison. Member

Nu,exp /Nul,ep

AS11 AS12 AS13 AS21 AS22 AS23 AS31 AS32 AS33 AS41 AS42 AS43 BS11 BS12 BS13 BS21 BS22 BS23 BS31 BS32 BS33 BS41 BS42 BS43

1.068 1.050 1.076 0.989 1.038 1.002 1.005 0.966 0.976 0.888 0.923 0.863 1.051 1.035 1.034 0.993 1.050 1.050 0.953 0.967 0.945 0.950 0.832 0.898

Nu,exp /Nu,z 1.065

1.010

0.982

0.891

1.040

1.031

0.955

0.894

1.068 1.050 1.076 1.003 1.053 1.016 1.026 0.986 0.997 0.909 0.946 0.884 1.064 1.048 1.047 1.024 1.082 1.081 0.991 1.006 0.983 0.997 0.873 0.943

1.065

1.024

1.003

Figure 11. Theoretical and experimental capacities comparison, material group B.

0.913

1.053

1.063

0.993

0.937

these limit loads are very close to the elastic-plastic post-critical limit loads Nul,ep . But when the real boundary conditions of members are considered in accordance with arrangement of the loading machine, the elastic-plastic post-critical behavior and interaction between thin webs and flanges may appear in conclusive rate. All of the tested members were failure by means of the local failure of flanges in consequence of webs’ local deflection in multiple half-waves with different shapes. The conclusive buckling of web and flanges was mainly concentrated in the ending areas of members—obviously because of concentrated loading transfer.

154

Very good consonance can by found from the comparison of determined experimental limit loads Nu,exp and the theoretical limit loads Nul,ep or Nu,z , if the web slenderness of tested members β w ≤ 150. However, more significant differences was registered between evaluated theoretical and experimental limit loads, if the web slenderness β w = 200. The results mention very clear on the effect of the local buckling and interaction of the member web and flanges subjected to compression. This effect is very significant first of all in the places of the direct transmission of the loading to compressed members. 7

CONCLUSIONS

The obtained and partially presented research knowledge and results have allowed following conclusions: – The results of presented research affirm and expand the knowledge of the previous research about the elastic-plastic behavior and load-carrying capacity of the thin-walled compression members with quasi-homogenous and combined cross-sections. – Elastic-plastic post-critical load-carrying capacity of compression members depends in a large scale on the initial web buckling shapes and their consecutive formation and changes during the loading process. – Post-critical load-carrying capacity of the thinwalled compression members increases by increasing the number of buckling waves during the elasticplastic and plastic stage of the loading. – Thin webs with slenderness βw ≤ 150 (A), event. βw ≤ 100 (B) prove a sufficient support of compression members’ flanges. Theoretical limit loads Nul,ep and Nu,z are in a good consonance with the obtained experimental limit loads Nu,exp . – In the case of members with webs’ slenderness β = 200, the influence of non-sufficient support of compression flanges by ultra thin web was also manifested here. This effect was significant near the members’ ending which can be caused by local transfer of loading to the flanges and web. In the case of these members, theoretical local load-carrying capacity Nul,ep and global buckling

155

load-carrying capacity Nu,z are a bit less than elasticplastic load-carrying capacity Nu,exp determined by experiments. – New obtained knowledge and results encourage to more consistent analysis of very slender webs influence and in this context encourage to adequate reduction of the local and global elastic-plastic loadcarrying capacity of the thin-walled compression members. ACKNOWLEDGEMENTS The presented research and preparation of the paper has been partially founded by the Slovak Ministry of Education and Slovak Academy of Sciences—project VEGA 1/4220/07. The financial support is gratefully acknowledged. REFERENCES Juhás, P. 1997, Thin-walled hybrid compressed elements. Publications of the University of Miskolc, Series C: Mechanical Engineering. Vol. 47: 123–129. Juhás, P. 1999, Load-carrying capacity of hybrid compressed steel elements. Proceedings of the 2nd European Conference of Steel Structures. Vol. 2: 697–700. Juhás, P. 2006, Buckling load-carrying capacity of steel hybrid thin-walled compressed members. Selected Scientific Papers—Journal of Civil Engineering. Vol. 1: 7–27. Juhás, P., Kokorud’ová, Z., Al Ali, M. 2007, Investigation of elastic-plastic load-carrying capacity of thin-walled compressed steel elements. Proceedings of the 8th Scientific Conference of the TU - Civil Engineering Faculty Košice. Civil Engineering Faculty, Košice: 99–106. Juhás, P., Al Ali, M., Kokorud’ová, Z. 2007, Experimental investigation of elastic-plastic carrying capacity of thin-walled compressed steel members. Proceedings of Czech—Slovak Conference Experiment’07. AP CERM, Ltd., Brno, Czech Republic: 139–146. Juhás, P., Al Ali, M., Kokorud’ová, Z. 2008, The elasticplastic load-carrying capacity of thin-walled steel members with quasi-homogenous and combined cross-sections. Selected Scient. Papers—Journal of Civil Engineering. Vol. 3: 7–18.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Finite element analysis of wind induced buckling of steel tank S. Borgersen SEBCorp, Eagan, MN, USA

S. Yazdani Department of Civil Engineering, North Dakota State University, Fargo, ND, USA

ABSTRACT: During construction, large diameter steel plate storage tanks could experience severe damage and buckling under environmental loading conditions, such as gusting wind loads, if inadequate bracing is provided. One such case is presented in this paper involving a 2,760 cubic-meter steel plate tank that experienced localized buckling under a severe straight line wind pressure with a velocity exceeding 100 km per hour. This paper discusses the failure mode, simulation of the failure using FEA with wall material non-linearity, and recommendations for design and construction methods which could be used to prevent similar occurrences on other tank erection projects. 1

INTRODUCTION

Construction of a 2,760 cubic-meter ethanol storage tank was nearly complete, requiring only the installation of a conical roof structure. The completed tank (Figure 1) was to be a welded plate cylindrical shell structure 15.24 m diameter and 15.24 m wall height, constructed of rolled stainless steel plates 1.22 m × 2.44 m sections with plate thickness varying with wall elevation, as shown. The tank wall was of welded plate construction, and supported on a reinforced concrete ring wall. Stainless steel threaded anchor rods, 32 mm diameter, were used to attach the tank to the reinforced concrete ring wall foundation by use of plate gussets. The gussets were welded to the tank wall at a circumferential spacing of 305 mm around the tank base perimeter. The interior tank floor was a 0.61 m inverted conical surface, constructed with parallel strips of 1.22 m ×

Figure 1.

Schematic representation of the tank.

2.44 m × 8 mm stainless steel plates, overlapping 25 mm and seam welded using a 5 mm fillet weld. A pipe drain and reinforcing pad was located at the center or apex of the conical floor. The tank floor extended approximately 38 mm past the base of the tank wall. The bottom wall course consisting of 38 mm wall plate was continuously welded to the 8 mm plate floor along its outer perimeter. A 2.7 m high conical plate roof, constructed of 5 mm welded stainless steel plate segments had been assembled at ground level, next to the tank wall and foundation. The roof structure was reinforced by eighteen, radially oriented, W10 × 17 beam members, on a 20 degrees circumferential spacing. The outer edge of the roof assembly included a circumferential 13 mm × 152 mm × 203 mm compression angle, with the 8 leg extended outward in the radial direction. The entire roof assembly, including access hatches, vents, peripheral equipment, and perimeter safety railing, was assembled on the ground next to the erected tank wall. The roof assembly was scheduled to be lifted into position and attached to the top of the tank wall, using a 9 kN capacity crane. As the crane operator prepared to initiate the installation, a weather warning, citing potentially very strong, straight-line winds, caused the installation process to be halted. The roof structure and crane boom were then securely anchored, as a safety precaution. Shortly thereafter, a series of severe, straight-line wind gusts struck the construction site. Unfortunately, the straight line wind gusts were localized, and peak gust magnitudes and durations were not recorded by the local airport weather station. However, the wind gust direction was directly perpendicular to the tank

157

Figure 2.

Schematic representation of the damage zone.

wall, as evidenced by the location of the damaged wall section and the presence of nearby surrounding structures. Security cameras allowed plant personnel to witness the event. Their statements indicated the wind gusts were so severe that the tank wall underwent large amplitude, dynamic deformations during the storm. The wind loading continued to affect the tank structure, finally causing the cylindrical, open-end tank wall to buckle along an elliptically shaped failure surface shown in Figure 2. Examination of the damaged tank wall structure determined that the damage consisted of large, plastic deformation along the elliptical wall segment covering approximately 30◦ around the top circumference and extending approximately 7-1/2 courses down from the top of the tank wall. The plates were so severely buckled that removal and replacement of several plates were required. The remaining plates were corrected in-place, using acceptable construction techniques. However, removal and replacement of the damaged wall plate segments created a situation where the new welds had to be re-inspected using radiographic methods. Also, hydrostatic leak testing of the tank shell had to be reconducted to certify the structure after the repairs were completed. Once the roof structure had been installed, a second series of hydrostatic pressure tests were required to ensure all safety valves, manhole access hatches and welds were sufficient to meet API code (American Petroleum Institute) requirements. Added costs associated with the delay included: added material costs, radiological inspection costs, labor, cost of water required for the tests and delays due to required code verification testing. However the largest cost was downtime and loss of plant production capability, due to inability to utilize the storage tank as scheduled.

2

THEORY

Shell buckling is a complex, stability based event, which takes into account geometry, loading and nonlinear material properties. There appears to be no

directly applicable closed form solutions for the large, plastic deformation of cylindrical thin wall shells open and unsupported at one end and cantilevered at the other. A forensic replication of the wind induced failure of the tank wall would therefore require the use of a finite element analysis (FEA) modeling technique using software that would be able to account for the applicable parameters, including distributions of pressure loading applied over portions of the structure, incrementally varying loading amplitudes, and non-linear material characteristics. In order to meet these analysis requirements, the MARC software provided by MSCSoftware Corporation was used. The MARC software is one of a family of technical analysis packages available from MSC Software Corporation, including: Patran, Nastran, Adams, Dytran, and others.

3

PRESSURE DISTRIBUTIONS

Evaluation of AISC, API and IBC Codes with their assumed design wind velocities, shape coefficients and associated design pressures were found to be not applicable to this type of construction geometry: a cylindrical shell structure cantilevered at the base and open at the top. The codes are design guidelines intended for fully enclosed structures, or structures with openings in one or more vertical surfaces. It was therefore decided that initial wind pressure distributions on the structure type being investigated could best be evaluated by using the computational fluid dynamics (CFD) capabilities in the MARC software. Preliminary wind pressure distributions were derived using 2-dimensional CFD models: the first using a two-dimensional plan view (Figure 3), representing a rigid circular structure to evaluate initial pressure distributions around the circumference of the tank wall perimeter; the second represented a twodimensional, vertical elevation cut through the center of the tank. Results obtained from the two-dimensional plan view wind model provided locations of peak positive, zero and negative wind pressure distributions around the tank perimeter. The two dimensional vertical elevation cut provided additional information with respect to vertical pressure distributions up the tank wall, as well as the effects of turbulence on pressure at the open ends of the tank, upwind, downwind and internal to the open end. The results obtained from these two CFD studies were then implemented as pressure loadings in the FEA model, and correlated with ASCE, API and IBC code design wind pressure recommendations for closed ended cylindrical structures of this geometric type.

158

Figure 3.

4

Study Case 4: The stepped wall tank model used in CASE 3, but modified with strut stiffeners tying the top of the center column to the top of the tank wall at 22.5◦ around the circumference. Study Case 5: The stepped wall tank model used in Study Case 2, but modified with a continuous 13 mm × 152 mm × 203 mm stiffener angle added to the top edge of the tank with the 203 mm leg being horizontal. The FEA model used for this investigation incorporated 12,240 nodes and 12,000 elements. The elements were a six-degree-of-freedom per node, type 75, thin shells. Analysis studies incorporating strut stiffeners and a cantilevered center post used additional nodes and element types, as needed. The axial only members used a three-degree-of-freedom per node, type 9 truss elements, and a three degree-of-freedom per node, type 7, solid element was used for the center post.

Wind pressure distribution on the tank.

MATERIAL PROPERTIES

The material used in the model was assumed to be stainless steel, isotropic, elastic-plastic with Young’s Modulus E = 200 GPa; Poisson’s ratio λ = 0.3; yield stress Fty = 206 MPa; ultimate stress Ftu = 310 MPa. A non-linear stress strain model was used for the stainless steel material.

7

MODEL CORRELATIONS

To simulate the over all wind effect, an incrementally increasing static loading condition using the wind pressure load was applied to the tank. The incremental loading was linearly increased until a stress level in the tank wall was developed which provided an assumed equivalent von Mises yield stress of 206 MPa, matching that of the observed damage on the tank wall.

The incremental application of wind pressure on the external surfaces of the tank wall correlating with the CFD analysis resulted in the representative deformation of the wall structure summarized in Table 1. The analysis shows that the wind pressure forms an elliptical yield surface extending from the top of the cylindrical shell towards the lower, thicker course of the tank wall. The FEA model predicts that the primary yield surface extends down to approximately 7-1/2 plate courses, approximately 8 m to 9 m, from the top of the tank wall. The plan view of the deformed surface also indicates that the maximum deformed shape of the wall structure occurs around the circumference of the tank wall.

6

8

5

APPLIED LOADS

ANALYSIS MODELS

Five FEA model studies were evaluated to determine the effects of various stiffening options which might have been used to prevent this type of failure during construction: Study Case 1: A constant thickness tank wall model using 13 mm plate thickness fixed at the base and cantilevered and open at the top edge was developed as a baseline for comparison purposes. Study Case 2: A stepped wall tank model reflecting actual plate wall thickness at 2.4 m course heights, as shown in Figure 1. Study Case 3: The stepped wall tank model used in CASE 2, but modified with a full height 457 mm diameter pipe column cantilevered at the base, and with strut stiffeners tying the top of the center column to the top of the tank wall at 45◦ around the circumference.

159

ANALYSIS RESULTS

FEA analysis studies evaluated three potential damage prevention options that may have prevented the tank wall buckling, had they been incorporated prior to installing the roof structure. Study Cases 1) & 2) are baseline studies using the constant 13 mm thickness tank wall and the actual, as constructed, stepped thickness tank wall which failed due to wind load. This produced a maximum calculated radial displacement Equivalent Stress of 2.1 meter, 214 MPa and 1.48 meter and 201 MPa respectively. Study Case 3) provided radial stiffening of the tank top perimeter wall, with the axial stiffeners located at 45◦ around the perimeter. This produced a maximum inward radial deflection of the wall with magnitude of 1.2 m, approximately 280 mm less than the unsupported tank wall, but still produced a yield surface in the tank wall with a predicted stress level of 208 MPa. Study Case 4)

consisting of tension rods or cables at the 1/8 points along the open end of the tank wall perimeter, or ii. Installing a circumferential stiffener flange of appropriate size on the perimeter of the tank wall open end.

provided radial stiffening of the tank top perimeter wall, with the axial stiffeners located at 22.5◦ around the perimeter. This produced a maximum 0.132 meter inward radial deflection of the wall and a peak wall plate stress of 89 MPa, providing a Factor of Safety of over 2.0 against yielding. Study Case 5) incorporated a circumferential stiffener angle installed at the top edge of the tank wall. This produced a maximum radial deflection of 0.460 meter and a maximum 97 MPa stress level in the tank wall plates with a Factor of Safety = 2.14 preventing yield and permanent damage to the tank wall. 9

6. Based on estimated construction costs for labor and materials, the installation of a center column support for radial stiffeners, and placement of radial stiffeners, would be a less cost effective construction procedure than installing a circumferential stiffener flange along the perimeter of the open top edge of the tank wall.

CONCLUSIONS

The FEA model representing the stepped thickness wall results correlated well with the actual damage resulting in the tank wall structure due to high wind gusts. Based on these study results, a number of conclusions are presented. 1. Wind pressure distributions specified for cylindrical, large volume tanks based on API, ASCE and IBC codes are not adequate for field construction conditions. 2. The codes assumptions are based on closed ended, rigid structure and small deformations. 3. Additional turbulence and higher wind pressures generated by the open end cylindrical plate structures subjected to severe wind gust conditions are not covered by the design codes. 4. Peak gust wind pressures on an open ended cylindrical structure, represented by the tank wall in this study case, may be 2 to 4 times higher than values predicted by the codes, due to turbulence effects at the open end. 5. Failure of the structure—i.e. tank wall snap-thru buckling, could have been prevented by: i. Installing a circumferential stiffening angle around the top perimeter of the tank wall consisting of either temporary construction bracing

REFERENCES ASCE/SEI 7-05, 2007. Minimum Design Loads for Buildings and Other Structures, American Society of Civil Engineers, Structural Engineering Institute. Argon, Ali S., 1975. Constitutive Equations in Plasticity, MIT Press. Baker, E.H., Kovalevsky, L., Rish, F.L. 1972. Structural Analysis of Shells. McGraw Hill Book Company. Calladine, C.R.1985. Plasticity for Engineers, John Wiley & Sons. Chen, W.F. & Saleeb, A.F., 1982. Constitutive Equations for Engineering Materials, Vol 1: Elasticity and Modeling, John Wiley & Sons Hodge, Jr. P.G., 1959. Plastic Analysis of Structures, McGraw Hill Book Company. International Building Code 2008. International Code Council, Inc. (ICC). Kollar, L. & Dulacska, E. 1984. Buckling of Shells for Engineers, John Wiley & Sons, Inc. Kraus, H. 1967. Thin Elastic Shells—An Introduction to the Theoretical Foundations and the Analysis of Their Static and Dynamic Behavior. John Wiley and Sons, Inc. MSC-MARC® 2007 r3, Volume A:Theory and User Information, MSC-Software Corporation. William, P. & Hodge Jr, P.G. 1968. Theory of Perfectly Plastic Solids, Dover Publications.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Performance based analysis of RBS steel frames P. Alexa & I. Ladar Technical University of Cluj – Napoca, Romania

ABSTRACT: Reduced beam section technique is an efficient technique in the larger objective of imposing a controlled behaviour of buildings under seismic actions. Several numerical comparative studies are conducted on six steel multi-storey planar frames equipped with reduced beam sections (dog bone) with the aim of selecting optimum design solutions for multi-storey real steel structures. The structures selected for the present study are part of an interactive design job. The first component of this job is to convince the beneficiary to agree to RBS technique. The second component of the job is to obtain real steel structures placed somewhere in the ‘‘immediate occupancy’’ performance level. Two cases of reductions in the beams’ sections (RBS)—of 20% and 30% reduction in the cross section area of beams—for two planar frames are considered. Also, the reference case of classical non reduced beam section for each frame is considered.

1

INTRODUCTION

The lessons delivered by the Northridge 1994 earthquake and the damages induced into multi-storey steel structures, mainly with welded beam—column connections emphasized the important role of ductile connections beam—column for both, the reduction of the level of damages (cracks in the welds) and the possibility of repairing of cracked welded connections. Among the important lessons that have been learned, is that the structure has to be equipped with several levels of ductility (structural level, member level, joint level). The need of ductility and the beneficial role it plays during an earthquake, raised the ductility requirement of seismically acted structures at the same level as the strength and serviceability requirements. Today ductility is, in structural design, a top priority the structure has to be ensured with and provided for. Reduced beam section connectivity of beam—column joints of steel frame type structures provides for a nonlinear bending moment—rotation relationship of different type of the classical beam—column connectivity. RBS frames force the formation of plastic hinges in locations generally accepted by structural design codes. If the rules of computing the reinforcement of the beams in the vicinity of the beam—column connections of R/C frames are brought to attention, then inducing the formation of plastic hinge into beams near the beam—column connection, in the case of steel frames, appears to be almost the same technique: the plastic hinge is going to form into the beam. The semirigidity may be viewed as aiming to the same result, though it has been adopted from totally another necessity. From semirigidity procedure to reduced beam section technique in order to

161

induce plasticization into beam end sections it has only been a natural step (Engelhardt & Winneberger 1995). The present contribution intends to introduce several results of numerical studies (statically nonlinear analyses of push-over type and time history type) specific to performance based analysis such as: base shear—lateral top displacement curves, ductility

IPE 180

IPE 180

IPE 180

IPE 180

IPE 220

IPE 220

IPE 220

IPE 220

IPE 220

IPE 220

IPE 220

IPE 220

HEA 180 HEA 180 HEA 180 HEA 200 HEA 200

HEA 200

Figure 1.

Six story two bay frame.

coefficients, story drifts, elastic-plastic collapse mechanisms. The analyses have been carried out on six skeletal steel planar frames. Two steel multi-storey frames (Figs. 1, 2) are considered in two situations of Reduced Beam Section (20% and 30% reduction in the area of beams). The case of not reduced beam section is, also, considered as the reference case/frame. The reduced beam sections are located into the vicinity of beam—column connections and are applied to all beams. The location of the buildings is in the Southern Romania, therefore the analyzed frames are subjected to a time history (Romania, Vrancea—1977 earthquake—a reference earthquake for this country) loading and, also to a pushover type both in the presence of corresponding gravitational loads. The gravitational loads are associated to a residential serviceability as it is the beneficiary of the design request. Relevant—for the authors’ objective—conclusions are drawn for the Reduced Beam Section technique and for the (seismic) performances they fulfil.

IPE 180

IPE 180

IPE 180

IPE 180

IPE 180

IPE 180

IPE 220

IPE 220

IPE 220

IPE 220

IPE 220

IPE 220

HEA 180 HEA 180

The loads are of gravitational and lateral (modal push-over type loads and time history type). The time history type analysis is conducted using the Vrancea (Romania) 1977 earthquake accelerogram (Fig. 3). Each Reduced Beam Section connection—as a potential plastic hinge—requires its own bending moment— relative rotation curve. The ductility levels are expressed through, already, classical numerical procedure: ductility coefficients computed as ratios between the values of kinematic parameters (displacements, rotations) associated with the initiating of plasticization of cross section and (last) hinge formation leading to structural collapse (Gioncu & Mazzolani 1995). 2 2.1

ANALYZED STRUCTURES General geometry, sections, loadings

The general geometry, the cross sections of the elements (beams, columns) and (gravity) loadings of the six storey two bay analyzed planar steel frame (Fig. 1) and of the six storey three bay (Fig. 2) are given below. The Vrancea (Romania) accelerogram is well known in this country (Vrancea 1977 earthquake became a reference earthquake, both emotionally and for structural design) and has been selected—in spite of not being as famous as other earthquakes that made history—at the request of the beneficiary of the (analyzed) structures.

HEA 180 HEA 200 IPE 220

IPE 220

IPE 220

IPE 220

IPE 220

IPE 220

HEA 200

HEA 200

Figure 2.

Six storey three bay frame.

Figure 3.

Vrancea, March 1977 earthquake accelerogram.

3

NUMERICAL RESULTS

The carried out analyses take into account the real state of beam—column connectivity with reference to the bending moment diagram in the vicinity of the beam end section (Fig. 4). The real value of the bending moment is associated to the RBS axis—not to the column geometrical axis as shown in Figure 4 (Alexa et al. 2006). The distance between the column geometrical axis and the RBS axis has been taken 30 cm. Regarding the pushover analysis, the target displacement is computed implicitly and is about 4.8% of the total height of the frame.

162

As it has been mentioned, each RBS is associated to a bending moment—rotation diagram corresponding to the material (steel) specifications. The numerical results for push over curves of base shear versus lateral top displacement (Chopra & Goel 2001) are presented in Figures 5, 6. The (non dimensional) values of rotation ductility coefficient η is associated to the joint located at top level right corner of each of the two frames and their values are presented in Table 1. The values of global ductility coefficients (associated to top lateral displacements) are presented in Table 2. Table 1. Figure 4. axis.

Rotation joint ductility coefficient.

Association of bending moment diagram to RBS Frame

Joint ductility level η

Six story two bay reference frame Six story two bay 20% RBS frame Six story two bay 30% RBS frame Six story three bay reference frame Six story three bay 20% RBS frame Six story three bay 30% RBS frame

12.6 14.9 15.5 10.3 11.1 11.7

Table 2.

Figure 5.

Pushover curves of six storey.

Figure 6.

Pushover curves of six storey three bay frames.

163

Global ductility coefficient.

Frame

Joint ductility level η

Six story two bay reference frame Six story two bay 20% RBS frame Six story two bay 30% RBS frame Six story three bay reference frame Six story three bay 20% RBS frame Six story three bay 30% RBS frame

12.6 14.9 15.5 10.3 11.1 11.7

(26)

(25) (27)

(24)

(22)

(18) (23)

(17)

(15)

(14) (16)

(13)

(25)

(24)

(19)

(18) (20)

(17)

(15)

(14) (16)

(13)

(11)

(8)

(12)

(6)

(11)

(8)

(12)

(3)

(7)

(4)

(9)

(2)

(7)

(5)

(9)

(2)

(5)

(3)

(10)

(1)

(6)

(4)

(10)

(1)

(21)

(19)

(23)

(20)

a) Reference frame Figure 7.

(21)

(19)

(18) (20)

(17)

(15)

(14) (16)

(13)

(11)

(8)

(12)

(3)

(7)

(5)

(9)

(2)

(6)

(4)

(10)

(1)

c) 30% RBS

Collapse mechanisms of six story two bay frames. History and locations of plastic hinges.

(31)

(33)

(29) (28)

(22) (27)

(25)

(23)

(21) (24)

(20) (26)

(19)

(13)

(17)

(18) (16)

(14) (15)

(7)

(9)

(6)

(13)

(4)

(11)

(2)

(8)

(5)

(12)

(3)

(10)

(1)

(32)

(29)

(30)

(31) (26)

(27) (25)

(28)

(30)

(19)

(22)

(18) (24)

(16) (20)

(14)

(14) (15)

(7)

(21)

(17) (23)

(15) (19)

(4)

(11)

(2)

(12)

(6)

(3)

(10)

(1)

(11)

(5)

(30)

(29) (28)

(22) (27)

(25)

(23)

(21) (24)

(20) (26)

(17)

(18) (16)

(9)

(6)

(13)

(8)

(5)

(12)

(10)

(4)

(8)

(2)

(9)

(3)

(7)

(1)

(34)

(33)

(36)

a) Reference frame

(35)

(34)

(38)

b) 20% RBS frame

(36)

c) 30% RBS frame

Collapse mechanisms of six story three bay frames.

a) two bay frames

b) three bay frames

Story drifts in percentages.

164

(34)

(37)

(39) (35)

(35) (32)

(33)

(31)

Figure 9.

(21)

(22)

b) 20% RBS

(32)

Figure 8.

(22)

a) Reference frame

Figure 10.

b) 20% RBS frame

c) 30% RBS frame

Base shear—lateral top displacement hysteretic curves of three bay frames.

The analyzed frames have been designed according to Romanian current seismic code (P100/1-2006) and having a location associated to a ground acceleration value ag = 0.18 g. The selected frames are transverse components of a steel structure of six storey office building. The total (live + dead loads) gravity loading is associated to a design value of about 1000,00 daN/m2 . Reduction in beams’ section refers to the area of beams cross section. The selected numerical results are a part of a larger design job carried out for an investment company. The RBS technique has been applied according to the current specifications (NEHRP 1997; NEHRP 2003). Also, the provisions of both, FEMA 273 and FEMA 450 are observed regarding target displacement. The quantity and the geometry of reduction in beam section follows current recommendations (Chen 1997; Anastadiadis 2000). 4

c) 30% RBS frame

Base shear—lateral top displacement hysteretic curves of two bay frames.

a) Reference frame

Figure 11.

b) 20% RBS frame

CONCLUDING REMARKS

The selected structures and the numerical results obtained are relevant for the objective of both, the authors and the beneficiary of the design job. RBS steel structures exhibit an improved behaviour under

seismic actions, though the reduction in the beams section carries an emotional refrain in accepting the weakening of the structure from the part of both, the designers and the beneficiary. The difficulty of accepting reduction in the area of beams sections is—implicitly—associated to the weakening of the entire structure. This makes the RBS technique a problem of mentality beyond its technical feature. What can be observed from the comparative study is that a reduction of 20% in the area of the beams is not associated with an improved collapse mechanism versus the collapse mechanism of the reference frame. A further reduction (30%) is able to lead to a desired collapse mechanism under seismic action. In what regards the ductility, a visible improvement can be observed (from both, the ductility coefficients and hysteretic base shear—lateral top displacements curves) in the case of RBS frames versus the reference frame. It has to be mentioned that due to the specificity of the accelerogram (having a peak value of about 2m/s2 between the 5th and 10th seconds, the time history behavioural curves presented in this study have been extracted from the total seismic response in the time interval from second 5 to second 10. The push over base shear—lateral top displacements curves show a slight difference in the deformation

165

capacity of reference frame and RBS frames. The maximum push over load of RBS six story frames decrease to about 80% of that of the reference frame. What has been the final decision of the beneficiary? The beneficiary accepted solutions of RBS type under the condition that the bearing capacity of the structures should be preserved. The solution that has been agreed upon is the strengthening of the beam section in the immediate vicinity of the column flange and leave the original beam section uncut. As it has been pointed out before, the RBS technique is a problem of both, the design procedure and mentality of the beneficiary and also, of the structural designer. REFERENCES Anastadiadis, A. et al. Improved ductile design of steel MR frames based on constructional details, Proceedings of 9th Conference on Metal Structures, Timisoara, Orizonturi Universitare, 367–376, 2000.

Alexa, P. et al. 2006. Ductility of RBS versus Base Isolated Steel Structures. Proceedings of IASS—APCAS Symposium, Beijing, 2006: 40–41. Chen S.J., Chu J. & Chou J.L.M., Dinamic behaviour of steel frames with beam flanges shaved around connections, J. Construct. Steel Research, Vol. 42, No. 1, 49–70, 1997. Chopra, A.K. & Goel, R.K. 2001. A Modal Pushover Analysis Procedure to Estimate Seismic Demands for Buildings: Theory and Preliminary evaluation’’, Report no. PEER 2001/3, Pacific Earthquake Engineering Research Centre, University of California, Berkeley, C. A., 2000. Engelhardt, M.D., Winneberger, T.,Z., A.J., Potyraj, T.J. 1995 The Dogebone Connection: Part II, Modern Steel Construction, August 1995. Gioncu, V. & Mazzolani, F.M. 1995. Alternative methods for assessing local ductility’’ Behaviour of Steel Structures in Seismic Areas. In E&FN Spon, UK (ed.) 1995: 182–190. NEHRP, Guidelines for the Seismic Rehabilitation of Buildings, FEMA 273, Washington D.C. October, 1996. NEHRP, Recommended provisions for seismic regulations for new buildings and other structures (FEMA 450), Washington D.C. 2003.

166

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Practical non-prismatic stiffness matrix for haunched-rafter pitched-roof steel portal frames H.K. Issa & F.A. Mohammad Nottingham Trent University, Nottingham, UK

ABSTRACT: Pitched roof steel portal frames are popular structures among single-storey buildings. Since the bending moment at the column-to-rafter joint is very high, the decision is made to haunch a part of the rafter adjacent to the joint. The haunch makes this part of the rafter linear non-prismatic. In this study, a stiffness matrix for non-prismatic members is derived and passed through regression analysis to set up a practical stiffness matrix. A column analogy is used to simulate the bending and shear effects whereas a virtual work method is used to involve axial force effect into the stiffness matrix. After a large amount of data was collected from regression analysis, quadratic coefficients have been obtained to generalize the stiffness matrix for both prismatic and non-prismatic members. The correctness of the obtained stiffness matrix is verified by a simple numerical example.

1

INTRODUCTION

Single-storey frame structures are extensively used in commercial, industrial, and leisure buildings. The nature of those buildings necessitates selecting a structural system which covers the area without intermediate columns. As steel provides an economical solution in those buildings, they are commonly constructed with steel frames (Saka 2003). Pitched roof steel portal frames are the most popular steelwork among the structures used in single-storey buildings. It is estimated that 50% of single-storey buildings are constructed with steel portal frames (Salter et al. 2004). Since the bending moment at the column-to-rafter joint is very high, the decision is made to haunch a part of the rafter adjacent to the joint. The haunch makes this part of the rafter linear non-prismatic. Because of non-uniform distribution of the bending moment in the non-prismatic member, material savings can be achieved, while the cross-section capacity is satisfied. Despite the additional cost of fabrication, the use of a tapered member in structural elements provides a more economical structure than the uniform member (Fraser 1983). When the direct stiffness method is adopted to fulfill the structure analysis, the stiffness matrix for each element of the structure should be necessarily constituted. The stiffness matrix for prismatic element is well documented in the text books. However, different developed stiffness matrices for non-prismatic members are not somehow practical to use. Saka (2003) implemented the stiffness matrix of non-prismatic member developed by Matheson et al. (1959; cited in Saka 2003) to carry out the optimization process on steel portal frames. Luo et al. (2007)

167

Figure 1.

Typical pitched roof steel portal frame.

adopted the transfer matrix method to a deduced general expression for the components of the stiffness matrix of non-prismatic members. Both developed stiffness matrices require long computation time as the matrices components are in integration form. Hence, in this study, it is attempted to develop a practical and generalized stiffness matrix for both prismatic and non-prismatic members so that it could be brought into office daily use. A typical pitched roof steel portal frame with a haunched-rafter is depicted in Figure 1.

2 2.1

STIFFNESS MATRIX Prismatic members

The slope deflection method is used to derive the stiffness matrix for prismatic members. For a twodimensional structure, a prismatic member has six degrees of freedom whereby a 6 × 6 matrix dimension is made. The derivation of the stiffness matrix of a prismatic member is well documented in standard structural analysis text books (see for example; Willems and Lucas (1978), McGuire et al. (2000), and

Ghali et al. (2003)). Accordingly, a stiffness matrix for such elements will take the form: ⎡ E [k] = 2 L

⎢ ⎢ ⎢ ⎢ ⎢ ⎢ ⎣

⎤ 0 −AL 0 0 6I 0 −12I /L 6I ⎥ ⎥ 4IL 0 −6I 2IL ⎥ ⎥ AL 0 0 ⎥ ⎥ Symmetric 12I /L −6I ⎦ 4IL

AL

0 12I /L

(1) Figure 2. Internal axial force and displacement of nonprismatic member.

where A = area of the member cross-section; E = modulus elasticity of the member; I = moment of inertia of the member; and L = length of the member. 2.2

which consequently gives  x x + A2 Ax = A 1 1 − L L Similarly (from Figure 2):

Non-prismatic members

The stiffness matrix of non-prismatic members is formed in practice by subdividing the member into a number of smaller prismatic member segments. This way of making a stiffness matrix is somewhat cumbersome and time-consuming as it substantially increases the global stiffness matrix of the frame. As a result an error is likely to occur and less accurate results are gained. It necessitates constructing a stiffness matrix for non-prismatic members to eliminate more members stiffness matrix. In this study, the Virtual Work Method has been implemented to derive the axial stiffness. Whereas Column Analogy (Ghalli et al. 2003) has been the guideline to constitute the stiffness matrix for bending and shear effects. The following procedures are applied to derive the stiffness matrix for non-prismatic members. Virtual work is work done by real forces acting through virtual displacement (Willems and Lucas 1978). The principle of virtual work is: δWint = δWext

δu =

x δum L

(7)

where A1 and A2 = cross sectional area of member ends. From Equation 3: L δWint =

d(δu)  x  du x E A1 1 − + A2 dx dx L L dx

Substituting Equation 7 into 8 and performing the integration, gives: δWint

L 

2 x2 δum x2 = 2 E A1 x − + A2 L 2L 2L 0 =

2 δum E 2 L



A1 L + A2 L 2

=

2 δum L



A 1 + A2 2

E (9)

(2)

Considering Equation 1 and substituting into Equation 4:

δu A1 + A2 2 Eδum Fδum = (10) L 2

L δεx EAεx dx

(8)

0

where: δWint =

(6)

(3)

0

δWext = Fδum

F=

(4)

where δWint = internal work; δWext = external work; E = modulus of elasticity; A = cross-sectional area of the member; εx = axial strain; F = applied axial force; and δu = axial displacement. As depicted in Figure 2. the relation between Ax , A1 and A2 can be defined as: (A1 − A2 )

L−x L

= (Ax − A2 )

(5)

(11)

and since F = kδum

(12)

therefore; k=



E (A1 + A2 )δum 2L

E (A1 + A2 ) 2L

where k = stiffness value.

168

(13)

Since the actual load for deriving the flexibility matrix is considered as unity, the vertical displacement will be:  



2 dx Mu1 Mu2 dx Mu1 (16) v= Fy + Mz EIx EIx and similarly for the rotation:  θz =

Mu1 Mu2 dx EIx





2 dx Mu2 EIx

Fy +

Mz

(17)

Since, {vθ} = [f ]{Fy Mz } Figure 3. (a) Simulation of column analogy on nonprismatic member. (b) Application of the unit load methods on the non-prismatic member.

Applications of force and displacement methods have, in some cases, been applied through the classical procedure known as column analogy and moment distribution. Column analogy can be applied for a plane framed analysis of one closed bent where the degree of redundancy does not exceed three. It involves calculations similar to those of stresses in column cross sections when subjected to combined bending moments and axial force (see Figure 3). The redundancy is chosen at a point called the elastic centre (Ghali et al. 2003). Regarding this condition, the column analogy has been used to derive the stiffness matrix of a single member. However, the column analogy imitates the force method and to form the stiffness matrix it is required to inverse the flexibility matrix generated from the force method. This is only true when the plane remains plane after deformation (elastic theory). Figure 3a shows a member of variable cross section idealized as a straight bar having a variable EI. Figure 3b refers to demonstration of a unit load and moment system acting on the member which has the shape of a strip with varying width = 1/EI and the length = l, where EI ≡ EI (x) the flexural rigidity at any point at a distance x from O, the centroid of the analogous column (the elastic centre). Figure 3 shows a non-prismatic member with varied depth along the length of the member. Figure 3b is used to determine the flexibility matrix of the member when the one end (j) is assumed fixed (force method). Using the virtual work, the moment due to unit load and unit moment are:

(18)

where {vθ} = displacement vector; [f ] = flexibility matrix; {Fy Mz } = force vector; v = vertical displacement of the joint; θz = rotation of the joint; Fy = shear force at the joint; and Mz = the bending moment at the joint, therefore the flexibility matrix could be constituted as: ⎤ ⎡   2 Mu1 dx Mu1 Mu2 dx ⎥ ⎢ EIx x ⎥ [f ] = ⎢  EI ⎣ Mu1 Mu2 dx M 2 dx ⎦ u2

⎡

EIx

(L1 + x)2 ⎢ EIx =⎢ ⎣  (L1 + x) − EIx

⎤ (L1 + x) ⎥  EIx ⎥ ⎦ dx EIx

(19)

The stiffness matrix is obtained by inversing the flexibility matrix: ⎡

1 ⎢  x2 dx ⎢ [k] = ⎢ EIx ⎣ L1  x2 EIx dx

 

1 1 EIx dx

⎤

L1 x2

EIx dx

+

⎥ ⎥

L12 ⎥ ⎦ x2 dx EIx

(20)

Considering the other member end (i) as fixed in Figure 3 and including the axial displacement, the stiffness matrix for the member could be formed as: ⎡ ⎤ k11 k12 k13 k14 k15 k16 ⎢ k22 k23 k24 k25 k26 ⎥ ⎢ ⎥ k33 k34 k35 k36 ⎥ ⎢ [k] = ⎢ (21) k44 k45 k46 ⎥ ⎢ ⎥ ⎣ Symmetric k55 k56 ⎦ k66

Mu1 = −(L1 + x)

(14)

where:

Mu2 = 1

(15)

k11 = −k14 = k44 =

169

 −

EIx

E (A1 + A2 ) 2L

k22 = −k24 = k54 =  k26 = −k56 =  k35 = −   

k33 = 

L2 x2 EIx dx

x2 EIx dx

1 1 EIx dx

+

L12

k4 = −k8 = d(6I  ) k6 = −k12 = c(6I  ) k7 = g(4I  L) k13 = e(4I  L)



x2 EIx dx

L1 L 2 +  x2 EIx dx   L22 1 +  x2 =  1 EIx dx EIx dx 1



1 EIx dx

k12 = k13 = k15 = k16 = k45 = k46 = 0 In the special case, when EI is constant, L1 = L2 and A1 = A2 then the stiffness matrix will be same as that for the prismatic members (Equation 1).

3

REGRESSION ANALYSIS

The derived stiffness matrix in Equation 22 has the components with integration. This will need covering more loops in the program subroutine to formulate the integration. To eliminate this, the stiffness matrix is passed through regression analysis. All eighty existing standard cross-sections available in BS 5950 have been engaged in analysis. Furthermore, the difference between the depths of both ends of the members has varied between 0 and 1.28 m. After performing the analysis, the stiffness matrix given in Equation 22 can be redefined as follows: ⎡ ⎤ k1 0 0 k2 0 0 ⎢ k3 k4 0 k5 k6 ⎥ ⎥ E ⎢ k7 0 k8 k9 ⎥ ⎢ (22) [k] = 2 ⎢ ⎥ k 0 0 ⎥ 10 L ⎢ ⎣ Symmetric ⎦ k11 k12 k13 where: k1 = −k2 = k10 = a(A L) k3 = −k5 = k11 = b(12I  /L) k9 = f (2I  L) If A1 ≥ A2 : k4 = −k8 = c(6I  ) k6 = −k12 = d(6I  )

k7 = e(4I  L) k13 = g(4I  L) and if A1 ≤ A2 :

L1

k36 = − 

k66

1 x2 EIx dx

where: A = A1 + A2 ; I  = I1 + I2 ; A1 & A2 = areas of the member ends; I1 & I2 = moments of inertia of the member ends; and a = 0.50 b = 0.50 − 0.53d + 0.17(d)2 c = 0.50 − 0.59d + 0.21(d)2 d = 0.50 − 0.47d + 0.13(d)2 e = 0.50 − 0.56d + 0.21(d)2 f = 0.50 − 0.64d + 0.23(d)2 g = 0.50 − 0.39d + 0.08(d)2 d = difference between the depths of member ends. In the case when the difference between both depths of the member ends (d) is zero (prismatic member), the stiffness matrix in Equation 23 will have the same components of the one in Equation 1. To have a more cost effective shape for haunched part of the steel portal frame, the depth of the haunch is taken as the same depth of the rafter. Having said this, the elements of the stiffness matrix given in Equation 23 can be refined to the following equations: where: k1 = −k2 = k10 = 0.50A L k3 = −k5 = k11 = 0.32(12I  /L) k9 = 0.28(2I  L) If A1 ≥ A2 : k4 = −k8 = 0.36(6I  ) k6 = −k12 = 0.29(6I  ) k7 = 0.39(4I  L) k13 = 0.29(4I  L) and if A1 ≤ A2 : k4 = −k8 = 0.29(6I  ) k6 = −k12 = 0.36(6I  ) k7 = 0.29(4I  L) k13 = 0.39(4I  L)

170

4

BENCHMARK EXAMPLE

22% decreased lateral displacements at the top of the columns (Figure 4, joint 6).

Implementing the direct stiffness method, structural analysis is conducted on a pitched roof haunchedrafter steel portal frame (given in Figure 4) to examine the suitability of the developed matrix. The steel portal frame is assumed to experience the gravity load of 14 kN generated by purlins spaced horizontally at 3 m in out-plane. A horizontal load of 0.7 kN is assumed to act on the frame to portray the sway behavior. As the column behaves as a beam-column section due to large value of the bending moment, the universal beam section (not universal column section) is almost used for the column in steel portal frames. The universal beam sections of 762 × 267 × 134 and 686 × 254 × 170 are used for the columns and rafters successively and the depth of the haunched is assumed as equal as the depth of the section of the rafter. The steel frame is analyzed twice named as case 1 and case 2. In case 1 the developed stiffness matrix (Equation 23) is used to set up global stiffness matrix whereas in case 2 the haunched part of the rafter is subdivided into eight smaller prismatic elements. The outputs of both cases is compared and tabulated in Table 1 including support reactions, nodal displacements, and elapsed time of analysis. After analysis of the structural response, it was found that using the idea of subdividing the non-prismatic member into prismatic elements to form global stiffness matrix uses three times more computation time than when Equation 22 is used. However, results show that using the developed stiffness matrix ends up with 6% increased vertical displacement at apex (Figure 4, joint 4), and

5

CONCLUSIONS

A column analogy was simulated and a virtual work method was adopted to set up and generalize the stiffness matrix for both prismatic and non-prismatic members. Through regression and structural analyses, it is found that the formed stiffness matrix comes up with 200% saving of time in determining the member forces. However, it increases vertical displacement of the apex by 6% and decreases the lateral displacement of the frame by 22%. As a result the developed stiffness matrix could be brought into office daily use by structural engineers. As massive iterations are required for non-linear analysis and the optimization process, it is concluded that using the developed stiffness matrix can assist the operation to reduce the computation cost. It can lead the design problem to a faster convergence into an optimum solution when the stiffness matrix is implemented in the analysis part of structural optimization. ACKNOWLEDGEMENTS The authors would like to thank Mr. Guy Birkin, a PhD student in Art & Design at Nottingham Trent University for devoting his time to help conducting the regression analysis. REFERENCES

Figure 4. Pitched roof steel portal frame used as benchmark example. Note: All loads are in ‘kN’ and dimensions are in ‘m’. Table 1.

Result of structural analysis.

Component

Case 1

Case 2

Difference (%)

FxL , kN FxR , kN FyL , kN FyR , kN δu2 , mm δu6 , mm δv4 , mm Elapsed Time, Sec

60.11 60.81 69.88 70.12 2.2 1.8 25.3 1

60.67 61.37 69.88 70.12 2.1 2.3 23.9 3

0.1 0.1 0 0 5 22 6 200

171

Fraser, D.J. 1983. Design of tapered member portal frames. Journal of Constructional Steel Research 3(3): 20–26. Ghali, A., Neville, A.M. and Brown, T.G. 2003. Structural Analysis; A Unified Classical and Matrix Approach. 5th ed. New York: Spon Press. Luo, Y.Z., Xu X. and Wu, F. 2007. Accurate stiffness matrix for non-prismatic members. ASCE, Journal of Structural Engineering. 133(8): 1168–1175. McCormac, J.C. 2007. Structural analysis; using classical and matrix method. 4th ed. New Jersey: John Wiley & Sons. McGuire, W., Gallegher, R.H. and Ziemian, R.D. 2000. Matrix structural analysis. 2nd ed. NewYork: John Wiley & Sons. Saka, M.P. 2003. Optimum design of pitched roof steel frames with haunched rafter by genetic algorithms. Computers and Structures 81: 1967–1978. Salter, P.R., Malik, A.S. and King, C.M. ed. 2004. Design of Single-Span Steel Portal Frames to BS 5950-1:2000. Berkshire: The steel Construction Institute. Willems, N. and Lucas, W.M. 1978. Structural Analysis for Engineers. Tokyo: McGraw-Hill.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Relationship between strength of scaffolds and shear rigidity of frames H. Takahashi, K. Ohdo & S. Takanashi National Institute of Occupational Safety and Health, Tokyo, Japan

ABSTRACT: Recent studies confirm that the shear rigidity of the vertical frame is the influencing factor when scaffolds are buckling. However, no such regulations exist and little research data is available with respect to the shear rigidity of the horizontal frame. It is thought that the shear rigidity of the horizontal frame influences the overall resistance of the scaffolds to buckling, but the importance of this phenomenon remains unclear. In this study we propose an equation for evaluating the strength of standard prefabricated scaffolds. A buckling analysis of such scaffolds is executed to act as a parameter in assessing the shear rigidity of the vertical and horizontal frames. The validity of the proposed evaluation equation is also examined. 1

INTRODUCTION

The standard prefabricated scaffolds might be used as the concrete support. Therefore, excessive vertical load is also likely to act on prefabricated scaffolds. The buckling modes of scaffolds are illustrated in Figure 1. They include member buckling, when each story of the scaffold curves, and total buckling, when the entire side of the scaffold curves (Mori et al. 1962). Recent studies (Mori et al. 1962) confirm that buckling comes about mainly when the stiffening member in the vertical frame is shorting. Therefore, total buckling happens when the shear rigidity in the vertical frame is deficient. On the other hand, it is thought that the shear rigidity of the horizontal frame influences the overall resistance of the scaffolds to buckling, but the importance of this phenomenon remains unclear. Load

Load

In this study a buckling analysis of standard prefabricated scaffolds was executed, using the shear rigidity of the vertical and the horizontal frames as parameters, to investigate the strength performance of prefabricated scaffolds. 2

OUTLINE OF NUMERICAL ANALYSIS

In the analysis, ANSYS 11.0, a general-purpose Finite Element Method, was used. The material was assumed to be the isotropic elastic-plastic model, and the yield criterion was assumed to be the von Mises yield. Figure 2 shows the relationship between the stress σ and the strain ε used for the analysis, and the material properties in the analysis are shown in Table 1. In the table, ν shows Poisson’s ratio. Referring to the actual material of the vertical frame, the horizontal member and leg member in the vertical frame were STK500, and the stiffening member in the vertical frame was STK400. The horizontal frame was assumed to be SS400, a typical structural material.

y

E st E (a) Member buckling

Figure 1.

(b) Total buckling

Figure 2. analysis.

Scaffolds buckling modes.

173

Relationship between stress σ and strain ε in

Material property in analysis.

Member (Material)

ν

Shear rigidity of the vertical frame ks (N/mm)

Table 1.

E Est σy (N/mm2 ) (N/mm2 ) (N/mm2 )

The leg member 0.3 205000 and the horizontal member of the vertical frame (STK500) The stiffening 0.3 205000 member of the vertical frame (STK400) The horizontal 0.3 205000 frame (SS400)

2050

355

2050

235

2050

235

160

120

80

40

0 0

500

1,000

1,500

Length of the stiffening member hs (mm) Displacement

Figure 4. Relationship between shear rigidity of the vertical frame ks and length of the stiffening member hs .

(Unit of size: mm) 900 50

Shear force Q

490

Figure 4 shows the analytical results. When the length of the stiffening member hs was 0 mm, the shear rigidity of the vertical frame, ks , was 0.013 kN/mm. The value of ks increased as the length of the stiffening member, hs , increased. h0=1,700

Stiffening member (Diameter × Thickness = 27.2 × 2.0)

550

hs

Horizontal member (Diameter × Thickness = 42.7 × 2.5)

4

EVALUATION OF THE BUCKLING LOAD ON THE VERTICAL FRAME

460

Z X

Leg member (Diameter × Thickness = 42.7 × 2.5) 200 120

Figure 3.

3

150

90

Y

Vertical frame in analysis.

CONSIDERATION OF THE ANALYTICAL MODEL FOR THE VERTICAL FRAME AND EVALUATION OF SHEAR RIGIDITY

The vertical frame in the analysis is illustrated in Figure 3; it represents the type of frame generally used on construction sites. The length of the stiffening member hs shown in Figure 3 is a typical length for those used in actual vertical frames. The beam elements with two nodes were used as a finite element. To examine the strength performance of the vertical frame, the length of the stiffening member hs was adjusted, and the corresponding shear rigidity of the vertical frame ks was examined in the analysis. The boundary condition at the bottom of the leg member in the vertical frame was assumed to be the pin node as the most risky case.

To examine the strength performance of the vertical frame, the length of the stiffening member hs was adjusted, and the corresponding buckling load of the vertical frame Ps was examined in the analysis. The boundary condition in this case, at the bottom of the leg member in the vertical frame was assumed to be the pin node as the most risky case. In setting the boundary condition, the upper and lower edge of the leg member in the vertical frame was taken as the pin node. The horizontal movement of the upper and lower edge of the leg member was held as shown in Figure 5. The vertical load was set from the upper part of the vertical frame. The buckling load of the one vertical frame Ps in this boundary condition more or less equaled the buckling load of prefabricated scaffolds, when the member buckling is occurred. Moreover, an actual steel member usually bends a little at the beginning, and this is known as the ‘initial crookedness’. The initial crookedness at the center of the member was assumed to be about 1/1,000 of length of the member from measurements of the actual member (Narioka et al. 1970). The initial crookedness was set by the sine wave referring to this value. In the Y direction of the vertical frame, the maximum displacement, due to the initial crookedness, was 1/1,000 (1.7 mm) of the height of the vertical frame.

174

Load P

60

60

1,800

Load P

Outside frame (Size of cross section: height × width bh= 50.0 × bh)

Displacement

Shear force Q

Figure 7.

Z

Buckling load of the vertical frame Ps (N)

Figure 5.

Buckling of the vertical frame.

96,000

92,000

88,000

Horizontal frame in analysis.

150 125 100 75 50 25 0 0

84,000

10

20

30

40

Width of the cross section bh (mm)

Figure 8. Relationship between shear rigidity of horizontal frame kh and width of cross-section of horizontal frame bh .

80,000 0

500

1,000

1,500

Length of the stiffening member hs (mm)

shear force as shown in Figure 7 was placed on the horizontal frame, and the frame’s shear rigidity was tested. Then, a cross-section of the outside frame in the horizontal frame, bh , was adjusted for width and the shear rigidity kh was tested again. Figure 8 shows the analytical results. The shear rigidity of the horizontal frame, kh , increased as the cross-section of the outside frame in the horizontal frame, bh , increased.

Figure 6. Relationship between buckling load of the vertical frame Ps and length of the stiffening member hs .

Figure 6 shows the analytical results. When the length of the stiffening member hs was 0 mm, the buckling load of the vertical frame, Ps , was 81,700 N. Ps increased as the length of the stiffening member, hs , increased. 5

(Unit of size: mm)

X

Shear rigidity of the horizontal frame kh (N/mm)

Y

740

Hook (Size of cross section: height × width = 50.0 × 8.0)

6

AN ANALYTICAL MODEL OF THE HORIZONTAL FRAME

An analytical model of the horizontal frame was modeled simply because an actual horizontal frame has a complex shape. The horizontal model used in the analysis is illustrated in Figure 7. Beam elements with two nodes were used as a finite element of the horizontal frame. The end point of the horizontal frame was assumed to be the pin joint because this generally forms the junction between the vertical frame and the horizontal frame in actual construction sites. The

6.1

BUCKLING ANALYSIS OF PREFABRICATED SCAFFOLDS Analytical model and method

Stays must be fastened to the scaffold in the current regulations. The intervals between stays must not be more than 8,000 mm in the horizontal direction and 9,000 mm vertically. The prefabricated scaffolds in this study had stays placed at intervals appropriate for a 5-story (1,700 mm [length of vertical frame] × 5 = 8,500 mm) and 4-bay (1,800 mm [length of horizontal frame] × 4 = 7,200 mm) scaffold,

175

Load P

Load P

Stay

and was set according to the sine wave. In the X direction the maximum displacement due to initial of the scaffolds. In the Y direction the maximum displacement was 1/1,000 (1.7 mm) of the height of each story of the scaffolds. The shear rigidity of the vertical and horizontal frames was adjusted and a buckling analysis was carried out on the models. 6.2

Z Y

X

Hold point

(a) Prefabricated scaffolds (b) Analytical model Figure 9.

Prefabricated scaffolds in analysis.

as shown in Figure 9 (a). When a vertical load was set on top of the vertical frame with stays in position, it is thought that the stay shared the load. However, we then set the vertical load on the top of the frame without stays in place, to simulate a situation of greatest risk. The analysis model of the scaffold was a 10 story structure. In the tested models of prefabricated scaffolds, the row of the vertical frame containing the stay was assumed to have no movement in a horizontal direction. The prefabricated scaffold, as shown in Figure 9(a), was simplified to form the 2-span models as shown in Figure 9(b). The X-axis represents the span direction of the vertical frame of one row and two horizontal frames; the Y-axis represents the longitudinal direction of the scaffold; and the Z-axis represents the height direction. Where the end of the horizontal frame is not connected with the vertical frame was fixed in the X direction. The Y (length) direction of the horizontal member in each story was fixed in terms of the effect of the brace, so the brace was not modeled as the element. The boundary of the bottom edge of the scaffolds was determined as the pin joint as the highest risk case. The joints of both vertical and horizontal frames were also determined as the pin joints, and the joint connecting sections of the vertical frame was determined as the rigid joint with reference to the actual construction sites. The vertical load was set from the top of the scaffolds on the assumption that standard prefabricated scaffolds are used with the concrete support. Initial crookedness was set both in the X direction, where total buckling is usually caused, and in the Y direction, where member buckling is usually caused

Analytical result for buckling mode and evaluation equation for scaffolds strength

Degree of deformation of the scaffolds at maximum load (buckling load) is illustrated in Figure 10 by numerical analysis. Figures 10(a) show the results of ks = 130 N/mm (hs = 1, 500 mm), kh = 30 N/mm. Figure 10(b) show the results of ks = 25 N/mm (hs = 300 mm), kh = 1 N/mm. Figure 10(a) became the member buckling, and Figure 10(b) did not become the total buckling. In case of no member buckling, the displacement was caused in the X direction, and the lowest story of the scaffolds was greatly deformed. This displacement does not have been the total buckling, but it is thought that this displacement was caused due to the lack of shear rigidity in the vertical frame. Moreover, it is thought that also influence at the boundary condition between the scaffolds and ground. The lowest story of the scaffolds is particular attention. Figure 11(a) shows a model of the vertical frame at the lowest story. Figure 11(b) is a simplified model of Figure 11(a), and here the vertical frame is shown as one member. The value k as shown in Figures 11(a) and 11(b) represents a spring constant occurring where the top point O of the leg member in the vertical

Load P

Load P

Z Y

(a) Member buck Figure 10.

176

Buckling mode.

X

(b) Total buck

Initial crookedness Displacement Load P

When it is solved with the spring constant k, k is shown as follows.

0

Load P

k=

O

Z X

Y

Figure 11a. Model of the vertical frame at the scaffold’s lowest story.

0

2P

k

kcr =

O

h0=1700 mm

Z

Ps h0

(4)

When the Equation 3 is expressed by the equivalent geometrical moment of inertia Ie that derives from the influence of the stiffing member of the vertical frame, kcr is shown in the following equation:   2π 2 E I0 + Is hh0s π 2 EIe kcr = = (5) h30 h30

Qk = k ( 0+ )

Y

(3)

The buckling load 2P is proportional to the spring constant k. When k becomes infinity, the buckling load 2P becomes infinity also. When 2P is the same value as the buckling load of the vertical frame Ps , the prefabricated scaffolds experience member buckling. At this time, point O doesn’t move. The buckling load always becomes Ps whatever the value of the spring constant k. When this Ps is substituted for 2P in the Equation 3, the next equation is obtained:

h0=1700mm

Spring constant k

2P h0

X

Qk = k ( 0+ ) 2P

Figure 11b. Simplified model of the vertical frame at the scaffold’s lowest story.

frame is braced. k takes all the horizontal force in the X direction of the vertical frame. When

When the shear rigidity of the vertical frame or the horizontal frame is the same as or more as kcr as shown in Equation 5, the scaffolds experience member buckling. For this situation Figure 12 shows a comparison between kcr of the Equation 5 and the shear rigidity of the vertical frame ks . The horizontal axis in Figure 12 is the slenderness ratio of the vertical frame λ calculated by using the equivalent geometrical moment of inertia Ie . λ is shown as follows: λ=

h0 h0 h0 =  =  Ie ie I0 +Is hs 2A0

k = ks + kh

A0

(1)

the horizontal frame is assumed to be a horizontal bracing member for the vertical frame, k is added, as well as the value for shear rigidity kh for the horizontal frame at the lowest story. Therefore, k is given as follows: When the vertical frame with initial crookedness buckles as shown in Figures 11(a) and 11(b), horizontal displacement δ is caused. Simultaneously, a spring reaction force Qk = k(δ0 + δ) is caused. The balance of the moment at point O as shown in Figure 11(b) then becomes as follows: 2P(δ0 + δ) = k(δ0 + δ)h0 = Qk h0

(6)

h0

where A0 is a cross-section of the leg member, and ie is the radius of gyration as influenced by the stiffening member of the vertical frame. Here, hs = 1, 500 mm is as about λ = 110.2, and hs = 0 mm is as about λ = 119.4. It is compare between kcr and the shear rigidity of the vertical frame. In the next equation, the prefabricated scaffolds become member buckling despite the shear rigidity of the horizontal frame. kcr < ks

(7)

(2) In the next equation we see that when shear rigidity is only found in the vertical frame, the prefabricated

where, h0 is the height of the vertical frame.

177

140

0.5 Shear rigidity of the vertical frame ks

120

Buckling load of the vertical frame

0.4

0.3

80

Pm/Py

k (N/mm)

100

kcrh (Equation (9))

60

0.2

40 20

: kh =30 N/mm : kh =20 N/mm : kh =10 N/mm : kh =1 N/mm

0.1

kcr (Equation (4))

0

0 110

112

114

116

118

120

Figure 12. Relationship between shear rigidity of the vertical and the horizontal frames and scaffolds strength.

110

112

114

116

118

120

Figure 13. Relationship between the yield ratio of the axial force and the slenderness ratio.

scaffolds do not become member buckling because ks is insufficient.

frame λ calculated by using the equivalent geometrical moment of inertia Ie . Py is shown as follows:

kcr > ks

Py = 2A0 σys

(8)

In case of the Equation 8, when the value of kh in Equation 1 becomes greater than the value of kcrh in the next equation, the prefabricated scaffolds do become member buckling. kcrh = kcr − ks

(9)

From Figure 12, when the scaffolds become member buckling, the shear rigidity of the horizontal frame kh is in the range of 0–36 and where λ = 115−120. In the models of Figure 9(b), the horizontal frame is set two spans to the one vertical layer. Thus, the shear force of one horizontal frame is half of that for one vertical frame. Therefore, when member buckling occurs, kcrh is within the range of 0–18. Referring to this value, the shear rigidity of the horizontal frame kh were set for four values of kh = 1, 10, 20, and 30 N/mm for both models shown in Figures 9(b). We also tested six values for ks = 14, 25, 44, 77, 117, and 130 N/mm (hs = 0, 300, 600, 900, 1200, 1500 mm) to analyze the shear rigidity of the vertical frame. 6.3 Result of buckling analysis and assessment of evaluation equation Figure 13 shows the analytical results. The vertical axis shown in Figure 13 is the ratio of the buckling load Pm to the yield axial force Py . The horizontal axis shown in Figure 13 is the slenderness ratio of the vertical

(10)

where A0 is a cross-section and σys is the yield stress of the leg member. The curve in Figure 13 represents the buckling load of the vertical frame Ps . The white points show the analytical results for the 5-story model, and the black points show the analytical result for the 10-story model. At the time when member buckling occurred, the value of Pm /Py for the white points and the black points were the almost same as for the curve. When no member buckling was occurring, the value of Pm /Py for the white points and the black points were less than the value of Pm /Py for the curve. From Figure 12 we see that when the analytical model is almost at λ < 115, the scaffolds experienced member buckling regardless of the value of shear rigidity of the horizontal frame kh . Figure 13 shows that when the value of Pm /Py of the white and black points at λ < 115 were the almost same as the Pm /Py value of the curve, regardless of the value of kh , the scaffolds experienced member buckling. In the case of λ > 115, when λ is larger and kh is smaller, the value of Pm /Py for both white and black points was less than the Pm /Py value of the curve. When λ is larger, it is necessary to enlarge kh to see member buckling occur. It is thought that the results of Figure 13 correspond to the results of Figure 12. Therefore, it was proven that ks and kh can be calculated from Equations 4–9, to ascertain when member buckling would occur. Moreover, when the value of λ and kh for the white and black points is the same, the value of Pm /Py in the white and black points is

178

almost the same. Therefore, when the boundary condition between the scaffold and the ground is marked by a pin and the scaffold is between 2 stories and 10 stories in height, we can assume that the strength of the scaffold is decided by the shear rigidity of the vertical and horizontal frames of the lowest story of the scaffolds, regardless of the number of stories. In this study, we assumed the position of the boundary at the bottom of the scaffolds to be the pin joint as in a situation of highest risk. However, the boundary between the scaffold and the ground does not always get pinned; actual scaffolds use a jack base at their lowest point; in that situation it is probable that the number of stories does influence the strength of scaffolds. 7

CONCLUSIONS

A buckling analysis of prefabricated scaffolds was conducted in our study to provide a parameter in the study of shear rigidity of the vertical and horizontal frames. We also investigated the validity of a proposed evaluation equation. The results of this study can be summarized as follows: 1. When the junction between the scaffold and the ground is pinned, as for a highest risk situation and the scaffold is between 2 and 10 stories, we conclude that the strength of the scaffold is decided

179

by the shear rigidity of the vertical and horizontal frames of the lowest story of the scaffold, regardless of the number of stories. 2. When the shear rigidity of the vertical frame ks is greater than kcr of the Equation 5, i.e. kcr < ks , the prefabricated scaffolds will become member buckling in spite of the shear rigidity of the horizontal frame kh . Member buckling will also occur in the case of kcr > ks , when the value of kh is kcr − ks (= kcrh ) or more. REFERENCES Mori, Y., Mae I. and Kunimori M. 1962. On the Loadcarrying Capabilities of the Steel Tubular Vertical Frames which are Used for Supporting the Concrete Bridge Mold, Research Report of the Research Institute of Industrial Safety, 3: 1–8, in Japanese. Narioka, M., Fukumoto, S. and Ito, K. 1970. Buckling Curves Studied by C. E. A. C. M. Commission VIII, JSSC, 6(55): 56–71, in Japanese.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Sequential failure analysis of tension braced MRFs M. Lotfollahi Department of Civil Engineering, Amirkabir University of Technology, Tehran, Iran Ministry of Petroleum, National Iranian Oil Engineering & Construction Company, Tehran, Iran

M.M. Alinia Department of Civil Engineering, Amirkabir University of Technology, Tehran, Iran

ABSTRACT: In this paper, Moment Resisting Frames (MRFs) reinforced by bracing elements, i.e. Braced Moment Resisting Frames (BMRFs) are studied. The main aim of this paper is to consider several characteristic references such as ductility, stiffness, optimum size of braces, load contribution shares in the elastic and inelastic regions, and yielding sequence of members. These properties are not readily fully understood and can only cautiously be associated with the current aspects of structural behavior and design. The main reason for this situation is, believed to be, the absence of an acceptable methodology to examine the philosophies of gradual collapse in a manner which aligns with the conventional approaches of structural design. This paper attempts to remedy this omission by proposing and illustrating a developed analytical model. 1

INTRODUCTION

The purpose of a proper structural design is to ensure a ductile performance without permitting structural collapse. However, plasticized zones and the resulting large permanent displacements induced by strong earthquakes can cause irreparable damages to conventional steel structural systems, such as Moment Resisting Frames (MRFs) and/or Concentrically Braced Frames (CBFs). In a MRF system, the inelastic deformations of beams result in large dissipation of energy (Della Corte et al. 2002) and substantial damages to gravity load carrying members are expected. Additionally, the flexibility of the MRF system may result in large drift-induced non-structural member damages under seismic loading and consequently, costly postearthquake retrofit of structures is required. On the other hand, CBF systems possess high elastic stiffness which prevents large drifts. Furthermore, the presence of bracings can lead to material savings since they are subjected to less bending (Gwozdz et al. 1997). CBFs are economical, and their strength and stiffness assist in achieving serviceability limit states in Performance Based Seismic Designs (PBSD). During severe earthquakes, brace yielding and buckling can occur. This behavior provides the deformation and energy dissipation capacities to satisfy life-safety (L.S.) and collapse-prevention (C.P.) performance objectives. The Special Concentrically Braced Frame (SCBF) design requirements in the AISC Seismic Design Provision (AISC 2005) address these objectives. Due to these advantages and the uncertainty of

181

the performance of special moment resisting frames (SMRFs) after the 1994 Northridge Earthquake, SCBFs have been increased in usage in recent years (Roeder 2002). Recent researches have shown that SCBF frames with connections designed to meet current practice may exhibit a behavior that is very different from that envisioned in the original design and may have relatively poor seismic performance (Uriz et al. 2004; Lehman et al. 2007). To improve the seismic response of these frames, an improved design concept has been proposed with an emphasis on identifying a primary yield mechanism (e.g., tensile yielding or buckling of braces) and balancing multiple secondary yield mechanisms (e.g., yielding of gusset plates) with the primary yield mechanism to enhance the overall drift capacity of the system (Roeder et al. 2005). In particular, increased flexibility and yielding in the connections at larger drift demands have been shown to improve the seismic performance of frames and permit the use of HSS braces (Lehman et al. 2007). In this investigation, at first, the behavior change of Braced Moment Resisting Frame (BMRF) system with respect to the employment of different diagonal tension braces in the form of various single story frames is investigated and the differences between the distributed and lumped plasticity approaches are considered. The formation and propagation of plastic hinges and the corresponding displacement histories for different single story BMRFs are displayed and the effects of brace stiffness on the behavior of MRFs and the drift reductions are discussed. Then by developing an analytical model and corresponding correction process, the behaviors of several multi story BMRFs in the

form of 2, 4 and 6 story buildings having different stiffness and brace sizes under regular lateral loadings are studied. Unsuitable mechanisms, such as frame mechanism (elastic brace behavior), brace mechanism (similar to MRF collapse manner), and soft story mechanism (inappropriate full plastic hinge formation in one story) are evaluated by developing an example problem in a 2 story BMRF system.

2

METHOD OF ANALYSIS

The geometrical configuration of a presumed generic single-story planar frame is assumed with a 3.0 m interstory height and a bay width of 4.0 m. Structural design is carried out according to the Eurocodes, accounting for the seismic action through the simplified response spectrum analysis. The yield strength and the modulus of elasticity were assumed to be 240 MPa (corresponding to steel grade ST37) and 200 GPa, respectively. European wide flange sections (IPB or HEB) were selected for frame members, and double channel sections were considered for braces. The characteristic gravity loads consisted of a total live load of 20 MPa on each floor and 15 MPa on the roof, plus a dead load of 50 MPa with 6 m spacing between transverse frames. In the design under gravity loads, factored loads were considered according to the Eurocode 1 (2004) and the Eurocode 3 (2004). In the seismic design analysis, live loads were reduced for the combination of seismic and other actions according to the Eurocode 8 (2004). By selecting IPB 240 section for frame members and limiting the stress ratio to 0.95, both strength and deformability requirements were accomplished. The design drift was limited to min /H = 0.003 to account for the stiffening effect of bracings. To investigate the effect of bracings on the nonlinear response of frames, various brace elements were considered in the form of various double channel box sections 2UNP30, 2UNP40, 2UNP60, 2UNP80, 2UNP100, 2UNP120 and 2UNP140. Different 2, 4 and 6 story frame models were analyzed while frames were given various member sections, namely IPB140, IPB180,

Figure 1.

IPB240, and IPB300. The frame dimensions were similar to the one story frames having, 4 m bay and 3 m height in each story. Multi story frames had different bracing configurations to allow various comparative relative rigidities. Nonlinear analyses were performed by two different plasticity approaches through two separate nonlinear finite element modeling. A series of analysis were carried out by the distributed plasticity approach using the shell element, S4R of ABAQUS. S4R is a 4-noded quadrilateral, stress/displacement element with reduced integration, large strain formulations and six degrees of freedom per node. Frames were modeled with rigid beam to column connections and incremental static pushover loads were applied to the top flange nodes of beams. In addition, the lumped plasticity approach was employed in ETABS using the FEMA steel beam and column elements, rigidplastic moment hinge elements, panel zone elements and inelastic bar elements. In the sequential failure analysis, plastic hinges were manually introduced to models by the linear P/V/M hinges in the form of reduced calibrated rotation stiffness within the FEMA steel beam and column elements. Also, linear elastic bar element was employed for braces, which was presumed to have a real stiffness in the elastic range and a reduced calibrated stiffness in the post-elastic phase.

3

DISTRIBUTED AND LUMPED PLASTICITY

For comparison and modeling verification purposes, several models were analyzed by the distributed and lumped plasticity approaches and typical results for frames having IPE 240 members are shown in Fig. 1. The analyses were carried out under FEMA 273 (1997) considerations regarding the development of plastic hinges. These results show that the maximum difference in the linear region between the two methods was 3.8%. This small difference is due to different gusset plate modeling in the two methods. On the other hand in the nonlinear regions, the maximum difference (i.e. 5.4%) is due to the presumed incorporated methods of plasticity and the formation of plastic hinges.

Distributed and Lumped plasticity results for typical single story BMRF system.

182

4

ANALYTICAL MODEL STRATEGY

In another rigorous study, several more frames were analyzed. The frames had similar stiffness in each story and the ultimate design capacity was employed to ensure all predominant yield modes. The formation and propagation of plastic hinges in brace and frame members via distributed and lumped plasticity approaches were evaluated and the results were classified in different categories. To follow, the most suitable collapse mechanism in each category was

assessed as discussed in following sections. The results of this section is utilized for the viability of using a summation curve (dual model = truss model + frame model) by introducing a new analytical model and developing an applicable correction process in the linear and nonlinear regions. The proposed analytical model, the ‘‘Truss-Frame’’ model shown in Fig. 2, contains the following features: (i) it develops a closed form stiffness and strength relationship between the individual truss and frame actions within the BMRF system, (ii) it finds the contribution shares of the frame and truss systems in multi story BMRFs via an empirical relationship in the linear stage and design curves in the nonlinear stage, and (iii) it enables a comparative scheme between the linear and nonlinear contribution shares of multistory BMRFs. In the linear stage, the multiple accounting of the axial stiffness of columns in the proposed model was dealt with introducing reduction factors α and β. These stiffness reduction factors were respectively applied to the columns of the frame and truss systems. The factors were calculated via satisfying two conditions: (i) the sum of axial stiffness of the truss and frame systems should be equal to the corresponding BMRF system (α + β = 1), (ii) the internal forces in the Truss-Frame and BMRF members should be similar.

Figure 2.

The Truss-Frame model.

Figure 3.

Comparison of pushover curves between Truss-Frame and BMRF system; 2 story BMRF with IPB240.

Figure 4.

Comparing the results of Truss-Frame model and BMRF system in linear and nonlinear stages.

183

The results of the implementation of this correction procedure for typical 2 story BMRFs and related Truss-Frame models are presented in Fig. 3. The results show well correlation between the two systems in the linear portion after the application of correction factors. However, in the nonlinear stage, especially in models with stiffer braces, there are some differences. In order to correct this discrepancy (due to the post-yield reserves of braces that affect the pattern of plastic hinge formation in frame members), step-by-step trend of plastic hinge formation in the BMRF system was dictated to the TrussFrame model during the sequential collapse analysis. Therefore, the post-yield stiffness and strength of moment frames after implementing plastic hinges in each sequence of collapse mechanism and their corresponding strain hardening was precisely re-created (Lotfollahi et al. 2008a, b). Fig. 4 shows the results of the proposed correction process for a typical four story BMRF with IPB 240 for the frame sections and 2UNP80 for the brace sections.

Figure 5.

5

SEQUENTIAL FAILURE ANALYSIS

The nonlinear behavior of single story BMRFs regarding their collapse mechanisms and their sequences of plastic hinge formations are illustrated in Fig. 5. The numbers inside the circles indicate the sequence and the numbers beside them show their corresponding displacements. The best frame-brace contribution provides the most suitable failure mechanism with sufficient strength and stiffness together with high energy dissipation. Similar sequential failure analyses were carried out for 2, 4 and 6 story BMRFs. The results for 2 story BMRFs are presented in Table 1 in accordance with the plastic hinge locations in Fig. 6. Table 1 is utilized to demonstrate the gradual collapse mechanisms for a wide variety of brace to frame stiffness ratios. Furthermore, Table 2 shows all possible collapse mechanisms in the presumed 2 story BMRF which can provide an appropriate plan for design procedures. Table 2 also demonstrates the effect of the relative rigidities

Sequence of plastic hinge formation and their related displacements in mm.

184

Table 1.

Plastic hinge formation pattern in various 2 story BMRFs. Sequence of plastic hinge formation*

Bracings

Frame members

1st

2nd

3rd

4th

5th

6th

7th

8th

9th

None

IPB 140 IPB 180 IPB 240 IPB 300 IPB 140 IPB 180 IPB 240 IPB 300 IPB 140 IPB 180 IPB 240 IPB 300 IPB 140 IPB 180 IPB 240 IPB 300 IPB 140 IPB 180 IPB 240 IPB 300 IPB 140 IPB 180 IPB 240 IPB 300 IPB 140 IPB 180 IPB 240 IPB 300 IPB 140 IPB 180 IPB 240 IPB 300 IPB 140 IPB 180 IPB 240 IPB 300

1.2 1.2 1.2 1.2 13 13 13 13,14 13 13 13 13 13 13 13 13 13 13 13 13 1,13 13 13 13 1 13 13 13 1,3,7 1 13 13 1,3 1,3 1 1

5,6 5,6 5,6 5,6 14 14 14 1,2 1 14 14 14 1 1 14 14 1 1 1,14 14 3 1 1 14,1 3,13 1 1 1 13 3,13 1 1 7 7 3 13

11,12 11,12 11,12 11,12 1,2 1,2 1,2 5,6 2 1 1,2 1,2 2,3 2 1 1 3 2 2 1 2 3,2 2 2 2 3 2 2,14 2 2 2,3 2 5 13 7 3

9,10 9,10 9,10 9,10 5,6 5,6 5,6 11,12 3,14 2 3 5,6 4 3 2 2 2 3 3 2 4 4 14 3 4 2 3 3 4 4 4 3 – 2 2 2

– – – – 3,4 3,4 11,12 9,10 4 3 4,5,6 11,12 – 14 3 6 4 4 4 5,6 – – 3 5,6 – 4 14 4,6 – – – 4 – – 13 4

– – – – – – 9,10 – – 6 6 3 – 4 6 5 – 14 6 3 – – 4 4 – – 4 5 – – – 14 – – 4 –

– – – – – – – – – 5 5 10 – – 5 3 – – 5 4 – – 6 – – – – – – – – – – – – –

– – – – – – – – – – – 9 – – 4 4 – – – – – – – – – – – – – – – – – – – –

– – – – – – – – – – – 4 – – – – – – – – – – – – – – – – – – – – – – – –

2UNP30

2UNP40

2UNP80

2UNP100

2UNP120

2UNP140

2UNP2O0

2UNP300

* The numbers 1 to 14 relate to the plastic hinge locations denoted in Figure 6.

of different brace to frame members and shows that structures with inappropriate distribution of strength and stiffness perform poorly under lateral loadings. As a result, six probable collapse mechanisms representing different characteristics of BMRFs, as well as the formation and propagation of plastic hinges are categorized as follows:

Figure 6. Numbering system of the location of plastic hinges referred to in Table 1.

185

i. Both brace and frame members yield. At first, plastic hinges develop in braces; then, frame members in the BMRF system yield in a pattern similar to the corresponding MRF. ii. Same as I, but frame members in the BMRF system do not follow the pattern of the corresponding MRF.

Table 2.

Classification of all probable collapse mechanism and definition of soft story mechanism. Formation of plastic hinges

Examples

Mechanism type

Primary

Secondary

Frame

Brace

I

Braces

Frame members similar to the unbraced MRF.

IPB240 IPB300

2UNP30 2UNP30, 2UNP40

II

Braces

Frame members NOT similar to the unbraced MRF.

III

Brace of 1st story

Frame members and braces of 2nd story

IV

Brace of 1st story

Frame members

V

Frame members

Braces

VI

Frame members

–

IPB140 IPB180 IPB240 IPB300 IPB140 IPB180 IPB240 IPB300 IPB140 IPB180 IPB240 IPB140 IPB180 IPB240 IPB300 IPB140

2UNP30 2UNP30, 2UNP40 2UNP40, 2UNP80, 2UNP100 2UNP80, 2UNP100, 2UNP120 2UNP40 2UNP80, 2UNP100 2UNP120, 2UNP140 2UNP140, 2UNP200 2UNP80, 2UNP100 2UNP120, 2UNP140 2UNP200 2UNP120, 2UNP140, 2UNP200 2UNP200, 2UNP300 2UNP300 2UNP300 UNP300

iii. Same as II, but more effective interaction between the truss and the frame systems develops. The collapse pattern of frame in the BMRF system is completely different to the corresponding MRF. iv. Widespread yielding occurs in frame members, but brace elements only yield in some stories. At first, brace elements in lower stories yield and then many frame members yield individually. v. Same as IV, except that frame members yield before brace elements. vi. Frame members yield but braces remain elastic until ultimate stage. It should be noted that as the brace to frame stiffness ratio increases, the collapse mechanism gradually changes from I to VI. The soft story mechanism due to a large concentrated deformation in an individual story can also develop in BMRF systems. This type of mechanism can occur in categories IV, V and VI. The results obtained from the collapse mechanisms of various multi story BMRFs are incorporated to offer a solution which can be utilized in evaluating the key performance characteristics of such dual system in the form of load contribution shares of brace and frame members. These data are in turn utilized to formulate a relationship for a proper and realistic design procedure.

6

CONCLUSIONS

A rigorous parametric study was performed on a wide variety of single and multi story BMRFs. Sequential

collapse analyses were executed to foresee all probable collapse mechanisms which were classified into six categories. Unsuitable mechanisms, such as frame mechanism (elastic brace behavior or mechanism type VI), the brace mechanism (similar to MRF collapse manner or mechanism type I), and the soft story mechanism (inappropriate full plastic hinge formation in a story or mechanism types IV, V and VI) were evaluated through illustrating an example problem in a 2 story BMRF system. In order to evaluate the load contribution shares an analytical Truss-Frame model accompanied by a correction process in the linear and nonlinear regions was proposed. The model can be used to evaluate the frame contribution share in the linear and nonlinear stages with respect to the relative frame to brace rigidity. REFERENCES ABAQUS. Analysis User’s Manual, version 6.5. Hibbitt, Karlsson and Sorensen Inc. American Institute of Steel Construction. AISC. 2005. Seismic pro-visions for structural steel buildings. AISC/ANSI Standard 341–05, Chicago. Della Corte, G., De Matties, G., Landolfo, R. & Mazzolani, F.M. 2002. Seismic analysis of MRF steel frames on refined models of connection, Journal of Constructional Steel Research, 58(10): 1331–1345. FEMA 273. 1997. NEHRP guidelines for the seismic rehabilitation buildings. Federal Emergency Management Agency. European Committee for standardization, EN 1991-1-1: 2004: Eurocode 1: Actions on Structures.

186

European Committee for standardization, EN 1993-1-1: 2004: Eurocode 3: Design of Steel Structures. European Committee for standardization, EN 1998-1-1: 2004: Eurocode 8: Design of structures for earthquake resistance. Gwozdz, M. & Machowski, A. 1997. Strengthening of building steel frames by modification of structural system, Architectural Civil Engineering, 43(1): 37–49. Lehman, D.E., Roeder, C.W., Herman, D., Johnson, S. & Kotulka, B. 2007. Improved seismic performance of gusset plate connections, ASCE, Journal of Structural Engineering, 134(6): 890–901. Lotfollahi, M. & Alinia, M.M. 2008a. Effect ot Tension Bracing on the collapse mechanism of steel moment frames, Journal of Constructional Steel Research, Submitted for Publication, Under Review.

187

Lotfollahi, M., Mofid, M. & Alinia, M.M. 2008b. BraceFrame interaction in tension braced MRFs. International Conference on Numerical Analysis and Applied Mathematics 2008, Proc., American Institute of physics: 364–367. RAM-Perform 3D. Element Description Manual, Version 1.19, Graham H. Powell Inc. 2003. Roeder, C.W. 2002. Connection performance for seismic design of steel moment frames, ASCE, Journal of Structural Engineering, 128(4): 517–525. Roeder, C., Lehman, D. & Yoo, J.H. 2005. Improved design of steel frame connections. International Journal of Steel Structures, 5(2): 141–153. Uriz, P. & Mahin, S. 2004. Seismic performance of concentrically braced steel frame buildings, Proc., 13th World Congress on Earthquake Engineering. Paper 1639.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

The optimization of the industrial steel building T. Žula & S. Kravanja Department of Civil Engineering, University of Maribor, Maribor, Slovenia

ABSTRACT: The paper presents the structural optimization of a single-storey industrial steel building. The structure is consisted from equal frames which are mutually connected with purlins and rails. All structural elements are proposed to be built up of standard hot rolled I sections. The structural optimization is performed by the Mixed-Integer Non-linear programming approach, MINLP. The MINLP performs a discrete topology, material and standard dimension optimization, while continuous parameters are simultaneously calculated inside the continuous space. Since the discrete/continuous optimization problem of the steel building structure is non-convex and highly non-linear, the Modified Outer-Approximation/Equality-Relaxation (OA/ER) algorithm has been used for the optimization. Alongside the optimal structure mass, the optimal topology, structural steel grade and standard I sections have been obtained. The paper includes the theoretical basis and a practical example with the results of the optimization. 1

INTRODUCTION

Single-storey frame structures are extensively used for industrial, leisure and commercial buildings. In order to obtain efficient frame designs, researchers have introduced various optimization techniques, appropriate either for the continuous or the discrete optimization. O’Brien & Dixon (1997) have proposed a linear programming approach for the optimal design of pitched roof frames. Guerlement et al. (2001) have introduced a practical method for single-storey steel structures, based on a discrete minimum weight design and Eurocode 3 (1992) design constraints. Recently, Saka (2003) has considered an optimum design of pitched roof steel frames with hunched rafters by using a genetic algorithm. One of the latest researches reported in this field is the work of Hernández et al. (2005), where authors have considered minimum weight design of steel portal frames with software developed for structural optimization. It should be noted that all the mentioned authors deal with the discrete sizes optimization at fixed structural topologies. This paper deals with topology, material and standard dimension optimization of the unbraced single-storey industrial steel building structures. The optimization of the portal frames, purlins, rails and secondary facade columns was performed by the Mixed-Integer Non-linear Programming approach (MINLP). The MINLP is a combined discrete and continuous optimization technique. In this way, the MINLP performs the discrete topology (numbers of frames, purlins, rails and secondary facade columns), different material and discrete dimension (standard cross-section sizes of the columns, beams, purlins,

189

rails and secondary facade columns) optimization simultaneously with the continuous optimization of parameters (structure mass, internal forces, deflections, etc.). The MINLP discrete/continuous optimization problems of frames are in most cases comprehensive, non-convex and highly non-linear. The optimization is proposed to be performed through three steps. The first one includes the generation of a mechanical superstructure of different topology, material and standard dimension alternatives, the second one involves the development of an MINLP model formulation and the last one consists of a solution for the defined MINLP optimization problem. The objective of the optimization is to minimize the mass of the single-storey industrial building. The mass objective function is subjected to the set of the equality and inequality constraints known from the structural analysis and dimensioning. The dimensioning of steel members is performed in accordance with Eurocode 3 (1992). The Modified Outer-Approximation/EqualityRelaxation algorithm is used to perform the optimization, see Kravanja & Grossmann 1994, Kravanja et al. (1998a, b, and c). The two-phase MINLP optimization is proposed.

2

MECHANICAL SUPERSTRUCTURE

The paper presents the topology, material and standard dimension optimization of the unbraced singlestorey industrial building steel structures, see Figure 1. Columns, beams, purlins, rails and secondary facade

er

H

er

q

Lf

P

Lf

P

Lf

P

Lf

LL

Lf

ec ec ec ec ec ec ec e c e c ec

Lf

L Figure 1.

Single-storey industrial building.

columns are proposed to be built up of standard hot rolled steel I sections. The considered portal frame structures are optimized under the combined effects of the self-weight of frame members, vertical uniformly distributed surface variable load (snow) and horizontal surface variable load (wind). The purlins as well as rails and secondary facade columns are designed to transfer permanent load (self-weight and weight of panels) and variable load (snow and wind) to the frame structure. Internal forces are calculated by the elastic first-order analysis. The dimensioning of steel members is performed in accordance with Eurocode 3 for the conditions of both ultimate limit state and serviceability limit state. When the ultimate limit state of structural members is considered, the elements are checked for axial resistance, shear resistance, bending moment resistance, interaction between bending moment and axial force, interaction between axial compression/buckling and buckling resistance moment. Considering the serviceability limit state, the vertical deflections of beams, purlins and rails are calculated by the force method. The total deflections δmax , subjected to the overall load, and the deflections δ2 , subjected to the variable imposed load, are calculated to be smaller than limited maximum values: span/200 and span/250, respectively. The horizontal deflections are also checked for the recommended limits: the relative horizontal deflection  should be smaller than the height/150 for portal frames, and span/150 for rails and the secondary facade columns.

3

MINLP MODEL FORMULATION FOR MECHANICAL SUPERSTRUCTURE

It is assumed that a general non-linear and non-convex continuous/discrete optimization problem can be formulated as an MINLP problem in the form: min z = cT y + f (x) s.t. : h(x) = 0 g(x) ≤ 0 (MINLP) By + Cx ≤ b x ∈ X = {x ∈ Rn : x lo ≤ x ≤ x up } y ∈ Y = {0, 1}m where x is a vector of continuous variables specified in the compact set X and y is a vector of discrete, binary 0–1 variables. Functions f (x), h(x) and g(x) are nonlinear functions involved in the objective function z, equality and inequality constraints, respectively. All functions f (x), h(x) and g(x) must be continuous and differentiable. Finally, By + Cx ≤ b represents a subset of mixed linear equality/inequality constraints. The above general MINLP model formulation has been adapted for structural optimization. It is postulated that it helps us construct an MINLP mathematical optimization model for any structure. In the context of structural optimization, continuous variables x define structural parameters (dimensions, strains, stresses, costs, mass...) and binary variables y represent the potential existence of structural elements within the defined superstructure. An extra binary

190

variable y is assigned to each structural element. The element is then selected to compose the structure if its subjected binary variable takes value one (y = 1), otherwise it is rejected (y = 0). Binary variables also define the choice of discrete/standard materials and sizes. The mass (economical) objective function z involves fixed mass (cost) in the term cT y, while the dimension dependant mass (cost) are included in the function f (x). Non-linear equality and inequality constraints h(x) = 0, g(x) ≤ 0 and the bounds of the continuous variables represent the rigorous system of the design, loading, resistance, stress, deflection, etc. constraints known from the structural analysis. Logical constraints that must be fulfilled for discrete decisions and structure configurations, which are selected from within the superstructure, are given by By + Cx ≤ b. These constraints describe relations between binary variables and define the structure’s topology, materials and standard dimensions. It should be noted, that the comprehensive MINLP model formulation for mechanical structures may be found elsewhere Kravanja et al. (1998a, b, c, 2005). 4

models that may cut off the global optimum. In order to reduce undesirable effects of non-convexities the Modified OA/ER algorithm was proposed by Kravanja & Grossmann (1994), see also Kravanja et al. (1998a) by which the following modifications are applied for the master problem: the deactivation of linearization’s, the decomposition and the deactivation of the objective function linearization, the use of the penalty function, the use of the upper bound on the objective function to be minimize as well as the global convexity test and the validation of the outer approximations. The optimal solution of complex non-convex and non-linear MINLP problem with a high number of discrete decisions is in general very difficult to be obtained. The optimization is thus proposed to be performed sequentially in two different phases to accelerate the convergence of the Modified OA/ER algorithm. The optimization starts with the topology optimization of a structure, while discrete dimensions are temporary relaxed into continuous parameters. When the optimal topology is found, discrete/standard dimensions are in the second phase re-established and the simultaneous topology and discrete dimension optimization of the structure is then continued until the optimal solution is found.

SOLVING THE MINLP PROBLEM

After the MINLP model formulation is developed, the defined MINLP optimization problem is solved by the use of a suitable MINLP algorithm and strategies. Since the MINLP optimization problems of steel frame structures are in most cases comprehensive, non-convex and highly non-linear, the Outer Approximation/Equality-Relaxation (OA/ER) algorithm by Kocis & Grossmann (1987) is selected for the optimization. The OA/ER algorithm consists of solving an alternative sequence of Non-linear Programming (NLP) optimization subproblems and Mixed-Integer Linear Programming (MILP) master problems. The former corresponds to continuous optimization of parameters for a mechanical structure with a fixed topology (and fixed discrete/standard dimensions) and yields an upper bound to the objective to be minimized. The latter involves a global approximation to the superstructure of alternatives in which a new topology and discrete/standard dimensions are identified so that its lower bound does not exceed the current best upper bound. The search of a convex problem is terminated when the predicted lower bound exceeds the upper bound, otherwise it is terminated when the NLP solution can be improved no more. The OA/ER algorithm guarantees the global optimality of solutions for convex and quasi-convex optimization problems. The OA/ER algorithm as well as all other mentioned MINLP algorithms do not generally guarantee that the solution found is the global optimum. This is due to the presence of non-convex functions in the

191

5

NUMERICAL EXAMPLE

The numerical example presents the MINLP topology, material and standard dimension optimization of a single-storey industrial steel building. The building is 22 meters wide, 44 meters long and 8.0 meters height. The structure is consisted from equal non-sway steel portal frames which are mutually connected with the purlins and rails. The portal frame is subjected to the self-weight and the surface variable load. Variable imposed load 2,0 kN/m2 (snow) and 1,0 kN/m2 (horizontal wind) are defined as the uniformly distributed surface load in the model input data. Both, the horizontal uniformly distributed load and the vertical uniformly distributed line load on the beams are calculated considering the intermediate distance between the portal frames. The portal frame superstructure was generated in which all possible structures were embedded by 60 portal frame alternatives, 50 purlin and 20 rail alternatives with different material and standard sizes variation. The superstructure comprised 18 different standard hot rolled European I sections, i.e. IPE sections (from IPE 80 to IPE 600), 3 different structural steel grades (S235, S275, S355) for each column, beam, purlin, rail and secondary facade column separately. The optimization was carried out by a user-friendly version of the MINLP computer package MIPSYN, the successor of PROSYN Kravanja & Grossmann (1994) and TOP Kravanja et al. (1992). As an interface

.3 x3

8.0 m

4x1.87 m 0.5 m

13

8m

44

.0

m

10 x 2.20 m 22.0 m

0.5 m

Optimal topology of the single-storey industrial building. IP E 1 2 0

IPE 550

IPE 550

8.0 m

IPE 120

IP E 5 0 0

IP E 5 0 0

4 x 1.87 m

Figure 2.

10 x 2.20 m 22.0 m

Figure 3.

Optimal steel sections.

for mathematical modeling and data inputs/outputs GAMS (General Algebraic Modeling System), a high level language by Brooke et al. (1988), was used. The Modified OA/ER algorithm and the two-phased optimization were applied, where GAMS/CONOPT2 Generalized Reduced-Gradient method Drudd (1994) was used to solve NLP sub problems and GAMS/Cplex (2001), Branch and Bound was used to solve MILP master problems. The optimization model contained 218 (in) equality constraints, 274 continuous and 400 binary variables. The final optimal solution of 73.42 tons was obtained in the 4th main MINLP iteration. The optimal result represents the mentioned optimal structure mass of 73.42 tons, the obtained optimal topology of 14 portal frames, 12 purlins with the same number of the secondary facade columns and 10 rails see Figure 2. Gained were also the optimal standard sizes of columns, beams, purlins, rails and the secondary facade columns see Figure 3. The optimal structural steel S355 was calculated.

6

CONCLUSIONS

The paper presents the Mixed-Integer Non-linear Programming approach (MINLP) to structural optimization. The Modified OA/ER algorithm and the two-phase MINLP optimization strategy were applied. The optimization is performed by a user-friendly version of the MINLP computer package MIPSYN. Beside the optimal structure mass, the optimal topology with the optimal number of structural elements, the optimal structural steel and the optimal standard I sections are obtained simultaneously. The example, presented at the end of the paper, clearly show the efficiency of the proposed MINLP approach.

REFERENCES O’Brien, E.J., Dixon, A.S. 1997. Optimal plastic design of pitched roof frames for multiple loading. Comput. Struct. 64: 737–740.

192

Gurlement, G., Targowski, R., Gutkowski, W., Zawidzka, J., Zawidzki, J. 2001. Discrete minimum weight design of steel structures using EC3 code. Struct. Multidisc. Optim. 22: 322–327. Eurocode 3, 1992. Design of steel structures, European Committee for Standardization. Saka, M.P. 2003. Optimum design of pitched roof steel frames with haunched rafters by genetic algorithm. Comput. Struct. 81: 1967–1978. Hernández, S., Fontán, A.N., Perezzán, J.C., Loscos, P. 2005. Design optimization of steel portal frames. Adv. Eng. Software. 36: 626–633. Kravanja, Z., Grossmann, I.E. 1994. New Developments and Capabilities in PROSYN—An Automated Topology and Parameter Process Synthesizer. Computers chem. Eng., 18: 1097–1114. Kravanja, S., Kravanja, Z., Bedenik, B.S. 1998a. The MINLP optimization approach to structural synthesis. Part I: A general view on simultaneous topology and parameter optimization. Int. J. Numer. Methods Eng. 43: 263–292. Kravanja, S., Kravanja, Z., Bedenik, B.S. 1998b. The MINLP optimization approach to structural synthesis. Part II: Simultaneous topology, parameter and standard dimension optimization by the use of the Linked two-phase MINLP strategy. Int. J. Numer. Methods Eng. 43: 293–328.

Kravanja, S., Kravanja, Z., Bedenik, B.S. 1998c. The MINLP optimization approach to structural synthesis. Part III: Synthesis of roller and sliding hydraulic steel gate structures. International Journal for Numerical Methods in Engineering, 43: 329–364. Kravanja, S., Kravanja, Z., Bedenik, B.S., Faith, S. 1992. Simultaneous Topology and Parameter Optimization of Mechanical Structures. Proceedings of the First European Conference on Numerical Methods in Engineering, Brussels, Belgium, ed. C. Hirsch et al., Elsevier, Amsterdam, pp. 487–495. Kravanja, S., Šilih, S., Kravanja, Z. 2005. The multilevel MINLP optimization approach to structural synthesis: the simultaneous topology, material, standard and rounded dimension optimization. Adv. eng. softw., 36 (9): 568–583. Kocis, G.R., Grossmann, I.E. 1987. Relaxation Strategy for the Structural Optimization of Process Flowsheets. Ind. Engng Chem. Res., 26: 1869–1880. Brooke, A., Kendrick, D., Meeraus, A. 1988. GAMS—A User’s Guide, Scientific Press, Redwood City, CA. Drudd, A.S. 1994. CONOPT—A Large-Scale GRG Code. ORSA J. Comput. 6: 207–216. CPLEX 2001. User Notes, ILOG inc.

193

Composite structures

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

A finite element model for double composite beam S. Duan, R. Niu, J. Xu & H. Zheng Research Institute of Structural Engineering, Shijiazhuang Railway Institute, Shijiazhuang, P.R. China

ABSTRACT: Double steel-concrete composite continuous beam is a new structural system developed on the basis of single steel-concrete composite beam. The mechanical properties of the composite beam obviously depend on their respective properties and interactions. The stiffness matrix of a double composite beam element was yielded for the negative applied bending moment area, in which the interface slips can be taken into account. A structural analysis method for the double composite continuous beam was presented, and a structural computing program was compiled. At last, as a numerical example, a two-span continuous beam specimen is investigated in detail. The load deflection curve is given, which verified with the experimental data and good agreement is obtained. It is shown that the proposed method is correct and convenient for the analysis of the double steel-concrete composite continuous beam.

1

2

INTRODUCTION

A double steel-concrete composite continuous beam (Reiner 1996; Stroh & Sen 2000) is a new structural system developed on the basis of single steel-concrete composite beam, in which there is also a bottom reinforced concrete slab connected to a steel profile in the negative moment regions through the shear connectors, therefore with two interfaces. Comparing with the traditional single steel-concrete composite continuous beam, its advantage is that effectively limits the crack width of the negative moment area, and also improve the stress state of section, so that it is suitable to the composite continuous beam with a larger span. The mechanical properties of the double composite beam obviously depend on their respective properties and interactions. In the negative applied bending moment area, the concrete slab cracks under tension and then the interface slip occurs between steel profile and concrete slab, with non-linear features, it makes great impact on the structure of the internal forces and deformation. Therefore, it is necessary to establish a finite element model to study the mechanical properties of the double steel-concrete composite beam in negative moment regions. Based on the partial collaboration theory (Newmark et al. 1951), the elastic stiffness matrix in the negative moment region for a double composite beam element is derived and used to study the double composite continuous beam actions in this paper.

197

PROBLEM STATEMENT AND CROSSSECTION BALANCE EQUATION

An example of a double steel-concrete composite continuous beam modal and its cross-section shape are shown in Figure 1. 2.1

Basic assumptions

Similar to the characteristics of single steel-concrete composite beam under working load, the steel profile and the tension reinforcement in the double composite beam behaves elastically, tensile concrete slab

P

1

P

1

1

1

(a) proposed structure

(b) cross-section for positive moment region

Figure 1.

(c) cross-section for negative moment region

Structure and its cross-section shape (mm).

out of work due to cracking, the concrete strain in compressed area is in the straight-line stage of the stress–strain relation curve. So the composite beam will be considered as an elastic body under working load, and bases the following assumptions: 1. Steel profile, reinforcements and concrete in compression are in isotropic elastic, the tensile strength of concrete and the cohesive strength along the interface between concrete and steel may be neglected. 2. During the loading to failure of the composite beam, the strain distribution over the depth of the top concrete flange, the steel profile and the bottom concrete flange is respectively assumed to be linear. 3. Horizontal shear force is proportional to the relative slipping value between steel profile and concrete slab, with the vertical separation will be neglected. 4. The slip between reinforcement and concrete slab is not considered; 5. Shear connectors between concrete slab and steel profile are in elastic phase and in uniformly distribution along the beam length.

2.2

with the tensile stiffness 1 1 1 + + Ec Ac Es As Er Ar 1 1 = + E A E EA1 c c s As 1

EA 1

=

in which, EI = elastic  bending stiffness of composite cross-section; EI = elastic bending stiffness of cross-section with no combination; EA = tensile stiffness of composite cross-section; and EA1 = tensile stiffness of cross-section with combination by the steel profile and the bottom concrete slab. The strains εtb at the bottom surface of the top concrete slab and εtt at the top surface of the steel profile as follows: εtb =

Tr − φyr Er Ar

εtt = φys +

The force acting on the small element of the double steel-concrete composite beams in negative moment regions and its deformation are shown in Figure 2. By the mechanics of materials, we can define the elastic bending stiffness of the cross-section for the beam in the case of double composite and without composite respectively as following.

in which, φ=

Mc Ms = E s Is E c Ic

εct =

Ts − φ(hs − ys ), Es As

εcb =

2 EI = EI + EA1 · dsc  EI = Ec Ic + Es Is + Er Ir

Ts Tr − Es As Er Ar

dx

Ms

yr

Tr+dTr

Vs sc

Mc

Vc st+

ys

Ts

Dc

V

rc vc

(2)

(3)

The relative slip strain of bottom interface can be obtained from Equation 2. εsc = sc = εcb − εct = φdsc −

rtdx vtdx

φhc Dc − 2 Ec Ac

Then, the relative slip strain of the upper interface can be obtained from Equation 1. εst = st = εtt − εtb = φdrs +



st

(1)

The strains εct at the bottom of the steel profile and εcb at the top surface of the bottom concrete slab as follows:

Differential equation of equilibrium

Tr Vc

Ts Es As

Ms+dMs

VsTs+dTs sc+

Mc+ dMc Dc+dDc

V

Figure 2. Force acting on a small element of the double composite beam and its deformation.

Dc Ts − Ec Ac Es As

(4)

From Figure 2 and Figure 3, we can write the equilibrium equation of the bending moment as M = MS + MC + Tr drs + Dc dsc + Ts ds = φ(Ec Ic + Es Is ) + Tr drs + Dc dsc + Ts ds

(5)

where drs = distance from the centre of the reinforcements cross-section to the centre of the steel profile cross-section, drs = yr + ys ; dsc = distance from the centre of the steel profile cross-section to the center of the compressive concrete slab cross-section,

198

yr

u

u

St

ys

u

where η = shear modulus of the interface contact, η = K/a, η = 0 for an overlapping beam and η = ∞ for a full composite beam.

3

x

3.1

u

u

Cross-section strain of small element of the beam.

dsc = hs − ys + hc /2; ds = distance from the centre of the steel profile cross-section to Ts . By the basic assumptions, each shear connector subjected to the shear force Q = V ·a, where V = horizontal shear force in unite length; a = space between the shear connectors. Hence, we have εst = −

a d 2 Tr · K dx2

(6)

a d 2 Dc εsc = − · K dx2

Substituting Equations 3 and 4 into Equation 6, the equilibrium equations of the axial forces for top and bottom interfaces can be obtained Ts Tr a d 2 Tr − =− · φdrs + Es As Er Ar K dx2 φdsc −

2

Dc Ts a d Dc − =− · Ec Ac Es As K dx2

Element stiffness matrix

The nodal force and deformation of the beam element is shown in Figure 4, the bending moment of arbitrary cross-section can be expressed by the bending moment M1 and shear force Q1 as:

S

dx Figure 3.

ELEMENT ANALYSIS

u

(7)

Combination the both of Equation 7, the equilibrium equation of the axial forces for the composite beam cross-section is,

M = M1 − Q 1 x

Substituting the above expression into Equation 9, then d 2y d4y − a2 2 − β(M1 − Q1 x) = 0 4 dx dx

where α and β are the constants related to the used materials, the cross-sectional size and the shear stiffness in the interface. α2 =

ηEI 

EA ·

+

+

η , Er Ar

β=

EA ·

η 

EI

(11)

y = C1 cosh ax + C2 sinh ax −

β (M1 − Q1 x) (12) α2

Then, the deformation equation can be obtained by integration of Equation 12

l 1

2

x

2

a d Tr a d Dc · + · =0 K dx2 K dx2

(8)

Finally, the equilibrium differential equation of the double composite beam is yielded as following   ηEI 1 d 2M η d4y d 2y  − + +  dx4 Er Ar dx2 EI dx2 EA · EI ηM − =0  EA · EI

EI

Solving the Equation 9, we obtain

Mc Ms Dc Tr dsc + drs − − Ec Ac Er Ar Ec Ac Er Ar 2

(10)

M1

1

M2 1 2

x y

(9) Figure 4.

199

Beam element.

2

2

2

y=

C1 C2 cosh ax + 2 sinh ax 2 a a   1 β 1 2 M1 x − Q1 x3 + C3 x + C4 − 2 α 2 6



M2 = (13)

By the following boundary deformation conditions:

EI a3 sinh al · 6(δ1 − δ2 ) Ks    α2 3l EI a3 sinh al − + θ1 Ks lβ    3l EI a3 sinh al α2 + + θ2 Ks lβ

M1 + M2 Q1 = −Q2 = l  EI a3 sinh al [12(δ1 − δ2 ) + 6l(θ1 + θ2 )] = Ks l



y=0 y  = θ1  y = δ2 − δ1 x = l,  y = θ2 x=0

C1 =

1 a2 D1 [hal cos(hal)θ1 − halD2 − sin(hal)θ1 + D3 ha]

where    2 α2  3 − EI Ks = EI a sinh al + 2 l β

× (12 + 6al sinh al − 12 cosh al)

C2 =

a2 (−lD2 + D3 + θ1 hal sinh al + cosh alθ1 − θ1 ) D1

Finally, the stiffness equation of the composite beam element can be expressed by,

C3 =

−h sin2 (hal)aθ1 + [h2 sin(hal)a2 − cos(hal)ha]D3 D1

s|x=0 = 0 C1 − C4

C4 =

and s|x=l = 0 can be determined

−

[ha cos2 (hal) + cos(hal)]θ1 D1

+

D2 [sin(hal) + ha cos(hal) − ha] D1

⎡ 12 4ϕ2 + 2ϕ3 ϕ1 ⎢ l2 l ⎢ ⎢ 6 ⎢ ϕ 4ϕ2 ⎢ l 1 α2 ⎢ ⎢ 12 4ϕ + 2ϕ3 lβ ⎢ ⎢− ϕ1 − 2 ⎢ l2 l ⎢ ⎢ 6 ⎣ 2ϕ3 ϕ1 l ⎧ ⎫ ⎧ ⎫ ⎪ ⎪ ⎪ ⎪Q1 ⎪ ⎪ ⎪δ1 ⎪ ⎪ ⎬ ⎪ ⎨M ⎪ ⎬ ⎨θ ⎪ 1 1 = ⎪ δ2 ⎪ ⎪ ⎪ ⎪Q 2 ⎪ ⎪ ⎪ ⎪ ⎭ ⎪ ⎩ ⎪ ⎭ ⎩ ⎪ θ2 M2

1 D1 [hal cos(hal)θ1 − halD2 − sinh alθ1 + haD3

where: D1 = h a l sin(hal) − ha cos(hal) + sin(hal) 2 2

+ ha cos(hal) − ha   β 1 1 D2 = δ2 − δ1 + 2 M1 l 2 − Q1 l 3 α 2 6   β 1 D3 = θ2 + 2 M1 l 2 − Q1 l 2 α 2 Solving the Equation 10, we obtain: 

M1 =

EI a3 sinh al · 6(δ1 − δ2 ) Ks    3l EI a3 sinh al α2 + + θ1 Ks lβ    3l EI a3 sinh al α2 + − θ2 Ks lβ

(14)

12 4ϕ2 + 2ϕ3 ⎤ ϕ1 2 ⎥ l l ⎥ ⎥ 6 ⎥ ϕ − 1 2ϕ3 ⎥ l ⎥ ⎥ 4ϕ2 + 2ϕ3⎥ 12 ⎥ ϕ − 1 ⎥ l2 l ⎥ ⎥ 6 ⎦ − ϕ1 4ϕ2 l

−

(15)

The above equation can be symbolically written as [k]{δc } = {fc }

(16)

where, [k] is the stiffness matrix, in which the interface slips can be taken into account, and  EI (al)3 sinh al ϕ1 =   2 EI (al)3 sinh al + 2( αβ − EI ) ×

1 (12 + 6al sinh al − 12 cosh al)

3 ϕ1 + 4 3 ϕ3 = ϕ1 − 2 ϕ2 =

200

1 4 1 2

3.2

Equivalent nodal force

as:

In finite element analysis, the non-joint loads should be transformed into equivalent nodal force. Two cases are considered in this paper. Figure 5 shows a double composite beam with both fixed ends subjected to the uniformly distributed force in the span. As we know, Q1 = Q2 =

1 ql, 2

M1 = −M2

C1 C2 cosh ax − 2 sinh ax a2 l a l   β 1 1 qx2  − + 2 a2 α2 EI   β 1 C2 1 1 Q1 x3 − M1 x2 − qx 4 + 2 + 2 lα 6 2 24 a l

C3 = −

C4 = −

C1 a2

The bending moment about the arbitrary location x is

Similarly, C1 and C2 can be derived from the interface slips condition,

1 M = Q1 x − M1 − qx2 2

s|x=0 = 0,

s|x=l = 0

as:    β β ql 1 1   − 2 = − 2 EI EI α 2a α   1 β ql(1 + cosh al)  − 2 C1 = 2a sinh al α EI 

Substituting above expression into Equation 9, we have:

C2 =

d4y 1 d2y − a2 2 −  q 4 dx dx EI   1 − β Q1 x − M1 − qx2 = 0 2

In addition, from the symmetrical deformation condition θ|x=0 = 0, we have C2 h + C3 = 0 a

Solving the above equation, we obtain y=

C1 C2 cosh ax + 2 sinh ax a2 a   β 1 qx2 1  − − EI 2 a2 α2   β 1 1 1 Q1 x3 − M1 x2 − qx 4 + C3 x + C4 − 2 α 6 2 24

Hence, finally: M1 = −M2 =

in which, C3 and C4 can be derived from the vertical displacement condition, y|x=0 = 0,

Q1 a

y|x=l = 0

1 2 ql 12

The results show that the equivalent nodal force for a uniformly distributed force acted at steel-concrete composite beam is not relative to interface shear stiffness, and it is the same as a common beam. Similarly, the equivalent node force for a concentrated force acted at an arbitrary location (see Figure 6) is:

y

y

p

q

M2

M1

M2

M1

x

x a

l 1

Figure 5.

b

2

A beam subjected to uniformly distributed force.

201

Figure 6.

A beam subjected to a concentrated force.

Table 1.

Parameters of the composite beam (unit: mm).

Steel-concrete composition beam

Studs Studs distance in distance Reinforcement Top Top Bottom Bottom negative moment Studs in positive slab slab slab slab region of distance in moment ① ② width depth width depth top slab bottom slab region

Huo et al. (2008) Zhou (2008) Double composite beam Single composite beam

7 8 5 8 7 8 5 8 7 8 5 8 7 8 –

80 80 80 80

100

Load/kN

150

200

600 600 600 600

present

50

Duan et al Ansys

0

single composite beam

0

1

2

3

4

5

6

7

Deflection/mm

Figure 7.

M1 = −

Load-deflection curves.

pab2 l2

M2 =

pb2 (l + 2a) Q1 = l3

pba2 l2

pa2 (l + 2b) Q2 = − l3

The results show that the equivalent nodel force for a symmetrical force acted in beam is not relative to the interface shear stiffness, it can be calculated by reference to the common beam. But for an unsymmetrical force, the equivalent nodel force is relative to the interface shear stiffness. 4

NUMERICAL EXAMPLES

A double steel-concrete composite beam structure and cross-section shape is shown in Figure 1, and the relative parameters are shown in Table 1. The obtained load-defecation curves are demonstrated in Figure 7. The present curve is very close to the results by the test (Duan et al. 2008) and by the Ansys software (Duan et al. 2007). The deflection of the double composite beam is deduced about 25% due to adding the bottom concrete flange. 5

CONCLUSION

350 350 350 –

80 80 80 –

242.5 235 235 235

84.6 84.6 85 –

115 135 115 115

can be taken into account. The proposed finite element method for the analysis of the double steel-concrete composite continuous beam is reliable. The obtained results show that the bending moment and the curvature is not in linear, that being related to the slipping value and distribution. The elastic bending stiffness of the cross-section for a double composite beam is also a variable along the beam length duo to the shear connector stiffness. By the example, the deflection is significantly reduced due to adding the bottom concrete slab, the stress state of the cross-section being improved in the negative moment regions over the interior piers. ACKNOWLEDGEMENTS The work in this paper was supported by the Natural Science Foundation of Hebei Province, China (Contract No. E2008000397). The authors would also like to thank the people who contribute the study in Shijiazhuang Railway Institute. REFERENCES Duan, S.J. Huo, J.H. & Zhou, Q.D. 2007. On calculation method of the ultimate bearing capacity of double steel-concrete composite beam (in Chinese). Journal of Shijiazhuang Railway Institute 20(4):1–4. Duan, S.J. Zhou, Q.D. et al. 2008. Experimental study on bearing capacity of double steel and concrete composite continuous beams (in Chinese). Journal of Railway Science and Engineering 5(5): 12–17.’’ Newmark, N.M. Siess, C.P. & Viest, I.M. 1951. Test and analysis of composite beams with incomplete interaction. Proceeding, Society of Experimental Stress Analysis 9:75–92. Reiner, S. 1996. Bridges with double composite action. Structural Engineering International 1: 32–36. Stroh, S.L. & Sen, R. 2000. Steel bridge with doublecomposite action: innovative design. In: 5th International Bridge Engineering Conference, April 3–5, 2000, Tampa, FL, United States. Transportation Research Record 1(1696): 299–309.

In this paper, the stiffness matrix of a double composite beam element is yielded, in which the interface slips

202

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

A new composite element for FRP-reinforced concrete slabs Y.X. Zhang School of Aerospace, Civil and Mechanical Engineering, The University of New South Wales, Australian Defence Force Academy, Canberra, Australia

Y. Zhu School of Civil Engineering, The University of Guang Zhou, Guangzhou, China

ABSTRACT: A newly developed composite layered plate element is presented in this paper and employed for finite element modeling of nonlinear structural behavior especially the cracking progress of FRP-reinforced slabs. The layered plate element is constructed based on Mindlin–Reissner plate theory and Timonshenko’s beam function. Both geometric nonlinearity and material nonlinearity incorporating tension, compression, concrete cracking and tension stiffening, are included in the nonlinear finite element analysis. Cracking progress of a clamped GFRP-RC slab and a simply-supported one-way FRP-RC slab is analyzed using the developed nonlinear finite element modeling technique in this paper. 1

INTRODUCTION

Fiber-reinforced polymers (FRPs), such as carbon fiber reinforced polymers (CFRPs) and glass fiber reinforced polymers (GFRPs) have been applied increasingly in concrete structural members as an internal reinforcement including bridges, parking garages, multi-storey buildings and other industrial structures. In comparison with the traditional reinforcement of steel, FRPs have lower weight, higher tensile capacity and higher resistance to corrosion. The characteristics of high corrosion resistance of FRPs make them especially suitable for use in concrete structures subjected to aggressive environmental conditions, and thus the high costs of repair and maintenance due to corrosion of steel reinforcement in concrete structures can be reduced significantly. However due to the relatively low modulus of elasticity of the FRPs, the FRP reinforced concrete member may undergo large deflection and suffer less margin of safety in comparison to that of its counterpart concrete element reinforced by steel bars and this reduces the serviceability of the flexural members. Thus the structural behaviour of the FRP reinforced concrete members is needed to obtain sufficient information for design to assure adequate safety and serviceability of the FRP concrete structures, especially the ductile behaviour and cracking progress process. A simple shear-flexible rectangular layered FRPreinforced composite plate element has been developed recently based on Mindlin–Reissner plate theory and Timoshenko’s composite beam functions (Zhang & Kim 2004) and has been validated to be efficient

203

and accurate for nonlinear finite element analysis of deformation behavior of FRP-reinforced concrete slabs (Zhang & Zhu 2009). The development element is used to analyze the cracking progress behavior of the FRP-reinforced concrete slabs in this paper. Both geometric nonlinearity and material nonlinearity incorporating tension, compression, concrete cracking and tension stiffening, are included in the nonlinear finite element analysis. 2

A LAYERED RECTANGULAR FRP-REINFORCED COMPOSITE PLATE ELEMENT

A layered rectangular FRP-reinforced composite plate element as shown in Figure 1 has been developed recently (Zhang & Zhu 2009) to analyse the deflection behaviour of FRP-reinforced concrete slabs. In the layered rectangular FRP-reinforced composite plate element, the composite cross section is divided into a series of concrete layers and the FRP reinforcement is treated as equivalent smeared FRP layers with the equivalent thickness of the FRP layer tFRP = AFRP /b, where AFRP is the area of one reinforcing FRP bar and b is the spacing of the bars. Each concrete layer and smeared reinforcement layer is assumed to be in a state of plane stress, and compatibility between the reinforcement layers and the concrete layers is assumed to be maintained through the analysis. The material properties for the element are calculated by summing algebraically the material property of each layer, which is assumed to be constant throughout the thickness of the layer.

Middle surface of element FRP,1

ZC,1 ZC,2 .

FRP,1

FRP-bar layers

Concrete layers

Figure 1.

The layered rectangular FRP-reinforced concrete element.

In the FRP-reinforced composite plate element, the bending strain matrix is obtained from the deflection and rotation functions based on Timoshenko’s composite beam functions (Zhang & Kim 2004), and plane displacement of a quadrilateral isoparametric element with drilling degrees of freedom (MacNeal & Harder 1988) are employed to calculate the membrane strain. Shear deformation effect is also included in the composite element and the transverse shear strain matrices employed in the reference (Zhang et al. 2007) will be used for the FRP-RC plate element.

3

NONLINEAR MATERIAL MODEL

The employment of an accurate material model plays a very important role in the finite element modeling of the structural behavior of reinforced concrete slabs, especially for the cracking progress process. Concrete cracking is one of the main reason for material nonlinearity, thus needs to be described accurately. In the layered finite element model, before cracking or crushing, concrete is assumed to be isotropic and linear elastic. The maximum principal stress criterion is used to detect concrete cracking and a fixed smeared crack approach is used to model the cracking once it has been identified. Once the maximum principal stress at a Gauss points reaches the concrete tensile strength, cracks are assumed to form in planes perpendicular to the direction of the maximum principal tensile stress, and the elastic modulus and Poisson’s ratio are reduced to zero in the maximum principal stress direction. Assuming that ‘1’ and ‘2’ are the directions of maximum and minimum principal stress respectively, when the maximum principal stress in direction 1 reaches the concrete tensile strength ft , the in-plane material property matrix in the principal coordinate system becomes

c Dc = diag[0 Ec G12 ]

(1)

When the minimum principal stress also reaches ft , a second crack plane perpendicular to the first one is assumed to form, and the in-plane property matrix then becomes Dc = diag[0 0

1 c G ] 2 12

(2)

After cracking, the out-of-plane material property matrix in the principal coordinate system becomes c c Dco = diag[G13 , G23 ]

(3)

c c c where G12 , G13 and G23 are the cracked shear modulus that accounts for aggregate interlock and dowel action in the smeared cracking model. In the present formulation, the crack shear modulus presented by Cedolin and Deipoli (Cedolin & Deipoli, 1977) is chosen, and it is assumed to be a function of the tensile strains. For concrete cracked in the maximum principal axis c is direction, the shear modulus G12

c G12

=

⎧   ⎨0.25 × G × 1 − ε1 /0.004 ⎩

ε1 < 0.004 ε1 ≥ 0.004 (4)

0

while for concrete cracked in both principal axis direcc tions, the shear modulus G12 is

c = G12

204

⎧ c ⎨0.5 × G13

c c G23 ≥ G13

⎩ c 0.5 × G23

c c G23 < G13

(5)

and   0.25 × G × 1 − ε1 /0.004 0

 c G13 =

where Ae is the original area of the element, D is the property matrix, B is the strain matrix of the element given by

ε1 < 0.004 ε1 ≥ 0.004

B = Bl + Bnl ,

(6)    0.25 × G × 1 − ε1 /0.004 c G23 = 0

ε2 < 0.004 ε2 ≥ 0.004

in which the linear component Bl can be expressed as (7)

Bl = [Bl1 , Bl2 , Bl3 , Bl4 ]

where ε1 and ε 2 are the principal tensile strains in the directions 1 and 2 respectively, and G is the elastic shear modulus of the concrete. The following formulation is used to represent the tension stiffening effects after cracking of concrete (Izumo et al., 1992)  σi =

εcr ≤ εi ≤ 2εcr

ft

(2εcr /εi )

0.4

with ⎡

Blmi ⎢ l Bi = ⎣ 0 0

(8)

εi > 2εcr ,

× ft

⎤ 0 ⎥ Bbi ⎦ Bsi

(i = 1, . . . 4)

(13)

whereBlmi , Bbi , Bsi (i = 1, . . . 4) is the components of membrane strain, bending strain and shear strain. Nonlinear strain matrix component is expressed as

in which εcr = ft /Ec and σi and εi (i = 1, 2) are the principal stress and strain respectively in the i direction, and ft is the tensile strength of concrete. The FRP reinforcement is assumed to be linearly elastic in tension and compression, with axial stiffness only in the bar direction. But when the tension stress of FRP bar reaches the material ultimate strength, the brittle rupture is occurred, and the stress will be reduced to zero immediately.

4

(12)

nl nl nl Bnl = [Bnl 1 , B2 , B3 , B4 ]

with ⎡

⎤ 0 Bnl mi ⎢ ⎥ Bnl 0 ⎦ i = ⎣0 0 0

(i = 1, . . . 4) .

(14)

NONLINEAR FINITE ELEMENT ANALYSIS The nonlinear membrane strain matrices Bnl mi used in the reference (Zhang et al., 2007) will be used for the FRP-RC plate element.

The finite element formulation based on U.L. approach can be written as (K + Kσ ) q = Pt+t − R t where K= Ke ,

(9) 4.2

Kσ =

e



Keσ ,

e

Rt =



Ret

(10)

Element geometric stiffness matrix

In component form, the element geometric stiffness matrix can be written as 

e

and Pt+t is the external load vector at time t + t in the analysis. The element stiffness matrix K e , the element geometric stiffness matrix Keσ and the element internal force vector R te are defined as follows.

σ Ke ij

=

Kmij

σ

(15)

GiT σL Gj dAe

(16)

 Nxy . Ny

(17)

Ae

BT DB dAe

0





Kmij =

The element stiffness matrix K e is expressed as: Ke =

0

where σ

4.1 Element stiffness matrix

0

(11)

Ae

205

 σL =

Nx Nyx

4.3

Element internal force vectors

The element internal force vector is given by

Ret = BT σ dAe

et al. (Zhang et al., 2004) was analysed using the nonlinear finite element modelling technique. The slab was 1000 mm wide and 250 mm thick, with a clear span of 3000 mm, and the slab was reinforced in the tension zone with NEFMAC C16 type CFRP grid, having an actual reinforcement ratio of 0.85 times the balanced reinforcement ratio. A 4 × 5 mesh was used to discretisize the quadrant of the slab due to symmetry, and the concrete of the cross section was divided into 10 concrete layers. The progress of cracking process at the cross section A-A (shown in Figure 2) was tracked and presented in Figure 2 for the external loading of 17.65 kN, 32.55 kN and 134.9 kN respectively. It can be seen that the cracking occurred only at the bottom layer of concrete when the external load was 17.65 kN, and the cracks approached further throughout the cross section of the slab with the increase of the external loading.

(18)

Ae

where σ = N, M, TT , N = Nx , Ny , Nxy T is the membrane force vector defined at the central-plane, M = Mx , My , Mxy T is the bending moment vector and T = Qx , Qy T is the transverse shear force vector. In the finite element analysis, the material properties of each layer and the element internal force will be updated continually with the change of the stress states at each Gauss point. The updated in-plane stress vectors for the ith concrete layer and jth FRP layer in the next (n + 1)th iteration due to the nodal displacement increment e qm , qbe are (σ i )n+1 = (σ i )n + σ i ,

5.2

(σ j )n+1 = (σ j )n + σ j (19)

Cracking analysis of a clamped GFRP-reinforced concrete slab

The cracking process of a 3 m × 2 m × 0.2 m clamped GFRP-reinforced rectangular concrete slab tested by

where i T σ i = σxi , σyi , τxy   l  e e = Dc,i · Bm qm + zBb qbe + Bnl m qb

Element mesh

(20)

and

A

A

j T σ j = σxj , σyj , τxy   l  e e = DFRP,j · Bm qm + zBb qbe + Bnl m qb

(21)

F=17.65kN

The updated out-of-plane shear stress vector in the next (n + 1)th iteration due to the shear stress increment τ is (τ i )n+1 = (τ i )n + τ i ,

XXX

(22)

with  τxz i = Dco,i Bs qbe . τ = τyz

F=32.55kN



5

(23)

FINITE ELEMENT MODELLING OF CRACKING PROGRESS OF FRP-REINFORCED CONCRETE SLABS

5.1 Cracking analysis of a simply supported one-way CFRP-reinforced concrete slab The cracking progress of a simply supported oneway CFRP-reinforced concrete slab tested by Zhang

XXX XXX X X X XXX X X X XXX X X X X XXX X X X X

X X

X X

XXX XXX XXX XXX XXX XXX XXX XXX XXX

X X X X X X X X X

X X X X X X X X

Figure 2.

206

X X X X X X X X X

X X X X X X X X X

X X X X X X X X X

X X X X X X X X X

F=134.9 X X X X X X X X

X X X X X X X X

X X X X X X X

X X X X

X X X

X X X

Progress of cracking at cross section A-A.

100.15kN

B

B

XXX XX X XX X XX X

XXX XXX XXX X XX X XXX XXX XX X XX X XXX

XX XX X

XX

499.34kN 70.97kN XX XX XXXXXXX X X XXX XXXXX XXXX XX XX XXXX XX XX X XXX X XXXX XX XX X X XXXXXXXXXX X X XX XXXXXX XXXX X XXXX XXX XXXXX X XXX X XXXXXXXX X XXX X XXXX XXXX X

XXX XXX

Figure 3.

X X X X X X X X X

XX X X X X X X X X X X X X X X X X X X X X X X X X X X X X

XX X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X X

X X X X X X X X X X

X X X X X X X X X X

X X X X X X X X X X

X X X X X X X X X

Progress of cracking at cross-section B-B.

EI-Gamal et al. (2007) was analysed herein. The slab was reinforced with GFRP bars in tension and compression region. Due to symmetry only one quarter of the square slab was analysed using a 5 × 8 mesh, and 10 concrete layers were used to divide the cross section of the model. The progress of cracking at the cross section B-B (shown in Figure 3) with the increase of external loading for this clamped slab is shown in Figure 3. It can be seen that concrete cracking occurred when external loading reached 70.97 kN, that when the external loading was 100.15 kN, the fourth layer of concrete cracked near central point of the slab, and that when the external loading reached 499.34 kN, the cracks occurred throughout the whole cross section. ACKNOWLEDGEMENT The supports from the Rector’s starting up Grants awarded to Y.X. Zhang and from the Rector’s Visiting Fellowship awarded to Y. Zhu by the University of New South Wales at the Australian Defence Force Academy to this research are acknowledged. REFERENCES Cedolin, L. & Deipoli, S. 1977. Finite element studies of shear-critical R/C beams. Journal of the Engineering Mechanics Division, ASCE 103: 395–410.

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El-Gamal, S., El-Salakawy, E. & Benmokrane, B. 2007. Influence of reinforcement on the behavior of concrete bridge deck slabs reinforced with FRP bars. J. Comp. Constr. ASCE; 11(5): 449–458. Izumo, J., Shin, H., Meakawa, K. & Okamura, H. 1992. An analytical model for PC panels subjected to in-plane stresses, Concrete Shear in Earthquakes, Elsevier Applied Science, London and New York, 206–215. MacNeal, R.H. & Harder, R.L. 1988. A refined four-noded membrane element with rotational degrees of freedom. Comput. Struct. 28: 75–84. Zhang, B., Masmoudi, R. & Benmokrane, B. 2004. Behavior of one-way concrete slabs reinforced with CFRP grid reinforcements. Constr. Build. Mater. 18: 625–635. Zhang, Y.X. & Kim, K.S. 2004. Two simple and efficient displacement-based quadrilateral elements for the analysis of composite laminated plates. Int. J. Num. Methods Eng. 61: 1771–1796. Zhang, Y.X. & Zhu, Y. 2009. A new shear-flexible FRPreinforced concrete slab element. In A.J.M. Ferreira (ed.), Proceeding of the 15th International Conference on Composite Structures (ICCS15), June 15–17, 2009, Porto, Portugal. Zhang, Y.X., Bradford, M.A. & Gilbert, R.I. 2007. A layered shear-flexural plate/shell element using Timoshenko beam functions for nonlinear analysis of reinforced concrete plates. Finite Elements in Analysis and Design, 43(11): 888–900.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

An experimental study on double steel-concrete composite beam specimens S.J. Duan & J.W. Wang Research Institute of Structural Engineering, Shijiazhuang Railway Institute, Shijiazhuang, P.R. China

Q.D. Zhou Research Institute of Structural Engineering, Shijiazhuang Railway Institute, Shijiazhuang, P.R. China Department of Civil Engineering, Tianjin Railway Technical and Vocational College, Tianjin, P.R. China

H.L. Wang Research Institute of Structural Engineering, Shijiazhuang Railway Institute, Shijiazhuang, P.R. China Department of Civil Engineering, Tianjin Institute of Urban Construction, Tianjin, P.R. China

ABSTRACT: Double steel-concrete composite continuous beam is a new structural system, in which, there is a concrete bottom slab connected with the bottom steel flange through studs where the cross-section suffers negative moment. Three pieces of 2 × 2.9 m span double steel-concrete composite continuous beam specimens with full shear connection are designed and tested, and are subjected to concentrated loads and their self-weight in order to reveal the mechanical behavior of the composite beam in the paper. If once the concrete cracks, the interface slip occurs and then increases with the load because the used stud is a kind of flexible shear connector. The load-deflection curve, the ultimate flexural capacity, the interface slips and the slip strain between steel and concrete along the span direction are measured and obtained. The bearing capacities of the beam’s cross-section by the tests are fairly close to the plastic moment by the simplified calculating formulae. Both the bending stiffness and the structural ultimate bearing capacity of the double steel-concrete composite continuous beam are greatly higher than that of the single steel-concrete composite continuous beam. 1

INTRODUCTION

Double steel-concrete composite continuous beam is a new structural system in which there is a concrete bottom slab connected with bottom steel flange through studs where the cross-section suffers hogging bending moment (Reiner 1996). Therefore it is a reasonable and effective measure to improve the performance of the single composite beams under hogging bending moment for the continuous beam. Although many experimental and theoretical studies for the traditional single steel-concrete composite beam have been done, few research studies have been found in references to the double steel-concrete composite continuous beam. Up to now, some theoretical studies can be found for the double steel-concrete composite beams focus on the problems of interface slip, deformation, ultimate bearing capacity and the effective flange width of concrete slab (Yen et al. 1986; Stroh et al. 2000; Wang et al. 2006; Duan et al. 2007a, b; Yang & Duan 2008). A double composite girder under pure hogging moment was tested and its ultimate bending moment strength was measured by Nagai et al. (2007).

Three pieces of 2×2.9 m span double steel-concrete composite continuous beam specimens with full shear connections were designed and tested at collapse, as seen in Figure 1. These beams were subjected to concentrated loads and their self-weight in order to study the mechanical behavior of the composite beam in this paper. For saving space, some experimental results are omitted.

2 2.1

TESTING PROGRAMME Beams and materials

The experimental test concerned three double steelconcrete composite beams, named SCB1, SCB2 and SCB3. These beams are designed base of the limited plastics state with the full shear connection, the properties of which are summarized as following: The cross-sections for all the specimens are constructed by a Chinese steel profile HW 150 × 150 × 150 × 7 × 10 (see Figure 1) by a top concrete slab along the whole beam length with tension reinforcement 78, and by a 1000 mm length bottom concrete

209

Figure 1.

Test specimen and its cross-sections.

Figure 2. Location of diagonal gauges for deformation measurements.

210

Figure 3.

Location of instrumentation for strain measurements.

slab over interior support. The concrete strength is C30 with fc = 47 MPa, Ec = 46.2 GPa, the steel being Q235 with ft = 235 MPa, Ec = 206 GPa and the head studs being 13 × 60. The heads were arranged on two lines with different numbers for each beam (top slab being 94 and bottom slab being 28 for SCB1; top slab being 82 and bottom slab being 28 for SCB2; top slab being 82 and bottom slab being 24 for SCB3). 2.2 Instrumentation During each test both global and local quantities, such as displacements, relative slips and slip strains were monitored. Dial gauges were located along the beam axis, as shown in Figure 2, for measuring the vertical displacements, which named by Ni or Hi, and they were also used for measuring the relative slip between slab and steel profile which named by Di. Strain gauges were applied on the steel profile surfaces and on the concrete slab surfaces in typical crosssection (see Fig. 3), in which the number represents

211

the cross-section. The gauges were also applied on the reinforcing bars before the concrete casting only in the section 5. 2.3

Experimental setup of the beams and data collection

The continuous beam was subjected to bending moment through a concentrated load applied on the top surface in each middle span section. The load was applied by a hydraulic jack with steps of 15 kN at the beginning, with 10 kN from steel profile edge yielding, and with 5 kN over about 70% of the collapse load.

3 3.1

EXPERIMENTAL RESULTS AND ANALYSIS Main results

The measured results for the three test beams are displayed in Table 1.

3.2

Testing phenomenon description

3.2.1 Loading process and crack extension One cycle loading style is applied to SCB1. The first flexural crack occurs when P = 40.35 kN on the top Table 1.

Main results.

Items

Control SCB1 SCB2 SCB3 cross section

Cracking load (kN)

40.35 −

46.85 Interior support section Elastic ultimate 189 188 187 Interior support load (kN) section Plastic ultimate 234 233 232 Interior support load (kN) section Max. deflection (mm) 14.61 15.37 14.61 Span center Max. slip (mm) 2.05 0.725 0.85 Interior support section

Figure 4.

surface of the top concrete slab over the interior support point. With the load increasing, the crack extends and other cracks initiate; the clear sound was made from the specimen when P = 89.3 kN, that the cohesion between steel and concrete being overcome can be conjectured. The width of the first crack reaches to 1.55 mm before the beam collapsed. Specimen SCB2 used three cycle loading style. The first cycle is to add the theoretical ultimate linear load and then unload to zero; the second one is to add the theoretical ultimate elastic load and then unload to zero; at last that is to collapse. One cycle loading style is also taken to SCB3. 3.2.2 The specimen deflection The measured load deflections of the middle section in every span for SCB1, SCB2 and SCB3 are shown in Figures 4 (a) to (f). Notice that the deformation by the

Loading-deflection curves by experiment.

212

Figure 5.

Load and slip relations.

Figure 6.

Load and slip strain relations.

213

weight itself is not included, and for SCB2, the result is one of the last loading cycle. The curve can be divided into concrete tensile cracking, elastic, elasto-plastic and plastic phases from the figures. They show that the composite beams have good ductility. 3.2.3 Slip character The interface slips between concrete and steel for SCB1, SCB2 and SCB3 are demonstrated in Figures 5 (a) to (f). The corresponding slip strain distribution along the beam length direction can be deduced from slip curve and is displayed in Figure 6. The curve shows the change is smoothed because of the flexural shear connector being used. 4

CONCLUSIONS

Through the beam collapse tests for three pieces of double steel-concrete composite continuous beam, the load-deflection curve, the ultimate flexural capacity, the distribution of cracks and their widths, the strain distribution over the depth of the typical crosssections, the interface slips and the slip strain between steel and concrete along the span direction, and stress distribution over the width were measured and obtained. It supplied some references to the studies and engineering practice for double steel-concrete composite beams.

REFERENCES Duan S.J., Duan, Y.J. & Zhang, Z.G. 2007A. The interface slip expression of double steel-concrete composite beam under concentrated load. Journal of Shijiazhuang Railway Institute 20(2): 1–4, 21 (in Chinese). Duan, S.J., Huo, J.H. & Zhou, Q.D. 2007B. The research on the calculating method for the ultimate bearing capacity of double steel-concrete composite beams. Journal of Shijiazhang Railway Institute 20(4): 1–4 (in Chinese). Nagai, M., Inaba, N., et al. 2007. Experimental study on ultimate strength of composite and double composite girders. In: Proceedings of 8th Pacific Structural Steel Conference-Steel Structures in Natural Hazards. pp. 329–334. Reiner, S. 1996. Bridges with double composite action. Structural Engineering International 1: 32–36. Stroh, S.L. & Sen, R. 2000. Steel bridge with doublecomposite action: innovative design. In: 5th International Bridge Engineering Conference, April 3–5, 2000, Tampa, FL, United States. Transportation Research Record 1(1696): 299–309. Wang, G., Wang, F.J., et al. 2006. Theoretical analysis of double composite beam deformation in elastic state by Goodman elastic sandwich method. China Railway Science 27(5): 66–70. (in Chinese) Yang, X.W. & Duan, S.J. 2008. The effective width of reinforcement bars for double steel-concrete composite beam. Engineering Mechanics 25(A1): 184–188 (in Chinese). Yen, B.T., Huang, T., et al. 1986. Steel box girders with composite bottom flanges. In: Official Proceedings-3rd Annual International Bridge Conference, Pittsburgh, PA, USA. pp.79–86.

ACKNOWLEDGEMENTS The work in this paper was supported by the Natural Science Foundation of Heibei Province, China (Contract No. E2008000397). The authors would also like to thank the people who contribute the study in Shijiazhuang Railway Institute.

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Behaviour of FRP wrapped circular reinforced concrete columns M.N.S. Hadi & V. Yazici School of Civil, Mining and Environmental Engineering, University of Wollongong, Australia

ABSTRACT: Columns are generally under bending moment and axial load at the same time. Studies done on solid concrete columns under concentric loading have shown that FRP wrapping improves the strength and ductility of columns. This paper presents results of testing 16 reinforced concrete columns, 8 solid and 8 hollow, under different eccentricities. The height and the diameter of the columns were 925 mm and 205 mm respectively. Each group of sample columns was divided into two sub-groups. The first sub-groups served as a reference and did not have any external FRP wrapping. The second sub-groups were wrapped with three layers of Carbon FRP in the hoop direction. Sample columns were tested under concentric, eccentric (25 m and 50 mm) and pure bending loading. Axial load—bending moment interaction diagrams were constructed for each of the sub-groups. The effectiveness of CFRP wrapping on solid and hollow sample was compared and discussed. 1

INTRODUCTION

Increasing costs of building materials has created a new trend of retrofitting the existing structural members which would otherwise be insufficient due to increased demand from the structure, adopting more stringent design codes, or deterioration of structural members in hostile environmental conditions. Fiber reinforced polymer (FRP) sheet wrapping has been a common method of strengthening reinforced concrete columns for the last two decades. This new material’s resistance to corrosion, excellent durability to harsh conditions, high tensile strength to weight ratio and ease of installation to existing structural members have made it a popular strengthening material. There have been various studies investigating the behaviour of FRP wrapped solid columns such as capacity or ductility increase. However, behaviuor of FRP wrapped hollow columns has been less investigated though their frequent use as structural members such as bridge piers. This paper investigates the effect of Carbon-FRP (CFRP) wrapping on both reinforced solid and hollow core concrete columns under different loading conditions.

2

LITERATURE REVIEW

Steel jacketing is assumed to provide a constant confinement to the columns after the yield strain. There are a number of steel confined concrete stress-strain models and most of them are based on Richart et al.’s study (1928). Since the FRP materials exhibit a linear stress-strain behaviour under tensile loads up to the failure, the confinement stress, fl , they provide to the columns is highly dependant on the strain level

of FRP wrapping. Models proposed by Samaan et al. (1998), Spoelstra & Monti (1999), Toutanji (1999) and Teng et al. (2002) have shown that previously defined stress-strain models for steel confined concrete such as Mander et al. (1988) cannot be used for FRP confined concrete. Although none of FRP confined concrete stress-strain models has been accepted as a standard model worldwide, various studies such as), Hadi (2007a, b) have shown that both strength and ductility of solid concrete columns are substantially improved after FRP wrapping. Hollow core columns are generally preferred to solid columns to reduce the self weight and cost of the structures. In spite of their widespread use, even modern design codes do not specify a special design procedure for hollow core columns. 3

THEORETICAL CONSIDERATIONS

Although the only apparent difference from an FRP wrapped solid column seems to be the hollow portion in the center, confinement mechanism of an FRP wrapped hollow core column under an axial load is quite different. For an axially loaded column, when FRP wrapping reaches its maximum tensile strength, ffrp , the confinement stress reaches its maximum value, fl (Figure 1). The maximum confinement stress provided by FRP wrapping, fl , can be calculated using the Equations 1 and 2 for the solid and hollow core columns, respectively, derived from the equilibrium of forces acting on the half-cut cross-sections shown in Figure 1.

215

fl,solid =

2ffrp tfrp D

(1)

Figure 1. Forces acting on half cut cross sections of a) solid column, b) hollow core column under axial loading.

fl,hollow =

2ffrp tfrp (D − d)

(2)

where tfrp is the total thickness of FRP wrapping, D is the outer diameter of solid and hollow core column, and d is the hollow core diameter. Equations 1 and 2 imply that, for the same D, ffrp , and tfrp ; fl , is larger on the hollow core columns because of the hollow part. However, for solid columns fl results in a triaxial stress state on any concrete element within the column whereas this confinement creates only a biaxial state of stress on a concrete element within a hollow column resulting in less capacity increase (Zahn et al. 1990; Yazici & Hadi 2008).

4

EXPERIMENTAL STUDY

The experimental part of this study was conducted at the laboratories of the School of Civil, Mining and Environmental Engineering at the University of Wollongong and involved testing of 16 circular reinforced concrete columns, half solid and half hollow core. The geometry of sample columns and testing conditions are given in Figure 2 and Table 1, respectively. The concrete used to cast the sample columns had a 28 day compressive strength of 60 MPa. All specimens were internally reinforced with the same amount of steel reinforcement. Longitudinal steel reinforcement consisted of evenly distributed six N12 (12 mm diameter deformed bar) bars and tied inside a helical steel reinforcement. The helical reinforcement was made of R10 bars (10 mm diameter plain bar) with a 50 mm pitch. The steel reinforcement was placed into the column moulds with 20 mm clearance to the outer moulds (Fig. 2). A summary of material testing is given in Table 2. The fl values were calculated using Equations 1 and 2 for groups of SF and HF sample columns respectively. Confinement due to helical steel was ignored. A PVC pipe having an inner diameter of 205 mm was used as the outer mould for both solid and hollow core columns. A 56 mm outer diameter PVC pipe was used to form the hollow part of Groups H and HF. After

Figure 2. a) General geometry of column samples, b) Crosssection of solid column samples, c) Cross section of hollow core column samples. Table 1. Geometry of sample columns and testing conditions. Inner Internal WrapSpeci- Diadia- reinping men meter Height meter force- configucode (mm) (mm) (mm) ment ration

Test eccentricity (mm)

205

925

−

Yes

None

SF0 205 SF25 SF50 SFB

925

−

Yes

Three layers 0 of carbon 25 FRP in hoop 50 direction Bending

H0 H25 H50 HB

205

925

56

Yes

None

HF0 205 HF25 HF50 HFB

925

56

Yes

Three layers 0 of carbon 25 FRP in hoop 50 direction Bending

S0 S25 S50 SB

0 25 50 Bending

0 25 50 Bending

setting of the concrete, the outer moulds were removed using the previously cut joints on the PVC pipe, and the inner PVC pipes for hollow core sample columns were pulled out by means of a hydraulic jack. Epoxy was used to adhere three layers of Carbon FRP sheet on to the surface of sample column Groups SF and HF with fibres in the hoop direction with 100 mm overlap and one CFRP layer onto another (Figure 3). The combined tensile strengh of three layers of CFRP was tested to have 920 MPa using the standard ASTM 3039-2006.

216

Table 2.

Summary of material tests and confinement stresses due to Carbon FRP wrapping.

Sample Group

Concrete strength, fco ,(MPa)

Longitudinal steel strength (MPa)

Helical steel strength (MPa)

Combined tensile strength of CFRP, ffrp , (MPa)

Combined thickness of CFRP, tfrp , (mm)

Maximum confinement stress due to CFRP wrapping, fl , (MPa)

S SF H HF

60 60 60 60

558.6 558.6 558.6 558.6

366.9 366.9 366.9 366.9

− 920 − 920

− 1.60 − 1.60

− 14.4 − 19.8

Figure 3. CFRP wrapping configurations for Sample Groups SF and HF.

Figure 5.

Four-point loading apparatus.

Figure 4.

Figure 6. samples.

Measurements taken for axially loaded column

Figure 7.

Measurements for bending.

a) Loading Head, b) Knife edge.

One specimen from each sub-group (Specimens S0, SF0, H0, HF0) was tested to failure under concentric loading, the next two specimens were tested under 25 mm (S25, SF25, H25, and HF25) and 50 mm (S50, SF50, H50, and HF50) eccentric loading. The final specimen of each sub-group (SB, SFB, HB, and HFB) was tested under pure bending. 25 mm and 50 mm eccentric loadings were applied to the columns by means of especially designed and manufactured loading heads and knife edges as shown in Figure 4. Pure bending load was applied to specimens (in fact beams) SB, SFB, HB, and HFB by means of a four point loading apparatus as shown in Figure 5. For concentric and eccentric loading the load, P, that was applied to the columns was measured by the internal load cell of the loading machine. For the axially loaded specimens, the axial deformations () of the

217

columns were monitored by an LVDT attached to the moving (lower) plate of the loading device. Lateral deformations of the specimens (δ) were monitored at the mid height of columns using a laser displacement sensor (Figure 6).

For pure bending loading, the flexural load, P, and corresponding mid-span bending deformation was measured (Figure 7). Bending deformation was measured by an LVDT attached to the lower plate of loading device. A hole was formed in the middle of lower part of four point loading device to let the laser through.

5

Figure 8–9 shows the axial load-deformation graphs of solid sample columns, namely sub-groups S and SF, respectively. Figure 10 shows flexural load-midspan deformation of sample columns SB and SFB. Figure 11–12 shows the axial load-deformation graphs of hollow core sample columns, namely subgroups H and HF, respectively.

RESULTS

All specimens were tested to failure. For 25 mm and 50 mm eccentrically loaded columns, bending moment capacities are calculated by multiplying the maximum load capacity (Pmax ) and the eccentricity (e). Bending moment capacities including the secondary moments (MII ) were also calculated as follows; MII = Pmax (e + δ)

Table 4.

Midspan Max. Load, Deformation at M (Bending Specimen Pmax (kN) Pmax , δ, (mm) Moment) (kNm)

(3)

Where Pmax = max axial load, e = eccentricity, and δ = lateral deflection at maximum load. Table 3 shows a summary of testing results for the column specimens tested under concentric and eccentric loading. Table 4 shows the test results for sample columns tested under four point loading (pure bending). Bending moment for sample columns tested under four point loading regime was calculated using Equation 4 derived from the moment equation for simply supported beams. Mmax =

0.280Pmax 2

SB SFB

288.0 370.1

9.59 26.47

40.32 51.82

HB HFB

207.2 327.7

23.04 35.65

29.01 45.88

(4)

where 0.280 is the distance between one lower support to the nearest upper loading point, and Pmax is the maximum flexural load applied to the column sample. Table 3.

Summary of results for pure bending.

Figure 8. Load-deformation graph for solid columns without CFRP wrapping.

Summary of test results for concentric and eccentric loading. Axial Midheight Eccen- deformation horizontal deformation at max. Max. Load, tricity at Pmax , Pmax (kN) e, (mm) , (mm) load, δ, (mm)

Axial Bending deformation Horizontal Moment MII = at failure deformation at MI = Pmax e Pmax (e + δ) (mm) failure (mm) (kNm) (kNm)

S0 S25 S50

2802.3 1440.4 1134.4

0 25 50

5.63 6.02 4.69

− 5.80 4.31

11.80 17.02 4.71

− 53.04 4.94

− 36.01 56.72

− 44.36 61.61

SF0 SF25 SF50

4504.5 3069.2 1527.0

0 25 50

9.70 16.10 8.62

− 14.32 12.14

12.95 24.79 17.11

− 72.26 63.02

− 76.73 76.35

− 120.68 84.50

H0 H25 H50

2145.4 1699.5 915.7

0 25 50

4.50 7.03 4.38

− 4.17 5.34

29.04 21.61 13.30

− 36.51 27.20

− 42.49 45.78

− 49.52 50.67

HF0 3237.7 HF25 1785.0 HF50 1119.3

0 25 50

12.54 7.18 8.98

− 7.64 12.48

29.90 15.40 19.18

− 53.06 53.36

− 44.63 55.97

− 58.26 69.93

218

Figure 9. Load-deformation graph for CFRP wrapped solid columns.

Figure 10. Flexural load-midspan deformation of sample columns SB and SFB.

Figure 11. Load-deformation graph for hollow core columns without CFRP wrapping.

Figure 12. Load-deformation graph for CFRP wrapped hollow core columns

Figure 13. Flexural load-midspan deformation of sample columns HB and HFB.

Figure 14. columns.

6 Figure 13 shows flexural load-midspan deformation of sample columns HB and HFB. Axial load-bending moment diagrams (P-M) of all sub-groups are shown in a single chart in Figure 14.

219

P-M diagrams of all sub-groups of sample

DISCUSSION OF RESULTS

Increase in the axial load carrying capacity due to CFRP wrapping for concentric loading was larger for solid column sample as expected (60.7% and 50.9% for sample columns SF0 and HF0, respectively).

For 25 mm eccentric loading, solid column sample SF25 exhibited an axial load capacity increase of 113.1% compared to S25 which implied a possible error in testing as a less capacity increase was expected as the eccentricity of loading increases. For the same eccentricity (25 mm), the hollow core column sample HF25 exhibited only a 5% axial load capacity increase compared to H25. For 50 mm eccentricity, SF50 exhibited an axial load carrying capacity increase of 34.6% compared to S50. The increase was only 22.2% for HF50 compared to H50. For pure bending loading, the increase in flexural load carrying capacity was more in hollow core column samples. The flexural load carrying capacity was increased by 58.2% for HFB compared to HB, whereas 28.5% for SFB compared to SB. Axial deformation capacities corrsponding to Pmax of hollow sample columns without CFRP reinforcement tend to be larger than that of solid columns without reinforcement. Likewise, the increase in the axial deformation capacities corresponding to Pmax of CFRP wrapped hollow core columns are generally higher than that of CFRP wrapped solid columns for the same level of eccentricity. However, horizontal deformation capacities of solid columns (wrapped or not), are generally higher than hollow core columns for the same level of eccentricity and CFRP wrapping configuration. 7

CONCLUSIONS

Testing results revealed that CFRP wrapping increased axial load carrying capacity, axial and horizontal deformation capacities of both solid and hollow core reinforced columns at the same time. However, the increase was more substantial for solid columns. The column specimens were designed as short columns, but after CFRP wrapping horizontal de-formation capacity increased in such amounts that the secondary

moments became considerable. In other words, the column specimens began to behave like slender columns under eccentric loads. This problem should be considered when the columns are being strengthened with FRP wrapping especially for the columns which will also be subjected to eccentric loads unavoidably. REFERENCES ASTM 3039-2006, Standard test method for tensile properties of polymer matrix composite Materials. Hadi, M.N.S. 2007a. Behaviour of FRP strengthened concrete columns under eccentric compression loading. Composite Structures 77(1), 92–96. Hadi, M.N.S. 2007b. The behaviour of FRP wrapped HSC columns under different eccentric loads. Composite Structures 78(1): 560–566. Mander, B.J., Priestley, M.J.N. & Park, R. 1988, Theoretical stress-strain model for confined concrete, Journal of Structural Engineering, ASCE 114(8),1804–1825. Samaan, M., Mirmiran, A. & Shahawy, M. 1998. Model of concrete confined by fiber composites. Journal of Structural Engineering, ASCE 124(9): 1025–1031. Spoelstra, M.R. & Monti, G. 1999. FRP-confined concrete model. Journal of Composites for Construction, ASCE 3(3): 143–150. Richart, F.E., Brandtzaeg, A. & Brown, R.L., A Study of the Failure of Concrete under Combined Compressive Stresses. Bulletin 185. University of Illinois Engineering Experimental Station, Champaign, Ill, 1928. Teng, J.G., Chen, J.F., Smith, S.T. & Lam, L. 2002. FRPStrengthened RC structures. John Wiley & Sons Ltd., West Sussex, England. Toutanji, H.A. 1999. Stress-strain characteristics of concrete columns externally confined with advanced fiber composite sheets. ACI Materials Journal 96(3): 397–404. Yazici, V. & Hadi, M.N.S., 2008, Interaction diagrams for FRP wrapped circular hollow columns, Proc.The 20th Australasian Conference on the mechanics of structures and materials, Toowoomba, Australia, 2–5-December 2008, London: Balkema.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Contribution of NSM CFRP bars in shear strengthening of concrete members A.K.M.A. Islam Youngstown State University, Youngstown, Ohio, USA

ABSTRACT: The author presents the results of an experimental study that investigated the shear strength contribution of carbon fiber reinforced polymer (CFRP) bars attached with concrete beams using near surface mounted (NSM) technique. In this research, four concrete beams were cast with regular steel reinforcement in flexure. The control beam had typical shear steel and the other three beams were strengthened in shear with CFRP bars. Strain gages were attached with the shear reinforcement of all four beams at various shear critical locations. Strains during loading to failure of the beams were recorded using a data acquisition system. Performance of NSM technique was found to be very effective with no occurrence of delamination, debonding or fracture of FRP. Effective strains in the NSM CFRP bars were determined analyzing the collected strain data. A new formula to calculate the nominal shear strength provided by NSM CFRP bars has also been proposed herein. 1

INTRODUCTION AND BACKGROUND

Shear failure is a common problem in concrete structures. Natural disasters, such as hurricanes and earthquakes, may also cause shear failure of structures before full flexural capacity is achieved (MCEER 2005). Reinforced concrete (RC) and prestressed concrete (PC) structures, such as buildings and bridges that were designed several decades ago, also exhibit shear cracks because of regular and unintended or unforeseen loads, unaccounted loads in the earlier designs, inferior material behavior, and loss of concrete strength due to aging (Bousselham & Chaallal 2004). Efficient and cost-effective method of strengthening concrete members in shear is of utmost importance to encounter shear-deficiency problem in RC and PC structures. Fiber reinforced polymer (FRP) systems have been used in the United States for almost 20 years and are becoming a widely accepted method for strengthening concrete structures in flexure and shear. American Concrete Institute (ACI) Committee 440 (ACI 2002) has developed guidelines for the design and construction of externally bonded FRP systems for strengthening concrete members. In this paper, the term ‘externally bonded FRP’ refers to FRP composites attached on the external surface of concrete members by means of epoxy or cement-based adhesive. In case of externally bonded FRP systems, FRP sheets, strips or laminates are externally attached on the concrete surface. The use of externally bonded FRP laminates has been derived from the practice of steel plate bonding. On the other hand, near surface mounted (NSM) FRP system is not a widely used method yet for strengthening deficient concrete members in flexure and shear (Alkhrdaji et al. 1999; De Lorenzis et al. 2000), as well as strengthening

221

unreinforced masonry walls (Tumialan et al. 2001). The NSM technique was developed in Europe for strengthening reinforced concrete structures in 1950s. In 1948, an RC bridge deck in Sweden needed to be upgraded in its negative moment region (Asplund 1949) due to an excessive settlement of a steel cage during construction. This was accomplished by inserting steel reinforcement bars in grooves made in the concrete surface and filling them with cement mortar. In the NSM technique, grooves are cut into the surface of the concrete members, and FRP bars, sheets or strips are inserted into the grooves and attached using epoxy or cement-based adhesive. The grooves are normally cut 1.5 to 2 times the diameter of the FRP bars to ensure adequate bonding between FRP composites and surrounding concrete. Unlike externally bonded FRP strengthening, NSM technique does not require additional precaution in surface preparation. In fact, the technique has been proved to be very practical, efficient, and economical in strengthening negative moment regions of beams and slabs. Different methods of strengthening concrete members in shear have been developed and are being used. The lighter weight and greater tensile strength of CFRP laminate greatly reduce the overall installation and maintenance costs as compared to steel plate bonding techniques. Moreover, FRP materials have gained popularity due to their non-corrosive nature and low maintenance cost. Shear steel reinforcement is internal; however, FRP composites are bonded externally, which makes the assessment of the FRP performance and bond mechanism more complex. RC and PC members strengthened in shear with externally bonded FRP composites show a number of failure modes that are largely related to the bond behavior at the concrete/ composite interface. Debonding, delamination, and

steel (MPa), s = spacing of shear reinforcement (mm), Af = area of FRP bars in shear on both sides of beam (mm2 ), Ef = tension modulus of elasticity of FRP bars (MPa), and εef = effective strain in FRP bars.

fracture of FRP composites at the ultimate load or thereafter are examples of such failure modes (Triantafillou and Antonopoulos 2000). Geometry of the member, type of applied loading, ratios of internal longitudinal and shear reinforcement, and shear span ratio (defined as the ratio of the shear distance to the effective depth of the member) are some of those parameters, some of which have been investigated in this research. The experimental and analytical study conducted in this research measured the corresponding strains in the FRP bars and shear steel at various shear critical locations along the beam, and developed methods to calculate the nominal shear strength provided by the NSM CFRP bars. The outcomes of this research are expected to provide useful guidance in developing design criteria for shear strengthening of concrete members with FRP bars attached using NSM technique. 2

3 3.1

Beam design and loading configuration

Typical beam dimensions and loading configurations used in this research are shown in Figs. 1 and 2. The beams were 2.134 m (7 ft) long with a cross-section of 254 mm by 305 mm. They were made of concrete with average compressive strength of 49.75 MPa (7.215 ksi). The concrete compressive strength was later determined by taking the average strength of three concrete cylinder specimens taken during the pouring of concrete into the beam formwork. All four beams had 4 No. 19 (#6) steel bars at the bottom and 2 No. 19 (#6) steel bars at the top as flexural reinforcement, as shown in the cross section presented in Fig. 2. In order to achieve full shear capacity before full flexural capacity at failure, the beams were designed with stated reinforcement and concrete strength to make them fail in shear slightly before they fail in flexure. Grade 60 reinforcing steel with tensile yield strength of 413.7 MPa (60 ksi) was used to strengthen the beams in flexure and shear. The nominal flexural and shear capacity of the control beam were calculated as 112 kN-m and 178.3 kN, respectively, based on

MODEL BEAM SHEAR STRENGTH

In recent years, many research programs (Malek & Saadatmanesh 1998; Triantafillou 1998; Triantafillou & Antonopoulos 2000; Matthys 2000; Khalifa & Nanni 2000) have been conducted to develop appropriate calculation models and general design guidelines for strengthening RC members with externally bonded FRP composites. The determination of the concrete and steel stirrup contribution is based on expressions provided by ACI 318-05 (ACI 2005). Thus the nominal shear capacity (Vn ) of a RC member strengthened in shear with NSM FRP composites can be given by Equation 1 below: Vn = Vc + Vs + Vf

EXPERIMENTAL PROGRAM

Strain Gage

P

P Steel Stirrups @ 152 mm o.c. (Control Beam) NSM CFRP Bars @ variable spacing (Test Beams)

(1)

where Vc stands for the nominal shear strength provided by concrete, and Vs and Vf stand for the nominal shear strength provided by steel and FRP bars, respectively, within a length equal to the effective depth. The nominal shear strength contribution by concrete, steel and FRP bars can be calculated by using the following Equations 2, 3 and 4, respectively.  fc bw d (ACI 2005) (2) Vc = 6 Av fy d (ACI 2005) (3) Vs = s Af Ef εef d Vf = (4) s where fc = specified compressive strength of concrete (MPa), bw = beam web width (mm), d = distance from extreme compression fiber to centroid of longitudinal tension reinforcement (mm), Av = area of shear steel (mm2 ), fy = tensile yield strength of shear

152

610

610

610

152

1829 2134 Elevation

Figure 1.

Beam dimensions and loading configurations. Steel Stirrups NSM CFRP Bars

305

Strain Gage 305

13

13 Flexural Steel Bars Plan

Figure 2.

222

Epoxy 254 Cross Section

Plan and cross section of model beam.

structural design. They were loaded under four-point bending within a span of 1.83 m, as shown in Fig. 1, until failure. The theoretical 4-point bending force required to fail the control beam in flexure was calculated as 364 kN, whereas to fail in shear was calculated as 353 kN. Therefore, it was theoretically predicted that the beam would fail in shear before it would fail in flexure. This was done in order to assess any shear strength gain by the model beams as a result of shear strengthening by using NSM CFRP bars. There were a total of four beams with Beam 1 as the ‘control beam,’ and Beam 2, Beam 3 and Beam 4 were referred herein as the ‘model beams.’ The control beam had single No. 10 (#3) double-legged steel stirrups at 152 mm on center, as shown in Fig. 1. All four beams had a typical clear cover of 25 mm on all four sides. The beams were cured using wet burlap for more than a month to ensure adequate strength gain before they were strengthened in shear. The model beams were strengthened in shear, using NSM techniques shown in Figs. 1 and 2, with No. 10 (#3) CFRP bars attached vertically at a spacing of 152 mm on both sides of the beam. The mechanical properties of No. 10 (#3) ASLAN 200 CFRP bars (Hughes Brothers 2007) reported by the manufacturer are shown in Table 1. The groove size was 13 mm wide and 13 mm deep, which are one and a half times the diameter of No. 10 (#3) CFRP bars. The vertical length of each groove is 305 mm, which is equal to the depth of the beam. The grooves were half-filled with Concresive 1420 (BASF 2006) structural epoxy, and the No. 10 (#3) round shaped deformed CFRP bars were inserted into the grooves and left for a week for adequate curing. Beam 2 was strengthened in shear by attaching NSM CFRP bars at a spacing of 152 mm on both sides of the beam. Beam 3 has single No. 10 (#3) double-legged steel stirrups at 305 mm spacing with NSM CFRP bars attached in between the steel stirrups on both sides of the beam. This beam had CFRP bars attached with a spacing of 305 mm on both sides of the beam. Beam 4 had single No. 10 (#3) steel stirrups at 610 mm spacing (to hold the flexural reinforcement during casting) with NSM CFRP bars attached at 152 mm spacing in between the steel stirrups on both sides of the beam. The average spacing of these CFRP bars were calculated as 191 mm (7.5 in.).

Table 1. bars.

Mechanical properties of No. 10 ASLAN CFRP

Cross Tensile sectional Nominal Tensile modulus Ultimate Bar area diameter strength of elasticity strain size (mm2 ) (mm) (MPa) (MPa) (%) 10

65.2

9

2068

124,000

0.017

223

Figure 3.

Beam #1 under four-point bending test.

Figure 4.

Beams after shear failure.

All four beams were 2134 mm long, 254 mm wide and 305 mm deep. Only the middle 1829 mm of the beam were simply supported leaving 152 mm at each end to provide adequate space for the supports. Shear reinforcement (steel stirrups and CFRP bars) with strain gages were located at the critical shear locations of a four-point bending test setup. Strain gages were also attached to all concrete beams at midspan, to measure the concrete strain during loading to failure. The loading set up and the control beam under four-point bending test is shown in Fig. 3. 3.2

Testing and data acquisition

All strain gages were connected to the data acquisition (DAX) system before applying the load. Loads were applied incrementally until the beam failed. All four beams have failed in shear, as shown in Fig. 4, at different failure loads. Strain data with corresponding loads for each beam were recorded from the beginning to the failure of the beams using the DAX system. The collected data were later transferred to a spreadsheet for further analysis. 4 4.1

TEST RESULTS Analysis of collected data

The shear span ratio (a/d), which is defined as the ratio of shear distance (a) to the effective depth (d) of the member, plays important roles in the effectiveness of

Table 2.

500 450 400 350 Load (kN)

CFRP bars used in shear strengthening. Shear steel is internal and integrally built with concrete beam, and on the other hand, CFRP bars, which have almost eight times higher ultimate strain compared to that of the shear steel, were attached externally. Therefore, ratio of effective strain to ultimate strain in the external CFRP bars compared to that of internal shear steel is expected to be less. Shear span ratios for the selected strain gages are shown in Table 2. Shear steel almost reached its ultimate strain before failure, whereas effective strains in the CFRP bars are within a range of 30 to 35% of its ultimate strain before shear failure of the beams. This ratio of effective strain to ultimate strain is very similar to that observed by Adhikari et al. (2004) (19 to 27%), Triantafillou and Antonopoulos (2000) (23 to 31%), and Khalifa et al. (1998) (26 to 36%), as reported by Adhikari et al. (2004), where externally bonded CFRP laminates were used in strengthening the concrete members. It is to be noted that the strains included in Table 2 represent the highest strain for each individual beam. The theoretical failure load on the control beam was 353 kN, and the beam failed at 365 kN, which is within a range of 3.4%. The nominal shear strength of the control beam was 178.3 kN and the beam failed at a shear force of 182.5 kN, which is within 2.3% of the prediction. The actual shear force at failure for the model beams are also within a range of less than 2% of the

300 250 Beam 2

200

Beam 3

150

Beam 4

100 50 0 0

1000

2000

3000

4000

5000

6000

7000

CFRP Strain (10E-6)

Figure 5.

Load versus vertical strain in CFRP bars.

predicted shear strength. The predicted and the failure data are presented in Table 2. The failure modes of all four beams were in shear with almost diagonal shear cracks, as indicated by these results and the failure pictures shown in Fig. 4. The above results provide validity of this experimental and analytical research. As a result of using the NSM technique in attaching CFRP bars to strengthen concrete beams in shear, there is a significant increase in shear strength of the concrete members, which ranges from 17 to almost 25% of its original strength according to the data shown in Table 2. Therefore, NSM technique of attaching FRP bars to strengthen concrete beams in shear seems quite effective, and may be considered as one of the efficient methods for shear strengthening of concrete members.

Analysis of experimental test data.

4.2

Beam

Conc. comp. strength, fc (MPa) Effective depth, d (mm) Shear span ratio, a/d Shear reinf. material Spacing, s (mm) Effect. strain, εe × 10−3 Ult. strain, εu × 10−3 Ratio, εe /εu (%) Shear strength by conc., Vc (kN)∗ Shear strength by reinf.∗∗ , V (kN) Nominal shear strength, Vn (kN) Failure load (kN) Shear failure load (kN) Shear strength increase (%)

1

2

3

49.75

49.75

49.75

260.76

260.76

260.76

1.17 Steel 152.4 1.959

1.76 CFRP 152.4 5.474

1.76 Steel 304.8 5.891

2.07 94.6 77.9

16.7 32.8 77.9

16.7 35.3 77.9

100.5

151.4

50.2

178.3

229.3

209.6

365 182.5

454 227

427 213.5

–

24.4

17

∗ Calculated using Equation 2. ∗∗ Calculated using Equations 3

and 4.

CFRP 304.8

81.5

Load strain relationship

The load versus CFRP strain relationship for the three model beams is shown Fig. 5. The maximum failure loads for Beam 2, 3, and 4 were recorded as 454, 427 and 436 kN, with corresponding maximum strain of 5,474, 5,891, and 5,103 micro-strain, respectively. These values are shown in Table 2. It is interesting to note that the strain is almost negligible up to a load of approximately 180 kN, which is close to the modulus of rupture of the concrete in tension. Once the concrete in the shear region cracked diagonally, the shear force was transferred to the CFRP bars, and a rapid increase in strain in the CFRP bars was observed. In almost vertical part of the load-strain curve, the beams behave as elastic. In the next part of the curve, strain in CFRP bars increases sharply with a little increase in the applied load, and the relationship is almost linear, which is identical to the behavior of CFRP bars under tensile stress. 5

DISCUSSIONS

The load versus concrete compressive strain relationship for all four beams is shown in Fig. 6. The strain in concrete under compression increases almost linearly

224

Table 3. Comparison of effective and proposed shear strengths provided by CFRP bars.

500 450 400 Load (kN)

350

Beam

300

Beam 1

250

Beam 2

200

Beam 4

100 50 0 0

500

1000

1500

2000

2500

Concrete Strain (10E-6)

Figure 6.

Load versus concrete strain.

with the increase in load, which is an accepted practice. The maximum strain for the four beams were recorded as 1,293, 1,923, 1,517, and 1,582 micro-strain for Beam 1, 2, 3, and 4, respectively. 5.1

3

4

Effective depth, d (mm) Shear reinf. material Spacing, s (mm) Area CFRP bars used, Af (mm2 ) Effective strain, εef × 10−3 Ultimate strain, εuf × 10−3 Shear strength*, Vf (kN) Shear strength proposed**, Vf (kN) Difference (%)

260.76 CFRP 152.4 130.4

260.76 CFRP 304.8 130.4

260.76 CFRP 190.5 130.4

5.474 16.7 151.4 153.8

5.891 16.7 81.5 76.9

5.103 16.7 112.9 123.0

∗

2

−6

9

Calculated using Eq. 4. Calculated using proposed formula in Eq. 5.

∗∗

Performance of NSM technique

No debonding or fracture of CFRP bars was observed at failure. Delamination, debonding or fracture of externally bonded FRP materials have been reported at failure by almost every researcher for concrete members strengthened in that manner. Therefore, NSM technique of attaching FRP bars to strengthen concrete members in shear seems more effective compared to the externally bonded cases. As a result of attaching CFRP bars using the NSM technique, the increase in shear strengths of the model beams compared to the control beam were calculated as 24.4, 17.0 and 19.5% for Beam 2, 3 and 4, respectively. The average gain in shear strength is more than 20%. Therefore, the NSM technique is expected to perform well in enhancing shear strength of sheardeficient concrete members. It is also noticeable from Table 2 that with the same spacing, shear strength provided by CFRP bars is almost 50% more than that provided by steel bars (Beam 1 and 2). The ratio of effective strain to ultimate strain of CFRP bars in Beams 2, 3 and 4 are 32.8, 35.3 and 30.6%, respectively, as shown in Table 2. On average at failure, almost one-third of the ultimate strain becomes effective in the CFRP bars attached with the concrete beams following the NSM technique. Therefore, the following formula in Equation 5 is being proposed to calculate, Vf , the nominal shear strength provided by NSM CFRP bars used in shear strengthening of concrete members. Vf =

2

Beam 3

150

1 Af fyf d 3 s

(5)

where fyf = tensile yield strength of FRP bars. Comparison of shear strengths calculated by Eq. 4 and the proposed formula in Eq. 5 is shown in Table 3. The proposed formula calculates the shear strength

225

provided by the NSM CFRP bars used in shear strengthening of concrete members within an acceptable range and accuracy of 2 to 9% of the experimental values. 6

CONCLUSIONS AND RECOMMENDATIONS

The experimental studies on shear strengthening of concrete members using NSM CFRP bars, as performed herein, are expected to provide useful information about FRP design procedures, methods and guidelines. The following conclusions can be made from the outcomes of this experimental and analytical research: – CFRP bars attached using NSM technique is very effective in shear strengthening of concrete members. As a result of strengthening the concrete beams in shear using NSM CFRP bars, increase of shear strength has been found in the range of 17 to 25%. The average gain in shear strength was calculated to be more than 20%, which is significant in strengthening of shear-deficient concrete members. Effective strains in the NSM CFRP bars until beam failure were almost one-third of their ultimate strain. – No delamination, debonding or fracture of FRP materials, which are very common in case of externally bonded FRP materials, was observed. – A formula was proposed to calculate the nominal shear strength provided by NSM CFRP bars used in strengthening concrete members in shear. Although the proposed formula is not supported by a large number of test samples, it is expected to provide important guidelines for the future researchers to refine the equation, if necessary. The study involved only four concrete beams that include one control beam and three model beams.

In order to establish and adopt the relationship proposed herein to calculate shear strength provided by NSM FRP bars used to strengthen concrete beams in shear, more tests with larger number of samples, different beam sizes and various FRP materials may be conducted. The proposed formula may only be applicable to NSM CFRP bars. More studies are required involving other types of FRP bars. ACKNOWLEDGEMENTS The author is grateful to Youngstown State University for providing funds for the project through the University Research Council Grant 2006-2007 #06. Materials and labor supports provided by AP O’Horo Construction, BASF Company, Hughes Brothers, one graduate and three undergraduate students, and college of engineering laboratory technicians are greatly acknowledged and appreciated. NOTATIONS Vn = nominal shear strength Vc = nominal shear strength provided by concrete Vs = nominal shear strength provided by shear steel Vf = nominal shear strength provided by FRP bars fc = specified compressive strength of concrete bw = beam web width d = distance from extreme compression fiber to centroid of longitudinal tension reinforcement Av = area of shear steel fy = tensile yield strength of shear steel fyf = tensile yield strength of FRP bars s = spacing of shear reinforcement Af = area of FRP bars in shear Ef = tension modulus of elasticity of FRP bars εef = effective strain in FRP bars εuf = ultimate strain in FRP bars εe = effective strain εu = ultimate strain

REFERENCES ACI Committee 440. 2002. ‘‘Guide for the Design and Construction of Externally Bonded FRP Systems for Strengthening Concrete Structures,’’ American Concrete Institute, Farmington Hills, Michigan. ACI Committee 318-05. 2005. ‘‘Building Code Requirements for Structural Concrete (ACI 318-05) and Commentary (ACI 318R-05),’’ American Concrete Institute, Farmington Hills, Michigan. Adhikary, B.B., Mutsuyoshi, H. and Ashraf, M. 2004. ‘‘Shear Strengthening of Reinforced Concrete Beams Using Fiber-Reinforced Polymer Sheets with Bonded Anchorage,’’ ACI Structural Journal, Vol. 101, No. 5, pp. 660–668.

Alkhrdaji, T., Nanni, A., Chen, G. and Barker, M. 1999. ‘‘Upgrading the Transportation Infrastructure: Solid RC Decks Strengthened with FRP,’’ Concrete International, American Concrete Institute, Farmington Hills, Michigan, Vol. 21, No. 10, pp. 37–41. Asplund, S.O. 1949. ‘‘Strengthening of Bridge Slabs with Grouted Reinforcement,’’ Journal of the American Concrete Institute, Vol. 20, No. 6, pp. 396–407. Atkinson, R.H. and Schuller, M.P. 1992. ‘‘Development of Injectible Grouts for the Repair of Unreinforced Masonry,’’ Proceedings of the Workshop on Effectiveness of Retrofitting of Stone and Brick Masonry Walls in Seismic Areas, Politecnico di Milano, Milan, Italy. BASF Building Systems, The Chemical Company. 2006. Product Data—Concresive 1420, Shakopee, Minnesota. Bousselham, A. and Chaallal, O. 2004. ‘‘Shear Strengthening Reinforced Concrete Beams with Fiber-Reinforced Polymer: Assessment of Influencing Parameters and Required Research,’’ ACI Structural Journal, Farmington Hills, Michigan, Vol. 101, No. 2, Mar.–Apr. pp. 219–227. CEN. 1991. ‘‘Eurocode 2: Design of Concrete Structures— Part 1-1. General Rules and Rules for Buildings,’’ ENV 1992-1-1, Comite’ Europee’n de Normalisation, Brussels, Belgium, 252 pp. Chaallal, O., Shahawy, M. and Hassan, M. 2002. ‘‘Performance of Reinforced Concrete T-Girders Strengthened in Shear with Carbon Fiber Reinforced Polymer Fabrics,’’ ACI Structural Journal, Farmington Hills, Michigan, Vol. 99, No. 3, May 1, pp. 335–343. Colotti, V., Spadea, G. and Swamy, N. 2004. ‘‘Analytical Model to Evaluate Failure Behavior of Plated Reinforced Concrete Beams Strengthened for Shear,’’ ACI Structural Journal, Farmington Hills, Michigan, Vol. 101, No. 6, Mar.–Apr. pp. 755–764. De Lorenzis, L., Nanni, A. and La Tegola, A. 2000. ‘‘Flexural and Shear Strengthening of Reinforced Concrete Structures with Near Surface Mounted FRP Bars,’’ Proc., Third Int. Conf. on Advanced Composite Materials in Bridges and Structures, Ottawa, Canada, pp. 521–528. El-Hacha, R. and Rizkalla, S.H. 2004. ‘‘Near-SurfaceMounted Fiber-Reinforced Polymer Reinforcements for Flexural Strengthening of Concrete Structures,’’ ACI Structural Journal, Farmington Hills, Michigan, Vol. 101, No. 5, Sep.–Oct. pp. 717–726. Garrity, S.W. 1995. ‘‘Retro-Reinforcement—A proposed repair System for Masonry Arch Bridges’’, Proceedings of the First International Conference on Arch Bridges, Bolton, UK, pp. 557–566. Hughes Brothers, Aslan 200 CFRP Bars. 2007. ‘‘Installation of Aslan 200 for NSM Strengthening,’’ Design Examples and Application Photos of Aslan 200 CFRP Bars, Seaward, Nevada.URL: Joint ACI-ASCE Committee 445. 1998. ‘‘Recent Approaches to Shear Design of Structural Concrete,’’ Journal of Structural Engineering, ASCE, Vol. 124, No. 12, pp. 1375–1417. Khalifa, A., Gold, W.J., Nanni, A. and Abdel Aziz, M.I. 1998. ‘‘Contribution of Externally Bonded FRP to Shear Capacity of Flexural Members,’’ Journal of Composites Construction, ASCE, Vol. 2, No. 4, pp. 195–203. Khalifa, A. and Nanni, A. 2000. ‘‘Improving Shear Capacity of Existing RC T-Section Beams Using CFRP Composites,’’ Cement and Concrete Composites, Vol. 22, No. 3, pp. 165–174.

226

Malek, A.M. and Saadatmanesh, H. 1998. ‘‘Ultimate Shear Capacity of Reinforced Concrete Beams Strengthened with Web-Bonded Fiber-Reinforced Plastic Plates,’’ ACI Structural Journal, Vol. 95, No. 4, July–Aug., pp. 391–399. Matthys, S. 2000. ‘‘Structural Behavior and Design of Concrete Members Strengthened with Externally Bonded FRP Reinforcement,’’ Ph.D. Dissertation, Department of Structural Engineering, Faculty of Applied Science, Ghent University, Belgium. Multidisciplinary Center for Earthquake Engineering Research (MCEER) at the University of Buffalo. 2005. ‘‘Preliminary Damage Reports on Bridges (due to Hurricane Katrina),’’ New York, September 2005. Nanni, A., Di Ludovico, M. and Parretti, R. 2004. ‘‘Shear Strengthening of a PC Bridge Girder with NSM CFRP Rectangular Bars,’’ Advances in Structural Engineering, Vol. 7, No. 4, pp. 97–109.

227

Parretti, R. and Nanni, A. 2004. ‘‘Strengthening of RC Members Using Near-Surface Mounted FRP Composites: Design Overview,’’ Final Draft Report Submitted to ACI Committee 440. Triantafillou, T.C. 1998. ‘‘Shear Strengthening of Reinforced Concrete Beams Using Epoxy-Bonded FRP Composites,’’ ACI Structural Journal, Vol. 95, No. 2, Mar.–Apr., pp. 107–115. Triantafillou, T.C. and Antonopoulos, C.P. 2000. ‘‘Design of Concrete Flexural Members Strengthened in Shear with FRP,’’ Journal of Composites for Construction, ASCE, V. 4, No. 4, pp. 198–205. Tumialan, G., Morbin, A., Nanni, A. and Modena, C. 2001. ‘‘Shear Strengthening of Masonry Walls with FRP Composites,’’ COMPOSITES 2001 Convention and Trade Show, Composites Fabricators Association, Tampa, FL, CD-ROM, October 3–6, 6 pp.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Effect of transverse reinforcing on circular columns confined with FRP G. Ghodrati Amiri Center of Excellence for Fundamental Studies in Structural Engineering, School of Civil Engineering, Iran University of Science & Technology, Tehran, Iran

A. Jaberi Jahromi & B. Mohebi School of Civil Engineering, Iran University of Science & Technology, Tehran, Iran

ABSTRACT: In recent years retrofitting of bridges against earthquake is interested and many works have been done for strengthening every part of bridges. One of very important parts of a bridge is piers. Piers in many bridges are main earthquake resisting system. Therefore they play an important rule during earthquakes and failure of them can cause collapse of entire deck and other parts of bridge. Strengthening of bridges piers’ can be done by several methods. One of the most common methods is retrofitting RC columns with FRP jacketing. Any piers of a bridge can be improved against lack of strength or ductility. Usually RC columns in bridges are circle shaped. In this study it has been tried to investigate effect of amount of transverse reinforcing before retrofitting columns on nonlinear behavior of confined columns with FRP. For performing this research, 4 Finite element models with different amount of transverse reinforcing has been created and results have been presented in this paper. 1

GENERAL INSTRUCTIONS

When reinforced concrete columns are subjected to seismic loading, the large lateral cyclic earthquake force will degrade the concrete and the reinforcing bar very quickly, and the columns will fail prematurely. Investigations of bridge failures during recent earthquakes, such as the 1987 Whittier, 1989 Loma Prieta, 1994 Northridge, and 1995 Kobe show that inadequate lateral reinforcement and insufficient lap length of the starter bars are among the major catastrophic causes of failure (Priestley & Whitter 1988; Buckle 1994; Priestley & Seible 1996; Parvin & Wang 2002). The main resisting system against earthquakes in many bridges is piers. In this kind of bridges all lateral force induced to bridge will be transferred to piers. Therefore any failure in this member results the overall collapse in bridge. In recent years, the use of externally bonded fiber-reinforced polymers (FRP) has been become increasingly popular for civil infrastructure applications, including wrapping of concrete columns. Significant research has been devoted to circular columns retrofitted with FRP and numerous models were proposed (Khaloo et al. 1991; Fardis & Khalili 1981; Miyaushi et al. 1997; Samman et al. 1998; Spoelstra et al. 1999). FRP wrapping of existing circular columns has proven to be an effective retrofitting technique (Seible et al. 1997; Chaallal et al. 2003). Retrofitting bridge columns is done because several reasons. One of these reasons is the lack of transverse reinforcing. In this case the shear

229

capacity of section is not sufficient for resisting against earthquake forces, transferred to the column from deck and other masses of bridge. On the other hand lacking of transverse reinforcing can cause decrease in deformation capacity in plastic hinges of bridges and it will lead to collapse of columns in plastic hinge regions. As mentioned before, for overcoming this deficiency, many engineers use FRP jacketing. FRP jacketing can increase strength of column and deformation capacity of it simultaneously. It has been shown that confinement with fiber reinforced polymer (FRP) improves the behavior of columns submitted to seismic loading (Priestley et al. 1992; Katsumata et al. 1987). FRP fabric wraps consisting of carbon, aramid, or glass fibers bonded by an epoxy resin have been successfully applied for seismic rehabilitation of bridge piers in the U.S. and Japan (Mufti et al. 1992; Kasei 1993). The growing use of FRP composites as confinement elements is attributed to the important mechanical and chemical properties of these materials. Some of these advantages are light weight, high-tensile strength and modulus, corrosion resistance, and durability. These advantages make FRP composites suitable for use in coastal and marine structures like river bridge piers. In additions, their low density is important because it adds less weight to the existing structures, and because the use of heavy equipment for repair with FRP composites is not necessary during rehabilitations. It has been shown that wrapping FRP fabrics around the perimeter of both circular and rectangular concrete columns to create a confinement effect

improves ductility and strength (Katsumata & Kimura 1990; Picher et al. 1996). Other FRP confinement techniques have been shown to improve the behavior of normal and high-strength concrete (Harmon et al. 1995). Retrofitting of concrete columns by lateral confinement with FRP wires has also resulted in an increase in strength and ductility under uniaxial compression (Nanni et al. 1994; Houssam & Toutanji 1999). Amount of lacking of transverse reinforcing is different in different case studies. In this research 4 samples with different amount of transverse reinforcing have been retrofitted by FRP jacketing method. The kind and thickness of FRP in all specimens are similar and the only difference between them is the amount of transverse reinforcing. In next part of this paper the properties of FRP is presented. After that specification of samples and analyze method have been explained. Finally the comparison between results has been done. 2 2.1

2.2

It is unavoidable to use of steel bars because the concrete is weak in tension. Strength and deformation properties of steel bars usually obtain from strainstress curves. Hot rolled steel bars often have a specific yield point and at the time of rupture show significant strain; for this reason sometimes called mild steels. For steel modeling, usually apply three plasticity models (Ansys Standard Users Manual): 2.2.1 Steel with isotropic hardening This model is very simple and often used when loads are monotonic. In the mentioned model, Bauschinger effect is neglected and yield area is changed isotropically. 2.2.2 Steel with kinematic hardening This state is a little more complicated than the isotropic model considers the Bauschinger effect and consisting of bilinear hardening model. In this case, only the center of yield area is transformed in space and not changed isotropically.

DESCRIPTION OF MECHANICAL PROPERTIES OF MATERIALS Mechanical properties of concrete

Essentially concrete is an inelastic material thus behaves elastic in many respects. Compression strength is only one specification of concrete materials. For explanation of concrete behavior, the strain-stress curve could be useful. This curve explains that in low stresses, concrete behaves elastic. This curve is inclined horizontally in higher stresses and strain reaches to 0.002 in ultimate stress. After ultimate stress, the curve generally is descending to rupture point. As much as concrete strength would be low, the ductility property is higher. The concrete used in the present study for all the specimens is 25 MPa and the module of elasticity is the same with value of 210 MPa. Thickness of concrete cover is 2.5 cm in models. Stress strain curve that have been used for confined concrete in finite elements model is mander curve which is shown in Figure 1 (Mander et al. 1988). 250

Mander

200 150 100 50 0 0

Figure 1.

0.002

0.004

0.006

0.008

Stress-strain curve for confined concrete.

Mechanical properties of steel bars

2.2.3 Steel with combined hardening This model is the most exact of the steel models and is combination of two previous models. This means that yield area is changed isotropically and transformed in space. It results to improve the loadings cycles in addition to consideration of Bauschinger effect. For determination of this behavior is needed much parameters and their calibrations are difficult. Because steel is the secondary material (Concrete is the primary material), then kinematic hardening model is used to consideration of cyclic loads and it is not complicated such as compound model. 2.3

Mechanical properties of FRP

Usually a composite material is defined as a physical combination of two or some different materials in macroscopic scale. These materials keep their mechanical and chemical properties and form a specific boundary with each other. These materials have better specification than their components. Fiber Reinforced Plastic (FRP) products were first used to reinforce concrete structure in the mid 1950s. Thoday, these FRP products take the form of bars, cables, 2-D and 3-D grids, sheet materials, plates, etc. FRP products may achieve the same or better reinforcement objective of commonly used metallic products such as steel reinforcing bars, prestressing tendons, and bonded plates. Application and product development efforts in FRP composites are widespread to address the many opportunities for reinforcing concrete members (ACI 1996). Some mechanical properties of FRP is shown in Table 1. All the details about FRP material that used in the present modeling have been shown in Table 2.

230

Table 1.

Specifications of carbon, glass and aramid fibers.

Material

Modulus of elasticity Strength (GPa) (MPa)

Ultimate tensile strain %

Carbon High strength Ultra high Strength High modulus Ultra high modulus

215–235 215–235 115–130 500–700

3500–4800 3500–6000 3500–4000 2100–2400

1.4–2 1.5–2.3 2.5–3.5 0.2–0.4

70 85–90

1900–3000 3–4.5 3500–4100 4.5–5.5

70–80 115–130

3500–4100 4.3–5 3500–4000 2.5–3.5

Glass E S Aramid Low modulus High modulus

Table 2.

Height of columns Diameter Concrete cover thickness

120 GPa 1.5 GPa 3 GPa 40 GPa 0.5 mm 0.012

Figure 2.

DESCRIPTION OF SPECIMENS

All the models have similar geometrics properties. All models were circular with diameter of 70 cm and height of 300 cm. These properties have been shown in Table 3. Thicknesses of CFRP in all models are similar (0.5 mm).

4

Geometric details of models. 300 cm 70 cm 205 cm

FRP material details.

Modulus of elasticity in fibers direction Tensile strength in fiber direction Modulus of elasticity in vertical fibers direction Tensile strength in vertical fiber direction Thickness Ultimate strain

3

Table 3.

Three-dimensional modeling with its mesh.

wrinkling of the element in compression in one or both orthogonal directions (Mirmiran et al. 2000). ANSYS (Ansys Structural Nonlinearities Manual), used in this study, has a parametric design language that is useful for parametric input and automatic mesh generation. The following steps were taken in the modeling: 1. Geometric input consists of core diameter, specimen height, and number of elements. 2. Concrete, FRP, and reinforcement bars properties include compressive strength, tensile strength, yield stress, Poisson’s ratio and modules of elasticity.

FINITE ELEMENT MODELING

Concrete has been modeled by an 8-noded SOLID65 element, which consist of a single solid material and up to three smeared reinforcing materials in three different orientations. The solid material, i.e., plain concrete, is treated as an initially isotropic homogeneous material with different tensile and compressive strengths. It is also capable of cracking in tension and crushing in compression. Cracking can occur in any of the three orthogonal directions. The element can also accommodate plastic deformations and creep. The jacket is modeled by 4-noded linear elastic membrane SHELL41 element, which is a three dimensional shell element with membrane stiffness and three translational degrees of freedom per node. However, it does not have any bending stiffness, nor any rotational degree of freedom. The element can accommodate variable thickness, orthotropic behavior, stress stiffening, large deflection, and cloth option, the last of which is a nonlinear feature that constitutes

231

5

LOADING AND ANALYZE METHOD

After making four 3-D finite element models of RC columns, confined with FRP, by ANSYS, axial load statically submitted to the top of the columns, then lateral load as cyclic load submitted to the model with displacement control. As mentioned before all the characteristics of four models are the same, except transverse reinforcement ratio, as it has shown in Table 4.

6

VALIDATION BY EXPERIMENTAL MODEL

The validity of the proposed analytical model is checked through extensive comparisons between analytical and experimental results of RC columns confined with FRP under compression load. The experimental

7

RESULT AND COMPARISONS

A comparison between four finite element models has been done. Force-Lateral Displacement curves have been shown in Figure 5. Also hysteresis curves derived from finite element nonlinear analysis have been shown in Figures 6–9. As it shown in Figure 5

Figure 3.

Cyclic load.

Table 4.

Transverse reinforcing of models.

Model name

Transverse reinforcement ratio

Model I Model II Model III Model IV

0 0.004 0.006 0.008

Table 5.

Maximum lateral displacement in models.

Models

Maximum lateral displacement

I II III IV

30 mm 47 mm 48 mm 68 mm

Figure 5.

Comparison between results.

Figure 6.

Hysteresis curve for model I.

Figure 7.

Hysteresis curve for model II.

Figure 4. Comparison between finite element model and Samaan’s model.

data used herein were derived from Samaan et al. test (1998). For this aim, Samaan’s model, exactly modeled in finite element software (ANSYS).comparison between analytical and experimental model have been shown in Figure 4. As it is shown, there is a good relation between them.

232

REFERENCES

Figure 8.

Hysteresis curve for model III.

Figure 9.

Hysteresis curve for model IV.

in range 0.004 and 0.006 in transverse reinforcement, there is no difference. After the ratio of 0.006 there is a significant increase in the strength and maximum displacement of specimen. Hysteresis loops of models II and III have the same shape, but in model IV, there exist a good hysteresis model with more energy dissipation.

8

CONCLUSIONS

As it shown in figures, the amount of transverse reinforcing of FRP retrofitted column has no significant effect in column’s behavior, if amount of transverse reinforcing be in a regular range (e.g. 0.004–0.006). If amount of transverse reinforcing be more than usual, it has a good effect on columns behavior. In fact, if transverse reinforcing of a column be sufficient, FRP retrofitting has no significant effect on column’s behavior.

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ACI Committee 440R. 1996. State of the Art Report on Fiber Reinforcement for Concrete Structures. American Concrete Institute, Detroit. ANSYS Standard Users Manual Help,Ver.11 and ANSYS Structural Nonlinearities Manual. Ahmad, S.H., Khaloo, A.R. and Irshaid, A. Behavior of concrete spirally confined by Fiberglass Filaments. Magazine of Concrete Research, V. 43, No.156, 1991, pp. 143–148. Buckle IG (Ed). The Northridge, California earthquake of January 17, 1994: performance of highway bridge. Tech. Rep. NCEER-94-0008, Nat. Ctr. for earthquake Engrg. Res., state University of New York at Buffalo, NY, 1994. Chaallal, O., Shahawy, M. and Hassan, M. Confinement Model for Axially Loaded Short Rectangular Columns Strengthened With Fiber-Reinforced Polymer Wrapping, ACI Structural Journal, March–April 2003. Fardis, M.N. and Khalili, H.H. Concrete Encased in Fiberglass-Reinforced Plastic, ACI Journal, Proceedings V. 78, No. 6, Nov/Dec. 1981, pp. 440–446. Harmon, T., Slattery, K. and Ramakrishnan, S. The Effect of Confinement for Concrete Structures, Proceedings of the second International RILEM symposium, 1995, pp. 584–592. Houssam A, Toutanji: Stress-Strain Characteristics of Concrete Columns Externally Confined with Advanced Fiber Composite Sheets.ACI Material Journal, 1999, pp. 397–404. Kasei, M. Carbon Fiber Reinforced Earthquake-Resistant Retrofitting, Mitsubishi Kasei Crop., Japan 1993. Katsumata, H. and Kimura, K. Applications of Retrofit Method with Carbon Fiber for Existing Reinforced Concrete Structures, 22nd joint UJNR Panel Meeting, U.S.Japan Workshop, Gaithersburg, MD, 1990, pp. 1–28. Katsumata, H., Kobatale, Y. and Takeda, T. A Study on the Strengthening with Carbon Fiber for EarthquakeResistance Capacity of Existing Reinforced Concrete Columns, Proceedings of the seminar on Repair and Retrofit of Structures, U.S.-Japan Panel on wired and Seismic Effects, UJNR 1987, pp. 1816–1823. Mander, J.B., Priestley, M.J.N. and Park, R. 1988: Theoretical Stress-Strain model for confined concrete, Journal of structural engineering, V. 114, No. 8, pp. 1804–1826. Mirmiran A., Zagers K. and Yuan, W. Nonlinear finite element modeling of concrete confined by fiber composites. Elsevier 2000, pp. 79–96 Mufti, A.A., Eriki, M.A. and Jaeger, L.C. Advanced Composite Materials in Bridge and Structures in Japan, Canadian Society of Civil Engineering, Montreal, Canada, 1992. Nanni, A. Norris, M.S. and Bradford, N.M. Lateral Confinement of Concrete Using FRP Reinforcement, Fiber Reinforced Plastic Reinforcement for Concrete Structures, SP-138, American Concrete Institute, Farmington Hills, Mich., 1994, pp. 193–209. Parvin azadeh, Wang Wei. Concrete columns confined by fiber composite wraps under combined axial and cyclic lateral loads, 2002. Picher, F. Rochette, P. and Labossiere, P. Confinement of Concrete Cylinders with CFRP, Fiber Composites in Infrastructure, Proceedings, First International Conference on Composites in Infrastructure, 1996, pp. 829–841.

Priestly, M.J.N., Seible, F. and Fyfe. E. Column Seismic Retrofit Using Fiber Glass/Epoxy Jackets, Proceedings of the first International Conference on Advanced Composite Material in Bridge and Structures, Sherbrooke, Canada, 1992, pp. 287–298 Priestley MJN., Seible F., Calvi GM. Seismic design and retrofit of bridges. New York: Wiley; 1996. Priestley MJN. Whittier narrows, California earthquake of October 1, 1987-damage to the I-5/I-605 separator. Earthquake Spectra J 1988; 4(2): 389–405.

Samaan, M., Mirmiran, A. and Shahawy, M. 1998: Model of concrete confined by fiber composites., J. of Stru. Eng., ASCE, V. 124, No. 9, pp. 1025–1031, September 1998. Seible, F., Priestly, N., Hegemier, G.A. and Innamorato, D.: Seismic Retrofit of RC Columns with Continuous Carbon Fiber Jackets, Journal of Composites for Constructon, V. 1, No. 2, 1997, pp. 52–62. Spoelstra, M.R., and Monti, G. FRP-Confined Concrete Model, Journal of composite for construction, V. 3, No. 3, August 1999, pp. 884–888.

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Experimental investigation of FRP wrapped RC circular and square hollow columns M.N.S. Hadi University of Wollongong, Wollongong, NSW, Australia

Y. Kusumawardaningsih State University of Semarang, Semarang, Central Java, Indonesia University of Wollongong, Wollongong, Australia

ABSTRACT: An investigation was conducted to identify the stress-strain behaviour of FRP wrapped hollow reinforced concrete (RC) columns under concentric loading, focusing on circular and rectangular column sections having different shape of hole. Twelve RC columns consisted of six circular columns (two solid columns, two hollow columns having circular hole, and two hollow columns having square hole) and six square columns (two solid columns, two hollow columns having circular hole, and two hollow columns having square hole) were cast and tested. Six columns from each configuration were left unwrapped as control specimens, while the others were wrapped with FRP. It was found that FRP confinement in lateral direction increases the hollow RC columns’ compressive strength and axial strain. FRP jackets prevent premature failure of concrete cover and buckling outwards of steel bars, leading to significant improved performance of FRP concrete column composite. 1

INTRODUCTION

Reinforced concrete is one of the most widely used materials for constructing a wide range of buildings, bridges and other structures; due to its advantages such as high strength and durability, easy to cast into particular shapes, and economy. An increase in the demand of constructing high rise concrete structures has led people to use high strength concrete in lieu of normal strength concrete, to get stronger concrete structures and to minimize the size and weight of concrete members in the structure. Even though by using high strength concrete stronger concrete can be gained, the problem of big size concrete members in the structures still exists. While this will lead to the increase of the structure’s overall weight; and finally as other impacts it would decrease the structure ductility, deflection capacity, and increase the construction overall budget. According to Warner et al. (1998) weak ductility, low deflection capacity and tensile strength will cause concrete structures to collapse without any warning. On the other hand, increased ductility in concrete structures will allow the structure to display failure signs and warns inhabitants that the structure could fail in turn to save lives. In order to increase the strength and ductility of concrete it is necessary to use advanced materials and construction technique to improve concrete strength, structural longevity, and to reduce the weight of concrete members. Fiber Reinforced Polymers (FRP) is

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a popular material to be used as external confinement of concrete members for both strengthening and retrofitting purposes. By applying FRP material in concrete members, no significant increase in weight of structure will occur, while FRP material significantly enhances the concrete structure’s performance especially in terms of strength and ductility (Teng et al. 2001; Hadi 2005). Recently, many researchers have been conducting studies to investigate the compressive strength and stress-strain behaviour of FRP confined composite members due to the huge potential market of FRP applications for example: Spoelstra & Monti 1999; Pessiki et al. 2001; Teng et al. 2001; Pinto et al. 2003; Lignola et al. 2007; and Yazici & Hadi 2009. One possible technique to reduce the weight of concrete members is by making hollow members. The minimal amount of concrete used in hollow concrete members might be considered economical to minimize the cost of concrete. Some studies about hollow composite members including Hsu & Liang 2003, Pinto et al. 2003, and Lignola et al. 2007, have shown that hollow concrete members with low axial load ratio, moderate amount of longitudinal steel and reasonably concrete cover were found to perform in a ductile manner at the flexural strength, similar to the solid ones. However, current standards of practice have not answered specific problems regarding new designs of hollow members to be used; while hollow RC members have already been used in many existing bridge piers.

Since the use of FRP for confinement on RC concrete is relatively new, theoretical work in this area is limited to the models that already existed. The present paper develops fundamental knowledge of FRP external confinement on hollow circular and square high strength RC columns with different shapes of hole under concentric loading, so that the different behavior can be identified. Specifically, this paper shows the effect of the same net cross section areas on both circular and square hollow RC columns under axial concentric loading.

2

EXPERIMENTAL WORK

2.1 Column specimen layout By considering the possibility of making column specimens that have the desired similar cross section areas based on the availability of PVC formwork in the market, and after being calculated; a net cross section of 29,252 mm2 was chosen to make all column specimens (except for Column CCC and CCF that are very difficult to be made for having the same net cross section area due to the diameter limitation of PVC formwork that are available in the market). For square columns, to achieve better efficiency and to eliminate regions with ineffectively confined concrete, the four corners of square columns were rounded by a corner radius of 20 mm (Mirmiran & Shahawy 1997). Concrete (provided by a local supplier) having an average 28-day cylinder concrete compressive strength of 71.80 MPa, deformed and plain steel bars having tensile strengths of 669 and 476 MPa, respectively; were used to make a total of 12 RC column specimens having a length of 925 mm, consisted of six circular and six square columns. From each configuration, a total of six columns were left unwrapped as control specimens, while the others were externally wrapped with two layers of CFRP (Carbon Fiber Reinforced Polymers). The CFRP strips used to externally wrap the columns were specified having a width of 75 mm, and the CFRP strips wrapped the columns at 10◦ angle to the horizontal; bonded with a combination of epoxy and hardener with a mix ratio of 5:1. Figures 1 and 2 show the cross section detail of the column specimens.

Figure 1. Reinforcement details of circular columns (all units in mm).

2.2 Instrumentation

Figure 2. Reinforcement details of square columns (all units in mm).

In order to investigate the stress-strain behavior of FRP wrapped hollow high strength RC columns on circular and rectangular column section having different shape of hole; an experimental investigation was carried out in the University of Wollongong. The Denison 500 tonne Compressive Testing Machine having a maximum loading capacity of 5000 kN was used to apply concentric loading to the

columns. All columns were loaded under an increased concentric load applied under displacement control, using displacement rate 0.3 mm/minute and adjusted to 0.5 and 0.7 mm/minute once the load change was insignificant. The data was recorded every two seconds. Strain gauges having a length of 5 mm were

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Figure 4. Stress–axial strain curves of Circular Columns CCC and CCF.

Figure 3.

Test setup.

attached on the outside of both the longitudinal and lateral bars, in the middle of columns (463 mm from the top of columns), to investigate the strain of columns. Figure 3 shows the test setup of one column specimen. 2.3

Figure 5. Stress–axial strain curves of Circular Columns CCSH and CCSHF.

Observed behavior

Failure of column specimens was generally marked by spalling of concrete cover at or near the midheight of the columns, followed by buckling outward of vertical reinforcement and rupture of lateral reinforcement around the middle of the columns. In columns wrapped with CFRP, snapping sounds were heard before the ultimate failure, revealing the rupture of CFRP composites and debonding between the layers of CFRP wrapping and concrete column. This failure was explosive but not sudden, and the reinforcement where the CFRP composites failed, experienced buckling outward. Column specimens wrapped with CFRP remained intact after failure. Failure near the top and bottom edges of the specimens was avoided by strengthening the edges of specimens with additional confinement of CFRP. 2.4

Comparison of columns’ behavior

2.4.1 Unwrapped and CFRP wrapped columns Figures 4–9 show the different stress-strain of columns unwrapped and wrapped with CFRP. Stresses were calculated simply by dividing the applied load by

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Figure 6. Stress–axial strain curves of Circular Columns CCCH and CCCHF.

the cross section area, while strains were defined as the column’s axial deformation divided by its original length. According to these figures, generally, columns wrapped with CFRP experienced higher maximum stresses capacity and larger axial strains

Figure 7. Stress–axial strain curves of Square Columns CSC and SCF.

Figure 8. Stress–axial strain curves of Square Columns SCSH and SCSHF.

Figure 9. Stress–axial strain curves of Square Columns SCCH and SCCHF.

compared to columns that are not wrapped with CFRP. However, circular columns have much better performance in stress carrying capacity compared to square columns. In columns without CFRP wrapping; when the stresses increase, the columns’ axial strains will

also increase. However, once columns reach their maximum stress capacity, the columns will experience significant decreases in their stress carrying capacity. The columns’ axial strains still expand until they reach their ultimate strain capacity and then fail. In circular columns wrapped with CFRP (see Figures 4–6), it is very interesting that compared to those without CFRP wrapping, circular columns wrapped with CFRP have more capability to experience larger axial strain (demonstrating an increased ductility) before they finally reach their maximum stress capacity and experience significant vertical decreases in their stress carrying capacity. Figure 4 shows that Column CCF (solid circular column wrapped with CFRP) has higher maximum stress and axial strain namely 114.66 MPa and 0.005526, respectively; compared to Column CCC (solid circular column without CFRP wrapping) that has a maximum stress of 76 MPa and an axial strain of 0.005286. Figure 5 shows that Column CCSHF (circular column with a square hole, wrapped with CFRP) has a higher maximum stress and axial strain namely 91.13 MPa and 0.007696, respectively; compared to Column CCSH (circular column with a square hole, without CFRP wrapping) that carries a maximum stress of 61.15 MPa and an axial strain of 0.004783. Figure 6 shows that Column CCCHF (circular column with a circular hole, wrapped with CFRP) carries a higher maximum stress and axial strain namely 91.19 MPa and 0.010334, respectively; compared to Column CCCH (circular column with a circular hole, without CFRP wrapping) that has a maximum stress of 68.07 MPa and an axial strain of 0.005209. In square columns wrapped with CFRP (see Figures 7–9), compared to those without CFRP wrapping, square columns wrapped with CFRP tend to have more capability to receive repeated large loadings and stresses (demonstrating an increased ductility) before they finally reach their maximum stress capacity and experience a significant vertical decrease in stress carrying capacity. Figure 7 illustrates that Column SCF (solid square column wrapped with CFRP) has a higher maximum stress and axial strain namely 87.24 MPa and 0.006175, respectively; compared to Column CSC (solid square column, without CFRP wrapping) that has a maximum stress of 75.07 MPa and an axial strain of 0.005637. Figure 8 shows that Column SCSH (square column with a square hole, without CFRP wrapping) has slightly higher maximum stress and axial strain namely 73.86 MPa and 0.005861, respectively; compared to Column SCSHF (square column with a square hole, wrapped with CFRP) that carries a maximum stress of 72.34 MPa and an axial strain of 0.005816. However, it can also be seen in Figure 8 that after its first maximum stress is reached, Column SCSHF can experience one more maximum stress (that has a value slightly below its first maximum stress) before its stress capacity

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decrease sharply. This condition shows that the CFRP confinement in Column SCSHF works to prevent re lated column from early failure. Figure 9 illustrates that Column SCCHF (square column with a circular hole, wrapped with CFRP) has a higher maximum stress and axial strain namely 78.43 MPa and 0.007147, respectively; compared to Column SCCH (square column with a circular hole, without CFRP wrapping) that has a maximum stress of 70.54 MPa and an axial strain of 0.005491. These analyses clearly demonstrate that CFRP external confinement in both circular and square columns can enhance the ultimate compressive strength and axial strain of the columns by delaying rupture of the concrete and reinforcement.

Figure 12. Stress–axial strain curves of Square Columns SCCH and SCSH.

2.4.2 Columns with different shape of hole Figures 10–13 show the stress-strain behavior of column specimens which have different shape of hole. According to these figures, generally, hollow columns

Figure 13. Stress–axial strain curves of Square Columns SCC HF and SCSHF.

Figure 10. Stress–axial strain curves of Circular Columns CCC H and CCSH.

Figure 11. Stress–axial strain curves of Circular Columns CCC HF and CCSHF.

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that have circular holes show higher maximum stresses and axial strain carrying capacity. According to Figures 10 and 11, in hollow circular columns; columns having circular holes (Columns CCCH and CCCHF) have better performance in terms of maximum stress and axial strain carrying capacity, compared to circular columns which have square holes (Columns CCSH and CCSHF). Figure 10 illustrates that Column CCCH (circular column with a circular hole) has higher maximum stress and axial strain namely 68.07 MPa and 0.005209, respectively; compared to Column CCSH (circular column with a square hole) that has a maximum stress of 61.15 MPa and an axial strain of 0.004783. Figure 11 shows that Column CCCHF (circular column with a circular hole, wrapped with CFRP) has slightly higher maximum stress and axial strain namely 91.19 MPa and 0.010334, respectively; compared to Column CCSHF (circular column with a square hole, wrapped with CFRP) that has a maximum stress of 91.13 MPa and an axial strain of 0.00766.

On the other hand, in this study, different behavior occurs in hollow square columns without CFRP wrapping (see Figure 12). According to Figure 12, Column SCSH (square column with a square hole) has better performance in maximum stress and axial strain carrying capacity, compared to Column SCCH (square column with a circular hole). Based on Figure 12, Column SCSH has slightly higher maximum stress and axial strain namely 73.86 MPa and 0.005861, respectively; compared to Column SCCH that has a maximum stress of 70.54 MPa and an axial strain of 0.005491. Similar to circular columns, in hollow square column (see Figure 13); Column SCCHF (square column with a circular hole, wrapped by CFRP) has better performance in terms of maximum stress and axial strain carrying capacity, compared to Column SCSHF (square column with square hole, wrapped by CFRP). Based on Figure 13, Column SCCHF has higher maximum stress and axial strain namely 78.43 MPa and 0.007147, respectively; compared to Column SCSHF that has a maximum stress of 72.34 MPa and an axial strain of 0.005816. 3

CONCLUSIONS

The following conclusions are drawn from the experimental work: 1. CFRP external confinement in both solid and hollow columns allows columns to experience larger loadings and demonstrate an increased stress and axial strain carrying capacity. 2. Under concentric loading, external confinement of RC columns with CFRP can significantly improve the columns’ performance, by delaying rupture of the concrete and reinforcement. 3. Generally speaking, hollow columns which have circular holes can carry higher maximum stresses and axial strain, compared to hollow columns which have square holes.

REFERENCES Hadi, M.N.S. 2005. Behaviour of FRP Wrapped HSC Columns under Loads with Different Eccentricities. Collaboration and Harmonization in Creative Systems; Proceeding of the 3rd International Structural Engineering and Construction Conference (ISEC-03), Shunan, 20–23 September 2005. Japan: Balkema. Hsu, H.L. & Liang, L.L. 2003. Performance of Hollow Composite Members Subjected to Cyclic Eccentric Loading. Earthquake Engineering and Structural Dynamic, Vol. 32: 443–461. Lignola, G.P., Prota, A., Manfredi, G. & Cosenza, E. 2007. Experimental Performance of RC Hollow Columns Confined with CFRP. Journal of Composites for Construction, Vol. 11: 42–49. Mirmiran, A. & Shahawy, M. 1997. Behavior of Concrete Columns Confined by Fiber Composites. Journal of Structural Engineering, Vol. 123: 583–590. Pessiki, S., Harries, K.A., Kestner, J.T., Sause, R., Ricles, J.M. 2001. Axial Behavior of Reinforced Concrete Columns Confined with FRP Jackets. Journal of Composites for Construction, Vol. 5: 237–245. Pinto, A.V., Molina, J., Tsionis, G. 2003. Cyclic Tests on Large-Scale Models of Existing Bridge Piers with Rectangular Hollow Cross-Section. Earthquake Engineering and Structural Dynamic, Vol. 32: 1995–2012. Spoelstra, M.R. & Monti, G. 1999. FRP—Confined Concrete Model. Journal of Composites for Construction, Vol. 3: 143–150. Teng, J.G., Chen, J.F., Smith, S.T., Lam, L. 2001. FRP— Strengthened RC Structures. Chichester: Wiley. Warner, R.F. Rangan, B.V. Hall, A.S., Faulkes, K.A. 1998. Concrete Structures. Melbourne: Longman. Yazici, V. & Hadi, M.N.S. 2009. Interaction Diagrams for FRP Wrapped Circular Hollow Columns. Proceedings of the 20th Australasian Conference on the Mechanics of Structures and Materials, Toowoomba, 2–5 Dec 2008. Australia: CRC Press.

ACKNOWLEDGMENTS The authors wish to thank the University of Wollongong for the financial support provided to this project. The contribution of Messrs Joel Cahill, Di Zhou, and Robert Rowlan is highly appreciated.

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Numerical study on strengthening composite bridges K. Narmashiri & M.Z. Jumaat University of Malaya, Kuala Lumpur, Malaysia

ABSTRACT: This research examines the effects of using CFRP strips and steel plates for flexural strengthening of composite bridge. Computer simulation is used and adopted from a computer program called ANSYS. The specimens are modeled in full 3D case, thus the concrete deck, concrete reinforcements, steel profile, steel plates, shear connectors, adhesive and CFRP strips are modeled in 3D solid elements (Tetra 10 node 187). Non-linear static analysis method is used in this research. The steel profile, concrete section and girder length measurements are the same for all models. Polymer strips and steel plate’s thicknesses are different for each model, though. The results are compared in stress levels on CFRP strips or steel sheets, strain on adhesive and concrete, and deflection of the bridge girder. The results indicate that stress, strain, deflection, and load bearing in different samples are affected by type of material and thicknesses. 1 1.1

BACKGROUND Introduction

The various sorts of methods exist for strengthening different steel structures, such as putting up additional steel parts, applying steel plates, including external pre-stressing of the parts, reducing or bridging the gap between the supports, and applying Fiber Reinforced Polymer (FRP). Under normal circumstances, strengthening steel-concrete composite bridges is carried out using steel plates or CFRP sheets, but what is widely used today for this purpose is Fiber Reinforced Polymer (FRP). FRP is a composite substance treated with resin with qualities of high tension, light weight, and a high resistance to corrosion. That explains why it has been used so often for strengthening steel and concrete structures. FRP can be produced from Carbon Fiber Reinforced Polymer (CFRP) or Glass Fiber Reinforced Polymer (GFRP). However, applying (CFRP) for the mentioned purpose has proven to be more satisfactory than GFRP. High Modulus (HM) resin which results in greater resistance has been more popularly used recently. In this research, examination on the effects of applying CFRP and steel plates on the flexural strengthening composite bridge is tried. An analysis of the comparison is also done. Some samples were considered for this study. Computer simulation adopted from a computer program called ANSYS was also used. The samples are modeled in 3D case; so concrete decks, concrete reinforcements, steel profile, steel plates, shear connectors, adhesive and CFRP strips are modeled in 3D solid elements (Tetra 10 node 187). Non linear static analysis method is used in this research. The steel profile, concrete section and girder length measurements are the same for all

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models. Polymer strips and steel plate’s thicknesses are different for each model, though. The results of strengthening composite bridge, with CFRP or steel sheets are measured as indicated by the figures in the diagrams, and they indicate the effects of strengthening the bridge girder by the mentioned methods. These results are compared based on the stress exerted on CFRP or steel sheets, and the strain on the adhesive and concrete, and the deflection of the bridge girder. 1.2

Literature review

Recently, applying FRP for strengthening structures has become more popular because other strengthening methods have limitations especially for structures that are under loading. Strengthening steel-concrete bridges have been researched by some scientists. Edberg et al. (1996) studied the application of CFRP for strengthening steel girders. An efficient way of strengthening the girder has also been found by Gillespie et al. (1996.) The strengthening of girders regarding the corrosion effects and section reduction was examined by Liu et al. (2001) & Tavakolizadeh et al. (2001, 2003.) Strengthening steel-concrete composite bridge girders using carbon polymer was examined too by Sen et al. (2001) and Tavakolizadeh et al. (2001, 2003.) Also, applying high modulus carbon polymer in order to strengthen the steel-concrete composite bridge girder was examined by Schnerch et al. (2004, 2006, and 2007) and Dawood et al. (2006.) For the abovementioned studies though, longitudinal sheets were used for the strengthening process. The abovementioned researches showed that before the maximum tension strength level was reached, a state of debonding occurred on the longitudinal polymer sheets.

The effects of CFRP cutting shape on steel-concrete composites were also looked into by Schnerch et al. (2004, 2006, and 2007). The findings were further reinforced by Narmashiri et al. (2009), especially in the effects of end cutting shape of HM-CFRP on the stress and strain intensity. The effects of 2D and 3D computer simulation of strengthened steel-concrete bridges on stress, strain, and deflection for different parts of the girder were examined by M.Z. Jumaat et al. (2009.) In this research comparison between the application CFRP sheets and steel plates for strengthening steelconcrete composite bridges were examined, and the results present a better understanding of the effects of applying steel plates or CFRP sheets on the different parts of the girder. 2

Figure 2.

Shear connector location.

Figure 3.

Strengthened sample by CFRP/Steel plate.

Figure 4.

Three dimensional simulated sample.

MODELING

2.1

Computer simulation method

Finite element method by ANSYS program is used for simulation and analysis of the samples. The samples are modeled in 3D case, and concrete deck, concrete reinforcements, steel profile, steel plates, shear connectors, adhesive and CFRP strips are modeled in 3D solid elements (Tetra 10 node 187). Non linear static analysis method is used in this research. 2.2

Specification of the samples

Specifications of the samples are shown in Figures 1–4. The maximum value of the load P is 200 KN (2 point load 100 KN) which is applied to the structure step by step using the nonlinear analysis method. The length of the girder is 3000 mm, and the length of constant moment region is 800 mm. The length of CFRP sheet or steel plates are 2000 mm, and their widths are 50 mm. CFRP thicknesses are 1.4 or 2 mm, and the elasticity modulus is 300 MPa. The steel plate thicknesses are 6 or 8 mm, and the elasticity modulus is 210 MPa. The steel section is U152*152*30. The steel section and steel plates have the same material

Concrete

Steel bars

Shear studs

Steel section

CFRP/Steel PL Figure 1.

Specifications of the girder.

Stiffener

of low carbon. The shear studs diameter is 10 mm with length 50 mm, and their distances are 100 mm. The compression strength of concrete for the deck is 25 MPa. The deck width is 400 mm, and the thickness is 70 mm. The steel bar diameter is 10 mm and it is

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3 3.1

NUMERICAL ANALYSIS Stress on CFRP or steel plate

In this section the effects of strengthening composite bridge girder with CFRP or steel plate on the maximum Von Mises stress of CFRP and steel plate are examined. Von Mises stress on different parts of the girder is shown in Figure 5. This figure shows the maximum Von Mises stress on CFRP or steel plates appears in the pure bending region especially under the point load at mid span. The effects of application CFRP sheets or steel plates on the maximum stress of them are shown in Figure 6. In this figure, the horizontal axis is the load that is increasingly applied to the structure using the nonlinear analysis method. The vertical axis indicates the maximum Von Mises stress on CFRP sheet or steel plate. The maximum Von Mises stress is located in the pure bending region. Figure 6 shows that in the primary steps of loading there is no difference between applying CFRP or steel plate with different thicknesses, but with increasing load, greater stress appears on the steel plates than on CFRP sheets. Also, the best stress result (lowest) appears on CFRP with the greater thickness, and the load bearing capacity of CFRP sheets is also better than steel plates. 3.2

M a x im u m V o n M is e s S tre s s o n C F R P /S te e l P L

of the same grade as the steel section and steel plates, and the distance of bars is between 75 to 100 mm. CFRP sheets and steel plates were pasted onto the bottom of steel section flanges by adhesive. It is noted that CFRP and steel plates do not have the same thicknesses. Steel plate thickness is more than that of the CFRP sheet, because in actual applications, the steel plates are thicker than CFRP sheets.

CFRP,t=1.4mm CFRP,t=2mm Steel PL,t=6mm Steel PL,t=8mm

P

Figure 6. plate.

Maximum Von Mises stress on CFRP or steel

Figure 7. Shear strain on adhesive at the end of CFRP or steel plate.

Strain on adhesive

In this section the effects of the application CFRP sheets or steel plates on the Maximum shear strain on the adhesive are examined.

Figure 5.

Von Mises stress for different parts of the girder.

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Figure 7 shows the maximum shear strain on the adhesive appears at the end of CFRP sheet or steel plate, and this problem gives rise and may have caused the de-bonding of CFRP sheets or steel plates in the pasted regions. Figure 8 shows the effects of the application CFRP sheets or steel plates on the maximum shear strain of adhesive. In Figure 8, the horizontal axis represents the load, and the vertical axis is the maximum shear strain on adhesive at the end of CFRP sheet or steel plate. It shows that at the primary steps of loading, the shear strain on the adhesive for steel plates with different thicknesses are the same, but after increasingly heavier loads, the shear strain on adhesive for the thicker steel plate is greater than thinner ones. Also, shear strain on the adhesive for thicker CFRP sheets is closer to the shear strain on the adhesive for the steel plates, but for the thinner CFRP sheets, shear strain on the adhesive is so much less than the thicker CFRP sheets or steel plates, and this difference for primary and final steps of loading are important. It means

M a x im u m S h e a r S tra in o n a d h e s iv e

CFRP, t=1.4mm CFRP, t=2mm Steel PL, t=6mm

Figure 8. Maximum shear strain on adhesive at the end of CFRP or steel plate.

3.4

Deflection of the girder

One of the most important parameters for strengthening structures is deflection and how to control it. In this section the effects of application CFRP sheets or steel plates on the maximum deflection of the girder are examined. Figure 11 shows the maximum deflection of the girder in the pure bending region and at the middle of the span. Figure 12 tabulates the effects of the application of CFRP sheets and steel plates on the maximum deflection of the girder. In this figure the horizontal axis represents the applied load, and the vertical axis represents the maximum deflection of the girder at the

Maximum Von Mises Strain on Concrete

Steel PL, t=8mm

P

Figure 9.

plate thicknesses and CFRP sheet strengthening using maximum Von Mises strain are the same, but maximum Von Mises strain on concrete for thinner CFRP strengths is greater than in other cases. Also, the load bearing CFRP sheets has a better capacity than steel plates.

Punching of concrete under the point load.

CFRP,t=1.4mm CFRP,t=2mm Steel PL,t=6mm Steel PL,t=8mm

P

application of thinner CFRP sheets to overcome the debonding problem is better than thicker CFRP sheets or steel plates. Also, this figure shows the load bearing capacity of thicker CFRP is better than the steel sheets. 3.3

Figure 10.

Maximum Von Mises strain on concrete.

Figure 11.

Deflection of the girder.

Strain on concrete

In this section the effects of the application CFRP sheets or steel plates on the maximum Von Mises strain of concrete are examined. Figure 9 shows that large deformation appears under the point load region, and it punches the deck caused by point loads. The effects of application CFRP sheets or steel plates on the maximum Von Mises strain of concrete are shown in Figure 10. In this figure, the horizontal axis represents the load, and the vertical axis represents the maximum Von Mises strain on concrete at the point loading region. Figure 10 shows that the results for steel plate strengthening with varying

244

Maximim Deflection of the Girder

CFRP,t=1.4mm CFRP,t=2mm Steel PL,t=6mm Steel PL,t=8mm

P

Figure 12. of span.

Maximum deflection of the girder at the middle

mid span. This figure shows for steel plate strengthening case with different thickness of plate the maximum deflection of the girder is so close to each other. Also, the maximum deflection of the girder in CFRP strengthening case with thinner CFRP is closer to the steel plate readings, but for the thicker CFRP strengthening case the maximum deflection of the girder is less than that of the steel plates. Also, load bearing CFRP sheets is more than steel plates.

4

– If the load increases, then maximum shear strain on adhesive at the end of steel plates for thicker steel plate will be more than that for thinner steel plates. – Maximum shear strain on adhesive at the end of CFRP sheets for thicker CFRP sheet is much more than the maximum shear strain of thinner CFRP sheets, and the difference start to occur from the primary steps of loading to the final steps. It means that the de-bonding problem for thinner CFRP sheets is less than the thicker CFRP sheets. – Maximum Von Mises strain on concrete appears under the point load that occurs punching for deck. (Not clear) – Application on the different thicknesses of steel plates has no noticeable effects on Von Mises strain on concrete. – If the thickness of CFRP is less, then there is more strain on the concrete. – Maximum deflection of the girder appears in the pure bending region which located in the mid span. – The different thicknesses for steel plates have no noticeable effects on the maximum deflection of the girder. – If thicker CFRP sheets are chosen, then the maximum deflection of the girder will be less noticeable. – Load bearing CFRP sheet is better than steel plate in rehabilitation. – Load bearing CFRP sheet is much greater than steel plate with the same thickness. – Load bearing of thicker CFRP sheets is greater than thinner CFRP sheets.

CONCLUSIONS

In review on the abovementioned highlights, figures, and graphs, some conclusions can be made: – Maximum Von Mises stress on CFRP sheet or steel plate at the pure bending region. – In the primary steps of loading Von Mises stress on CFRP sheets or steel plates with different thicknesses are the same. – If the load increases, then Von Mises stress on steel plates will be greater than CFRP sheets. – The different thicknesses of steel plates have no noticeable effects on the Von Mises stress of steel plates. – If the thicker CFRP sheets are chosen, then Von Mises stress on CFRP sheets are much lesser than thinner CFRP sheets. – Maximum shear strain on the adhesive appears at the end of CFRP sheets or steel plates and may result in de-bonding, and it seems not to let CFRP sheets or steel plates bear the load up to their final capacity. It seems better to use thicker adhesive for steel plates or thinner CFRP sheets. – Maximum shear strain on adhesive at the end of steel plates for different thicknesses is the same.

245

REFERENCES Dawood, M., Sumner, E. & Rizkalla, S. 2006. Fundamental Characteristics of High Modulus CFRP Materials for Strengthening of Steel-Concrete Composite Beams. Proceedings for Structural Faults & Repair, 13–15 June 2006. Edinburgh: Scotland: CD-ROM. Edberg, W., Dennis, M. & John Gillespie, Jr. 1996. Rehabilitation of steel beams using composite materials, Material for the new millennium. Proceeding of the ASCE fourth materials engineering conference. 10–14 November 1996. Washington, D.C: 502–508. Gillespie, J.W., D.R. Mertz, K. Kasai, W.M. Edberg, J.R. Demitz & I. Hodgson 1996. Rehabilitation of steel bridge girder: Large scale testing. Proceeding of the American society for composites 11th technical conference. 4–7 November 1996. Atlanta, Georgia: 1249–1257. Jumaat, M. Z. & Narmashiri, K. 2009. Compare between 2D and 3D computer modeling for strengthening composite steel-concrete bridges with HM-CFRP and steel sheets. Proceedings of 9th international conference on steel concrete composite and hybrid structures. 8–10 July 2009. Leeds, UK: (in press.) Liu, X., P.F. Silva & A. Nanni 2001. Rehabilitation of steel bridge member with FRP composite materials. Proceeding of the international conference on composites in construction. 10–12 October 2001: 613–617.

Narmashiri, K. & Jumaat, M.Z. 2009. Effect of the HMCFRP sheets cutting shape on the stress and strain intensity of strengthened composite bridges. Proceedings of 2nd international conference on Geo-Information Technology for Natural Disaster Management and Rehabilitation. 30–31 January 2009. Bangkok, Thailand: 156–160. Schnerch, D. & Rizkalla, S. 2004. Behavior of scaled SteelConcrete composite girders and steel Monopole Towers strengthened with CFRP. Proceedings of the Innovative Materials and Technologies for Construction and Restoration conference (IMTCR). 6–9 June 2004. Leece, Italy: 42–60. Schnerch, D., Dawood, M., Sumner, E. & Rizkalla, S. 2006. Design guidelines for Strengthening of Steel-Concrete Composite Beams with High Modulus CFRP Materials. Proceedings of the 7th International Conference on Short and Medium Span Bridges. 23–25 August 2006. Montreal, Quebec, Canada: CD-ROM.

Schnerch, D., Dawood, M. & Rizkalla, S. 2007. Design guidelines for the use of HM-Strips: Strengthening of steel composite bridges with high modulus carbon fiber reinforced polymer (CFRP) strips. Technical Report, NC State University: No. IS-06-02. Sen, Rajan, Larry Liby. & Gray Mullins 2001. Strengthening steel bridge sections using CFRP laminates. Composites Part B: Engineering. 32 (4): 309–322. Tavakkolizadeh, M. & Saadatmanesh, H. 2001. Repair of cracked steel girder using CFRP sheet Creative systems in structural and construction engineering. Proceeding of the 1st international structural engineering and construction conference. 24–27 January. Honolulu, Hawaii: 461–466. Tavakkolizadeh, M. & Saadatmanesh, H. 2003. Strengthening of steel-concrete polymer sheets. Journal of structural engineering, ASCE. 129 (1): 30–40.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Repair systems for unbonded post-tensioned 1-way & 2-way slabs with CFRP P.R. Chakrabarti, U. Kim, M. Busciano & V. Dao California State University, Fullerton, CA, USA

ABSTRACT: In this research a total of six unbonded post-tensioned slab specimens were tested. Three of the specimens were done in phase-1. These were simply supported two-way slabs with two-way post-tensioning (specimen PTS-1, PTS-2 and PTS-6). Three other one-way slabs with different boundary conditions were completed in phase-2 (specimen PTS-3, PTS-4 and PTS-5). The specimens were loaded to develop extensive cracks. Each of specimens was repaired with CFRP (carbon fiber reinforcing plastics) using different patterns. The repaired specimens (PTS-1CR to PTS-6CR) were tested again to reach their ultimate loads. This research was focused on the study of cracking, deflection and ultimate strength behavior of control slab specimens and the investigation was extended to the repair of these slabs with CFRP and to find the flexural capacity, deflection and cracking pattern of the repaired slabs. Cracking appearance was postulated and the collapse loads were estimated using Yield Line Theory. Comparisons were made between control specimens and repaired specimens.

1

2

INTRODUCTION

In recent decades, a large number of structures have been built with one way and two way unbonded posttensioned slabs. Many of these slabs need to be repaired or retrofitted. Fiber Reinforced Polymer (especially CFRP) repairing and retrofitting system is very suitable for this purpose. Researches on strengthening of the pre-stressed member using FRP laminates were conduced in recent decades (Meier & Kaiser 1991; Tomas et al. 1995; Tadros et al. 1998). Extensive research on repair and retrofit of unbonded pre-stressed structures has been conducted at California State University, Fullerton by Chakrabarti, since 1998. All the tests mentioned above were focused on the study of pre-stressed beams. In this paper reports are made for the unbonded post-tensioned one-way and two-way slabs and repaired slabs (with CFRP) tested under uniform loads. The objectives of this research project were: i) to find the ultimate strength of un-repaired and repaired slabs, ii) to observe the general behavior of the slabs (during the test) before and after the application of CFRP, iii) to understand the relationship between internal reinforcement (mild steel and unbonded post tensioning) and externally bonded CFRP, iv) to observe and record crack propagation, strain, load, and deflection during testing and to comply with ACI 318-05 and ACI 440.2(R)-02 standards.

MATERIALS

– Concrete: Quality concrete of approximately 5500 psi compressive strength was used. Concrete mixes were prepared according to ASTM C 94-65. – Post-Tensioning Steel: 16-1/4 Ø 7-wire strands were used in each direction of the slabs (ASTM A-416 with nominal fpu = 270 ksi). – Plastic tubing: The individual pre-stressing strands were inserted through 9/64 diameter plastic tubes. This process eliminated any bonding between the wires and the concrete. – Mild Steel: Welded Wire Mesh with nominal yield strength 60 ksi (ASTM A-185) was used at the top of each of the two-way-slabs. In one-way slabs, #3 bars were used. – FRP and Adhesive: Wabo® MBrace Fiber Reinforcement Systems (CF130 and CF160 High Tensile Carbon) was supplied by Watson Bowman Acme Corp.

247

3

SPECIMEN PREPARATION AND STRENGTHENING

The two-way, simply-supported, unbonded posttensioned slabs (with or without CFRP) were tested with uniformly distributed external loading. The 9 − 0½ × 9 − 0½ two way slabs (PTS-1, PTS-2

Table 1.

Summary of reinforcement parameters.

Specimens

Pre-stressing Steel (Aps), in2 /in

PTS-1 PTS-2 PTS-3

16-1/4 Stands ea. way 16-1/4 Stands ea. way (0.00554) 16-1/4 Stands (0.00554)

PTS-4

16-1/4 Stands (0.00554)

PTS-5

16-1/4 Stands (0.00554)

PTS-6

16-1/4 Stands ea. way (0.00554)

Mild Steel (in Tensile zone) (As), in2 /in

Mild Steel (in Comp zone) (As), in2 /in

fc (psi)

0.0100 (4 × 4-4 × 4) 0.0024 (6 × 6-10 × 10) 0.0265 25-#3 at top and 9-#3 at bottom at each end ASTM A615, Gr. 60 0.0276 26-#3 at top and 9-#3 at bottom at each end ASTM A615, Gr. 60 0.0287 27-#3 at top and 9-#3 at bottom at each end ASTM A615, Gr. 60 0.0024 (6 × 6-10 × 10)

0.0024 (6 × 6-10 × 10) 0.0024 (6 × 6-10 × 10) 0.0024 (6 × 6-10 × 10)

5931 5963 5726

0.0024 (6 × 6-10 × 10)

5959

0.0024 (6 × 6-10 × 10)

5362

0.0024 (6 × 6-10 × 10)

5223

Table 2.

Figure 1.

Values for Wabo® -M-Brace Fiber.

Wabo® MBrace fiber

Ultimate strength (ksi)

Design strength (ksi)

Tensile modulus (ksi)

CF130 High Tensile Carbon

620 (slab 1 and 2)

550

33,000

CF530 High Modulus Carbon

580

510

54,000

CF160 High Modulus Carbon

620 (slab 3 and 4)

550

33,000

The information above was obtained from Wabo® MBrace Composite Strengthening System Design Guide. CF160 (7.14 k/in) has twice the thickness of CF130 (3.57 k/in). Properties of resins not shown—available upon request.

Typical placement of strands, Wire Mesh.

and PTS-6) were internally reinforced using mild steel and un-bonded post-tensioning strands each-way Table 1 (Figure 1 & Figure 13). The slabs repaired with CFRP are indicated as ‘‘CR’’. The two-way slabs were repaired using different strengthening schemes. Diagonal schemes are shown in Figure 4 and Figure 15, which was used for PTS-2CR (perpendicular to cracks). Orthogonal schemes were used in PTS-1CR & PTS-6 (parallel to slab edges). Slabs PTS-3 was an one-way slab with both ends fixed, PTS-4 was an one-way slab with one end fixed and PTS-5 was a simply supported one way slab. Pre-stressed strands and reinforcements for these slabs are shown in Table 1. The one-way slabs were strengthened (PTS-3CR, PTS4CR and PTS-5CR) with straight CFRP sheets at the top tension zone and at the support at bottom tension zone (1/4th clear span) as shown (Figures 5 & 6 and Figure 16). Details of CFRP materials are given in Table 2.

Figure 2.

248

Crack on PTS-1 (PTS2 & PTS6 similar).

4

TEST SETUP AND LOADING CRITERIA

Loading was gradually increased at approximately 1 psi step. The ultimate load was determined using

one of the following three criteria: 1) Excessive cracks were visually observed, 2) Deflection reached close to L/120, and 3) Post-Tensioning force reached nearly 75% to 80% of the ultimate load. During the testing, load, deflection, strain (in mild steel), and change in post-tensioning forces were recorded. Same criteria were used for the repaired slabs.

5

TEST RESULTS AND DISCUSSION

Unlike bonded pre-stressed or plain reinforced members, the pre-stressing strands in the unbonded posttensioned members do not reach their ultimate strength. As the slabs started to crack heavily the stress in the mild steel started to increase and became close to yield stresses. Ultimate loads of the test specimens, strengthened with CFRP composites, were always higher than the ultimate loads of control specimens. Test results and comparisons are given in the following tables. Figure 3.

Orthogonal CFRP placement (PTS-1CR).

Figure 4.

Diagonal CFRP placement (PTS-2CR).

Table 3.

Figure 5.

Crack line in control specimen PTS-3.

First-crack loads-ultimate loads and comparisons.

Specimens

First-crack loads (Measured) q1 (psi)

Ultimate loads (measured) qu (psi)

FCL/UL qu/ql

PTS-1 PTS-2 PTS-3 PTS-4 PTS-5 PTS-6 PTS-1CR PTS-2CR PTS-3CR PTS-4CR PTS-5CR PTS-6CR

6.35 4.8 5.65 4 3.5 3.5 5.5 5.2 6.7 4.9 4.4 6.2

6.5 5.2 6.5 5.4 6.4 5.5 9.2 9.8 10.9 7.9 7.48* 8.5

1.02 1.08 1.15 1.35 1.83 1.57 1.67 1.88 1.36 1.61 1.55 1.37

UL(Repaired)/UL(Constrol) qu/qu (control) – – – – – –

9.2/6.5 = 1.42 9.8/5.2 = 1.88 10.9/6.5 = 1.677 7.9/5.4 = 1.463 7.48/6.4 = 1.17 8.5/5.5 = 1.545

* The testing was prematurely stopped for safety reason hence 10% of last observed value is added.

249

10 9 8

Applied Load (psi)

7 6 5

PTS-1

4

PTS-1CR

3 2 1 0 0.2

0

0.4

1

0.8

0.6

Deflection at Mid Span (in)

Figure 7.

Defln. vs. Ext. load at PTS-1 & PTS-1CR.

11 10 9 8

Applied Load (psi)

The failure loads calculated for the specimens are summarized in Table 3. Good agreement was reached between the estimated and measured values of the loads of control specimens calculated by Yield Line Theory. The measured collapse loads of all repaired test specimens were lower than the loads estimated based on yield line theory. The test result confirmed that the analysis of repaired slabs by yield line theory was not conservative (values obtained upon request). The critical yield line pattern developed in the control specimens could not propagated because of the presence of the CFRP across the crack lines. It seems, at the time the test was stopped, the slabs were in a semi-elastic condition. Clear rigid plate movement along crack line did not happen yet. Most likely crushing would have happened at the bottom surface along the Yield Lines if the load was further increased. In all cases ultimate loads of the repaired specimens were larger than the corresponding ultimate loads of the control specimens. Effective placement of CFRP (as in PTS-2CR) can increase load carrying capacity.

7 6

PTS-2

5

PTS-2CR 4 3 2 1 0

6

DEFLECTION AND CRACKING OF CONTROL SPECIMENS AND STRENGTHENED SLABS

0.1

0

0.2

0.3

0.4

0.6

0.5

0.7

Deflection at Mid Span (in)

Figure 8.

In general, the control slabs behaved linearly during the first stages of load. Once the slabs cracked and load-deflection curve showed bi-linear behavior, Figures 6 to 12. The second stage of linear behavior of cracked elastic slab continued until the yielding of bonded non-pre-stressed mild steel began. As the flexural tensile cracks increased in number and width, the load deflection curves started to show tri-linearity and slope reduction (not seen in the repaired slabs).

Defln. Vs. Ext. load at PTS-2 & PTS-2CR.

12

A ppliedLoad (ps i)

10 8 6

PTS-3 PTS-3CR

4 2 0 0

0.2

0.4

0.6

1

0.8

1.2

1.4

1.6

Deflection (in.)

Figure 9.

Deflection vs. Ext. load PTS-3 & PTS-3CR.

9 8 7

Applied Load (psi)

6 5

PTS-4

4

PTS-4CR

3 2 1 0 0

0.5

1

1.5

2

2.5

Deflection (in.)

Figure 6.

CFRP placement on PTS-3CR.

Figure 10.

250

Deflection vs. Ext. load at PTS-4 & PTS-4CR.

8 7 6 Appiled Load (psi)

5 4

PTS-5 PTS-5CR

3 2 1 0 0

0.2

0.4

0.6

0.8

1

1.2

1.4

1.6

1.8

2

2.2

Deflection (in.)

Figure 11.

Deflection vs. Ext. load PTS-5 & PTS-5CR.

Figure 15.

Diagonal Layout of CFRP Sheets on PTS-2CR.

Figure 16.

PTS-3 at ultimate load (Cracking).

10 9 8 Applied Load (psi)

7 6 5

PTS-6

4

PTS-6CR

3 2 1 0 0

0.2

0.4

0.6

0.8

1

1.2

1.4

Deflection (in.)

Figure 12.

Deflection vs. Ext. load PTS-6 & PTS-6CR.

Table 4. Comparison of measured deflection at center of slabs at 5.2 psi.

Figure 13.

Form- post-strands- reinforcements.

Specimens

Ext. loads q (psi)

Measured deflection 

 repaired/  control

PTS-1 PTS-1CR PTS-2 PTS-2CR PTS-1 PTS-1CR PTS-3 PTS-3CR PTS-4 PTS-4CR PTS-5* PTS-5CR PTS-6 PTS-6CR

5.2 5.2 5.2 5.2 6.5 6.5 5.2 5.2 5.2 5.2 5.2 5.2 5.2 5.2

0.451 0.404 0.312 0.201 0.753 0.563 0.850 0.600 1.520 1.054 2.000 0.688 0.450 0.385

1 0.9 1 0.64 1 0.75 1 0.71 1 0.69 1 0.34 1 0.86

* Excessive cracking.

Figure 14.

In general the repaired slabs were stiffer and they had less deflection. As expected, diagonal cracking modes were observed in the two-way control slabs (PTS-1, PTS-2 and PTS-6). In general, cracks were not visible on the top surface of the repaired slabs during testing (covered by CFRP). In some of the exposed areas, the old cracks opened up. The comparison of deflection

PTS-1 (similar PTS-2) at Ultimate Load.

251

values of control specimens and repaired specimens are shown in Table 4.

7

CONCLUSIONS

1. The post-tensioned concrete slabs, repaired with CFRP fabric and bonded to the tension surfaces, gained considerable strength. Flexural capacity of the slabs strengthened with CFRP composite materials increased by 42%, 88% and 55% for the two-way slabs PTS-1CR, PTS-2CR and PTS6CR respectively and 68%, 46% and 6% for the one-way slabs PTS-3CR, PTS-4CR and PTS-5CR respectively. 2. The measured deflection ratios between the control post-tensioned slabs and repaired specimens, at a specific load, indicate considerable increase in stiffness for the repaired slabs. These ratios further reduced as the loads were more increased for each of the slabs. 3. Slabs repaired with properly designed CFRP schemes showed larger load carrying capacities. The placement scheme was most effective when the CFRP fibers were perpendicular to the crack lines. 4. Strengthening of the slabs with the CFRP composites reduced the crack width and frequency at the high moment region at all load level. The slabs siffer even at a much higher load. Yield line theory used to calculate the collapse load of the control specimens provided good correlations. Estimated ultimate loads obtained by using Yield Line Theory, for the repaired slabs, were much larger than that were obtained by measurement. Additional testing is necessary. 5. Test results indicated better serviceability conditions for the repaired slabs. Flexural capacity of the slabs significantly increased due to the presence of the CFRP sheets. No de-bonding of the CFRP sheets was observed at the ultimate loading stage. 6. Large increases in flexural strength of one-way post-tensioned concrete slabs were obtained by using CFRP fabric bonded to the tension surfaces. The results from this project indicate that different end supports of one-way slabs caused large variations in performance. 7. As anticipated, test results indicate that both serviceability and flexural strength improved in the slabs repaired with CFRP.

8

may be taken to total collapse to realize full capacity. Additional safety precautions must be taken in the tests of an unbonded post-tensioned structure (strands remain unbonded).

RECOMMENDATION

Extensive testing of unbonded post-tensioned slabs with different variables is necessary. Repaired slabs

REFERENCES ACI committee 318, ‘‘Building Code Requirements for Structural Concrete and Commentary (ACI 318-02/318R-02),’’ ACI, Detroit, 1992 ACI Committee 440, ‘‘Guide for the Design and Construction of Externally Bonded FRP Systems for Strengthening Concrete Structures (ACI 440.2R02)’’, American Concrete Institute, Detroit 2002. Chakrabarti, P.R., ‘‘Retrofitting and Repairing of Heavily Cracked Unbonded Post-Tensioned Structural Systems with Composite Materials’’, Edward Nawy Symposium, American Concrete Institute Spring Convention, NY, NY, April 17–20, 2005 (ACI special publication, ACI SP-225). Chakrabarti, P.R. and Thornton, K., ‘‘Application of Unbonded Post Tensioning in Infrastructure—(a review on research and Development)’’, Proc. CONMAT-2003, International Workshop/Conference on Construction Management, and Materials, Jan 9–11, 2003, IITKharagpur, India pp. 29–44. Chakrabarti, P.R., ‘‘Ultimate Stress for Unbonded PostTensioning Tendons in Partially Pre-Stressed Beam,’’ ACI Structural Journal, v. 92, No. 6, Nov–Dec 1995, pp. 689–697 Chakrabarti, P.R., ‘‘Repairing and Retrofitting of PostTensioned Beams’’, ACI Concrete International, Feb. 2005, pp. 1–4. Chakrabarti, P.R., ‘‘Ultimate Strength of Corner Panels of Post-Tensioned Flat Plates,’’ ACI Structural Journal, V. 84, No. 1, Jan–Feb 1987, pp. 54–87. Chakrabarti, P.R., Miller, D. and Bandyopadhayay, S., ‘‘Application of Composites in Infrastructure—Part I, II and Part III—(a brief report on materials, and construction)’’ Proc. ICCI-2002, the Third International Conference on Composites in Infrastructure, June 10–12, 2002, San Francisco, pp. 1–10. Meier, U. and Kaiser, H., ‘‘Reprinted from Advanced Composite Materials in Civil Engineering Structures Proceedings,’’ MT Div/ASCE/Las Vegas, Jan. 31, 1991, pp. 224–229. Nawy, E.G. and Chakrabarti, P.R., ‘‘Deflection of PreStressed Concrete Flat Plate,’’ Pre-Stressed Concrete Institute Journal, March–April, 1976, pp. 86–102. Structural Preservation Systems, Inc. ‘‘Wabo® MBrace Composite Strengthening System Engineering Design Guidelines,’’ (05/02). Seim, W., Vasquez, A., Karbhari, V. and Seible, F., ‘‘Poststrengthening of Concrete Slabs: Full-Scale Testing and Design Recommendations’’ Univ. of California, San Diego. Tadros, G., Rizkalla, S.H., Michaluk, C.R. and Benmokrane, B., ‘‘Flexural Behavior of One-Way Concrete Slabs Reinforced by Fiber Reinforced Plastics Reinforcements,’’ ACI Structural Journal, V. 95, No. 3, May–June 1998, pp. 353–365. Timoshenko, and Krieger, ‘‘Theory of Plates and Shells,’’ McGraw-Hill, Inc.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Strengthening a concrete slab bridge using CFRP composites S.H. Petro, J.T. Peaslee & T.G. Leech Gannett Fleming, Inc, Morgantown, WV, USA

ABSTRACT: This paper describes the feasibility and efficiency of using CFRP composites as a strengthening system to upgrade the load carrying capacity of a three-span continuous (17.17 ft/17.33 ft/17.25 ft) [5.23 m/ 5.28 m/5.26 m] concrete slab bridge constructed (ca. 1946) in the City of Phoenix spanning the canal system and carrying an average daily traffic of 30,000 vehicles. Given the relatively short canal dry-up period (three weeks), a fast-paced approach to strengthen and improve the load carrying capacity of the existing structure was required. Externally applied CFRP wraps were used for the soffit in the positive moment regions; and carbon fiber rebars embedded in the concrete were used over the pier locations in the negative moment regions. Based on load test results, the externally applied CFRP composite materials significantly enhanced the structural performance of the 19th avenue bridge and improved its load rating at a fraction of the cost of total replacement and at a fraction of the time. 1

INTRODUCTION

1.1 General The retrofit and rehabilitation of bridges and structures by externally bonded fiber reinforced polymers (FRP) composite materials has become popular due to their superior characteristics such as high tensile strength and modulus combined with light weight, high resistance to corrosion from deicing salts and other corrosive agents, ease of fabrication, ease of shipping and jobsite handling, and ease and speed of installation. The majority of FRP applications for bridge structures use longitudinally oriented carbon fibers (CFRP). CFRP composites have been proven to be strong, non-corrosive, and a long-term durable composite system suitable for bridge strengthening to increase the load carrying capacity. This paper summarizes the results of a completed bridge strengthening project using CFRP composites with live load testing to validate the effectiveness of the CFRP system. 1.2 Bridge description The 19th Avenue Bridge is a three-span, continuous concrete slab bridge originally constructed (ca. 1946) with enhancements and widening implemented in the 1960s. The 19th Avenue bridge is a typical type slab bridge found in the City of Phoenix spanning the canal system and carries an estimated average daily traffic of 30,000 vehicles (Figure 1). Designed to accommodate traffic loads smaller than currently permitted, the structure had reached the end of its service life and demanded an immediate upgrade. An evaluation of the 19th Avenue Bridge prior to rehabilitation revealed that the structure was

253

Figure 1.

19th avenue bridge with dry canal.

deficient by Federal Highway Administration (FHWA) standards. In addition, an initial analysis, using current American Association of State Highway and Transportation Officials (AASHTO) code, indicated that the bridge required upgrading or strengthening for flexure to meet the current Arizona legal load capacities. A recent load rating performed by the City of Phoenix (COP) resulted in an Inventory Rating (IR) of 15.6 tons and an Operating Rating (OR) of 26 tons for an HS-20 (36 ton) vehicle. Given the relatively short canal dry-up period (three weeks), a fast-paced approach to strengthening and improving the load carrying capacity of the existing structure was required.

2

EXISTING CONDITIONS

The material properties of the existing reinforcing bars and concrete were determined by non-destructive evaluation methods and mild destructive coring. Rebar spacing and depth of cover were determined using

ground penetrating radar (GPR); while coring was performed to determine rebar size, concrete slab depth, and concrete compressive strength (f c). Considering the age of the bridge, the yield strength of the reinforcing bars was assumed to be 33 ksi (228 MPa), as no samples were available for testing; while concrete compressive strength (f c) was determined to be 3500 psi (24 MPa) from concrete core testing. To verify the results of the GPR for the existing reinforcement, mild destructive testing (i.e., exposing rebar by removing cover) was performed by the contractor. The results indicated that for positive moment regions (bottom reinforcement) at midspan, the reinforcement consisted of No. 8 bars spaced at 6 (152 mm) with 1.4 (36 mm) cover; while for negative moment regions (top reinforcement) over piers, the reinforcement consisted of No. 8 (25.4 mm) bars spaced at 17.75 (451 mm) with 3.5 (89 mm) cover indicating insufficient negative moment reinforcement and a departure from the design intent of the construction drawings.

3

Modulus of Elasticity = 9.93×106 psi (68,000 MPa) Tensile Strength = 104,000 psi (716 MPa) Thickness = 0.04 inch (1 mm) Fracture Strain = 0.98%

Aslan 200 Bars – Tensile Strength = 300,000 psi (2068 MPa) – Modulus of Elasticity = 18 × 106 psi (124 Gpa) – Bar Diameter = 0.47 inch (12 mm) 4

Application of CFRP fabric strips at soffit.

Figure 3. deck.

CFRP bars surface mounted in grooves at top of

CFRP SYSTEM

For the 19th avenue bridge rehabilitation, a high strength, unidirectional carbon fiber (SiKaWrap Hex 103C) was used for the soffit; and carbon fiber rebar (Aslan 200 Bar by Hughes Brothers) embedded in the concrete was used over the pier locations. For the positive moment regions, two (2) layers of SiKaWrap Hex 103C 12 × 15 (305 mm × 4.5 m) long wide spaced at 36 (0.9 m) on center were bonded to the soffit using a high strength adhesive (Sikadur Hex 300) as recommended by the manufacturer. For negative moment regions over the piers, 63 ½-in (12 mm) diameter Aslan 200 CFRP bars were near surface mounted in grooves using high strength adhesive (Sikadur 32) as recommended by the manufacturer. The following are the CFRP mechanical properties: SiKaWrap Hex 103C – – – –

Figure 2.

was restored with non-shrink mortar (Sikadur 30 or Sikatop 123) and after surface preparation that included air blasting; CFRP strips were bonded as shown in Figure 2. The required number of near surface mounted CFRP bars was determined to be 63 at 12 (305 mm) groove spacing. The 13 feet (4 m) long bars were embedded in 1 (25.4 mm) deep by 3/4 (19 mm) wide grooves cut into the bridge deck over the piers as shown in Figure 3. Prior to installation of the CFRP system, it was required to minimize deck vibrations due to traffic to allow the epoxy to cure. COP personnel decided that closing the bridge for any length of time would not be feasible. Instead, efforts were made to limit the effects of traffic loads during strengthening operations. This was achieved by diverting traffic away from the lane directly above the CFRP installation each day.

CFRP INSTALLATION 5

The design of the externally bonded CFRP fabrics called for 12 × 15 long wide strips spaced at 36 on center for the positive moment regions. The concrete surface that contained irregularities and/or cracks

DESIGN PHILOSOPHY

The design of the CFRP systems was carried out according to ACI 440.2R-02 Guide for the Design and

254

Construction of Externally Bonded FRP Systems for Strengthening Concrete Structures, (ACI 440.2R-02, 2002). The flexural capacity of strengthened sections with CFRP was computed based on force equilibrium, strain compatibility, and the governing mode of failure (GangaRao and Vijay 1998). A trial and error procedure was followed to satisfy these conditions. Since the use of externally bonded CFRP reinforcement tends to reduce the ductility of the original section, a strength reduction factor φ was used to convert nominal values to design capacities. Adequate ductility is achieved if the strain in the steel reinforcement at the point of concrete crushing or failure of the CFRP including delamination or bonding is at least 0.005 (ACI 440.2R-02, 2002). Furthermore, material properties of the CFRP reinforcement reported by the manufacturers, such as the ultimate tensile strength, do not consider long term exposure to environmental conditions. Therefore, an additional strength reduction factor, f , was applied to the flexural strength contribution of the CFRP reinforcement. A factor of f = 0.85 was used. The nominal flexural strength of a section reinforced with CFRP reinforcement was computed from Equation 9–13 (ACI 440.2R-02) as follows:     β1 c β1 c + ϕf Af ffe h − Mn = As fs d − 2 2

(1)

In addition, because FRP materials are linearly elastic until failure, the level of strain in the FRP determines the level of stress developed in the FRP (ffe ). The design procedure to prevent debonding failure according to ACI 440.2R-02 limits the CFRP strain levels by introducing a bond dependent coefficient to compute the nominal capacity of the section. Similarly, to limit deflections (i.e., serviceability) under service loads, the existing steel reinforcement is prevented from yielding by limiting the steel stress (fs ) to 80% of the yield strength.

6

LOAD RATING

Bridge load rating provides a basis for determining the safe load carrying capacity of a bridge. As mentioned previously, a load rating analysis was performed by COP for the 19th Avenue over grand canal bridge and determined the need for strengthening. The team’s goal was to raise the bridge’s inventory rating to HS-20 using the CFRP strengthening. The governing ratings were developed following the Load Factor Method. The following rating equation was used:  RF =

MCAP − MDL MLL+I

 (2)

255

Table 1.

Load ratings. Before retrofit

After retrofit

Vehicle

RF

Tons

RF

Tons

HS-20—IR HS-20—OR

0.43 0.72

15.6 26

1.0 1.67

36 60

where RF = rating factor; MCAP = moment capacity of strengthened section with CFRP; MDL = dead load moment; and MLL+I = live load plus impact. For the 19th avenue over grand canal bridge, the load rating was calculated for an HS-20 vehicle (AASHTO, 2007). Two load ratings levels were calculated, the maximum load level called the Operating Rating (OR), and a lower load level called the Inventory Rating (IR) which is the load level the bridge can support vehicles on a daily basis without damaging the bridge. The computed load ratings before and after retrofit are as follows: The maximum force responses associated with these rating factors are flexural due to negative moments occurring over the piers. After retrofit the rating factors and associated load limits were increased by 130%. The governing inventory and operating rating factors became 1.0 and 1.67, respectively. Since the load rating factors (RF) are greater or equal to 1.0, the bridge does not need to be load posted.

7

INSTRUMENTATION

To evaluate the performance of the CFRP strengthening, a continuous monitoring program was installed to measure strains in real-time making the 19th Avenue Bridge a smart structure. The bridge was instrumented and tested before retrofit and shortly after installation of the strengthening system. The initial tests established a number of critical benchmark responses of the bridge; while the other tests provided information regarding the participation of the CFRP system. The instrumentation program was implemented by a third party which included a continuous monitoring system using propriety displacement gages (i.e., strain gages or sensors) placed on both the CFRP and concrete soffit and gages embedded over the CFRP rebar and in the concrete regions over the piers. Thermal sensors were also installed on the deck to measure thermal fluctuations. The monitoring system is powered with a solar array and data is uploaded wirelessly to a Web site which can be easily accessed by the City of Phoenix for continuous monitoring (Figure 4). The 19th Avenue bridge was instrumented to measure concrete surface strains at 8 locations (S1, S3, S5, S8, S10, S12, S13, and S15); and strains in the

CFRP system at 8 locations (S2, S4, S6, S7, S9, S11, S14, and S16). The strain gage sensors on the CFRP system were installed close to the locations where the maximum concrete surface strains in each span had been measured during testing before retrofit. Sensor locations were distributed to measure maximum responses in positive moment regions and negative moment regions for all spans. Two COP loaded water trucks were used for the static load testing. These trucks had a three-axle configuration with a gross vehicle weight of 60 kips (267 kN) distributed between the front axle (20 kips [89 kN]) and rear two axles (40 kips [178 kN]). One of the 3-axle trucks was 18 (5.48 m) long while the second truck was 16.5 (5 m) long. For the First Load Position, a single truck was positioned with the center of the rear axles at midspan of each of the three spans of the bridge. This position was chosen to produce the maximum positive moment in each of the three spans (Figure 5). For the Second Load Position, the two rear axles from both trucks were placed on either side of Pier 1 or Pier 2 in the same lane to produce the maximum negative moment over Pier 1 and/or Pier 2.

The load tests included fifteen (15) truck positions and were conducted in the center turning lane and the adjacent fast north bound lane. The full series of the static load tests at each position were repeated two times (i.e., two measurements were taken) both before and after the installation of the CFRP system.

8

RESULTS OF LOAD TESTS

Figure 4.

19th Avenue bridge—solar panel and antenna.

The impact of the CFRP retrofitting system on the performance of the bridge is illustrated by comparing strains on the concrete before and after installation of the CFRP system. In Figure 6, the maximum measured strains in the concrete (S1, S3, S5, S7, and S10) are compared to the corresponding concrete strains before and after CFRP installation. The maximum strains are from different tests, but the strains for the concrete are from the same Truck Load Position 1 and 2, intended to produce maximum positive moment in spans 1, 2, and 3; and maximum negative moments over Pier 1 and Pier 2. After installation of the CFRP wraps on the soffit, the concrete strain at all locations dropped (Fig. 6). For example, the strain at location S5 (sensor 5) where the largest strain in the concrete soffit was measured before the CFRP wraps were installed, was reduced significantly. Similar results were achieved from all other locations in both lanes from load tests at Truck Position 1. This trend points to the participation of the CFRP system toward resisting the applied live loads. Similarly, the measured concrete surface strains from the embedded sensors over the piers indicated a drop in tensile strain after CFRP installation due to truck load Position 2. For example, the strain at location S7 (sensor 7) where the largest strain in the concrete over the Pier 1 was measured before the CFRP bars were installed, was reduced significantly. This trend also points to the participation of the embedded CFRP bars toward resisting the applied live loads.

Figure 5.

Truck position to produce max positive moment.

Figure 6.

256

Maximum strains in concrete slab.

The measured concrete surface strain values are very small indicating that the trucks used for testing did not produce large demands in the bridge. The maximum concrete strain was about one third of that causing cracking, and the CFRP systems were strained to only 0.05% of the useable strain at which the capacity of the bridge with CFRP was computed. The small strains do not imply that the CFRP system cannot increase the overall strength of the bridge and its rating factors, it simply implies that the loads for the truck load tests are less than those used for rating (i.e., HS-20); and the capacity of the members strengthened with CFRP was computed based on a strength design approach in which significantly larger strains were used in the calculations. 9

CONCLUSIONS

The retrofit of an existing concrete slab bridge using externally applied CFRP composite materials and field testing to investigate their effectiveness were presented. Based on the results, the following conclusions are drawn: – Externally applied CFRP composite materials significantly enhanced the structural performance of the 19th Avenue bridge and improved its load rating at a fraction of the cost of total replacement and at a fraction of the time. – This technology has demonstrated that a bridge structure can be rehabilitated in-situ and put into service in a matter of days, as opposed to weeks or months. The FRP strengthening system using CFRP composite materials allowed the upgrading of the load carrying capacity and improving the load rating of this structure, thereby eliminating the need to load post the structure without major modifications to the superstructure. – This project provided a unique opportunity to examine short- and long-term performance measures of the CFRP system. Specific static truck load tests were conducted to obtain benchmark responses of

257

the original structure without CFRP reinforcement and shortly after CFRP installation. Results indicate that concrete strain in the slab dropped as the CFRP system contributed to the load carrying capacity of the bridge and most importantly, debonding did not occur. – The accelerated construction schedule of the 19th Avenue Bridge retrofit was accomplished within the time constraints of the three-week canal dry-up period and on budget. A construction crew ranging from 8 to 20 workers participated on the project. Utilizing CFRP to strengthen the 19th Avenue Bridge, the project team was able to deliver an innovative and cost-effective solution providing the benefits of no interruption of utility lines, limited environmental pollution, limited construction time, and minimal traffic disruptions, and most importantly, the construction cost represented a fraction of the cost of total replacement. ACKNOWLEDGEMENTS The authors would like to acknowledge the contribution of Gannett Fleming, Inc staff from the Morgantown, WV; Phoenix, AZ; and Pittsburgh, PA offices. Assistance from the City of Phoenix is also gratefully acknowledged. REFERENCES AASHTO. 2007. Standard Specifications for Highway Bridges, 17th Edition, American Association of State Highway and Transportation Officials, Washington, D.C. ACI COMMITTEE 440.2R-02. 2002. Guide for the Design and Construction of Externally Bonded FRP systems for Strengthening Concrete structures, American Concrete Institute, Detroit, MI. GangaRao, H.V.S. and Vijay, P.V. 1998. Bending Behavior of Concrete Beams Wrapped With Carbon Fabric, Journal of Structural Engineering, V. 124, No. 1 pp. 3–10.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Strengthening effect of CCFP for RC member under negative bending I. Yoshitake & S. Hamada Yamaguchi University, Yamaguchi, Japan

K. Yumikura Inai conex co. ltd. Osaka, Japan

Y. Mimura Kure National College of Technology, Hiroshima, Japan

ABSTRACT: ` Consolidated Carbon Fiber Plate (CCFP) has significant tensile strength and higher ductility, so it has often been employed for strengthening of RC member in recent years. The present study employed CCFP for strengthening cantilever RC slabs which subjected to negative bending. In order to obtain a rational anchorage method for CCFP, bending tests were conducted by using beam specimen with various anchorage methods. Based on these experiments, we have devised a simplified method that embeds CCFP in the surface of RC member. Additionally, fatigue properties of RC member with such anchorage were investigated by the wheel running test and negative bending beam test.

1 1.1

method using carbon fiber plates is proposed in the present paper.

INTRODUCTION Background and purpose

RC slabs of road bridges are directly loaded by wheel load, so RC slabs are gradually deteriorated by such fatigue loading. In other words, life time of RC slabs is mainly determined by such fatigue deteriorations. Even such deteriorated RC slabs must be used as long as possible so that budget for infrastructures is limited in spite of increase of old bridges. Road Bridge is one of the most important infrastructures for transportation and distribution. Strengthening of such RC slabs of bridge will be more and more important for sustainable civil infrastructure. Various strengthening methods for deteriorated RC slabs have been developed, e.g. steel plate, fiber sheet and so on. These have been often employed for strengthening of bottom of RC slabs between main girders; those are subjected to normal bending moment. On the other hand, cantilever of RC slabs are often subjected to negative bending due to wheel loads and wind load. Thus, top of slabs at main girder must be strengthened in these cases. In most cases, such strengthening works need prohibition of traffic on bridges. Bridge engineers should finish it as fast as possible in order to provide usual transportation. In previous studies, there are little discussions about strengthening of cantilever RC slabs subjected to negative bending moment. The purpose of the present study is to develop a rational strengthening method for such cantilever slabs. Especially, a simple strengthening

259

1.2

Consolidated carbon fiber plate

There are many strengthening methods using various materials for existing concrete structures. Especially, carbon fiber sheet has significant strength and high utility, so it is employed for strengthening of various

Figure 1. Table 1.

Consolidated Carbon Fiber Plate (CCFP). Properties of CCFP and CF sheet.

Size Tens. strength Young’s Modulus Thermal coef. Strain capacity CF content

CCFP

CF Sheet

50 × 1.2 mm 2664 N/mm2 188 kN/mm2 0.7 × 10−6 /C◦ 1.42% 60%

t = 0.111 mm 3400 N/mm2 245 kN/mm2 0.3 × 10−8 /C◦ 1.39% 100%

simply attached by epoxy adhesive. FP1-H, FP1-HA, FP1-Pl-A, FP1-S and FP1-L are RC beams employing reinforcements of edge of CCFP respectively. In these names of specimen, ‘‘H’’ after first hyphen means hooks. Similarly, ‘‘A’’ and ‘‘Pl’’ means anchor-bolt and steel-plate which has thickness of 6 mm respectively. ‘‘S" means external plates made of CCFP, which has shape of triangle, in order to extend bond area on the concrete surface. And ‘‘L’’ means L-shaped anchor made of CCFP, which bonded on the edge of CCFP. FP0’ is normal RC beam similar to FP0; it is employed for comparison with FP1-G. FP1-G is RC beam which include CCFP in ditch of full length. The ditch on concrete surface has 2 functions that are to bond CCFP into tougher layer of concrete and to embed CCFP for flatness of upper surface of RC slab.

structures. Carbon fiber sheet (CF sheet), however, needs complicated works of epoxy-resin (typical bond material). It has a possibility to cause longer traffic control, when CF sheet is employed for cantilever slab which is main target of this study. Consolidated carbon fiber plate (CCFP) is a prefabricating material which is made with CF sheet and epoxy-resin. The sample of CCFP is presented in Figure 1 and the properties of CCFP and typical CF sheet are given in Table 1. The mechanical properties of CCFP are almost equal to CF sheet and CCFP may contribute more rapid works for strengthening. 2 2.1

ANCHORAGE METHOD FOR CCFP CCFP anchorage on cantilever RC slabs

Anchorage method of CCFP is very important for strengthening; we cannot obtain appropriate effect by CCFP if the anchorage effect is insufficient. Additionally, the flat surface of slabs must be maintained for reconstruction of pavement. So, we cannot employ some external special attachment for strengthening of cantilever slabs in many cases. In order to investigate appropriate anchorages of CCFP for cantilever RC slabs, the present study set 2 objects shown in below. I. II.

2.3

The employed specimens were typical RC beam with steel ratio of 0.4%, and all of these specimens were 200 × 200 × 1900 mm size. RC beam model is illustrated in Figure 3 and materials employed in such specimen are given in Table 2. These RC beams were designed as bending failure type even if RC beams include CCFP or any reinforcement. Thus, the steel ratio was designed as smaller than normal ones of RC slab for bridges.

Simple and economic anchorage work To keep flatness of upper surface of RC slabs

We conducted bending test of RC beam with various anchorage of CCFP for obtaining above objects. 2.2

2.4

Various anchorage methods employed in the present study are given in Figure 2. FP0 is the RC beam without CCFP and FP1 is the RC beam with CCFP C.L.

Bottom of RC beam

Layer

CCFP Epoxy Adhesive

CCFP Epoxy Adhesive Concrete

FP0 (FP0 )

C.L.

Steel Plate CCFP Ste Plate Steel CCFP CF Epoxy ox Adhesive

Layer

CCFP

CCFP Plate

C.L.

CCFP

CCFP CCFP Epoxy Adhesive Concrete

Layer

FP1-S

Layer CCFP CF Epoxy ox Adhesive Concrete

C.L.

Bottom of RC beam

CCFP

CCFP CCFP Epoxy Adhesive Concrete

RC beam with various anchorage of CCFP.

260

C.L.

CCFP

Ditch

Layer yer

FP1-L

C.L.

CCFP

FP1-H-A

Bottom of RC beam

CCFP P Plate ate

Bottom of RC beam

A.B.

Hook CCFP Epoxy Adhesive

FP1-H

Bottom of RC beam

Concrete

FP1-PL-A

C.L.

Concrete

Concrete

FP1

Bottom of RC beam

Bottom of RC beam

Hook

Layer

Layer

Figure 2.

C.L.

CCFP

CCFP

Experimental procedure

The experimental conditions of the present study are illustrated in Figure 3. These bending tests were conducted in order to obtain mainly ultimate flexural strength of RC beam with various anchorages of CCFP. In the experimental cases except FP0 and FP1-G, the bending span was 1500 mm. However, such bending

Anchorage methods

Bottom of RC beam

RC beam specimen

Edge

FP1-G

Layer CCFP Epoxy Adhesive Concrete

C.L.

600mm

Table 3.

Experimental results. 2M

Specimen

1M

FP0 FP1 FP1-H FP1-H-A FP1-Pl-A FP1-S FP1-L FP0’ FP1-G

15.0 kNm 15.1 kNm 15.6 kNm 15.6 kNm 18.5 kNm 14.9 kNm 14.8 kNm 12.1 kNm 21.1 kNm

u

3ε

y

D10(SD295A)

910mm

30mm

950mm

75mm

-SpecimensFP0, FP1, FP1-H, FP1-H-A, FP1-PL-A, FP1-S, FP1-L CCFP 750mm

8.2 kNm 11.7 kNm 12.9 kNm 11.3 kNm 14.6 kNm 12.7 kNm 12.0 kNm 7.6 kNm 13.5 kNm

4ε

u

– 3480 × 10−6 2780 × 10−6 – 4145 × 10−6 3020 × 10−6 2650 × 10−6 – 6215 × 10−6

y

– 2049 × 10−6 1791 × 10−6 – 2580 × 10−6 1777 × 10−6 1822 × 10−6 – 2087 × 10−6

75mm

1M

u

2M

y

Ultimate bending moment (=Maximum moment). Bending moment at re-bars yielding. 3 ε Maximum strain of CCFP. u 4 ε Strain of CCFP at re-bars yielding. y

-SpecimensCCFP 850mm

Detail of RC beam specimen. Properties of materials for RC beam.

Reinforcing bar (D10) Diameter Yielding Strength Tens. Strength Concrete Comp. Strength Splitting Tens. Strength Young’s Modulus Epoxy Adhesive Density Comp. Strength Tens. Strength (Tens.) Shear Strength Young’s Modulus

10 mm 350 N/mm2 483 N/mm2

20

FP1-Pl-A FP0

15

FP1

10 5

Deflection at Center 0 0

25.6 N/mm2 1.85 N/mm2 33 kN/mm2

5

10

15

20

25

30

35

40

45

Deflection (mm)

I) Bending tests with span of 1500 mm 25

1.64 g/cm3 79.2 N/mm2 36.5 N/mm2 17.9 N/mm2 5.8 kN/mm2

Bending Moment (kNm)

Table 2.

Bending Moment (kNm)

25

Figure 3.

FP1-G 20 15

FP0'

10 5

Deflection at Center

span test may not provide bending failure for FP1G, so the bending tests with the span of 1700 mm were conducted in case of FP0 and FP1-G. In these experiments, we measured the deflection at the center span and strain of re-bar or CCFP for quantification of strengthening-effect by CCFP bonded. 2.5 Experimental results Table 3 gives the ultimate bending strength of each specimen and the bending moment when re-bar yielded. Additionally, Table 3 gives the strains of CCFP at the maximum moment or the strains of CCFP at the re-bar yielding. According to the experiments with bending span of 1500 mm, the ultimate strength of FP1 which has simply bonded of CCFP was almost equal to the strength of RC beam without CCFP. Other specimens

261

0 0

5

10

15

20

25

30

35

40

45

Deflection (mm)

II) Bending tests with span of 1700 mm

Figure 4.

Bending moment—deflection at center.

with reinforcing the edge of CCFP had strength almost equal to or slightly higher than that of FP0. FP1-Pl-A had obviously higher strength than simple RC beam, but it used anchor bolts which may disturb the flatness of cantilever RC slabs. Some experimental results of the relation between bending moment and deflection at center are shown in Figure 4 (I). FP0 without CCFP had the largest deflection in the test series of 1500 mm span. The maximum deflection of FP1 with simply bonding CCFP

was smaller than that of FP0; however, it indicated brittle failure with de-bonding of CCFP at the maximum load. When the yielding strain of re-bar is assumed as 1800 × 10−6 , Bernolli-Euler theory was kept in all specimens bonded with CCFP under the bending moment of M y . The strains of CCFP, however, were lower than the strains of re-bar after yielding. The results imply that CCFPs were anchored insufficiently on the concrete in such higher load condition. Most of these failures were caused from debonding between CCFP and surface which was relatively weak mortar. For obtaining tougher anchorage of CCFP, such weak layer must be removed as much as possible. FP1-G which includes CCFP in ditch of full length had the highest bending strength among the experiments conducted in this study. Figure 4 (II) indicates FP1-G had extremely smaller deflection and higher strength than normal RC beam. Based on these results, we decided that to embed plates in ditch of concrete surface is one of the most effective anchorage-methods for CCFP.

3.2

Loading program

Initial load in the fatigue test was set as 78.4 kN that is almost equal to a rear wheel load of large trucks. After the cantilever slab without CCFP was subjected to the initial load with repetition of 60000 cycles, cracks with width of 0.3 mm was occurred. Such cracked slab was strengthened by embedding CCFP on the surface, the initial wheel load was supplied again. The providing load was increased to 88.2 kN after 240000 cycles of the initial load so the slab had little deteriorated by repetition of the wheel load. However, the second loading could not provide obviously the deterioration, so the load was increased again to 98.0 kN. Each repetition of the increased wheel-load was 60000 cycles respectively.

WHEEL RUNNING FATIGUE TEST

3.1

3.3

Wheel running fatigue test

In order to obtain behavior under real load condition, the wheel running fatigue test was conducted by employing the RC slab model strengthened by CCFP. In the present experiment, we designed the RC model based on former specification (1964) in Japan; many bridges including such RC slabs deteriorated by wheel load have been used even now. The outline of the wheel running fatigue test is illustrated in Figure 5. As shown in this figure, the loading line was planned as the outside of 700 mm from the main girder (support B). Thus, the employed RC slab model was certainly cantilever slab subjected to negative bending. CCFPs were embedded in the surface of the slab model as same

190mm

200mm

1000mm

500mm

700mm

450mm

5

CC FP

fixe d support Deflection Support A

4

Support B Hard Rubber

. . . . . .

1 6@200mm

CCFP

4500mm

Experimental result

The cantilever slab was not fractured even by the cyclic loading of the sum of 360000 cycles that included 60000 + 180000 (the initial load) and each 60000 (the second or third loads). Such loading history was almost equivalent to 1.48 million cycles of the initial wheel load as based on Miner’s rule. The results represented that cantilever RC slab strengthened by CCFP has extremely higher durability to cyclic loading. Figure 6 presents the relation between the cyclic loading numbers and the deflection at the center of the cantilever slab. The graph shows the deflections at loading or unloading. As shown in this figure, deflection after bonding with CCFP was decreased as compared with normal RC cantilever slab (before strengthening by CCFP). The present study evaluated

Deflection (mm)

3

way of FP1-G, which had the most effective anchorage method in the above experiments. The bond length of CCFP and the interval of each plate were designed as 1550 mm and 200 mm respectively. In order to design the arrangement of CCFP, the present study referred to the bending theory of RC member.

Loading

Loading

(without CCFP)

(with CCFP)

3

Unloading (with CCFP)

2

Unloading

1

(without CCFP)

78.4 kN 88.2 kN 98.0 kN

0 0

Figure 5.

Outline of the wheel running fatigue test.

6

12

18

24

30

36

Number of Cycles ( 104)

1550mm

Figure 6.

262

Deflection at the center of cantilever slab.

42

Static Load 200mm 190mm

1,000 800

100mm

500mm

700mm

450mm

CCFP

fixed support Support B

Support A

600

Deflection Line of Load CCFP

200

78.4 kN 88.2 kN 98.0 kN 0 0

6

12

18

24

30

36

5@150mm

400

700mm

Elastic Strain of CCFP ( 10 -6)

1,200

42

Number of Cycles ( 10 4)

Figure 7.

Top surface

Elastic strain of CCFP (loading minus unloading).

1550mm

a) Static strength test

4.1

ULTIMATE FATIGUE STRENGTH

500mm

190mm

100mm

700mm

450mm

CCFP

Deflection

Support C Line of Load CCFP

5@150mm

4

Cyclic Load 200mm

700mm

the elastic strain of CCFP by subtracting the strain at unloading from the strain at loading. Elastic strain of CCFP with cyclic loading was presented in Figure 7. Figure 7 indicates the strain was little increased or decreased with numbers of loading. This experimental result implies that CCFP had sufficiently bonded on the surface and could strengthen RC slab under cyclic loading.

Purpose of fatigue test Top surface

4.2

Static strength under negative bending

In the present study, static strength was evaluated experimentally for determination of load for fatigue test. Another purpose of the static test was to investigate the ultimate failure mode of cantilever member. The negative bending test for cantilever beams is illustrated in Figure 8. The employed cantilever specimens were 700 × 1900 × 190 mm as shown in Figure 8a. Such specimen was designed as the same spec of the large slab specimen for wheel running fatigue test. The static tests were conducted twice, i.e. normal RC beam and RC beam with CCFP. Figure 9 shows the relation between the bending moment and deflections. In addition, the specimens after bending test are given in Figure 10. Deflection of CCFP beam was smaller than that of RC beam. Such tendency

263

1550mm

b) Fatigue strength test

Figure 8. Negative bending test for cantilever beam.

120

Bending Moment (kNm)

The fatigue properties under wheel loading were obtained from the above experiment. However, the ultimate fatigue strength could not be estimated from the wheel running fatigue test so the fatigue failure was not distinct. Thus, the fatigue test by using RC beam with CCFP was conducted in order to evaluate the ultimate fatigue strength of such structures. The present study especially investigated the fatigue behavior of cantilever slabs subjected to negative bending.

with CCFP

100 80 60

without CCFP

40 20 0 0

5

10

15

20

25

30

35

40

Deflection (mm)

Figure 9.

Static bending moment and deflection.

is almost similar to the experimental results mentioned in the above chapter. The maximum moment of CCFP cantilever slab was 112.5 kNm as the ultimate strength. The strength was extremely higher than the

7

Support A

Loading (with CCFP)

Deflection (mm)

6

a) RC cantilever beam without CCFP Support B

5

Loading

4

(without CCFP)

3

Unloading (with CCFP)

Unloading

2

(without CCFP)

1

37.9~6.69 kNm

53.2~6.69 kNm

0 0

5 10 15 20 25 30 35 40 45 50 55 60 65 70 75 80 85 90 95

Number of Cycles (×104)

Figure 11.

Deflection in the bending fatigue test.

b) RC cantilever beam with CCFP

Figure 10. Crack conditions after negative bending test.

ultimate strength (66.8 kNm) of the normal RC specimen. Figure 10 represents that RC beam with CCFP had smaller and more cracks than RC beam without CCFP. In other words, the crack condition shows that CCFP was able to prevent harmful wider cracks. 4.3

Fatigue strength under pseudo-negative bending

The ultimate fatigue test of negative bending was conducted as shown in Figure 8b. For providing the negative bending, a cantilever slab specimen was set reversely. So the providing force from the fatigue machine is almost equivalent to a reacting force on main girders nearest cantilever slabs. The initial maximum force and the minimum force were determined as moment of 37.9 kNm and 6.69 kNm, which were equivalent to 57 or 10 percent of the static strength of RC specimen without CCFP. In the fatigue test of RC beam without CCFP, rebar was broken down at the numbers of 140000 loadings. On the other hand, RC beam with CCFP had little fatigue-deterioration after the loading of 550000 cycles. Thus, the maximum moment was determined as 53.2 kNm that was equivalent to 80 percent of the static strength of RC beam. After loading 350000 cycles, the specimen was fractured by debonding CCFPs from the concrete surface. Such loading history was almost equivalent to 26.7 million cycles of the initial moment as based on Miner’s rule. The experimental result implies that CCFP can provide RC member extremely higher durability under fatigue loading. Figure 11 presents

the change of deflections at the loading point. This graph indicates that the deflection had little changed before upgrading the load. Additionally the result represents the deflection had gradually increased after the upgrading the load. The ultimate fatigue failure was due to exfoliation of CCFP as above mentioned, so CCFP may strengthen cantilever slabs more and more if CCFP was anchored on concrete more toughly. 5

CONCLUSIONS

The present study focused on cantilever RC slabs subjected to negative bending, and reported the effective anchorage method and its fatigue properties. The conclusions of the present study are listed as follows: 1. CCFP simply bonded on concrete could hardly strengthen RC member under bending. Reinforcing the edge of CCFP with various materials was not so effective. 2. To embed CCFP in ditch of concrete surface could provide the highest strength and fatigue durability of RC beam or slab in the present experiments. 3. CCFPs bonded on the cantilever slab had little fatigue deterioration even though the wheel was loaded directly many times. 4. CCFP can provide extremely longer fatiguedurability to cantilever RC slabs. REFERENCE Japan Road Association 1964. Specification for Road Bridge.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Structural behaviour of reinforced palm kernel shell foamed concrete beams U.J. Alengaram SEGi University College, Malaysia

M.Z. Jumaat & H. Mahmud University of Malaya, Malaysia

ABSTRACT: Experimental results of an investigation conducted to study the structural behaviour of reinforced palm kernel shell foamed concrete (PKSFC) beams prepared using palm kernel shell (PKS) as lightweight aggregates (LWA) are reported. The use of foam in the PKS concrete was to reduce the density of PKS concrete from 1900 kg/m3 to 1650 kg/m3 and at the same time to produce grade 20 concrete using mineral admixtures. 10% silica fume and 5% class F fly ash were used as additional and cement replacement materials respectively in PKSFC, but normal weight concrete (NWC) contained none. The test results show that flexural and shear capacities of the PKSFC beams were found close to that of NWC beams. The PKSFC beams exhibited more cracks within the flexural zone and shear cracks in the shear zone than NWC beams. The deflection of PKSFC was found to be higher than that of NWC beams.

1

INTRODUCTION

1.1 Use of waste material as lightweight aggregate Malaysia being the second largest palm oil producer in the world and has total palm tree planted area coverage of 3.8 million hectares. Every year, the production of palm kernel shell (PKS) as a waste from palm oil industry is approximately 4 million tones. The PKS has the potential to be used as coarse aggregate in concrete. PKS is hard and light and hence it can be utilized to replace conventional coarse aggregate to produce lightweight concrete. These types of waste materials, when properly processed, have shown to be effective as construction material. 1.2 Lightweight aggregate foamed concrete RILEM classifies lightweight concrete (LWC) with compressive strength of 15 N/mm2 or more as structural grade concrete—class-I, and strength between 3.5 and 15 N/mm2 as structural and insulating concrete— class-II (FIP 1983). LWC in the form of aerated concrete is either a cement or lime mortar, classified as lightweight concrete in which air-voids are entrapped in the mortar matrix by means of suitable aerating agent (Valore 1954). The main advantage of aerated concrete is its lightweight, which economizes the design of supporting structures including foundation and walls of lower floors. For cellular concrete the voids may occupy

265

anything from 25 to 80 percent of the total volume. As with conventional concrete, the strength of aerated concrete depends on its density; in addition, water-cement ratio, addition of admixtures and curing conditions also play a role in strength development. LWC in the forms of LWAC and foamed concrete played a dominant role in the development of LWC as useful construction material. However, only very few literatures are available on the properties of concrete containing foamed mortar and LWA. Weighler and Karl (1980) conducted experiments on such concrete that contained light expanded clay aggregates (leca) with foamed mortar. They concluded that the foamed concrete incorporating LWA has lighter density and good thermal insulating property. This was due to the voids created by the foam inside the concrete. Thus, for the concrete tested, they found the density and the compressive strength in the range between 700 to 1200 kg/m3 and 5 to 30 N/mm2 , respectively. This proved to be structural and insulating concrete. However, further works on using such structural and insulating concrete hadn’t been carried out. This necessitates further works on lightweight aggregate foamed concrete. Generally manufactured lightweight aggregates cost more than dense natural/ gravel aggregates. Also, manufactured and natural LWA are not available in Asia and use of such LWA is not common. But studies on the use of organic waste materials such as palm kernel shells (PKS) as lightweight aggregates as construction materials are on the rise in Asia and Africa (Abdullah 1984; Okafor 1988; Basri et al. 1999; Ata et al. 2006).

1.3

Objectives

The objective of this work was to produce palm kernel shell foamed concrete (PKSFC) of grade 20 and to study its structural behaviour with respect to flexure and shear and to compare with that of normal weight concrete (NWC). Two beams one each on PKSFC and NWC were prepared for testing in flexure. However, to study the shear behaviour, four beams were prepared. In addition two beams on NWC were also prepared for comparison. The variable in shear test was the shear span to effective depth ratio (a/d).

2 2.1

MATERIALS

replacement. Similarly, 10% of Silica fume (SF) in undensified form with specific gravity of 2.10 was used as additional cementitious material. The use of SF was to develop a good bond between PKS and the cement matrix. 2.2

Mining sand of relative density of 2.7 and of size between 0.15 and 2.36 mm was used as fine aggregates. The PKS obtained from local crude palm oil producing mill were used as coarse aggregates. Figure 1 (a) shows the shells and it can be seen that they possess curved and irregular surfaces. 2.3

Cement and mineral admixtures

Ordinary Portland cement conforming to MS 522; Part-1:2003 with specific gravity of 3.10 was used for all mixes in this investigation. 5% of class F fly ash (FA) by cement weight was used as cement

3.1

Figure 1.

b) Pre-formed foam

Palm kernel shell and foam.

SAMPLE PREPARATIONS AND TESTING Mix proportion and mixing

The mix design for the PKSFC specimens based on the specific gravities of constituent materials was done. The quantity of foam added was calculated based on the target density of 1650 kg/m3 . However, P

P

Hanger Bar

Foaming agent and superplasticizer

Synthetic foaming agent (polyoxyethylene alkyl ether tenside) with specific gravity of 1.02 was used in the investigation. It was supplied by BASF, a German based chemical company and diluted in the water in the ratio of 1:19 to produce foam. The generated stable foam is shown in Figure 1 (b). The superplasticizer (SP), Rheobuild 1000 M with specific gravity of 1.21 was used at about 0.5% by cement weight.

3

a) Palm kernel shell

Fine and coarse aggregates

R6 @ 70 mm c/c

2T10

a) Reinforcement cage for beam in flexure

P

Note: The clear cover to reinforcement = 30 mm. Effective depth = 209mm

Hanger Bar R6 @ 70 mm c/c

2T10

b) Reinforcement cage for beam in shear for a/d = 1.0 (For beams with a/d = 2.0, the link spacing was 150 mm c/c)

Figure 2.

Reinforcement details of test beams.

266

for NWC concrete, the department of environment (DOE) method was adopted. The following mix ratios were used for PKFSC beams: sand to cement ratio (s/c) = 1.6; aggregate to cement ratio (a/c) = 0.8; water to binder ratio (w/b) = 0.35; superplasticizer (SP) content = about 0.5% by cement weight. The cement content for PKSFC was about 370 kg/m3 . However for NWC, the ratios were in the following order: s/c and a/c = 3.6; w/c = 0.75; SP = 0.5%. Cement content for NWC = 260 kg/m3 . The mixing was done in the following order: firstly PKS in saturated surface dry condition was added with dry sand and mixed in mixer for about 2 minutes. Then one-half of cement and cementitious materials were added and part of water with superplasticizer was added. Then remaining materials were added and mixed. 3.2 Beam preparation and Instrumentation The reinforcement details of beams are shown in Figure 2. It can be seen from the sketch that there was no holding bar provided in the pure moment zone for beams in flexure. For beams in shear, the minimum shear reinforcement was provided based on the a/d ratio of 1.0 and 2.0. The clear cover for all beams was 30 mm. Companion concrete specimens were cast to study the mechanical properties of PKSC. Compressive strength of 100 mm cubes, flexural strength of 100 × 100 × 500 mm prisms and splitting tensile strength on 150 mm diameter × 300 mm height cylinders have been carried out. All beams designed using BS 8110-Part 1:1997 were cast in steel moulds. They were vibrated using internal vibrator and covered with jute clothes for 28 days and cured. Afterwards the beams were kept under laboratory condition till the

Table 1.

Details of beams tested in flexure and shear.

day of testing at the age of about 60 days. An Instron machine with a built in capacity of 500 kN load cell was used in testing of beams. The tensile and compressive strains of both reinforcement and concrete were measured through electrical resistance gauges. All the strains were recorded using data logger. In addition, the strain distribution on the vertical face of the beams in the flexural zone was determined using de-mountable digital extensometer with a sensitivity of 0.001 mm. Three linear voltage displacement transducers (LVDT) were placed, one at centre of beams, the other two under load points, to measure the deflections at centre and under load points for beams in flexure. For beams in shear, the vertical deflections were measured under the load single point load was applied for beams tested in shear. However, two-point loads were applied for beams in flexure. All strain, crack width and deflection measurements were measured at every load increment. Table 1 and Table 2, respectively show the details and concrete and steel properties of beams tested in flexure and shear. The beams tested in flexure and shear are designated as FB and SH, respectively. Similarly, the materials are indicated as NWC and PKSFC for normal weight concrete and palm kernel shell foamed concrete, respectively. 3.3

Properties of PKS

The natural moisture content and 24 hour water absorption of PKS were determined. The thicknesses and the size of PKS were also measured. The particle size distribution and the specific gravity were determined. The loose and compacted densities were also found. The results of PKS are shown in Table 3.

Table 3.

Beam designation

Material

Type of test

No of beams

a/d*

FB-P1 FB-N1 SH-P1 & P2 SH-N1 SH-P3 & P4 SH-N2

PKSFC NWC PKSFC NWC PKSFC NWC

Flexure Flexure Shear Shear Shear Shear

1 1 2 1 2 1

3.11 3.11 1.0 1.0 2.0 2.0

Physical property of palm kernel shells.

Properties

Values

Size (mm) Thickness (mm) Bulk density (loose) (kg/m3 ) Bulk density (compacted) (kg/m3 ) Specific gravity Water absorption (1 hour) (%) Water absorption (24 hour) (%)

2–15 1.0–3.0 568 620 1.27 12 25

* Shear span to effect depth ratio. Table 2.

Details of concrete properties and reinforcement details (for beams in flexure and shear).

Material

Saturated density (kg/m3 )

Slump (mm)

Cube strength (N/mm2 )

Modulus of rupture (N/mm2 )

Young’s modulus (kN/mm2 )

Overall beam size (mm)

Main bar size (mm) yield strength (N/mm2 )

NWC PKSFC

2341 1675

76 120

23.78 18.71

3.44 2.33

27.0 8.0

152 × 251 149 × 252

2T-10 for flexure and shear; fy = 483

267

Steel ratio, ρ = As/bd (%) 0.5

4 4.1

RESULTS AND DISCUSSION

for higher shear resistance could be the aggregate interlock of the PKS that contribute to shear resistance. In addition, the link spacing that was provided based on BS had sufficient shear resistance to prevent shear cracks. The modes of shear failure for beams are shown in Figure 4. The failure of NWC was sudden and brittle in contrast to the ductile failure of the PKSFC. This was the one of the main findings and also an advantage as PKSFC gives adequate warning before failure. As reported by Kong (1995), for shear span to effective depth ratio between 1 and 2, the shear cracks start independently and not as development of flexural cracks. This was observed in experiments and thus PKSFC and NWC beams behaved similar manner. In shear test, some cracks started as flexural cracks, and then converted into shear crack. Therefore, a lot of cracks, so called flexural-shear cracks occurred. Shear crack was usually propagated faster than flexural crack although the links had been provided to resist it. Thus shear failure behaved in brittle manner compared to that of ductile failure mode of flexural failure.

Mode of failure

4.1.1 Beams in flexure The flexural behaviour test on two beams, FB-P1 (PKSFC) and FB-N1 (NWC) were conducted using Instron testing machine. A comparison of test results between the PKSFC and the NWC was also made. In order to ensure flexural failure, the shear and compression reinforcement were not provided in the middle-third. The spacing of link was kept at 70 mm to obtain typical flexural failure. The beams were designed as under-reinforced, so steel bar yields first before the compressive failure occurs in the concrete. In the experiment, both beam shown typical flexural failures as yielding of steel took place followed by concrete crushing at compression zone. Since PKSFC was weaker than NWC, the compression zone of PKSFC had larger crushing area compared to NWC as shown in Figure 3. 4.1.2 Beams in shear For the shear capacity test, the varying shear span (a) to effective depth (d) ratio of, a/d = 1 and a/d = 2, was adopted. Three beams for each case have been tested to study the shear behaviour. The spacing of link for beam with a/d = 1 and a/d = 2 has been calculated based on BS 8110: 1997. It has been found from shear testing, that flexural cracks occur in the beginning followed by shear cracks near the support. The shear cracks start in the shear zone near the support and propagated towards the load point, but ultimately the beams failed in flexure due to the higher shear resistance of the beams. One possible explanation

Figure 3.

Flexural failure of beam.

Figure 4.

Shear crack patterns for beams in shear.

4.2

Failure load

4.2.1 Beams in flexure From Table 4, it can be seen that the first crack load of PKSFC beam, FB-P1 was lower than NWC. This is significant as it was shown earlier that PKSFC had lower modulus of rupture than NWC. In addition, the PKSFC was ductile material as compared to NWC that fails in brittle manner. Ductile behaviour of PKS made PKSFC to behave more elastic than NWC. Therefore, the first crack load of PKSFC was found lower than NWC. However, the number of cracks

268

Table 4.

Results on flexure test. Ultimate moment, (kNm)

Beam

First crack load (kN)

Experimental failure load (kN)

Experimental, Mexp

FB-N1 FB-P1

16 13

60.5 58.5

19.21 18.70

Table 5.

Shear capacity (a/d = 1).

Theoretical, Mtheo

Table 6.

Shear capacity, V (kN) Theoretical, Vtheo

BS

ACI

EURO

Mexp / Mtheo(BS)

13.36 13.01

13.14 12.91

12.67 12.43

1.44 1.44

Shear capacity (a/d = 2). Shear capacity, V (kN)

Beam

BS

ACI

EURO

Experimental, Vexp

SH-P1 SH-P2 SH-N1

60.57 60.57 71.75

68.07 66.58 73.59

61.57 57.12 68.09

78.14 70.83 92.77

was found more in PKSFC beams than NWC. This consequently resulted in smaller crack widths in PKSFC than NWC beams. The addition of silica fume and fly ash to the PKSFC increased the cohesiveness of the concrete. This enhanced the bond strength which enabled more number of cracks. The failure load and hence the experimental ultimate moment of NWC were higher than PKSFC.

Theoretical, Vtheo Beam

BS

ACI

EURO

Experimental, Vexp

SH-P3 SH-P4 SH-N2

34.06 34.06 40.96

41.47 42.30 47.97

40.74 41.66 48.95

40.29 41.08 47.60

and a/d = 2 shows that the shear capacity for a/d = 1 was higher than the shear capacity for a/d = 2. This is because the load point for a/d = 1 was nearer to the support, so it has to resist higher load. The ratio of experimental shear capacities between a/d = 1 and a/d = 2 was found as 1.83 and 1.94, respectively for PKSFC and NWC. 4.3 Strains of beams in flexure and shear

4.2.2 Beams in shear Table 5 shows the comparison between theoretical and experimental shear capacities for beams tested for shear of a/d = 1. The theoretical shear capacity has been calculated by referring to different design codes such as British Standard (BS), American Code (ACI), and Euro Code (EC). From the results, it can be seen that the experimental shear capacities of all three beams were found higher than the theoretical values. The theoretical calculation is made by assuming the procedure of testing is perfect and no error would occur during testing. Further, the NWC beams recorded higher shear capacities compared to that of PKSFC. The poor adhesion between PKS aggregate and cement matrix and the smooth surface of PKS were some of the factors that affect the compressive strength. Though the addition of SF imparts cohesion to the mix, the formation of foam in the vicinity of PKS makes the bond weaker. Thus, the strength of PKSFC was lower than the NWC. The values of theoretical and experimental of shear capacities for a/d = 2 are shown in Table 6. The results show similar trend as that of a/d = 1, as the experimental value was always higher than theoretical value based on BS code. The comparison for both shear capacity test of a/d = 1

269

Both beams of PKSFC and NWC recorded higher reinforcement strains. The strain at yielding of steel in both beams was close to 3000 microstrains. However, after yielding the PKSFC behaved in ductile manner thus, the final strain before failure was found about 7000 microstrains. As mentioned due to lower modulus of rupture PKSFC beam cracked earlier than NWC beam. Higher steel strain in PKSFC is also an indication that good bonding exists between steel and concrete. In addition, due to the ductile behaviour of PKSFC, the beams undergo higher strain compared to NWC. The maximum steel strain recorded for a/d = 2 was in the range of 2500 to 2780 microstrains. This shows that the reinforcement was close to yielding. Higher compressive strains were recorded for concrete for the PKSFC beam than for the NWC beam soon after the first crack. This was possibly due lower modulus of elasticity of PKSFC as it recorded only 30% of NWC. Thus, PKSFC beam had undergone larger deflection and resulted in larger concrete strains. 4.4

Deflection of beams in flexure and shear

Higher deflections were recorded for beams in flexure compared to beams in shear. The ductility ratio,

5. The lower modulus of elasticity of PKSFC resulted in higher deflection than NWC. However, the deflections at serviceability limit state were within limits. 6. The ductility ratio of PKSFC was higher than NWC, and thus proved the ductile nature of PKSFC. 7. PKSFC beams had more number of crack compared to NWC beams. Thus, the average crack spacing and crack width were found lower in PKSFC beams. 8. Experimental shear capacities generally found higher than theoretical values.

FB-P1

FB-N1 70 60

Load (kN)

50 40 30 20 10 0 0

10

20

30 40 Deflection (mm)

50

60

ACKNOWLEDGEMENTS Figure 5.

Table 7.

Mid-span deflection of beams in flexure. Deflection of beams in shear (a/d = 2).

This research work is funded by Ministry of Science, Technology and Innovation (MOSTI) under Science Fund Grant no. 03-01-03-SF0309.

Service deflection under load point (mm) Theoretical, δtheo

REFERENCES

Beam

BS

ACI

EURO

Experimental, δexp

SH-P3 SH-P4 SH-N2

6.84 7.02 4.55

6.97 7.06 4.02

8.09 8.28 5.10

6.01 4.29 2.78

defined as the ratio of deflection at yield stage to ultimate stage was found twice for PKSFC beams compared to NWC beam in both flexure and shear. Figure 5 shows the mid-span deflection of beams in flexure. Table 7 shows the service deflection of beams in shear under the load point for a/d = 2.0. The experimental deflections of PKSFC beams are higher than the NWC beams. Based on the ductility ratio of beams in flexure, it can be concluded that the behaviour of PKSFC beam has higher ductility characteristics than the corresponding NWC beam. The PKSFC will give ample warning before failure and such behaviour is best suited for structures in earthquake prone region. 5

CONCLUSIONS

1. The first crack load of PKSFC beams was found to be lower than the corresponding NWC beams. 2. Both the NWC and PKSFC beams failed in typical flexural mode. 3. The experimental flexural capacities of the shear link spacing calculated using BS8110: Part 1: 1997 was sufficient to resist shear crack from widening and prevent the shear failure in PKSFC beams. 4. The higher reinforcement strain recorded for PKSFC beams in flexure shows that the bond between PKS and cement paste is stronger.

Abdullah, A.A.A., 1984. Basic Strength Properties of Lightweight Concrete Using Agricultural Wastes as Aggregates, Proceedings of International Conference on Low-cost Housing for Developing Countries, Roorkee, India. ACI manual of concrete practice, Part 3-2002, American Concrete Institute 2002. Ata, O., Olanipekun, E.A., Oluola, K.O. A comparative study of concrete properties using coconut shell and palm kernel shell as coarse aggregates. Building and Environment 2006; 41: 297–301. Basri, H.B., Mannan, M.A., Zain, M.F.M. Concrete using waste oil palm shells as aggregates. Cement and Concrete Research 1999; 29: 619–622. BS 8110: Part 1:1997. Structural use of concrete. Part 1. Code of practice for design and construction, British Standards Institution 1997. BS 8110: Part 2:1985. Structural use of concrete. Part 2. Code of practice for special circumstances. British Standards Institution 1985. Eurocode 2: Design of concrete structures; Part 1 General rules and rules for buildings, 1992. FIP Manual of Lightweight Aggregate Concrete, 2nd ed. London: Surrey University Press; 1983. Kong, F.K., Evans, R.H. Reinforced and Prestressed Concrete, 3rd edn. Chapman & Hall; 1995. Okafor, F.O. Palm kernel shell as a lightweight aggregate for concrete. Cement and Concrete Research 1988; 18: 901–910. Valore, R.C. Cellular concretes-physical properties. J Am Concr Institute. 1954; 25: 817–836. Weighler, H., Karl, S. Structural lightweight aggregate concrete with reduced density-lightweight aggregate foamed concrete. International Journal of Lightweight Concrete 1980; 2: 101–104.

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Dynamic impact and earthquake engineering

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

A note on the model based on the constant Q damping assumption and its corrected models A.S. Takahashi Nagoya Institute of Technology, Nagoya, Japan

ABSTRACT: As Crandall indicated, the structural damping (constant Q damping) assumption does not satisfy the causality law. In addition, we cannot find the expression about its relaxation function. To improve these defects, its corrected models have been proposed. But these models, including constant Q model, have a problem which it is difficult to calculate these time domains function because of logarithmic features as complex functions. In this article, we solve these problems by using well-defined distribution function. We clarify the relaxation function of constant Q model, and consider its features as a viscoelastic model. Furthermore, we express these models collectively in both frequency and time domains, and by comparing each others, we evaluate corrected models of constant Q model.

1

INTRODUCTIONS

There are many ways to determine the damping of structures, and Figure 1 shows the relation between natural period and damping coefficient. Proportional damping is useful for analysis, but in high frequency range, it evaluates the effect of damping higher. Then, in the case that we cannot ignore the effect of higher mode, we use constant Q model (Q model, in the following) to estimate the damping of structures. The model seems simple one because real and imaginary parts of its impedance are constant in any frequency range, but the model has fatal defects. As Crandall indicated in 1963, this model doesn’t satisfy the causality law, that is, there is a response prior to the excitation. Furthermore, we cannot find the expression about its relaxation function.

To improve these defects, its corrected models were proposed. In 1958, Biot proposed a model (B model, in the following) which is based on the infinite combination of generalized Maxwell model. In 1990, Makris proposed a model (M model, in the following) which based on Q model (1997). He defined the real part of M model’s impedance so as to make Hilbert transform pair with imaginary part of Q model. In this article, by making full use of mathematics, we compare Q model with these corrected models, and make all the details of it clear (Crandall 1963).

2 2.1

COMPLEX FUNCTION THEORY Q model

Impedance of Q model is usually expressed in the following Equation.   1 G(j) = G0 1 + j (sgn ) Q

(1)

In Equation 1, j is the imaginary unit and we assume  is independent frequency. Before transferring it into complex function, we introduce the complex frequency function concerned with Q model. FP (s) = Figure 1. Relation between natural period and damping coefficient (Takizawa 2006).

273

log(s/ω0 ) [ω0 > 0] s

(2)

Associate with it, we also introduce the time domains function.

1 fp (t) = 2π j

+0+j∞ 

f (t) = fp (t) + fn (t) = − {γ + log(ω0 |t|)}

FP (s) exp(st)ds +0−j∞

= − {γ + log(ω0 |t|)} 1(t)

(3)

In Equation 3, γ is Euler’s constant and 1(t) is Heaviside function. On this Bromwich integration, we need the distribution function related to the logarithmic feature. It is defined in reference 4, we don’t describe the detail. Figure 2 shows the singularity of FP (s) and shape of fp (t). Furthermore, we introduce similar complex frequency function. FN (s) = −

log(−s/ω0 ) [ω0 > 0] s

(4)

This function is symmetric to FP (s) with respect to imaginary axis. Similarly, the time domains function associate with it is below. 1 fn (t) = 2π j

+0+j∞ 

FN (s) exp(st)ds

= − {γ + log(ω0 |t|)} 1(−t)

(5)

Figure 3 shows the singularity of FN (s) and shape of fn (t). Then, we add these two functions in frequency and time domains respectively. F(s) = FP (s) + FN (s) = = jπ

log(s/ω0 ) − log(−s/ω0 ) s

sgn(Im s) s

Figure 2.

Singularity of FP (s) and shape of fp (t).

Figure 3.

Singularity of FN (s) and shape of fn (t).

Figure 4 shows the singularity of F(s) and shape of f (t). We have to pay attention to the fact that there is not the parameter ω0 in F(s), but there is the parameter ω0 in f (t). This is the reason of the misunderstanding we cannot define the relaxation function of Q model. This feature makes huge difficulty in transferring between time and frequency domains. By using these relations, we transfer Equation 1 into complex function so as to gain the relaxation function from it. The complex function is below.   G(s) = G0 1 + j(sgn(Im s))Q−1   Q−1 {log s − log(−s)} (8) = G0 1 + π 2.2

B and M model

B model is based on the infinite combination of generalized Maxwell model (Figure 5). The impedance is expressed in the following form. G(s) = G0 {1 + q log(1 + sτU )}

+0−j∞

(7)

(9)

The singularity of this model is expressed in Equation 10, and it is superior to Q model in which it doesn’t include the origin. −∞ ≤ Re s ≤ −1/τU & Im s = 0

(10)

On the other hand, Makris proposed a model which based on the Q model. He made the real part of its

(6)

Figure 4.

Singularity of F(s) and shape of f (t).

Figure 5.

Viscoelastic model proposed by Biot.

274

impedance as the Hilbert transform of the imaginary part. As a result, the impedance of M model is given by Equation 11. G(s) = G0 {1 + q log(sτQ )}

(11)

The singularity of this model is expressed in Equation 12, and it is identical to aforementioned function FN (s). −∞ ≤ Re s ≤ 0 & Im s = 0

3

(12)

⎧ ⎪ ⎪ ⎨−q log χ

VISCOELASTIC THEORY

3.1 Comparison in impedance of damping models Already we got impedance of three models as complex functions, and in this section, we compare them with each other. For making it easy to understand differences among these models, we divide impedance into real and imaginary parts (Table 1). For convenience in comparison, we introduced several parameters. First, we defined the material damping coefficient ζ by the ratio of loss modulus to storage modulus. ⎧ 1 ⎪   ⎨ (Q) 1 Im G(j) 2Q (13) = ζ = 2 Re G(j) =ω ⎪ ⎩ π q (B, M) 4 Second, we introduced dimensionless parameters α (= ωτU ) and χ (= ωτQ ), and these are significant for these models to be well approximations of Q model. For making them the best approximation of Q model, we determined these parameters so as to the difference in real and imaginary parts become smallest. In the case of B model, real and imaginary parts increase monotonously, and in the range of ω <  < ωU , the difference in real part εR and imginarypart εI are given by Equation 14 and 15.

 ω 2  q U (14) εR = log 1 + α 2 2 ω εI = 1 −

2 tan−1 α π

The value of α depends on ζ hardly, and Figure 6 shows the relations between these values and dimensionless parameter in the case of ωU /ω = 10. According to our research, the influence of ωU /ω is small when the value is realistic one. Then we determined α = 1, 2, 3, 7, 12 for ζ = 0.01, 0.02, 0.05, 0.1, 0.2 respectively to make both εR and εI small simultaneously. In the case of M model, imaginary part is equal to that of Q model, and similary to the case of B model, εR is given by Equation 16. 1 for χ < √ ωU /ω εR = 1 ⎪ ⎪ ⎩q{log χ + log(ωU /ω)} for χ > √ ωU /ω (16) The value of χ is almost constant toward any value of ζ , and the value is χ = 0.4. According with intro˜ duction of these parameters,  should be (= /ω). Figure 6 shows their real and imaginary part under ˜ is small, both B and M ζ = 0.2. In the range  model are not well approximations of Q model. Real part of M model and imaginary part of B model are too small compare to these of Q model. According

Figure 6.

Difference between B and Q model.

Figure 7.

Difference between M and Q model.

(15)

Table 1. Comparison in real and imaginary parts of impedance. Re G(j)/G0

Im G(j)/G0

Q model B model

1 1+

M model

1+

˜ 2ζ sgn 4 −1 ˜ π ζ tan (α ) ˜ 2ζ sgn

2 π 4 π

˜ 2) ζ log(1 + (α ) ˜ ζ log(χ||)

275

˜ becomes larger, real parts of B and M model to  become larger than Q model, but the degree is not large because of logarithmic function. Generally, even if we determine dimensionless parameters appropriately, we cannot declare they are well approximations. 3.2 Relaxation function The relaxation function is calculated from impedance by using the relation as following. 1 g(t) = 2π j

+0+j∞ 

+0−j∞

Figure 10.

G(s) exp(st)ds s

In Equation 18, χ appears in the relaxation function of Q model. It is caused by ω0 in Equation 7, and we assumed ω0 = τQ−1 . We introduced τ = ω t, and Ei is exponential integral.

(17)

Equation 17 is known as Bromwich integration. In this calculation, we have to pay attention to singularities of these models, B and M model are regular in the right half of complex plane, so we can easily understand they satisfy the causality law. But Q model is singular on the whole of real axis, so it isn’t causal model. As results of Bromwich integration, we gained relaxation functions of three models. ⎧   2  ⎪ 1(τ ) − ζ γ + log |τ |/χ ⎪ ⎪ ⎪ π ⎪ ⎪  ⎪ 4 g(t) ⎨ 1 − ζ Ei (−|τ |/α) 1(τ ) = π ⎪ G0 ⎪ ⎪  ⎪ ⎪ ⎪ 4 ⎪ ⎩ 1 − ζ {γ + log (|τ |/χ)} 1(τ ) π

x Ei(x) = −∞

(Q) (B) (M) (18)

4.1

Comparison in real part of impedance.

Comparison in imaginary part of impedance.

[x < 0]

(19)

DYNAMICS Impulse response of single degree of freedom systems

In this section, we add the inertia to these models for argue impulse response of single degree of freedom system. So we assume aforementioned parameter ω is natural frequency of no dumped model. In 1964, Caughey argued the impulse response of B model. Complex frequency function H (s) of the model added the inertia G0 /ω2 is given by Equation 20. H (s) =

Figure 9.

exp y dy y

Figure 4 shows these relaxation functions. The relaxation function of Q model can be determined clearly by using the relation in Equation 6 and 7. But it exposes the defect that its relaxation function in not equal to zero in the range of t < 0. The relaxation functions of B and M model are equal to zero in the range of t < 0, so we can make sure that they satisfy the causality law. And in B model, relaxation function is divergent at t = 0 and converges on static stiffness quickly. On the other hand, that of M model becomes negative value according to ω t becomes larger. In addition, it diverges according to t becomes larger. It can be said M model has another defect whereas it is causal. 4

Figure 8.

Comparison in relaxation functions.

1 s2 + ω2 G(s)/G0

(20)

We can obtain this function of each model by substituting its impedance respectively, and it is featured by poles and brunch cut. In these cases, all poles are simple poles. For convenience, we transfer Equation 20 to 21.

276

H (s) =

1 1 where z = s/ω ω2 f (z)

(21)

It is important for calculating impulse response to understand characteristics of function f (z), and these functions of three models are expressed in Equation 22. ⎧ 2 −1 ⎪ ⎨z + 1 + jQ sgn(Im z) (Q) 2 f (z) = z + 1 + q log(1 + αz) (B) ⎪ ⎩z 2 + 1 + q log(χ z) (M)

(22)

In these models, complex frequency function satisfies Equation 23. H (s) = H (s)

(23)

Then, all poles are complex conjugates pairs (including poles on the real axis). Furthermore, there is one complex conjugate pair in these cases. Figure 11 shows the position of pole corresponding to ζ . Figure 11 (a) is in the case of the parameter determined in 3.1, and (b) is the case of the parameter that of (a) multiplied by 7.5. Impulse response function h(τ ) associated with model can be expressed in Equation 24.

ωh(τ ) =

=

1 2π j

+0+j∞ 

+0−j∞



1 f

b + a

a + b

 (z

i)

In Equation 24, 1/f  (zi ) is the residue at i-th pole, and second and third terms express the influence of branch cut. From the view of numerical value, impulse response is almost the influence of complex conjugates pair poles, and effects of pole on the real axis and branch cut are too small. Figure 12 shows impulse response corresponding to ζ = 0.2 and the dimensionless parameters are similar to Figure 11. From these figures, parameters determined in 3.1 are not appropriate in the view of dynamics. And impulse response of M model in t < 0 is not zero, it means that this model is not causal in dynamics, whereas B model is causal in dynamics. The reason is a pole on real axis move to right half of complex plane in dynamics. But the effect of it is small, as abovementioned, so if we determined parameter χ appropriately, it can be ignored (Figure 12 (b)). Generally, for B and M model to be well approximation of Q model, values of dimensionless parameters are matters of great concerns. 4.2 Impulse response of multi degree of freedom systems We assume N degree of freedom system, and proportional damping in this section. And complex frequency function as vector in expressed in Equation 25. {H (s)} = (s2 [M ] + κ(s)[K])−1 [M ]{e}

1 exp(sτ )ds f (z)

where κ(s) = G(s)/G0 And via Equation 26, {H (s)} = (s2 [M ] + κ(s)[K])−1 [M ]{e}

exp(zi τ )

= [M ]−1/2 (s2 [I ] + κ(s)[M ]−1/2 [K][M ]−1/2 )

1 exp(σ τ )dτ f (σ + j 0) ⎫ ⎬ 1 exp(σ τ )dτ 1(τ ) ⎭ f (σ − j 0)

(24)

× [M ]1/2 {e}  N −1  s2 + ω2 κ(s) r −1/2 T {Vr }{Vr } = [M ] {Vr }T {Vr } r=1

(a) α = 1, x = 0.4

Figure 11.

(25)

× [M ]1/2 {e}

(b) α = 7.5, x = 3

Positions of poles corresponding to material dumping factor.

277

(a) α = 1, x = 0.4

Figure 12.

=

(b) α = 7.5, x = 3

Impulse response corresponding to material dumping factor. N 

1 {Ur }{Ur }T (s2 + ωr2 κ(s)){Ur }T [M ]{Ur }

r=1

× [M ]{e} =

N 

5

βr {Ur } s2 + ωr2 κ(s)

r=1

(26)

{Vr } = [M ]1/2 {Ur } {Ur }T [M ]{e} {Ur }T [M ]{Ur }

βr =

ωr , ω1

χr = χ

ωr ω1

(30)

CONCLUSIONS

Results of this paper are summarized as follows. 1. Relaxation function of Q model can be determined clearly by using the relations Equation 6 and 7. 2. M model doesn’t satisfy the causality law in dynamics. 3. Values of dimensionless parameters are important for B and M model to be well approximations of Q model.

ωr2 {Vr } = ([M ]−1/2 [K][M ]−1/2 ){Vr } where

αr = α

ACKNOWLEDGMENTS impulse response become following Equation 27. 1 {(τ )} = 2π j =

N 

+0+j∞ 

{H (s)} exp(sτ )ds

REFERENCES

+0−j∞

βr {Ur }hr (τ )

(27)

r=1

In this Equation, hr (t) is given by. 1 hr (t) = 2π j

This work is supported by Dr. Haruo Takizawa, associate professor of Hokkaido University.

+0+j∞ 

+0−j∞

1 exp(st)ds s2 + ωr2 κ(s)

(28)

and in Equation 28, κ(s) becomes as following. ⎧ (Q) ⎨1 + j2ζ (sgn(Im s)) κ(s) = 1 + q log(1 + αr (s/ωr )) (B) ⎩ 1 + q log(χr (s/ωr )) (M) Then, the parameter should be as follows.

(29)

Biot, M.A., ‘‘Linear thermodynamics and the mechanics of solids,’’ Proc. 3rd U.S. Nat. Cong. Appl. Mech., June. 1958, pp. 1–18. Caughey, T.K., ‘‘Vibration of dynamic systems with linear hysteretic damping(linear theory),’’ Proc. 4th U.S. Nat. Cong. Appl. Mech., June 1962, Vol. 1, pp. 3–18. Clough, R.W. and Penzien, J., ‘‘Dynamics of Structures,’’ MCGRAW-HILL International Edition, 1986. Crandall, S.H., ‘‘Dynamic response of systems with structural damping,’’ S. Lees(ed.), Air, Space, and Instruments—Draper Anniversary Volume, McGraw-Hill, 1963, pp. 183–193. Makris, N., ‘‘Causal Hysteretic Element,’’ Jour. Eng. Mech., Am. Soc. Civil. Engrs, Vol. 123, No. 11, Nov. 1997, pp. 1209–1214. Takizawa, H., ‘‘Logarithmic singularity of complex frequency functions examined at the infinity and origin points (With relation to the spectral expressions)’’, Journal of Structural Engneering (in Japanese), Vol. 52B, 2006. 3, pp. 175–188.

278

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

A neural-oscillator model for human-induced lateral vibration on footbridges M. Yoneda Kinki University, Osaka, Japan

ABSTRACT: Locomotion of animals, such as walking and running, is generated and controlled by the central nervous system called as the central pattern generator (CPG). Recently, the CPG framework has been utilized to develop locomotion controllers for autonomous walking robots. Therefore, a neural-oscillator model proposed by Matsuoka is investigated to grasp some useful information for human-induced lateral vibration on congested pedestrian bridges. The dynamic response analysis is also carried out taking into account the neural-oscillator model. Based on these results, it is confirmed that the neural-oscillator might be one of the effective models to explain the synchronization (the lock-in phenomenon) of a fairly high part of all pedestrians being on the bridge. 1

INTRODUCTION

The phenomenon of synchronous lateral excitation caused by pedestrians walking on footbridges such as the London Millennium Bridge has increasingly attracted public attention (Dallard et al. 2001). Up to date, this kind of dynamic oscillation has not led to structural failure, but has often caused discomfort for the users and the temporary closure of the footbridges in order to provide proper countermeasures. This excitation phenomenon for example the triggering of the lock-in phenomenon has not been fully understood or modeled (Fujino et al. 1993). By the way, locomotion of animals, such as walking and running, is generated and controlled by the central nervous system called as the central pattern generator (CPG). Recently, the CPG framework has been utilized to develop locomotion controllers for autonomous walking robots. Therefore, a neural-oscillator model is investigated to grasp some useful information for the human-induced lateral vibration on congested pedestrian bridges. The dynamic response analysis is also carried out taking into account the neural-oscillator model from the bridge design viewpoint.

τ2 v˙ 1 = −v1 + [x1 ]+

(2)

τ1 x˙ 2 = −x2 − βv2 − γ [x1 ]+  − + c − (−1.0) × uf

(3)

τ2 v˙ 2 = −v2 + [x2 ]+

(4)

[x]+ = max(x, 0)

(5)

yout = y1 − y2 = max(x1 , 0) − max(x2 , 0)

(6)

where xi is the firing rate, vi is a variable representing the self-inhibition of the neuron (modulated by the adaption constant β), and the mutual inhibition is controlled by the parameter γ . The output of each neuron yi is taken as the positive part of xi , and the output of the whole oscillator as yout · uf is the input applied to the oscillator. The tonic excitation c determines the amplitude of the oscillation, with amplitude proportional to c. There is no oscillation if c = 0. uf

2

+

c

v1 y1 = x1

NEURAL-OSCILLATOR MODEL

The neural-oscillator model consists of two simulated neurons arranged in mutual inhibition, as shown in Figure 1. The model for the neuron is taken from Matsuoka (Matsuoka 1985), and describes the firing rate of a real biological neuron with self-inhibition. The firing rate is governed by the following equations.  + τ1 x˙ 1 = −x1 − βv1 − γ [x2 ]+ + c − uf

(1)

279

input

x1

x2

+

+

output

+

y = y1-y2

uf

c

Figure 1.

v

y2 = x

2

+

2

Oscillator model proposed by Matsuoka (1985).

The two time constants τ1 and τ2 determine the speed and shape of the oscillator output. For stable oscillation, τ1 /τ2 should be in the range of 0.1∼0.5 (Williamson 1998).

Output (Oscillator) frequency (Hz)

3

4.0

NEURAL-OSCILLATOR BEHAVIOR

When no input is applied to the neural-oscillator, it oscillates at a natural frequency f , as shown in Figure 2. Natural frequency of the oscillator when β = 2.5, γ = 2.5, c = 1.5 is calculated under changing two time constants τ1 and τ2 . Figure 3 shows the output frequency of the oscillator plotted against 1/τ1 . It can be seen from this figure that the oscillator frequency is proportion to the magnitude of 1/τ1 under the condition that the ratio of τ1 /τ2 is constant. Therefore, the natural frequency of the oscillator when β = 2.5, γ = 2.5, c = 1.5 (CASE-252515) can be evaluated by using the proposed equation as shown in Table 1. The evaluation equation for β = 1.5, γ = 1.5, c = 1.5 (CASE-151515) is also shown in Table 2 (Yoneda et al. 2008). Based on the results of comparison between Table 1 and Table 2, it is assumed that the natural frequency of the oscillator slightly depend on the values of β and γ . On the other hand, the effect of the tonic excitation c on the output amplitude of the oscillator is investigated. Figure 4 shows the oscillator output under changing the value of the tonic excitation c. It can be seen from this figure that the value of amplitude/c and the output frequency are not varied although the output amplitude of oscillator increases according to the increase of the tonic excitation c. Based on these results, it is noted that the both peak output values are almost same under the condition that the ratio of τ1 /τ2 is constant. When an oscillatory input is applied, the oscillator can entrain the input, locking in the input frequency. This is illustrated in Figure 6 which shows the output of the oscillator as the size of the input signal is increased. It can be seen from this figure that the oscillator is not entrained for small input (top graph), and that the oscillator is almost entrained when the input is larger in the middle graph. Moreover, it can be said from

1/ 1/ 1/ 1/ 1/

3.0

2 = 0.1 2 = 0.2 2 = 0.3 2 = 0.4 2 = 0.5

2.0

1.0

0.0 0.0

10.0

20.0

30.0

Figure 3. Output frequency of the oscillator plotted against 1/τ1 (β = 2.5, γ = 2.5, c = 1.5). Table 1. Evaluation equation for natural frequency (β = γ = 2.5, c = 1.5). τ1 /τ2

Natural frequency (Hz)

0.1 0.2 0.3 0.4 0.5

f f f f f

= 0.0395×1/τ1 = 0.0665×1/τ1 = 0.0850×1/τ1 = 0.0991×1/τ1 = 0.1121×1/τ1

Table 2. Evaluation equation for natural frequency (β = γ = 1.5, c = 1.5). τ1 /τ2

Natural frequency (Hz)

0.1 0.2 0.3 0.4 0.5

f f f f f

= 0.0454×1/τ1 = 0.0689×1/τ1 = 0.0861×1/τ1 = 0.1012×1/τ1 = 0.1125×1/τ1

1.5 amplitude c = 0.48

1.00

1.0

0.721

Oscillator output

natural frequency = 0.449 Hz

oscillator output

0.50

time (sec) 0.00 0

1

2

3

4

5

6

-0.50

C = 1.0 C = 2.0

0.5

time (sec) 0.0 0

5

10

15

20

25

-0.5 -1.0 -1.5

-1.00

Figure 2. A sample output from the oscillator (τ1 = 0.25, τ2 = 0.50, β = 2.5, γ = 2.5, c = 1.5).

40.0

1/ 1

Figure 4. Output behavior under changing the tonic excitation c (τ1 /τ2 = 1.00/2.00 = 0.5, β = 2.5, γ = 2.5).

280

1.5

τ1=0.1 τ2=1.0

1.4

same peak

t1 / t2 = 0.1(t1= 0.0454 t2 = 0.0454) = ? =1.5 c = 1.5

τ1=0.03125 τ2=0.3125

1.3 output(oscillator) frequency (Hz)

1.0

output

0.5 0.0 0

1

2

3

4

5

6

7

8

9

10

-0.5 -1.0 -1.5

time (sec)

Figure 5.

1.2 1.1

oscillator frequency = 1.000 Hz

1.0 input amplitude 0.2 input amplitude 0.1 input amplitude 0.05

0.9 0.8

input amplitude 0.03 input amplitude 0.02 input amplitude 0.01

0.7

Oscillator output (τ1 /τ2 = 0.1).

0.6 0.4

0.6

0.8

1.0

input frequency

1.4

1.6

Figure 7. Relationship between input and oscillator frequencies (τ1 /τ2 = 0.1).

1.0

output input

A=0.0716

1.2 (Hz)

input/output

0.5

the lower graph that the oscillator locked in the input frequency. Numerical simulations were carried out taking into account various combinations of the parameters of the neural-oscillator model, placing much emphasis on the lock-in phenomenon. Figure 7 shows the relationship between input and output (oscillator) frequencies. It can be seen from this Figure 7 that there is the positive slope in which output oscillator frequency is nearly equal to input frequency, though oscillator frequency is just set up to be 1.000 Hz. These analytical results mean that the neural-oscillator might be one of the useful models to explain the synchronization (the lockin phenomenon) of a fairly high part of all pedestrians being on the bridge.

0.0 0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

4.0

4.5

5.0

4.5

5.0

time (sec)

(a) Input amplitude = 0.0716 1.0

output input

A=0.239

input/output

0.5

0.0 0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

4.0

4

time (sec)

(b) Input amplitude = 0.239

4.1

1.0

output input

A= 0.477

input/output

0.5

0.0 0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

4.0

4.5

5.0

time (sec)

(c) Input amplitude = 0.477

Figure 6. Effect of increasing input signal on oscillator output (τ1 = 0.1, τ 2 = 0.2, β = γ = 2.5, c = 1.0 Cnatural frequency of the oscillator = 1.122 Hz, output frequency = 1.738 Hz).

281

DYNAMIC RESPONSE ANALYSIS BASED ON A NEURAL-OSCILLATOR MODEL Lateral vibration caused by a neural-oscillator model with the natural frequency of 0.970 Hz

A pedestrian bridge with the span of 50 m was selected as an analytical model. The 1st symmetric lateral natural frequency of this bridge model, which is assumed to be simply supported in lateral direction, is just to be 1.000 Hz. Structural dimensions are shown in Table 3. The dynamic response analysis was carried out taking into account the above-mentioned neural-oscillator model (β = γ = 1.5, c = 1.5) with the natural frequency of 0.970 Hz (τ1 = 0.046804, τ2 = 0.46804). It is appended that the dynamic lateral loading caused by the neural-oscillator in these analyses is almost equal to the force induced by 30 walking persons. Figure 8 shows the oscillator frequency which is obtained by zero-crossing method for the time history oscillator

Table 3.

Structural dimensions of the pedestrian bridge model.

Model

Length

Weight

Young’s modulus of elasticity

geometrical moment of inertia

1st natural frequency

MODEL-100

50 m

14.7 kN/m

20.58×107 kN/m2

0.01844 m4

1.000 Hz

0.03

1.040

original oscillator frequency = 0.970Hz

1.020

max = 0.014416 m

0.02

Dynamic response (m)

output(oscillator) frequency (Hz)

Input frequency = 0.970 Hz (Constant)

bridge frequency = 1.000Hz

1.000

0.980

0.01

time (sec) 0.00 0

5

10

15

20

25

30

35

40

-0.01 -0.02 -0.03

0.960 0

5

10

15

20

25

30

35

40

time (sec)

Figure 10. Time history response at the center of the span obtained by the usual dynamic response analysis (Bridge frequency = 1.000 Hz, Input frequency = 0.970 Hz).

Figure 8. Output (oscillator) frequency obtained by the dynamic response analysis (Bridge frequency = 1.000 Hz, Original oscillator frequency = 0.970 Hz). Non-dimensional maximum amplitude

1.0

0.03 ymax = 0.02625 m

dynamic response (m)

0.02 0.01 time (sec) 0.00 0

5

10

15

20

25

30

35

40

-0.01 -0.02

0.8

0.6

0.4

0.2

0.0 0.94

-0.03

0.96

0.98

1.00

1.02

1.04

1.06

Original oscillator frequency (Hz)

Figure 9. Time history response at the center of the span obtained by the dynamic response analysis taking into account the neural-oscillator model (Bridge frequency = 1.000 Hz, Original oscillator Frequency = 0.970 Hz).

output. It can be seen from this figure that the oscillator frequency varies gradually, and is entrained to be equal to 1.000 Hz which is the lateral natural frequency of the bridge. Consequently, the dynamic response of the girder at the center of the span increases as shown in Figure 9 when the logarithmic decrement δ is 0.02. For reference, Figure 10 shows the result obtained by the usual dynamic analysis in case that the frequency of the dynamic lateral loading does not vary with the original natural frequency of 0.970 Hz. It is found that the maximum response in this case is about half of the maximum displacement as shown in

Figure 11. Relationship between original oscillator frequency and non-dimensional maximum amplitude.

Figure 9 obtained by the analysis taken into account the neuron-oscillator. 4.2

Lateral vibration caused by a neural-oscillator model with the natural frequency of 0.950∼1.05 Hz

The dynamic response analysis was also carried out taking into account the neural-oscillator model with the natural frequency of 0.950∼1.050 Hz. Figure 11 shows the relationship between original oscillator frequency and non-dimensional maximum amplitude at the middle point of the bridge with span length of

282

0.4

bridge frequency = 1.000 Hz

1.03

input wave frequency = 1.000 Hz initial amplitude A = 0.01 log.decrement d = -0.10

0.3

original oscillator frequency = 0.980 Hz 1.02

input amplitude

output(oscillator)frequency (Hz)

1.04

1.01 1.00 0.99

f15

0.98

20

= 1.0035 Hz

0.2 0.1

time (sec)

0.0 -0.1

01

02

03

0

40

-0.2 -0.3

0.97

-0.4 0.96 0

5

10

15

20

25

30

35

40

Figure 13. Developing input wave according to the increase of time (Input wave = exponential function).

time (sec)

(a) Output frequency (Original oscillator frequency = 0.980 Hz)

1.04 output (oscillator) frequency (Hz)

output(oscillator)frequency (Hz)

1.04

bridge frequency = 1.000 Hz

1.03

original oscillator frequency = 1.000 Hz

1.02 1.01 1.00

f15

20

= 1.0142 Hz

0.99 0.98 0.97

input wave frequency = 1.000 Hz initial amplitude A = 0.01 log.decrement d = -0.10

1.03 1.02 1.01 1.00

original oscillator frequency = 0.970 Hz

0.99 0.98 0.97 0.96 0

0.96 0

5

10

15

20

25

30

35

40

10

20 time (sec)

30

40

time (sec)

Figure 14. Output (oscillator) frequency in case that the input wave is an exponential function.

(b) Output frequency (Original oscillator frequency = 1.000Hz) Figure 12. Output (oscillator) frequency obtained by the dynamic response analysis (Bridge frequency = 1.000 Hz, Original oscillator frequency = 0.980 Hz, 1.000 Hz).

50 m when the logarithmic decrement δ is 0.02. It is appended that the vertical axis in this figure is non-dimensionalized by the maximum displacement obtained by the analysis using the output from the oscillator with natural frequency of 1.000 Hz which is assumed not to be entrained. Based on this figure, it is noticed that the output value in case of 1.000 Hz is smaller than that in case of 0.980 Hz. Therefore, the oscillator frequency is investigated as shown in Figure 12 which is obtained by zero-crossing method for the time history oscillator output. It can be seen from this Figure that the average f15∼20 of frequencies from 15 sec to 20 sec for the oscillator with the original frequency of 0.980 Hz is 1.0035 Hz which is more closer than the bridge frequency of 1.000 Hz compared with the value of 1.0142 Hz which is the average f15∼20 for the oscillator with the original frequency of 1.000 Hz. The amplitude of input wave to the oscillator varies because the walking position of the pedestrian along

283

the footbridge depends on time. Then, the characteristics of the output wave from the oscillator with the original frequency of 0.970 Hz is investigated through the numerical simulations taking into account the developing input wave uf according to the increase of time t as shown in the following equation. uf = Ae−hωt sin ωt  0.01e−(−0.1)×2π×1.0 sin(2π × 1.0)t

(7)

where A is the amplitude, ω is the circular frequency, h is the damping constant and δ is the logarithmic decrement. Figure 13 shows the developing input wave according to the increase of time t when A = 0.01 and δ = −0.1, respectively. Figure 14 also shows the relationship between the output oscillator frequency and the time. It can be seen from this Figure that the oscillator frequency does not lock in the input frequency perfectly and becomes greater than 1.000 Hz which is the bridge frequency in case that the input wave is an exponential function. It is supposed that this phenomenon might be the reason why the maximum bridge displacement obtained by the dynamic

analysis in case of 1.000 Hz is smaller than that in case of 0.980 Hz as the above-mentioned result (see Figure 11). 5

CONCLUDING REMARKS

This paper deals with a neural-oscillator model to grasp some useful information for the human-induced lateral vibration on congested pedestrian bridges. The dynamic response analysis is also carried out taking into account the neural-oscillator model. The results are summarized bellow: 1. The oscillator frequency is proportion to the magnitude of 1/τ1 under the condition that the ratio of τ1 /τ2 is constant. 2. A simplified equation to evaluate the natural frequency of the neural-oscillator is proposed. 3. The value of amplitude/c and the output frequency are not varied although the output amplitude of oscillator increases according to the increase of the tonic excitation c. 4. Based on the results of dynamic response analyses, it is confirmed that the neural-oscillator might be one of the effective models to explain the synchronization (the lock-in phenomenon) of a fairly high part of all pedestrians being on the bridge. 5. The oscillator frequency does not lock in the input frequency perfectly and becomes greater than 1.000 Hz which is the bridge frequency in case that the input wave is an exponential function.

Needless to say, a lot of further study might be necessary to investigate the human-induced lateral vibration on congested pedestrian bridges. It is hoped that this study will provide useful information for bridge engineers in investigating dynamic lateral behavior of pedestrian bridges. REFERENCES Dallard, P., Fitzpatrick, A.J., Flint, A., Bourva, S. Le, Low, A., Ridsdill Smith, R.M. & willford, M. 2001. The London Millennium Footbridge, The structural Engineer, Vol. 79, No. 22, pp. 17–33. Fujino, Y., Pacheco, M.B., Nakamura, S. & Pennung, W. 1993. Synchronization of Human Walking Observed during Lateral Vibration of a Congested Pedestrian Bridge, Earthquake Engineering and Structural Dynamics, Vol. 22, pp.741–758. Matsuoka, K. 1985. Sustained Oscillations Generated by Mutually Inhibiting Neurons with Adaptations, Biological Cybernetics, Vol. 52, pp. 367–376. Williamson, M. 1998. Neural Control of Rhythmic Arm Movements, Neural Networks, Vol. 11, pp.1379–1394. Yoneda, M. & Fukae, M. 2008. A Fundamental Study on Neural-oscillator-based Algorithm for Human-induced Lateral Vibration on Congested Pedestrian Bridges, Journal of Structural Engineering, Vol. 54A, pp. 218–227 (in Japanese).

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Analysis of large dynamic structures in the entertainment industry D.P. Cook & R.T. Robinson Entertainment Engineering and Design, University of Nevada, Las Vegas, USA

ABSTRACT: The Gantry Lift system at Cirque du Soleil’s show Kà is the largest hydraulically powered lifting system in North America with over 7000 HP of lifting power and 140 tons of moving mass. The Hammerhead components of the Gantry Lift, which are responsible for guiding the vertical motion of the Gantry, are experiencing deflection due to the static and dynamic loading conditions on the system. This deflection interferes with the smooth operation of the lifting system and potentially could cause catastrophic system failure. Analysis of these components was performed and the calculated deflections compared well with measurements made on the actual hammerhead units. Further modeling was conducted to determine placement of stiffening plates which would reduce or eliminate the deflection of the hammerheads, and these results will be the basis for proposed modifications to the as-built structure.

1

INTRODUCTION

The modern live entertainment industry poses a number of fairly unique structural and mechanical engineering problems in the design and use of theatrical stage components. Traditional structural engineering focuses mainly on large static structures, e.g. bridges, highways, building, etc. These structures are design predicated on the assumption that they will not move, at least not significantly. Traditional machine design which investigates dynamic devices whose purpose involves motion, are usually on a human scale, i.e. 1 meter in length scale and 100’s of kilograms of moving mass or smaller. Dynamic stage components in many current theatrical productions are on the order of 10’s of meters in length and 100’s of metric tons in moving mass. These components thus call for the combined expertise of both the structural and mechanical engineer in order for a design to be successful. A prime example of these types of stage components can be seen at the Cirque du Soleil show Kà, which is located at the MGM Grand Hotel in Las Vegas, NV. Cirque du Soleil is a Montreal, Canada based entertainment company. Founded in 1986 by a group of street performing acrobats, clowns, and gymnasts, Cirque has grown into a multi-billion dollar enterprise employing over 3000 people globally. Based in the traveling circus paradigm, the original Cirque productions were touring shows. The first permanent Cirque production, Mystère, was created in 1993 at the Treasure Island Hotel and Casino in Las Vegas. The success of Mystère led Cirque to subsequently create a number of other permanent shows. Kà was the 4th such show, opening in 2005. The Kà production cost approximately $200 million to develop (Huntington 2005).

285

Figure 1. Photograph of the Sand Cliff Deck in the Kà theater at the MGM Grand Hotel.

The Kà theater has no stage in the traditional sense. Instead the artists perform on an interacting system of seven moving lifts and decks. The largest of these is the Sand Cliff Deck and Gantry Lift system, shown in Figure 1. The Sand Cliff Deck measures 25 by 50 by 6 and weighs roughly 80,000 pounds. It is mounted on a large wrist or knuckle-like mechanism that allows it to rotate (slew) continuously at speeds up to 2 revolutions per minute, and tilt from a horizontal position to 110 degrees, i.e. 20 degrees past vertical. The Sand Cliff Deck and knuckle are connected through a cantilevered outrigger arm and torsion tube to the Gantry Lift, which provides for the vertical motion of the system. The Gantry Lift is a hydraulic lifting system which is capable of moving the Sand Cliff Deck surface 70 vertically at a rate of 2 feet per second. The

2. the automation subsystem which allows for precise and repeatable control of the positioning of the SCD surface.

Figure 2.

Architectural rendering of the Kà Gantry Lift.

total, combined moving mass of this system is roughly 280,000 pounds (Tomlinson 2007). The work presented in this paper focuses on the Hammerhead components of the Sand Cliff Deck and Gantry Lift system. The Hammerheads, shown in Figure 2, guide the vertical motion of the Gantry Lift. During operation these components were exhibiting elastic flexure that interfered with the smooth operation of the lift. Structural analysis of the system was performed using standard FEA tools to determine potential modifications which would alleviate these operational issues. This paper will present a complete description of the design and operation of the Sand Cliff and Gantry Lift system, and a discussion of the operational issues that arose during the first three years of the shows tenure at the MGM Grand. Details on the structural analysis methods employed will be given, and the results from the calculations will be presented. Finally, comparisons will be made with the actually flexure seen during operation.

2

DESCRIPTION OF THE GANTRY LIFT

The Gantry Lift system consists of 6 major structural subcomponents: 1. 2. 3. 4. 5. 6.

the Sand Cliff Deck (SCD), the knuckle, the outrigger arm, the torsion tube, the Hammerheads, and the support columns.

In addition, there are two non-structural subsystems that are part of the Gantry Lift: 1. the hydraulic power subsystem responsible for the lifting and tilting action of the SCD, and

Figure 2 shows an architectural rendering of the Gantry Lift showing all structural components except the SCD. For much of the discussion that takes place in this paper, a traditional theatrical coordinate system will be used, so it is important to quickly review these. ‘‘Stage right’’ and ‘‘stage left’’ refer to the directions as seen from the artist’s perspective while looking out at the audience. Conversely, ‘‘house right’’ and ‘‘house left’’ refer to the directions as seen from the audience perspective. ‘‘Upstage’’ refers the portion of the stage furthest from the audience, while ‘‘downstage’’ refers to the portion closest to the audience. ‘‘Onstage’’ and ‘‘offstage’’ refer to directions from a specific point. For example, for an artist standing at the very front of the stage, ‘‘onstage’’ would also be ‘‘upstage’’, while ‘‘offstage’’ would also be ‘‘downstage’’. As mentioned earlier, the Sand Cliff Deck is the main performance platform for the Kà production. The face, or performance surface, of the deck measures 25 × 50 . The SCD is 6 thick, constructed from structural steel, and weighs approximately 80,000 lbs. There are three lifts inside the SCD which descend from its face where artists are hidden during portions of the show. There are 80 automated pegs which extend from the face of the SCD. These pegs are used in several climbing scenes and also mimic spears or arrows striking the SCD. The physical mechanisms of these lifts and pegs are contained within the deck, as well as their automation and control hardware. The face of the SCD is covered with a layer of capacitive pads which effectively turn it into a giant touch screen, so the position of each artist can be tracked during the performance (Tomlinson 2007). The bottom surface of the SCD is attached to the knuckle, an articulated joint that provides 2 of the 3 degrees of motion to the SCD. Four hydraulic cylinders are located in the knuckle which tilt the face of the SCD from 0 degrees (horizontal) to 110 degrees, (20 degrees past vertical). Tilting the SCD allows it to be used as a climbing wall and as a projection surface during the show. An electric motor, also located in the knuckle, can provide continuous rotation (slew) at a rate of 2 RPM at any tilt angle (Wendlandt 2006). The SCD and knuckle rest on the down stage end of the outrigger arm, a cantilevered beam structure whose upstage end is attached to the torsion tube. The purpose of the outrigger arm is to provide the necessary clearance for the motion of the SCD. The vertical load of the SCD is transferred through the outrigger arm and torsion tube to the two hammerheads, which guide the vertical motion of the Gantry Lift on the two support columns.

286

The twin support columns are located stage right and stage left. They are constructed of 0.75 steel pipe with a 4 diameter. The bottom end of each column is bolted into the foundation of the theater, 30 below the stage level. The columns extend upwards from the foundation 80 . Located on the tops of the columns are the four hydraulic cylinders that provide the vertical motion of the Gantry Lift, two cylinders per column. The columns are maintained vertical by a steel support frame which also isolates the Gantry Lift seismically from the remainder of the casino structure. Located on the onstage side of each column are the brake rail and roller guide rails. In addition, the two hammerheads are positioned on the onstage side of the support columns. The brake mechanisms and guide rollers are mounted on the hammerheads, and the lower end of the hydraulic piston rods are attached to the top of each hammerhead. The hammerheads measure 30 in the vertical direction, 6 in each of their horizontal dimensions, and are constructed of 0.5 thick, welded A36 steel plate. In summary, the weight of the SCD is transferred through the outrigger arm, torsion tube, hammerheads, and guide rollers, reacting against the faces of the roller guide rails. Under these loading conditions, the hammerhead bodies exhibit sufficient elastic deformation so that 1) the guide rollers are not correctly aligned on the rails, and 2) there is potential for contact between the hammerheads and the brake rail. This contact would cause catastrophic failure of the Gantry Lift system. The casino would see significant loss of revenue, roughly $250 k per show, until the Gantry could be rebuilt.

3

3.1

The commercial FEA package COSMOSWorks was used to perform a linear, static analysis of the hammerheads. The computational domain was meshed using 50,000, 2nd order, tetrahedral elements with an average size of 10 cm. The set of linear equations was solved using the default iterative solver (FFEPlus). The boundary conditions that were applied to the model were: 1) a counter-clockwise, torsional load of 2.2 × 106 ft-lb acting on the cut-plane of the torsion tube, 2) a vertically, downwards force of 100,000 pounds representing the weight of the torsion tube, outrigger, knuckle, and deck acting on the cut-plane of the torsion tube, 3) a bearing load to counteract the weight of the system acting vertically upwards at the piston rod attachment points, 4) non-penetratative restraints applied at the points where the hammerheads are in contact with the guide rails. The model material was A36 structural steel with the properties given in Table 1.

3.2

Geometry of cases considered

Calculations for three cases were made: 1. As-built: This is the current configuration at the Kà theater. The geometry data was taken from drawings of the original design and measurements performed by the author(s). 2. Capped pigeon holes: In this case, the as-built geometry was modified by the addition of 0.5 inch (12.7 mm) thick, horizontally-oriented steel plates in the openings at the top and bottom of the hammerheads, referred to as ‘‘pigeon holes’’ by Kà staff. Two plates were located in each pigeon hole at distances 0.5 and 24 inches from the ends of the hammerhead. 3. Capped pigeon holes and stair steps: In this case, in addition to the plates in the pigeon holes, verticallyoriented, 0.5 inch thick steel plates were located in the openings in the off-stage legs of the hammerheads, referred to as ‘‘stair steps’’ by Kà staff. These plates were located 0.5 inches in from the off-stage surface.

ANALYSIS OF THE HAMMERHEADS

Structural analysis of the load and deformation profile in the hammerheads was performed using the commercial finite element analysis (FEA) package COSMOSWorks. Three cases were considered: 1) the current configuration of the hammerheads, i.e. the asbuilt case, 2) the effects of adding horizontally oriented stiffening plates in the ‘‘pigeon holes’’ at the top and bottom of the hammerhead, and 3) the effects of adding vertically orientated stiffening plates in the ‘‘stair steps’’ in the offstage legs of the hammerheads. The calculations were made considering the stage left hammerhead. The coordinate system used defined the positive x-direction as downstage, the positive y-direction as vertically up, and the positive z-direction as onstage. Thus when looking at the onstage face of the hammerhead, the torsional load is acting in a clockwise direction. For these calculations, it is assumed that the stage right hammerhead would be the mirror image of the stage left hammerhead.

Model methods and parameters

Table 1.

Materials properties of A36 steel.

Property

British

SI

Young’s modulus Poisson’s ratio Shear modulus Yield strength Ultimate tensile strength

2.9 × 107 psi 0.26 1.1502 × 107 psi 36,259 psi 58,015 psi

200 GPa 0.26 79.3 GPa 400 MPa 400 MPa

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4

RESULTS AND DISCUSSION

The von Mises-Hencky yield criterion states that an elastic material begins to fail when the von Mises Stress reaches the yield stress of the material (von Mises 1913). Figure 3 shows the calculated Von Mises stress distribution on three of the vertical edges of the hammerhead for the as-built case. The peak value for the von Mises stress occurs on the down-stage, on-stage edge at approximately the midheight of the hammerhead, near the center of the torsion tube. The peak value approaches the yield stress of A36 but does

not in fact exceed it. The calculated results thus indicate that the hammerheads are not in danger of failure from exceeding the yield strength of the material. More problematic than failure of the hammerheads due to yielding, is the chance of their failure due to excessive deflection of the steel plates that make

Figure 5. Calculated z-direction deflection at top and bottom, off-stage edges of hammerhead for as-built case.

Figure 6. Calculated z-direction deflection at bottom, offstage edge of hammerhead for all cases.

Figure 3. Calculated von Mises stress distribution on vertical edges of hammerhead.

Figure 4. Photograph of bottom of hammerhead showing body plate deflection.

Figure 7. Calculated y-direction deflection at vertical down-stage, on-stage edge of hammerhead for all cases.

288

up the body of the hammerheads. Figure 4 shows a photograph of the bottom of the stage right hammerhead while in operation. The onstage direction is upwards in this picture, while the offstage direction is downwards. The bottom end of the brake rail is seen in the lower left hand corner of the photograph. In this photograph, the deformation of the body plates is evident, as is the proximity of the offstage, flexing plates to the brake rail. Figure 5 shows a plot of the calculated z-direction deflection in the as-built hammerhead as a function of position along the top and bottom edges of the offstage face of the hammerhead. The actual bottom edge of the hammerhead was seen in Figure 4. Positive values in this figure indicate offstage deflection. Clearly, deflection is much greater at the bottom of the hammerhead than at the top. This is due to the presence of the brake mounting fin at the top of the hammerhead, which effectively acts as a stiffening member. It was due to observations of the difference in deflection between the top and bottom edges of the hammerhead that led to the idea of using horizontal stiffening plates to reduce the overall deflection of the hammerheads. Figures 6 and 7 compare calculated deflection data for all three cases considered in this study. Figure 6 shows a plot of the calculated z-direction deflection along the bottom, offstage edge of the hammerhead, while Figure 7 shows calculated y-direction deflection along the vertical downstage, onstage edge of the hammerhead. In Figure 6, positive values indicate offstage deflection, and in Figure 7, positive values indicate downstage direction. In both of these figures, it is seen that the addition of stiffening plates in the pigeonholes has a dramatic effect on reducing the deflection. However, additional stiffening plates in the stairsteps have little positive effect on reducing deflection. 5

CONCLUSIONS

Structural analysis calculations of the hammerhead components of the Gantry Lift system at the Kà theater

289

were carried out using the commercial FEA software COSMOSWorks. In particular, these calculations consisted of a linear static analysis of the deformation of the hammerheads while under the static load created by the weight of the Sand Cliff Deck. The following recommendations were made based on these calculations: 1. The addition of stiffening plates in the pigeon holes at the top and bottom of the hammerheads is recommended. The calculations carried out in-dicated that these would have significant effects on the reducing the deformation of the Hammerhead body plates. 2. The addition of stiffening plates in the stair-steps is not recommended as the calculations indicated these would have little effect on the deformation of the Hammerhead body plates. 3. Further calculations that investigate the effects of dynamic loading in the Gantry lift system are recommended as a single, large dynamic loading instance could have a significant effect on the structure integrity of the system, and lower magnitude, cyclical loading could lead to fatigue failure. REFERENCES American Institute of Steel Construction (1986). Steel Construction Manual (8 ed.). Huntington, J. (2005, April). How did they do that?, Lighting and Sound, 27–35. Tomlinson, J.T. (2007). Personal communication. Head of Automation, Kà. Von Mises, R. (1913). Mechanik der festen korper im plastisch deformablen zustand. Gottin. Nachr. Math. Phys. I. 582–292. Wendlandt, W. (2006). Personal communication. Asst. Head of Automation, Kà.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Comparison of different standards for progressive collapse evaluation procedures A. Saad, A. Said & Y. Tian University of Nevada, Las Vegas, USA

ABSTRACT: The chain reaction failure of a major portion of a structure that is initiated by the failure of a relatively small portion is referred to as the progressive collapse of a structure. The main approach used for the evaluation of a structure’s vulnerability to progressive collapse is the instantaneous removal of a load bearing element of the structure, followed by studying its effect on other structural elements. An analytical study using a finite element model (FEM) is used in this investigation with the aim of comparing the main two standards that address progressive collapse. A three-dimensional nonlinear model of a concrete generic frame structure is used in the study. The structure is designed according to different seismic zones in order to evaluate the effect of the seismic region on its vulnerability to progressive collapse.

1

INTRODUCTION

Progressive collapse is the spread of an initial local failure from element to element, resulting in the collapse of the entire structure or a large portion of it. Structural Engineers’ attention was drawn to its threat after the partial collapse of Ronan Point apartment building in the UK in 1968 (Nair 2007). Consequently, increasing the overall structural integrity was addressed in different building codes, specifically for the purpose of progressive collapse mitigation. Recent terrorist attacks on the Alfred Murrah building in Oklahoma City and the World Trade Center further highlighted the importance of progressive collapse mitigation. These attacks indicated that most of the casualties are due to building collapses rather than the initial explosion or impact. To reduce the risk of progressive collapse resulting from local failure, precautions should be considered in a design. For this purpose two design methods were defined by Breen (1975) and Ellingwood and Leyendecker (1978). These methods were termed as direct and indirect design methods. The indirect design method is an implicit approach to ensure progressive collapse resistance by providing the minimum levels of strength, continuity and ductility. The direct design method is an explicit consideration of progressive collapse resistance through one of two techniques: (1) the alternative load path method and (2) the specific load resistance method. The first technique provides an alternative load path in case of local failure so that the damage is absorbed and collapse is prevented. The second technique aims at providing resistance to failure (Ellingwood and Leyendecker 1978). Building codes and standards generally address the increase of the

291

overall integrity of structures, which is considered in the indirect design method. The direct design method, particularly the alternative load path method, is used in the design provisions for progressive collapse analyses, design, and evaluation. The main objective of this investigation is to perform a comparison between two of the main guidelines that address progressive collapse namely: GSA Design Guidelines (2003) and the DOD Unified Facility Criteria (2005). This analytical comparison is made at various seismic zones using a three-dimensional FEM of a generic concrete structure using nonlinear static (pushover) analysis. 2

PREVIOUS RESEARCH

Various studies have discussed different issues related to the progressive collapse. Here are some examples of the research done recently. Marjanishvili introduced the different analysis procedures that can be used for the progressive collapse evaluation and design (2004). Bescemi and Marjanishvili discussed the analysis of the column removal scenario as a SDOF system (2005). Sasani and Sagiroglu studied the effect of seismic detailing, design and rehabilitation of structures on progressive collapse resistance (2008). Examinations of different seismic design and strengthening were conducted by Sozen et al. on the Alfred Murrah Building and indicated the positive influences of seismic design on progressive collapse resistance (1998). Different retrofit techniques to mitigate the progressive collapse were also discussed by Crawford 2002; Astaneh 2003; Orton 2007 and others.

3

STANDARDS AND CODES

Different standards and codes address abnormal loads and progressive collapse analysis. The focus on this issue started after the progressive collapse of Ronan Point apartment tower in 1968 (Nair 2007). Although the ASCE/SEI 7-05 includes a definition for the progressive collapse, it does not incorporate any specific steps or design criteria during the design process to prevent or minimize the risk of progressive collapse. Instead, guidelines for the provision of general integrity are included. These guidelines emphasize providing the required integrity to carry loads around the severely damaged walls, trusses, beams, columns and floors. Similarly, the ACI 318-08 includes requirements for structural integrity in the details of reinforcement recommendations. It states that the structural members shall be tied together to improve the overall structural integrity. The intent of this section of the code is to improve the redundancy and ductility in structures so that in the event of damage to a major supporting element or an abnormal loading event, the resulting damage is contained and the structure will have a better chance to maintain an overall stability. The Federal Emergency Management Agency introduced its two publications FEMA 273 and 274 (1997) as guidelines for the seismic rehabilitation of buildings. Despite the fact that progressive collapse analysis was not clearly discussed in the guidelines, comprehensive guidelines were presented for the selection of the analysis procedure, which is generally valuable for use by engineers in estimating structural response.

4 4.1

STANDARDS USED IN THE STUDY General Services Administration (GSA 2003)

The Progressive Collapse Analysis and Design Guidelines introduced by the General Services Administration (GSA 2003) to assist in reducing potential for progressive collapse in new buildings and for evaluating the potential of existing ones. It starts with a process to determine whether a building is exempt from progressive collapse considerations or not based on the type, usage, size and occupation of the structure. This is followed by an evaluation process described for concrete and steel non-exempt structures. The evaluation is done by performing structural analysis for the following the removal of one column or a 30 ft length of bearing wall. The potential of progressive collapse is determined using the acceptance criterion described for each analysis technique. Different analysis techniques are considered in the GSA guidelines including: linear elastic static and dynamic analysis and nonlinear static and dynamic analysis techniques. For each of

these techniques the GSA guidelines mandates loading values and acceptance criteria for evaluation. For static analysis procedures the loading value is taken as: Load = 2(DL + 0.25 LL)

(1)

while for the dynamic analysis procedures the loading value is: Load = DL + 0.25 LL

(2)

where DL and LL are the dead and live loads of the structure, respectively. An amplification factor of 2 is used in the static analysis loading equation to account for the dynamic effect. Evaluation in the linear elastic analysis procedures is based on the demand capacity ratio (DCR), while in the nonlinear analysis procedures it is based on the plastic hinge rotation and displacement ductility ratios. 4.2

Department of Defense (DOD 2005)

The Department of Defense (DOD) introduced the Unified Facility Criteria (2005) to provide the necessary design requirements to reduce the potential of progressive collapse for new and existing DOD facilities. The document provides detailed guidelines for analysis procedures for RC, steel, masonry and wood structures. Detailed steps of the analytical techniques and evaluation criteria are described in the document following the same scenario of the load bearing element removal from the structure in the GSA. Three analysis techniques are presented as the linear static, nonlinear static and the nonlinear dynamic analysis procedures. For linear and nonlinear static analysis procedures, the following amplified load combination is applied to the bays adjacent to the removed element: Load = 2[(0.9 or 1.2)DL + (0.5LL or 0.2S)] + 0.2W (3) While for the nonlinear dynamic analysis procedure, the following load combination is used: Load = (0.9 or 1.2)DL + (0.5LL or 0.2S) + 0.2W (4) where S and W are the snow and wind loads, respectively. For linear static analysis procedures, iterations are requited since elements are removed from the model if their ultimate capacities are exceeded. For the nonlinear analyses, the evaluation is performed based on the stresses and forces in the elements and connections as well as deflection and plastic hinge rotation values which may require additional analysis iterations with new initial.

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5 5.1

DESCRIPTION OF PROTOTYPE BUILDING Geometry

The building used in the analysis is a seven storey typical reinforced concrete generic building (6 stories + roof) with a story height of 12 ft (Figure 1). The reason for choosing this number of stories is to have the earthquake load as the governing lateral load in the design rather than the wind load, so that having the building in various seismic zones would affect the cross sectional capacities and accordingly the geometric nonlinearity definitions. Another reason is to distinguish between the two standards, since the GSA does not account for wind load (eqs. 1 and 2) to the contrary of DOD (eqs. 3 and 4). In plan, the building consists of four 25 ft bays in each direction. The slab thickness was assumed to be 8 in. (per ACI provisions for 2-way slabs) and columns are 24 × 24 in. 5.2 Modeling The slabs were modeled using plate elements except for the two bays adjacent to the removed column where equivalent frame elements were used instead. Beams were modeled using frame elements. The nonlinearity

in beams and slab elements in the two bays where the majority of plastic deformation takes place was defined using the designed cross-sections full momentcurvature relationship. The foundations at the ground floor were modeled as fixed support (restrained in the six degrees of freedom) in all seismic zones. 5.3

Design loads

The building was designed in accordance to the Uniform Building Code (UBC 97) and the ACI 318-02 (2002). In addition to the self-weight, the superimposed dead load plus the distributed load equivalent to partitions load was estimated to be 60 lb/ft2 and a live load to be 50 lb/ft2. A wind speed of 70 mph with ‘‘Exposure B’’ and an importance factor of 1.15 were used in the design. Three different seismic zones (Zero, 2B and 4) were used independently resulting in three different buildings compared in two different standards. The soil profile type was selected to be Sc (Very dense soil or soft rock) for all seismic zones with an importance factor of 1.25. Table 1 shows

25 ft

Interior Frame 25 ft

Exterior Frame

Figure 1. analysis.

Lost Column

3D view of the generic building used in the Figure 2.

Table 1.

Plan of the typical floor.

Longitudinal reinforcement in frames at different seismic zones.

Seismic zone

0

2B

4

Location

Exterior frame

Interior frame

Exterior frame

Interior frame

Exterior frame

Interior frame

Beam size* Top Rft‡ Bottom Rft

24 × 12 in. 3#6 2#6

24 × 12 in. 4#6 2#6

24 × 12in. 4#6 2#6

24 × 12 in. 4#7 2#6

24 × 12 in. 4#7 2#6

24 × 12 in. 4#8 2#6

* Beam dimensions are depth x width. Note: 1 in = 25.4 mm; #6 = D19; #7 = D22 and #8 = D25.

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longitudinal (top and bottom) reinforcement in the interior and exterior frames (refer to Figure 2) connected to the lost column.

6

ANALYSIS PROCEDURES

Three dimensional finite element models for the building’s different design configurations using the SAP 2000 program (2002) were developed to perform the analyses. The Response 2000 program (2000) was used in the sectional analysis for the hinge definition used in the nonlinear input of SAP2000. The nonlinear static analysis procedures were used according to the two standards in the analysis of the buildings with the following assumptions: 1. The building was chosen to be symmetric to avoid complications of asymmetry. 2. Zero initial condition methodology was used in the analysis. 3. Geometric nonlinearity (P-delta plus large deformations) was considered. 4. Strength increase factor of 1.25 was used in the material definition (in accordance with the GSA and the DOD standards). 5. The building frames were considered to be special moment resisting frames in case of seismic zone 4, intermediate moment-resisting frames in case of seismic zone 2B and ordinary moment-resisting frames in case of zero seismic zone. 6. The response of the structure was monitored step by step (not only the final stage) during analysis. 6.1

Nonlinear static (pushover) procedure

The first analysis procedure used in the comparisons is the static nonlinear (pushover) analysis. As shown in Figures 3 and 4, Equations 1 and 3 are used for the load definitions of GSA and DOD, respectively. The loads are applied statically taking geometric and material nonlinearities. The analysis procedure involves increasing the applied load monotonically and incrementally until maximum amplified loads are achieved or collapse occurs (Marjanishvili and Agnew 2006). The displacement at node A (the top point of the failed column, see Figure 1) was monitored in each case and plotted in Figures 5 and 6 with the percentage of loading (100% of loading represents the removed column load in Equations 1 and 3 for GSA and DOD, respectively). 6.2 Analysis of results Through the computer runs, a numerical analysis divergence is noted. This is likely due to the significant stiffness degradation at a large portion of the structural members based on the observation of the inelastic

Figure 3.

Nonlinear static case definition for GSA.

Figure 4.

Nonlinear static case definition for DOD.

deformation in the slab (Refer to Figure 8); it is likely that the structure is at the verge of collapse. Each one of the plots in Figures 5a through 5c represents the pushover curve for one of the six buildings (two standards for three seismic zones). During the incrementally increasing load, each curve encounters three stages before the potential collapse. The first stage (stage 1 in Figure 7) is the elastic stage before the formation of the first hinge; the second stage (stage 2 in Figure 7) is represented by the curve after the straight line where most of the hinges are formed successively, and finally the third stage (stage 3 in Figure 7) represented by the straight part of the curve after the hinge formation stage. The deformation in the third stage is the highest of the three stages as the strain hardening occurs until collapse point. In the comparison of the GSA to the DOD (Figures 5a, b and c), the value of deflection at the collapse point in case of the DOD in slightly higher than those in case of the GSA, even though the same building is used in the analysis in each case. This is attributed to the use of the absolute deflection of node A and not the relative deflection. As a result of including the wind load in

294

20

40

60

80

0

100

20

40

GSA DoD

-2 -3

2B zone 4 zone

-7 -8

% of Load

20

0 -1 -2 -3 -4 -5 -6 -7 -8 -9 -10

40

60

80

Figure 6a. Percentage of load vs. displacement for GSA at various seismic zones. 0

100

Deflection (in)

0

Deflection (in)

100

-9 -10

Figure 5a. Percentage of Load vs. displacement for GSA and DOD at zero seismic zone.

GSA DoD

20

0 -1 -2 -3 -4 -5 -6 -7 -8 -9 -10

40

40

60

100

2B zone 4 zone

80

Figure 6b. Percentage of load vs. displacement for DOD at various seismic zones. 0

100

Deflection (in)

20

80

% of Load

Figure 5b. Percentage of Load vs. displacement for GSA and DOD at 2 B seismic zone.

0

60

Zero zone

% of Load

Deflection (in)

80

Zero zone

-4 -5 -6

% of Load

0 -1 -2 -3 -4 -5 -6 -7 -8 -9 -10

60

0 -1 Deflection (in)

Deflection (in)

0 0 -1 -2 -3 -4 -5 -6 -7 -8 -9 -10

GSA DoD

0 -1 -2 -3 -4 -5 -6 -7 -8 -9 -10

20

40

60

80

100

Collapse Point

Stage 1

Stage 2

Stage 3

% of Load

% of Load

Figure 7.

Load-displacement curves interpretation.

Figure 5c. Percentage of Load vs. displacement for GSA and DOD at 4 seismic zone.

the loading equation of the DOD standard, the overall deflection of this side of the building is higher than the other side (Leeward side has more deflection than the windward side) and accordingly higher than the GSA standard. With reference to Figures 6a and 6b representing the three seismic levels in each standard, the curves are closer to each other in case of the DOD standard than those in the case of GSA. The reason for that is the application of higher load intensity in the DOD case than in case of the GSA. Consequently,

295

Figure 8. Deformed shape and hinges formed in beams and slab strips at collapse.

the percentage of load at failure in the case of DOD varies from 74.02% to 77.6% while in the case of GSA varies from 86.4% to 94.4%. 7

CONCLUSIONS

A study was performed to compare the two main standards for evaluating structures’ vulnerability to progressive collapse, namely the GSA and DOD standards. It was concluded that when comparing the two standards used in the study, the percentage of load at which the potential collapse occurs (for nonlinear static analyses) in the DOD procedures are less than the percentage of load at which the collapse occurs in the GSA procedures. This is due to the higher applied load intensity in the DOD than that of the GSA. This conclusion is very important for the fact that at a certain point a structure may have a huge difference in its design or evaluation results if the two standards are used. One of the main reasons resulting in such a difference is the application of the wind load in the DOD, even though the wind speed used in the analyses of the study is a basic wind speed with a moderate exposure. Accordingly, the loading equations used in the GSA standard should be reconsidered. ACKNOWLEDGMENT This study is based on work supported by the Institute for Security Studies (ISS) at University of Nevada, Las Vegas. The authors greatly appreciate this support. REFERENCES ACI Committee 318, ‘‘Building Code Requirement for Structural Concrete and Commentary (ACI 318-08),’’ American Concrete Institute, Farmington Hills, MI, 2008. ACI Committee 318, ‘‘Building Code Requirement for Structural Concrete and Commentary (ACI 318-02),’’ American Concrete Institute, Farmington Hills, MI, 2002. ASCE/SEI 7, ‘‘Minimum Design Loads for Buildings and Other Structures,’’ Structural Engineering Institute, American Society of Civil Engineers, Reston, VA, 2005. Bentz E.C., Collins M.P. Response-2000. Software Program for Load-Deformation Response of Reinforced Concrete Section, , 2000. Breen, J.E., (1975). ‘‘Research Workshop on Progressive Collapse of Building Structures Held at the University of Texas at Austin,’’ National Bureau of Standards, Washington, DC, 1975. Buscemi, N., and Marjanishvili, S., (2005). ‘‘SDOF Model for Progressive Collapse Analysis,’’ Proceedings of the 2005 Structures Congress and the 2005 Forensic Engineering Symposium April 20–24, New York, New York. Corley, G.W., (2004). ‘‘Lessons Learned on Improving Resistance of Buildings to Terrorist Attacks.’’ Journal of Performance of Constructed Facilities, American Society of Civil Engineers (ASCE), V. 18, No. 2, pp. 68–78.

Crawford, J.E., (2002). ‘‘Retrofit Measures to Mitigate Progressive Collapse,’’ NIST/NIBS, Multihazard Mitigation Council National Workshop on Prevention of Progressive Collapse, Chicago, IL, July 2002. Ellingwood, B., and Leyendecker, E. V., (1978). ‘‘Approaches for Design Against Progressive Collapse,’’ Journal of the Structural Division, ASCE, V. 104, No. ST3, pp. 413–423. Federal Emergency Management Agency, ‘‘NEHRP Guidelines for the Seismic Rehabilitation of Buildings,’’ FEMA 273, Washington D.C., October 1997. Federal Emergency Management Agency, ‘‘NEHRP Commentary on Guidelines for the Seismic Rehabilitation of Buildings,’’ FEMA 274, Washington D.C., 1997. Hayes, J.R., Woodson, S.C., Pekelnicky, R.G.; Poland, C.D.; Corley, W. G.; and Sozen, M., ‘‘Can Strengthening for Earthquake Improve Blast and Progressive Collapse Resistance?’’ Journal of Structural Engineering, ASCE, V. 131, No. 8, 2005, pp. 1157–1177. International Conference of Building Officials ICBO (1997). 1997 Uniform Building Code. Volume 2. International Conference of Building Officials, Whittier, California. Marjanshvili, S. and Agnew, E., (2006). ‘‘Comparison of Various Procedure for Progressive Collapse Analysis,’’ Journal of performance of constructed facilities, American Society of Civil Engineers (ASCE), V. 20, No. 4, pp. 365–374. Marjanshvili, S.M., (2004). ‘‘Progressive Analysis Procedure for Progressive Collapse,’’ Journal of performance of constructed facilities, American Society of Civil Engineers (ASCE), V. 18, No. 2, pp.79–85. Nair, R.S., (2007). ‘‘Progressive Collapse Basics,’’ North American Steel Construction Conference, March 24–27, Long Beach, CA (April 2nd, 2008) National Institute of Standard and Technology (NIST). (2007). ‘‘Best Practices for Reducing the Potential for Progressive Collapse in Buildings,’’ NISTIR 7396, February 2007. Orton, S.L., (2007). ‘‘Development of a CFRP System to Provide Continuity in Existing Reinforced Concrete Buildings Vulnerable to Progressive Collapse,’’ University of Texas at Austin, August 2007, 363 pp. SAP2000 Version 8, Analysis Reference Manual. Computers and Structures, Inc. Berkeley, California, July 2002. Sasani, M. and Sagiroglu, S., (2008). ‘‘Progressive Collapse of Reinforced Concrete Structures: A Multihazard Perspective,’’ ACI Structural Journal, V. 105, No. 1, pp. 96–103. Sozen, M.A., Thornton, C.H., Corley, W.G., and Mlakar, P.F., (1998). ‘‘The Oklahoma City Bombing: Structure and Mechanisms of the Murrah Building,’’ Journal of Performance of Constructed Facilities, American Society of Civil Engineers (ASCE), V. 12, No. 3, pp. 120–136. U.S. Department of Defense (DOD). (2005). ‘‘Design of Buildings to Resist Progressive Collapse’’. UFC 4-02303, Unified Facility Criteria, Washington, D.C. U.S. General Services Administration (GSA). (2003). ‘‘Progressive Collapse Analysis and Design Guidelines for New Federal Office Buildings and Major Modernization Projects,’’ Washington, D.C.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Correlation between minimum building strength and the response modification factor L.G. Daza University of Puerto Rico, Río Piedras Campus, Río Piedras, Puerto Rico

ABSTRACT: The response modification factor plays an important role in the process of estimating the seismic forces of buildings. The dynamic response of buildings beyond the elastic range activates at least three mechanisms that reduce elastic forces into inelastic loads: ductility (μ), over-strength () and redundancy (ρ). This study correlates the R-factor and the minimum required strength of the building (C ) based on these mechanisms and the pushover analysis of the building. The results of this study conclude that building codes do not address rationally the estimation of the R-factor, because they assign a particular value based on the structural system and ‘‘experience’’. This study brings a procedure to estimate the R-factor for buildings based on parameters such as ductility ratio (μ), maximum lateral resistance (Vu ) and redundancy (ρ). The results show that not necessarily the R-factor assigned by the code is the same value that the structure would develop. INTRODUCTION

RESPONSE MODIFICATION FACTOR DUE TO DUCTILITY (R ) 6 DUCTILITY RATIO ( )

5 4 3 2 1

1.3

1.2

1

1.1

0.8

0.9

0.7

0.6

0.5

0.4

0.3

0.2

0 0.1

When a structure is subjected to a specific onedirection lateral load, the overall behavior of the building can be studied using the load .vs. displacement curve. During the loading process the response modification factor is developed and may be computed if at least three parameters can be evaluated: ductility (μ), over-strength () and redundancy (ρ). The response modification factor will be computed as: R = Rμ · R · Rρ

0

1

PERIOD (seconds)

1.1

Modification factor due to ductility (Rμ )

The modification factor due to ductility (Rμ ) can be calculated from the estimation of the translational ductility ratio. The Rμ factor is defined as the relation between the maximum elastic load (Vue ) and the maximum inelastic load (Vu ) of the same structure under inelastic behavior. Newmark & Hall (1982) did one of the most important studies about Response Modification Factor due to ductility. They found that Rμ is sensitive to the natural period of the structure and there are five period ranges where different values of Rμ can be found. Figure 1 shows the Rμ –μ–T for several ductility ratios and periods. Equations 1–5 serve to compute the ductility modification factor (Rμ ) for any natural period of the structure. Periods ≤ 0.03 sec: Rμ = 1.0

Figure 1. Rμ –T –μ curves (Newmark & Hall).

Periods 0.03 < T < 0.12 sec: √  (T − 0.03) · (2μ − 1) − 1 Rμ = 1 + 0.09 Periods 0.12 ≤ T ≤ 0.5 sec:  Rμ = (2μ − 1)

(2)

(3)

Periods 0.5 < T < 1.0 sec:     Rμ = (2μ − 1) + 2 (T − 0.5) · μ − (2μ − 1) (4) Periods T ≥ 1.0 sec:

(1)

297

Rμ = μ

(5)

1.2

Modification factor due to over-strength (R )

The actual building strength is greater than the base shear calculated during the design process. Certainly, the design methods, minimum code provisions, control of inter-story displacements, and many others, lead to provide structural elements with greater size and reinforcement than those sections designed exclusively for lateral and gravity loads. A valuable tool to determine de excess of strength is the non-linear pushover analysis. The R factor is calculated as the maximum base shear that the building can withstand without any structural failure (Vu ), divided by the design base shear (Vd ) at service level. R =

Vu C · W = Vd Vd

(6)

(7)

2. Structures with periods (0.03 < T < 0.12) seconds:

C = Rρ ·

0.16 · Ca · I √  0.09 + (T − 0.03) · (2μ − 1) − 1



(8) 3. Structures with periods (0.12 ≤ T ≤ 0.5) seconds:   1.79 · Ca · I C = Rρ · √ (9) (2μ − 1) 4. Structures with periods (0.5 < T < 1) seconds:

Where: C = percent of maximum base shear that could resist the structure respect to its total weight (W ). This parameter will play an important role during the verification or estimation of the final Response modification Factor.

1.3

1. Structures with periods T ≤ 0.03 seconds:   Ca · I C = · (Rρ ) 0.56

Modification factor due to redundancy (Rρ )

 C = Rρ ·

√

1.79 · Ca · I   √ 2μ − 1 + 2(T − 0.5) μ − 2μ − 1



(10) 5. Structures with periods (T ≥ 1) second:   1.79 · Ca · I C = Rρ · μ

This parameter is the most difficult factor to be determined. There are few studies dedicated to address it. Some research works indicate the relationship between redundancy with the amount of vertical lines for seismic resistance and the number of plastic hinges required to produce the collapse mechanism (Moses 1974). The computation of the reliability/redundancy factor (ρ) as mandated by UBC-1997 code in section 1630, could lead to proper values for Rρ .

(11)

REQUIRED STRENGTH COEFFICIENT (C ) (Structures w/ period less than 0.03 secs) 3.0

C

= V u /W

2.5 2.0 1.5 1.0 0.5 0.0 0.4

2

0.7

0.9

1.1

1.2

1.4

1.5

1.6

REDUNDANCY FACTOR (R ρ )

REQUIRED STRENGTH COEFFICENT VS. THE RESPONSE MODIFICATION FACTOR Figure 2.

REQUIRED STRENGTH (C ) .vs. DUCTILITY Periods between 0.03 and 0.12 secs (This graph for T=0.10 secs)

2.00 1.75 1.50 C = V u /W

There is a correlation between the maximum base shear capacity that a building can develop and the actual Response Modification Factor. This means that a suggested R-factor assigned to a structural system not necessarily works as assumed if the structure does not have the ability to reach a minimum ultimate base shear (Vu )—obtained from the pushover analysis. The strength coefficient (C ) is defined as the ratio between maximum base shear (Vu ) and the total weight (W ) of the structure. This parameter varies according with natural period. The following equations illustrate the minimum strength coefficient for the complete period ranges. Figures 2 to 6 were made for seismic zone 3, soil profile E and importance factor I = 1.0.

Required strength coefficient (T ≤ 0.03 sec.).

1.25 1.00 0.75 0.50 0.25 0.00 0.43

0.70

0.90

1.07

1.21

1.35

1.48

1.60

REDUNDANCY FACTOR (R )

Figure 3. Required strength coefficient (0.03 ≤ T ≤ 0.12 sec.).

298

depending on the natural period and the minimum strength coefficient (C ) assigned to each family of structures. Example: Lets assume a structure with T = 0.6 seconds; target ductility factors μ = 4, 5

REQUIRED STRENGTH (C ).vs. DUCTILITY (Structures w/ periods between 0.12 and 0.5 secs) 2.00 1.75

1.25 1.00

RESPONSE MODIFICATION FACTOR .vs. PERIOD [C = 0.05; R target = 8.5; R = 1.07]

0.75 0.50

10

0.25

9

0.00

8

0.43

0.70

0.90

1.07

1.21

1.35

1.48

1.60

REDUNDANCY FACTOR (R )

Figure 4. 0.5 sec.).

Required strength coefficient (0.12 ≤ T ≤

RESPONSE MODIFICATION FACTOR

C = Vu/W

1.50

R target= 8.5

7 6 5 4 3 2 1

1.2

1.3

1.1

1.1

1.0

0.8

0.9

0.8

0.7

0.6

0.5

0.5

0.3

0.4

0.1

0.2

0.0

0.2

0

REQUIRED STRENGTH (C ) .vs. DUCTILITY P eriods between 0.5 and 1.0 secs

PERIOD (seconds)

(This graph for T=0.60 secs)

Figure 7.

2.00

Expected R-factor for C = 0.05.

1.75

C =V u/W

1.50

RESPONSE MODIFICATION FACTOR .vs. PERIOD [C = 0.1; R target = 8.5; R = 1.07]

1.25 1.00

18

0.75 16

0.25 0.00 0.43

0.70

0.90

1.07

1.21

1.35

1.48

1.60

REDUNDANCY FACTOR (R )

Required strength coefficient (0.5 ≤ T ≤ 1 sec.).

Figure 5.

RESPONSE MODIFICATION FACTOR

0.50 14 12 10

R target = 8.5

8 6 4 2

Structures w/ periods greater than 1 sec

1.3

1.2

1.1

1.0

0.9

0.8

0.7

0.6

0.5

0.4

0.3

0.2

0.1

0.0

0

REQUIRED STRENGTH (C ) .vs. DUCTILITY

PERIOD (seconds)

2.00 1.75

Figure 8.

C =Vu/W

1.50

Expected R-factor for C = 0.1.

1.25 1.00

RESPONSE MODIFICATION FACTOR .vs. PERIOD [C = 0.2; R target = 8.5; R = 1.07]

0.75 0.50 32

299

12

T target = 8.5 8 4

PERIOD (seconds)

Figure 9.

Expected R-factor for C = 0.20.

1.3

1.2

1.1

1.1

1.0

0 0.9

The following examples about the computed R-values have been done for buildings in seismic zone 3, soil type SE , importance factor I = 1, following the UBC1997 code. Each Figure shows different behavior

16

0.8

EXAMPLES AND APPLICATIONS

20

0.8

3

Required strength coefficient (T > 1 sec).

0.7

Figure 6.

24

0.6

REDUNDANCY FACTOR (R )

28

0.5

1.60

0.5

1.48

0.4

1.35

0.3

1.21

0.2

1.07

0.2

0.90

0.1

0.70

0.0

0.00 0.43

RESPONSE MODIFICATION FACTOR

0.25

R-factors for a structure with target R-factors of 4.5, 7.0 and 8.5 (similar values as suggested by UBC-97 for Shear Wall Systems, steel Eccentrically Braced Frames (EBF) and Special Moment Resisting Frames (SMRF), respectively. Notice that in each case the ductility factor (μ) is kept constant but the strength coefficient (C ) varies from 0.1 up to 0.5.

Table 1. Computed R-factor for several strength coefficients (C ) and target R-factors (Target ductility ratio μ = 3). Strength coefficient (C ) C C C C C C

= 0.10 = 0.15 = 0.20 = 0.30 = 0.40∗ = 0.50

Target R = 4.5

Target R = 7.0

Target R = 8.5

1.23 1.84 2.46 3.69 4.51 6.14

1.91 2.87 3.82 5.73 7.64 9.56

2.32 3.48 4.64 6.96 9.28 11.60

4

PROCEDURE TO ESTIMATE THE R-FACTOR

Table 3. Computed R-factor for several strength coefficients (C ) and target R-factors (Ductility ratio μ = 5).

Step 1: During the design process: a) select of the structural system and preliminary size of members and reinforcement; b) assume a ‘‘target’’ translational ductility ratio (μ); c) select the ‘‘target’’ Response Modification Factor (R). Initially, assume the R-factor suggested by current code; d) selection of the strength coefficient (C ) that the structure should develop. Step 2: Define local seismic parameters: a) Importance factor (I), b) Seismic coefficients (Ca , Cv ), c) Calculation of building redundancy (Rρ ), d) Computation of Seismic base shear (Vd ) and e) Natural period (T). Step 3: Verify the preliminary design. Make adjustments to the structural design and calculate of the capacity curve by means of the non-linear pushover analysis. Step 4: Compute Rμ = Rρ = R as described in previous sections. If the structure has a computed R-factor greater than the ‘‘target R-factor’’, the proposed structural design may be considered as adequate. Otherwise, the structure needs improvements, such as greater values for the ductility ratio (μ), minimum strength coefficient (C ) and redundancy factor (Rρ ).

Strength coefficient (C )

5

∗

Note: The ‘‘computed’’ and ‘‘target’’ R-factors have approximately the same values when C = 0.4 and μ = 3. Table 2. Computed R-factor for several strength coefficients (C ) and target R-factors (Target ductility ratio μ = 4). Strength coefficient (C ) C C C C C C

= 0.10 = 0.15 = 0.20 = 0.30∗ = 0.40 = 0.50

Target R = 4.5

Target R = 7.0

Target R = 8.5

1.50 2.25 3.00 4.50 6.00 7.50

2.33 3.50 4.67 7.0 9.33 11.67

2.83 4.25 5.67 8.50 11.33 14.17

∗ Note: The ‘‘computed’’ and ‘‘target’’ R-factors have the same values when C = 0.3 and μ = 4.

C C C C C C

= 0.10 = 0.15 = 0.20 = 0.30 = 0.40 = 0.50

Target R = 4.5

Target R = 7.0

Target R = 8.5

1.75 2.62 3.50 5.25 6.99 8.74

2.72 4.08 5.44 8.16 10.88 13.60

3.30 4.95 6.61 9.91 13.21 16.51

∗ Note: The ‘‘computed’’ and ‘‘target’’ R-factors have approximately the same values when C is between 0.2 and 0.3 for a ductility ratio μ = 5.

and 6, and Redundancy Factor Rρ = 1.07. The structure may develop actual R-factors not necessarily equal to the assumed R-factor (target value), depending on the required coefficient strength (C ). Figures 7–9 show that structures in the lower period range, calculated assuming constant target R-factor and redundancy factor (Rρ ), are exposed to develop actual Rfactors below the target R-factors. This effect is especially significant for structures with low strength coefficients (C ). Tables 1–3 summarize the calculated

CONCLUSIONS

The response modification factor due to redundancy (Rρ ) may be computed following the same requirements proposed by the current Building Code, specifically following the suggested procedure for the redundancy/reliability factor (ρ). The response modification factor is sensitive to the natural period of the structure (T), the ductility ratio (μ) and the strength coefficient (C ). Structures in the lower period range—with constant target R-factor and redundancy factor (Rρ ), are prone to develop actual R-factors below the target values. This effect is especially significant for structures with low strength coefficients (C ). During the design process not necessarily the R-factor provided by the code (target R-factor) will be the same that the structure will develop. It is a mistake to assume that all structures having the same structural system, will behave equally, and consequently, should have the same Response Modification Factor. The strength coefficient (C ) represents the maximum lateral building resistance compared against

300

the total dead load of the structure. The results of this study indicate that main reason to have differences between the ‘‘target’’ and ‘‘actual’’ (computed) R-factors is due to the strength coefficient (C ). Greater C values will guarantee computed R-factors beyond the ‘‘target’’ R-factors. A particular R-factor can be reached adjusting the translational ductility ratio (μ) and the strength coefficient (C ). In other words, several (μ, C ) combinations can lead to the same Response Modification Factor. Building codes, in the near future, should revise the procedure to compute the response modification factor, using a more rational approach. More research is required to address some parameters such as the modification factor due to redundancy (Rρ ).

301

REFERENCES Applied Technology Council ATC-19. 1995. Structural Response Modification Factors. Redwood City, California. Moses, F. 1974. Reliability of Structural Systems. Journal of Structural Division, ASCE,100 (ST9). Newmark, N. & Hall, W.J. 1982. EarthquakeSpectra and Design. EERI Monograph Series, EERI, Oakland. UBC-97. 1997. International Conference of Building Officials, Whittier, California.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Effect of infill walls in structural response of RC buildings I. Idrizi & N. Idrizi Design and Construction Company, ‘‘Arting5’’- Shkup, Macedonia

Z. Idrizi University of Prishtina, Civil Engineering Faculty, Prishtina, Kosova

S. Idrizi Fund of National and Regional Roads, Shkup, Macedonia

I. Idrizi Technical Collaborator in the Federal Polytechnic Faculty of Lausanne, Switzerland

ABSTRACT: This paper elaborates the study of the effect of infill walls in the seismic response of reinforced concrete (RC) buildings. For this study it has been used a typical 9 story RC building, considering three different lateral bearing systems, (frame type—FT, wall type—WT, and a dual type–DT). Further on, this study includes the consideration of two types of infill wall systems (IW1, IW2) that are more vastly applied in engineering practice, and one inventive wall system (IW3). The main analysis technique used for this study was pushover analysis with and without P-Delta effect. This paper aims at demonstrating, qualitatively and quantitatively, the effect of different infill walls in different lateral bearing structural systems. 1

INTRODUCTION

Infill panel elements, as part of the building RC structures, play a very important role on the seismic performance of the building structure. If rightfully accounted for its mechanical properties in the analysis and design process, as well as in their implementation in practice, they can markedly increase the total structural strength and energy dissipation abilities of buildings. In this paper, it is elaborated the effect of infill panel walls in the seismic response of a 9 storey RC building, considering 3 different lateral bearing systems (frame type—FT, wall type—WT and a dual type—DT) (Figure 1 a–c). In this paper, for simplicity of analysis results, a characteristic 2D frame of the building model will be thoroughly examined. For each of the three lateral bearing systems, it is investigated the effect of 3 different types of infill panels, namely IW1 (solid clay brick infill panel), IW2 (calcium silicate block panel), and IW3 (inventive infill panel) (Figure 1 a3–c3). Finally, there are 12 generated frame models, i.e. 4 models (Bare, IW1, IW2 and IW3) for each of the three different lateral bearing systems.

303

Figure 1.

2

Geometry of 2D frame models.

GENERATION OF MATHEMATICAL MODELS

The generation of the 12 mathematical models and their analysis, is performed using CSI SAP2000

computer program. As it is seen in Figure 1, the frame models represent a 9 story building frame with a height of 26.4 m and a width of 15.8 m. The bearing elements are of Reinforced concrete (RC) material. The ideal compressive strength of the concrete material is taken 27 Mpa (4000 kip), while ideal yielding tensile strength of rebar material (steel) is adopted 400 Mpa and its ultimate tensile strength 600 Mpa. Frame models are loaded in each story with a distributed loading along the beam elements with 15 kN/m intensity. In this loading are included the super-imposed dead loads, live loads and the weight of the infill elements. Each story has load intensity of 23.25 t, or about 210 t totally for all of the stories. The self weight of the structure is automatically generated by SAP2000 program using the assigned self weight of the elements properties. The foundation of the frames is assumed fully rigid; hence the columns at the base are assigned fixed joint restrains in all of their six degrees of freedom. Considering that, in practice, the beam elements are monolithically joined with slab RC elements, the axial rigidity of beams is very high and this effect is considered by applying ‘‘Rod’’ joint constraints for all the beam joints of a story. The model frames are consisted of different bearing elements, namely columns, beams, structural walls and infill walls. In Figure 2 are shown the characteristic crosssection dimensions of columns, beams and RC structural walls for the first story. Because the scope of this study is to show comparisons between model types, the bearing element proportions and their reinforcement is kept unchanged throughout different frame types, thus design optimization for proper element proportioning and detailing is not considered in this study. The rebar confinement for all the elements is adopted similarly. The assumed plastic hinges of columns and beams are confined with Ø8/10 cm while their assumed elastic zones with Ø8/20 cm.

Figure 3. walls.

In order to account for the nonlinear properties of bearing elements, SAP2000 automatically generates the column and beam hinge nonlinear properties in compliance to ‘‘FEMA356’’ and ‘‘ACI 318-02’’, and, in addition, the effect of confinement in these elements is considered using Section-design (SD) elements, where fiber analysis of element cross-sections is considered using ‘‘confined-reinforcement Mander model’’. Unfortunately, due to the great variability of wall section types, their automatic generation of nonlinear hinges using axial-moment interaction diagrams is of a more complex nature and their implementation in SAP2000 program can be done indirectly through use of column elements with SD (sectiondesign) cross-section properties, or by application of nonlinear joints. For this study, the definition of nonlinear properties for structural walls is done by use of traditional concepts of equilibrium and strain compatibility, consistent with the plane section hypothesis (Paulay & Pristley 1992). These concepts were applied in this study for modeling the nonlinear flexural behavior of structural wall elements along the height of the model frames. In account to the actual load intensities in the wall elements, obtained from simple linear static analysis, and considering the confinement effect in walls, in Figure 3 are shown the generated nonlinear models for nonlinear flexural behavior of walls for all model frames in consideration. The generated nonlinear (axial force-strain) hysteretic envelopes (Figure 3), are assigned to nonlinear link elements with multi-linear plasticity behavior, which in SAP2000 program are used as contact elements between wall elements.

3

Figure 2. Proportions and reinforcement schemes for bearing element cross-sections.

Nonlinear hysteretic envelope curves for RC

NONLINEARITY CONSIDERATIONS OF INFILL WALLS

The inclusion of infill panel walls in the seismic response of the buildings is usually done by applying

304

and the generated nonlinearity characteristics of frame bearing elements (columns, beams, structural walls and infill walls). In the following will be shown the analytical results obtained, for all of the 12 generated model frames, from Modal, response spectra and nonlinear static pushover analysis. These results will serve to estimate the influence of the three particular types of infill walls on the overall structural response due to seismic lateral forces. 4.1

Figure 4. Mathematical models (shear force-lateral displacement) for IW1, IW2 and IW3 wall panels.

the concept of equivalent diagonal compression strut model (Figure 4), which replaces the infill wall panels with two equivalent diagonal compression struts. Based on this concept, there are well established mathematical formulations under which is made possible to describe the nonlinear behavior of the infill wall panels due to lateral forces. The same models are then easily implemented as an integral part of the overall building mathematical models. This well established approach has been used for the purpose of this study, for which are generated 3 different mathematical models of infill wall panels. The first model type ‘‘IW1’’ (for all of the three structural systems FT, WT and DT) considers the effect of infill walls consisted of ‘‘solid clay brick’’ material. The second model ‘‘IW2’’ represents the effect of calcium silicate blocks, and the third type ‘‘IW3’’ considers an infill wall consisted of inventive panel construction having different nonlinear behavior from the first two infill models. The evaluated axial compressive strength for IW1 panel is obtained 44 kg/cm2 , for IW2 is 22.33 kg/cm2 and for IW3 is 19 kg/cm2 . The generated mathematical models, describing the ‘‘shear force-displacement’’ behavior of infill panels of all the three types of infill walls are presented in Figure 4. 4

Modal analysis

In Figure 6 are shown the first three modes of vibration for all the 12 building models. It can be seen that the FT models are about twice more flexible than the WT and DT models, while WT and DT models have similar flexibility between each other. Another parameter worth observing is how the infill panel stiffness affects the flexibility of the three structural types FT, WT and DT. In reference to Figure 4 and Figure 5, can be observed how the ‘‘stronger’’ IW1 (in comparison to IW2 and IW3) wall panel increases the structural stiffness (reduces the vibration period) of FT, WT and DT more than the ‘‘weaker’’ IW2 and IW3.

Figure 5. frames.

First three periods of vibrations for all model

STRUCTURAL ANALYSIS OF BUILDING MODELS

So far, have been presented the general assumptions made on the modeling phase of the model frames,

305

Figure 6. Pushover curves for FT, WT and DT (‘Bare’ type only), with and without consideration of P-D effect.

4.2

Pushover analysis & response spectra analysis

The focus of this study is based on nonlinear static pushover analysis, with and without consideration of the P-D effect. The results obtained from these analyses (performed for each of the 12 generated frame models) are presented in the following figures.

Figure 7. Pushover curves for all FT, WT, and DT model frames (including three different infill wall panels), with and without consideration of P-D effect.

In Figure 6, are shown the pushover curves of FT, WT and DT ‘Bare’ structural systems, with and without consideration of P-delta effect. In reference to the effect of P-Delta one can observe a reduction of ductility level on FT-Bare model, while on DT-Bare and on WT-Bare models it is seen a reduction in ‘both’ ductility and strength capacity. The maximal shear strength capacity for ‘FT-Bare’ type is 50 t, for ‘DT-Bare’ is 100 t and for ‘WT-Bare’ is 220 t. Figure 7 is consisted of three graphs which altogether present the generated pushover curves for all of the 12 generated frame models. First graph of Figure 7 shows the pushover curves for Frame models (FT-Bare, FT-IW1, FT-IW2 and FT-IW3). Similarly, second graph is showing the generated pushover curves for WT frame models and the third graph shows the results for DT frame models. According to these graphs of Figure 7, it is obvious that the inclusion of infill wall panels in the overall lateral resisting mechanism of structural elements will generally result in increasing of lateral stiffness and strength for the structure. The lateral strength and stiffness increase in model frames, due to infill wall panels, is higher in FT model frames and DT models while it is lower in WT models. Obviously, the ratio of lateral strength increase goes in proportion to the amount of ‘infill wall panel’s area’ participating in model frames. Moreover, considering nonlinear models of infill wall panels in Figure 4 and comparing the pushover curves of FT and DT model frames of Figure 7, one can observe that the gain in lateral strength capacity in DT model frames is twice higher than in FT model frames. Additionally, the inclusion of infill wall panels in DT model frames shows increase in global ductility level as well, as seen in the lowest (third) graph of Figure 7. These positive effects are shown more clearly in Figures 8, 9 and 10. Figure 8 presents the displacements, relative story drifts and shear forces acting along the stories of FT model frames. The same results for WT model frames are shown in Figure 9, while for DT model frames are given in Figure 10. These obtained results are referring to both pushover and response spectrum analysis. In reference to Figure 8, it can be observed how the presence of infill wall panels in FT model frames tends to concentrate the lateral resisting mechanism at the lower stories of the model. This effect is very dangerous and an insufficient lateral resistance capacity at these story levels would result in activation of a soft-story mechanism (a very dangerous failure mechanism which represents a severe damage and collapse of building structures). On the other hand, in Figure 9 and especially in Figure 10, it is seen how this dangerous failure mechanism is avoided by the presence of RC structural walls and, moreover, the lateral resistance capacity of infill wall panels is more successfully engaged along the entire height of model frames.

306

Figure 8. Displacements, relative story drifts and shear forces along story heights for FT model frames.

307

Figure 9. Displacements, relative story drifts and shear forces along story heights for WT model frames.

5

CONCLUSIONS

Finally, the obtained results, as presented in previous figures, suggest that the infill walls, when properly implemented in buildings, tend to improve the structural resistance from seismic forces. The improvements on seismic resistance depend not only on the lateral resistance of infill walls and their level of participation in the building, but also on the structural configuration in elevation and the type of the lateral resistance structural system. In addition, the presence of RC structural walls in buildings prevents the occurrence of soft story-sway mechanism, and successfully engages the lateral resistance capacity of infill wall panels along the entire height of the building. At this point, in reference to Figures 7, 8, 9, and 10, it is very important to note the remarkable improvements in structural response for DT model frames, where one can observe the effects of added stiffness, strength and ductility along the entire height of the building. Moreover, the gains in lateral strength, by inclusion of infill walls in model frames, are twice larger in DT model frames in comparison to FT and WT model frames. In general, by inclusion of infill wall panels in the overall building model, it is observed an increase in stiffness and lateral bearing strength of the building, while the ductility capacity of the structure mainly depends on the ductility of the primary lateral resisting elements (columns, beams and RC walls). A specific care must be taken on locations of openings in infill wall panels, especially when these openings are located at the bottom story levels, which would make the building vulnerable to soft story-sway failure under earthquake events. REFERENCES

Figure 10. Displacements, relative story drifts and shear forces along story heights for DT model frames.

CEB Bulletin-236, 1997. Seismic design of RC Structures, Switzerland: Lausanne, International Federation for Structural Concrete (fib). FEMA 356, 2000. Prestandard and commentary for the seismic rehabilitation of buildings, Federal emergency management agency. FIB/CEB-FIP Bulletin 25, 2003. Displacement-based seismic design of Reinforced concrete buildings, Switzerland: Lausanne, International Federation for Structural Concrete (fib). Huebner, K. 1975. The finite element method for engineers, USA, John Wiley & Sons. Madan, A. et al. 1997. Modeling of masonry infill panels for structural analysis, Journal of structural division, ASCE, Vol. 114, No. 8, pp. 1827–1849. Park, T. & Paulay, T. 1975. Reinforced concrete structures, USA. John Wiley & Sons. Paulay, T. & Priestley, M.J.N. 1992. Seismic design of reinforced concrete and masonry buildins, USA: San Diego, John Wiley & Sons. Saneinejad, A. & Hobbs, B. 1995. Inelastic design of infilled frames, Journal of structural engineering, ASCE, 121(4), pp. 634–650.

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Estimation of statically equivalent seismic forces of single layer reticular domes H. Abdolpour Islamic Azad University, Ahar Branch, Ahar, Iran

Z. Zamanzadeh Islamic Azad University, Bostanabd Branch, Bostanabad, Iran

A. Behravesh Islamic Azad University, Mahabad Branch, Mahabad, Iran

ABSTRACT: Structural design of space structures in areas under great seismic activities presents problems on how to analyze the dynamic response under earthquake motions. On contrary few results are available in the case of reticular dome form of space structures to estimate seismic forces. In the present paper, dynamic responses on single layer reticular domes subjected to horizontal earthquake motion by using scaled near field earthquake accelerations, different masses, densities and different members of domes have been investigated. The seismic forces on different levels of domes have been obtained. In order to understand the dynamic response of domes under earthquake motions, graphs have been obtained. Finally, relations have been achieved for estimating base shear force and seismic forces in different levels of dome. By using these relations, seismic forces could be estimated accurately without any need to time-consuming dynamic analysis and complicated mathematical calculations. 1

INTRODUCTION

Structural design of single layer reticular domes in areas under great seismic activities presents problems on how to analyze the dynamic responses under earthquake motions and how to estimate the magnitudes of statically equivalent seismic forces to be applied. A sizable amount of work has been dedicated to assess the seismic behaviour of space structures. Issues such as study of the dynamic strength of reticulated domes under severe earthquake loading (Kato et al. 1997), horizontal earthquake loading and linear/nonlinear seismic behaviour of double layer barrel vaults (Sadegi 2004) and relations for estimation of seismic forces of such structures have been achieved. Furthermore wind-induced behavior and its load estimation of a single layer lattice dome with a long span (Ltd 2001) have been thoroughly investigated. In the case of a seismic design of ordinary high rise buildings, we may use several codes. In the most cases, the static horizontal seismic forces can be estimated by design rules without any complicated analyzes. On the contrary, few results are available in case of reticular domes to estimate the seismic forces. The present paper discusses initially the linear vibration characteristics of single layer reticular domes subjected to horizontal earthquake motions. The characteristics are followed

309

by a simplified method for estimation of statically equivalent seismic forces. Discussions go on to the validity of the estimation method by comparing the responses based on a more precise dynamic analysis. The paper finally proposes graphs to understand the dynamic response of domes under earthquake motions. Therefore, the geometrical modeling of structure has been performed using the written software by the researchers and coordinates of joints of dome have been obtained. The form is a spherical surface with a base diameter of 60 m, and each dome is assumed to be pin-supported at its base. As for the rise of the domes, the half open angle φ is varied from 60 to 120 degrees by 15 degrees step to investigate its effects to the response.

2 2.1

STRUCTURAL MODELS FOR RETICULAR DOMES AND MODELING FOR MEMBERS Configuration of domes

The geometrical modeling of structures has been performed using the written software by the researchers and coordinates of joints of dome have been obtained. The configurations of single layer reticular domes studied here are shown in figure 1. The form is a

Table 1.

Member size for domes.

P–d × t

A (Cm2 )

I (Cm4 )

P-114.3*3.5 P-165.2*3.5 P-216.3*5.8

12.18 22.72 38.36

187 734 2130

Table 2.

2.2

Material constants.

Y (Kgf/cm2 )

E (Kgf/cm2 )

2.1*105

2400

Modeling of members and analytical method

The members in the domes are assumed to be rigidly connected at the nodes. As for the members in the domes, one common size selected from Table 1 is applied to every member within the same dome. The symbol P-d×t means that the member is a circular tube with external diameter d and wall thickness t in mm unit. The material constants are shown in Table 2.

3 3.1

Figure 1. Configuration for reticular domes in terms of half open angle φ.

spherical surface with a base diameter of 60 m, and each dome is assumed to be pin-supported at its base. Previous studies show that the responses are very different for low and high rises (Kato et al. 1997). As for the rise of domes, the half open angle φ is varied from 60 to 120 degrees, step by 15 degrees to investigate its effects to the responses. Four cases depending on the magnitude of the half open angle are studied. Figure 1 shows that the shape and the member arrangement of a rigidly jointed reticular dome simply supported at the boundary nodes. As for the boundary condition, the nodes of the outer ring are pin supported.

EARTHQUAKE MOTIONS AND LOADING Earthquake motions

The horizontal earthquake ground motion is assumed to act at the base of the dome also the response on the vertical loading may have interesting dynamic response characteristics. Especially in case of extremely low rise domes with a small half open angle less than 30 degrees, vertical earthquake motions might greatly affect the response as found in the studies (Kato et al. 1997). Applied accelerations adopted in this study are three near-field accelerograms: KOBE, TABAS and ERZINJAN. The acceleration response spectra are given in figure 2 using damping factor hd = 0.05 according to the EUROCODE 3 the acceleration needs to be scaled so that acceleration scaled with software written by researchers. Ultimately scaled responses have been obtained as shown in Figures 2a, 2b and 2c which left figures show acceleration response spectra and right figures illustrate scaled one. 3.2

Loading

As for the gravity loads, three cases are assumed; 600 kg, 700 kg and 800 kg for each dome to investigate the effect of self weight. The uniform load is applied vertically at each of the free joints as shown in Figure 3.

310

Response Acceleration (g)

Response Acceleration (g)

0.9 0.85 0.8 0.75 0.7 0.65 0.6 0.55 0.5 0.45 0.4 0.35 0.3 0.25 0.2 0.15 0.1 0.05 0 0

1

2

3

4

Table 3.

1.3 1.2

1 0.9 0.8

Member

0.7 0.6

0.4

114.3 × 3.5 600 700 800

0.3 0.2 0.1 0 0

1

2

3

4

Period (s)

Response Acceleration (g)

Response Acceleration (g)

5 2 1.9 1.8 1.7 1.6 1.5 1.4 1.3 1.2 1.1 1 0.9 0.8 0.7 0.6 0.5 0.4 0.3 0.2 0.1 1

2

3

1.2 1.1 1

0.7

0.5 0.4 0.3 0.2 0.1 0 0

1

3

4

Acceleration response spectra for KOBE.

Response Acceleration (g)

Response Acceleration (g)

2

Period (s)

1.1 1 0.9 0.8 0.7 0.6 0.5 0.4 0.3 0.2 0.1 0

3.6 3.4 3.2 3 2.8 2.6 2.4

3

4

ESTATICALLY EQUAVALENT SEISMIC FORCES FOR HIGH RISE DOMES BASED ON LINEAR RESPONSES

2.2 2 1.8

Vb = CHO × Wt

1.6 1.4 1.2 1

(1)

0.8 0.6 0.4

where Vb = shear; Wt = effective weight and CHO = shear coefficient.

0.2 0

2

Period (s)

Figure 2c.

0.485 0.516 0.546

The present study repeats linear dynamic response analysis for the domes with different rise and members in table 1 depending on the three kinds of mass density. The domes are assumes to be subjected to the aforementioned three horizontal accelerograms as shown in figure 2 where the horizontal accelerations are applied at the base in the x direction. On the base of result, the distribution of storey shear forces occurring in the domes is investigated. Following the seismic formulas which given in different codes such as EUROCODE, the base shear is defined as follows:

0.6

4

1.2

1

0.366 0.390 0.412

0.8

1.3

0

0.290 0.314 0.333

0.9

Period (s)

Figure 2b.

0.221 0.237 0.251

Acceleration response spectra for ERZINJAN.

2.1

0

Joint’s Mass φ = 60 φ = 90 φ = 105 φ = 120

0.5

Period (s)

Figure 2a.

First natural periods for domes.

1.1

0

1

2

3

4

Period (s)

Acceleration response spectra for TABAS.

6

DISTRIBUTION OF HORIZONTAL SESMIC FORCES ON DIFFERENT LEVELS OF DOMES

The horizontal seismic forces on different levels of domes are calculated by Eq. 2      n   Fi =  mj × U¨ j + U¨ g  (2)  j=1  Figure 3.

4

Loading.

NATURAL PERIODS AND VIBRATION MODES

In the case of high rise with φ larger than 60 degrees the first natural vibration mode becomes predominant in the responses (Kato et al. 1997). Examples of the first natural vibration modes of the domes classified as high rise are given in Table 3. The value of T1 has a tendency to become long with increasing rise.

311

where U¨ j = horizontal acceleration at points j on dome surface; mj = the masses for nodal points j and g = gravity acceleration 980 cm/sec2 . Horizontal seismic forces of dome with φ = 60, p − 114.3 × 3.5 and 600 kg under ERZINJAN earthquake accelerations are shown in Table 4. Figures 4 to 6 give the horizontal seismic forces on different levels of domes. The horizontal axis stand for shear forces and vertical axis represent height of domes. The shear forces increase proportional to the height up to 7/8 height and then it decreases. This is caused by the flexibility of the dome with the increasing number of grids.

Table 4. Horizontal seismic forces of dome with φ = 60.

Figure 4.

Story

Vi x (Kgf)

STORY12 STORY11 STORY10 STORY9 STORY8 STORY7 STORY6 STORY5 STORY4 STORY3 STORY2 STORY1 BASE

419.74 3205.59 6280.49 6422.8 11890.13 11372.72 10511.92 9373.51 8665.83 7002.12 7138.59 3933.51 0

Distribution of shear force on dome with φ = 60.

Figure 6. φ = 120.

Distribution of shear force on dome with

Figure 7. φ = 60.

Distribution of β coefficient on dome with

where Fi = horizontal seismic force of I story; Vb = shear base; hi = height of the i-th layer and β = coefficient that depends on dome’s rise. The magnitude of β is inversely computed by Eq. 4 as follows: β=

Figure 5.

7

Distribution of shear force on dome with φ = 90.

ESTIMATION OF SESMIC FORCES IN DIFFERENT LEVELS OF DOMES

Fi Vb ×

(4)

nWi ×hi i=1 Wi ×hi

Figures 7 to 9 give the values of β on different levels of domes in which horizontal axis represent h/H and vertical axis represent β. By curve fitting of various values of β in figures 7 to 9, the β coefficient is estimated in a simple form that is found to be function of height as follows:

The above characteristics of the results lead to the following approximate expression:

β = a × e3( H ) + b × e2( H ) + c × e H + d

Wi × hi Fi = Vb × n ×β j=1 Wi × hi

where a, b, c and d are the coefficients that depend on half open angle φ, h is the height of the i-th layer and H is the total height of dome. Calculated values

(3)

h

312

h

h

(5)

Figure 8. φ = 90.

Figure 9. φ = 105.

Figure 11.

Distribution of b coefficient on domes.

Figure 12.

Distribution of c coefficient on domes.

Figure 13.

Distribution of d on domes.

Distribution of β coefficient on dome with

Distribution of β coefficient on dome with

8 Figure 10.

CONCLUSIONS

The present study investigated the dynamic response of reticular domes subjected to horizontal earthquake motions. Results obtained may be as follows.

Distribution of a coefficient on domes.

of coefficients a, b, c and d are shown in figures 10 to 13 in which horizontal axis represent angles and vertical axis represent different coefficients (a, b, c, d). By curve fitting of various values of a, b, c and d in figures 10 to 13, a, b, c and d coefficients are estimated in a simple form that is found to be functions of angle of dome as follows.

313

1. The statically equivalent seismic forces for different levels of high rise domes are estimated by Eqns (1–5) that could be estimated accurately without any need to time-consuming analysis and complicated mathematical calculations. 2. The horizontal seismic forces increase proportional to the height up to 7/8 height and then it decreases.

This caused by flexibility of the dome with the increase number of grids. 3. In this study, many restraints were assumed on the domes, and accordingly further research will be necessary to investigate the effects of soil structure interactions and supporting lower structures and other related problems. REFERENCES Co-Ltd, Specification for the Design and Construction of Space trusses jgj 7–9. International journal of space structures. Vol. 16, No. 3, 2001. Fan, F., Shen, S.Z. and Parke, G.A.R. 2005. Study of the Dynamic Strength of Reticulated Domes under Sever Earthquake Loading. International journal of space structures. Vol. 20, No. 4, pp. 235–244.

Kato, S., Ueki, T. and Mukaiyama, Y., 1997. Study of Dynamic Collapse of Single Layer Reticular Domes Subjected to Earthquake Motion and the Estimation of Statically Equivalent Seismic Forces. International journal of space structures. Vol. 12, No. 3 pp. 191–203. Sadegi, A., 2004. Horizontal Earthquake Loading and Linear/Nonlinear seismic Behaviour of Double Layer Barrel Vaults. International journal of space structures. Vol. 19 No. 1, pp. 21–37. Uematsu, Y., Kuribara, O., Ymada, M., Sasaki, A. and Hongo, T. 2001. Wind-induced Dynamic Behaviour and its Load Estimation of a Single Layer Latticed Dome with a long Span. Journal of wind engineering and industrial aerodynamics. Vol. 89, No. 14, pp. 1671–1687.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Experimental study of RC slab-CFT column connections under seismic deformations Y. Su Hebei Polytechnic University, Tangshan, China

Y. Tian University of Nevada, Las Vegas, USA

ABSTRACT: A new type of hybrid concrete and steel structural system used for high-rise buildings in seismic regions was proposed. The system consisted of a primary lateral load-resisting system to limit the earthquakeinduced lateral deformations and a secondary floor system employing reinforced concrete (RC) flat-plate floors supported on concrete-filled steel tube (CFT) columns to mainly carry gravity loads. Experimental study was carried out to examine the capability of a type of slab-column connection to sustain seismic lateral deformations. Two isolated slab-column connection specimens were constructed and tested. The specimens were subjected to high levels of simulated gravity loads but demonstrated desirable lateral drift capacity, indicating the potential of the suggested connection detailing to be used in the proposed hybrid structural system. 1 1.1

INSTRUCTIONS Concrete-filled steel tube (CFT) columns and reinforced concrete (RC) flat plates

CFT columns possess distinctive advantages over conventional RC columns and have gained increasing applications in high-rise buildings. The steel tube of a CFT column confines the encased concrete that in return constraints local buckling of the steel tube. In addition, the steel tube functions as formwork when concrete is cast in the column and eliminates the needs of using longitudinal and transverse steel reinforcing bars. The use of CFT columns can significantly reduce the size of columns required to resist high magnitude of axial forces in the lower stories of a high-rise building and thus provides more usable floor space. The behavior of axially and laterally loaded CFT columns has been extensively studied by researchers such as Furlong (1967), Schneider (1998), and Varma et al. (2000). Compared with reinforced concrete columns, CFT columns can exhibit excellent load resistance and ductility when proportioned adequately. More recent studies of CFT columns focused on the development of fully-restrained connections between steel girders and CFT columns arranged as momentresisting frames. A variety of bolted and welded connections for wide-flange girders framing into circular, square, or rectangular CFT columns have been tested by Schneider and Alostaz (1998), Roeder et al. (1999), and Peng et al. (2000). The studies indicated that properly designed and detailed connections could provide

315

excellent hysteretic response and ductility under cyclic lateral loading. The use of CFT columns in momentresisting frames, however, generally involves complex detailing of the connections. A conventional RC flat-plate system consists of RC slab with uniform thickness supported directly on RC columns without using beams or drop panels. Due to the low construction cost associated with simple formwork and simple arrangement of slab flexural reinforcement, flat-plate systems are widely used for residential and office buildings. An additional advantage of flat-plate systems is the reduced story heights that lead to increased number of stories for a given building height. Flat-plate systems, however, are not effective to resist earthquake-induced lateral deformation due to the low rotational stiffness built into the slab-column connections. Consequently, a separate lateral load resisting system such as reinforced concrete shear walls or perimeter moment frames must be used in conjunction with flat-plates to control lateral deformations. Under seismic loading, the lateral story drift causes significantly concentrated bending moment and vertical shear in RC slab in the vicinity of the column. The stress and strain concentration may result in a brittle premature punching shear failure of the slab due to diagonal tension cracking. The lateral deformation capacity of RC slab-column connections under seismic loading has been experimentally investigated by Hawkins et al. (1974), Morrison et al. (1983), Pan and Moehle (1989, 1992), Durrani et al. (1995), and Tian et al. (2008) among many others.

CFT Column RC Flat-Plate

Reinforced Concrete Flat-Plate Floor

Concrete-Filled Steel Tube

Reinforced Concrete Shear Wall

Plan View

Figure 1. Schematic illustration of a hybrid RC flat plateCFT column structural system.

1.2

Hybrid RC slab-CFT column system

The studies of CFT columns and RC flat-plates have been carried out separately in the fields of reinforced concrete and steel structures. Very limited information is available for the performance of a structural system jointly using RC flat-plates and CFT columns. The research underlying the tests presented herein was to develop a new hybrid concrete and steel structural system for high-rise buildings used in moderate-tohigh seismic regions. The new system, as shown in Figure 1, consists of a primary lateral load system to limit the earthquake-induced lateral deformations and a secondary floor system that employs RC flatplates supported on square or circular CFT columns to mainly carry gravity loads. Such a hybrid system takes full advantage of the key mechanical, economical, and architectural benefits offered by CFT columns and RC flat-plates. The system also has the potential to substantially improve the poor behavior of conventional RC flat-plate structures subjected to earthquake loading.

2 2.1

deformation capacity of interior RC slab-CFT column connections of the hybrid structural system subjected to combined gravity and lateral loading. For a conventional RC slab-column connection, the complex state of stress of RC slab surrounding the column results from the combination of several internal forces. Under lateral loading, both vertical shear and bending moments about the two orthogonal principle axes exist in the RC slab. Punching failure of slab is generally assumed to be initiated at the back face of the column relative to the lateral loading direction due to either high stress or large inelastic deformation of the concrete slab. The key concept for the connections of RC slab and CFT column is to reduce the stress concentration caused by negative bending in the plane of lateral loading while maintaining simple connection detailing. This can be achieved by designing a connection where the slab is simply-supported by the CFT column. Following this strategy, a type of interior slabcolumn connection that limits slab bending moment demand and increases the connection shear capacity in the vicinity of CFT column was proposed and tested. As shown in Figure 2, the RC flat plate is supported on a circular steel plate welded to the CFT column. Because the slab flexural reinforcement is not continuous through the circular column, the connection behaves as a pin connection, leading to only limited negative bending moment in the slab when gravity and lateral loads are applied. In addition, although the steel plate supporting the RC slab provides the slab with little extra flexural capacity, the steel plate attracts a significant portion of the shear transferred from the RC slab to the CFT column and thus release the heavy burden of shear stress placed on concrete situated directly above the steel plate. The critical section of RC slab for punching shear is thereby pushed away from the region immediately adjacent to the column. The RC slab outside the steel plate, due to the enlarged area for carrying shear, is expected to fail in punching only after the connection has sustained sufficient lateral deformation.

EXPERIMENTAL PROGRAM RC flat plate-CFT column connection

The connections of RC flat plate and CFT column are critical structural components affecting the performance of the proposed hybrid system under strong ground motions. Although the slab-column framing is taken as the secondary lateral system, the connections must have a lateral deformation capacity greater than that of the primary lateral load system so that premature punching shear failure of RC slabs can be prevented and the gravity load capacity can be maintained. The experimental study focused on the lateral

Circular CFT column

Circular steel plate

Figure 2. column.

316

RC flat-plate

Proposed interior connection of RC slab and CFT

2.2

Test setup, specimens, and measurements

The tests were carried out on large-scale isolated slabcolumn connection subassemblies consisting of a portion of RC slab and a circular CFT column located in the center. Figure 3 shows the test setup used to simulate the gravity and lateral loading effects on the slabcolumn connections. Loading consists of two steps: a vertical loading was applied first and followed by lateral loading. Eight vertical struts were symmetrically installed around the slab center to restrain vertical displacement and to ensure symmetrically distributed stresses in the slab around the CFT under gravity loading. The vertical struts were connected with the slab at one end and with the strong floor at another end using clevises. The vertical struts allowed the slab to displace laterally but provide no rotational restraint at the slab boundaries. It is noted that, although the slab was nominally simply supported on the steel plates, small amount of bending moments about an axis perpendicular to the lateral loading direction still existed and it would be difficult to determine the exact location of the contraflexural points of the continuous slab of the prototype structure. Consequently, the location of the vertical struts did not represent exactly the slab inflection points in the prototype structure. During a test, the vertical load simulating gravity load effects was applied upward through a jack placed underneath the lower column while a servo-controlled hydraulic actuator, as shown in Figure 3, was controlled so that no lateral displacement was induced at the column. The seismic movement was then simulated by applying displacement to desired drift levels at the top column through the actuator when the simulated gravity load was controlled at constant levels. A horizontal strut was used to laterally restrain the lower end of column and partially transfer the horizontal reaction to a rigid support anchored to the

strong floor. Another portion of the horizontal force was transfer to the strong floor by means of the friction force developed at the interface of the jack and the clevis at the lower column. Two specimens, SP1 and SP2, were tested as preliminary study to verify the feasibility of the suggested detailing of slab-column connection and to identify critical variables for further experimental study. In these specimens, the RC slab measured 2100 mm in each plan direction and 150 mm in thickness. The vertical distance between the two clevises connected to the column top and bottom ends defined a 1600 mm effective column height. Because the CFT column was much stiffer than the RC column, the lateral drift determined from the tests can be approximated as the slab-column joint rotation. Cylinder compressive strength of slab concrete was measured at the commencement of testing as 33.7 and 31.9 MPa for Specimen SP1 and SP2, respectively. The slab flexural reinforcement, with a diameter of 12 mm and measured yield strength of 360 MPa, was placed with a uniform spacing of 200 mm for Specimen SP1 and 100 mm for Specimen SP2. The resulting flexural reinforcement ratio (for both top and bottom mats) was 0.42% for Specimen SP1 and 0.84% for Specimen SP2. The specimens had a clear concrete cover of 15 mm for both the top and bottom layers of the flexural reinforcement. Circular steel tubes with an outer diameter of 194 mm and a thickness of 8 mm was used for the CFT columns. The bottom circular steel plate had a diameter of 320 mm, a thickness of 20 mm, and yield strength of 235 MPa. The connection details of the specimens are shown in Figure 4. Applied loads were measured using load cells, displacements using displacement transducers, and steel strains by strain gauges installed at selected locations. 2.3

Specimen SP1—A constant vertical load of Vg = 200 kN was applied and maintained on Specimen SP1. Following the gravity loading, monotonic lateral loading was applied. Figure 5 shows the response of lateral load versus lateral displacement measured at the top column. Lateral drift was defined as the ratio of lateral displacement to the effective column height that was 1600 mm. Based on the test data presented in Figure 5, the initial lateral stiffness of the specimen was determined as 1.44 kN/mm. The specimen survived a lateral drift of 6% without brittle punching shear failure. At this drift level, the secant lateral stiffness was reduced to 34% of the initial value due to slab concrete cracking. Specimen SP2—Figure 6 shows the test of Specimen SP2 subjected to the combined vertical and lateral loading. Three loading stages corresponding to three levels of gravity loads were applied to this specimen. The lateral load versus lateral deformation

Actuator CFT Column Reaction Wall

1600

Slab

Vertical Strut

Strong Floor

Horizontal Strut 700

Figure 3.

Jack

700

Loading history, test results, and discussion

700

Test setup (dimension: mm).

317

194 Specimen SP1

Circular CFT column

ø12@200

ø12@200

Circular steel plate (20 mm thick) 320 (a) 194 Specimen SP2

Circular CFT column

ø12@100

ø12@100

Figure 6. Testing of Specimen SP2 subjected to combined gravity and lateral loading.

Circular steel plate (20 mm thick)

Top Displacement (mm) 320

-64

(b)

-32

0

32

64

40

Figure 4. Reinforcing details: (a) Specimen SP1 and (b) Specimen SP2 (dimension: mm). Lateral Load (KN)

20 Lateral Drift (%) 0

1

2

3

4

5

6

7

60

Lateral Load (kN)

50 40

0

-20

30

Gravity Shear = 200 KN -40

20

-4 10

-2

0

2

4

Lateral Drift (%)

0 0

20

40

60

80

100

120

Lateral Displacement at Top Column (mm)

Figure 5.

Figure 7. Lateral load vs. lateral deformation response of Specimen SP2 (simulated gravity load = 200 kN).

Lateral load versus lateral drift response of SP1.

response is illustrated in Figures 7 to 9. In the first stage, a vertical load was applied and maintained at Vg = 200 kN. Following vertical loading, the top column was cyclically displaced to 0.5%, 1.0%, 1.5%, 2.0%, and 2.5% lateral drift. Three loading cycles were applied at each drift level. At the completion of 2.5% drift, although the RC slab had cracked on the top surface, no punching failure occurred. During the second loading stage, the vertical load was increased to Vg = 240 kN and the specimen was then laterally loaded to produce 1%, 1.5%, 2%, and 3% drift.

The connection survived 3% lateral drift without any shear distress. No further lateral loading was applied to SP2 at this gravity load level. This was because the lateral drift of reinforced concrete shear walls proposed to be used in conjunction with RC slab-CFT column framing would be limited to 3% under strong earthquakes. During the third loading stage, a vertical load of Vg = 270 kN was applied to the specimen. The same lateral loading protocol used in the second stage was attempted. Punching failure of the specimen, as indicated by the sudden drop of lateral load in Figure 9, occurred during the third cycle of 2% drift.

318

Top Displacement (mm) -64

-32

0

32

64

40

Lateral Load (KN)

20

0

-20

Figure 10. Effects of gravity load level on connection lateral drift capacity (Hueste and Wight, 1999).

Gravity Shear = 240 KN -40 -4

-2

0

2

4

Lateral Drift (%)

Figure 8. Lateral load vs. lateral deformation response of Specimen SP2 (simulated gravity load = 240 kN). Top Displacement (mm) -64

-32

0

32

64

40

Lateral Load (KN)

20

0

-20

Gravity Shear = 270 KN -40 -4

-2

0

2

4

Lateral Drift (%)

Figure 9. Lateral load vs. lateral deformation response of Specimen SP2 (simulated gravity load = 270 kN).

3

DISCUSSION OF TEST RESULTS

The preliminary study revealed valuable information about the deformation capacity of the proposed RC slab-CFT column connection. For a conventional RC slab-column connection, its lateral drift capacity is primarily affected by the gravity load level applied on the slab, typically measured as the ratio of gravity shear Vg to nominal two-way shear capacity Vc . ACI 318-08 (2008) defines Vc as  Vc = vc Ac = 0.33 fc Ac (in SI units)

(1)

where fc is concrete compressive strength and Ac is the area of shear critical section assumed to be located at a distance of half effective depth of the slab. The effect of gravity load on connection deformation capacity is shown in Figure 10 (Vc is denoted as Vo ). If Equation (1) is used for the specimens tested in this study, a nominal two-way shear capacity of 369 kN for SP1 and 359 kN for SP2 results. Specimen SP1 was applied a gravity shear of 200 kN, corresponding to a gravity shear ratio of 0.54. According to figure 10, the lateral drift capacity is likely limited to 1.5% for a conventional RC slab-column connection at this level of gravity load. Specimen SP1, however, demonstrated a 6% lateral drift. For specimen SP2, the gravity shear ratio was 0.67 and 0.75 when a vertical load of 240 kN (loading stage 2) and 270 kN (loading stage 3) was applied. At these levels of gravity loading, the drift capacity for a conventional RC slabcolumn connection would be no more than 1.25%. Specimens SP2, however, achieved a much greater drift capacity. Thus, the tests indicated that a properly detailed RC slab-CFT column connection has great potential to avoid premature punching shear failure and achieve a superior lateral deformation capacity. It is noted that, the gravity load ratios adopted in this study for the specimens were much higher than those used in practical situations, which are generally about 0.3 when the gravity load accounts for the total dead load and 25% of the design live load. It can be expected that, at lower levels of gravity loading, the proposed RC slab-CFT column connection would be able to accommodate greater lateral drift than the deformation capacity demonstrated in this study. 4

CONCLUSIONS

This study proposed a type of hybrid concrete and steel structure consisting of a primary lateral loadresisting system and a RC slab-CFT column framing. The performance of slab-column connections, the critical components for the slab-column framing, was

319

preliminarily studied by testing two large-scale isolated connection subassemblies. The tests indicated that the use of pin connection between the RC slab and CFT column enabled the connections to achieve desirable lateral drift capacity. REFERENCES ACI Committee 318, 2008. Building Code Requirements for Structural Concrete (ACI 318-08) and Commentary, American Concrete Institute, Farmington Hills, Michigan. Durrani, A.J., Du, Y. & Luo, Y.H. 1995. Seismic Resistance of Nonductile Slab-Column Connections in Existing FlatSlab Buildings. ACI Structural Journal 92(4): 479–487. Furlong, R.W. 1967. Strength of Steel-Encased Concrete Beam-Columns. J. Struct. Div., ASCE 93(5): 113–124. Hawkins, N.M., Mitchell, D. & Sheu, M.S. 1974. Cyclic Behavior of Six Reinforced Concrete Slab-Column Specimens Transferring Moment and Shear, Progressive Report 1973–1974, NSF Project GI-38717, Section II, Department of Civil Engineering, University of Washington, Seattle. Hueste, M.B. & Wight J.K. 1999. Nonlinear Punching Shear Failure Model for Interior Slab-Column Connections. Journal of Structural Engineering, ASCE 125(9): 997–1008. Morrison, D.G., Hirasawa, A.M. & Sozen, M.A. 1983. Lateral-Load Tests of R/C Slab-Column Connections. Journal Structural Engineering, ASCE 109(11): 2698–2714.

Pan, A. & Moehle, J.P. 1992. An Experimental Study of Slab-Column Connections. ACI Structural Journal 89(6): 626–638. Peng, S.W. Ricles, J.M. & Lu, L.W. 2000. Full-Scale Testing of Seismically Resistant Moment Connections for Concrete Filled Tube Column-to-WF Beam Hybrid Systems. In: Xiao, Y. and Mahin, S.A., editors. Composite and Hybrid Structures. Los Angeles, California: Association for International Cooperation and Research in Steel-Concrete Composite Structures: 591–598. Roeder, C.W., Cameron, B. & Brown, C.B. 1999. Composite Action in Concrete Filled Tubes. Journal of Structural Engineering, ASCE 125(5): 477–484. Schneider, S.P. 1998. Axially Loaded Concrete-Filled Steel Tubes. Journal of Structural Engineering, ASCE 124(10): 1125–1138. Schneider, S.P. & Alostaz, Y.M. 1998. Experimental behavior of connections to concrete-filled steel tubes. Journal of Constructional Steel Research 45(3): 321–352. Tian, Y., Jirsa, J.O., Bayrak, O., Widianto, & Argudo, J.F. 2008. Behavior of Slab-Column Connections of Existing Flat-Plate Structures. ACI Structural Journal 105(5): 561–569. Varma, A.H., Ricles, J.M., Sause, R. & Lu, L.W. 2000. Seismic Behavior of High Strength Square CFT BeamColumns. In: Xiao, Y. and Mahin, S.A., editors. Composite and Hybrid Structures. Los Angeles, California: Association for International Cooperation and Research in Steel-Concrete Composite Structures: 547–556.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Identification of frequency dependency of quality factor in subsurface ground O. Tsujihara Department of Civil Engineering, Wakayama National College of Technology, Gobo city, Wakayama, Japan

ABSTRACT: Quality factor, which is the seismological term and denotes the degree of damping of earthquake motions in the subsurface layers of ground, has not been well understood. In this study, an analytical method is proposed to identify the frequency-dependent quality factor using vertical array records of earthquake ground motions. Comparing the results of identification at the identical site using the records of ground motions obtained in different events of earthquake, the coherency of estimated quality factor is focused on and the reliability of the proposed method is discussed. 1

2

INTRODUCTION

Strong ground motions are largely affected by the amplification effect of subsurface layers of the ground. Therefore, it is very important to estimate dynamic soil properties of subsurface ground in order to predict the characteristics of strong ground motions that influence the behavior of structures based on the ground or lifeline facilities buried underground. Recently, several studies (Ohta 1975; Tsujihara 1996; Sato 1994; Annaka 1994; Yoshida 1995; Nakamura 2002; etc.) have been done on the identification of dynamic soil properties of subsurface ground using vertical array records of ground motions. Among the properties, the damping is known to be difficult in particular to be identified. Shear wave velocity and quality factor are generally identified as the stiffness and damping parameter, respectively, supposing one-dimensional multiple reflection of shear wave in the horizontally laminated soil deposits. The accuracy of identification of shear wave velocity has been improved. But, the improvement of accuracy is not very notable in the identification of quality factor. Lately, assuming the soil deposits as a layer in the identification of quality factor, the detection of the frequency dependency of quality factor has become of major interest. Tsujihara & Sawada (2007) proposed the sweeping method in which the values of quality factor were estimated only at the frequency points where they were sensitive.In this study, the sweeping method is modified so as to obtain the more reliable results of identification. Comparing the results of identification by this method at the identical site using the records of ground motions obtained in different events of earthquake, the coherency of estimated quality factor is focused on and the reliability of the proposed method is discussed.

321

2.1

THEORY AND METHOD Identification of shear wave velocity and quality factor

Horizontally laminated soil deposits are assumed to be excited by vertical incident SH wave. Consider the identification of subsurface ground model as shown in Figure 1, in which H , ρ, V and Q denote the thickness, density, shear wave velocity and quality factor, respectively. Denoting Fourier spectra of the vertical array records at the points p and q (p < q) by Xp ( f ) and Xq ( f ), the amplitude of quasi transfer function between p and q can be obtained by

; sensor locations

1

H1

1

, V1 , Q1

2

H2

2

, V2 , Q2

p

Hp

p

,V p , Q p

q

Hq

q

, Vq , Qq

Zp

q +1 Figure 1. Analytical model of subsurface ground and sensor locations.

Upq ( fj ) = Xp ( fj )/Xq ( fj )

(1)

where fj is the discrete frequency. The identification problem of unknown parameters such as shear wave velocity and quality factor of the layers above the point q can be reduced to the problem of optimization, which is represented by S(α) =

Nf  2  X˜ p ( fj , α) − Xp ( fj ) → min

(2)

j=1

where α denotes the unkown parameters to be identified and X˜ p (fj , α) is obtained by X˜ p ( fj , α) = U˜ pq ( fj , α)Xq ( fj )

(3)

Nf is the total number of discrete frequency points. U˜ p ( fj , α) is the transfer function between the points p and q, which is caculated by the multiple reflection theory (Haskell 1960). The identification problem is schematically shown in Figure 2. The minimization of Equation 2 is carried out by the scheme of MSLP (Modified Successive Linear Programming; Sawada 1992). 2.2

is assumed to be independent of the frequency in this stage. Quality factor at every frequency point is identified in the second stage with the shear wave velocity fixed to the values estimated in the first stage. In the second stage, quality factor at the frequency points where it is not sensitive is swept out in the process of the optimization by the iterative manner, because the residuals of spectra in Equation 2 at these frequency points can not be minimized however drastically quality factor may be modified. In practice, setting upper and lower limits for them, the identification is performed. Quality factor exceeds the limits in the iterations at the frequency points where it is not sensitive. Eventually, only in the significant frequency bands, quality factor remains in between the limits.

Outline of sweeping method

The difficulties in the identification of quality factor have mainly its origin in the sensitivity of it to the transfer function. In the sweeping method, quality factor is swept out of feasible range. The procedure is shown in the followings. Shear wave velocity of each layer and the average value of quality factor through the layers are identified in the first stage by Equation 2. Quality factor

2.3

Modified sweeping method

Nevertheless, quality factor even at the insensitive frequency points is often left in between the limits without being swept out, which makes it difficult to show clear and consistent frequency-dependent quality factor. Then, the third stage is added in the procedure of identification. The values of quality factor are estimated by Equation 4 only at the frequency points where the sensitivity curve of quality factor has peaks. i+2 Q( fi ) =

 j=i−2 c( fj ) · Q ( fj ) i+2 j=i−2 c( fj )

where Q( fi ) is finally estimated values at frequency points fi , i = 1, 2, . . . , n where the sensitivity curve

Xp( f ) ; sensor location

; Density Vi ; S-wave velocity i

~ X p( f )

Qi ; Quality factor

X p( f ) Fourier Spectrum

~ ~ X p ( f ) = U pq ( f ) ⋅ X q ( f ) Fourier Spectrum

~ U pq ( f )

Xq( f )

1

H1

1

,V1 , Q1

2

H2

2

,V2 , Q2

p

Hp

p

,V p , Q p

q

Hq

q

,Vq , Qq

q +1

Observed ground motions Subsurface ground Figure 2.

(4)

Ground model

Schematic diagram of identification problem.

322

60 30 0 30 60 0

APPLICATIONS

20

40

60

10 0

60 0 30 60 0

20

60

(b) G.L.-100 m Figure 3. Acceleration of ground motions recorded at IWTH08 in the event EQ3. Intitial

(5)

1000

2000

0

100 100 150 156 100 103 150 100

Figure 4. Identified shear wave velocity and quality factor in the first stage of identification.

3.2 Profiles of earthquake.

Event no. Date

Latitude Longitude Depth degree degree Magitude km

EQ1 EQ2 EQ3 EQ4 EQ5 EQ6 EQ7 EQ8

38.45 38.15 40.68 38.40 38.89 41.15 41.01 40.15

142.18 142.28 142.60 141.17 142.14 142.28 142.42 143.30

6.1 7.2 5.7 6.2 6.1 5.9 6.2 5.9

Depth (m)

40 100

141.1744 141.7867 140.9500 141.0508 141.9372 141.8267 141.0153 141.0047

15

80

40.2583 40.2658 39.6408 39.3406 39.4706 39.2717 39.1950 38.9661

10

100

IWTH06 IWTH08 IWTH16 IWTH20 IWTH21 IWTH23 IWTH24 IWTH26

5

20

0

Latitude Longitude Depth degree degree m

80

Site code Site name

60

Table 1. Locations of observation sites and depth of lower sensor locations.

Depth (m)

20

where Vp denotes primary wave velocity.

Q

Estimated

3000 0

0

40

ρ = 0.3Vp1/4

40

Time (sec)

S-wave Velocity (m/sec)

2005/12/17 2005/8/16 2005/2/26 2003/7/26 2002/11/3 2002/10/14 2001/8/14 2000/10/3

80

30

Acceleraion(gal)

The network of the digital strong-motion seismographs, so called KiK-net, is deployed at nearly 700 sites in Japan by National Research Institute for Earth Science and Disaster Prevention (NIED). Each site has six channels of strong-motion seismograph. The sensors of 1–3 and 4–6 channels are installed at the bottom of the borehole and on the ground surface, respectively. The sensors of 1 and 4 channels are installed in the North-South direction, 2 and 5 channels in the East-West direction and 3 and 6 channels in the UpDown direction. Borehole tests were carried out at all the stations, and the values of shear wave and primary wave velocity in the layers were estimated. Since the density of soil in the layers is not available, it is approximated by the following equation (Gardner 1974).

Table 2.

100

(a) Ground surface

3.1 About KiK-net

NINOHE-W KUJI-N SHIZUKUISHI HANAMAKI-S YAMADA KAMAISHI KANAGASAKI ICHINOSEKI-E

80

Time (sec)

60

3

Acceleraion(gal)

of quality factor has peaks. c( fj ) and Q  ( fj ) are the values of sensitivity and quality factor estimated in the second stage of identification, respectively. If Q ( fj ) is not in between the upper and lower limits, they are omitted in Equation 4.

40 42 45 12 46 53 43 0

323

Example of identification

The modified sweeping method is applied to the identification at 8 sites in KiK-net as shown in Table 1. All of the sites are in Iwate Prefecture which is located in the northern part in Japan and is the quake-prone area. Many of the KiK-net sites in this prefecture, like in other prefecture, are located in the mountain-ringed region. The selected 8 sites are in or near the plain field. The data of the events of earthquake are shown in Table 2. As an example, the identification at the site IWTH08 in the event EQ3 is described in detail. The records of ground motions at the ground surface

30

100

10

Q

20 60

10

Quality factor 5

10

15

0

0

0

0

20

Senseitivity

Sensitivity

80

Initial

40

Fourier spectrum (gal • sec)

Observed

20

0

5

Frequency (Hz)

10

15

20

Frequency (Hz)

Figure 8. Quality factor estimated in the 2nd stage identification and its sensitivity.

10

30

10 0

Figure 5. Target Fourier spectrum and its estimation calculated with the initial ground model.

Q

20 10

0

0

20

Senseitivity

80

Sensitivity

60

Quality factor

Estimated

40

Fourier spectrum (gal • sec)

Observed

0

0

0

5

10

15

5

20

10

15

20

Frequency (Hz)

Frequency (Hz)

Figure 6. Target Fourier spectrum and its estimation calculated with the ground model identified in the 2nd stage.

Figure 9. Quality factor estimated in the 3rd stage identification and its sensitivity. 100

EQ2

60

EQ3 EQ4

40

EQ6 20

EQ7

0

60

0

5

10

15

20

40

Frequency (Hz)

Figure 10. Estimated values of quality factor at IWTH08 in the third-stage identification using 6 events of earthquake.

0

20

Quality factor

80

10 0

Quality factor

EQ1 80

0

5

10

15

20

Frequency (Hz) Figure 7. fication.

Quality factor estimated in the 2nd stage identi-

and G.L.-100 m are shown in Figures 3 (a) and (b), respectively. They are the transverse components to the epicentral direction. The intervals of strong ground motions are selected, which are to be used in the identification. Shear wave velocity and quality factor, which are estimated in the first stage identification, are shown in Figure 4.

324

The values of shear wave velocity estimated by borehole test are given as the initial values. In the first stage, quality factor is assumed to be constant throughout the frequency band between 0 and 20 Hz. Fourier spectrum of the ground motions at the surface calculated with the initial ground model is shown in Figure 5 together with the target spectrum. The estimated Fourier spectrum calculated with the model which is identified in the second stage is shown in Figure 6. The fitness between the target and estimated spectra gets better in

the second stage identification. The values of quality factor identified in the second stage are shown in Figure 7. Many of the values are as large as 100 which is set as the upper limit or as small as 3 which is set as the lower limit. Figure 8 shows those which are estimated in between the upper and lower limits with the sensitivity of quality factor to the transfer function. Most of the values at the frequency bands, where the sensitivity of quality factor is very small, are not seen in the figure because they exceed the limits. Figure 9 shows quality factor estimated in the third 100

100 80

EQ2

60

EQ3

EQ1 Quality factor

Quality factor

EQ1

EQ6

40

EQ7

80

EQ3

60

EQ5 EQ6

40

EQ7

20

20

0

0 0

5

10

15

20

0

5

10

15

20

Frequency (Hz)

Frequency (Hz)

(a) IWTH06

(e) IWTH21

100

100

EQ2

60

EQ3

EQ1 Quality factor

Quality factor

EQ1 80

EQ4

40

EQ6

20

80

EQ2

60

EQ3 EQ5

40

EQ6

20

EQ7

0

EQ7

0 0

5

10

15

20

0

5

Frequency (Hz)

10

15

20

Frequency (Hz)

(b) IWTH08 100

100

80

EQ2

60

EQ3

EQ1 Quality factor

Quality factor

EQ1

EQ6

40

EQ7

20

80

EQ2

60

EQ7 EQ8

40 20

EQ8

0

0 0

5

10

15

20

0

5

Frequency (Hz)

10

15

20

Frequency (Hz)

(g) IWTH24

(c) IWTH16 100

100

EQ2

60

EQ3

EQ1 Quality factor

Qualityfactor

EQ1 80

EQ5

40

EQ7

20

EQ2

60

EQ3 EQ5

40

EQ6

20

EQ8

0

EQ7

0 0

Figure 11. 8 sites.

80

5

10

15

20

0

5

10

Frequency (Hz)

Frequency (Hz)

(d) IWTH20

(h) IWTH26

15

20

Estimated values in the third-stage identification and approximated frequency-dependency of quality factor at

325

stage identification. In the same way, the identification at IWTH08 is carried out using the records of ground motions obtained in other events, namely EQ 1, 2, 4, 6 and 7. Figure 10 shows the estimated values of quality factor. A fairly good agreement can be seen in the estimated values. The clear trend can also be recognized that up to about 5 Hz quality factor becomes small and then gradually becomes large with the frequency. The approximation of the frequency-dependent quality factor is shown by the solid line in Figure 11 (b). The results of identification at other sites are also shown in Figure 11.Though the trends of quality factor are not same, looking at the estimated values at each site, the variation of the estimated values of quality factor around the approximated solid line is not so large. It is indicated by this results that rather reliable estimates of quality factor can be obtained by the proposed method.

4

CONCLUSIONS

In this study, modified sweeping method is proposed to identify the quality factor of subsurface ground using earthquake ground motions recorded by vertical array of seismographs. The frequency-dependent quality factor is estimated by three stages in the process of identification. Quality factor in the frequency bands where it is not sensitivity is swept out. As a result the values of quality factor at significant frequency points are highlighted. Moreover, the weighted averages are taken at the frequency points where the sensitivity curve of the quality factor has peaks. The proposed method is applied to the identification at 8 sites in KiK-net using the ground motions observed in several different events of earthquake. The major results in this study are as follows. 1. The trends of the frequency dependency of quality factor can be recognized. 2. The similarity of the estimated values of quality factor can be recognized at the identical site even using different events of earthquake.

REFERENCES Annaka, T., Tsuzuki, T., Masuda, T., Shimada, M. & Okadome, K., 1994, Strain dependence of shear modulus and damping factor of soil deposit inferred from the strong motion accelerograms recorded by a vertical array, Proc. of 9th Japan Earthquake Engineering Symposium: 493–498 (in Japanese). Gardner, G.H.F., Garder, L.W. & Gregory, A.R., 1974, Formation velocity and density—The diagnostic basics for stratigraphic traps, Geophysics, 39: 770–780. Haskell, N.A., 1960, Crustal reflection of plane SH wave, Journal of Geophys. Res., 65(12): 4147–4150. Nakamura, S., Sawada, S., Yoshida, N. & Suetomi, I., 2002, Damping Characteristics of surface layer identified by extended Bayesian method, Proc. of 11th Japan Earthquake Engineering Symposium: 211–216 (in Japanese). National Research Institute for Earth Science and Disaster Prevention (NIED), Digital strong-motion seismograph network KiK-net, Ohta, Y., 1975, Application of optimization algorithm to earthquake engineering, Journal of Architectural Institute of Japan, 229: 35–41 (in Japanese). Sato, T., Kawase, H. & Sato, T., 1994, Engineering bedrock waves obtained through the identification analysis based on borehole records and their statistical envelope characteristics, Journal Struct. Constr., Architectural Institute of Japan, 461: 19–28 (in Japanese). Sawada, T., Tsujihara, O., Hiaro, K. & Yamamoto H., 1992, Modification of SLP and its application to identification of shear wave velocity and quality factor of soil, Journal Structural Mechanics and Earthquake Engineering, Japan Society of Civil Engineering, 446(I-19): 205–213 (in Japanese). Tsujihara, O. & Sawada, T., 1996, A localized identification of dynamic soil properties of subsurface layers in ground by vertical array records, Proc. of 11th World Conference on Earthquake Engineering: 1–8 (in CD-ROM). Tsujihara, O. & Sawada, T., 2007, Quality factor identified using Kik-net in Japan, Proc. of Ninth Canadian Conference on Earthquake Engineering, (in printing). Yoshida, I. & Kurita, T., 1995, Back analysis of dynamic soil properties of Port Island with observation data during Hyogo-ken Nanbu earthquake, Soils and Foundations, 43–9: 44–48 (in Japanese).

Accumulation of the analytical results of identification is necessary for the goal to propose the model of the frequency-dependent quality factor.

ACKNOWLEDGMENTS National Research Institute for Earth Science and Disaster Prevention in Japan is greatly appreciated for providing seismic ground motion records of KiK-net. This study was supported by a Grant-in-Aid for Scientific Research under project No. 20560456.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Integrated design and construction to mitigate wind-induced motions of tall buildings K. Moon School of Architecture, Yale University, New Haven, CT, USA

ABSTRACT: The direction of the evolution of tall building structural systems, based on new structural concepts with newly adopted high-strength materials and construction methods, has been towards augmented efficiency. Consequently, tall building structural systems have become much lighter than earlier ones. This direction of the structural evolution toward lightness, however, often causes serious structural motion problems primarily due to wind loads. This paper presents various design and construction strategies to mitigate wind-induced structural motions of tall buildings. The impact of recently-emerging relatively stiff structural systems, such as diagrids, is investigated. Innovative design and construction of damping systems for tall buildings are studied through systems integration. Further, retrospective installation strategies for damping devices are discussed. 1

INTRODUCTION

Buildings, subjected to various loads, move. In tall buildings, lateral movements due to winds are very critical. This paper investigates various integrated design and construction strategies to mitigate windinduced structural motions of tall buildings. Generally, in tall buildings, lateral vibration in the across-wind direction caused by vortex-sheddinginduced lock-in condition results in the most serious motion problem. A stiffer structure reduces the probability of vortex-shedding-induced lock-in condition because as a structure’s fundamental frequency increases, wind velocity that causes the lock-in condition also increases. Since the natural direction of structural evolution towards lightness is not likely to be reversed, higher stiffness should be obtained with a minimum amount of structural material. Achieving higher stiffness in tall buildings is related to the configuration of primary structural systems. For example, recently developed structural systems for tall buildings such as diagrid structures, in general, produce higher lateral stiffness than traditional orthogonal structures. This paper investigates the contribution of stiffer structural systems to controlling the motion of tall buildings. When severe wind-induced vibration problems are expected, installing auxiliary damping devices can be a reliable solution. This paper also investigates innovative damping strategies for tall buildings. A damping system introduced first in this paper is through systems integration between structures and facades, considering the fact that wind loads are initially applied to the building facades and then transmitted to the

327

structures. Among many façade systems available for tall buildings today, double skin façade (DSF) systems are more effective environmental mediators than conventional single skin façade systems, and due to this reason, their use has been increased despite their higher initial construction costs. While many studies have been performed regarding environmental aspects of the DSF system, no research has been done on the structural capability of the DSF system. The potential of the DSF system as a structural motion control device is presented in this paper. Another damping strategy studied is vertically distributed multiple tuned mass dampers (TMDs). TMDs are generally located near the top of tall buildings to reduce wind-induced dynamic responses of tall buildings, because they perform more effectively when they are located where more lateral displacement occurs. However, as a result, very valuable space near the top floor is sacrificed to contain TMDs. This paper studies the potential of vertically distributed multiple small TMDs, compared to the conventional systems composed of one or two large TMDs at the top of the building. Damping devices are often installed retrospectively during building occupancy. For example, the tuned mass dampers in the John Hancock Building in Boston and the viscoelastic dampers in the destroyed World Trade Center in New York were installed during building occupancy. Installing damping devices during building occupancy adversely affects the building use. This paper also presents more innovative design and installation strategies for retrospective application of damping devices for tall buildings.

2 2.1

STIFFER STRUCTURES Tall building structural systems with diagonals

The recognition of structural efficiency of diagonals to carry lateral loads and use of them combined with conventional orthogonal structural members date back to the time of very early tall building developments in the late nineteenth century. The efficiency of diagonals was augmented with the development of tubular structures in the late 1960s. By locating diagonal bracings on the building perimeter instead of confining them within the interior building core, higher lateral stiffness is achieved with less amount of structural material. While the conventional braced tube concept is still employed for today’s tall buildings, diagrids, another way of using diagonals on building perimeters without vertical columns, has emerged as a new design trend for tall buildings. As in the conventional braced tube, lateral stiffness in diagrid structures is also obtained very efficiently by their diagonal members’ axial action (Ali and Moon 2007).

Figure 1.

69˚

63˚

53˚

34˚

82˚

76˚

90˚

60-story diagrids with various angles.

2.50

2.00

meter

1.50 Horiz. Displ. @ Top (m) 1.00

0.50

0.00 34

2.2

Impact of higher lateral stiffness

53

63

69

76

82

90

degree

Higher lateral stiffness of braced tubes or diagrid structures is desirable not only for static loads but also for dynamic loads which generate responses in both the windward and across-wind directions. In general, the lateral motion in the across-wind direction due to vortex shedding is greater than the motion in the windward direction. Stiffer structures have a lower probability of the vortex frequency locking on a modal frequency (i.e., a lock-in condition) because, as a structure’s fundamental frequency increases, the wind velocity required to cause a lock-in condition also increases (Moon et al. 2007). Thus, due to their higher lateral stiffness, braced tubes or diagrid structures are less prone to lock-in conditions than orthogonal structures. In order to estimate the relative stiffness and wind speed causing vortex-shedding induced lockin condition, a 60-story building was designed using diagrids with various diagonal angles ranging from 34 degrees to 82 degrees as well as using a conventional orthogonal moment frames (Fig. 1). The building’s typical plan dimensions are 36 × 36 meters with an 18 × 18-meter gravity core at the center and typical story heights of 3.9 meters. Each structure was designed with almost identical amount of structural steel. The structures are assumed to be in New York and within category III, which implies that there is a substantial hazard to human life in the event of failure. Based on the code, the basic wind speed is 110 mph. Member sizes were generated to satisfy the maximum lateral displacement requirement of a five hundredth of the building height for the structures with diagonal angles of 53, 63, 69 and 76 degrees. The diagrids having angles of 34 and 82 degrees and the orthogonal

Figure 2. Maximum lateral displacements of 60-story diagrids with various angles

Table 1. Fundamental period (T1 ) and lock-in wind speed for the 60-story diagrids with various angles. Diagrid angle (deg)

T1 (sec)

Lock-In wind speed (mph)

34 53 63 69 76 82 90

7.44 4.55 3.86 3.73 3.96 5.18 7.95

54 89 105 109 102 78 51

Notations: fv = VS/D fv: vortex shedding frequency V : the mean wind velocity at the top of the building S: Strouhal number D: across-wind dimension of the building plan

moment frame do not meet the design parameter, but still designed with the same amount of structural steel for the purpose of comparison. Figure 2 shows each structure’s maximum displacement at the top of the structure. Appropriately configured diagrids with near optimal angles (i.e., about 69 degrees in this case) produce stiffer structures than conventional orthogonal structures. If the diagrid angles deviate too much from optimal, its

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stiffness reduces substantially. Table 1 summarizes each structure’s fundamental period and approximate vortex-shedding lock-in wind velocity based on the equation generally used to estimate vortex shedding frequencies (Simiu & Scanlan 1996; Taranath 1998). Stiffer structures require higher wind speed for lock-in condition. For example, while the lock-in wind velocity for the diagrid structure with a diagonal angle of 69 degrees is about 110 mph, that for the orthogonal moment frame is about 50 mph. Thus, as a building f v ( VORTEX SHEDDING FREQUENCY )

0.134 Hz (T=7.4sec)

LOCK - IN

structure becomes stiffer, it has less probability for a lock-in condition. Figure 3 shows lock-in conditions for the diagrids having 34 degrees and 69 degrees and the orthogonal moment frame.

3 3.1

INTEGRATIVE DAMPING SYSTEMS Damping systems for tall buildings

Damping is related to the design of primary structural systems and selected structural materials. However, the inherent damping ratio provided by the primary structure is quite uncertain until the building construction is completed. A more rigorous and reliable increase in damping, to mitigate tall building motion problems, could be achieved by installing auxiliary damping devices. The effect of such damping can be relatively accurately estimated. Thus, when severe wind-induced vibration problems are expected, installing auxiliary damping devices can be a reliable solution.

V 54 MPH

3.2

DIAGRID (34 DEGREES)

f v ( VORTEX SHEDDING FREQUENCY )

0.268 Hz (T=3.7sec)

LOCK - IN

V 109 MPH DIAGRID (69 DEGREES)

f v ( VORTEX SHEDDING FREQUENCY )

0.126 Hz (T=8.0sec)

LOCK - IN

V 51 MPH MOMENT FRAME

Figure 3. Vortex-shedding induced lock-in condition for the 60-story diagrid and moment resisting frame structures.

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Damping through systems integration between structures and facades

Wind loads are initially applied to the building facades and then transmitted to the structures. Considering this fact, a concept to control tall building vibration through design integration between structures and facades is investigated in this section. Double skin façade (DSF) systems, which have substantial cavity space between the inner and outer façade layers, have been obtaining increased interest mainly due to their contribution to energy savings by enhanced performance as environmental mediators. Through the cavity, for example, hot air can be effectively removed in summertime, and also natural ventilation can be introduced even in tall buildings’ higher levels because of the additional exterior skin which acts as a wind buffer. These functions of the DSF system reduce energy usage in building operation, resulting in economic benefits potentially in the long run, even though their initial construction cost is higher than conventional single skin façades. Another potential explored here is obtaining structural damping through DSF systems. The concept is designing the connectors between the DSF outer skin and the building’s primary structure to have very low axial stiffness so that the transmissibility of the dynamic wind load can be reduced through them. As a result, the DSF outer skin moves back and forth substantially, but the vibration of the primary structure, which is enclosed by the inner skin and contains occupants within it, is reduced significantly. Dynamic motion control for a tall building is achieved through this mechanism. Figure 4 illustrates the concept and simplified model of the system.

If a 1% primary structure damping ratio is assumed, the maximum dynamic amplification factor of the primary structure subjected to a harmonic load, which represents a vortex-shedding-induced lock-in condition, and without the proposed DSF system is about 50. Assuming the DSF outer skin mass to be 1% of the primary structure mass, the system shown in Figure 4 was designed with various DSF outer skin to primary structure frequency ratios in order to reduce the dynamic amplification factor. The system’s connector damping ratios in this study range from 0 to 40%. Figure 5 shows the reduced dynamic amplification factors when the frequency ratio is 0.5. With this very low stiffness of the DSF connectors, the proposed system substantially reduces the transmissibility of the applied dynamic loads from the DSF outer skin to the primary structure as intended. The maximum dynamic amplification factors of the primary structure in the resonance condition are about 18 and 22 when the DSF connector damping ratios are 20% and 40%

Primary Structure

Proposed DSF Connectors

DSF Outer Skin Vibration due to Dynamic Wind Loads (Dotted Lines)

3.3

DSF Outer Skin Mass Primary Structure Mass k

respectively, less than half of the maximum value of 50 in the case without the proposed DSF system. If the frequency ratio is further reduced to 0.1, the dynamic response of the primary structure is further reduced by more than 90% compared to the conventional case without the proposed DSF system. However, as the frequency ratio decreases, the DSF outer skin’s vibration increases, which may results in serious constructional, visual and psychological concerns. Even though the excessive movement of the DSF outer skin can be reduced by increasing connector damping ratio, it will also reduce the effectiveness of the system because the higher connector damping ratio increases the primary structure’s motion, as can be seen in Figure 5. Other ways to reduce the vibration of the DSF outer skin are to increase its mass ratio or to introduce an active control system. However, there are practical limitations in doing so as well. Further research is required for the practical application of the proposed system.

md

m kd

p cd

c u

u+ud

Figure 4. Concept diagram and simplified model of structural motion control using double skin facades.

Figure 5. Dynamic amplification factor (y-axis) of the primary structure with the proposed DSF system (x-axis: forcing frequency to primary structure frequency ratio).

Vertically distributed multiple tuned mass dampers

TMDs are used to reduce wind-load-induced dynamic response of tall buildings. TMDs are usually located near the top of tall buildings because they work more effectively when they are located where more lateral displacement occurs. However, as a result, very valuable top floor space is sacrificed merely as TMD room. This section investigates the potential of vertically distributed multiple small TMDs, compared to the conventional systems composed of one or two large TMDs. The idea is to install relatively small multiple TMDs on multiple levels. By distributing multiple small TMDs along the building height, valuable space near the top can be saved for more desirable functions which can maximize the advantage of great views. In addition, by distributing multiple small TMDs horizontally on each level as well as vertically along the building height, greater reliability can be obtained in case some of them do not function properly or have some tuning errors. Further, not only the first mode but also higher mode responses can be effectively controlled if necessary. A 240 m tall 60-story structure in New York is modeled as a 60-degree-of-freedom system and comparatively studied with a conventional large TMD located at the top as well as with vertically distributed multiple TMDs. Its plan dimension is 40 m × 40 m, story height is 4 m typical, and mass density is 190 kg/cubic meter. The stiffness is configured for the building to have a fundamental period of about 6.3 seconds. This is achieved using quasi-parabolic stiffness with 2,200,000,000 N/m at the base. A relatively low inherent damping ratio of 0.5% regarding the first mode is assumed for the structure. This low inherent damping ratio results in the need for structural motion control

330

not only for the first mode but also for the second mode response to meet the lateral displacement and acceleration design parameters. A sinusoidal load, having peak value of 20,000 N and a period of about 6.3 seconds, is applied to each node of the 60-degree-of-freedom system model to simulate a vortex-shedding induced lock-in condition. The original structure, which has an assumed inherent first mode damping ratio of 0.5%, does not meet the acceleration requirement of 0.02 g and the maximum displacement parameter, a five hundredth of the building height, usually employed in practice. Studies suggest that 6.3% and 3.2% equivalent damping ratios are required to meet the both displacement and acceleration design parameters for the first and second mode resonance conditions respectively. In conventional TMD design, a large single TMD with a modal mass ratio of 2.5% is located at the top of the structure and tuned to the first mode. By installing this TMD at the 60th node, both maximum displacement and acceleration meet the design requirements. However, when the TMD is tuned for the first mode and the building’s second mode is primarily excited, the structure does not meet the acceleration criteria. TMDs are now distributed from node 60 to node 31 for the fundamental mode control (Fig. 6). In order to meet the design parameters, this configuration requires a total 67% more mass compared with the conventional system that has a single TMD at the 60th floor. However, compared with the conventional system, only 5.6% TMD mass is required for each level with this distribution. Furthermore, the total required TMD mass for each level represents multiple small TMDs manufactured possibly as packages for easier installation. If the TMD mass for each level is distributed to 50 small TMDs, for example, each TMD mass is only about 0.1% of the conventional huge single TMD located near the top of the building. Small TMDs can also be distributed according to the mode shapes for the second mode control. An example distribution is shown in Figure 6. Figures 7 shows time history of the nodal and damper displacement at node 60, and Figure 8 shows

the maximum inter-nodal displacement of TMDs installed from node 60 to node 31 and tuned to the first mode in the first mode resonance condition. 3.4

Retrospective application of damping devices

Damping devices are often installed retrospectively during building occupancy as are the cases with the tuned mass dampers in the John Hancock Building in Boston and the viscoelastic dampers in the destroyed World Trade Center in New York. TMDs, designed in the previous section, can be more easily installed by making them very small. Especially in the retrospective installation case during building occupancy, installing small TMDs is much more desirable than conventional large TMD installation. In fact, a distributed small TMD may be

Figure 7. Time history of the nodal and damper displacement at node 60 in the vertically distributed multiple TMD system design.

60th FL.

TMDS TUNED TO THE 2ND MODE 50th FL.

40th FL.

TMDS TUNED TO THE 1ST MODE

30th FL.

TMDS TUNED TO THE 2ND MODE 20th FL.

10th FL.

-1.0

-0.5

0

0.5

1.0

Figure 6. Example of vertically distributed multiple TMDs based on mode shapes.

331

Figure 8. Profile of the maximum inter-nodal displacement of TMDs in the vertically distributed multiple TMD system design.

4

Figure 9. Profile of the maximum inter-nodal displacement of TMDs with increased damping ratio to limit the TMD motion.

constructed as a package and placed at any desired location in this case. When making a small package type TMD, the motion of the TMD mass inside the package, especially installed near the top of the structure, could act as a limiting factor of design. One way to reduce the TMD motion is to increase the TMD damping ratio over the optimal value. In the vertically distributed multiple TMD study presented in the previous section, when the first mode of the primary structure is excited, the motions of the TMDs tuned to the first mode range from ±0.4 meters at the mid-height to ±0.75 meters at the top of the building, and the TMD damping ratios range from 10.8% to 11.5% for optimal performance. Suppose the motion of the TMDs is desired to be limited to a maximum of ±0.5 meters in order to reduce the required space for TMD package placements. This motion limit can be achieved by increasing the TMD damping ratios ranging from 15.5% at the mid-height to 29.5% at the top, which is much higher than the previous case. Figure 9 shows the maximum inter-nodal displacement of TMDs installed from node 60 to node 31 with these increased damping ratios. In this case, however, the system does not produce optimal performance.

CONCLUSIONS

This paper presented various design and construction strategies to mitigate wind-induced structural motions of tall buildings. Relatively stiff structural systems, such as diagrids, reduce the probability of the lockin condition. With careful geometric configuration of the structural members, higher lateral stiffness can be obtained more efficiently. Innovative design and construction strategies for damping systems were also studied through systems integration. Instead of investigating damping solutions only within the structural systems and their components, integrative design approach between the relevant systems may result in more desirable damping solutions. Even though there is always a need for further research for practical application of new technologies and their best performance, a better built environment can be produced through this process.

REFERENCES Ali, M.M. & Moon, K. (2007). Structural Developments in Tall Buildings: Currents Trends and Future Prospects. Architectural Science Review, 50.3, pp. 205–223. Connor, J.J. (2003). Introduction to Structural Motion Control. New York: Prentice Hall. Moon, K., Connor, J.J. & Fernandez, J.E. (2007). Diagrid Structural Systems for Tall Buildings: Characteristics and Methodology for Preliminary Design, The Structural Design of Tall and Special Buildings, 16.2, pp. 205–230. Simiu, E. & Scanlan, R.H. (1996). Wind Effects on Structures: Fundamentals and Applications to Design. 3rd Edition. New York: Wiley. Taranath, B. (1998). Steel, Concrete & Composite Design of Tall Buildings. New York: McGraw-Hill.

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Mitigation of high acceleration shock waves in hybrid structures S.G. Ladkany & S. Sueki University of Nevada, Las Vegas, NV, USA

ABSTRACT: The speed of a shock wave depends on the properties of the medium in which it travels, however if a material junction exists within a structure, it results in a wave speed mismatch and impedance discontinuity. Part of the shock wave is transmitted and others reflected backward at the junction. The backward reflection of a part of the wave helps to reduce the intensity and magnitude of the peaks of the waves in a structure. Based on this concept, the mitigation of high acceleration shock waves in structures is considered using the wave speed/impedance mismatch at the junctions between structural materials that would exist in the path of the shock wave such as steel, concrete, wood or plastic. The paper presents numerical results from shock mitigation research in materially inhomogeneous and segmented long rods.

1

2

INTRODUCTION

High speed and high acceleration shock waves may cause damage to structures such as undesirable motions, misalignments, joint failure and possibly a destruction of structures. Published research considers attenuation of vibrations using different types through the active, passive and semi-active damping methods. One way of passively mitigating the vibration is the use of differences in material properties. Promising results were reported in the reduction of high acceleration vibration using differences in material stiffness and geometry (Solaroli et al. 2003 and Toso & Baz 2004). Under such conditions, attenuation of shock wave vibrations may be explained based on the wave reflection and transmission at a boundary. The speed of a shock wave depends on the properties of the medium in which it travels, however if a material or structural junction exists within a structure, it results in a wave speed mismatch and impedance discontinuity. The difference in impedance at the junction between two materials results from a mismatch in their wave speed or a sudden change in the junction geometry or both. Part of the shock wave is transmitted and the others reflected backward at the junction. The backward reflection of a part of the wave helps to reduce the intensity and magnitude of the peaks of the waves in a structure. Based on this concept, the mitigation of high acceleration shock waves in structures may be attained. High acceleration shock wave may be caused by collision and bomb blasting among other causes in building structures. In this paper, the combination of common building materials, steel, concrete or wood is considered in order to study the possibl application of the concept in hybrid structural design.

333

REVIEW OF EXPERIMENTAL INVESTIGATIONS

As reported in Sueki et al. (2007b), experimental investigation was conducted by suspending a segmented 1930 mm long rod (Fig. 1) from a steel frame then applying an impact load using a calibrated impact hammer at one end and measuring the accelerations at the other end using a calibrated accelerometer attached to a data acquisition system. The segmented long

(a) (b)

(c)

(d)

(e)

Figure 1. (a) the experimental setup, (b) the experimental segmented rod and (c)–(e) its connections.

rod consists of three parts fastened to each other by 5/16–24 thread. The rod may be made of either one material, such as aluminum only or two materials such as aluminum and nylon. The dimensions of each part are given in Table 1. The results showed that when the segmented aluminum and nylon rod was impacted, the high frequency accelerations were reduced by 30% at the onset of impact, during the initial 2 ms, and by 50% within 8 ms compared to the segmented all aluminum rod (Fig. 2). The frequency domain response analysis obtained from the acceleration results of segmented rods, using Fast Fourier Transform (FFT) showed the suppression of a high range of frequency response spectral peaks, when the segmented aluminum and nylon rod was tested and compared to the segmented all aluminum rod (Fig. 3). Table 1. Dimensions of the experimental segmented rod components. Part 2 Part 3 Part 1 Aluminum Aluminum Nylon Aluminum

Acceleration (m/s2)

Length (mm) 718 Diameter (mm) 22.5

521 22.5

514 692 25.1 22.5

3 3.1

Simulation studies of experiments

In this study, Finite Element Analysis (FEA) is conducted using the computer software program, LS-DYNA (Livermore 2006) with models constructed based on the experimental tests discussed earlier. Schematic diagrams of the experimental models for one and two material cases are shown in Figure 4(a) and (b), respectively. In the analysis, the segmented rods are modeled with 17,952 eight-node solid hexahedron elements. Impact is applied using the experimentally obtained force-versus-time curve. No contact surface is defined between different material connections. All materials are defined as elastic, using the material properties shown in Table 2. The simulation results are compared with the experimental results in the cases of aluminum and nylon rods and FFT analysis is conducted for all cases. 3.2

Shock propagation simulation studies using steel, concrete and wood

Since the simulation results showed analogous trends as the one observed from the experimental results, the

All Al Al and Nylon

10000 8000 6000 4000 2000 0 -2000 -4000 -6000 -8000

FINITE ELEMENT SIMULATION STUDIES

(A) (1) (B)

692 mm 0.002

0.004

0.006

0.008

518 mm

718 mm

(C)

0.01

(2)

(1)

102 mm

1724 mm

Time (s)

Figure 2. Experimental acceleration response of segmented rods made of aluminum only (All Al) and of aluminum and nylon (Al and Nylon).

(2)

22.5 mm

102 mm

(D) (1) 913 mm

All Al Al and Nylon

1000 900 800 700 600 500 400 300 200 100 0

22.5 mm

(1)

(2)

(1)

0

Magnitude

22.5 mm

1928 mm

(2)

(1)

102 mm

913 mm

22.5 mm

Figure 4. Four different configurations used in the simulations. The different numbers in parenthesis represent different materials. Table 2.

0

2000

4000

6000

8000

10000

Frequency (Hz)

Figure 3. Frequency domain response spectra obtained from the experimental acceleration response of segmented rods made aluminum only (All Al) and of aluminum and nylon (Al and Nylon).

334

Material properties.

Material

Density kg/m3

Young’s modulus GPa

Poisson’s ratio

Yield strength MPa

Aluminum Nylon Steel Concrete Wood

2700.0 1120.0 7849.0 2322.7 380.0

70 1.9 200 21.7 10.1

0.33 0.35 0.30 0.20 0.329

250.0 62.4 344.7 21.6 34.7

same experimental model discussed earlier is utilized to evaluate the shock propagation behavior of other material combinations such as steel, concrete and wood. In those studies, the impact force applied in the simulations is identical to the impact force obtained from the experiments. The parametric variations made to the models are the length of each segment in the rod and the material properties of the segments. First, the segmented rod simulations are conducted to evaluate the mitigation of wave propagations in the steel or the concrete structures by adding a different material. Configurations shown in Figure 4(a) and 4(c) are used to run the simulations. If steel is used for material 1, then material 2 is modeled as concrete, and vice versa. The length of material 2 in Figure 4(c) is chosen based on the ratio of a column width to a beam length, generally used in buildings. Next, wood is inserted at the middle of either the steel or concrete rods as shown in Figure 4(d). Generally, a soft material such as rubber can be a good damping material. Therefore, the effect of wood, which is soft compared to other common construction materials, is considered as an acceleration damper. Material properties of wood used in the simulation (Table 2) is selected based on the material properties of western white pine in the fiber directions as reported by Green et al. (1999). The wave propagation in this study is considered only in axial directions. Therefore, wood is modeled as an isotropic material even though wood is an orthotropic material. 3.3

Comparison between the experimental impact loading and blast loading

As reported by Longinow & Mniszewski (1996), the peak blast load caused by the World Trade Center Bomb at a distance of 1.52 m to bulding walls is approximately 16.9 MPa. The applied peak impact force obtained from the experiment (Sueki et al. 2007b) is about 2500 N which corresponds to 6.29 kPa based on the applied impact surface area of 397.6 mm2 .

Figure 5. Experimentally obtained impact curve (Experimental impact) and scaled up impact curve matching the intensity of blast loading (2.69 × Experimental Impact).

335

Therefore, the ratio of the peak pressure caused by the World Trade Center blast to the peak of the pressure caused by load curve (impulse) used in our experimental investigations is 2.69. The ratio was used in this research to augment the peaks of the experimental impulse load curve shown in Figure 5. The two load curves were applied to a model of configuration type b shown in Figure 4. In this model, material 1 is steel and material 2 is concrete.

4 4.1

RESULTS AND DISCUSSION Simulation studies of experiments

As shown in Figure 6, the acceleration response of a segmented rod made of aluminum and nylon has substantially lower accelerations than that of an aluminum only rod. The maximum acceleration of the aluminum rod was 10,667 m/s2 while the maximum acceleration of the aluminum and nylon rod was 6,614 m/s2 . Since no damping effect were considered in the model, the accelerations did not phase out (Fig. 6), as was observed in the experimental results (Fig. 2). However, the maximum accelerations are reduced by approximately 38% which is close to the reduction observed in the experimental results. The frequency response domain analysis obtained from the acceleration responses using FFT (Fig. 7) showed the suppression of the peaks in high frequency range accelerations when a nylon segment is inserted in the middle of the aluminum rod. Computer simulations using LS-DYNA of the segmented rod matched quite closely with the experimental results and its FFT analysis, which gave confidence in the ability of the finite element analysis to accurately reproduce the shock wave propagation and mitigation studies. 4.2

Simulation using steel, concrete and wood

When 102 mm length steel segments are placed at both ends of the concrete as shown in Figure 4(c),

Figure 6. Acceleration response of segmented rods made of aluminum only (All Al) and of aluminum and nylon (Al and Nylon), obtained from FEA without damping.

significant reduction of acceleration responses are observed as shown in Figure 8. However, when the material order is reversed, acceleration increased compared to the rod made of one material (Fig. 9). As shown in Table 3, wave speed of steel is higher than that of concrete. Thus, when a wave travels from a material with higher wave speed to a material with lower wave speed and again goes back to a material with higher wave speed, the mitigation effect is observed. When a wave travels in materials with opposite placement order, no acceleration mitigation effect is observed but on the contrary the accelerations increase. The same phenomenon was reported by Sueki et al. (2007a) using aluminum and titanium carbide. Note that when the accelerations obtained from the rod made of concrete only are compared with the ones obtained from a rod made of steel only, the rod made of steel only shows significantly less than the one made of concrete. However, the importance of this study is to obtain the design which can reduce accelerations compared to an original one material structure.

Figure 7. Frequency domain response spectra obtained from the acceleration responses using FFT for segmented rods made of aluminum (All Al) and of aluminum and nylon (Al and Nylon), obtained from FEA without damping.

Figure 8. Acceleration response of segmented rods made of concrete only (All Concrete) and of steel and concrete (4 Steel-Concrete-4 Steel), obtained from FEA without damping.

Figure 9. Acceleration response of segmented rods made of steel only (All steel) and of concrete and steel (4 ConcreteSteel-4 Concrete), obtained from FEA without damping.

Table 3.

Wave speed.

Material

Wave speed m/s

Aluminum Nylon Steel Concrete Wood

5092 1302 5048 3057 5155

The mitigation effects are also seen in the frequency domain responses obtained from the acceleration response using FFT as shown in Figures 10 and 11. When 102 mm length steel segments are added to the end of the concrete rod, the magnitudes of peaks are significantly reduced and peaks at higher frequencies are suppressed (Fig. 10). On the other hand, when 102 mm length concrete segments are added to the ends of the steel rod, increases in magnitudes are clearly seen in Figure 11. Also, strong peaks are observed in higher frequencies when concrete is added in the middle segment compared to the rod made of steel only (Fig. 11). As shown in Figures 12 and 13, inserting a wood segment in between two concrete or steel segments did not reduce accelerations significantly. Wood has much lower stiffness compared to steel and concrete. However the wave speed of materials is calculated as the square root of Young’s modulus divided by density. Since wood has much lower density, the wave speed of wood is 5155 m/s which is the highest among the considered materials in this study (Table 3). Therefore, inserting wood at the middle of a steel rod does not produce any acceleration mitigation effects. The case of a concrete rod inserting wood in the middle of the rod forces the waves to propagate from materials with lower wave speed to higher speed and back to a lower wave speed, which is found to increase the response as discussed earlier in this paper. The mitigation effect

336

Figure 10. Frequency domain response spectra obtained from the acceleration responses using FFT for segmented rods made of concrete (All Concrete) and of steel and concrete (4 Steel-Concrete-4 Steel), obtained from FEA without damping.

Figure 11. Frequency domain response spectra obtained from the acceleration responses using FFT for segmented rods made of steel (All Steel) and of concrete and steel (4 Concrete-Steel-4 Concrete), obtained from FEA without damping.

Figure 12. Acceleration response of segmented rods made of concrete only (All Concrete) and of concrete and wood (Concrete-4 Wood-Concrete), obtained from FEA without damping.

might be observed when the wood segments are placed at the end of concrete rod. However, such condition represents a wood frame with concrete beam or slabs, which is not structurally practical.

337

Figure 13. Acceleration response of segmented rods made of steel only (All Steel) and of steel and wood (Steel-4 WoodSteel), obtained from FEA without damping.

Figure 14. Acceleration response of a segmented rod made of steel and concrete using the experimental impulse load and an impulse whose magnitude is scaled by a factor of 2.69, obtained from FEA without damping.

4.3

Comparison between impact loading and blast loading

Figure 14 shows the acceleration responses obtained from two different impacts using a rod made of two 102 mm steel segments and a concrete segment. Two impact forces are used in the analysis; an impact obtained from the experiment and an impact scaled up from the experimental impact to match the intensity of blast loading obtained from Longinow and Mniszewski (1996). The maximum acceleration obtained using the experimental impact was 6,075 m/s2 while the one using the scale up impact was 16,856 m/s2 . The ratio of accelerations from the experimental and scaled up impacts is approximately 2.8, which is similar to the factor, 2.69, used to scale up the impacts. Thus the acceleration response increases linearly with the increase in the impulse load curve.

5

CONCLUSIONS

The mitigation of accelerations in wave propagation is observed when short steel segments are added at

the ends of the concrete segment. However, a material combination in the opposite order increases the accelerations in shock wave propagation. Based on these results, there is a possibility to reduce accelerations in a building using the wave speed mismatch concept. For instance, by adding steel cover or joints at the end of concrete beam, the shock wave propagates from steel to concrete and again back to steel thus reducing its intensity. Wood has lower Young’s modulus among the common construction materials. However, its density is very low giving it a high wave speed propagation. Therefore wood is not a practical material to use in hybrid structures for acceleration mitigation purposes. REFERENCES Green, D.W., Winandy, J.E. & Kretschmann, D.E. 1999. (Chapter 4: Mechanical properties of wood), Wood handbook, wood as an engineering material, Madison: U.S. Department of agriculture, forest service, forest products laboratory. Retrieved November 19, 2008, from http://www.fpl.fs.fed.us/documnts/fplgtr/fplgtr113

Livermore Software Technology Corporation. 2006. LS-DYNA [computer software]. Livermore, California. Longinow, A. & Mniszewski, K.R. 1996. Protecting buildings against vehicle bomb attacks. Practice periodical on structural design and construction 1 (1): 51–54. Solaroli, G., Gu, Z., Baz, A. & Ruzzene, M. 2003. Wave propagation in periodic stiffened shells: spectral finite element modeling and experiments. Journal of Vibration and Control, 9, 1057–1081. Sueki, S., Ladkany, S.G., O’Toole, B.J. & Karpanan, K. (2007a). Mitigation of high ‘‘g’’ impact vibrations in a projectile using material wave speed mismatch. Proceedings of 21st Canadian Congress of Applied Mechanics, CANCAM 2007, Toronto, Ontario, 3–7 June 2007. Sueki, S., Ladkany, S.G., Karpanan, K., O’Toole, B.J., Baz, A. & Berman, M. 2007b. Material wave speed mismatch for high-g acceleration mitigation. The 78th shock and vibration symposium, Philadelphia, PA, 4–8 November 2007. Toso, M. & Baz, A. (2004). Wave propagation in periodic shells with tapered wall thickness and changing material properties. Shock and Vibration, 11, 411–432.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Prefabricated multi-story structure exposed to engineering seismicity ˇ J. Witzany, T. Cejka & R. Zigler Department of Building Structures, Czech Technical University, Prague, Czech Republic

ABSTRACT: The paper presents results of an experimental and theoretical analysis of the response of a prefabricated wall structure of a multi-story building to technical seismicity effects. The experimental investigation employed a model of a seven-story carrying prefabricated wall structure of a prefabricated panel building in the scale of 1:3. The model of the prefabricated structure was gradually exposed to eight loading states, including the cyclic and monotonously rising loading. The cyclic load induced by Tira-vib electrodynamic exciter simulated technical seismicity effects. The drop of stiffness of the prefabricated system was monitored in the first place by measuring relative displacements between the wall units and the deformation of the whole system in horizontal, as well as vertical direction. 1

EXPERIMENTAL RESEARCH OF THE RESPONSE OF A MODEL OF A PREFABRICATED WALL STRUCTURE TO THE EFFECTS OF TECHNICAL SEISMICITY

In 2007, the research plan MSM6840770001 ‘‘Reliability, optimization and durability of building materials and structures’’ included the implementation of the second phase of research of the residual structural safety of a model of a prefabricated wall structure on

Figure 1. a) Experimental system, diagram of a plan and elevation arrangement of a model of a prefabricated wall structure on a 1:3 scale; b) Joint of wall and floor unit, linking bar, wall units faces coated with separation paint.

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Legend: The model of a prefabricated structure (Fig. 1) was composed of three transverse walls with an axial distance of 1.4 m (corresponding to a span of 4.2 m) and a longitudinal wall weakened by a door opening located in one transverse module. The arrangement of the prefabricated units, reinforcement of wall and floor units, reinforcement of the floor slab, and shaping of the contact interfaces corresponds to the pre-cast panel system T06B. The structural height of a story was 0.933 m (corresponding to a structural height of 2.8 m). The wall and floor units with a thickness 50 mm (corresponding to the unit thickness of 150 mm) were made of C16/20 concrete. The grout was made of concrete C16/20, and the grout reinforcement of the steel was of a E 10216 quality. The composition of the units and the arrangement of the bearing system are evident from Figure 1. Prior to assembly, the contact interfaces of the floor and wall units were finished with 2 coats of separation paint (simulation of the shrinkage crack at the contact of the units).

a 1:3 scale exposed to the effects of repetitive loading and the effects of technical seismicity (Fig. 1) (for results of the first part of the research see Witzany et al. 2001). The plan for the experimental research on the 7-story prefabricated wall structure’s model is shown in Table 1. The repetitive loading of the experimental model was carried out by a pair of steel tie rods exerting an inclined force with the horizontal and vertical components acting at the upper free end of the model. The vertical component stabilized the experimental model of the structure against tilting (exerting compressive stresses in the bed joints and substituting the effect of a vertical load, Figure 2). The repetitive loading was exerted by a pair of ‘‘enerpac RCH-603’’ hydraulic cylinders mounted on inclined steel tie rods. The dynamic load was exerted by the ‘‘TIRA vib’’ electrodynamic exciter, type TV5550/LS, 750 kg in

Table 1.

Overview of states of loading. 1st state of loading

2nd state of loading

3rd state of loading

4th state of loading

Type of loading

repetitive monotonously growing static and dynamic step-by-step load

repetitive dynamic load with an oscillation frequency of 15 Hz and a total number of oscillations 8.104 between individual static cycles

repetitive monotonously growing static and dynamic step-by-step load

repetitive monotonously growing static load, increasing in each successive loading cycle until the failure of the structure

Load [kN]

0; 10; 20; 30; 20; 10; 0

0; 10; 20; 30; 20; 10; 0 + 80 000 vibrations

0; 10; 20; 30; 20; 10; 0

0; 10; 20; 30; 20; 10; 0 0; 30; 0 0; 30; 40; 0 0; 30; 40; 50; 0 0; 30; 40; 50; 60; 0 0; 30; 40; 50; 60; 70; 0 0; 30; 40; 50; 60; 70; 80

Total

7 × repetition of static load

10 × vibration + 12 × repetition of static load

5 × repetition of static load

7 × repetition of static load

Figure 2. Diagram of loading a model of an experimental structure with inclined forces exerted by steel tie rods; picture of the mounted electrodynamic exciter.

weight fitted with a mobile weight of 13.2 kg. The exciter’s frequency range was 0–3 kHz; the maximum deflection of the mobile weight was 50.8 mm. The exciter was mounted by means of mandrels onto the floor structure of the topmost story of the model and adjusted to a horizontal oscillation. 2

RESPONSE OF A PREFABRICATED STRUCTURE TO A REPETITIVE DYNAMIC LOAD

In the individual cycles and states of loading, relative shifts in the selected vertical joints of wall units,

Figure 3.

Arrangement of the measuring devices.

horizontal deformations at three levels, vertical deformations at the gable wall footing and normal stresses in wall units were measured on the experimental model (Fig. 3). In the course of the dynamic loading, the horizontal response of the structure was measured by three Wilcoxon accelerometers, model CMMS 793 L, with an output sensitivity of 51 mV/ms−2 . One of the accelerometers was mounted onto mobile exciter parts scanning the weight motion, while the

340

Figure 5. Growth in the horizontal deformation (deflection fh ) in the individual states of loading at the 7th story level (fh × NT ), time pattern of the relationship of the total (horizontal) deformations in individual cycles (fh , ftotal × NT , fh,perm × NT ) on the upper free end.

reached a value of 2.64 mm, i.e. by 173.7% and by 1785.1% higher as compared to the first loading cycle of the 1st state of loading, and by 15.5%, or by 247.3% higher as compared to the first loading cycle of the 4th state of loading. The failure of the joints of the load-bearing units after the 1st and 2nd state of loading (a), or after the 3rd and 4th state or loading (b) is schematically displayed in Figure 5.

Figure 4. Shape of the first and second natural frequency (a, b), oscillation record for experimental determination of natural frequencies (c), record of structure’s oscillation due to dynamic load (d).

second accelerometer was mounted onto the 4th story of the model and the third onto the 7th story of the model. The test served for the determination of the first and the second natural frequencies (Fig. 4). In the 1st state of loading there was a growth in total deformation of 12.3% and permanent deformation of 50% as compared to the total and permanent deformation in the first loading cycle of the 1st state of loading. In the 2nd state of loading in the 10th loading cycle under loading by 2×30 kN, there was an increase in the total horizontal deformation by 121.1% as compared to the total deformation in the first loading cycle of the 1st state of loading and in permanent deformation by 200%. After five loading cycles of the 3rd state of loading there was a growth in the total horizontal deformation and the permanent horizontal deformation by 82.5% or by 387.5% respectively as compared to the first loading cycle of the 1st state of loading. In the 4th state of loading the total horizontal deformation under loading by 2 × 30 kN reached a value of 3.12 mm, and the permanent horizontal deformation

3

DISCUSSION OF EXPERIMENTAL RESEARCH RESULTS

The results of the experimental research carried out on an experimental model exposed to the effects of a repetitive monotonously growing load ranging between 0−2×30 kN and 0−2×80 kN and a dynamic load with a frequency of 15 Hz (8 × 105 cycles in all), the aim of which was an investigation of the response and impact of these effects on the residual rigidity and structural safety of a prefabricated system may be summed up into the following conclusions:

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– The analysis and comparison of the experimentally determined increments of the total and permanent deformations on the top of the experimental model in individual loading cycles of the 1st, 2nd and 3rd states of loading for the case of a selected frequency value of 15 Hz (experimentally measured 1st and 2nd natural frequencies at the 7th story level are f1 = 5.62 Hz and f2 = 13.92 Hz, while at the

4th story level f1 = 5.37 Hz and f2 = 13.92 Hz) suggests the relatively low impact of the dynamic effects on a gradual decrease in the rigidity of the load-bearing system resulting from joint degradation (appearance of structural cracks and their propagation in the joints of load-bearing units). The relatively high frequency of the dynamic load exerted by the electrodynamic exciter with a very low oscillation amplitude to which the experimental model was exposed in the 2nd state of loading did not cause any prominent, visually observable, failure of the joints of the load-bearing prefabricated units. – In the course of the 3rd loading cycle, no progressive growth in deformations and no prominent decrease in the rigidity and resistance of the loadbearing wall system was recorded. The growth in the total and permanent deformations caused by the impact of the repetitive low-cyclic load during the 1st–3rd state of loading, i.e. after 24 cycles, amounted to 82.5% and 387.5% as compared to the first loading cycle of the 1st state of loading. The growth of the total and permanent deformations caused by the impact of the dynamic load in the 2nd state of loading, i.e. after 80 × 104 cycles with a high frequency and a very low amplitude, which amounted to 90.9% and 166%, proves

Figure 7. (a), (b) Theoretically obtained course of vertical normal stresses σx due to the effect of loading by an inclined force 2 × 30 kN acting at the upper free end of the test structure, at the toe of the assembly at a level x = 0 m (a) and at the toe of the assembly on 4th over-ground story at a level x = 3.7 m (b). (c) The course of horizontal deformations (deflections) along the height of the test assembly at the point of the vertical joint of the longitudinal and transverse walls exposed to loading by an inclined force 2 × 30 kN on the upper free end of the test structure. Figure 6. Diagram of decreased rigidity in the vertical joints of load-bearing wall units of an experimental system in individual states of loading.

Note: The values in brackets are normal stresses σx (kPa) after a drop in rigidity of vertical joints of wall members on 4th–7th story to a value Kst = 0.08 of initial rigidity Kst = 1.0.

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that the investigated effect of the dynamic load did not cause stresses in the vertical joints of the wall units exceeding the limit of their linearly elastic action (Fig. 5). – Unlike a dynamic load with a high frequency, a low-cyclic repetitive shear load, where at least in some cycles load exceeding the proportional.

– elastic limit of the T × δ relationship is reached in vertical joints of wall units (Witzany 1987), causes a progressive decrease in the joint rigidity having consequently substantially more serious effect on a gradual decrease of the structural safety of the load-bearing system as compared to the dynamic effects caused by technical seismicity (e.g. effects of traffic, Figure 6). – Figure 7 and Figure 8 show a relatively small growth in vertical normal stresses σx and deformations during the loading cycles. The reduced rigidity value introduced in the FEM theoretical analysis Kst = 0.08 on 4th to 7th over-ground story was derived from the Kst relation displayed in Figure 6. The increase (decrease) in normal stresses σx after a drop in rigidity of vertical joints displayed in Figure 7 and Figure 8 ranges in the interval from 2 to the maximum of 10% manifesting a relatively high rigidity of the verified prefabricated wall structure to dynamic effects—i.e. a relatively low ‘‘softening’’ of vertical joints due to the effect of repetitive dynamic and static loads exerted within the scope of the respective experiment. Figure 8 clearly shows the weakest points

Figure 8. Theoretically obtained isolines of vertical normal stresses σx (kPa) due to the effect of loading by an inclined force 2 × 30 kN acting on the upper free end of the test structure in a cutaway of the structure in the case of undamaged vertical joints and in the case of a drop in rigidity of vertical joints on 4th–7th over-ground story.

Figure 9. Experimentally measured values of vertical normal stresses σx (kPa) in the course of 3rd loading cycle.

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Figure 10. Diagram of the gradual failure of the joints and units of an experimental model in the course of 1st and 2nd (a) and in the course of 3rd and 4th (b) state of loading, cracks during testing.

longitudinal walls even in the phase where they were separated on the top story of the experimental model from adjoining transverse walls and the floor structure by a continuous crack in the joints (Fig. 10). At this phase, the separated longitudinal walls acted to stiffen loosely—inserted diaphragms with a characteristic effect of a compressive diagonal prominently stabilizing the structure in the longitudinal direction against the effects of the horizontal load (Fig. 11 and Fig. 12). 4 Figure 11. Time pattern of the principal stresses in the longitudinal wall loaded by an inclined force 2 × 30 kN acting on the upper free end (disintegrated joints with lowered rigidity 10−3 ).

CONCLUSIONS

Based on the analysis of the experimental results of the response of a prefabricated experimental system to a dynamic load, a relatively high level of the reliability and resistance of similar load-bearing prefabricated wall systems of multi-story buildings to the effects of standard technical seismicity with the frequency spectrum of seismic response and the magnitude of the seismic load within the verified scope may be reported. These conclusions can be applied in their full scope to all prefabricated wall systems with sufficient horizontal reinforcement of the floor slab. ACKNOWLEDGEMENTS

Figure 12. Isolines of principal stresses in the longitudinal wall loaded by inclined force 2 × 30 kN on the upper free end (disintegrated joints with lowered rigidity 10−3 ).

The paper was written with support from the Research plan MSM 6840770001 ‘‘Reliability, optimization and durability of building materials and structures’’.

of the test structure—the joints between the longitudinal and transverse walls, lintels and subtle door pillars. Figure 7 to Figure 9 clearly show a very good accord of experimentally obtained and theoretically computed values. – The experimental loading of the experimental model within the 4th state of loading manifested an exceptionally serious (stabilization) effect of the

REFERENCES Witzany, J. 1987. Behavior of Joints of Concrete Units Loaded by Shear under Repetitive Loading. In: Pozemní stavby journal No. 8, pp. 343–348. Witzany, J., Zigler, R. & Pašek, J. 2001. Experimental Research of 3D Behavior of a 1:3 Model of a Prefabricated Wall Structure of a Multi-Story Building. In: Stavební obzor, Vol. 10, No. 12, pp. 21–23. ISSN 1210-4027.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Satisfying drift and acceleration criteria with double FP bearing M. Malekzadeh & T. Taghikhany Amirkabir University of Technology, Tehran, Iran

ABSTRACT: Double Concave Friction Pendulum (DCFP) bearing is new generation of friction isolators which contains two separate concave sliding surfaces with different properties. To accommodate enhanced performance compare to Friction Pendulum System (FPS) is one of the most important benefits of DCFP. Herein, the seismic behaviors of structures isolated by DCFP bearings are compared with the response of same buildings using FPS bearing. Accordingly, series of nonlinear dynamic analysis are carried out under ensembles of ground motions in three different hazard levels (SLE, DBE and MCE). Moreover, the adaptive behavior of DCFP and its advantageous in protecting secondary system are investigated. The probability of exceedance curves of peak roof acceleration, peak inter-story drift and peak isolator displacement are compared for two types of isolation systems. The result supports advantageous of DCFP isolation systems. 1

2

INTRODUCTION

One of the most widely implemented and accepted seismic protection systems is base isolation (Skinner et al. 1993; Naeim & Kelly 1999). The goal of base isolation is to simultaneously reduce inter-story drifts and floor accelerations to limit or avoid damage, not only to the structure but also to its contents, in a cost-effective manner. Based on observations from the January 17, 1994 Northridge earthquakes, some researchers (e.g., Hall et al. 1995; Heaton et al. 1995) have raised concerns as to the efficacy of seismic isolation during such events. With refer to these reports, design of seismic isolated buildings located at nearfault sites; the design engineer is faced with very large design displacements for the isolators. To reduce these displacements, supplementary dampers are often prescribed. These dampers reduce displacements, but at the expense of significant increases in inter-storey drifts and floor accelerations in the superstructure (Kelly 1999). The dilemma with regard to conventional isolation systems is the need to specify large amount of damping to mitigate very rare displacements, while this damping can be detrimental to the performance of the structure under occasional and rare events. A new innovative isolation system called double concave friction pendulum bearing has the ability to progressively exhibit different hysteretic properties at different stages of displacement response (Fenz et al. 2006; 2008a). This study is concerned with how this innovative isolation system can solve the dilemma related to conventional isolation systems and enhances seismic performance of isolated structure by exhibiting multi-stage behavior.

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2.1

DOUBLE CONCAVE FRICTION PENDULUM BEARING Mechanism of Double Concave Friction pendulum

The isolation bearing such as Friction Pendulum System (FPS) exhibit constant stiffness and damping under different hazard level and this behavior cause problem for design engineer to limit and control displacement at Maximum Credible Earthquake (MCE) while maintain good and desirable performance under more frequent and moderate seismic event (SLE, DBE; Kelly 1999). To improve performance of FPS bearing subjected to small and moderate seismic event an innovative friction bearing termed Double Concave Friction Pendulum (DCFP) have been introduces. This system comprise from two sliding concaved surfaces with an articulated slider. DCFP bearing has several advantageous over FPS bearing which are: 1. The lateral deformation is divided between the top and bottom concave surfaces, and consequently the required plan diameter of each concave dish is significantly less than the equivalent single-concave FP bearing (FPS). 2. The Double Concave Friction pendulum (FP) bearing exhibits desirable changes in stiffness and damping with increasing amplitude of displacement. The behavior and equations governing the forcedisplacement relationship of Double Concave Friction Pendulum in each stage are summarized in Table 1. The dynamic characteristics of DCFP bearing due to

Table 1.

Summary of double concave friction pendulum bearing behavior.

Figure

Stage Stage I: Sliding initiates on surface 1 but external force cannot overcome friction on surface 2

Force-displacement relationship

F=

W u + Ff 1 Reff 1

Stage II: External forces overcome the friction force along surface 2 and sliding occur along both surfaces

F=

F f 1 ( Reff 1 ) + Ff 2 ( Reff 2 ) W u+ Reff 1 + Reff 2 Reff 1 + Reff 2

Stage III: Slider contact displacement restrainer on surface 2 and continues sliding on surface 1

F =

Figure 1. Section through a typical double concave friction pendulum.

F W 1 Reff1

1 Reff1 + Reff2

2 eq

1

1 Reff1

the characteristic define for a good isolation system. It means that the overall force-displacement relationship is very stiff at low input shaking, softens with increasing input reaching minimum at the DBE, and then stiffens again at higher levels of input. With refer to single FP bearings there are just two target parameters which characterize isolation performance over different hazard levels but in the terms of double FP bearing due to having two different concaves which have different properties, there are different effective parameters can influence multi level performance of system. In fact design engineer is open handed in selecting different target parameters to do multi level design and enhance performance of structure over different hazard levels as it recommended in performance based-design. As an advice approach one can set lower concave parameters to characterize isolation performance over low level of excitation while performance of isolation system over moderate and high level of excitation can be define by optimizing upper concave parameters. 2.2

u

*

u eq

=

1

Reff1+ 2Reff2 Reff1+ Reff2

Figure 2. Overall force-displacement relationship of double concave friction pendulum.

action of two independent friction pendulum mechanisms, is function of seismic input level. The seismic behavior of DCFP bearing is termed adaptive because its stiffness and damping vary in proportion to displacement amplitudes. This allows the design of the isolation system to be separately optimized for multiple performance objectives and multiple levels of input. A section through a typical DCFP and overall forcedisplacement relationship are shown in Figures 1 and 2 respectively. According to Figure 2, DCFP satisfy all

W u + Ff1 R eff 1

Mathematical model

The overall hysteretic behavior of DCFP can be represented by two connected springs in series. The properties of springs are related to geometry of each concave respectively. Figure 3 shows a schematic of two single FP elements connected in series to model overall hysteretic behavior of double FP bearing. The small mass of articulated slider is shown by ms . Considering this small mass result in achieving displacements and velocities on each concave separately which are the primary parameters in order to model each FP. Gap element include to simulate stiffness exerted by displacement restrainer beyond the displacement capacity. Equation of motion for the elastic superstructure, with respect to vertical axis at center of mass of the base is as follow: [M ][¨u] + [c][˙u] + [k][u] = −[M ]R{¨ug + u¨ b }

(1)

In which M is the diagonal superstructure mass matrix, c is superstructure damping matrix, k is the

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Lastly, for the articulated slider mass ms the equation of motion is: ms (¨us + u¨ g ) + F1 − F2 = 0

(7)

where, u¨ s , u˙ s and us represent slider acceleration, velocity and displacement relative to the ground and F1 is nonlinear force which exerted by FP element 1 (FPS in Figure 3) and represent by F1 =

w (us ) + μ1 wz1 + Fr1 Reff 1

(8)

where μ1 , Z1 and Fr1 are governed by following equations: μ1 = fmax 1 − (fmax 1 − fmin 1 ) exp(−a1 |˙us |)

(9)

Z˙ 1 Y1 = A1 (˙us ) − |Z1 | (γ1 sgn((˙us )Z1 ) + β1 )(˙us ) η1

(10) Fr1 = kr1 (|us | − d1 )sign(us )H (|us | − d1 ) Figure 3. Mathematical model of structure isolated with double concave friction pendulum.

superstructure stiffness matrix (for shear building) and R is the matrix of earthquake influence coefficient. Furthermore, u¨ , u˙ and u represent the floor acceleration, velocity and displacement vector relative to the base, u¨ b is the base acceleration relative to the ground and u¨ g is the vector of ground acceleration. The equations of motion for the base are as R [M ]{{¨u} + R{¨ub + u¨ g }} + Mb {¨ub + u¨ g } T

+ Cb {˙ub } + {F2 } = 0

(2)

Mb is the mass of the rigid base, Cb is damping c efficient of any dampers included at the isolation level and F2 is nonlinear force which exerted by FP element 2 (FPS 2 in Figure 3) and represent by F2 =

w (ub − us ) + μ2 wz2 + Fr2 Reff

(3)

where μ2 , z2 and Fr2 are governed by following equations μ2 = fmax 2 − ( fmax 2 − fmax 2 ) exp(−a|˙ub − u˙ s |) (4) Z˙ 2 Y2 = A2 (˙ub − u˙ s ) − |Z2 |η2 × (γ2 sgn((˙ub − u˙ s )Z2 ) + β2 )(˙ub − u˙ s )

(5)

Fr2 = kr2 (|ub − us | − d2 )sign(ub − us ) × H (|ub − us | − d2 )

(6)

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(11)

Replacing F1 and F2 in equation (11), an additional equation of motion can be obtained in the form of: ms (¨ug + u¨ s ) +

W (us ) + fmax 1 − (fmax 1 − fmin 1 ) Reff 1

× exp(−a1 |˙us |)Z1 −

W (ub − us ) − fmax 2 Reff 2

+ (fmax 2 − fmin 2 ) exp(−a|˙ub − u˙ s |) = 0

(12)

Different approach can be implementing to solve these differential equations, such as Runge kutta method or taking advantageous of ODE functions which introduce in MATLAB for solving differential equation numerically. 3D-BASIS algorithm can be also applicable to solve above-differential equation by implementing some changes. In order to use the algorithm, three differential equations which govern state of motion related to articulated slider should be solved simultaneously in each step. In this way velocity and displacement on each surface can be calculate individually.

3

EARTHQUAKE GROUND MOTION

Study presented here set goals for investigated multistage performance of double FP bearing over different hazard levels. Achieving this goal require ground motion time history related to multiple hazard levels. As a part of SAC steel Project Somerville et al (1998) generating suites of time histories for use in performance based-design. Suites of time history are

to Nonlinear Dynamic Analysis. Figure 4 presents 5% damped absolute acceleration response spectra for SLE, DBE and MCE level. Red lines depict median response acceleration for each group. 0.1 0.08

LA56(SLE-74 year)

0.06

Normilized Force(F/W)

provided at three probabilities of occurrence: SLE (50% in 50 year), DBE (10% in 50 year) and MCE (2% in 50 year) and developed for Boston, Seattle and Los Angeles which represent a range of seismic hazard levels from seismic Zone 2 to Zone 4. These records have a wide variety of intensities and frequency content, providing an effective mean of studying multi-stage performance of double FP bearing and comparing structure over different hazard. So in this study the suite of 60 time histories developed by Somerville et al for Los Angeles used as input

3

0.04 0.02 0 -0.02 -0.04 -0.06

Response Acceleration (SLE Time Histories)

Response Acceleration

2.5

-0.08 -0.1 -60

2

0

20

40

60

80

(a) Hysteretic behavior of the double FP bearing under LA56 (SLE-74 year) Median

1

0.15

0.5

0

0.5

1

1.5

2

2.5

3

3.5

4

4.5

Normilized Force(F/W)

0.1

5

Period(sec)

4

Response Acceleration (DBE Time Histories)

3.5 3

Response Acceleration

-20

Bearing displacement(mm)

1.5

0

-40

LA01(DBE-475 year)

0.05 0 -0.05 -0.1 -0.15

2.5 2

-0.2 -300

Median

-250

-200

-150

-100

-50

0

50

100

150

200

Bearing displacement(mm)

1.5

(b) Hysteretic behavior of the double FP bearing under LA01 (DBE-475 year)

1 0.5 0

0.4 0

0.5

1

1.5

2

2.5

3

3.5

4

4.5

5

Period(sec)

0.3

Response Acceleration (MCE Time Histories)

4

Response Acceleration

Normilized Force(F/W)

5 4.5

3.5 3 2.5

LA28(MCE-2475 year)

0.2 0.1 0 -0.1

Median -0.2

2 1.5

-0.3 -600

1

-400

-200

0

200

400

600

800

Bearing displacement(mm)

0.5 0

0

0.5

1

1.5

2

2.5

3

3.5

4

4.5

5

Period(sec)

Figure 4.

Acceleration response spectra.

Figure 5. Multi stage performance of the double FP bearing under multi level time histories.

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4

BEHAVIOUR OF DOUBLE CONCAVE FRICTION PENDULUM BEARING UNDER DIFERENT LEVEL EXCITATION

To show all possible sliding stage of double FP bearing, a single story building isolated with DCFP investigated for three different level time-history: LA56 (SLE-72 year), LA01 (DBE-475 year) and LA28 (MCE-2475 year). The DCFP properties are taken as: TI = 1.5, TII = 3.5, μ1 min = 0.02, μ2 min = 0.05, μ1 max = 0.04, μ2 max = 0.1, d1 = 127 mm and d2 = No limit. Fig. 5 shows hysteretic behavior of above double FP bearing under selected multi level time histories. As it shown in Figures 5(a), 5(b) and 5(c), although double FP bearing is fully passive device, its hysteretic behavior over different seismic hazard level is compatible to a characteristic of input excitation and because of this fact the behavior called adaptive behavior. The behavior of this system under LA56 which relates to SLE with return period equal to 74 year is exactly the same as single pendulum system. Bilinear behavior of the double FP bearing is shown through Figure 4(a) under LA56. The external force which generate by LA56 is just adequate enough to move slider on a surface with lower coefficient of friction. On the other side, the external force is not able to overcome friction force caused by upper concave. By increasing seismic level up to DBE (LA01), the external force which excites the isolation system is now cable of overcoming friction force of both upper and lower concave and due to this fact the hysteretic behavior of DCFP change to tri linear under LA01 with return period of 475 year. Red dash-line in Figure 4(b) makes difference between stage I and stage II of sliding which described trough Table 1. And finally along with increasing level of excitation, Figure 4(c) represents behavior of the double FP bearing under LA28 which relate to MCE seismic level with return period of 2475 year. LA28 generate such a high external force that moves the slider on both concaves up to the level that slider contacts with restrainer displacement on lower concave. In following section performance of single pendulum bearing and double FP bearing under different hazard level are compared with each others.

5

The DCFP bearing was designed to exhibit the same median isolator displacements with the FPS under MCE hazard level. The DCFP bearing lengths were selected R1−h1 = 40 in and R2−h2 = 80 in, respectively. This corresponds to natural periods in each stage of sliding of T1 = 2 sec and T2 = 3.5 sec for sliding stages I and II respectively. The selected friction coefficients for each pendulum mechanism were μ1 min = 0.02, μ1 max = 0.04 and μ2 min = 0.06, μ2 max = 0.12 (selected coefficient of friction and radius of curvature result in μeq = 0.093).The single pendulum system (FPS) design to exhibit pendulum period of 3.5 sec (same as TII for the DCFP system) and a friction coefficient μ = 0.08. The nonlinear time-history analysis is conducted using precise mathematical model for both systems under ensembles of ground motions (60 records in three categories: SLE, DBE and MCE). The response quantities of interest are roof absolute acceleration, the relative bearing displacement and the inter-story drift of the superstructure. Roof acceleration might be critical for rigidly attached equipments and braced ceiling systems. The relative bearing displacement is crucial from the design point of view of the isolation system. Structural inter-story drift is a response quantity of importance in the assessment of performance of nonstructural components such as vertical piping, cladding, and architectural glass. The median calculated values of isolator displacement and absolute peak acceleration are summarized in Table 2 for each of the three hazard levels. Demand curves for two buildings isolated with FPS and DCFP are described in Figures 6. The results indicate that there is significant reduction in the top floor acceleration of the building isolated with DCFP under SLE and DBE hazard level while performance of both buildings during MCE is approximately the same. In fact double concave friction pendulum bearing not only protects structures against extreme earthquakes but also guaranties their performance during

PERFORMANCE OF SINGLE AND DOUBLE FP BEARING

5.1 Comparing seismic performance of DCFP and FPS

Table 2. Mean peak quantities of interest under three different hazard levels. Mean peak Seismic roof hazard Isolation acceleration level type (a/g)

Mean peak inter-story drift (mm)

Mean Peak isolation displacement (mm)

SLE

11.968 8.971 14.358 12.374 21.44 21.55

56.642 74.676 214.604 247.904 688.594 716.28

DBE

The base isolation effect of single FP bearing and double FP bearing are compared and investigate on one story superstructure with fix period of 0.5 sec.

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MCE

FPS DCFP FPS DCFP FPS DCFP

0.1995 0.1412 0.2325 0.1911 0.3587 0.3598

-1

10

10

-1

-1

10

FPS DCFP

-2

10

-3

10

-4

10

0

100

200

300

400

500

600

700

800

Mean Peak Isolator Displacement(mm)

(a) Demand curve for isolator displacement

Figure 6.

10

-2

10

-3

10

-4

0.1

-2

10

-3

10

-4

0.15

0.2

0.25

0.3

0.35

0.4

0.45

10

8

10

12

14

16

18

20

22

Mean Peak Roof Acceleration/g

Mean Peak Inter story Drift(mm)

(b) Demand curve for Roof Acceleration

(c) Demand curve for inter story drift

Median demand hazard curves for two buildings isolated with FPS and DCFP.

the frequent and moderate seismic events in acceptable level. 6

FPS DCFP Probability of Exceedance

Probability of Exceedance

Probability of Exceedance

FPS DCFP

CONCLUSIONS

The effectiveness of a recently developed isolation system, the double concave friction pendulum, for vibration control of systems has been investigated in this paper. Mathematical formulations involving differential equations have been proposed for analysis of the structure isolated by the DCFP subjected to ground motions. This study has given an exposition of dilemma in design of FPS system related to sacrificing performance during the more frequent, moderate seismic events in order to control and limit displacement under extreme earthquakes (MCE). Two isolated structures with FPS and DCFP have been analyzed in three different hazard level (SLE, DBE and MCE) under 60 records. The peak roof acceleration, peak inter-story drift and peak isolation displacement are considered as response quantities of interest. Results exhibit approximately the same performance in MCE levels however in SLE, and DBE levels, DCFP shows significant reduction in peak floor acceleration and peak inter-story drift of super-structure in compare to isolated building with FPS bearing. In fact DCFP act as an adaptive isolation system since stiffness and damping vary in proportion to level of input ground motion and can control peak floor acceleration and inter story drift together.

REFERENCES Fenz, D.M. & Constantinou, M.C. (2006). ‘‘Behaviour of the Double Concave Friction Pendulum Bearing.’’ Earthq. Engng. Struct. Dyn., 35(11), 1403–1422. Fenz, D.M. & Constantinou, M.C. (2008a). ‘‘Spherical Sliding Isolation Bearings with Adaptive Behavior’’: Theory, Earthq. Eng. Struct. Dyn., 37, 163–183. Fenz, D.M. & Constantinou, M.C. (2008b). ‘‘Modeling Triple Friction Pendulum Bearings for Response-History Analysis.’’ Earthquake Spectra, 24, 1011–1028. Hall, J.F., ed. (1995). ‘‘Northridge Earthquake of January 17, 1994 Reconnaissance Report.’’ Earthquake Spectra, 11, Supplement C, Vol. 1. Heaton, T.H., Hall, J.F., Wald, D.J. & Halling, M.V. (1995). ‘‘Response of High-Rise and Base-Isolated Buildings in a Hypothetical Mw 7.0 Blind Thrust Earthquake.’’ Science, 267, 206–211. Kelly, J.M. (1999). ‘‘The Role of Damping in Seismic Isolation.’’ Earthq. Eng. Struct. Dyn., 28(1), 3–20. Naeim, F. & Kelly, J.M. (1999) ‘‘Design of Seismic Isolated Structures.’’ John Wiley & Sons Ltd., New York, NY. Skinner, R.I., Robinson, W.H. & McVerry, G.H. (1993). ‘‘An Introduction to Seismic Isolation.’’ Wiley, Chichester, England. Somerville, P., Anderson, D., Sun, J., Punyamurthula, S. & Smith, N. (1998). ‘‘Generation of Ground Motion Time histories for Performance-Based Seismic Engineering.’’ Proc., 6th National Earthq. Eng. Conf., Seattle, Washington.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Seismic analysis of interlocking block in wall–foundation–soil system M.S. Jaafar, F. Hejazi, A.A. Abang Ali & J. Noorzaei Universiti Putra Malaysia, Selangor, Malaysia

ABSTRACT: An interlocking mortar less concrete masonry block is a innovative structural component for masonry building construction. In this study an attempt was made to analyse a special interlocking mortarless hollow concrete block system which was developed by the Housing Research Center of University Putra Malaysia. The system was subjected to earthquake loading using the finite element method. An analysis was conducted on the hollow block wall, foundation and soil interaction. For this purpose, a finite element program was developed to analyze the masonry system under earthquake excitation. In order to account for the dry joint contact between the blocks, foundation and soil mass, an interface element was used. The response of the mortarless block wall with respect to displacement, stress and acceleration subjected to earthquake was studied and the effect of applying dry interlocking joints for connection of the block on seismic response of the system investigated.

1

INTRODUCTION

An interlocking mortar less concrete masonry block building system was developed as a new structural component for masonry building construction. The interlocking mortar less block system will reduce construction time and cost of construction. Interlocking mortar less load bearing hollow block system is different from conventional mortared masonry system as the mortar layers are eliminated and instead the block units are interconnected through interlocking protrusions and grooves. Numerous analytical models have been developed to simulate the behaviour of the different types of structural masonry systems using the finite element (FE) technique. Two main approaches have been employed in the masonry modelling depending on the type of the problem and the level of accuracy required. The macromodeling approach intentionally makes no distinction between units and joints but smears the effect of joints through the formulation of a fictitious homogeneous and continuous material, equivalent to the actual one which is discrete and composite (Lotfi & Shing 1991; Cerioni & Doinda 1994; Zhuge et al. 1998). The alternative micro-modeling approach analyzes the masonry material as a discontinuous assembly of blocks, connected to each other by joints at their actual positions, the latter being simulated by appropriate constitutive models of interface (Suwalski & Drysdale 1986; Ali & Page 1988; Riddington & Noam 1994). An extensive critical review of the analytical models of different masonry systems can be found in the literature (Alwathaf et al. 2003).

351

The complex interaction between block units, dry joint and grouting material (if any) has to be well understood under different stages of loading; i.e. elastic, inelastic and failure. For interlocking mortarless masonry system, a very limited number of FE analyses have been reported in the literature (Oh 1994; Alpa et al. 1998). However, the characteristics of dry joints under seismic excitation, and their effect on the overall behavior of the interlocking mortarless wall— foundation—soil system, are still not well understood. This paper covers the analysis of a special interlocking mortar less hollow concrete block system developed by the Housing Research Center of Universiti Putra Malaysia, subjected to earthquake loads using finite element method, to study the effect of using dry interlocking joints for connection of blocks under seismic loading of the hollow block wall. The interface element is used for modelling of these joints. Also an attempt was made to model the foundation and soil and a seismic analysis of overall system was carried out. 2

PUTRA BLOCK BUILDING SYSTEM

The traditional masonry construction method which requires the use of mortar is labor intensive, and hence slow, due to the presence of a large number of the mortar joints. Therefore, there have been several attempts to develop interlocking mortarless hollow blocks in different parts of the world in the recent years. These blocks vary widely in dimension, shape and interlocking mechanism. In Malaysia, the Putra interlocking

(a) Figure 2.

Parabolic interface element.

Also it can be write:

(b) Figure 1. block.

(c)

(a) Stretcher block , (b) Corner block, (c) Half

mortar less load bearing hollow concrete block system has been developed recently by the Housing Research Centre (HRC) of Universiti Putra Malaysia. (Thanoon et al. 2004; Saleh Jaafar et al. 2006; Thanoon et al. 2008). This building system as seen in Figure 1 consists of three types of blocks: Stretcher block, Corner block and Half block. The Putra block building system does not require the mortar layer, but is capable of withstanding the vertical and horizontal loads; it also eliminates the need for steel reinforcement and therefore is very effective in reducing both cost and time of construction. 3

A typical curved parabolic interface element which is sandwiched between two continuums, as shown in Figure 2, the pair of nodes 1–1, 2–2, 3–3 and the middle line nodes a, b, c are defined by the same coordinates respectively. Brief descriptions on the formulation of the interface element are presented herein:  x= Ni xi (1)

u= v=

  

Ni yi

ub = u2T − u2B

and vb = v2T − v2B

uc = u3T − u3B

and vc = v3T − v3B

(6)

u1T , u1B , v1T , v1B , u2T · · · etc, are the nodal displacements in x and y directions, and the indices T and B indicates top & bottom continuum respectively. The above relationship can be expressed as. ⎧ T⎫ ⎪ ⎪u1 ⎪ ⎪ ⎪ ⎪   ⎪  ⎨v T ⎪ ⎬ 1 0 −1 0 ua 1 {a } = = B va 0 1 0 −1 ⎪ ⎪u1 ⎪ ⎪ ⎪ ⎪ ⎪ ⎩ B⎪ ⎭ v1 = [T ] {δa }

(7)

Similar expression can be written for b , c (8)

Here again [T] is transfer matrix. As in the case of isoparametric elements, the relative displacement

   u Na = v 0

0 Na

Nb 0

(2)

0 Nb

Nc 0

⎧ ⎫ ua ⎪ ⎪ ⎪ ⎪ ⎪ ⎪ ⎪ va ⎪ ⎪ ⎪ ⎪ ⎪ ⎪ ⎪ ⎪ ⎪ ⎨ ⎬ u b 0 Nc ⎪ vb ⎪ ⎪ ⎪ ⎪ ⎪ ⎪ ⎪ ⎪ ⎪ u c ⎪ ⎪ ⎪ ⎪ ⎪ ⎩ v ⎪ ⎭ c

Ni ui Ni νi

(3)

1 1 ξ(ξ − 1), Nb = (1 − ξ 2 ), Nc = (ξ + 1) 2 2

  u = [N ]2∗6 {}6∗1 v

(4)

Where Na , Nb , Nc are the shape function at nodes as follows: Na =

and va = v1T − v1B

{} = {a b c }T = [T ]6∗12 {δ}12∗1

INTERFACE ELEMENT

y=

ua = u1T − u1B

(5)

(9) (10)

By substituting Equation 8 into Equation 10 to get the following equation:   u = [N ] [T ] {δ} = [N ]δ {δ} (11) v

352

The strain at any point in the joint defined by the local coordinate system:     1 u 1 εt [R] [Nδ ] {δ} = [Bi ] {δ} =  = εn t v t 

Wall 5m

(12) 5m



5m

5m

Figure 4. system. Table 1.

where [Bj ] is the strain displacement matrix, [Di ] is the elasticity matrix for the joint and ds is a small length of the joint and   τ [σ ] = = [Di ] {ε} σn   Kss 0 [D] = 0 Knn

3m

5m

5m

Wall, foundation and soil properties. Modulus of elasticity (kN/m2 )

Poisson’s ratio

Density (kg/m3 )

1.4285e7 1.4285e7 5e4

0.3 0.3 0.2

2000 2000 1500

(15)

FINITE ELEMENT MODEL

In the present study the hollow prisms modelled using an eight-node isoparametric element to simulate the masonry constituents Figure 3-a and six-node isoparametric interface element of zero to model the interface characteristics of the dry joint and bond between blocks Figure 3-b. An eight-node isoparametric element is also used for modelling of foundation and soil Figure 3-c. Figure 4 shows the finite element model of the wallfoundation—soil system. In this model an interface

(a)

5m

Geometry of considered wall, foundation and soil

Wall Foundation Soil

(14)

where Kss, and Knn are the shear and normal stiffness respectively.

Hollow Block

Soil

0.3m

where u , and v are the displacements in the local co-ordinate ξ and η directions respectively. εt and εn are the tangential and normal strains respectively at the point, [R] is a rotation matrix and transfers global strains to local strains, (t) is the thickness of the element and [B] is the strain–displacement matrix of the joint. The stiffness matrix of the interface element can be written as:  [K] = [Bi ]T [Di ] [Bi ] ds (13)

4

Foundation

Interface

(b)

Foundation & Soil

element has been used for modelling of connection between wall and foundation. The material properties for the modeling of the wall, foundation and soil are summarized in Table 1. The two models prepared to investigate the effectiveness of the interlocking system for connecting the block prisms are as follow: 4.1

Model 1

In this model, the wall was modelled without considering the interlocking joints between prisms and the blocks directly connect to each others from their nodes. 4.2

Model 2

The wall was modelled by considering dry joints between blocks. As explained earlier, the interface element was used for modeling of these joints. During the first step, this analysis was carried out under static loading with self weight of blocks and loads coming from the roof and the resulting stresses were treated as initial stresses which were then imposed on the structure during seismic analysis.

(c)

Figure 3. (a) 8-noded elements used for modelling of hollow block prism, (b) 6-noded elements used for modelling of the interface, (c) 8-noded elements used for modelling of foundation and soil.

353

5

RESULTS AND DISCUSSION

Both models were subjected to earthquake in the form of ground acceleration time series previously recorded

Acceleration (m/s^2)

2 1.5 1 0.5 0

-0.5 -1 0

5

10 Time (Sec)

15

20

(a) – Model 1

Figure 5. Earthquake acceleration record for Malaysia (PGA = 0.15 g).

(b) – Model 2

Figure 7. tion (m).

Displacement of top node of the wall in Y direc-

(a) – Model 1

(b) – Model 2

Figure 6. tion (m).

Displacement of top node of the wall in X direc-

in Malaysia as shown in Figure 5 and the response of the wall was evaluated in terms of displacement, stresses and accelerations. Figure 6 shows the time history displacement of the top node of the wall in X direction during earthquake excitation. Response of both models in the X direction are plotted in Figures 6-a and 6-b respectively. As seen in these figures maximum displacements of models (1) and (2) were about 0.06 and 0.025 meters respectively. The displacement of model (2) in X direction was almost 55% less than model (1) during the imposed earthquake. It was seen that the interlocking dry joints for connecting the block prisms can effectively reduced the horizontal displacement of the wall. The displacement of model (2) in X direction was almost 55% less than model (1) and it was shown that

the interlocking dry joints for connecting the block prisms can effectively reduce the horizontal displacement of the wall. The permissible displacement value for the horizontal movement of the wall was about 0.03 meters, and the displacement of the model (2) wall in X direction with the interlocking joints was less than the permissible amount. Figure 7 shows the displacement of the top node of both models in the Y direction subjected to earthquake load. It was clear from these figure that the amount of displacement in Y direction was very small. The time history for principle stresses (S1) and (S2) at the nearest gauss point at the bottom of both model walls are shown in Figures 8 and 9 during earthquake loading. Maximum principal stress variation in the element at the bottom of the wall subjected to earthquake are plotted in figures (8-a) and (8-b) in model (1) and model (2) wall respectively. As seen in the figures the maximum nominal stress in model (1) was 84 Mpa and for model (2) was 8 Mpa. Therefore the maximum principal stress was reduced to about 90% in model (2) compared with model (1) due implementation of the dry joints for connecting the block prisms. The compressive strength of Putra block was about 23 Mpa, therefore model (2) shows it can resist the imposing earthquake because the stress was only 8 Mpa and less than the ultimate compression strength of the block of 23 Mpa; so it is in an acceptable range of stresses. But the stresses in model (1), exceeds the permissible strength of block and will lead to failure. Also Figure 9-a and 9-b show the minimum principal stresses in the same element for Model (1) and model (2), respectively. The minimum nominal stress in model (1) was about 82 Mpa and in model (2) is

354

(a) Model 1

(a) – Model 1

(b) Model 2

(b) – Model 2

Figure 8. (MPa).

Principal stress of bottom element of the wall—S1 Figure 10. Acceleration of top node of the wall in X direction (m/s2 ).

the wall as shown in Figure 10-a subjected to earthquake were plotted in Figure 10-b for model (1) and Figure 10-c for model (2) wall. From these plots it was obvious that the maximum acceleration was about 17 g for model (1) and 26 g for model (2). The acceleration was increased by about 50% in some time steps of load excitation (Earthquake acceleration) on model (2) wall compared with that in model (1). Therefore from the results it was obvious that the interlocking wall system was suitable for low earthquake area by just adding some strengthening system to resist the tensile stresses as mentioned above. But in high intensity earthquake area this system is not recommended.

(a) – Model 1

(b) – Model 2

Figure 9. Principal stress of bottom element of the wall—S2 (MPa).

8.1 Mpa. It was clear that the amount of minimum stress is about 90% smaller than that in model (1) in most time step of earthquake excitation. Therefore, the use of dry interlocking mechanism for connecting of the blocks leads to reduced stress in the wall. The tensile strength of the block was about 2.06 Mpa. So the minimum stress in model (2) was more than the tensile strength of the blocks and it would fail. As observed the use of interlocking dry joints for connecting the blocks effectively reduce the amount of minimum stress in the wall but again it is more than tensile strength of the block. Thus it is necessary to provide some additional strengthening system such as reinforcement for increasing the tensile strength of the wall to resist tensile stress. Time series acceleration of the top node of

355

6

CONCLUSIONS

Based on the foregoing analysis and discussions on the test results from this investigation, several conclusions can be drawn as follows: 1. The finite element model of the mortar less block masonry wall-foundation-soil system has been successfully developed and includes the modelling of masonry materials, mortarless dry joint and block-grout interface behaviour. 2. The interlocking keys provided for the building system were able to integrate the blocks into a sturdy wall and can replace the mortar layers that are used for conventional masonry construction in low seismic area. 3. Application of dry interlocking joints for connecting the blocks effectively reduces the seismic response of the system. Maximum displacement of

the wall is reduced by about 50% in the X direction. Also the maximum and minimum principal stresses were reduced by about 90%. 4. The results show that the interlocking hollow block wall system displacement subjected to earthquake was within acceptable range. 5. The maximum stress in the model (2) wall was less than the compressive strength of the block prism and it was within an acceptable margin of compressive strength but the minimum stress exceeded the tensile strength of the block, and hence some reinforcement need to be provided for this. ACKNOWLEDGMENTS The financial support granted by the Academy of Sciences, Malaysia through the Housing Research Center of Universiti Putra Malaysia is gratefully acknowledged. REFERENCES Ali SS, Page AW. 1988. Finite element model for ma-sonry subjected to concentrated loads. Journal of Structural Engineering, ASCE; 114(8): 1761–84. Alpa G, Gambarotta L, Monetto I. 1998. Dry block as-sembly continuum modeling for the in-plane analysis of shear walls. In: Proceeding of the 4th international symposium on computer methods in structural masonry. E & FN, Spon; pp. 111–8. Alwathaf AH, Thanoon WAM, Noorzaei J, Jaafar MS, Abdulkadir MR. 2003. Analytical models for different masonry systems: Critical review. In: Proc. of IBS2003 conf.

Cerioni R, Doinda G. 1994. A finite element model for the nonlinear analysis ofreinforced and prestressed masonry wall. Computer and Structures; 53: 1291–306. Lotfi H, Shing P. 1991. An appraisal of smeared crack model for masonry shear wall analysis. Computer and Structures; 41: 413–25. Oh K. 1994. Development and investigation of failure mechanism of interlocking mortarless block masonry system. Ph.D. thesis. Philadelphia: Drexel University. Mohd Saleh Jaafar , Waleed A. Thanoon, Amad M.S. Najm, Mohd Razali Abdulkadir, Abang Abdul-lah Abang Ali. 2006. Strength correlation between individ-ual block, prism and basic wall panel for load bear-ing interlocking mortarless hollow block masonry. Journal of Construction and Building Materials. Vol. 20; p. 492–498. Riddington JR, Noam NF. 1994. Finite element predic-tion of masonry compressive strength. Computer and Structures; 52(1): 113–9. Suwalski P, Drysdale R. 1986. Influence of slenderness on the capacity of concrete block walls. In: Proceed-ing of 4th Canadian masonry symposium; p. 122–35. Thanoon, W.A., Jaafar M.S., Abdul Kadir, M.R., Ali, A.A., Trikha, D.N. and Najm, A. M. 2004. De-velopment of an Innovative Interlocking Load Bear-ing Hollow Block System in Malaysia, Construction and Building Materials, 18: 445–454. Waleed A.M. Thanoon, Ahmed H. Alwathaf, Jamaloddin Noorzaei, Mohd. Saleh Jaafar, Mohd. Razali Abdulkadir. 2008. Finite element analysis of inter-locking mortarless hollow block masonry prism. Journal of Computers and Structures. 86: 520–528. Zhuge Y, Thambiratnam D, Coreroy J. 1998. Nonlinear dynamic analysis of unreinforced masonry. Journal of Structural Engineering, ASCE; 124(3): 270–7.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Shear crack width of RC column with cut-off rebar under cyclic loading T. Tsubaki Department of Civil Engineering, Yokohama National University, Yokohama, Japan

M. Dragoi & J. Onishi Yokohama National University, Yokohama, Japan

ABSTRACT: In determining the degree of damage of a reinforced concrete (RC) column caused by earthquake loading it is important to know the maximum load applied to the RC column during the earthquake. In this study it is examined if the data of shear cracks in a RC column with cut-off longitudinal reinforcement can be used as an indicator of the damage of the column to estimate the maximum load applied to the column. The cracks in a RC column are considered to reflect the effect of loading and they are easily observed and analyzed after earthquake. Starting from shear crack width, empirical relationships are determined to estimate the maximum load that causes the shear cracks. Those relationships are established by analyzing experimental data of RC column specimens with axial load and shear reinforcement ratio in the typical range of values.

1

INTRODUCTION

RC columns with cut-off reinforcement are known to suffer severe damage due to earthquake loading. The damage tends to concentrates around the position of cut-off and many cracks are observed in that zone. In this study simple empirical linear relationships to estimate the maximum load experienced by RC columns with cut-off reinforcement subjected to reversed cyclic loading are to be obtained by analyzing the results of the test data of RC columns subjected to cyclic loading and using the shear crack width of the column in unloaded states. Up to now the existing formulas for shear crack width are derived mostly from the test data of RC beams. CEB-FIP Model Code 1990 (CEB 1993) and Hussein, Sabry & Ueda (1991) determined the prediction formulas for shear crack width which are based on the stress of the shear reinforcement. Using the experimental results of RC columns (Ohtaka et al. 2004), a set of formulas for shear crack width for a column were proposed by Dragoi & Tsubaki (2006b). The appropriateness of using the data of shear crack to estimate the maximum load was examined by analytical methods as well (Tsubaki & Dragoi 2005; Dragoi & Tsubaki 2006a). The design importance of shear crack width is also stated in JSCE standard specification (JSCE 2002). Tsubaki et al. (2007) proposed the prediction formulas to estimate the maximum load applied to RC columns based on the shear crack width. This study is the extension of the above work to the case of RC columns with cut-off reinforcement. The effect of cutoff reinforcement on the relationships for the shear crack width at loaded and unloaded states, the average

357

strain in a shear reinforcing bar at loading state and the maximum shear force applied to the column is clarified.

2 2.1

EXPERIMENT OF RC COLUMNS Specimen

As in the experiment by Ohtaka et al. (2004) two column specimens with or without cut-off reinforcement have the same size, i.e., cross section 300 mm × 300 mm, and the height (the distance from the base position of a column to the loading point) is 900 mm. The size of footing is 1300 × 400 × 600 mm. For concrete the average compressive strength is 30 MPa, the maximum aggregate size is 20 mm, the water cement ratio is 58.5% and the sand aggregate ratio is 45.9% by volume. For the longitudinal reinforcement, D13 deformed bars with the yield stress 445 MPa, the yield strain 2301 × 10−6 are used and the reinforcement ratio is 2.25%. The shear reinforcement ratioρs is 0.42% with D6 deformed bars with the yield stress 332 MPa and the yield strain 1777 × 10−6 , and the spacing between shear reinforcing bars is 50 mm. The details of the specimen size and the loading system are given in Figures 1–4. The arrangement of reinforcement is shown in Figures 1–2. Two types of specimens are used. The specimen with cut-off reinforcement and that without cut-off reinforcement are shown in Figure 1 and Figure 2 respectively. The cut-off position is mid-height of the column part of the specimen. The amount of reinforcement is

Figure 3. Position of strain gauges for longitudinal and shear reinforcement. Figure 1.

RC column specimen with cut-off reinforcement.

Horizontal Actuator

900

1500

Vertical Actuator

400

300

[mm]

1300 Figure 4.

Figure 2. RC reinforcement.

column

specimen

without

cut-off

reinforcing bars is the same for all specimens. The length of the protection covering for the strain gauges is around 30 mm.The strain gauges are also placed on the longitudinal reinforcing bars in the zone at the base part of the column in order to determine the yielding of the longitudinal reinforcing bars. 2.2

reduced to one-half of that of the base position of the column. The arrangement of strain gauges is shown in Figure 3. Strain gauges are placed on three shear reinforcing bars with 150 mm interval from the base of the column part of a specimen in the zone where shear cracks are expected. The position of these three shear

Loading system.

Loading

The cantilever type loading system used in this study is shown in Figure 4. The horizontal and vertical loads are applied by hydraulic actuators. The compressive stress by the vertical applied load is 1 MPa close to the axial load level of actual bridge piers. In this experiment the specimens are subjected to reversed cyclic loading by gradually increasing the load each cycle

358

3.2

Figure 5.

For the specimen without cut-off reinforcement cracks are concentrated in the base zone of the column where many flexural cracks are observed. Cracks are observed even at the height of 300 mm and 450 mm from the base. When the displacement is small, only hair-line shear cracks are observed. In the specimen with cut-off reinforcement many shear cracks are observed around the zone of cut-off reinforcement. Shear cracks with large crack width are observed even for small displacement. In the zone below the cut-off zone shear cracks are also observed. When the rotation angle is 8/200 rad, spalling of concrete is observed at the cut-off zone.

Shear crack measurement.

up to the yield load, then controlling the deflection of the column, increasing the rotation angle by 1/200 rad each cycle, i.e., 2/200, 3/200, 4/200, 6/200 and 8/200 rad. These positive horizontal displacements (rotation angles) are associated with the loading in the left direction in Figure 4. The load is applied at 900 mm from the base position of the column by controlling the deflection. The displacement at the loading point is repeated for three cycles and then increased by 1/200 rad for another set of three cycles up to the failure of the specimens. 2.3

Crack distribution

Shear crack measurement

The shear crack width is measured by using a digital microscope for loading and unloading stages of each loading cycle. The images of shear cracks have been taken at the intersection of the shear crack with shear reinforcement as shown in Figure 5. Analyzing the images the shear crack width along a shear reinforcing bar Wt , local crack angle θcl , shear slip δ, shear slip angle θs and crack opening W are determined as shown in Figure 5 where X , Y , D, Wl and Wt are directly measured.

3.3

Shear crack width

The starting point in the estimation of the maximum load experienced by RC columns during reversed cyclic loading is cracking pattern and crack width. Because the crack width is obtained after an earthquake a relationship between crack width for unloaded state (after the earthquake) and crack width for loaded state (during the earthquake) must be determined to connect these states. The shear crack width for loaded state is used to determine the strain in shear reinforcement produced by the shear force during loading. The relationship between the shear crack width of loaded state and that of unloaded state is shown in Figure 6 for loading stages up to 3/200 rad. The shear crack width shown in the figure means the total sum of the crack width of all the shear cracks crossing the specified shear reinforcement. The test data indicates that there is a linear relationship between the crack width at loaded state and that at unloaded state. 700 600

[Average]

500

3

EXPERIMENTAL RESULTS 400

S = 2.37S0

[No Cut-off 300mm]

3.1 Load-displacement relationship

[Cut-off 450mm]

S = 6.03S0

For the specimen without cut-off reinforcement the load is kept increasing as the displacement at the loading point increases from 1/200 rad to 3/200 rad. On the other hand the load becomes almost constant although the displacement is increasing. For the specimen with cut-off reinforcement the load is kept increasing as the displacement increases up to 2/200 rad. For larger displacement, however, the load does not increase. Moreover the load tends to decrease with increasing displacement. When the rotation angle is 8/200 rad, the load decreases up to 70 to 80% of the maximum load.

359

300 200

S = 2.57S0

[Cut-off 300mm]

100 S = 2.82S0 [No Cut-off 450mm]

0 0

50

100

150

200

250

S0 Figure 6. Relationship between shear crack width at loading and that at unloading.

In Figure 6 there is a scatter in the data of no cut-off at 300 mm height. This is considered to be attributed to the fact that the crack width in that zone is relatively small compared to that in other zones. It is confirmed from the test data that for the displacement up to 3/200 rad, the effect of load repetition on the shear crack width seems to be negligible. To determine the relationship between the values of unloaded and loaded states for shear crack width the sum of the shear crack width measured along a shear reinforcing bar is considered. The total crack width crossing the shear reinforcement is assumed to be related to the average strain in the shear reinforcement. From the test data it is observed that the relationship for shear crack width in loaded and unloaded states is related to the yielding of shear reinforcement. The data used in the regression analysis in this study are those obtained before yielding of shear reinforcing bars. A linear function is used in the regression analysis as follows. S = aS0

(1)

In Equation 1 S and S0 are the total sum of shear crack width of shear cracks along a shear reinforcing bar for loaded state and unloaded state respectively. The relationships between shear crack width of loaded and unloaded states are shown in Figure 6 for all measured positions and the average. Equation 1 is valid up to the yielding of shear reinforcing bars. When the stress state of shear reinforcement is beyond yielding, the relationship between the shear crack width of loaded state and that of unloaded state becomes a bilinear function (Dragoi & Tsubaki 2006b). The bilinear relationship indicates the change of tendency by yielding of shear reinforcement. The slope of the second part after yielding in the bilinear relationship is smaller than that of the first part before yielding as long as the test data of Ohtaka et al. are concerned (Dragoi & Tsubaki 2006b). In addition to the measurement of shear crack width, the shear slip along the shear crack has also been measured. In case of a shear crack the crack width is related to the magnitude of the diagonal tensile stress and the shear slip along the shear crack surface. The asperity or the surface roughness of the shear crack surface increases the opening of crack by shear slip. Therefore shear slip can also be considered as an indicator of shear deformation. The test data indicate that the shear slip is a function of shear crack width and the ratio between them during load repetition is almost constant. This means that the shear crack width is a good indicator of shear deformation as the shear slip. From the point of view of obtaining a relationship to estimate the maximum load experienced by RC columns based on the strain in shear reinforcement, the shear crack width is more directly related to the

strain in shear reinforcement than the shear slip which does not contribute the change in the stress of shear reinforcement. Therefore, the shear crack width is considered appropriate to be used for the purpose of this study. In order to use the shear crack width as an indicator of load history it is important to confirm the monotonically increasing nature of shear crack width as the loading proceeds. It has been confirmed that the test data used in this study shows the monotonically increasing relationship. 3.4

Strain in shear reinforcement

In the analysis the average strain in the shear reinforcement at the same level where crack width was measured and the sum of crack width crossing the shear reinforcement are considered. The average strain has been obtained using the following equation. ε = (εA + εB + εC + εD + εE )/5

(2)

where εA , εB , εC , εD , εE are the strain measured by five equally spaced strain gauges along one shear reinforcement as shown in Figure 3. These strains of the shear reinforcing bars are the average value of two strain gauges pasted on the opposite sides of the reinforcing bar at the specified points. The average strain in shear reinforcement is considered proportional to the sum of the crack width of a shear crack crossing a shear reinforcing bar. The proportionality coefficient is given by coefficient b expressing the effect of influencing factors. It is assumed that the shear reinforcement has not yielded. ε = bS

(3)

The results of the regression analysis are given in Figure 7 for all measured positions and the average. The data of strain are obtained for the same stages as the shear crack width. For these stages a small effect of load repetition on the relationship between shear crack width and strain in shear reinforcement has been observed. Analyzing the experimental data the relationship between the shear crack width and the strain in shear reinforcement can be understood to be considered as a relationship expressing the effect of bond degradation mechanism. When the shear crack width is small, the ratio between shear crack width and strain in shear reinforcement is large. When the shear crack width increases, the ratio becomes smaller and almost constant for a certain value of crack width. If the stress state of shear reinforcement is beyond yielding, the relationship between shear crack width and strain in shear reinforcement becomes a bilinear relationship (Dragoi & Tsubaki 2006b). After yielding

360

1200

160

[Average]

Load (Test) (kN)

S

800

[Cut-off 300 mm]

[Cut-off 300 mm] [No Cut-off 450 mm]

600 400

120 100 80 60

[No Cut-off 300 mm]

40

200 0

(a) No cut-off R = 0.745

140

1000

20 0

200

400

0

600

0

S m)

Figure 7. Relationship between strain in shear reinforcement and shear crack width at loading.

160

V = 0.128ε + 70.7

120 100

Load (Test) (kN)

V(kN)

[No Cut-off 450 mm]

[Average] [No Cut-off 300 mm] V = 0.117ε + 66.3 V = 0.104ε + 60.5 [Cut-off 300 mm] V = 0.0542 ε + 56.8 [Cut-off 450 mm]

80 60 40

160

120 100 80 60 40

20 0 0

200

400

600

800

1000

20

1200

ε(μ)

0 0

Figure 8. Relationship between maximum load and average strain in shear reinforcement. 160

the shear crack width becomes significantly large with the increase of strain in shear reinforcement. Estimate of maximum load

In order to estimate the maximum load experienced by RC columns a linear function is used to express the increase of load by the increase of strain in shear reinforcement. The load V has been obtained from the average strain in shear reinforcementε using the following linear equation with empirical coefficients c1 and c2 . V = c1 ε + c2

(4)

In the elastic range the shear force acting on the column is related to the stress in shear reinforcement that depends on the area of shear reinforcement in the vicinity of the crack and the value of the Young’s modulus of the shear reinforcement. Coefficient c1 obtained for each column specimen is found to be dependent on the axial load and the properties of the shear reinforcement. Another influence on the maximum load is given by the longitudinal reinforcement ratio that is expressed by coefficient c2

361

40

80 120 Load (Predicted) (kN)

160

(c) Position at cut-off R = 0.894

140 Load (Test) (kN)

3.5

80 120 Load (Predicted) (kN)

(b) Position below cut-off R = 0.960

140 160 140

40

120 100 80 60 40 20 0 0

40

80 120 Load (Predicted) (kN)

160

Figure 9. Comparison between predicted maximum load and test data.

which shows the load at which shear cracks occur. This effect can be included in Equation 4 by making the constant term of the equation a function of the crosssectional area of concrete. Thus the size effect on the formula is also considered. In Figure 8 the results of the regression analysis for all the data obtained in this study are shown.

The results of the regression analysis are obtained for the same loading stages used for the relationship between the shear crack width and the strain in shear reinforcement discussed before. The regression lines are obtained for each cycle of load repetition. The slope of a regression line is increasing with higher value of strain at the beginning of each loading cycle. From the results shown in Figure 8 for all the data obtained in this study to estimate the maximum load from the strain in shear reinforcement, it is confirmed that the regression lines are different depending on the position of shear cracks, i.e., the position without cut-off reinforcement, that below cut-off and that at cut-off. This is considered due to the different stress condition at these positions. The comparison between the predicted maximum load experienced by the specimens and the test data is shown in Figure 9. The coefficient of correlation is in between 0.745 and 0.960. It can be said that the assumptions made for the relationship between the average strain of shear reinforcement and the shear crack width and the relationship between the maximum load experienced by a RC column and the average strain in shear reinforcement are appropriate if the stress state of shear reinforcement is within the elastic range. In general the empirical coefficients c1 and c2 in Equation 4 are functions of the area of shear reinforcement, the cross-sectional area of concrete, the longitudinal reinforcement ratio, and the axial compressive stress.

3. On the relationship between the maximum load and the strain of shear reinforcement: This relationship is linear for all the positions where the shear crack width was measured. The slope of the relationship is smaller at the position of cut-off. It is confirmed that the effect of cut-off reinforcement is significant in this relationship compared to the above two relationships. Therefore the slope of the regression line is different depending on the position of shear cracks. 4. On the range of applicability of the proposed relationships: The range of applicability of the proposed prediction formulas for the maximum load experienced by RC columns with cut-off reinforcement under reversed cyclic loading is that the compressive strength of concrete is around 30 MPa, the shear reinforcement ratio is 0.42%, the longitudinal reinforcement ratio is 2.25%, the average axial compressive stress to the column due to the axial load is 1 MPa, the number of load repetition is up to three times, and the strain in shear reinforcement is assumed to be less than the yield strain. This range of applicability of the proposed empirical formulas covers the RC columns with typical values of parameters. 5. Remarks: To obtain a universal empirical formula for shear crack width it is necessary to accumulate experimental data considering the parameters such as concrete cover, reinforcing bar diameter, size effect, compressive strength of concrete and type of the structural element.

4

REFERENCES

CONCLUSIONS

A set of empirical relationships to estimate the maximum load applied to RC columns with and without cut-off reinforcement from the shear crack width in unloaded states are shown in the present study. The conclusions obtained from the present study are summarized as follows. 1. On the relationship between the shear crack width for unloaded state and that for loaded state: It is confirmed that the relationship between the shear crack width for unloaded state and that for loaded state is linearly proportional. In case of cut-off reinforcement the crack width is larger than the case of no cut-off for the same load. 2. On the relationship between the shear crack width for loaded state and the strain of shear reinforcement: The relationship between the shear crack width for the loaded state and the strain of shear reinforcement is turned out to be linearly proportional for the loading state examined in this study. The bond between the shear reinforcement and the surrounding concrete is considered to be still sound.

CEB 1993. CEB-FIB Model Code 1990. Thomas Telford. Dragoi, M. & Tsubaki, T. 2006a. Strain estimate method of RC columns subjected to cyclic loading. Journal of Applied Mechanics, JSCE 9: 437–444. Dragoi, M. & Tsubaki, T. 2006b. Shear crack width and maximum load of RC columns subjected to cyclic loading. Proc. of JCI 28(2): 811–816. Hussein, M.H., Sabry, A.F. & Ueda, T. 1991. Displacements at shear crack in beams with shear reinforcement under static and fatigue loading, Proc. of JSCE 433(V-15): 215–222. JSCE 2002. Standard Specification for Design and Construction of Concrete Structures-2002 ‘‘Structural Performance Verification,’’ Chapter 7, Verification of Serviceability. JSCE Guidelines for Concrete 3: 107–130. Ohtaka, M., Hayashi, K. & Tsubaki, T. 2004. Shear crack width of RC columns subjected to reversed cyclic loading. Proc. of JCI 26: 1033–1038. Tsubaki, T. & Dragoi, M. 2005. Indices for geometrical properties of cracks of RC structures. Journal of Applied Mechanics, JSCE 8: 481–488. Tsubaki, T. & Dragoi, M. 2007. Shear crack width of concrete member under axial load and transverse reversed cyclic load. Proc., ISEC-4, Innovations in Structural Engineering and Construction 1: 421–426.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Structural behavior of steel frame connections subjected to blast G.S. Urgessa & T. Arciszewski George Mason University, Fairfax, Virginia, USA

ABSTRACT: When a structural steel frame is subjected to blast, the beam-to-column connections, which are responsible for load transfer between different members within the frame, play a major role in structural response. This paper presents results of a comparative finite element analysis of a steel frame subjected to a blast loading from a vehicular threat. The study compared three connection systems referred as standard, TA and SidePlateTM . Connection plate thickness variations were also considered. Strain rate effects were included in the material constitutive model. The pressure-time histories were determined using FEFLO, a general purpose computational fluid dynamics program. A maximum sway criterion is used to determine comparative advantages and disadvantages of the TA and SidePlateTM connection types with respect to the standard connection. 1

INTRODUCTION

Many federal agencies and large private building owners require that new buildings be designed and existing buildings be upgraded to resist the effects of blast loadings. However, current design guidelines for steel connections regarding structures subjected to blast loads do not impose definitive detailing requirements (NRC 1995; DOD 2003). Earlier studies of steel connections subjected to dynamic loads were conducted for steel momentresisting frames subjected to seismic events following the 1994 Northridge earthquake. These studies focused on the behavior of steel connections under cyclic loading. In recent years, there is a great interest in investigating the performance and behavior of steel connections that are subjected to blast loadings. The most widely referenced document for blast resistant design is the Army Technical Manual (TM5-1300). The manual only outlines considerations necessary for connection design and it does not provide any specific design guidelines for the connection systems (Dept. Army 1990). Houghton and Karns (2003) studied a finite element analysis of a SidePlateTM connection system for blast resistance. The side plates act as discrete structural continuity elements to sandwich and connect girders and columns. The researchers found out that the side plate connections exceed the full strain-hardened flexural capacity of the connected elements and the systems cyclic rotational capacity satisfies inter-story drift requirements. The paper asserts the imperative need for more research to develop new types of connection systems that will provide essential structural linkage of beam-column connections. Hamburger and Whittaker (2004) studied the design of steel structures for blast related progressive collapse. Their research findings acknowledge the need to develop

363

new connection types to make blast-resistant designs more efficient. They concluded that the standard web plate connection is not capable of allowing joints to develop the large inelastic rotations and tensile strains necessary for resisting progressive collapse through large deformation behavior. Sabuwala et al. (2005) studied the behavior of fully restrained steel connections subjected to blast loads. They studied a web side-plate connection system and a welded flange plate connection system. The investigation was conducted using a general purpose finite element code and the analytical results were compared with results of a prior experimental research. The welded flange plates showed improved performance under blast loads with reduced stresses in the connection regions. The paper also highlights the need for incorporating a strength based criterion into the existing recommendations of TM5-1300. Chen and Richard Liew (2005) presented a numerical approach for the analysis of rigid steel frames subjected to explosion and fire. Although the approach was focused on the global response of the frames, they used member end forces to determine the resistance of joint components to ensure the validity of rigid frame analysis. The approach included detailed behavioral effects associated with high-strain rate loading and the effect of elevated temperature. Krauthammer and Cipolla (2007) found that finite element analysis of steel frame structures is sensitive to failure modes of connections. They studied the non-linear momentrotation relationship of joints in a steel frame with rigid and semi-rigid connection. The paper argued that current structural design considerations such as those presented in TM5-1300 underestimate load carrying capacity of frames for severe blast loading. The work was conducted using an isotropic elasto-plastic material model for connection components.

The objective of this paper is to present results of a comparative study of structural behavior under blast for a steel rigid frame with three different types of beam-column connections. The analysis was conducted for several connection plate thicknesses using blast pressures calculated from a computational fluid dynamics code. Strains rate effects were also considered.

2

FINITE ELEMENT ANALYSIS

Figure 1 shows an eight story steel frame in a typical office building which was considered for this study. Frame members were designed and optimized to carry a live and dead load for a column spacing of 9.14 m using the International Building Code (IBC 2003). The story height was 3.81 m from the centerline of floor beams. Table 1 shows the selected member sizes and their locations. ABAQUS/Explicit finite element analysis software was selected for this study. The program is a specialpurpose analysis tool that uses explicit dynamic finite element formulation. It is proven to be suitable for modeling brief, transient dynamic events such as blast and impact problems (ABAQUS 2004). In the explicit approach, acceleration of any node is determined completely by its mass and net force acting on it. The corresponding velocity and displacement are calculated by using numerical integrations of the

acceleration function. Element type S4R has been selected for analysis after careful consideration of a trade-off between computational time and relative accuracy. S4R is a fully integrated general purpose conventional shell element in ABAQUS and it accounts for finite membrane strains and arbitrarily large rotations which makes it suitable for largerotation problems. The selected element type uses a reduced (lower-order) integration to form the element stiffness with only one integration location per element. The reduced integration significantly reduces model running time with a relatively reasonable result (Cook 2002). The only disadvantage of using S4R element is its susceptibility to hourglass distortions. Reduced integration elements tend to be too flexible and create a zero-energy mode distortion with no strain energy related to the distortions. However, the ABAQUS/Explicit solver has hourglass control which minimizes the problem without introducing excessive constraints on the element’s physical response. 2.1

For the analysis presented here, threat parameters were determined using the recommendations provided by Conrath et al. (1995). The aggressor tactics was assumed to be a stationary vehicle bomb with an equivalent TNT value of 90.7 kg at 3.96 m from the face of the column. The threat parameter satisfies the criterion of a medium design basis threat severity level. The corresponding blast pressure-time histories were determined by FEFLO, a general purpose computational fluid dynamics program developed by the Center for Computational Fluid Dynamics at George Mason University. FEFLO is based on adaptive unstructured grid methodology and is capable of producing timeaccurate solutions for non-linear geometries and wide range of flow regimes (Löhner et al. 2001). Pressuretime histories were calculated for 606 target points throughout the height of the column. Figure 2 shows the plot of the maximum (peak) pressure-time history observed at the column next to the charge. 2.2

Figure 1. Table 1.

Eight-story steel frame model. Member sizes.

Story number

Beam size

Column size

1 and 2 3, 4 and 5 6, 7 and 8

W30 × 116 W24 × 104 W24 × 104

W14 × 120 W14 × 90 W14 × 61

Determination of blast pressure

Material properties

Stress-strain relationships of materials for high-strain loadings such as blast are not widely available when compared to stress-strain relationships for static loadings. One approach to determine the dynamic stressstrain relationships is applying a factor called a Dynamic Increase Factor (DIF) to the static stressstrain relationships (Dept. Army 1990). Steel experiences an increase in strength under rapidly applied loads. It cannot respond at the same rate as which the load is applied. Thus the yield strength increases and results in the reduced plastic deformation. Therefore at a fast strain rate, a greater load is required to produce the same deformation than

364

at a lower rate. Strain rate is usually measured at logarithmic intervals. The only disadvantage of inputting rate-sensitive material constitutive behavior is the possibility of introducing nonphysical high-frequency oscillations. Fortunately, the ABAQUS/Explicit solver is capable of filtering these oscillations through an assigned rate-sensitive factor (ABAQUS 2004). For the finite element analysis, strain-rate parameters were selected that satisfy the Cowper-Symond overstress power law. The law relating the dynamic yield stress to the static yield stress has the form  σd = σ o

ε˙ 1+ D

The material constants D = 40.0 and n = 5.0 are selected in order to match Cowper-Symond overstress power law constants. Figure 3 shows the rate-sensitive constitutive model used in the analysis which is based on the recommendations of the ASCE task committee on blast resistant design (ASCE 1997).

 1q (1)

where σd = dynamic yield stress, σo = static yield stress, ε˙ = strain rate and, D and q = Cowper-Symond coefficients. Rate-dependent behavior is expressed in terms of equivalent plastic strain rate, ε˙ pl = D (R − 1)n

(2)

where R = ratio of the dynamic yield stress to the static yield stress and, n and D = material constants.

Figure 4a.

Standard connection.

Figure 4b.

TA connection.

Figure 4c.

SidePlateTM connection.

Figure 2. Peak pressure-time history at the column closest to the charge.

Figure 3.

Material constitutive model.

365

2.3

Connection types

The analysis was conducted for three types of beamcolumn moment resisting connections. Figure 4a shows the first connection type referred as the ‘‘standard connection’’. In this case, there are separate connections between individual beams and columns. These connections are in the form of horizontally positioned plates on the top and bottom of the individual beams and the plates are welded to both beams and columns. Such connections are widely used in steel construction in the post-Northridge earthquake era. Figure 4b shows the second connection type referred as the ‘‘TA connection’’. It has been developed and patented by the second author in the mid1970s. It is an integrated connection, i.e. at a given location a single system connects the column and all beams. In the case of a rigid frame, the connection is in the form of two horizontal and two vertical plates welded together and connected to the column. Individual beams are connected to this system through fillet welds or through bolts. The horizontal plates are intended to ‘‘shift’’ plastic zones away from columns as the frame undergoes large elastoplastic deformations during blast. Figure 4c shows the third connection type referred as the ‘‘SidePlateTM connection’’. This connection system was developed after the events of September 11, 2001. The connection is also an integrated connection, as the TA connection, although in this case the integration of connecting beams and the column takes place only in the plane of bending. The connection can be described as an expanded and stiffened standard connection. In addition to the horizontal flange plates used in the standard connection, two vertical plates are positioned at both sides of the column and connected with the horizontal flange plates. Also, a number of stiffeners are used, vertical between the beams and the side plates, and horizontal between the column and the side plates. In this way, a rigid box is created which significantly increases the stability of the entire connection and its bending and torsional moment capacities with respect to the standard connection. The comparative analysis of the considered rigid frame with three types of connections was conducted for six model cases. It was a computationally costly analysis with each run conducted for 0.4 seconds Table 2.

time period and requiring approximately 24 hours of CPU time. In all three types of compared connections, horizontal plates are used which transfer the bending moments from the beams to the column. Such plates are refereed as ‘‘connection plates’’ in the paper. The connection plate thickness is usually assumed as comparable with the thickness of the flanges of the beams. However, we have investigated a research question about the impact of overdesigning the connection plates on the behavior of the entire frame described by its deformations and stresses. It is obvious that overdesigning of connection plates is easier and less expensive than overdesigning the entire frame. Therefore, any positive results of the reported analysis would not only improve our understanding of the behavior of a steel frame under blast but also help develop better design guidelines resulting in more economic design of steel frames under blast. Table 2 provides the relative thickness of the analyzed connection plates with respect to the thickness of beam flanges.

3

ANALYTICAL RESULTS AND DISCUSSION

The research is still undergoing but few analytical results are presented in this paper. A typical in-plane global response of the investigated steel frame is shown in Figure 5 at the end of the model run which is 0.4 seconds. It reveals the nature of the frame’s elastoplastic deformations with large plastic deformations located near the blast zone and ground floor columns exhibiting deformation characteristics for the ground floor plastic sway collapse mechanism. In the conducted analysis, the maximum horizontal displacements (sways) have been determined for the connections located at the right end of the top of the first floor. The maximum horizontal displacement of a structure is a good measure of deformations of the

Model runs and relative connection thicknesses.

Case

Conn. type

Conn. plate relative thickness

1 2 3 4 5 6

Standard Standard TA TA SidePlateTM SidePlateTM

1 2 1 2 1 2

Figure 5.

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Typical in-plane global response.

The study showed that a SidePlateTM connection type, that strengthens column and beam ends, provides a better performance against blast loading when compared to a standard connection. The TA connection type, which included horizontal plates intended to shift plastic zones away from columns, also performs better as compared to standard connections. The study also showed that doubling the connection plate thicknesses reduces in-plane although it will entail additional cost. Cost comparisons were not considered in the study. Figure 6. Displacement comparison for frames with connection plate relative thicknesses of 1.

ACKNOWLEDGMENT The authors gratefully acknowledge the support of Dr. Rainald Löhner, the Director of the Center for Computational Fluid Dynamics at George Mason University. REFERENCES

Figure 7. Displacement comparison for frames with connection plate relative thicknesses of 2.

entire structural system, particularly in the case of steel skeleton structures. Figure 6 and Figure 7 represent preliminary results of sway histories for the frames for all considered cases. There is a qualitative difference between the standard connection and the integrated connections in terms of the connection plate thickness impact on the sway. In the case of the standard connection, doubling the connection plate thickness reduces sway by 14% while in the case of the TA connection the reduction is about 16% and for the SidePlateTM connection is 20%. The results clearly demonstrate that overdesigning plates in the integrated connections reduce deformations of the frame, as measured by the sway.

4

CONCLUSIONS

This paper presents preliminary results of a finite element analysis of steel frames with different connection types subjected to blast. Understanding the behavior of connections is critical in keeping the structural continuity of steel frames during blast scenarios. Connections must provide the robustness and essential structural linkage across a column that is needed to prevent failure and progressive collapse.

ABAQUS Inc. 2004. Users’s manual version 6.5. Providence, RI. Chen, H. and Richard Liew, J.Y. 2005. Explosion and fire analysis of steel frames using mixed element approach. J. of Eng. Mech. 131(6): 606–616. Conrath, E.J., Krauthammer, T., Marchand, K.A. & Mlakar, P.F. 1999. Structural design for physical security: state of the practice. Reston, VA: ASCE. Cook, R.D., Malkus, D.S., Plesha, M.E. & Witt, R.J. 2002. Concepts and applications of finite element analysis. New York, NY: John Wiley and Sons. Department of the Army. 1990. Structures to resist the effects of accidental explosions, technical manual TM5-1300. Washington, DC. Department of Defense (DOD). 2003. Unified Facilities Criteria (UFC 4-010-01) DoD Minimum Antiterrorism Standards for Buildings. Washington, DC. Hamburger, R. & Whittaker A. 2004. Design of steel structures for blast-related progressive collapse resistance. Proc.NASCC conference, Long Beach, CA. Houghton, D.L. & Karns, J.E. 2003. SidePlateTM steel frame connection technology. Proc. 73rd shock and vibration symposium. San Diego, CA. IBC. 2003. International Building Code. Country Club Hills, IL: International Code Council. Krauthammer, T. & Cipolla, J. 2007. Building blast simulation and progressive collapse analysis. Proc. 11th NAFEMS world congress. Vancouver, Canada. Löhner, R., Yang, C., Cebral, J., Soto, O., Camelli, F., Baum, J.D., Luo, H., Mestreau, E. & Sharov D. 2001. Advances in FEFLO. Proc. 40th AIAA Aerospace Sciences Meeting and Exhibit. Reno, NV. National Research Council (NRC). 1995. Protecting People from Bomb Damage: Transfer of Blast-Effects Mitigation from Military to Civilian Applications. Washington, DC: National Academy Press. Sabuwala, D.L., Linzell, D. & Krauthammer, T. 2005. Finite element analysis of steel beam to column connections subjected to blast loads. Int. J. of Impact Eng. 31: 861–876.

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Bridges and special structures

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Applicability of AASHTO LRFD live load distribution factors for nonstandard truck load Y.J. Kim Department of Civil Engineering, North Dakota State University, Fargo, ND, USA

R. Tanovic & R.G. Wight Department of Civil Engineering, Royal Military College of Canada, Kingston, ON, Canada

ABSTRACT: This paper presents the applicability of live load distribution factors provided by the American Association for State Highway and Transportation Officials Load and Resistance Factor Design (AASHTO LRFD) Specifications for the evaluation of capacities of civilian bridges when standard military vehicles are applied. The standard military vehicles are prescribed by the NATO Standardization Agreement 2021 (STANAG 2021). Three-dimensional finite element analysis models are developed to examine the AASHTO LRFD distribution factors for the wheeled Military Load Classification (MLC) truck loads. A total of 27 different loading scenarios are considered to evaluate the LRFD formulas for a typical steel I-girder bridge having a loading span of 36 m with three traffic lanes. The AASHTO LRFD formulas for lateral load distributions are conservative up to approximately 30% for this particular bridge.

1

INTRODUCTION

An adequate evaluation of the load-carrying capacity of a bridge is an important issue to determine the allowable traffic loads on a bridge. The design and construction of modern highway bridges are based on the load effect of standard design trucks. However, constructed bridges may be subject to nonstandard truck loads such as military trucks or logging-industry trucks. The flexural response of a bridge under such nonstandard loads may be different from that under the standard truck load. Inadequate predictions of the load effect may cause serious damage in the bridge. Because the traffic loads in bridges are not uniformly distributed to the individual components of the superstructure, one of the first steps to evaluate the capacity of the bridge superstructure to carry nonstandard vehicles is to make a realistic assessment of the lateral load distribution to the structural components such as main girders. Numerous research efforts have been made to predict the load distributions on a bridge (Zokaie 2000; Song et al. 2003; Kim et al. 2008). Zokaie et al. (1991) conducted an intensive survey of over 800 bridges across the United States and developed design formulas to predict the lateral load distributions. The American Association of State Highway and Transportation Officials (AASHTO) adopted the developed formulas for a new version of the bridge code called the Load Resistance Factor Design (LRFD) Bridge Design Specifications (AASHTO 1994, 2007). The applicability of the design provisions should be

371

reviewed when a bridge is subjected to nonstandard truck loads, given that the formulas have been developed based on the standard trucks. This paper examines the applicability of the live load distribution factors of the AASHTO LRFD to nonstandard military truck loads.

2 2.1

BACKGROUND Live load distribution

Distribution of axle loads in slab-on-girder bridges is significantly influenced by the configuration of truck axles and wheels as well as the structural characteristics of the main structural components such as the span length, girder spacing, girder stiffness, etc. Three methods are typically used in North America to predict the lateral load distributions, including the AASHTO Standard, the AASHTO LRFD, and the Canadian Highway Bridge Design Code (CHBDC). The AASHTO Standard (AASHTO 1996) uses a simple formula called the S/D method. According to the S/D formula, the load distribution is simply based on the girder spacing (S) and the characteristic constant (D) that depends on the type of superstructure (e.g., D = 5.5 for concrete slab on steel girders, S is in feet). No other parameters are considered. Zokaie (2000) reported that the S/D formulas were acceptable for bridges having girder spacing near 6 ft. and span length close to 60 ft., whereas the predictability of the

formulas was not accurate when the properties of a bridge were varied. Kocsis (2004) also reported the inaccuracy of the S/D method, based on an analysis program using the semicontinuum method. The AASHTO LRFD (AASHTO 1994, 2007) specifications provide more advanced format of distribution formulas, as typically shown in Eq. 1.      S 0.6 S 0.2 Kg 0.1 = 0.075 + 2900 L Lts3

girder determined by sharing equally the total moment on the bridge cross section among all girders in the cross section. Further details are available elsewhere (CSA 2006). The live load distribution from a refined analysis can be calculated using Eq. 4.



LDF

MI

LDF = (1)

where LDF MI is the load distribution factor for exterior girders subject to two or more (multiple) design lanes, S is the girder spacing (mm), L is the span length (mm), ts is the slab thickness (mm), and Kg is the stiffness term consisting of a modula ratio between the girder and the deck, moment of inertia and area of the girder, and girder eccentricities. The distribution for exterior girders is based on the Lever Rule and the Rigid Method. The Lever Rule means the statical summation of moments about one point to calculate the reaction at a second point (AASHTO 1994). The load distribution factor is the reaction divided by the applied load. This simple method based on mechanics does not account for the properties of a bridge (i.e., only girder spacing and truck locations are taken into account). The AASHTO Standard specifications also use the Lever Rule for exterior girders. The Rigid Method is shown in Eq. 2. R=

 NL Xext NL e + N b 2 Nb x

(2)

where R is the reaction on an exterior beam in terms of lanes, NL is the number of loaded lanes under consideration, e is the eccentricity of a design truck or a design lane load from the center of gravity of the pattern of girders (mm), x is the horizontal distance from the center of gravity of the pattern of girders to each girder (mm), xext is the horizontal distance from the center of gravity of the pattern of girders to the exterior girder (mm), and Nb is the number of girders. The Rigid Method may overestimate the live load effect because the method does not include the transverse and torsional stiffness of a superstructure (Tobias et al. 2004). The CHBDC (CSA 2006) assumes a uniform distribution of the applied load effect and applies amplification factors to allocate the live load effect to the exterior and interior girders, as shown in Eq. 3. Mg = Fm Mg−average

(3)

where Mg is the maximum bending moment per girder due to live load, Fm is the amplification factor to account for the transverse variation of the moment, and Mg−average is the average bending moment per

Mrefined Mbeamline

(4)

where Mrefined is the maximum bending moment that a girder may experience and Mbeamline is the largest bending moment obtained from a simple beamline analysis with a single lane of traffic. The live load distributions can also be obtained by the measured values from load testing using Eq. 5 (Klaiber et al. 2001). δi LFi = α n j

δj

(5)

where LF i is the load fraction factor for ith girder, α is the number of wheels, δi is the deflection of ith girder, δj is the deflection of jth girder, and n is the total number of girders. 2.2

Effect of nonstandard truck load

Although the flexural response of a bridge subject to nonstandard trucks is an important issue, very limited information is available. Heavy truck loads that have not been considered in a bridge design may affect the performance of existing bridges in terms of load-carrying capacity (Du & Han 2008). Papavizas & Kostem (1985) reported that deflections and midspan bending moments of the bridges subject to nonstandard vehicles increased up to 208% and 199%, respectively, in comparison to those under the standard truck. Wide wheel-line spacing of nonstandard trucks may result in different load distributions in comparison to those under standard trucks. According to a field test combined with a refined model (Bridge Diagnostics 1999), the lateral load distributions of tested bridges using a Heavy Equipment Transporter System (HETS) truck were significantly different from those under the HS20 truck. Keating et al. (1995) proposed a reduction factor to adjust the different load effects induced by nonstandard trucks.

3 3.1

NONSTANDARD MILITARY TRUCK LOAD Military load classification

STANAG 2021 (STANAG 2002) provides standard classifications of military trucks, namely, the Military Load Classification (MLC) series that consist of tracked and wheeled design vehicles. For this study,

372

62.3 kN

Weight and wheel-line spacing

145 kN

145 kN

35 kN

115.65 kN 115.65 kN

MLC-40 3660

Truck

Weight (kN)

Wheel-line (mm)

HS-20

325

1800

214

4300

249

2150

MLC-30

303

2100

MLC-40

418

2100

MLC-50

516

2050

MLC-60

623

2380

MLC-70

716

2320

MLC-80

819

2260

4880

4300 71.2 kN

Standard 75.6 kN 75.6 kN

26.8 kN

2150

MLC-24

1220

Standard

35.6 kN

MLC-20

124.6 kN

133.53 kN 133.53 kN

177.94 kN

MLC-50 3660

1220

4880

MLC-20 3050

1220

3660 71.2 kN

44.5 kN

88.95 kN

26.8 kN

88.95 kN

160.2 kN

160.2 kN

115.65 kN 115.65 kN

MLC-60 3660

1520

4570

122

MLC-24 3050

1220

3660 93.4 kN

53.4 kN

97.85 kN

97.85 kN 53.4 kN

186.88 kN 186.88 kN

124.57 kN 124.57 kN

1520

122

MLC-70 3660

4570

MLC-30 3050

1220

3660 106.8 kN 213.56 kN 213.56 kN

142.34 kN 142.34 kN

MLC-80 3660

Figure 1.

1520

5490

1520

Details of the MLC wheeled trucks.

the wheeled vehicles are considered. Typical information of the MLC wheeled trucks is shown in Figure 1, including the weight and the wheel-line and axle spacing. The weight of the MLC wheeled trucks varies from 66% to 252% when compared to that of the standard truck (HS20 weighing 325 kN). The number of axles is more than that of standard truck and the wheelline spacing varies from 14% to 32% greater than the 1.8 m spacing of HS20. The military trucks are well controlled during the operation, thus spacing between the trucks is assumed to be 0.5 m (STANAG 2002).

380

13420

380

205

(a)

(b)

(c) 3.2

Load combination

The MLC wheeled trucks may be positioned to generate the maximum bending moment in a bridge. Multilanes of the MLC trucks may be operated if necessary (e.g., emergency situation during war). Typical load combinations for this study are shown in Figure 2. To provide an adequate evaluation of the AASHTO LRFD provisions on the live load distribution (AASHTO 1994, 2007), two cases of load combinations are considered herein such as single-lane and two-or-more-lane loaded. The effect of dynamic load allowance was not taken into account. The focus of this paper is on the interior girders subject to various types of MLC trucks, as shown in Figure 2. In particular, the trucks were loaded to induce the maximum stress of girder 3 (Fig. 2). The identification code to represent the loading cases (Table 1) includes the truck type (IS = standard truck and INS = nonstandard truck), class of the MLC truck, and number of the loaded lanes. For instance, a load case of IS-2 means 2 lanes loaded with two standard trucks (e.g., Fig. 2b), whereas INS-40-3 indicates three lanes of MLC40 trucks (e.g., Fig. 2c).

1 990

2

3 4 5 5@2440 = 12200

6 990

14180

Figure 2. Load combination of MLC trucks (unit: mm): (a) one-lane loaded; (b) two-lane loaded; (c) three-lane loaded.

4

BRIDGE DETAILS

A steel plate girder bridge was designed to simulate the MLC load effect, based on Barker and Puckett (1997). The designed bridge was simply supported and was 36 m long and 14.2 m wide (3 traffic lanes). The bridge consisted of six I-girders with a height of 1,540 mm (i.e., top flange = 400 mm × 15 mm, web = 1500 mm × 10 mm, and bottom flange = 400 mm × 25 mm) and a deck thickness of 205 mm with a hunch depth of 25 mm, as shown in Figure 2. Full composite action was assumed between the girders and the deck. The concrete strength was 30 MPa with a modulus of

373

Table 1.

Live load distribution factors (interior girder).

Loading case IS-1 IS-2 IS-3 INS-W20-1 INS-W20-2 INS-W20-3 INS-W24-1 INS-W24-2 INS-W24-3 INS-W30-1 INS-W30-2 INS-W30-3 INS-W40-1 INS-W40-2 INS-W40-3 INS-W50-1 INS-W50-2 INS-W50-3 INS-W60-1 INS-W60-2 INS-W60-3 INS-W70-1 INS-W70-2 INS-W70-3 INS-W80-1 INS-W80-2 INS-W80-3 Average error of HS20 Average error of MLC

LRFD (a)

FEA (b)

Error (%) (a−b)/a

0.348 0.605 0.712 0.348 0.605 0.712 0.348 0.605 0.712 0.348 0.605 0.712 0.348 0.605 0.712 0.348 0.605 0.712 0.348 0.605 0.712 0.348 0.605 0.712 0.348 0.605 0.712

0.275 0.495 0.628 0.283 0.478 0.610 0.287 0.479 0.611 0.285 0.476 0.610 0.279 0.467 0.601 0.271 0.489 0.624 0.271 0.497 0.603 0.271 0.426 0.541 0.272 0.458 0.576

21.0 18.2 11.8 18.7 21.0 14.3 17.5 20.8 14.2 18.1 21.3 14.3 19.8 22.8 15.6 22.1 19.2 12.4 22.1 17.9 15.3 22.1 29.6 24.0 21.8 24.3 19.1 17.0% 19.5%

A perfect connection between the structural members was assumed to represent the full composite action as was designed. Detailed material properties are shown in the previous section. The solved FEA results (stresses from the elements) were converted to equivalent moments using structural analysis. To determine the live load distribution factors, Eq. 4 was used. The AASHTO LRFD provisions were then compared to the refined analysis results. It should be noted that the effect of multiple presence factor was not included in the comparison. 5.2

Model validation

The accuracy of the FEA model was validated using a conventional structural analysis method. A threeaxle standard truck (325 kN) was loaded on the designed bridge where the maximum bending moment occurred and the deflections were compared, as shown in Figure 4. The agreement was satisfactory with a maximum error of less than 3%.

6

APPLICABILITY OF THE AASHTO LRFD PROVISION

A comparison between the AASHTO LRFD provision and the refined analysis results is made in Table 1. It should be noted that only governing load factors that represent the load factors of 6 girders are shown in Table 1. The LRFD predictions for MLC trucks were

elasticity of 26 GPa. The yield strength of the steel was 350 MPa with an elastic modulus of 200 GPa. The Poisson’s ratios were assumed to be 0.25 and 0.3 for the deck and the girders, respectively.

5 5.1

MODEL DEVELOPMENT Figure 3.

Developed FEA model.

Figure 4.

Validation of the FEA model.

Finite element analysis

A refined analysis using the finite element package ANSYS was conducted to examine the response of the bridge (Fig. 2) under the standard and nonstandard truck loads, rather than a simple grillage method. The 3-dimensional bridge model is shown in Figure 3. An elastic analysis was performed, given that the bridge may not experience any inelastic behavior under the loading range provided by the trucks considered herein. Concrete cracking in the deck was ignored. The bridge model included 4-node shell elements (SHELL 63), consisting of 6 degrees of freedom per node, to represent the deck and the girders. The cross bracings were represented by 3-dimensional spar element (LINK 8) to provide stability of the girders.

374

in general conservative with an absolute average error of 19.5% that is 14.7% higher than the error under the standard truck of 17.0%. On the contrary to the AASHTO LRFD classifications (i.e., one-lane and multiple lane loaded), the refined analysis showed constant increases in load distribution factors when more trucks were loaded, as shown in Table 1. The greatest error between the LRFD and the FEA was found in the case of two-lane loaded and the smallest error was under the three-lane loaded cases. This is attributed to the fact that the AASHTO LRFD provision do not distinguish the load effect from two-lane and three-lane loaded cases, whereas the multiple presence

factors (1.0 for two-lane and 0.85 for three-lane) in the formula that have been extracted influenced the comparison. Figure 5 shows typical profiles of the distribution factors across the critical section. For the singlelane loading case, there was no notable difference in the response under the trucks (Fig. 5a), whereas a trend showing reduced distribution factors was observed when more lanes were loaded. This is due to the fact that more girders shared the applied load effect.

7

CONCLUDING REMARKS

This paper has presented the applicability of the AASHTO LRFD provisions to predict the live load distributions on a girder-type bridge, based on the calibrated 3-dimensional FEA models. The AASHTO LRFD approach provided conservative predictions of the load distributions for the military trucks within the error range of 14.2% to 29.6% for this particular I-girder bridge. The on-going research includes an expansion of the current model to cover various loading spans, different sizes of the girders, and more loading combinations to examine the nonstandard load effect for the interior and exterior girders. An improved predictive design equation will be proposed.

(a) ACKNOWLEDGMENTS The authors gratefully acknowledge the financial support of the Natural Sciences and Engineering Research Council of Canada (NSERC) to the first author.

REFERENCES

(b)

(c) Figure 5. Profile of load distribution factors across the critical section: (a) one-lane loaded; (b) two-lane loaded; (c) three lane loaded.

375

AASHTO. 1994. AASHTO LRFD bridge design specifications, 1st ed. American Association of State Highway and Transportation Officials, Washington, D.C. AASHTO. 1996. AASHTO standard specifications for highway bridges, 16th ed. American Association of State Highway and Transportation Officials, Washington, D.C. AASHTO. 2007. AASHTO LRFD bridge design specifications, 4th ed. American Association of State Highway and Transportation Officials, Washington, D.C. Bridge Diagnostics Inc. 1999. Bridge response investigation: U.S. army heavy equipment transporter system (HETS). Report, New Mexico State University, Engineering Research Center. Barker, R.M. and Puckett, J.A. 1997. Design of highway bridges based on AASHTO LRFD bridge design specifications. John Wiley & Sons, Inc., New York, NY. CSA. 2006. Canadian highway bridge design code (CHBDC): CAN/CSA S6-06. Canadian Standard Association, Toronto, ON, Canada. Du, J. & Han, D.J. 2008. Bridge safety analysis considering heavy truck loading, Intern. Conf. on Bridge Maintenance, Safety, Health Monitoring and Information (IABMAS08), Seoul, Korea: 2442–2449.

Keating, P.B., Litchfield, S.C., & Zhou, M. 1995. Overweight permit rules. Texas Transportation Institute, College Station, Texas. Kim, Y.J., Green, M.F., & Wight, R.G. 2008. Live load distributions on impact-damaged prestressed concrete girder bridge repaired using prestressed CFRP sheets. Journal of Bridge Engineering, ASCE, 13(2): 202–210. Klaiber, F.W., Wipf, T.J., Nahra, M.J., Ingersoll, J.S., Sardo, A.G., & Qin, X. 2001. Field and laboratory evaluation of precast concrete bridges. Iowa Department of Transportation, Final Repost, Project No.TR-440, Ames, Iowa. Kocsis, P. 2004. Evaluation of AASHTO live load and line load distribution factors for I-girder bridge decks. Practice Periodical on Structural Design and Construction, ASCE, 9(4): 211–215. Papavizas, P. & Kostem, C.N. 1985. Structural response of simple span bridges to nonstandard vehicles. The 2nd Annual Intern. Bridge Conf., Pittsburgh, PA: 184–188.

Song, S.T., Chai, Y.H., & Hida, S. 2003. Live-load distribution factors for concrete box-girder bridges. Journal of Bridge Engineering, ASCE, 8(5): 273–280. STANAG .2002. Standardization agreement (STANAG) 2021, 6th ed. NATO Standardization Agency. Tobias, D.H., Anderson, R.E., Khayyat, S.Y., Uzman, Z.B., & Riechers, K.L. 2004. Simplified AASHTO load and resistance factor design girder live load distribution in Illinois. Journal of Bridge Engineering, ASCE, 9(6): 606–613. Zokaie, T., Osterkamp, T.A., and Imbsen, R.A. 1991. Distribution of wheel loads on highway bridges. NCHRP 12-26/1 Final Report, National Cooperative Highway Research Program, Washington, D.C. Zokaie, T. 2000. AASHTO-LRFD live load distribution specifications. Journal of Bridge Engineering, ASCE, 5(2): 131–138.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Full scale test on a bridge PC box girder C. Mircea National Building Research Institute [INCERC], Cluj-Napoca Branch, Cluj-Napoca, Romania

A. Ioani & Z. Kiss Technical University of Cluj-Napoca, Cluj-Napoca, Romania

ABSTRACT: The paper presents the largest destructive test, ever performed in Romania on a full scale PC prefabricated bridge box girder (22.45 × 3.28 × 37.10 m), made for the Transylvania Motorway, Romania. Experimental data is obtained through an advanced acquisition technique. The general frame and testing conditions are comprehensive presented. The testing procedure is based on the design documentation, the Romanian specification for bridge girder testing, and a preliminary non-linear analysis of the concrete member. Results are presented in a synthetic manner using graphical figures resulted from digital processing. The paper includes commentaries and comparisons regarding the structural response in the elastic and post-elastic stage of the unit, as well as important conclusions for designers and precasters. 1

INTRODUCTION

The paper deals with the destructive test performed on a 37.1 m span PC box girder, in order to certify the new type of girder for the use at the Transylvania motorway bridges. The girder was manufactured by Bechtel International Inc. Reno-Nevada, Cluj-Napoca Subsidiary, in accordance with the design completed by Iptana SA Bucharest. The test was performed at the site of the producing plant, between 15–18 of September 2008, the mean temperature being 11◦ C and the average relative moisture of 80%. Figures 1 and 2 present aspects from the test. The PC box girder was assembled by a U prefabricated unit and a top slab of 25 cm thickness made of Figure 2.

Details of the full scale test setup.

cast in situ concrete, as shown in Figure 3. Prestressing was induced by 84 strands T 15.2 mm positioned on the tensioned area of the girder, and 4 strands T 6.35 mm placed in the compressed area. The specimen age at testing was four month. 2

Figure 1.

TESTING PROCEDURE AND EQUIPMENT

Girder 96–41 was gradually loaded and unloaded in 6 cycles. The last cycle included only the loading steps up to the predicted failure load, and was continued until the member collapsed. Concentrated loads were applied in four sections and were divided in eight punctual loads, as shown in Figure 3. Cycles C1 and C2 correspond to the level

Box girder (ID 96-41) on the testing stand.

377

Figure 3.

Table 1.

Basic geometry of the girder, control and loading sections.

Cycles, loading stages and load increments.

Cycles C1 C1 C1 C1 C1 C1 C1 C1 − − − − − − − − − − − − − −

C2 C2 C2 C2 C2 C2 C2 C2 − − − − − − − − − − − − − −

C3 C3 C3 C3 C3 C3 C3 C3 C3 C3 C3 C3 C3 C3 C3 − − − − − − −

C4 C4 C4 C4 C4 C4 C4 C4 C4 C4 C4 C4 C4 C4 C4 C4 C4 C4 C4 − − −

C5 C5 C5 C5 C5 C5 C5 C5 C5 C5 C5 C5 C5 C5 C5 C5 C5 C5 C5 − − −

C6 C6 C6 C6 C6 C6 C6 C6 C6 C6 C6 C6 C6 C6 C6 C6 C6 C6 C6 C6 C6 C6

of the service load, the next cycle (C3 ) corresponds to the cracking load, cycles C4 and C5 considered the design load, and in the last one, the load was increased up to failure. Loading steps corresponding to cycles C1 –C6 are show in Table 1. During unloading, only half of the loading steps were considered. Loading step 0 corresponds to the weight of the loading equipment, and loading step 26 corresponds to the predicted failure load, considering the characteristic values of the materials strength.

Loading step

P kN

P kN

0 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 . . . 17 18 19 20 . . . 26 27 . . . 30 31

– 47.50 38.25 21.63 21.63 21.63 21.63 21.63 45.28 45.28 45.28 45.28 15.10 15.10 15.08 52.70 2×50.00 49.00 54.35 7×107.41 4×50.00 42.25

62.50 110.00 148.25 169.88 191.51 213.14 234.77 256.40 301.68 346.96 392.24 437.52 452.62 467.72 482.80 535.50 585.50 . . . 635.50 684.50 738.85 846.26 . . . 1490.75 1540.75 . . . 1690.75 1733.00

Monitoring was done by real time acquisition digital equipment and mechanical devices (see Fig. 2). The monitored parameters were: deflections, cracking load, cracks closure and re-opening load, cracks spacing, cracks width, bonding of the prestressed strands and failure load. Forces were applied through two hydraulic jacks of 5000 kN located on a steel beam supported at the ends (see Figures 1 and 2). Loads were controlled by two force transducers C6A/5000 kN (placed between the hydraulic jacks and the steel

378

Figure 4.

Displacement and settlement monitoring devices in the control sections.

Figure 5.

Strain and slip monitoring devices.

beams. One force transducer was connected to a portable electronic device Scout 55, and the other one to a channel of a Spider 8 electronic PC parallel measurement unit. Displacements and end settlements were measured by 10 inductive transducers (T) HBM-WA/300 mm linked to the same Spider 8 unit, and 14 dial gauge mechanical devices (F). Figure 4 shows the equipment setup. Half-bridge LY 41-11/150 strain gages (B) were used to asses the stress state of concrete in the mid-span cross-section and in inclined end sections (see Figure 5). Compensation was done by strain gauges attached to witness concrete specimens. End slips were measured by for dial rigid mechanical devices (S). Crack widths were measured with field microscopes, and crack heights were noticed every two loading steps after cracking. Digital recording was performed up to the loading step 25 (i.e., the level of the load P = 1386.04 kN). After that, in order to prevent the destruction of the equipment, all digital

sensors were removed and only mechanical devices were used for subsequent steps. Application of the force was controlled through the pressure dial gauge of the electric pump.

3 3.1

DESIGN SPECIFICATIONS AND PRELIMINARY NON-LINEAR ANALYSIS Material properties

Design concrete classes were C 35/45 for the precast U unit and C 25/30 for the top cast in situ slab. Laboratory tests revealed that the effective strength of concrete at 28 days was fc,cube = 50.0 MPa (i.e., fc,cil = 38.5 MPa) for the prefab U unit, and fc,cube = 41.0 MPa (i.e., fc,cil = 32.3 MPa) in the top slab. Transfer of the prestressing forces to concrete was designed to be made for a minimum concrete strength

379

fci,cube = 41.5 MPa. Passive longitudinal and shear reinforcement was made of high ductility steel, with the yielding strength fy = 255.0 MPa, yielding strain εy = 1.21%, tensile strength of 318.8 MPa and its associated elongation of 20%. Prestressed strands were made of high strength prestressing steel of low relaxation class, with fpy = 1636.0 MPa, tensile strength fpu = 1860.0 MPa and its associated strain 2%. Bonding was interrupted in two sections, resulting a longitudinal layout with three significant cross-sections (i.e., margin, middle and central segments). The control prestressing stress was 1440.0 MPa. Table 2 presents the total reinforcing ratios in the three significant longitudinal segments of the box girder. Table 2.

Reinforcing ratios for the box girder segments.

Girder segment

Prestressed strands ρp

Longitudinal reinforcement ρ

Shear reinforcement ρw

Margin Intermediary Central

0.0032 0.0046 0.0054

0.0050 0.0052 0.0052

0.0057 0.0057 0.0057

3.2

Design values

Figure 6 shows the stress state in the control crosssection C before testing. Tables 3 and 4 show the analytical predicted values of the displacements and stresses for several loading steps. The limit crack width of 0.1 mm, related to the Serviceability Limit States, corresponds to the characteristic load level Pcrack = 482.80 kN. The computed failure load is Pfailure = 1490.76 kN. Table 4.

Concrete stress state in the control cross-section C. Stresses due to P

Total stresses

Loading P step kN

Top MPa

Bottom MPa

Top MPa

Bottom MPa

2 3 4 5 6 7 8 9 10 11 12 13 14

+1.587 +1.818 +2.050 +2.281 +2.513 +2.745 +3.229 +3.714 +4.199 +4.683 +4.845 +5.007 +5.168

−3.045 −3.489 −3.933 −4.738 −4.822 −5.266 −6.196 −7.126 −8.056 −8.986 −9.296 −9.606 −9.916

+16.714 +16.945 +17.177 +17.408 +17.640 +17.872 +18.356 +18.841 +19.326 +19.810 +19.972 +20.134 +20.295

+4.516 +4.072 +3.628 +3.183 +2.739 +2.294 +1.365 +0.435 −0.495 −1.425 −1.735 −2.045 −2.355

148.25 169.88 191.51 213.14 234.77 256.40 301.68 346.96 392.24 437.52 452.62 467.72 482.80

45,000 40,000 35,000

Table 3.

Bending moment [kNm]

30,000

Figure 6.

Stress state in cross-section C prior loading. Displacements of the control cross-sections. Displacements

Loading step

P kN

δA = δE mm

δB = δD mm

C

2 3 4 5 6 7 8 9 10 11 12 13 14

148.25 169.88 191.51 213.14 234.77 256.40 301.68 346.96 392.24 437.52 452.62 467.72 482.80

6.59 7.55 8.51 9.48 10.44 11.40 13.41 15.42 17.44 19.45 20.12 20.79 21.46

8.98 10.29 11.60 12.91 14.22 15.53 18.27 21.02 23.76 26.50 27.41 28.33 29.24

9.29 10.65 12.00 13.36 14.71 16.07 18.91 21.74 24.57 27.42 28.37 29.31 30.26

25,000 20,000 15,000 10,000

Margin cross-section Intermediary cross-section Mid span cross-section

5,000 0 -1 0 -5,000

1

2

3

4

5

6

7

8

9

10

-10,000 -15,000 Curvature 10 [mm ]

Figure 7. Moment-curvature relationships for three characteristic cross-sections.

380

3.3

Non-linear analysis

A non-linear FE analysis was done before testing, in order to get supplementary information in preparing the test. FE analysis was done based on the procedure developed by Mircea & Petrovay 2005 for the reference cross-sections, considering the initial strain concept for the initial prestressing state of the strands, and following a traditional Newton-Raphson incremental scheme. The actual properties (i.e., cylinder strengths based on the cubic strengths given by laboratory tests on sample specimens) of concrete were implemented in the CEB-FIB constitutive model. Bilinear relations were introduced for the characteristic properties of steel. Figure 7 shows the typical moment-curvature relationships for the three segments of the box girder, implemented in the FE analysis. Table 5.

Tables 5 and 6 present the displacements and stresses for the relevant service loading steps, resulted from the FE analysis. Non-linear FE analysis revealed a cracking load Pcrack = 545.0 kN and an ultimate load Pfailure = 1822.5 kN. The predicted failure occurs by crushing of concrete after the passive reinforcement yields, and the ultimate deflection was found to be 585.8 mm.

FEA displacements of the control cross-sections. Displacements P kN

δA = δE mm

δB = δD mm

C mm

2 3 4 5 6 7 8 9 10 11 12 13 14

148.25 169.88 191.51 213.14 234.77 256.40 301.68 346.96 392.24 437.52 452.62 467.72 482.80

2.9 4.3 5.1 5.7 6.3 6.8 7.8 7.4 8.3 9.3 9.6 9.9 10.2

4.1 6.0 7.2 8.1 8.8 9.5 11.0 10.4 11.7 13.0 13.5 13.9 14.4

4.3 6.0 7.2 8.1 8.9 9.6 11.2 10.7 12.0 13.4 13.9 14.3 14.8

Table 6. FEA concrete stress state in control crosssection C.

P kN

Top MPa

Bottom MPa

2 3 4 5 6 7 8 9 10 11 12 13 14

148.25 169.88 191.51 213.14 234.77 256.40 301.68 346.96 392.24 437.52 452.62 467.72 482.80

+12.90 +13.31 +14.16 +14.30 +14.42 +14.56 +14.67 +14.81 +15.06 +15.31 +15.62 +15.92 +16.06

+3.76 +2.31 +2.11 +1.90 +1.66 +1.45 +1.24 +0.79 +0.37 −0.13 −0.67 −0.92 −1.09

4.1

Displacement state

The camber of the member was 11 mm before starting the testing procedure. Figure 8 presents a comparison of the experimental displacements and the theoretical ones (Exp—test, FEA—finite element analysis, Design—analytically predicted). As exposed by Figure 8, there is a fine correspondence between the theoretical values and experiment. Thus, it can be concluded that the member has a predictable behavior which can be adequately modeled by current design techniques. Table 7 presents the experimental values of the displacement registered under the service loading and unloading cycles. The total elastic deflection 2000 1800 1600 1400 1200 1000 800

Total stresses in concrete Loading step

RESULTS AND COMMENTARIES

P [kN]

Loading step

4

600 400 200 0 0

50 100 150 200 250 300 350 400 450 500 Displacement [mm] Exp-A Exp-B Exp-C Exp-D Exp-E FEA-C FEA-B,D FEA-A,E Design-A,E Design-B,D Design-C

Figure 8. Experimental and theoretical load-displacement curves in the control sections.

381

after the service cycles (i.e., cycle C2 ) is 9.97 mm, less than the analytically predicted, of 16.07 mm, and the difference in respect with the numerically determined one, of 9.60 mm, is well beyond 10 %. After the same service cycles, the maximum service deflection is 4.84 mm, much less than the admissible deflection (i.e., span/800 = 45.1 mm). After cycle C2, it results

Table 7. cycles.

Experimental displacements under the service δA mm

Characteristic displacements

δB mm

C mm

δD mm

δE mm

Cycle C1 Under service load 3.42 4.92 5.06 5.00 3.41 Residual 0.32 0.56 0.55 0.53 0.30 Elastic under total load 7.01 9.71 9.92 9.53 6.94 Cycle C2 Under service load 3.23 4.82 4.84 4.78 3.25 Residual 0.44 0.77 0.76 0.70 0.47 Elastic under total load 7.03 9.71 9.97 9.63 6.90

Table 8.

Experimental displacements at failure.

Loading step

P kN

δA mm

δB mm

C mm

δD mm

δE mm

31

1733.0

295.0

383.3

403.5

381.3

290.5

a residual coefficient of 0.02, less than the standard limit value equal with 0.10. Table 8 presents the displacements registered at the last loading step, considered to correpond to the conventional failure. The recorded failure deflection is 403.5 mm, which corresponds to a deflection/span ratio of 1/89.5 (standard criterion is 1/100). 4.2

Cracking state

Under the service load level (i.e., loading step 7 with P = 256.40 kN ) cracks were closed at all loadingunloading cycles. On the loading branch of cycle C3 , up to the load corresponding to the admissible crack width (i.e., 0.1 mm) considered in design, three cracks occurred (i.e., w = 0.02 mm) at the loading step 12, under P = 452.62 kN. Under P = 482.80 kN (i.e., the maximum step for cycle C3 ) the maximum cracks width reached 0.04 mm. The cracks closed at the unloading (i.e., the descending) branch of cycle C3 when P = 437.52 kN . Re-opening of the cracks in cycle C4 took place for P = 392.24 kN , but the same maximum cracks width was recorded for P = 482.80 kN . Figure 9 shows the final cracking pattern. The maximum crack width in the central

Figure 9.

Table 9.

Cracking pattern of box girder 96–41 up to failure.

Test strains recorded during service cycles on the control cross-section C. Loading step

P kN

B0 μm/m

B1 μm/m

B2 μm/m

B3 μm/m

B4 μm/m

B5 μm/m

C1

1 2 3 4 5 6 7 5 3 1 −

110.00 148.25 169.88 191.51 213.14 234.77 256.40 213.14 169.88 110.00 62.50

−8.9 −14.2 −18.6 −27.0 −37.2 −37.9 −47.4 −53.2 −37.0 −14.6 −0.1

−4.5 −9.8 −11.8 −15.7 −18.0 −19.6 −21.3 −23.4 −18.1 −7.2 −0.1

0.7 1.3 1.5 3.1 7.3 7.4 11.3 9.7 5.6 1.5 1.1

6.2 13.0 17.8 19.7 22.1 22.3 24.1 31.2 16.6 9.2 3.0

11.5 21.8 29.3 33.6 39.0 38.9 43.4 52.8 28.4 15.0 3.8

12.6 24.6 32.0 37.3 43.9 44.3 49.4 59.5 33.6 18.4 5.8

C2

1 2 3 4 5 6 7 5 3 1 −

110.00 148.25 169.88 191.51 213.14 234.77 256.40 213.14 169.88 110.00 62.50

−9.0 −14.3 −18.8 −27.2 −37.4 −38.1 −47.6 −53.3 −37.1 −14.8 −0.3

−4.6 −9.9 −11.9 −15.8 −18.2 −19.8 −21.4 −23.5 −18.2 −7.3 −0.2

1.8 2.4 2.6 4.2 8.4 8.5 12.4 10.8 6.7 2.6 2.2

9.3 16.1 20.8 22.8 25.2 25.4 27.2 34.3 19.6 12.3 6.1

15.4 25.7 33.1 37.5 42.9 42.7 47.3 56.7 32.3 18.9 7.7

18.4 30.4 37.8 43.1 49.7 50.1 55.2 65.3 39.4 24.1 11.5

Cycle

382

2

3

4

5

6

Loading step 7 8 9 10 11 12 13 14

6.0

Concrete stress [MPa]

4.0 2.0 0.0 -2.0 -4.0

4.4

Failure

As shown in Figure 9, cracks path indicates a failure type like the one predicted by the non-linear FE analysis (i.e., yielding of steel followed by crushing of concrete). As shown above, conventional failure load corresponds to P = 173.0 kN . Under this load level, the deflection reached the ratio deflection/span of 1/89.5, the maximum cracks width was 0.1 mm with a cracks spacing of 90 mm. The test failure load is higher than the failure load predicted by analytical models in design, and the safety ratio is 2.51, more than the allowable value of 1.5.

-6.0

ACKNOWLEDGEMENT -8.0

This test was carried out with the support of the Structural Division from the Romanian branch of Bechtel International Inc. Ltd and ICECON Bucharest. The authors wish to express appreciation for their support regarding this work.

-10.0 Test - top

Test - bottom

Analytic - top

Analytic - bottom

FEA - top

FEA - bottom

Figure 10. Theoretical and experimental stress states at in the control cross-section C.

REFERENCE

segment was 1.0 mm, and the average crack spacing is 90 mm. The maximum inclined cracks width was 0.4 mm.

Mircea, C. & Petrovay, G. 2005. New Approach in NonLinear Sectional Analysis of RC and PC Members. In Proceedings of FIB International Symposium ‘‘Keep Concrete Attractive’’, vol. II, Budapest–Hungary: 736–741.

4.3

Strain state, stress state and ends bonding

Table 9 shows the strain registered in the control crosssection C at the service cycles (i.e., C1 and C2 ), and Figure 10 presents a comparison between theoretical stress states and the stresses resulted on the ground of the strains registered during loading branches of the test cycles C1 –C3 . No slips were recorded during the test between the strands and surrounding concrete. A close correspondence can be noticed and the same predictable behavior is found.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Long-term deflections of long-span bridges J. Navrátil SCIA CZ, Scientific Application Group, Brno, Czech Republic

M. Zich Department of Civil Engineering, Brno University of Technology, Brno, Czech Republic

ABSTRACT: The long-term deflections of cast-in-place segmental bridges constructed using the cantilever method are often larger than the deflections expected in the design. A detailed structural analysis and monitoring of these types of bridges has therefore become a matter of interest. The paper gives a comprehensive analysis of phenomena that can conspire to cause long-span prestressed concrete bridges to deflect more than predicted. The analysis is based on a parametric study and long-term monitoring of the behaviour of a Motorway Bridge across the Vltava River in the Czech Republic, Europe. To reveal and quantify possible reasons for excessive bridge deflections, a detailed time-dependent analysis was carried out. The results of the study were compared with in-situ measurements that have been regularly carried out since the very early stage of construction. Theoretical values of the increments of deflections, prestressing forces and concrete strains were compared with the measurements. 1

INTRODUCTION

In practice we often encounter the fact that the longterm deflections of prestressed bridges are greater than the deflections expected in the design. Especially in the case of cast-in-place segmental bridges constructed using the cantilever method, the excessive deflections might occur sometime after completion. Explanation of possible reasons for such large deflections therefore became a matter of interest of many engineers and researches. Consequently, some of them were specified. In spite of the effort no final quantification of the specified reasons and no unambiguous and definite statements have been concluded. 2

POSSIBLE REASONS FOR EXCESSIVE BRIDGE DEFLECTIONS

The phenomena that can, and often do, conspire to cause long-span prestressed concrete bridges to deflect more than predicted, have been identified based on the recommendations CEB (1997), and considering the author’s observations of the bridge and the results of stochastical analysis, Florian & Navrátil (1998). A substantial number of reasons for excessive deflection arise from technological errors. They are in particular the enlargement of water content in concrete mixture, insufficient modulus of elasticity, strength or unit weight resulting from a poor quality of concrete, a wrong sequence or time schedule of construction steps, geometrical imperfections and higher

385

prestressing losses, unsatisfactory stiffness of temporary supports or poor anchorage, improper and short curing and wrong estimate of the relative ambient humidity, etc. Another reason for unexpected deflection is the omission of some phenomena during structural modelling of a structure. It concerns for example shear lag and the influence of shear on deformation generally, eccentric position of prestressed and non-prestressed reinforcement, the contribution of non-structural members, such as parapets and asphalt wearing surface, to the load carrying capacity, stiffness reduction due to cracks development, friction in bearings, etc. A frequently discussed reason is the underestimation of the rheological effects, which can be caused especially by using an inappropriate physical model, or which can result from uncertainty of the creep and shrinkage prediction calculated only from the composition and strength of concrete. Not fully appreciated is excessive creep and shrinkage of concrete resulting from the exposition of the structure to the harsh conditions in situ, e.g. longterm high temperatures of concrete, cyclic load, cyclic humidity or acceleration of drying due to microcracking. The underestimation of long-term losses of the prestressing caused especially either by excessive shrinkage or creep of concrete can decrease the level of applied prestressing and consecutively increase the deflection. Creep and shrinkage related is also the underestimation of non-uniform drying of concrete in the

section: the differential drying of differently sized parts of the cross-section, non-uniform drying across the thickness of slabs and walls or the influence of bridge deck waterproofing on drying of concrete cross-section.

A series of measurements have been executed since the end of construction in 1996, with the most recent measurement done in November 2008. Some of them are compared with theoretical values in this paper. 3.2

3

BRIDGE ACROSS THE VLTAVA RIVER

To reveal and quantify the possible reasons for excessive bridge deflections the parametric study of longterm behaviour of the Motorway Bridge across the River Vltava was carried out. In parallel extensive technical supervision has been performed, followed by long-term monitoring since the construction of the bridge started in 1995. 3.1

Long-term monitoring

The construction of the bridge was carefully documented and an extensive testing program was carried out during the construction. The instrumentation for long-term measurements has been placed in the main span of the bridge, see Figure 1. In total five sections have been instrumented by mechanical or electricresistance concrete strain gauges and two sections by electric-resistance thermal sensors. The sensors have been placed in cross-section so that it would make the measurement of non-uniform distribution of temperature possible. Magneto-elastic force sensors have been used for long-term monitoring of changes of prestressing. In total 37 geodetic points have been fixed in the box of the main girder for the measurements of the deflection.

Figure 1. bridge.

Instrumented sections in the main span of the

Figure 2. Layout of measuring instrumentation in section IV.

Parametric study

The detailed finite element model was developed for the analysis. The deterministic analysis of the bridge was carried out repeatedly for different input parameters, which enabled to quantify the phenomena significant for bridge deflection analysis. The full stochastical approach was not used due to the necessity to deal with a vast amount of data. The analysis aimed especially at the calculation of the long-term deflection of the main span. 3.2.1 Structural model Certainly the structural model of bridges constructed using the cantilever method must respect the changes of static system and boundary conditions. New structural members are assembled or cast, post-tensioning is applied and temporary support elements are removed. Concrete of structural elements of various ages is combined. Therefore, during both construction and throughout the service life, account must be taken of the creep and shrinkage of concrete. Variety of static systems and the effects of creep and shrinkage make the structural analysis complicated. That is why sophisticated general methods are needed for the structural analysis, especially for the verification at intermediate stages of construction in structures where properties vary along the length. Such realistic modelling of structural behaviour can be achieved using the software SCIA ENGINEER, the module TDA, SCIA 2008. Limited space does not allow for the description of the details about the method used for the time-dependent analysis. Therefore we concentrate on the main features of modelling only. The finite elements on eccentricity represent the concrete box girder, prestressed tendons, diaphragms, piers, and temporary anchoring ties, non-prestressed reinforcement and the elements representing the formwork traveller. All operations in the construction are respected in the structural analysis according to the real production schedule. Real dead weight and real prestressing including the losses at transfer are considered in the analysis according to the records of both the contractor and independent observer. The contribution of the parapets to the stiffness of the structure and the friction in the bearings are neglected. 3.2.2 The studied input parameters The structural model described above was a basis for the parametric study and it is marked as Variant 0 in this paper. In this variant the B3 rheological model is used in the basic formulation, see Bažant & Baweja (1995). The modulus of elasticity is calculated

386

for stress duration  = 0.01 days and for designed composition of concrete, see Table 1. The proper curing of concrete is assumed for the first three days. The average effective cross-section thickness is calculated for the whole cross-section with respect to the influence of bridge deck waterproofing. The average measured ambient humidity h = 0.77 is taken into account. The composition, strength, and curing of concrete are the only material input properties for the majority of rheological prediction models. The composition and strength of concrete cast-in-situ always differ from the designed values, and they also differ for different segments. Detailed measurements of the quality of concrete during the construction of the bridge were collected by the authors, and used for the updating of input data in Variant 1 of the study. In Variant 2 the real composition of concrete is modified in the way that water-cement ratio is kept constant, but both the water and cement contents are increased by 10% (simultaneously the aggregate content is decreased to retain the unit weight of fresh concrete). The lack of work discipline at the site very often leads to an increase in the water content, which makes the concrete easier to work with. To keep constant strength of concrete cement concrete is also proportionally increased, which increases the creep and shrinkage of concrete. The type of curing is studied in Variant 4. In the case of the analysed structure, the curing was carried out through moistening of the fabric covering the horizontal parts of the box girder. The frequency of moistening was decreased during the first three days after casting. Therefore the effectiveness of the curing is questionable. For curing in water the value 1.0 of parameter α2 is recommended by Bažant & Baweja (1995). In Variant 4 α2 is considered 1.2, which is the value recommended for sealed specimens. The largest source of uncertainty of creep and shrinkage prediction is the dependence of model parameters on the composition and strength of concrete. To reduce this uncertainty the short-time laboratory measurements of creep and shrinkage of small size specimens were carried out and were used for the updating of the prediction, Navrátil (1998a). The updated B3 rheological model is used with parameters τsh = 245 days, p6 = 1.4, p1 = 0.13383 ∗ 10−7 , and p2 = 0.778397 in Variant 5. The quality of updated prediction was checked through the measurements of creep and shrinkage of specimens placed inside of the box of the box girder. Table 1.

Input parameters of B3 Model.

w kgm−3

c kgm−3

a kgm−3

α1

α2

ks

 fcyl MPa

170

410

1810

1.1

1.0

1.0

36.5

The influence of shear deformation on long-term deflection is tested in Variant 3. The contribution of shear deformation is respected in all variants of this study. The shear reduction factor can be used for the calculation of effective area of the cross-section. In practice usually the area of the walls of the box girder is used as the effective area. But using this approach the effect of shear lag is neglected, which may lead to underestimation of deflections, see Bažant et al. (2008). For simplicity the modified effective area is used according to the recommendations in Petˇrík & Kˇrístek (1998). The value of reduction factor 0.5 is used in Variant 3. All other input data are identical with Variant 1. The effect of the drying of concrete is approximately expressed in terms of effective cross-section thickness. It depends on volume to surface ratio and, in our analysis, it is calculated for the whole cross-section with respect to the influence of bridge deck waterproofing. The calculation itself is simple but still the calculated value is ambiguous. The uncertainties result from the differences in size between the laboratory specimens and a real bridge cross-section, and from the differences in ambient humidity inside and outside of the box. The sensitivity of long-term deflections on effective cross-section thickness can be deduced from Variant 8 and 9 respectively, in which the effective thickness is decreased and increased by 10%. All other input data are again identical with Variant 1. The creep and shrinkage of concrete exposed to the harsh conditions is accelerated and increased. These effects are approximately expressed by the extensions of the B3 model and were analysed in Navrátil, (1998b). The updated B3 rheological model with extension for cyclic humidity is used in Variant 7, in which all other input data come from Variant 5. The environmental humidity cycles were monitored, recorded and analysed by the authors. For this study the period of humidity cycles is considered 210 days, and the amplitude of environmental relative humidity is taken as 0.27. Variant 5 is also a basis for Variant 6, in which the effect of microcracking on the drying of concrete is studied. The updated B3 rheological model with no extensions is used. The effect of microcracking is respected via the correction of shrinkage half-time τsh . The diffusivity C1 of given intact concrete can be calculated from measured shrinkage half-time. The microcracks will increase the diffusivity as much as twice. Using new diffusivity we calculate new τsh and the new value of drying penetration depth Dp . Variant 18 links together the preceding Variants 5, 6, 7. The updated B3 model with all extensions and with respect to microcracks is used together with real composition and measured strength of concrete. In Variant 19 the mean values of measured relative environmental humidity were supplied in time intervals used for the time-dependent analysis. All other

387

input data are again identical with Variant 18. This variant should therefore be the most sophisticated and closest to the reality so far. The selection of a proper rheological model is of crucial importance for the realistic prediction of deflection. It is known that the models often reasonably differ. For this reason comparative calculations were performed for various models used in Czech ˇ ˇ (CSN 736207 and CSN 731201) and international (CEB-FIP MC 1978 and CEB-FIP MC 1990) codes. Two variants for each model were calculated. One for designed and one for measured composition and strength of concrete. From this point of view the input data are comparable in Variants 0, 10, 12, 14, and 16, and in Variants 1, 11, 13, 15, and 17. The input data were adapted for each rheological model specifically, including the effective cross-section thickness, concrete strength and modulus of elasticity, see Table 2. For the rheological model according to the CEBFIP Model Code 1990 the type of aggregate was determined as basalt. The influence of differential shrinkage and creep of differently sized parts of the cross-section is usually neglected in design practice. Based on the study in Navrátil et al. (1999) and Kˇrístek et al. 2006, the question arose of what the impact of this effect would be on real structures with more complicated geometrical shapes and construction. That is why a new structural model of the Motorway Bridge across the River Vltava was created based on Variant 5. The model therefore contains real composition and strength of concrete, B3 rheological model updated using short-time laboratory measurements, and mean values of measured relative environmental humidity. The cross-section of the main span of the bridge is split into nine structural elements placed on appropriate eccentricity related to the reference axis, see Figure 3. The cross-sections of other spans of the bridge are split into three structural elements (top slab, web, bottom slab). Hence this adaptation of

Table 2. models.

the structural model results in a significant increase of finite elements. The effective cross-section thickness is calculated in three variants (i) for the crosssection (not split into elements) with bridge deck waterproofing—Variant 20, (ii) for the cross-section split into elements with bridge deck waterproofing— Variant 21, and (iii) for the cross-section split into elements without bridge deck waterproofing—Variant 22. It is assumed that the waterproofing is applied immediately after the end of curing (for simplicity) and it influences the drying of the cross-section. 3.3

Results

For this paper we limit the outputs to the vertical deflection in the middle of the main span and to an illustration of force variation in one prestressing tendon. The variation of midspan deflection over time is shown in Figure 4 for two (most comprehensive) variants of the time-dependent analysis, for more variants see Navrátil (1999). The calculated vertical deflections are compared with measured values, which were adjusted for reference temperature and zero deformation of the piers. The deflections are related to the date of 21st June 1996, when the first geodesic measurement was performed after the connection of two ends of both double-cantilevers. From that time, the uncertainties resulting from temporary supporting were eliminated, as the structure became continuous. It should be emphasised that the prestressing, parapets, barriers and surfacing had not yet been completed and the ballast had not yet been removed at that time. For the purpose of clear arrangement the final deflections at the time of 100 years are shown in Figure 5. Differential shrinkage and creep cause the deflection, which increases in time with the maximum reached approximately 5000 days after the structure is made continuous. Further ahead the deflection decreases and it is close to zero at 100 years. It can be seen in Figure 6. The difference in deflections calculated using the models in which the cross-section is and is not split into elements (e.g., Variants 20 and 21) is drawn on the vertical axis. The models with bridge

Moduli of elasticity for various rheological

Code 736207 731201 CEB-FIP 78 CEB-FIP 1990 Variant 10 11 12 13 14 15 16 17 E/GPa/ 36.0 38.5 34.5 37.5 32.4 35.1 39.7 42.9

Figure 3.

Cross-section and the structural model.

Figure 4.

388

Relative deflection at midspan.

deck waterproofing show the increase of the deflection approximately by 11.5 mm. Conversely disregarding the effect of waterproofing we obtain the decrease of the deflection by 22.3 mm. It documents the error made by the designer, who neglects the effect of waterproofing for the whole lifespan of the structure even though he takes into account differential shrinkage and creep. The underestimation of the deflections would be quite significant. This finding can help us to explain why the gradient of measured deflections is higher than the gradient of values calculated in the first year of service. The acceleration of deflection might be caused by the application of waterproofing on the top slab of the box girder. Consequently the drying (and the effective cross-section thickness) is changed, which cannot be modelled by software used for the time-dependent analysis yet. Considering the results in Figure 6 we may conclude that differential shrinkage has no effect on the final value of the deflection, but it has significant impact on its in-time development. It is obvious that taking it into account, the agreement between measured and calculated values improved. Figure 7 demonstrates the agreement between the prediction of prestressing forces and measured values. The tendon shown in the picture was investigated in segment 3B, close to the anchor. The relative force shown in the diagram was obtained as a ratio of

Figure 5.

Final relative deflection at midspan.

Figure 6. Relative deflection at midspan caused by differential shrinkage.

389

Figure 7.

Prestressing force in tendon.

prestressing force, which was calculated and measured respectively at a specific time, to the prestressing force calculated and measured at the moment of transfer. In this way it was possible to determine the relative level of long-term losses. Measured values of prestressing forces were adjusted for reference temperature and corrected with respect to the change of the temperature coefficient of tendon magnetoelastic characteristics, due to the force in the tendon.

4

CONCLUSIONS

The parametric study made it possible to quantify the phenomena significant for bridge deflections and to reveal possible reasons for excessive long-term deflections of long-span bridges. From the early stage of construction, extensive measurements of the Motorway Bridge across the Vltava River have been performed. The agreement between the theoretical results and insitu measurements confirmed the quality of both the material and the structural model. Both the theoretical and in-situ investigations confirmed the expected total values of bridge deflection. The gradient of deflections is slightly higher than expected. Time-dependent analysis was performed for various input parameters and for various material models. Among the studied input parameters the water and cement contents have the greatest influence on longterm deflections. The importance of the selection of the rheological model for the realistic prediction of deflection is manifested in Figures 4 and 5. The most realistic predictions of both the gradients and total values of deflection are obtained by the B3 Model, especially in updated version with extensions (Variants 18 and 19). The updating of the creep and shrinkage model for specific concrete is of crucial importance. The method for the updating of the prediction described in Bažant & Baweja (1995) and Navrátil (1998a) is efficient and it is usable even at the phase of the preparation of design documentation with relatively low expenses compared to costs of important long-span

bridges. For these reasons the incorporation of the method into present codes of practice can be recommended, so that it can be used in future as standard procedure. As it is shown in Figure 5, the scatter of the final deflections is less than expected (from 70 mm to 160 mm) although the range of input parameters was relatively high. It might therefore be concluded that common variability of input data cannot be solely a reason for excessive deflections of the structure. Hence the susceptibility of long-span bridges to excessive deflection due to their sensitivity to input parameters is not as high as has been assumed to date. Based on this study we may therefore conclude that the excessive deflections, which have appeared in some bridges, resulted from the combination of serious errors in design (dead load not sufficiently balanced by prestressing), insufficient work discipline (imperfections in tendon profiles, higher losses) and adverse service conditions (corrosion of materials). We should remark that neither abrupt drops of the prestressing forces nor the significant increase of loads were investigated in this study. These events would have to result from disastrous errors in either design or construction and in their consequence both the upward loads caused by prestressing and the stiffness of structural members would rapidly decrease. In such a case the increase of deflections is very impressive and any (even sophisticated) tool for the analysis of such a structure would only help us to reveal those errors. To minimize the danger of such errors appearing, the design of prestressing using the load balancing method is strongly recommended. It is the advantage and glory of prestressed concrete that the distribution of internal forces in the structure can be actively modified. In this concept of prestressing, the engineer can use his or her inventiveness, ingeniousness and creativity, especially in the design of statically indeterminate structures. This philosophy was probably used, perhaps intuitively, even earlier, but it was described for the first time by T.Y. Lin in 1963, see Lin & Burns (1982). Nowadays it is recommended by leading engineers worldwide, e.g. Favre & Markey (1994), even though at first sight it is not economical. It requires a greater number of tendons and sometimes even more

complicated arrangement of these. On the other hand, it prolongs the lifespan of the structure and improves its serviceability.

REFERENCES Bažant, Z.P. & Baweja, S. 1995. Creep and Shrinkage Prediction Model for Analysis and Design of Concrete Structures—Model B3. Mater. Struct., 28: 357–365. Bažant, Z.P., Li, G.H., Yu, Q. Klein, G., Kˇrístek, V. 2008. Explanation of Excessive Long-Time Deflections of Collapsed Record-Span Box Girder Bridge in Palau. Preliminary report, 8th International Conference on Creep and Shrinkage of Concrete, Ise-Shima, Japan. Comité Euro-International du Béton 1997. Bulletin d‘Inform. No 235—Serviceability Models, Lausanne: CEB. Favre, R. & Markey, I. 1994. Generalization of the load balancing method. 12th FIP Congress, Béton Précontraint en Suisse: 32–37, Washington. Florian, A. & Navrátil, J. 1998. Stochastical Analysis of Highway Bridge across Vltava River. In Shiraishi, Shinozuka & Wen (eds), Structural Safety and Reliability; Proc. of the 7th International Conference Icossar ’97: 1957–1960. Rotterdam: Balkema. K˘rìstek, V., Bažant, Z.P., Zich, M., Kohoutková A. 2006. Box girder bridge deflections. ACI Concrete International Jour., 28: 55–63. Lin, T.Y. & Burns, N.H. 1982. Design of Prestressed Concrete Structures. New York: John Wiley & Sons. Navrátil, J. 1998a. Updating of Concrete Creep and Shrinkage Prediction (in Czech). Stavební obzor, 2: 44–50. Navrátil, J. 1998b. The Use of the B3 Model Extension for the Analysis of Bridge Structures (in Czech), Stavební obzor, 4: 110–116. Navrátil, J. 1999. Study of long-term behaviour of cast-inplace segmental bridge. In Proceedings of the fib Symposium 1999: 469–474. Prague: Viacon Agency. Navrátil, J., Zich, M., K˘rìstek, V. 1999. Influence of differential shrinkage on deflection of box girders (in Czech). In Proc. of the Conference Concrete Days 1999: 174–181. ˇ ˇ Pardubice: CBS CSSI. Pet˘rìk, V. & K˘rìstek, V. 1998. The deflection analysis of box girder bridges with respect to shear lag (in Czech). Project S303/120/602 Dpt. of Transport and Communications of Czech Republic, Prague: Czech Technical University. SCIA 2008. SCIA ENGINEER—Software System for Analysis, Design and Drawings of Steel, Concrete, Timber and Plastic Structures, Herk-de-Stad, Belgium: SCIA Group nv.

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Structural optimization and computation

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

A revised BESO method for structures with design-dependent gravity loads X. Huang & Y.M. Xie School of Civil, Environmental and Chemical Engineering, RMIT University, Melbourne, Australia

ABSTRACT: The effectiveness and efficiency of the bi-directional evolutionary structural optimization (BESO) method has been demonstrated on the problems of minimizing compliance of structures with fixed external loads. This paper considers the minimization of the mean compliance of continuum structure subjected to design-dependent gravity loads. Due to the non-monotonous behaviour for this type of the optimization problems, the extended BESO method using discrete design variables has difficulty in obtaining convergent solutions for such problems. In this paper, a new BESO method is developed based on the sensitivity number computation utilizing an alternative material interpolation scheme. Several examples are presented to demonstrate the capabilities of the proposed method for achieving convergent optimal solutions for structures with design-dependent gravity loads. 1

INTRODUCTION

Topology optimization for continuum structures is to find the best distribution of the available material in the predetermined design domain. For the maximum stiffness topology design, minimizing the mean compliance of the structure is commonly used as objective function and the material volume constrained is imposed which limits the maximum allowable material for the design. Typically the topology optimization problem is treated by dividing the design domain into N finite elements and each element is taken as the design variables. Therefore, a finite element-based topology optimization problem becomes 1 Minimize: C = f T u 2 Subject to: V ∗ −

N 

Vi xi = 0

(1)

i=1

xi = xmin or 1 where C is the mean compliance of the structure and f and u are the external force and displacement vectors. Vi is the volume of an individual element and V ∗ the prescribed total structural volume. The binary design variable xi denotes the density of ith element. Normally a small value of xmin e.g. 0.001 is used to denote the void elements. Topology optimization methods are able to redistribute the material so as to achieving the defined objectives. One of the most commonly used optimization methods is the Solid Isotropic Material and Penalization (SIMP) method (Bendsoe 1989; Zhou and Rozvany 1991; Rietz 2001; Sigmund 2001; Bendsoe

393

and Sigmund 2003). In this approach, the discrete design variable is relaxed and allowed to attain all values between its lower bound xmin and its upper bound 1. The intermediate material stiffness can be easily penalized with the power-law as p

E(xi ) = E 0 xi

(2)

where E 0 denotes the Young’s modulus for solid material, p is the penalty exponent. The gradient of the objective function with respect to individual element density can be easily derived by the adjoint method as p−1

px ∂C =− i ∂xi 2

uiT Ki0 ui

(3)

where Ki0 are the elemental stiffness matrix for solid elements. By specifying a value of p higher than one (normally p ≥ 3), the SIMP method leads to nearly solid-void optimal designs. Another popular and simple topology optimization method is the evolutionary structural optimization (ESO) method (Chu et al. 1996; Xie and Steven 1997) and its later version, the bi-directional ESO (BESO) method (Yang et al. 1999; Querin et al. 2000). The original ESO and BESO methods search a better topology by gradually removing and adding materials. Some critical comments on the original ESO and BESO methods can be found in the recent paper by Rozvany (2009). Typically, the original ESO and BESO methods hardly results in a convergent solution which indicates that the resulting solution can not be further improved under the given update algorithms. The recent BESO method utilizing the SIMP material model has been developed by authors (Huang and

Xie 2009) to directly attack the discrete topology optimization problem as Equation 1. The elemental sensitivity number which denotes the relative ranking of the elemental sensitivity can be simply Expressed by ⎧ 1 T ⎪ ⎪ u Ku ⎪ ⎨2 i i i

1 ∂C αi = − = ⎪ p−1 p ∂xi ⎪ ⎪ ⎩ xmin uT Ki ui 2 i

when xi = 1 when xi = xmin (4)

Normally, a large value p is applied where void elements always denotes by soft elements. Therefore, we often called this method as the soft-kill BESO method. When p tends to infinity, the elemental stiffness becomes zero according to Equation 2 and the elemental sensitivity number can also be simplified by ⎧ ⎨ 1 uT K 0 u i i αi = 2 ⎩ 0

when xi = 1

(5)

when xi = 0

where xi = xmin is replaced by xi = 0 because a non-stiffness soft element is equivalent to a void element. The original hard-kill ESO method was driven by the above sensitivity number. With the above sensitivity numbers void elements may also re-enter into the design domain when the filter scheme is applied (Huang and Xie 2007). This optimization approach is normally called as the hard-kill BESO method. Numerical examples indicate that the both softkill and hard-kill BESO methods with the meshdependency filter produce convergent solutions which are very close to that of the SIMP method embedded with a filter. Also, the solution of the BESO method is independent on the selection of the penalty exponent mainly due to the fact that the sensitivity numbers of the solid elements does not change with the penalty exponent as shown in Equations 4 and 5. The qualities of solutions and computational efficiency of both hard-kill and soft-kill BESO methods refers to our recent papers (Huang and Xie 2007; Huang and Xie 2009). Obviously, the feasibility of the BESO method using discrete design variables is based on the fact that the SIMP method with continuous design variables is really convergent to a void-solid design when p is large enough. It has been proved that the topology optimization with design-dependent body forces and especially self-weight loads cannot be solved by directly extending the SIMP procedure used for minimumcompliance topology optimization with fixed external loads (Bruyneel & Duysinx 2005). In practice, taking account of self-weight loads is extremely important in many engineering problems such as the optimizaiton of large civil structures. The particular difficulties

arising in such topology optimization problems are the non-monotonous behaviour of the compliance and the parasitic effect for low densities when using the power law material model (Bruyneel & Duysinx 2005). Therefore, it is also impossible to solve the topology optimization including self-weight loads based on a direct extension of the above BESO procedure. It is undoubtedly a challenging task for developing a new BESO procedure with discrete design variables which leads to convergent solutions for such problems. Topology optimization of continuum structures with self-weight loads has been studied by Yang et al. (2005) and Ansola et al. (2006) using the modified ESO procedure. An obvious shortcoming of the ESO procedure is that the material can not be recovered once it is wrongly removed. Therefore, there is an urgent and practical need to develop a sophisticated BESO procedure for this topology optimization problem. In this paper, we will develop a new BESO method utilizing an alternative material interpolation scheme proposed by Stolpe and Svanberg (2001). A number of examples will be given to demonstrate the capabilities of the proposed method for achieving convergent optimal solutions for structures with self-weight loads. 2

SENSITIVITY ANALYSIS

The basic formulation of a topology optimization problem may also consist of minimizing the mean compliance of forces including self-weight loads subjected to a given volume of the material as shown in Equation 1. Differing from the topology optimization problem for fixed external forces, the applied force vector, f , includes the design-dependent self-weight loads. The adjoint method can be used to determine the sensitivity of displacement and force vectors by introducing a vector of Lagrangian multiplier λ. The modified objective can be expressed by C=

1 T f u + λT (f − Ku) 2

(6)

where the term λT (f – Ku) is equal to zero according to the equilibrium equations and therefore the modified objective function is identical to the original one. The variation of the modified objective function can be written as dC 1 df T 1 du = u + fT dxi 2 dxi 2 dxi   df dK du + λT − u−K dxi dxi dxi

(7)

where the first term denotes the variation of the force vector with the design variable, xi . The above equation

394



1 1 1 1 T fi = Vi ρi gf = Vi ρi g 0, − , 0, − , 0, − , 0, − 4 4 4 4

can be rewritten as   dC df T 1 u+λ = dxi dxi 2   du 1 T dK T − λT u + f −λ K 2 dxi dxi

(12)

(8)

To eliminate the unknown du/dxi from the sensitivity expression λ is chosen as λ=

1 u 2

(9)

Thus, the sensitivity of the objective function becomes df T 1 dK dC = u − uT u dxi dxi 2 dxi

(10)

It can be seen that the sensitivity of the mean compliance can be either positive or negative. Sometimes it even changes sign when changing the value of the design variable, and therefore the mean compliance experiences a non-monotonous behaviour.

where g is the gravity acceleration. Therefore, the sensitivity of the mean compliance can be found to be: dC 1+q T = Vi ρ 0 gf ui − uiT Ki0 ui (13) dxi 2 [1 + q(1 − xi )]2 It can be seen that the elemental sensitivity depends on the value of penalty factor, q. Also, the ratio between the first and second terms becomes a constant as the design variable tends to zero. The sensitivity number in the BESO method denotes the relative ranking of solid and void elemental sensitivity and can be expressed explicitly as αi = −

1 dC q + 1 dxi

⎧ Vi ρ 0 g T ⎪ 1 T 0 ⎪ ⎪ ⎨− q + 1 f ui + 2 ui Ki ui = ⎪ Vi ρ 0 g T uiT Ki0 ui ⎪ ⎪ ⎩− f ui + q+1 2[1 + q(1 − xmin )]2

xi = 1 xi = xmin (14)

3

THE MATERIAL INTERPOLATION MODEL AND SENSITIVITY NUMBER

When the power law material interpolation scheme is applied, the ratio between the first and second terms becomes unbounded in the low-density regions. Therefore, it is almost impossible to obtain a 0/1 design under the given material model (Bruyneel and Duysinx 2005). Here, we consider an alternative interpolation scheme proposed by Stolpe and Svanberg (2001) which overcomes the above shortcoming of the powerlaw material model. The density and Young’s modulus of the material model for a 0/1 design take as ρi = xi ρ 0 Ei =

xi E0 1 + q(1 − xi )

(11)

where ρ 0 and E 0 denotes the density and Young’s modulus of the solid material. q is the penalty factor which is larger than 0 for topology optimization problems. When the structure is meshed with finite elements, for instance four node quadrangular elements, the elemental load vector subjected to its self-weight can be obtained as the following equation assuming that the gravity is aligned with the global y direction.

395

To minimizing the structural mean compliance, we should update design variables by switching xi from 1 to xmin for lowest sensitivity number and from xmin to 1 for highest sensitivity number. Similar to the powerlaw material interpolation scheme, a large penalty factor should be selected so that the BESO approach using discrete design variables can be convergent to a stable 0/1 design. We also noticed that the Young’s modulus of soft elements becomes zero according the material interpolation scheme Equation 11 when the penalty factor, q, tends to infinity. Assuming with the finite displacement field, the sensitivity number can be expressed by the following equation as the penalty factor tends to infinity. ⎧ ⎨ 1 uT K 0 u i αi = 2 i i ⎩ 0

xi = 1

(15)

xi = 0

where xi = xmin is replaced by xi = 0 because a soft element is equivalent to a void element. The above equation indicates the hard-kill BESO method for structures with self-weight loads has no difference from that for structures with fixed external forces. The sensitivity analysis indicates that the non-monotonous behaviour of the objective function depends on the value of penalty factor, q. With a larger penalty factor,

the non-monotonous behaviour of the objective function becomes weak. In the extreme case with an infinite penalty, the non-monotonous behaviour totally disappears as shown in Equation 15. It should be pointed out that the penalty factor here also weights the both effects from the self-weight loads and elemental strain energy as shown in the Equation 14. Differing from the topology optimization problems with fixed external forces, the selection of the penalty factor does affect the relative ranking of the sensitivity number for solid elements. The total elimination of the first term in the hard-kill BESO method may result in a different solution from that of the soft-kill BESO method. However, the high computational efficiency of the hard-kill BESO method still incurs us to explore its possibility.

4

Figure 1.

THE FILTER SCHEME AND STABILIZATION OF THE EVOLUTION PROCEDURE

Topology optimization usually encounters checkerboard patterns and mesh-dependency problems. To circumvent those problems, we use a BESO meshindependency filter (Huang and Xie 2007, 2009) by averaging the elemental sensitivity number with its neighbouring elements based on image-process techniques. The BESO mesh-independency filter works as a low-pass filter that eliminates features below a certain length-scale in the optimal topologies. In the BESO algorithm, only two discrete design variables xmin and 1 can be used. It is almost impossible to obtain a convergent solution unless an extreme fine mesh is used. Therefore, the current elemental sensitivity number is modified by averaging with its historical information (Huang and Xie 2007, 2009). This algorithm suppresses the changes of design variables for solid elements with higher historical sensitivity numbers and void elements with lower historical sensitivity numbers, and stabilizes the whole evolution process. Once the solution is convergent, the above averaging process almost has no effect on the original elemental sensitivity numbers. The detailed BESO procedure may refer to the papers (Huang and Xie 2007, 2009).

5 5.1

Design domain of example 1.

Figure 2. Evolution histories of mean compliances using soft-kill and hard-kill BESO methods.

EXAMPLES Example 1

A rectangular plate of 1 m × 0.5 m is simply supported at its bottom ends as illustrated in Fig. 1. The objective of the problem is to find the optimal topology of the structure withstanding its self-weight. The amount of available material was constraint to 15% of the design domain. The following material properties are assumed: Young’s modulus 200 GPa, Poisson’s

ratio 0.3 and the density 78 kg/m3 . Due to the symmetry, only half of the design domain is analyzed using a mesh of 100 × 50 four-node plane stress elements. Upon completion of the evolutionary process, the symmetric half region is mirrored up to clarify the whole structure. In this example, both the soft-kill and hardkill BESO methods are applied. The parameters for the soft-kill BESO are ER = 2%, ARmax = 2%, q = 5 and rmin = 30 mm. The parameters for the hard-kill BESO are same to those for the soft-kill BESO except for an inherent infinite penalty factor, q. The evolutionary optimization is carried out based on the above derived sensitivity numbers. Initially, all elements in the design domain are assigned to be solid elements. The total volume of the structure gradually decreases and then keeps constantly when the prescribed volume constraint is achieved. Figure 2 shows the evolution histories of the mean compliance for the soft-kill and hard-kill BESO methods. It can be seen that the mean compliances gradually decrease and then are convergent to an almost constant value within less 80 iterations. The evolution histories of the mean compliance for both BESO methods are almost identical and the hard-kill BESO method even gives a more stable history of the mean compliance. Figure 3 shows the optimized topologies resulted from the soft-kill and hard-kill BESO methods respectively. In both cases the optimization process yields

396

(a)

(b)

Figure 4.

Design domain of example 2.

Figure 3. Optimal designs from various BESO methods (a) the soft-kill BESO method with q = 5 and C = 3.82 × 10−8 Nm; (b) the hard-kill BESO method and C = 3.81 × 10−8 Nm. (a)

an arch structure which spans the both end supports to optimally withstand its self-weight. Differing from the minimization compliance of applied loads, the resulted optimal shapes from the soft-kill and hard-kill BESO methods have a little difference. The objective functions are 3.82 × 10−8 Nm and 3.81 × 10−8 Nm for the soft-kill and hard-kill designs respectively. The close values of the objective function indicate that both designs can be looked as final optimal designs.

(b)

(c)

(d)

5.2 Example 2 In this example, the classic MBB beam subjected to a concentrated load and its self-weight is to be optimized. The dimensions and support conditions of the design domain are shown in Fig. 4. Comparing with the first example, this structure is more flexible under its self-weight loads. Due to the symmetry, only half of the design domain is discretized with 100 × 50 four node plane stress elements. The material volume constraint is set to be 40% of the whole design domain. The Young’s modulus E = 200 GPa, Poisson’s ratio v = 0.3 and density ρ = 78 kg/m3 are assumed. Therefore, the total weight of the final design is expected to be 3120 kg. The following parameters are used in the following BESO procedure: ER = 2%, ARmax = 2%, q = 5 and rmin = 0.3 m. When self-weight is not taken into account, the amplitude of the load F has no effect on the resulting optimal topology as shown in Fig. 5(a). In this case, 76 iterations are needed to reach a convergent solution. It indicates that the soft-kill BESO method utilizing the alternative material interpolation model is also efficient to obtain an optimal solution for the minimization compliance problem with fixed external loads. When the self-weight loads are taken into account, several magnitudes of the applied force F are tested. The optimal designs and their corresponding mean compliances are shown Fig. 5. The total numbers of the

Figure 5. Optimal designs for different ratios between the applied load and the self-weight (a) no self-weight; (b) F = 100% of the weight and C = 0.282 Nm; (c) F = 50% of the weight and C = 0.131 Nm; (d) self-weight only and C = 0.034 Nm.

397

needed iterations are 57, 58, and 57 for Fig. 5(b-d) respectively. The same conclusions from (Bruyneel and Duysinx 2005) can be drawn that the resulting topology depends on the ratio between the applied load and the structural weight. When the applied load decreases, the shape of the structure tends to be an arch, which makes senses from an engineering point of view. To further verify the effectiveness of the hard-kill BESO method for the non-monotonous problem, it is applied for the above optimization problem with only self-weight loads. The same parameters except for the penalty factor are used. Figure 6 shows the optimal topology resulting from the hard-kill BESO method which is similar to that of the soft-kill BESO method. The evolution histories of the mean compliance are shown in Fig. 7. It indicates that the soft-kill BESO method produces a little better solution than the hard-kill BESO method. The mean compliance of the hard-kill design is 0.037 Nm which is about 9% higher than that of the soft-kill design. If we assumed a guess design to be a uniform plate with the objective weight, its mean compliance is 0.068 Nm. Comparing with

Figure 6. Optimal design of structures with self-weight only using the hard-kill BESO method.

The results demonstrate that both BESO methods produce convergent optimal solutions although the objective functions of the hard-kill designs maybe a few percent above those of soft-kill designs. Even so, the hard-kill BESO method is still very useful for topology optimization problems of complex 3D structures due to its high computational efficiency. REFERENCES

Figure 7. Evolution histories of mean compliances using soft-kill and hard-kill BESO methods.

the improvement over the guess design, the hard-kill solution can still be looked as an approximate optimal design.

6

CONCLUSIONS

The previous BESO method has been successfully applied for topology optimization of structures with fixed external forces. However, the direct extension of the BESO procedure does not give good results when it is used for structures with design-dependent loads such as gravity. In this paper, the sensitivity numbers used in the BESO procedure are derived based on an alternative material interpolation model so as to avoid the unbounded effect for low densities in the power-law model. The resulted sensitivity number is composed of two components: one comes from the self-weight loads and another from the element strain energy. The penalty factor weights the effects of two components and the non-monotonous behaviour of the mean compliance becomes weak as the penalty factor is increased. The hard-kill BESO method is also established as the penalty factor tends to infinity and the corresponding optimization problem becomes totally monotonous. Both the soft-kill and hard-kill BESO methods have been successfully applied to several topology optimization problems including self-weight loads.

Ansola, R., Canales, J. and Tarrago, J.A. (2006). ‘‘An efficient sensitivity computation strategy for the evolutionary structural optimization (ESO) of continuum structures subjected to self-weight loads." Finite Elements in Analysis and Design 42: 1220–1230. Bendsoe, M.P. (1989). ‘‘Optimal shape design as a material distribution problem.’’ Struct. Optim. 1: 193–202. Bendsoe, M.P. and Sigmund, O. (2003). Topology Optimization: Theory, Method and Application, Springer-Verlag, Berlin Heidelberg. Bruyneel, M. and Duysinx, P. (2005). ‘‘Note on topology optimization of continuum structures including selfweight.’’ Struct. Multidisc. Optim. 29: 245–256. Chu, D.N., Xie, Y.M., Hira, A. and Steven, G.P. (1996). ‘‘Evolutionary structural optimization for problems with stiffness constraints.’’ Finite Elements in Analysis and Design 21: 239–251. Huang, X. and Xie, Y.M. (2007). ‘‘Convergent and meshindependent solutions for bi-directional evolutionary structural optimization method.’’ Finite Elements in Analysis and Design 43(14): 1039–1049. Huang, X. and Xie, Y.M. (2009). ‘‘Bi-directional evolutionary topology optimization of continuum structures with one or multiple materials.’’ Comput. Mech. 43: 393–401. Querin, O.M., Young, V., Steven, G.P. and Xie, Y.M. (2000). ‘‘Computational efficiency and validation of bi-directional evolutionary structural optimization.’’ Comput. Meth. appl. Mech. Engng. 189: 559–573. Rietz, A. (2001). ‘‘Sufficiency of a finite exponent in SIMP (power law) methods.’’ Struct. Multidisc. Optim. 21: 159–163. Rozvany, G.I.N. (2009). ‘‘A critical review of established methods of structrual topology optimization.’’ Struct. Multidisc. Optim. 37(3): 217–237. Sigmund, O. (2001). ‘‘A 99 line topology optimization code written in Matlab.’’ Struct. Multidisc. Optim. 21: 120–127. Stolpe, M. and Svanberg, K. (2001). ‘‘An alternative interpolation model for minimum compliance topology optimizaiton.’’ Struct. Multidisc. Optim. 22: 116–124. Xie, Y.M. and Steven, G.P. (1997). Evolutionary Structural Optimization, London: Springer. Yang, X.Y., Xie, Y.M. and Steven, G.P. (2005). ‘‘Evolutionary methods for topology optimisation of continuous structures with design dependent loads.’’ Computers and Structures 83: 956–963. Yang, X.Y., Xie, Y.M., Steven, G.P. and Querin, O.M. (1999). ‘‘Bidirectional evolutionary method for stiffness optimization.’’ AIAA Journal 37: 1483–1488. Zhou, M. and Rozvany, G.I.N. (1991). ‘‘The COC algorithm, Part II: topological, geometrical and generalized shape optimization.’’ Comput. Meth. appl. Mech. Engng. 89: 309–336.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Investigating the buckling behaviour of single layer dome form of space structures Z. Zamanzadeh Islamic Azad University, Bostanabad Branch, Bostanabad, Iran

H. Abdolpour Islamic Azad University, Ahar Branch, Ahar, Iran

A. Behravesh Islamic Azad University, Mahabad Branch, Mahabad, Iran

ABSTRACT: In the present paper, the buckling behaviour of single layer geodesic and reticulated dome form of space structures is investigated. The types of buckling concerned here are the general buckling, the local buckling and the buckling of a member. As to the geometric parameter of a dome, the slenderness factor S is adopted which represents the openness and slenderness of the dome. The buckling loads are computed by the linear and nonlinear buckling analysis using the finite element method. The main point of interest is earning relations for estimating critical buckling load of single layer geodesic and reticulated dome form of space structures. By using these relations, buckling loads of domes could be estimated accurately without any need to time-consuming non-linear analysis and complicated mathematical calculations. 1

INTRODUCTION

Single layer space structures are often used in moderate span buildings, sport halls and exhibition centers (Sadegi 2004). As these structures have less bending stiffness, a sizable amount of work has been dedicated to assess the buckling behaviour of single layer space structures. Issues such as investigating the static stability of space structures using genetic algorithm (EI-Lishani et al. 2005), estimation of buckling loads of Elliptic Parabloidal single layer lattice domes under vertical loads (Kato et al. 2006) and optimization for maximum buckling load of a lattice space frame with non-linear sensitivity analysis have been thoroughly investigated (Chen et al. 2006). However there is still an uncertainty at the time of performing the buckling analysis of single layer dome form of space structures. The reason is that the approximate design criteria for buckling analysis established by the different codes are generally applicable to planner frames, but uncertain for space structures. In the present paper the criteria of the general buckling, local buckling and the member buckling are given in a simple form that are found to be function of a geometric characteristic parameter of a dome. Furthermore numerical and experimental studies of the buckling behaviour of geodesic and reticulated dome form of space structures have been carried out. In these studies, the geometrical nonlinearity due to large displacements could be included. The results of linear and non-linear buckling analysis

399

have been compared by finite element method and the relation of buckling load of geodesic and reticulated domes has been revealed through these studies. In order to arrive to these goals, the mathematical model of the structures must be selected, that sufficiently be able to show the act of the real model. Therefore, the geometrical modeling of structures has been performed using the written software by the researchers and coordinates of dome’s joints have been gained. 2

STRUCTURAL MODELS FOR DOMES AND MODELING FOR MEMBERS

Figures 1a and 1b show the shape of rigidly jointed geodesic and reticulated domes simply supported at the boundary nodes. The half meridian arc is divided into three (3P), four (4P) and five (5P) grids. A slenderness factor (α) is defined as a geometric parameter of a dome (Hangai et al. 1988).

Figure 1a.

Dome geometry.

3P reticulated dome

4P reticulated dome

3P geodesic dome Figure 1b.

4P geodesic dome

5P geodesic dome

Dome geometry.

Table 1.

Dimensions of domes.

S

ψ(d)

R(cm)

a(cm)

H(cm)

17 13 27 23 36 35 32 36

9 12 18 24 30 36 42 48

66583 49943 33307 24989 2000 16676 14303 12525

1047 1035 1022 1014 1000 980.2 954.7 930.8

82 109.2 163 216 268 318.5 367.4 414.4

Table 2.

Member properties.

Model

E(MPa)

A(c)

I(cm4 )

L(cm)

3P 4P 5P

2.06 × 05 2.06 × 05 2.06 × 05

17.7 17.7 17.7

758.036 240.03 98.31

348.6 261.5 209.2

S= √

5P reticulated dome

L r·R

(1)

where R is the dome radius, L is the typical member length and r is the radius of gyration. In order to earn comprehensive relations for estimating buckling loads of single layer geodesic and reticulated domes, considering domes with different radius and frequencies was necessary. The dimensions of the domes with the span = 2a, the radius = R, the height = H and the half open angle = ψ are shown in table 1.

The member properties of the tubular section as shown in table 2, where E= Young’s modulus, A = the sectional area and I = the moment of inertia.

3 3.1

BUCKLING CHARACTERISTICS OF DOMES Method of analysis and loading

Two methods of finite element analysis are used here to determine the buckling load. One is the linear buckling analysis containing linear static analysis and linear eigenvalue buckling analysis that is known as classical Euler buckling analysis and other is the geometric nonlinear analysis combined with the eigenvalue analysis repeated for different load levels. Nonlinear buckling analysis is more accurate than eigenvalue buckling analysis because it employs nonlinear large deformation static analysis to predict buckling loads. It gradually increases the applied load until a load level is found whereby the structure becomes unstable (Wen et al. 2006). The true nonlinear nature of this analysis thus permits the modeling of initial imperfection. The uniform load is applied vertically at each of the free joints as shown in figure 2. 3.2

Buckling load and mode

Figures 3a and 3b give the buckling load q(N/cm2 ) obtained by the linear eigenvalue analysis and the geometric nonlinear analysis for 3P, 4P and 5P models of geodesic and reticulated domes. The buckling load increases in proportional to the slenderness factor S. the typical buckling modes of 3P model are also shown in Figures 4a, 4b and 4c.

400

Figure 2.

Uniform loading.

12

q ( N / c m^ 2 )

10

Figure 4a.

Buckling mode of (3P-9).

Figure 4b.

Buckling mode of (3P-24).

Figure 4c.

Buckling mode of (3P-42).

3PL

8

4PL 5PL

6

3PN 4PN

4

5PN

2

0 0

1

2

3

4

5

S

Figure 3a.

Buckling load of reticulated domes.

10

9

8

7 3PL

q ( N / c m^ 2 )

6

4PL 5PL 5

3PN 4PN

4

5PN

3

2

1

0 0

1

2

3

4

5

S

Figure 3b.

Buckling load of geodesic domes.

The general buckling modes are observed for 3P9(3P model, ψ = 9◦ ) and 3P-12. As the slenderness factor S increases, the dimple buckling and finally the buckling of a member are recognized. The buckling load of 4P and 5P models is smaller than that of 3P models for every values of S. This is caused by the flexibility of the dome with the increasing number of grids. The buckling model of 4P and 5P models do not change compared with that of 3P models in the same range of S. Consequently, the buckling characteristics of reticulated and geodesic domes are classified into three types according to the slenderness parameter. The ratio of the linear buckling load to the non-linear buckling load is given in Figures 5a and 5b. The linear buckling load is 1.3 to 1.7 times large as the nonlinear buckling load for the dome of S < 3.0. As to the dome of S > 3.0 the ratio of linear buckling load to the nonlinear buckling load is approximately

401

1.0 this means that the linear buckling load is the good estimate of the nonlinear buckling load in this range of S, where the buckling of a member is observed. 3.3

Critical constant of the buckling load

Following the buckling formulas given in Reference (Linda 1969), the critical constant of the buckling load C is defined as follows. √ q = 4CE AI /R2 L

(2)

1.2

1.8 1.6

1

0.8

1.2 3P

C3R

1

C

q Linear/q Nonlinear

1.4

4P

C4R

0.6

0.8

C5R

5P 0.4

0.6 0.4

0.2

0.2 0

0

0

0

1

2

3

4

1

2

5

3

4

5

S

S

Figure 5a. Comparison of buckling load by linear and nonlinear analysis (geodesic domes).

Buckling load coefficient (reticulated domes).

buckling load is estimated in a simple form that is found to be function of slenderness factor of dome as follows. For geodesic domes:

1.8 1.6 1.4 q Linear/q Nonlinear

Figure 6b.

1.2 3P 1 4P 0.8 5P 0.6

 C = 0.947067 C = 0.316226 S 2 − 2.496532 S + 5.55259

0.4 0.2

S≺3 S≥3 (3)

0 0

1

2

3

4

5

S

For reticulated domes:

Figure 5b. Comparison of buckling load by linear and nonlinear analysis (reticulated domes).

 C = 937058

S≺3

C = 0.317413 S 2 − 2.5551699 S + 5.76457 S ≥ 3 (4)

1.4

1.2

3.4

1

0.8

C

C3G C4G C5G

0.6

0.4

0.2

0 0

1

2

3

4

Figure 6a.

⎧ √ 2 ⎪ ⎨3.788E AI /R L

5

S

Buckling load coefficient (geodesic domes).

Estimation of the buckling load relation

Substituting equations 3 & 4 into equation 2, The elastic buckling load of the reticulated and geodesic dome form of space structures is summarized as follows. For geodesic domes: S≺3 q = (1.264904 S 2 − 9.9861 S + 22.21036) ⎪ ⎩ E √AI /R2 L S≥3 (5)

Figure 6 gives the critical constant C obtained by the nonlinear analysis. Critical constant C is approximately 0.9 for the both geodesic and reticulated domes of S < 3.3, and decreases in proportional to S. The broken line shows the buckling criteria of a member resulting from direct forces. By curve fitting of various values of critical buckling constants in fig. 5, the critical constant of the

For reticulated domes: ⎧ √ 2 ⎪ ⎨3.748E AI /R L

S≺3 q = (1.269652 S 2 − 10.2206 S + 23.05831) ⎪ ⎩ E √AI /R2 L S≥3 (6)

402

4

CONCLUSIONS

The statistical analysis of elastic buckling load of rigidly jointed single layer reticulated and geodesic domes has been developed using the linear eigenvalue analysis and the geometrical nonlinear analysis. Summaries obtained may be as follows. 1. According to the buckling characteristics of the rigidly jointed single layer dome form of space structures, the general buckling modes are observed for S < 2.5. As the slenderness factor S increases, the dimple buckling and finally the buckling of a member are recognized. 2. The buckling load is well estimated by the linear eigenvalue analysis for the both geodesic and reticulated domes of S < 3.0. The reason may be due to the fact that the buckling mode obtained by the nonlinear analysis shows the finite vertical displacement of the nodes that do not appear by the linear eigenvalue analysis. It may conclude that the buckling sensitivity is understood to some extend by the linear eigenvalue analysis except the dome that shows the member buckling mode. 3. The static buckling load of the reticulated and geodesic dome form of space structures is expressed by Equations 5 and 6. REFERENCES Chen, P. & Kawaguchi, M. 2006. Optimization for Maximum Buckling Load of a Lattice Space frame with Nonlinear Sensitivity Analysis. International journal of space structures. Vol. 21, No. 2: 111–118.

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EI-Lishani, S., nooshin, H. & Disney, P. 2005. Investigating the Statical Stability of Pin-jointed Structures Using Genetic Algorithm. International journal of space structures. Vol. 20, No. 1: 53–68. Hangai, Y., Takayama, M. & Ohya, S. 1988. Comparison of Buckling Behaviours of Reticulated Single-Layer Domes and Reinforced Concrete Domes, Domes: Proceeding of the IASS-MSU International Symposium, Mimar Sinan University, Istanbul: 453–462. Kato, S., Ueki, T. & Nakazawa, S. 2006. Estimation of Buckling Loads of Elliptic Paraboloidal Single Layer Lattice Domes under Vertical Loads. International journal of space structures. Vol. 21, No. 4: 173–182. Lind, N.C., 1969. Local Instability Analysis of Triangular Dome Framework. The Structural Engineer. 47(8), pp. 317–324. Sadegi, A. 2004. Horizontal Earthquake Loading and Linear/Nonlinear seismic Behaviour of Double Layer Barrel Vaults. International journal of space structures. Vol. 19 No. 1: 21–37. Wen, P., Aliabadi, M. & Young, A. 2006. Post Buckling Analysis of Reissner plates by the Boundary Element Method. The journal of strain Analysis for Engineering. Vol. 41, No. 3: 239–252.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Reasoning on structural timber design for target reliability L. Ozola Latvia University of Agriculture, Jelgava, Latvia

ABSTRACT: The paper presents a consideration on some topical problems dealt with design of timber structures related to the safety. Basing to the results of probability calculations of load bearing capacity values regarding the variability of factors affecting the behaviour of timber elements a partial factor has been proposed for stability conditions. The toughness indice as a criteria for the comparable assessment of structural compositions has been suggested. A reasonable consistency between the real structure and the design model has been discussed according to a practical example which illustrates the case when hard consequencies may be expected due to disability of the design model defined to reveal a more unfavourable internal forces in system. The study is aimed for the refining the understanding in design and to promote an extensive judgement of structural solutions accepted, especially when the structure has been designed for covering wide spans of public building areas. 1

INTRODUCTION

Serious theoretical and experimental research works devoted both to the evaluation of strength and stiffness properties of wood materials and the analysis in detail of the behaviour of structural elements in complex loading situations promoting the onset of ultimate limit state have been carried out by the researchers worldwide during the last decades. However due to uncertainty the value of bearing capacity remains normally in a significantly wide interval around the deterministic value calculated according to the code conditions. And the question about the structural reliability of timber structures becomes more urgent than it is for steel and concrete structures. As regard to the timber structures the complexity of reliability problem is coincident with the wood specific properties and significant variability of strength and stiffness data sets. Regardless of the strong standard methods used for the determination of the characteristics a variation conserves peculiar to a design values in some range. Classification and quality requirements for structural wood products only partly reduce the variability of characteristics. Thereby structural timber design causes a number of problems dealt with target reliability which are not encountered for steel and concrete structures, at least not to a significant degree. The structural failures, although rarely being real, should be avoided generally, and may occur for various reasons including the defective materials and/or an attempt excessively saving materials, errors in design, empty areas in design codes, poor supervision and quality control carelessly. Each assessment method contains some degree of error, too. Structural engineer should comprise an available information,

405

computational methods and accumulated data from the experience in the best way to carry out more effective solution as it is proposed by the known authorities (Ditlevsen & Madsen 2005) in the reliability theory and methodology: Engineering judgment is the art of being able to decide whether results obtained from a structural analysis or design model is sufficiently realistic that the engineer dare base his or her practical decisions on these results’’. The sensitivity of a structure to different types of defects and errors is not the same for steel, concrete and timber structures. Some questionable topics in structural design area dealt with the conformity of a design model to the realistic behaviour of system and the variation of property values have been studied in this paper related to the safety assurance problem of timber structures particularly. The objective of this paper is to point out some specific features and characteristics for assessment of structural compositions, to promote the implementation of probability-based design for wide span structures, making it possible to ensure the safety of timber structures, and thus to improve the quality of designs and constructions subsequently. 2

VARIATION OF BEARING CAPACITY OF STRUCTURAL TIMBER ELEMENTS

The confidence of deterministic capacity value is always open to the question, particularly for timber structures. Every numerical quantity involved in the calculation of bearing capacity has been characterized by the variation in a specific range. Evidently for more adjustment factors introduced a wider range of capacity values may be expected, and thus the reliability

of structure may be reduced. An aspect of current chapter is to prove this significance by some practical example. The probability calculations of load bearing capacity values for structural timber elements has been carried out by assuming that an independent variation of the factors involved, i.e., geometric characteristics and material resistance and values of stiffness modulus exists. Using the results provided by some former investigations on the variation of wood strength (Ozola 2005) and of geometric characteristics of structural timber elements (Keskküla & Ozola 2001) the probabilistic expectations of load bearing capacity for structural timber elements have been obtained. The factors involved are assumed to be the representatives in a normal data set characterised by the coefficient of variation (COV) outlined in Table 1. The probability of bearing capacity value for solid timber beams in normal bending and longitudinal buckling for columns is given by the equations (1) and (2) as follows

0.25 0.20 0.15

0.4 0.2

1.3

1.2

1.1

1.0

0.9

0.8

0.7

0

0.10

1.0

0.08 0.07 0.06 0.05 0.04 0.03

0.8 0.6 0.4

2.0

1.8

1.6

1.4

1.2

1.0

0 0.8

0.2

0.01 0 0.4

0.02

0.6

(2)

Cumulative probability

0.09

b) Relative capacity Fdi/ Fd

0.10

1.0

0.09

0.07

0.05 0.04

Coefficient of variation

0.03

0.8

Cumulative probability

Probability P(Mcri/Mcr)

0.08

Statistical characteristics for input data samples.

0.6

0.4

0.02

0.025 0.05

0.2

0.01 0

0.05 0.05 0.05 0.05 0.20

1.8

1.6

1.4

1.2

1.0

0.8

0.6

0.4

0

0.15

0.3

Design strength value for solid wood Characteristic value of modulus of elasticity for wood Characteristic value of shear modulus for wood Design section area value Design section modulus value Design modulus of inertia value Design torsion modulus value for glulam section Effective length in buckling

0.6

a) Relative bending capacity Mdi/ Md

0.06

Variable

0.8

0

where P(MR,i /Md ) = probability of relative bearing capacity in bending; P(FR,i /Fd ) = the same in longitudinal buckling; P(fi /fm,d ) and P(fi /fc,0,d ) are probabilities of relative bending and compression strength values correspondingly; P(Wi /Wd ) and P(Ai /Ad ) are probabilities of geometric characteristics; P(kc,i /kc ) = probability of relative buckling factor value; fm,d , fc,0,d , kc , Ad , Wd are the design values (Eurocode 5). The outcomes of numerical tests of relative bearing capacity values in bending and longitudinal buckling of timber elements have been analysed and distributions are presented in Figure 1.

Table 1.

Cumulative probability

Probability P(Mdi/Md)

0.30

0.05

(1)

P(FR,i /Fd ) = P( fi /fc,0,d ) · P(Ai /Ad ) · P(kc,i /kc )

0.35

0.10

Probability P(Fdi/Fd)

P(MR,i /Md ) = P( fi /fm,d ) · P(Wi /Wd )

1.0

0.40

c) Relative effective moment capacity M cri/ Mcr

Figure 1. Probability distribution of relative load bearing capacity values in bending (a) and longitudinal buckling of solid timber elements (b) and for effective moment capacity in lateral buckling of glulam beams (c).

406

structural solutions when there are no specific indices implemented for the characterisation of structural efficiency as regard safety precautions near limit state. Despite the proper account of code conditions it would be reasonable to consider the potentials of a structure when being subjected to overloading and/or unfavourable service conditions. Progressing limit state procedures may be different in forms and for the velocities of collapse. Evidently, the growth of plastic deformations in compression always takes more time and the rupture in normal bending as well. The brittle failure in tension or shear occurs unexpectedly without any warnings through deformations. For the shortest time some more dangerous failure modes proceed when the column has been collapsed in longitudinal buckling or the loss of stability of compressed web take place when narrow beam section has been turned out from the bending plane. Taking into consideration the reasons above it is proposed to involve two more comparable criteria for a closer inspection of structural designs.

Statistics of relative capacity data generated.

Limit state

Mean value

Coefficient of variation

Lower capacity

Partial factor

Normal bending Strength Buckling

0.998 0.997

0.053 0.207

0.974346 0.903982

1.02 1.10

The probability of bearing capacity in lateral buckling for glulam timber elements is given by:  P

MR,i Mcrit



 =P

ef ef ,i



 P

Ei Gi Jz,i Jtor,i E05 G Jz Jtor

 (3)

where P(MR,i /Mcrit ) = probability of effective bending moment in relative terms; P( √ef /ef ,i ) = probability of relative buckling length; P ((Ei /E05 )(Gi /G)) = probability of root √ value from product of relative stiffnesss modulus; P ((Jtor,i /Jtor )(Jz,i /Jz )) = the same for the product of relative torsion and inertia modulus. The relative capacity data scatters obtained for the strength in bending and stability ensured for both longitudinal and lateral buckling results being equal taking into account the variation of input data values have been characterised by general statistics –mean value, coefficient of variation (COV ) and lower capacity value RC 99 at the lower bond of 99% confidence interval (Christensen 1996) as presented in Table 2. Partial factor γb represents the ratio of mean and lower bond values. While common design practice is based on the deterministic codes we need to involve some partial factors to the best of safety needs. The probabilistic analysis of load bearing capacity data distributions proves that there is a substantiated reason for the correction of partial safety factors regarding the complexity of element’s behaviour and of different material properties varying in different scatters. 3

3.1

Limit state toughness of structure

The modulus of toughness has been outlined as an important property found for structural material when we need the measurable indice for fracture mode (Hibbeler 2003). The part of area under stress strain diagram measured from the proportional limit up to the point of failure (Fig. 2) may be defined as limit state toughness uf . It is clear that there is a more detailed necessary consideration in order to define the certain specific

CONSIDERATION ON STRUCTURAL COMPOSITION

Comprehensive wood production technologies have achieved a high level of capacity. At present there is no problem to produce the glue laminated construction units of extremely large size sections. And computer programs provided enable to carry out the complicated calculations during a short time. Why the problems arise when doubtful solutions are advanced for the construction, as well as collapses of structures take place? Unfortunately sometimes accidents take place in large span beam type structures designed seemingly correctly according to the code conditions. This is a question for many practical discussions on

407

60 Normal stress, MPa

Table 2.

50 uf 40 30 20 uf 10 0 0

10

20

30

Strain of longitudinal grain, ε x 104 bending Figure 2.

tension

Definition of limit state toughness uf .

values for proposed toughness indice regarding both to the failure mode and the anticipated consequences of collapse, nevertheless that may be used for the assessment and comparison of different variants. The comment for clarifying the above mentioned: in safety meaning as a suggested structural solution out of different ones may be chosen that for which an onset of the limit state is expected to be accompanied by progressive growth of plastic deformations. 3.2

A rough value of strain energy per one glulam beam (grade Gl28h) structure has been calculated by using the equation (4) and resulting in u = 38.5 kNm. L u= 0

Mechanical energy stored by structure

V2 dx 2GA

(4)

0

u=

n  Ni2 2EA

(5)

i=1

where Ni = axial force of individual bar. It is proposed to use the strain energy as an indice for making decision in a structural design. It is significant to provide the minimum amount of mechanical energy to be stored in some height from the ground level. 2420

4200

6000

30000

3000

b) 6000

200x176 200x308

L

where M , V = bending moment and shear force for corresponding beam section; J , A = geometric characteristics of corresponding beam section–moment of inertia about stiff axis and area; E, G = stiffness modulus (elasticity and shear) of material. The strain energy stored by truss system has been compiled by the values supplied as the elastic strain energy of bars in compression or the tension by using the formulae (5) resulting in a total value of u = 17 kNm per one truss what is more than two times less when comparing with the beam structure.

Strain energy stored by the system under action of external loads may be proposed as one more applicable criteria considering the effectiveness and safety of structure. The significant differences in values of strain energy comprised by bearing structure may be illustrated by the theoretical test of two structural solutions designed for covering the span L = 30 m (Fig. 3). The design value of permanent and variable (snow) load combination is q = 14 kN/m.

a)

M2 dx + 2EJ

30000

Figure 3. Structural systems: (a)—glue-laminated timber post-beam frame, (b)—glue-laminated timber truss frame.

4

DISCUSSION ON DESIGN MODELS

A more serious task for a designer is the definition of an adequate structural model. Indeed, it is not possible to achieve a complete similarity between the assumed design model and the real timber structure, however we should need a maximum completion. In this regard it is useful to study some practical design in detail. Some time ago the world was troubled by the event of building collapse (Sennewald, 2008). It is useful to analyse the results of statics for some system (not correctly the same) which is composed for simulation of a box type beam system (Fig. 4a) by the space lattice structure including bracings (Fig. 4c). The lattice system has been treated by using the program ‘‘Analysis for Windows’’. Two loading variants have appeared giving the results exceptionally intriguing for discussion: 1. the full vertical load q = 18 kN/m applied to frame systems on axis 6–10, decreased load qn = 13 kN/m applied to other frames, and the upper chord of frontal frame is loaded by the horizontal force caused by moderate wind w = 3 kN/m 2. vertical load q = 18 kN/m applied to the frame systems at the middle part (on axis 4 to 7) and load qn = 13 kN/m applied to other frames.

408

Upper chord

a)

Outplane bending moment action

d)

Web element simulated as beam type with stiff ends

l a

Bracing element assumed having hinged ends Bottom chord

h

L

b) a

ho

h

e)

c)

ho

b

q

w

10

s

8 7

6 5

L 4 3

More stressed element in outplane bending

2 1

Figure 4. Design model of timber post-beam system: a)—glue-laminated timber post-beam frame, b)—section of main beam, c)—similar-sized lattice structure, d)—unit cut from system with linear and rotational displacement lines, e)—more stressed part in outplane bending; L = 40 m, a = 4 m, s = 7.5 m, h = 2870 mm, h0 = 2670 mm, b = 280 mm, l = 8.8 m.

The values obtained and the distribution of internal forces within the proposed lattice model enable to disclose the expectations of weakest section location in a real structure, and to evaluate the values of internal stresses more correctly.

409

So, the results of greater interest from this large numerical test are as follows: – generally strength conditions for cantilever beam with box type section are satisfied if only the stability

of compressed chord has been ensured by the roof structure; the normal stresses in chords are about 10 MPa – the statistical calculation of a proposed space lattice structure under the first loading situation encountering the moderate wind action prove that the maximal outplane bending moment arises in the elements of windward frame at a small distance from columns and reaches the value of 30.3 kNm which produces approximately 11.6 MPa of additional compression stresses in upper chord (Fig. 4c, 4e), total normal stress peaks up to 21.9 MPa – the second loading situation shows that maximal normal stresses in chords attain the value of 18 MPa at the middle span elements, yet the outplane bending moment achieves the value of 7.2 kNm producing approximately 3 MPa of additional stresses in the bottom chord near support on column. It is clear that neglecting the effects produced by the lateral actions may be the factor leading to some hard problems near the limit state for plane structure especially when large scale depth/width ratio of beam section has been accepted. Also it is useful to consider on suitability of box type sections having empty area around the stiff axis. As follows from the previous static trials in specific loading situations, the stresses arose by the outplane bending moment may exploit almost full strength capacity of webs (usually made of weaker material than chords), thereby destroying stiffness of connection between parts of section and producing instantaneous collapse. In the case of I-type section in use the situation would be another, and limit state toughness may be expected higher. There are some other questions related to adequacy of design model not discussed in this paper in detail yet one more comment as follows. Neglecting of joint stiffness effects during design of timber structure may be factor leading to hard in service problems the most frequently occurring. Considering existing uncertainties in rotational stiffness of joints it is significant to do compilation of design model being critical for structure in order to check for maximal values of internal forces and more unfavourable loading situations for heavy loaded elements and joints. 5

CONCLUSIONS

– The refining analysis of designs and probabilitybased evaluation of bearing capacity for timber structures designed and/or erected promotes the desired improvements in the safety assurance.

– Distinctive variation scatter of bearing capacity values inherent high complexity as regard to the composition and/or the mechanical model and/or the variation of material properties suggest that the probability-based analysis for structures is needed. Otherwise it is proposed to supplement the stability conditions with one more partial factor regarding complexity of stress state and factors affecting. – The structural compositions comprising minimal amount of mechanical energy and expected for progressing the plastic deformations when limit state goes on, must be advanced for construction. – It is thoughtless to use the large span timber beams of extremely narrow sections without refined structural analysis of whole space structure taking into account every practicable loading situation. – One important point to remember is that generally in code provisions the critical moment methodology has been elaborated considering the bending moment in one plane only. When some lateral force components are transferred to the beam from the bracing elements, the extreme loading situation arises reducing the flexural rigidity of beam significantly. – The conservative approach should be maintained in design of timber structures— to perform all really possible design models in order to find the unfavourable combinations and the values of internal forces and displacements. REFERENCES Christensen, R. 1996. Analysis of variance, design and regression. Applied statistical methods. London, . . . [etc.]: Chapman & Hall. Ditlevsen, O. & Madsen, H.O. 2005. Structural reliability methods. Copenhagen: Technical University of Denmark. EN 1995-1-1. Eurocode 5-Design of timber structures—Part 1-1: General—Common rules and rules for buildings. Brussels: CEN. Hibbeler, R.C. 2003. Mechanics of materials. Singapore, London . . . [etc.]: Prentice Hall. Keskküla, T. & Ozola, L. 2001. Factors’ variability in respect to reliability of timber framings. IABSE Reports, Volume 85. IABSE Conference. Innovative wooden structures and bridges: 131–135. Ozola, L. 2005. Statistical estimates of strength and stiffness properties of timber. Proceedings of the conference on probabilistic models in timber engineering. Tests, models, applications: 159–167. Sennewald, R. 2008. Der Einsturz der Eissporthalle in Bad Reichenhall—Wie hätte er verhindert werden können? Fachtagung Bauwerksdiagnose 2008—Vortrag 08. Praktische Anwendungen Zerstörungsfreier Prüfungen und Zukunftsaufgaben: 1–10.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Shape optimization of shell structures with variable thickness M. Kegl & D. Dinevski University of Maribor, Slovenia

B. Brank University of Ljubljana, Slovenia

ABSTRACT: This paper presents an effective approach to shape optimal design of statically loaded elastic shell structures. The shape parameterization is based on a design element technique. The chosen design element is a rational Bézier body, enhanced with a smoothly varying scalar field. The scalar field of the design element is obtained by attaching to each control point a scalar quantity, which is an add-on to the position and weight of the control point. This scalar field is linked to the shell thickness distribution, which can be optimized simultaneously with the shape of the shell. For the analysis of the structure a nonlinear 4-node shell finite element formulation is utilized. The presented optimization approach assumes the employment of a gradient-based optimization algorithm and the use of the discrete method of direct differentiation to perform the sensitivity analysis. A numerical example is presented to illustrate the performance of the proposed approach. 1

INTRODUCTION

2

In order to find effective shapes of shell structures many ideas have been investigated and proposed (see e.g. references in Bletzinger 1999). The shape optimization approach discussed in this paper is based on a design element technique. As a design element a rational Bézier body is chosen. The thickness of the shell may also be variable. This is achieved by enhancing the design element with a smooth scalar field and by further linking it with the shell thickness. The scalar field is introduced by attaching an additional scalar quantity (called value of the control point) to each control point of the Bézier body and by employing the interpolation functions of the design element. By adopting the above approach, the shape of variable thickness shell structures can be optimized in a unified and relatively simple way. The design variables may be related to the positions, weights and values of the design element control points as well as to the cross-sections of truss finite elements. The optimization is assumed to be performed by a gradient-based optimization algorithm and the sensitivity analysis is done by the discrete approach of direct differentiation. The outline of the paper is as follows: in Section 2 the employed shell finite element is presented very briefly. Section 3 discusses the shape and thickness parametrization concept. In Section 4 the optimization problem and its solution are addressed briefly, and Section 5 presents a numerical example. The approach addressed in this paper is presented in much more detailed form in Kegl and Brank (2006).

411

THE EMPLOYED SHELL FINITE ELEMENT

Nonlinear shell finite element that may undergo large displacements and large rotations (but small strains) are used in the optimization process. The 4-node shell finite element is based on an ANS (assumed natural strain) concept of Bathe and Dvorkin (1985); see also Brank et al. (1995). Finite rotations are parameterized by a constrained rotation vector, e.g. Brank and Ibrahimbegovi´c (2001), which is consistent with the standard incremental solution scheme for nonlinear problems. Since it maintains additive iterative rotational updates it is also very suitable for optimization problems (Ibrahimbegovi´c et al. 2004). The usual 4-node ANS shell element formulation is modified to suit the analysis of shells with variable thickness. According to the isoperimetric concept, the position vector of a shell middle surface point r, the normal vector to the middle surface n and the shell thickness t are given (for the undeformed configuration) as: r=

4  n=1

Nn rn ,

n=

4  n=1

N n nn ,

t=

4 

N n tn

n=1

(1) where Nn = 1/4 (1 + ξ ξn ) (1 + ηηn ) , n = 1, . . . , 4, are the usual bi-linear interpolation functions with (ξn , ηn ) ∈ {(−1, −1) , (1, −1) , (1, 1) , (−1, 1)}. The geometry of the element is fully specified by Eq. (1) and by the position of its nodes rn , nodal normals nn and nodal thicknesses tn , n = 1, 2, 3, 4 (see Figure 1).

Shell FE

nn tn Figure 1. element.

Design element

rn

Geometrical quantities of truss and shell finite

These quantities are needed for computation of the internal force vector Fe , the loading vector R e and the tangential stiffness matrix K e of the element. For further details please refer to Kegl and Brank (2006).

3

Figure 2.

A body-like design element.

Control point Pijk

s3

s2

SHAPE PARAMETRIZATION

In order to parameterize the shape of the structure, the design element technique is employed with a rational Bézier body as a chosen design element (DE). The Bézier body exhibits good shape flexibility, automatic prevention of excessive shape oscillations and possible representation of classical shapes. A body-like DE (see Figure 2) is preferred over a surface-like DE in order to enable for more complex structures (Kegl and Brank 2006). The Bézier body is defined by a topologically rectangular scheme of N1 × N2 × N3 control points Pijk . Conventionally, the attributes of each control point Pijk are its position qijk and the corresponding weight wijk . Here, an additional scalar quantity hijk , called the value of the control point Pijk , is attached to each control point (see Figure 3). By adopting such an arrangement, the shape of the Bézier body is still defined in the conventional way (by the control point positions and weights). The above mentioned values of the control points are used to introduce a smooth scalar field, defined over the whole Bézier body, which can be linked for example to the shell thickness. The DE shape depends on the design variables b1 , b2 , . . . , bN , assembled in the vector b ∈ RN . This is achieved by assuming that qijk , wijk and hijk may depend on b. A Bézier body has a local curvilinear coordinate system with its coordinates s1 , s2 and s3 running from 0 to 1 (see Figure 3). Thus, vector s = [s1 s2 s3 ]T defines a point in the DE coordinate system. For any point s there is an image r in the real 3-D space: N1 N2 N3 r = r (s, b) =

i=1

N1 N2 N3 k=1 Bi Bj Bk wijk qijk N1 N2 N3 j=1 k=1 Bi Bj Bk wijk

s1 Figure 3. Control point Pijk , its position qijk , weight wijk and value hijk .

and the corresponding scalar quantity t N1 N2 N3 t = t (s, b) =

i=1

N1 N2 N3 k=1 Bi Bj Bk wijk hijk N1 N2 N3 j=1 k=1 Bi Bj Bk wijk

j=1

N1 N2 N3 i=1

(3) where BiN1 = BiN1 (s1 ), BNj 2 = BNj 2 (s2 ) and BkN3 = BkN3 (s3 ) are the Bernstein’s blending polynomials of the orders of N1 − 1, N2 − 1 and N3 − 1, respectively. When using body-like DE for parameterization of a shell structure, it is convenient that the shell middle surface coincides with some parametric surface of the DE. Here, the shell middle surface is defined as s3 = s3c surface of the DE, where s3c is any fixed value between 0 and 1. It follows from Eq. (2) that a point on the shell middle surface can be expressed as r = r(sc , b), where sc = [s1 s2 s3c ]T . By using Eq. (3), a smoothly varying thickness of the shell is given as t = t(sc , b). And the shell director vector field n, being normal to the middle surface, is given as: 

j=1

N1 N2 N3 i=1

Design element

n=

(2)

412

∂r (sc , b) ∂r (sc , b) × ∂s1 ∂s2

   ∂r (sc , b) ∂r (sc , b)   ×  ∂s ∂s2  1 (4)

By the proposed design element technique, the position of a particular finite element node is fully specified by its position s in the parametric space (Kegl and Oblak 2000). Once the position s of the shell finite element (SFE) node is known, Eq. (2) is used to calculate its position r in the real 3-D space for any given values of the design variables b. By using Eq. (3) the thickness of the shell at the node under consideration can be calculated, while Eq. (4) is used to calculate the corresponding nodal shell director. Thus, the shape of the structure can be varied by varying the quantities qijk , wijk and hijk . For further details refer to Kegl and Brank (2006).

4

FORMULATION AND SOLUTION OF THE OPTIMAL DESIGN PROBLEM

An optimization problem of a statically loaded structure can usually be written in the following form:

s.t.

min f0 fi ≤ 0, i = 1, . . ., M bLi ≤ bi ≤ bU i = 1, . . ., N i ,

(5)

where f0 denotes the objective function, fi are the constrained quantities, the symbols bLi and bU i denote the lower and the upper limits of the design variables. The response variables (nodal displacements and rotations) are related to the design by the structural equilibrium equation: F−R =0

Figure 4. Ground plan of the structure and the positions of the control points.

(6)

where F and R denote the vectors of structural internal and external forces. Since the design variables b are continuous, and the equilibrium equation can be differentiated with respect to b, a gradient-based optimization algorithm can be used to solve the optimization problem efficiently. For the details refer to Kegl and Brank (2006).

The structure is loaded by its own weight and by a uniform snow load of w = 5 kN/m2 . The objective is to optimize the shape and the thickness distribution of the shell by minimizing its relative strain energy:

5

where the symbol C Ini denotes the strain energy of the shell at the initial (starting) design (constant thickness of 150 mm). The constraints imposed constraints are related to the vertical positions yA and yB of points A and B respectively (yA ≥ 8 m and yB = 8 m). Additionally, the structural volume is constrained to be equal to V = 60 m3 . A single design element with 5 × 5 × 1 = 25 control points is used to parameterize the structure, Figure 4b. The design variables are related to the control point positions (variable shape) and control point values (variable thickness) as given in Kegl and Brank (2006). For all design variables the lower limit is −1 and the upper limit is 2. The initial and final shapes

NUMERICAL EXAMPLE

The approach described above is illustrated by an example of shape optimization of a free-form shell structure with variable thickness. Full nonlinear response/sensitivity e analysis is performed and the gradient-based approximation method, described in (Kegl and Oblak 1997 and Kegl et al. 2002) is used to optimize the structure. The example is partially taken from Kegl and Brank (2006). The ground plan of the structure is shown in Figure 4. The four corner points of the structure have hinged supports. The material properties are E = 30000 N/mm2 , ν = 0.3 and ρ = 2500 kg/m3 .

 c = C C Ini

413

(7)

Initial

Optimal

Optimal thickness

Figure 5. Initial and optimal shell shapes and optimal thickness distribution.

as well as the final thickness distribution are shown in Figure 5. The optimization process was stable and smooth. The objective function c and maximal constraint violation (MCV) of initial and optimal designs are as follows: cini = 1, MCVini = 0.4999 and copt = 0.1117, MCVopt ≤ 0.

As a concluding remark one can say that the proposed enhanced design element enables consistent shape and thickness variations of the shell structure. The smoothly varying thickness of the shell is introduced in a simple manner by a scalar field defined over the design element. The presented numerical example indicates that the proposed approach performs well and the employed nonlinear shell finite element represents a reasonable choice. REFERENCES Bathe KJ, Dvorkin E 1985. A four-node plate bending element based on Reissner-Mindlin plate theory and a mixed interpolation, Int. J. Numer. Methods Engng.; 21: 367–383. Bletzinger K-U 1999. Structural optimization and form finding of light weight structures. Proceedings of the European Conference on Computational Mechanics (ed. Wunderlich W). Muenchen, Germany. Brank B, Ibrahimbegovi´c A 2001. On the relation between different parametrizations of finite rotations for shells, Engineering Computations; 18: 950–973. Brank B, Peri´c D, Damjani´c FB 1995. On implementation of a nonlinear four node shell finite element for thin multilayered elastic shells. Computational Mechanics; 16: 341–359. Ibrahimbegovi´c A, Knopf-Lenoir C, Kuˇcerova A, Villon P 2004., Optimal design and optimal control of structures undergoing finite rotations and elastic deformations, Int. J. Numer. Methods Engng.; 61: 2428–2460. Kegl M, Brank B 2006. Shape optimization of truss-stiffened shell structures with variable thickness, Comput. Methods Appl. Mech. Engrg.; 195: 2611–2634. Kegl M, Butinar BJ, Kegl B 2002. An efficient gradientbased optimization algorithm for mechanical systems. Communications in Numerical Methods in Engineering; 18: 363–371. Kegl M, Oblak MM 1997. Optimization of mechanical systems: on non-linear first-order approximation with an additive convex term. Communications in Numerical Methods in Engineering; 13: 13–20. Kegl M 2000. Shape optimal design of structures: an efficient shape representation concept. International Journal for Numerical Methods in Engineering; 49: 1571–1588.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Structural damage detection in plates using wavelet transform G. Ghodrati Amiri Iran University of Science & Technolog, Tehran, Iran Islamic Azad University of Shahrekord, Shahrekord, Iran

A. Bagheri, S.A. Seyed Razzaghi & A. Asadi School of Civil Engineering, Iran University of Science & Technology, Tehran, Iran

ABSTRACT: Detection of damage in engineering structures during their service life has received increasing attention from researchers in the last years. Exact structural damage detection is a challenging problem due to the nonlinear behavior of structures, incomplete sensed data and presence of noise in the data. Information on the location and extent of damages could assist in the diagnosis of the structural health conditions. In this study, a new method is proposed to detection of damage in plate structures using discrete wavelet transform. The wavelet transform was developed in the last years. The proposed method is applied to a four fixed supported rectangular plate containing damage with arbitrary length, depth and location. Numerical results of identifying the damage location are compared with the actual results to demonstrate the effectiveness of the method. 1

INTRODUCTION

Detection of damage in civil and mechanical engineering structures during their service life has drawn wide attention during last few decades. Structural damage can be identified as weakening of the structure that cause negative changes in its performance. Damage may also be considered as any change in property of material and original geometry of structure that make undesirable stress or displacement and vibration in structure. Structural damage will progressively impact on the dynamic property of structure such as stiffness and damping at damage location (Doebling et al. 1996). Therefore, these changes cause alteration in dynamic response behavior of structure. Identification of alteration in mode shape and natural frequency in damaged structural element in comparison with pre-damaged state of element is one of the popular methods in damage detection. These changes are often small and measurements are polluted by noise which cause the method be inefficient in detecting proper location of damage, therefore the method needs robust methodology to identify damage location and so forth. The Fourier transform spectral analysis methods is often used to analyze signal characters thus the lack of flexibility to deal with such a local changes and discontinuity in the mode shape of damaged element exhibit inefficiency of this method to damage detection. The wavelet transform is a new method to precisely analyze the signals, which overcomes the problems that other signal processing techniques exhibit. Applying wavelet transform to analyze the damaged mode shape of structure produce satisfying result

415

in damage identification. Sharp changes in wavelet coefficient near the damage exhibit the presence and location of damage. The literature on wavelet transforms in the one-dimensional case is very wide. The aspect of singularity detection through wavelets has been discussed in detail by Mallat (2001). Applicability of various wavelets in detection of cracks in beams has been studied by Poudel et al. (2007), Sun & Chang (2004) and Han et al. (2005). Frame structures have been analyzed by Ovanesova & Suarez (2004). The damage detection in plate structures were addressed by Rucka & Wilde (2006). They proposed a method for estimating damage localization in a beam and plate using continuous wavelet transform. Bayissa & Haritos (2007) proposed a new damage identification technique based on the statistical moments of the energy density function of the vibration responses in the time–scale domain. Also, Kim et al. (2006) and Chang & Chen (2004) studied on damage detection of plate structure. In this paper, new method for damage detection in plate structure based on discrete wavelet transform proposed. The proposed method is validated with a four fixed supported rectangular plate containing damage. 2

OVERVIEW ON ONE-DIMENSIONAL WAVELET TRANSFORM

The continuous wavelet transform is defined as:    +∞ 1 t−b Ca,b = a− 2 f (t)ψ ∗ dt (1) a −∞

where a and b are scale and translation parameters, respectively and ψ ∗ is the complex conjugate of ψ. The basis function ψ is represented as: ψj,k (t) = 2j/2 ψ(2j t − k)

W ( j , m , n ) W H ( j , m, n )

W ( j + 1, m, n)

(2)

W V ( j , m, n ) W D ( j , m , n )

Equation 1 can be represented as: ∗ Ca,b = f (t), ψa,b (t)

(3)

Therefore, continuous wavelet transform is a collection of inner products of a signal f (t) and the translated and dilated wavelets ψa,b (t). The main idea of discrete wavelet transform is the same as that of continuous wavelet transform. In discrete wavelet transform the signals can be represented by approximations and details. The detail at level j is defined as:  Dj (t) = cDj,k ψj,k (t) (4) k∈Z

where ψ j,k is wavelet functions and cDj,k is wavelet coefficients at level j . The approximation at level jis defined as Aj (t) =

+∞ 

cAj (k) φj,k (t)

(5)

k=−∞

where φj,k is scaling functions and cAj,k is scaling coefficients at level j. Finally, the signal f (t) can be represented by f (t) = AJ +



Dj

(6)

TWO-DIMENSIONAL WAVELET TRANSFORM

The one-dimensional wavelet transform can be extended to any dimensions. In this section, the twodimensional case for plate structures damage localization is studied. In two dimensions, a two-dimensional scaling function, φ(x, y), and three two-dimensional wavelet ψ H (x, y), ψ V (x, y) and ψ D (x, y), are required. Each is the product of a one-dimensional scaling function φ and corresponding wavelet ψ (Daubechies 1992). φ(x, y) = φ(x)φ( y)

(7)

ψ (x, y) = ψ(x)φ( y)

(8)

ψ V (x, y) = φ( y)ψ(x)

(9)

ψ D (x, y) = ψ(x)ψ( y)

(10)

H

where ψ H measures variations along columns (like horizontal edges), ψ V responds to variations along rows (like vertical edges), and ψ D corresponds to variations along diagonals. Like the one-dimensional discrete wavelet transform, the two-dimensional discrete wavelet transform can be implemented using digital filters and downsamplers. With separable two-dimensional scaling and wavelet functions, we simply take the one-dimensional fast wavelet transform of the rows of f (x, y), followed by the one-dimensional fast wavelet transform of the resulting columns. As in the two-dimensional case, image f (x, y) is used as the first scale input, and output four quartersize subimages Wφ , WψH , WψV , and WψD . These subimages are shown in the right of Figure 1. 4

j<J

3

Figure 1. One-scale of tow-dimensional decomposition in wavelet transform.

THE DAMAGE IDENTIFICATION METHOD

This paper suggests the use of the discrate wavelet transform for detection of damage in plate. This procedure uses the decomposing of mode shape of the plate using wavelet multiresolution analysis. The mode shapes of the rectangular plate with damage are derived in by finite element method. The damage is represented as the elements with reduced thickness and then the mode shapes are analyzed by the wavelet transform. The mode shapes of the rectangular plate with damage and without damage are derived using finite element method. Geometry of a four fixed rectangular plate with damage showed in Figure 2. The dimensions of the plate are L×B×t. The damage is represented as the elements with reduced thickness. The dimensions of the damaged region are W × D and location coordinates of the damage is L1, B1. The damage region is represented as the element with reduced thickness t. The equation of free vibration of a plate is as fallows: M x¨ + C x˙ + K x = 0

(11)

where M , C and K are mass, damping and stiffness matrices, respectively. Solving Equation 11, one can

416

Figure 2.

Geometry of plate with damage.

get mode shape i , i = 1; 2; . . . ; n, where n is the number of structural mode. In this method, the fundamental mode shape of the damaged plate are decomposed into three twodimensional wavelet ψDH , ψDV and ψDD then the mode shape of the undamaged plate are decomposed into H V D , ψUD and ψUD . three two-dimensional wavelet ψUD In this study, the ratio of wavelet coefficients of the damaged structure to the undamaged structure has been used for damage detection in plate structures. IDHR index is introduced as the ratio of the horizontal wavelet coefficients (ψDH ) of the damaged plate H ) of the to the horizontal wavelet coefficients (ψUD undamaged plate: IDHR =

ψDH H ψUD

elements with reduced thickness. The dimensions of the damaged region are 0.25 m × 0.02 m and location coordinates of the damage is 1 m, 1.75 m. The damage region is represented as the element with reduced thickness 2 cm.The material properties for the plate include Young’s modulus of 2 × 105 kg.f /cm2 , mass density of 2500 kg/m3 and Poisson’s ratio of 0.2. The finite element model of the plate consists of 16 × 16 elements meshed and the geometry of plate and location of damage showed in Figure 3. Damage is assumed to alter only the section properties. It is simulated by reducing the thickness at pre-determined element locations. Then analysis of finite element mode of plate determines the mode shapes of plate and this proposed method for damage detection use fundamental mode shape of the plate. A comparison between the undamaged plate and the damaged plate fundamental mode shapes are shown in Figures 3 and 4. Then, using discrete wavelet transform the mode shape of the damaged plate are decomposed into three two-dimensional wavelets ψDH , ψDV and ψDD . Also,

(12)

V With corresponding definitions for ψDV and ψUD where superscript V stands for vertical direction wavelet coefficients, the index of IDVR is defined:

ψV IDVR = VD ψUD

(13)

Figure 3. Location of damage in the four fixed supported rectangular plate.

Also, for the index IDDR, superscript D highlights directional wavelet coefficients: IDDR =

5

ψDD D ψUD

(14)

NUMERICAL ANALYSIS STUDY

To illustrate the effectiveness of the proposed method, a fixed supported rectangular concrete plate is considered for damage identification of single simulated damage condition. The dimensions of the plate are 4 m × 4 m × 0.2 m. The damage is represented as the

417

Figure 4.

Mode shape for the plate without damage.

Figure 5.

Mode shape for the plate with damage.

Figure 6. 3D view.

Damage identification in the plate using IDVR:

Figure 7. top view.

Damage identification in the plate using IDVR:

the mode shape of the undamaged plate are decomH V , ψUD posed into three two-dimensional wavelets ψUD D and ψUD . The index of IDVR computed and the result showed in Figures 6 and 7. Using the index of IDVR damage

Figure 8. 3D view.

Damage identification in the plate using IDHR:

Figure 9. top view.

Damage identification in the plate using IDHR:

Figure 10. 3D view.

Damage identification in the plate using IDDR:

is detected in plate structure. Figures 6 and 7 show coincidence of the damage location from the proposed method with the damage location in structure. Also, the results of the indexes of IDHR and IDDR showed in Figures 8 to 11.

418

REFERENCES

Figure 11. top view.

Damage identification in the plate using IDDR:

Finally, it can be concluded that this damage identification method is much more sensitive to damage. This is due to the use of two-dimensional discrete wavelet transform in the method for processing of data. 6

CONCLUSIONS

In this study, a method using wavelet multiresolution analysis is developed to damage detection in plate structures. The technique is based on using the twodimensional discrete wavelet transform for detection of liner damage method for damage detection in plate structures. Finally, the numerical results of the damage location are compared with the actual results to demonstrate the effectiveness of the method. The results obtained from the numerical studies indicate that the proposed method is strong and damage sensitive.

419

Bayissa, W.L. & Haritos, N. 2007. Structural damage identification in plates using spectral strain energy analysis. Journal of Sound and Vibration; 307: 226–249. Chang, C.C. & Chen, L.W. 2004. Damage detection of a rectangular plate by spatial wavelet based approach. Applied Acoustics; 65: 819–832. Daubechies, I. 1992. Ten lectures on wavelets. CBMS-NSF Conference Series in Applied Mathematics, Montpelier, Vermon. Doebling, S.W., Farrar, C.R., Prime, M.B. & Shevitz, D.W. 1996. Damage identification and health monitoring of structural and mechanical systems from change in their vibration characteristics: a literature review. Las Alamos National Laboratory Report LA-13070-MS, Las Alamos, New Mexico. Han, J., Ren, W. & Sun, Z. 2005. Wavelet packet based damage identification of beam structures. International Journal of Solids and Structures; 42: 6610–6627. Kim, B.H., Kim, H. & Park, T. 2006. Nondestructive damage evaluation of plates using the multi-resolution analysis of two-dimensional Haar wavelet. Journal of Sound and Vibration; 292: 82–104. Mallat, S. 2001. A wavelet tour on signal processing. NewYork, Academic Press. Ovanesova, A.V. & Suarez, L.E. 2004. Application of wavelet transform to damage detection in frame structures, Engineering Structures; 26: 39–49. Poudel, U.P., Fu, G. & Ye, J. 2007. Wavelet transformation of mode shape difference function for structural damage location identification. Earthquake Engineering & Structural Dynamics; 36: 1089–1107. Rucka, M. & Wilde, K. 2006. Application of continuous wavelet transform in vibration based damage detection method for beams and plates. Journal of Sound and Vibration; 297: 536–550. Sun, Z. & Chang, C.C. 2004. Statistical Wavelet-Based Method for Structural Health Monitoring. Journal of Structural Engineering, ASCE; 130(7): 1055–1062.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

The non-local theories based on different types of weighted functions and its application Z. Wang & Q. Yang Institute of Geotechnical Engineering, School of Civil and Hydraulic Engineering, Dalian University of Technology, Dalian, China

ABSTRACT: The effects of the stress intensity factor KI and KII on the all components of strains at the crack tip are analyzed by the non-local theories based on different types of weighted functions. The non-local strain will be considerably reduced. The larger the internal length scale is, the more significant the reduction of non-local strain with respect to the local strain predicted conventionally will be. The size of non-local strain field with the bell-shaped weighted functions is larger than that obtained by either Green’s or Gaussian weighted functions. The non-local normal stress-strain components depends on the stress intensity factor KI and KII , the circumferential stress is related to the stress intensity factor KI . As the distance of the point and crack tip increases, the non-local strain gets to decrease strongly, then it gets to increase and arrives to the biggest value, Finally, The non-local normal stress-strain gets to decrease. 1

INSTRUCTION

Classical mechanics are founded on the basic assumption that physical state of the material at a given point is determined only by the state lying within an arbitrary small neighborhood of that point. This hypothesis is valid only when the characteristic length scale of the object is much larger than the intrinsic length scale of media. Since micro-cracks inside rock-like materials exist, it can no longer be regarded as homogeneous. The process of degradation produce an unrealistic phenomenon of singularity at crack tip, cannot reproduce gradual degradation of stiffness and strength. The conventional assumptions on stress and strain states are not acceptable. In order to solve the problem and to reproduce interaction behavior of microstructures, the concept of non-locality has been incorporated with the theories of damage and fracture mechanics (Eringen et al. 1977; Pijaudier-Cabot & Bazant 1987; Aifant 1984). Based on the non-local fracture theory, the stress and strain state at a given point in a body relies on not only the deformation state at the point but also that of the neighborhood of the point. The size of the influence domain is related to the degree of discontinuity of the non-homogeneous material. Alternatively, the scale of this domain is associated with the intrinsic length scale of it. In order to reproduce the non-homogeneous feature, the parameter of the intrinsic length scale is introduced to remedy the defect of classical theory. To introduce the concept of non-local fracture to the constitutive model, the local values of the related physical variables should be replaced by their weighted average values of certain form in space.

421

The spatial average value of some physical variables in the form of weighted integration is defined by Bazant (1988) and it is stated that the physical variables related to non-linear strain softening behavior in constitutive p relation, such as incremental plastic strain εij , plastic multiplier λ or others should take their average values. The elastic variables which do not participate in the localization are still used in the local continuous models. The stress singularity at the crack of I-, II-, III-mode were discussed by Eringen (1977) using non-local theory within a wide range of frequency and wavelength. The effect of the lattice parameter of anisotropic materials on the stress field of crack tip under a uniform anti-plane shear loading is investigated by Sun et al. (2004) by means of non-local theory, and dependence of stress field at crack tip on crack length and crystal lattice dimension by means of Fourier transformation are discussed. The stress and strain fields near the crack tip of I mode crack are analyzed by Peerlings (2001) with use of non-local theory based on Gaussian weighted function and the results are compared with those obtained by strain-gradient model. Distributions of a certain physical variables are determined by weighted functions introduced in nonlocal fracture theory. Therefore, the choice of weighted function is a key to constitute to non-local strain softening model. In this paper, the relationship of different weighted functions and intrinsic length scale and their influence on non-local theory are discussed. Different distributions of stress field near the crack of the I-II mixed mode crack are compared by means of non-local model with different types of weighted functions.

COMPARISON OF WEIGHTED FUNCTIONS IN NON-LOCAL THEORY

In non-local model, the non-local equivalent strain relies on not only the local strain at the point but also that of the neighborhood of the point. The non-local equivalent variable is defined as a weighted average of the local equivalent variable x in the entire domain V (Pijaudier-Cabot 1987). The types of weighted functions substantially affect the spatial distribution of the corresponding variables. Thus it is meaningful to discuss the basic characteristics of different types of weighted functions and their effect on the nonlocal theory. In practice, Gaussian function, Green’s function and bell-shaped function as following are commonly used (Aifant 1984).   1 ρ2 ψ (x, ξ ) = (1a) exp − 3 2l 2 (2π) 2 l 3  ρ 1 exp − ψ (x, ξ ) = (1b) 2 4πρl l ⎧  2 ⎪ ρ2 ⎨ c 1 − 2 , ρ ≤ R; (1c) ψ (x, ξ ) = R ⎪ ⎩ 0, ρ≥R

 3 2 where ψ (x, x) = 1, ρ = |x − ξ | = i [xi − ξi ] and l is intrinsic length scale of the material.

( )/mm-3

0.8

-4

-2

l = 0.5 0

2

4 /l

0.6 l = 0.5 0.4 l =1 0.2 l =2 0 -6

-4

-2

0

2

4 /l 6

Figure 4. Distribution of Bell-shaped weighted function for different values of the internal length scale. 8 ( )/mm-1

Green’s function

6

Bell-shaped function

4 2

( )/mm-3

l=2 l=1

Figure 3. Distribution of Green’s weighted function for different values of the internal length scale.

0.4 0.2

2

Bell-shaped function 0

-2

0 /mm

2

0 4

Figure 1. Comparison among three types of weighted functions. 0.3 0.2 0.1

l = 0.5 l =1

0

l =2

-

-4

-2

0

2

Gaussian s function

Green s function

Gaussian’s function -4

( )/mm-3

0.4

0

0.8 0.6

0.6

0.2

( )/mm-3

2

4 /l

Figure 2. Distribution of Gaussian weighted function for different values of the internal length scale.

0

1

2

/mm

3

4

5

Figure 5. Distribution of the common multiplier 4πρ 2 for different types of weighted functions for internal length scale of l = 1(mm).

The weighted function contains at least one parameter related to scale of length such as l in Equations 1a, 1b and R in Equation 1c. This parameter incorporates in the simplest possible way some information about the microstructure and frequently controls the size of the localized plastic zone. The value of the parameter is related to the intrinsic length scale, which is dictated by the size and spacing of the dominant heterogeneities. Distribution of the three types of weighted functions under different intrinsic length scales is shown in Figures 1–4. The extremums and descending rates

422

of the Gaussian and bell-shaped weighted functions increase with the decrease of intrinsic length scale. The influence domain of the Gaussian weighted function tends to infinity while the influence of bellshaped weighted function is a finite value decreasing with decrease of intrinsic length scale. For Green’s weighted function with different intrinsic length scale, the extremum tends to infinity, and the decending rates decreases with decrease of intrinsic length scale. The influence domain tends to infinity. Distribution of the common multiplier 4πρ 2 for different types of weighted functions is shown in Fig. 5. Under the same intrinsic length scale condition, the size of the extremum of the common multiplier 4πρ 2 for the bell-shaped weighted function is much larger than that of the green’s function and that of the Gauss weighted function.

of different types of weighted functions on the strain singularity near crack tip, the I-II mixed mode crack shown in Figure 6 is analyzed as an example. The strain distribution near crack tip is discussed by non-local model with different types of weighted functions. In the polar coordinates, the origin of coordinate systems coincides with the crack tip. The stress expression at crack tip given by local elastic fracture mechanics can be written as

(3 − cos θ) cos θ2 KI + (3 cos θ − 1) sin θ2 KII σr = √ 2 2π r (2a)   cos θ2 cos2 θ2 KI − 32 sin θKII σθ = (2b) √ 2 2π r σrθ =

3

EFFECT OF WEIGHTED FUNCTION ON STRAIN AT CRACK TIP

cos θ2 [(1 + cos θ ) KI − (3 cos θ − 1) KII ] √ 2 2π r (2c)

The strain singularity near crack tip is an acceptable fact in classical theory. To illuminate the influence

The local equivalent strain at crack tip are defined as    θ 1 θ εr = √ cos 3 − cos θ − ν cos2 KI 2 2 2E 2π r    θ 2 θ cos θ − 1 + ν cos KII + 3 sin 2 2 εθ =

H

o a W

Figure 6. Table 1.

(3a)

   θ 1 θ cos cos2 − 3ν + ν cos θ KI √ 2 2 2E 2π r    θ θ cos2 − 3ν cos θ + 3ν KII + 3 sin 2 2 (3b)

  θ θ (1 + ν) γrθ = √ cos 2 cos2 KI − KII (3 cos θ − 1) 2 2 E 2π r (3c)

Model of calculation.

The expression of non-local strain for different types of weighted functions. Type of weighted function

Non-local variable

Gaussian’s function

ε¯ r ε¯ θ γ¯rθ



 ρ2 exp − 2 3 2l (2π) 2 l 3 3   8 

/4 3 KI − 3.7KII √ 3 3 24 π 2 E l 3 

/4 (−0.53KI + 8.8KII ) 3 3 √ 4 × 24π 2 E l 3   16  0.2 × /4 3 KI 3 3 √ 24 π 2 E l 1

Green’s function  ρ 1 exp − 2 4πρl l    

2 3/4 83 KI − 3.7KII √ 2π E 2l  

2 3/4 (−0.53KI + 8.8KII ) √ 8π E 2l     0.1 2 3/4 16 KI √ 3 πE 2l

423

Bell-shaped function ⎧ ⎨

2  ρ2 c 1 − 2 , ρ ≤ R; R ⎩ 0, ρ≥R   K − 14.8K 64 32 II 3 I 3

1

1

77π 2 El 2 6 6 64 (−0.533KI + 8.8KII ) 3

1

1

77π 2 El 2 6 6   51.2 16 3 KI 3

1

1

77π 2 El 2 6 6

The local and non-local equivalent strain with different angles is shown in Figures 8–10.

The non-local stress-strain field for different types of weighted functions is given in Table 1. The following founding can be obtained from a comparison of the results as stated above. 1. When interaction of microstructure is considered, the value of stress at crack tip is finite. It is implied that the singularity at crack tip can be smoothed. When microstructure interaction is not taken account, the results are identical to those obtained by the macro-fracture theory. The larger the intrinsic length scale is, the smaller the non-local strain value at crack tip is, and the larger the weaken tendency of strain field is. 2. The non-local normal strain at crack tip is related to stress intensity factors KI and KII . The non-local shear strain is related to stress intensity factor KI and independent of KII . 3. The size of the non-local strain obtained by different types of weighted functions is different. The size of the non-local strain obtained by the bell-shaped weighted functions is larger than that obtained by the Green weighted function and larger than that obtained by the Gauss weighted function. As the stress intensity factor KI increases, non-local normal strain and shear strain increases, the circumferential strain decreases. As the stress intensity factor KII increases, the non-local normal strain decreases and the circumferential strain increases, the shear strain is independent of the stress intensity factor KII .

4

1. The local equivalent strain of crack tip inclines to ∞. The larger the distance of the corresponding point with crack tip is, the smaller the local equivalent strain is, the smaller the weaken tendency of non-local stress field is. The Local equivalent strain at crack tip for different orientation is different. The larger The Local equivalent strain near crack tip is, the larger the angle of traction-straight line with the face of the crack is. The value of the strain for crack tip is finite owing to the introduction of internal length scale. 2. As the distance of the point and crack tip increases, The non-local strain gets to decrease strongly, then it gets to increase and arrives to the biggest value, Finally, The non-local normal stress-strain gets to decrease. 3. The size of the non-local strain obtained by the bell-shaped weighted functions is larger than those obtained by the Green’s and Gauss weighted functions. These are smaller than the size of local strain of the point. The larger the intrinsic length scale is,

APPLICATION OF WEIGHTED FUNCTION IN NON-LOCAL DAMAGE MODEL

To illuminate the influence of different types of weighted functions on the strain singularity near crack tip, the I-II mixed mode crack shown in Figure 6 is analyzed as an example. The specimen with a crack length of 25 cm consists of a rectangular solid 50 cm long and 60 cm tall, deforming in plane strain as depicted in Figure 6. The material is assumed to obey the VonMises yield criterion with associative flow rule, with Young’s modulus E = 0.3 GPa, Poisson’s ratio ν = 0.3 under an equal stress of σ = 0.1 MPa. The isoparametric finite element meshes are constructed for this problem. The singular finite element meshes are constructed for the crack tip. To consider the effect of internal length scale on strain at the crack tip, the local strain ε and the non-local strain ε¯ derived by different types of weighted function for the corresponding point is analyzed by deriving the straight line with angle of 0◦ , 15◦ , 30◦ , 45◦ , 60◦ , 90◦ and x-axis from the configuration center of crack element. The distribution of the local equivalent strain is shown in Figure 7. The singularity of the stress at the crack tip cannot be smoothed under the loaded condition.

Figure 7. crack tip.

Distribution of local equivalent strain near

eq 1.0 0.0 8 0.0 6 9

0.0 4 0.0 2 0

0

0. 5

1. 0 x/mm

1. 5

2.0

Figure 8. Local equivalent strain at crack tip for different orientation.

424

eq

0.06

0.08 0.06

l = 1 mm

0.04

eq

0.04 0.02

0.02 0

0

0.5

1.0 x/mm

1.5

Bell Green

0

2.0

0

30

angle/0

Gauss 90

60

Figure 12. Comparison of local equivalent strain and non-local equivalent strain for different types of weighted functions at crack tip for different orientation.

Figure 9. Non-local equivalent strain for bell-shaped weighted function at crack tip for different orientation. eq

0.03 0.06

0.02

Bell-shaped function

0.04

eq

0.01 9

0

0

0.5

0.02

1.0 1.5 x/mm

eq eq

l = 1 mm l = 2 mm

2.0

0 0

30

60

90

angle/0

Figure 10. Non-local equivalent strain for Green weighted function at crack tip for different orientation.

Figure 13. Comparison of local equivalent strain and nonlocal equivalent strain for bell-shaped weighted function at crack tip for different orientation.

eq 0.02

Table 2. Non-local strain for different types of weighted functions at crack tip.

0.01

Non-local strain Types of weighted Intrinsic function length scale ε¯ r ε¯ θ

0 0

0.5

1.0 x/mm

1.5

2.0

Figure 11. Non-local equivalent strain for Gauss weighted function at crack tip for different orientation.

Bell-shaped function Green’s function Gaussian’s function

the larger the weaken tendency of non-local strain field is. The non-local strain field for different types of weighted functions and different intrinsic length scale is given in Table 2. The size of the non-local normal strain obtained by the same weighted function is larger than the size of the non-local circumferential strain and the size of the non-local shear strain. The larger the intrinsic length scale is, the larger the weaken tendency of non-local strain field is.

425

5

l l l l l l

= 1 mm = 2 mm = 1 mm = 2 mm = 1 mm = 2 mm

0.101 0.072 0.039 0.027 0.030 0.021

−0.0050 −0.0036 −0.0019 −0.0012 −0.0015 −0.0011

γ¯rθ 0.0405 0.0286 0.0126 0.0089 0.0120 0.0085

CONCLUDING REMARKS

Distribution of strain field of I-II mixed mode crack is analyzed by non-local theory with different types of weighted functions. The effects of the stress intensity factor KI and KII on the all components of strains at the crack tip are analyzed. The following conclusions can be obtained. 1. The non-local theory is instructive to avoid the trouble resulting from stress-strain singularity at crack tip. As the distance of the point and crack

tip increases, the non-local strain gets to decrease strongly, then it gets to increase and arrives to the biggest value, Finally, the non-local normal stressstrain gets to decrease. The size of the non-local strain obtained by the bell-shaped weighted functions is larger than those obtained by the Green’s and Gauss weighted functions. These are smaller than the size of local strain of the point. The larger the intrinsic length scale is, the larger the weaken tendency of non-local strain field is. 2. The non-local radial strain at crack tip is related to stress intensity factors KI and KII . The influence of stress intensity factor KI on strain is positive while the influence of KII is negative. The nonlocal circumferential strain at crack tip is related to stress intensity factors KI and KII . The influence of stress intensity factor KI on strain is negative while the influence of KII is positive. The non-local shear strain is related to stress intensity factor KI and independent of KII . The influence of stress intensity factor KI on strain is positive. Gradual degradation of the parameter of the material is obtained at plastic zone near crack tip on account of the intrinsic length scale. These can simulate the strain-softening phenomena of materials used in the analysis of strain localization.

REFERENCES Aifant E.C. 1984. On the microstructural origin of certain inelastic models[J]. Journal of Engineering Materials and Technology[J]. 106(4): 326–330. Bazant Z.P. and Lin F.B. 1988. Non-local yield limit degradation[J]. International Journal for Numerical Methods in Engineering, 26(8): 1805–1823. Eringen A.C., Speziale C.G., Kim B.S. 1977. Crack tip problem in non-local elasticity[J]. Journal of Mechanics and Physics of solids, 25(5): 339–351. Eringen A.C., Edelen, D.G.B. 1972. On nonlocal elasticity[J]. International Journal of Engineering Science, 10(3): 233–248. Peerlings R.H.J., Geers M.G.D., de Borst R., Brekelmans W.A.M. 2001. A critical comparison of non-local and gradient-enhanced softening continua[J]. International Journal of Solids and Structures, 38(44/45): 7723–7746. Pijaudier-Cabot G., Bazant Z.P. 1987. Nonlocal damage theory[J]. Journal of Engineering Mechanics, ASCE, 113(10): 1512–1533. Yu-Guo Sun, Zhen-Gong Zhou. 2004. Stress field near the crack tip in nonlocal anisotropic elasticity[J]. European Journal of Mechanics A/Solids, 23(2): 259–269.

ACKNOWLEDGMENTS This work was funded by the National Natural Science Foundation of China (No. 50679015, 50574016) and Education Foundation of Liaoning (No. 2008Z030). This support is gratefully acknowledged.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

The MINLP approach to structural synthesis S. Kravanja & T. Žula University of Maribor, Maribor, Slovenia

ABSTRACT: The paper presents the Mixed-Integer Non-linear Programming optimization approach (MINLP) to structural synthesis. The MINLP is a combined discrete/continuous optimization technique, where discrete binary 0–1 variables are defined for optimization of discrete alternatives and continuous variables for optimization of parameters. The MINLP optimization to structural synthesis is performed through three steps: i.e. the generation of a mechanical superstructure, the modeling of an MINLP model formulation and the solution of the defined MINLP problem. As the discrete/continuous optimization problems are usually non-convex and highly non-linear, the Modified Outer-Approximation/Equality-Relaxation (OA/ER) algorithm is selected to be used for the optimization. The accompanied Linked Multilevel Hierarchical Strategy (LMHS) is developed to accelerate the convergence of the mentioned algorithm. Three numerical examples are presented at the end of the paper. 1

INTRODUCTION

The paper presents the Mixed-Integer Non-linear Programming optimization approach (MINLP) to structural synthesis. The optimization is performed by the Mixed-Integer Non-linear Programming approach (MINLP). The MINLP is a combined continuous and discrete optimization technique. It handles with continuous and discrete binary 0–1 variables simultaneously. While the continuous variables are defined for the continuous optimization of parameters (dimensions, stresses, strains, weights, costs, etc.), the discrete variables are used to express discrete decisions, i.e. usually the existence or non-existence of structural elements inside the defined structure (topology). Different standard sizes, materials and rounded dimensions may also be defined as discrete alternatives. Since the MINLP performs both the continuous and discrete optimizations simultaneously, the MINLP approach also finds the optimal continuous parameters (mass, costs, stresses, etc.), the structural topology (with an optimal number and a configuration of structural elements), standard sizes, materials and rounded dimensions simultaneously. Since some structural elements are added or removed from the structure within the optimization process, so that they on the basis of a defined algorithm form various structural alternatives automatically, the paper concentrates on the discussion of structural synthesis. The MINLP discrete/continuous optimization problems of structural synthesis are in general difficult to be solved. The structural synthesis is thus proposed to be performed through three steps: the first one includes the generation of a mechanical superstructure of different topology, material, standard and rounded

427

dimension alternatives, the second one involves the development of an MINLP model formulation and the last one consists of a solution for the defined MINLP optimization problem. The MINLP optimization problems of structures are in most cases comprehensive, non-convex and highly non-linear. This paper reports the experience in solving this type of problems by using the Modified Outer-Approximation/ Equality-Relaxation algorithm by Kravanja & Grossmann (1994), see also Kravanja et al. (1998a). The Linked Multilevel Hierarchical Strategy (LMHS) has been developed to accelerate the convergence of the mentioned algorithm. Since the number of discrete alternatives and defined binary 0–1 variables are usually too high for normal solution of the MINLP, a special prescreening procedure is developed to automatically reduce the number of binary variables on a reasonable level. Three examples are presented at the end of the paper. The first one shows the material and standard dimension optimization of a composite floor system, the second one introduces the standard section optimization of a three-story steel frame and the third one presents the simultaneous topology and standard section optimization of a single-story industrial steel building.

2

MINLP SUPERSTRUCTURE

The MINLP optimization approach to structural synthesis requires the generation of an MINLP superstructure composed of various topology and design alternatives that are all candidates for a feasible and optimal solution. While topology alternatives represent

f (x). Non-linear equality and inequality constraints h(x) = 0, g(x) ≤ 0 and the bounds of the continuous variables represent the rigorous system of the design, load, resistance, stress, deflection, dimensioning, etc. constraints known from structural analysis. Logical constraints that must be fulfilled for discrete decisions and structure configurations are given by By + Cx ≤ b. These constraints describe relations between binary variables and define the structure’s topology, materials, standard and rounded dimensions. It should be noted, that the comprehensive MINLP model formulation for engineering structures may be found elsewhere, see Kravanja et al. (1998a,b,c, 2005).

different selections and interconnections of corresponding structural elements, design alternatives include different materials, standard and rounded dimensions. The superstructure is typically described by means of unit representation: i.e. structural elements and their interconnection nodes. Each potential topology alternative is represented by a special number and a configuration of selected structural elements and their interconnections; each structural element may in addition have different material, standard and rounded dimension alternatives. Therefore, the main goal is to find within the given superstructure a feasible structure that is optimal with respect to topology, material, standard and rounded dimensions as well as to all defined continuous parameters.

3

4

After the MINLP model formulation is developed, the defined MINLP optimization problem is solved by the use of a suitable MINLP algorithm and strategies. A general MINLP class of optimization problems can be solved in principle by the following algorithms and their extensions: the Generalized Benders Decomposition, presented by Benders (1962) and Geoffrion (1972); the Non-linear Branch and Bound, presented by Gupta & Ravindran (1985); the Outer-Approximation algorithm, presented by Duran & Grossmann (1986); the Feasibility Technique by Mawengkang & Murtagh (1986); the Sequential Linear Discrete Programming, presented by Olsen & Vanderplaats (1989), Bremicker et al. (1990); the LP/NLP based Branch and Bound, presented by Quesada & Grossmann (1992); and the Extended Cutting Plane by Westerlund & Pettersson (1995).

MINLP MODEL FORMULATION

A general non-linear and non-convex continuous and discrete optimization problem can be formulated as an MINLP problem in the following form, see Equations 1 to 6: min z = cT y + f (x)

(1)

s.t. : h(x) = 0

(2)

g(x) ≤ 0

(3)

By + Cx ≤ b

(4)

x ∈ X = {x  Rn : x lo ≤ x ≤ x up }

(5)

y ∈ Y = {0, 1}

(6)

m

SOLVING THE MINLP PROBLEM

4.1

where x is a vector of continuous variables specified in the compact set X and y is a vector of discrete, binary 0–1 variables. Functions f (x), h(x) and g(x) in Equations 1, 2 and 3 are non-linear functions involved in the objective function z, equality and inequality constraints, respectively. All functions f (x), h(x) and g(x) must be continuous and differentiable. Finally, By + Cx ≤ b represents a subset of mixed linear equality/inequality constraints. The above general MINLP model formulation has been adapted for the optimization of structures. In the context of structural optimization, continuous variables x define structural parameters (dimensions, strains, stresses, costs, weight...) and binary variables y represent the potential existence of structural elements as well as the choice of materials, standard and rounded dimensions. The economical (mass) objective function z involves fixed costs charges in the term cT y for manufacturing, while the dimension dependant costs are included in the function

Modified OA/ER algorithm

Since the MINLP optimization problems of structures are in most cases comprehensive, non-convex and highly non-linear, the Outer-Approximation/ EqualityRelaxation (OA/ER) algorithm by Kocis & Grossmann (1987) is selected for the optimization. The OA/ER algorithm consists of solving an alternative sequence of Non-linear Programming (NLP) optimization subproblems and Mixed-Integer Linear Programming (MILP) master problems, see Figure 1. The former corresponds to continuous optimization of parameters for a structure with a fixed topology (and fixed standard dimensions, discrete materials) and yields an upper bound to the objective to be minimized. The latter involves a global approximation to the superstructure of alternatives in which a new topology, standard dimensions, discrete materials are identified so that its lower bound does not exceed the current best upper bound. The search of a convex problem is terminated when the predicted lower bound exceeds the upper bound, otherwise it is terminated when the NLP solution can be improved no more. The

428

decompose the original integer space and original MINLP problem in a hierarchical manner into several subspaces and corresponding MINLP levels. Each time the next MINLP optimization level is performed, the current integer subspace is extended by the next integer subspace and prescreened, while the discrete decisions belonging to all of the remaining subspaces are approximated by the relaxed 0–1 variables. The levels are linked by accumulating outer-approximations and yield lower bounds to their next level objective functions to be minimized, which considerably improve the efficiency of the search. Decision levels are hierarchically classified as:

MINLP Superstructure Combined optimization Fixed binary variables NLP Subproblem Continuous optimization New binary variables

– the level of discrete topology and material alternatives (the highest level), – the level of discrete standard dimension decisions (the middle level), – the level of rounded dimension decisions (the lower level).

MILP Master problem Discrete optimization Convergence ? NO YES STOP Figure 1.

Steps of the OA/ER algorithm.

OA/ER algorithm guarantees the global optimality of solutions for convex and quasi-convex optimization problems. The OA/ER algorithm, as well as all other mentioned MINLP algorithms, does not generally guarantee that the solution found is the global optimum. This is due to the presence of non-convex functions in the models that may cut off the global optimum. In order to reduce undesirable effects of nonconvexities, the Modified OA/ER algorithm was proposed by Kravanja & Grossmann (1994), see also Kravanja et al. (1998a), by which the following modifications are applied for the master problem: the deactivation of linearizations, the decomposition and the deactivation of the objective function linearization, the use of the penalty function, the use of the upper bound on the objective function to be minimized as well as the global convexity test and the validation of the outerapproximations. 4.2

Linked multilevel hierarchical strategy

Higher levels give lower bounds to the original objective function to be minimized while lower levels give upper bounds. The MINLP subproblems are iterated about each level until there are no improvements in the NLP solution. Thus, we start with the discrete topology and material optimization at the relaxed standard dimensions. When the optimal topology and materials are reached, we proceed with simultaneous discrete topology, material and standard dimension optimization at the second level. Finally, after the optimal topology, materials and standard dimensions are obtained, the MINLP is carried out once more for complete discrete decisions at the third level. The optimization model may contain up to some ten thousands of binary 0–1 variables of alternatives. Most of them are subjected to rounded dimensions. Since this number of 0–1 variables is too high for a normal solution of the MINLP, we developed a prescreening procedure, which automatically reduces binary variables for rounded dimension alternatives into a reasonable number. In the optimization at the third level are included only those 0–1 variables which determine rounded dimension alternatives close to continuous dimensions, obtained at previous MINLP optimization iterations. It should be noted, that the LMHS strategy can solve convex problems to global optimal solutions. More about multilevel MINLP strategies also see Kravanja et al. (2003, 2005). 5

The optimal solution of comprehensive MINLP problem with a high number of discrete decisions is in general very difficult to be obtained. For this purpose, the Linked Multilevel Hierarchical Strategy (LMHS) has been developed to accelerate theconvergence of the OA/ER algorithm. Using the LMHS strategy, we

429

NUMERICAL EXAMPLES

The MINLP optimization approach is illustrated by three numerical examples. The first one shows the material and standard dimension optimization of a composite I beam floor system, the second one introduces the standard section optimization of a three-story

As = 23.57 cm2/m1

10 2020

10

8 183

C 25/30

100

8

S 355

100 e = 4967 mm

Optimal topology of the single-story industrial building.

5.1

Material and standard dimension optimization of a composite I beam floor system

The first example presents the simultaneous material and standard dimension optimization of a 27 m long composite I beam floor system, subjected to the selfweight and to the uniformly distributed imposed load of 7.5 kN/m2 . The MINLP optimization model COMBOPT was developed for the optimization of composite I beams. The material and labor costs for composite beams were accounted for in the economical type of the objective function, subjected to the given design, material, stress, deflection and stability constraints, defined in accordance with Eurocodes 4 (1992). The superstructure comprised 6 different concrete strengths (C25, C30, C35, C40, C45, C50), 3 different structural steel grades (S 235, S 275, S 355), 48 various standard

5.0 m

q = 50.0 kN/m

q = 50.0 kN/m

5.0 m

and three bay steel frame and the third one presents the simultaneous topology and standard section optimization of a single-story industrial steel building. While the task of the first example was to find the minimal manufacturing costs of the structure, the mass optimization was carried out for both steel structures. The optimization of the structures was carried out by a user-friendly version of the MINLP computer package MIPSYN, the successor of PROSYN (Kravanja & Grossmann 1994) and TOP (Kravanja et al. 1992, 1998a). GAMS/ CONOPT by Drudd (1994), the Generalized Reduced-Gradient method, was used to solve NLP subproblems and GAMS/Cplex (2001), Branch and Bound, was used to solve MILP master problems. For each type of structure, a special optimization model was developed. Each model was constructed on the basis of the mentioned general MINLP model formulation. As an interface for mathematical modeling and data inputs/outputs GAMS (General Algebraic Modeling System), a high level language by Brooke et al. (1998), was used.

q = 50.0 kN/m

5.0 m

Figure 2.

5.0 m

Figure 3.

5.0 m

5.0 m

Three-story, three-bay steel frame.

reinforcing steel sections as well as 9 different standard thickness of sheet-iron plates (from 8 mm to 40 mm) for webs and flanges separately. The Modified OA/ER algorithm and the LMHS MINLP strategy were applied. The optimal result of 95.85 EUR/m2 was obtained in the 2nd MINLP iteration, see Figure 2. Beside the optimal self-manufacturing costs, the optimal concrete strength C25/30, steel grade S 355, the reinforcement, intermediate distances between I sections, depth of the slab and optimal standard thickness of webs and flanges also were obtained. 5.2

Standard sizes optimization of a three-story steel frame

The second example shows the standard sizes optimization of a three-story, three-bay steel frame, see Figure 3. The frame was subjected to the self-weight

430

and to the uniformly distributed variable load of 50 kN/m. The task of the optimization was to find the minimal structure mass and the discrete standard steel sections. The MINLP optimization model for steel frames FRAMEOPT (FRAME OPTimization) was developed. The mass objective functions was subjected to structural analysis constraints. The finite element equations were defined as the equality constraints for the calculation of internal forces and deflections of the structure. (In)equality constraints for the dimensioning of steel members were determined in accordance with Eurocode 3 (1992). Both, the ultimate and serviceability limit states were checked. Second-order elastic structural optimization was performed by considering a geometric nonlinearity due to P-δ and P- effects.

The frame superstructure was generated in which all possible structures were embedded by different standard section variation. The superstructure comprised 17 different standard hot rolled European wide flange HEA sections (from HEA 100 to HEA 500) for each beam and column separately. The material used was steel S 355. The Modified OA/ER algorithm and the two-phase MINLP optimization were applied. The minimal structure mass of 4569 kg was obtained in the 27th main MINLP iteration (the subsequent feasible results were not so good). The optimal structure of the frame is shown in Figure 4. Beside the minimal frame structure mass, the optimal standard hot rolled HEA sections of steel members (columns and beams) were also obtained. 5.3

HEA 120

5.0 m 5.0 m

HEA 120

5.0 m

HEA 260 HEA 180

HEA 180

HEA 140

5.0 m

Optimal design of the steel frame.

7.5 m 0.75

Figure 4.

5.0 m

HEA 120

HEA 120

HEA 160 HEA 260

HEA 260

The third example introduces the topology and standard section optimization of a single-story industrial steel building structure. The building is 32 meters wide, 65 meters long and 8.25 meters high. The structure is consisted from equal non-sway steel portal frames, which are mutually connected with purlins and rails. The structure was subjected to the self-weight and to the variable load of snow and wind. The mass of the roof was 0.20 kg/m2 and of the facade cladding 0.15 kg/m2 . The variable imposed loads 1.40 kN/m2 (snow) and 0.40 kN/m2 (horizontal wind) were defined in the model input data. The material used was steel S 355. The task of the optimization was to find the minimal structure mass, the optimal topology and all standard cross-sections. The building superstructure was generated in which all possible structures were embedded by 50 portal frame alternatives, 50 purlin alternatives,

HEA 260 HEA 160

HEA 120

HEA 120 HEA 260

HEA 260

5.0 m

H E A 26 0

HEA 260

HEA 140

H E A 2 60

Topology and standard section optimization of a single-story industrial steel building

2 m 8 .1 8 x m .0 65

14 x 2.29 m 32.0 m

Figure 5.

Optimal design of the steel frame.

431

20 rail alternatives and 24 different alternatives of standard hot rolled European wide flange HEA sections (from HEA 100 to HEA 1000) for each column, beam, purlin and rail separately. The MINLP optimization model HALLOPT was developed for the optimization of the single-story industrial buildings. Internal forces were calculated by the elastic first-order theory for the non-sway frame mode. The dimensioning of steel members was performed in accordance with Eurocode 3 for the conditions of both the ultimate and serviceability limit states. The MINLP optimization of the structure mass was performed by the computer package MIPSYN/TOP (GAMS/CONOPT2 and GAMS/CPLEX 7.0). The Modified OA/ER algorithm and the LMHS strategy were applied. The optimal result of 145.33 tons was obtained in the 11th main MINLP iteration. The optimal solution includes the obtained optimal topology of 9 portal frames, 16 purlins with the same number of secondary facade columns, and 8 rails, see Figure 5. The main frame columns are designed from HEA 700, beams from HEA 800, purlins from HEA 160, rails from HEA 100 and the secondary facade columns from HEA 220 standard sections.

6

CONCLUSIONS

The paper presents the Mixed-Integer Non-linear Programming approach (MINLP) to structural synthesis. The Modified OA/ER algorithm and the LMHS strategy are applied. The optimization is performed by the user-friendly MINLP computer package MIPSYN. Beside the optimal structure mass or costs, the optimal topology with the optimal number of structural elements, the optimal discrete materials and the optimal standard dimensions are obtained simultaneously. The examples, presented at the end of the paper, clearly show the efficiency of the proposed MINLP approach.

REFERENCES Benders, J.F. 1962. Partitioning Procedures for Solving Mixed-variables Programming Problems. Numer. Math. 4: 238–52. Bremicker, M., Papalambros, P.Y., Loh, H.T. 1990. Solution of Mixed-Discrete Structural Optimization Problems with a New Sequental Linearization Method. Comput. Struct. 37(4): 451–61. Brooke, A., Kendrick, D., Meeraus, A., Ramesh R. 1998. Tutorial by Rosenthal R.E. GAMS - A User’s Guide. GAMS Development Corporation, Washington. Drudd, A.S. 1994. CONOPT—A Large-Scale GRG Code, ORSA Journal on Computing 6(2): 207–216. Duran, M.A., Grossmann, I.E. 1986. An OuterApproximation Algorithm for a Class of Mixed-Integer Nonlinear Programs. Math. Prog. 36: 307–39.

Eurocodes 3, 1992. Design of steel structures. European Comitee for Standardization. Eurocodes 4, 1992. Design of composite structures. European Comitee for Standardization. GAMS/CPLEX 7.0, 2001. User Notes. ILOG inc. Geoffrion, A.M. 1972. Generalized Benders Decomposition. J. Optim. Theory Appl. 10(4): 237–60. Gupta, O.K., Ravindran, A. 1985. Branch and bound experiments in convex nonlinear integer programming, Management Science 31(12): 1533–1546. Kocis, G.R., Grossmann, I.E. 1987. Relaxation Strategy for the Structural Optimization of Process Flowsheets, Ind. Engng Chem. Res. 26: 1869–1880. Kravanja, S., Kravanja, Z., Bedenik, B.S., Faith, S. 1992. Simultaneous Topology and Parameter Optimization of Mechanical Structures. In Hirsch, C. et al. (ed.), Proceedings of the First European Conference on Numerical Methods in Engineering, Brussels, Belgium: 487–495. Elsevier, Amsterdam. Kravanja, S., Kravanja, Z., Bedenik, B.S. 1998a. The MINLP optimization approach to structural synthesis. Part I: A general view on simultaneous topology and parameter optimization. International Journal for Numerical Methods in Engineering 43: 263–292. Kravanja, S., Kravanja, Z., Bedenik, B.S. 1998b. The MINLP optimization approach to structural synthesis. Part II: Simultaneous topology, parameter and standard dimension optimization by the use of the Linked two-phase MINLP strategy. International Journal for Numerical Methods in Engineering 43: 293–328. Kravanja, S., Kravanja, Z., Bedenik, B.S. 1998c. The MINLP optimization approach to structural synthesis. Part III: Synthesis of roller and sliding hydraulic steel gate structures. International Journal for Numerical Methods in Engineering 43: 329–364. Kravanja, S., Soršak, A., Kravanja, Z. 2003. Efficient Multilevel MINLP Strategies for Solving Large Combinatorial Problems in Engineering. Optim. Engng. 4: 97–151. Kravanja, S., Šilih, S., Kravanja, Z. 2005. The multilevel MINLP optimization approach to structural synthesis: the simultaneous topology, material, standard and rounded dimension optimization. Adv. Eng. Softw. 36 (9): 568–583. Kravanja, Z., Grossmann, I.E. 1994. New Developments and Capabilities in PROSYN—An Automated Topology and Parameter Process Synthesizer. Computers Chem. Engng. 18(11/12): 1097–1114. Mawengkang, H., Murtagh, B.A. 1986. Solving Nonlinear Integer Programs with Large-Scale Optimization Software. Ann. Oper. Res. 5: 425–37. Olsen, G.R., Vanderplaats, G.N. 1989. Method for Nonlinear Optimization with Discrete Design Variables. AIAA J 27(11): 1584–9. Quesada, I., Grossmann, I.E. 1992. An LP/NLP Based Branch and Bound Algorithm for Convex MINLP Optimization Problems. Computers Chem. Engng. 16: 937–47. Westerlund, T., Pettersson, F. 1995, An extended cutting plane method for solving convex MINLP problems. In: European Symposium on Computer Aided Process Engineering-5, Supplement to Computers Chem. Engng. Bled, Slovenia: 131–6.

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Construction materials

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Basic study on physical property for calcium-solidification material and on Ca-based concrete A. Shimabukuro & K. Hashimoto Tokuyama College of Technology, Shunan, Japan

ABSTRACT: The Ca-based solidification material has neutral character for pH and natural color like soils, so that it is paid attention to as a new material in civil engineering field instead of cement. In this study, using this material, the physical property tests in consideration of influence of weathering, density test and bending test, were performed to investigate whether to be an effective material similar to cement. Ca-concretes were made by Ca-based solidification material using results of the density test and the strength tests for compression, tension, and bending were conducted to consider strength properties for Ca-concretes. As a result, Ca concrete has sufficient strength similar to cement concrete. Therefore, it is clear that Ca concrete is new promising material in civil engineering field. 1

INTRODUCTION

In the construction industry, many concretes are used for the civil engineering structure from the 19th century to the present and are fundamental subsistent for our daily life. However, when the concrete loses the function by deterioration and is dismantled, the application of recycled aggregate made by the dismantled concrete is limited because compression strength of concrete made by recycled aggregate comes down than concrete made by the ordinary aggregate (Kawano et al. 1994). In addition, concrete is pointed out about some demerits to the environment and the landscape because cement has frosty and heavy image from ash color, so that it inflicts bad influence for the landscape. In recent years, the study of the eco-cement which used rubbish and industrial waste (Ohsumi 2003), eco-concrete (Japan Concrete Institute 1998), and concrete in multi-natural type, which have the function of afforestation or creation of underwater forest (Akida et al. 1996; Advanced Construction Technology Center 2001; Tanaka et al. 2005; Yoshida et al. 2006), is promoted. So, we think that we must find the new concrete in consideration of the environment and the landscape. Since the purpose in this study considers about civil engineering material which considered the environment and solidification material instead of cement, we focus attention on Ca-based solidification material which has neutral character for pH and natural color like soil. First, the density test of Ca-based solidification material and slag cement and the strength test of mortar for each materials are performed to examine whether this Ca-based solidification materials is effective as civil engineering material like cement. Secondly, using

435

result of density test for Ca-based solidification material, the design of mix is carried out to make the specimen of concrete used by Ca-based solidification material (the following, Ca-concrete). The specimens of Ca-concrete are made by this result to perform some strength tests which are compression, tension, and bending test. These specimens cure for one week or four weeks at water curing after making specimens. The strength characteristics of Ca-concrete from each strength tests after water curing are comprehended and we evaluate whether Ca-concrete is effective on strength as civil engineering material. 2

CA-BASED SOLIDIFICATION MATERIAL

Ca-based solidification material (Japan Conservation Engineers Co., Ltd. 2002) has brown similar to natural soil and is solidification material made by mineral component. The main purpose is application to recreation trail and grass proofing at slope. The characteristics of this material are permeability to slowly percolate rain water and water retentivity to prevent desiccation, therefore, this material is effective for countermeasure of heat island. In addition, other characteristics following; 1. This material has no efflorescence phenomenon such as the cement. 2. This material is able to recycle as recycled spraying material. 3. The pH at the time of the spraying is approximately neutral. Thus, Ca-based solidification material is a challenging material for the environment and the landscape.

Table 1. Results of chemical composition analyses for Ca-based solidification material.

Table 2.

Mix proportion quantity of Ca concrete.

Material

Quantity kg/m3

Chemical composition

Results of analyses %

SiO2 CaO Al2 O3 MgO Fe2 O3 K2 O SO3 TiO2 Na2 O MnO P2 O5 BaO Cr2 O3 ZrO2

35.91 28.48 25.37 2.84 2.29 1.48 1.34 1.18 0.71 0.12 0.10 0.09 0.05 0.04

Water Ca-based solidification material Fine aggregate Coarse aggregate

165 275 810 996

Here, Table 1 is shown the analysis results of qualitative and quantitative in this solidification material.

specimen. Here, the sand for mortar specimen uses Toyoura sand and the mixture ratio of this specimen is weight ratio of solidification material 1, water 0.5, and Toyoura sand 3. In addition, the method of making this specimen is made by tamping method and the size of this specimen is cross section 40 mm2 and 160 mm in length. This specimen cures in water for period of 7-day or 28-day after making it, then the bending strength and compression strength at each period is measured. Water curing cures at state that whole of specimen is satisfied by water and water temperature is 22 degrees Celsius on the average.

3

3.2

3.1

EXPERIMENTAL OVERVIEW

The design of mix for Ca-concrete has no provision, so that it conforms to design of mix that mixed condition of cement concrete is water-cement ratio of 60%, slump of 8 cm, maximum size of coarse aggregate of 20 mm, and air content of 6.0%. Table 2 is shown about required amount of each material based on this condition for Ca-concrete. Here, used fine aggregate is sea sand and this fineness ratio is 2.70.

Density test and strength test

Cement is not effective material to long-term use because weathering deteriorates its density, strength, and quality. Therefore, in this study, the physical property tests of cement ruled by Japanese Industrial Standard, which are density test and strength test, are performed by using Ca-based solidification material to examine how the properties of density and strength for Ca-based solidification material change by weathering and whether this material must pay attention to reposition like cement. Here, in the above strength test, although mixture ratio and the methods of bending tests and compression tests after bending tests conform to Japanese Industrial Standard, the method of making specimen and curing method do not conform to Japanese Industrial Standard because we do not have need equipments. First, Ca-based solidification material is left naturally to examine influence of weathering to predefined period after conservation bag in Ca-based solidification material is opened, then density test is performed at every this period. Predefined period of natural leaving is 0-day, immediately after opening, 7-day, 28-day, and 56-day and temperature and humidity at 13:00 in leaving place are respectively 30.6 degrees Celsius and 75.6% on average. Secondly, the design of mix of mortar specimens for bending test is carried out by using result of that density test, and then the strength tests of bending and compression after bending test are performed to this

The design of mix for Ca-concrete

3.3

The strength tests for Ca-concrete

The strength tests for Ca-concrete is performed for compression, tension and bending. From results of these tests, we comprehend the strength characteristics of Ca-concrete. 3.3.1 Compression test Compression test is carried out by method of displacement control. The velocity of displacement is 0.6*10−2 mm/sec. Specimen is loaded until complete fracture. Compression strength is calculated from Equation 1 by maximum load shown by testing machine. The measurement of Acoustic Emission (AE) is also carried out to examine microfracture in this test. Fc =

P π(d/2)2

(1)

where Fc = compression strength; P = maximum load; and d = diameter.

436

Ft =

2P π dl

(2)

3.15 3.1 3

material density (g/cm )

3.3.2 Tension test Tension test is carried out based on the method of tension test for concrete conformed to Japanese Industrial Standard (JIS A 1113-2006). Specimen is loaded by the method of stress control until reached to maximum loading. The velocity of loading is controlled to become 0.06 ± 0.04 N/mm2 per second. Tension strength is calculated from Equation 2.

where Ft = tension strength; P = maximum load; d = diameter; and l = specimen length.

3.05 3 2.95 2.9

Ca-based solidification material slag cement

2.85 2.8

Pl Fb = 2 bh

(3)

where Fb = bending strength; P = maximum load; l span; b = width of fracture cross section; and h = height of fracture cross section. 4

EXPERIMENTAL RESULTS AND DISCUSSION

0 1 (immediately after opening)

2

3

4

5

6

7

8

leaving period in air (week)

Figure 1. Relationship between density of each materials and leaving period in air. 6

2

28-days bending strength (N/mm )

3.3.3 Bending test Bending test is carried out based on the method of bending test for concrete conformed to Japanese Industrial Standard (JIS A 1106-2006). Specimen is loaded by the method of stress control until reached to maximum loading. The velocity of loading is controlled to become region between 1.2 to 6.0 N/mm2 per a minute at edge stress intensity. Bending strength is calculated from Equation 3.

5

4

3

2 Ca-based solidification material slag cement Ca-based solidification material(avg.) slag cement (avg.)

1

0

4.1 Result of physical property test

0

4.1.1 Influence of weathering to solidification material for density Figure 1 is shown about relationship between density of Ca-based solidification material and leaving period in air. Figure 1 is also shown similar experimental result that compared with slag cement. Here, used slag cement is type of B. In this figure, density of both materials decreases when weeks pass. But, density of Ca-based solidification material is hard to deteriorate than slag cement if result is estimated until 56-day. Therefore, weathering of cement typically occurs after cement becomes solid form by absorbing moisture and carbon dioxide in air, so that density of cement decreases by influence of weathering. However, it is considered that density of Ca-based solidification material does not decrease than slag cement because density of Ca-based solidification material does not particularly have influence described above, solid form made by absorbing moisture and carbon dioxide in air, on weathering.

437

(immediately after opening)

1

2

3

4

5

6

7

8

leaving period in air (week)

Figure 2. Relationship between 28-days bending strength and leaving period in air.

4.1.2 Influence of weathering to solidification material for bending strength of mortar Figure 2 is shown about relationship between 28-day bending strength of mortar and leaving period in air for both materials. Solid line and dashed line in this figure links the average strength of three specimens for the 28-day bending test. In this figure, bending strength of slag cement immediately after opening is bigger than Ca-based solidification material. But, in result from 0-day to 28-day, bending strength of slag cement goes down with elapsing of leaving period. On the other hand, bending strength of Ca-based solidification material does not go down with elapsing of leaving period.

30

2.2

compression stress (N/mm )

25

2

slag cement

3

average specimen density (g/cm )

Ca-based solidification material

2.15

2.1

20

15

14.57 N/mm2 10

5

2.05 0 -3000

vertical strain lateral strain volumetric strain

-2000

-1000

28-day specimen 0

1

2

3

4

5

6

7

2000

1000

2000

(a) 7-day specimen

8 30

leaving period in air (week)

Therefore, Ca-based solidification material is considered as material that strength deterioration does not occur easily if it is left in air. Figure 3 is shown about relationship between average specimen density of three specimens for bending test at 28-day and leaving period in air. Specimen density and strength have close relationship. Comparing this figure and Figure 2, it shows that bending strength of slag cement also deteriorates because specimen density of slag cement obviously goes down with elapsing of leaving period. But, specimen density of Ca-based solidification material does not show significant change with elapsing of leaving period. Therefore, Ca-based solidification material does not particularly have influence to making specimen even if it is left in air. It is considered that Ca-based solidification material slightly has influence of weathering because its density goes down at Figure 1. However, it is insusceptible materials to weathering because deterioration of strength and specimen density by weathering does not be shown like slag cement. Thus, Ca-based solidification material is thought material that does not need to much consider for safekeeping. Result of strength test for Ca-concrete

4.2.1 Compression characteristic Density of Ca-based solidification material is 3.10 g/cm3 from result of density test at immediately after opening. Specimens of Ca-concrete for strength test are made by using this density from foregoing design of mix.

25

2

compression stress (N/mm )

Figure 3. Relationship between average density of specimen and leaving period in air.

4.2

1000 -6

2 (immediately after opening)

0

strain (*10 )

2

25.76 N/mm

20

15

10

5

0 -3000

vertical strain lateral strain volumetric strain

-2000

-1000

0 -6

strain (*10 ) (b) 28-day specimen

Figure 4. The examples of relationship between compression stress and strain for Ca concrete specimen.

Figure 4 is shown the representative example for Ca-concrete about relationship between compression stress and strain and (a) and (b) are shown as 7-day specimen and 28-day specimen, respectively. In relationship between compression stress and vertical strain, the behavior under compression loading for Ca-concrete shows tendency similar to cement concrete because its behavior shows non-linear behavior. In addition, when microfracture is considered, estimation of dilatancy can estimate microfracture because dilatancy occurs by microfracture before macro fracture (Patterson 1986). Here, dilatancy is defined as inelastic increase of volume, so that if relationship between compression stress and volumetric strain in Figure 4 is used, it is possible to consider about microfracture. Therefore, when linear gradient changes to non-linear gradient in this relation, the dilatancy is defined in this study as Dilatancy stress point.

438

30

25

25

2

compression stress (N/mm )

2

compression strength (N/mm )

30

20

15

10

20

15

14.37N/mm2 10

5

5 0 0

0 7-day

50

28-day

100

150

200

AE count rate (1/sec) (a) 7-day specimen

curing period 30

From this stress point, it is considered that Caconcrete for 7-day specimen begins to break at about 15 N/mm2 because curve line of stress-volumetric strain relation changes near this point. And peak stress of this specimen is about 16 N/mm2 , so that this specimen can not maintain only 1 N/mm2 extent from microfracture. 28-day specimen also show similar tendency. Therefore, fracture behavior of Ca-concrete shows brittle behavior that specimen is broken as soon as when microfracture occurs. Figure 5 is shown about relationship between compression strength and curing period. This compression strength is average value of three specimens in each curing period and is defined as the maximum value of compression stress in compression stress-strain relation shown in Figure 4. In this figure, it shows on compression strength of Ca-concrete that 28-day strength is stronger than 7-day strength and strength increasing ratio after 28-day is 1.57 extents. Furthermore, referential strength of cement concrete under same mixture condition with Ca-based concrete is presumed as 27 N/mm2 on plain concrete and as 24 N/mm2 on AE concrete (Kobayashi 2006). Consequently, it can presume that compression strength of Ca-concrete and AE concrete have equivalent strength under same mixture condition. Therefore, Ca-concrete will be effective material on strength as new civil engineering material. 4.2.2 Compression stress and AE relation On the other method to consider microfracture, method estimating Acoustic Emission is also used in this study. Figure 6 is shown representative example for Caconcrete about relationship between compression stress and AE and (a) and (b) are shown as 7-day

439

compression stress (N/mm2)

Figure 5. Relationship between compression strength and curing period.

25

2

24.60N/mm 20

15

10

5

0 0

50

100

150

200

AE count rate (1/sec) (b) 28-day specimen

Figure 6. The examples of relationship between compression stress and AE count rate for Ca concrete specimen.

specimen and 28-day specimen, respectively. Numerical value in both figures is stress point that AE count rate increases rapidly. Microfracture occurs at this point and this point is defined as AE stress point in this study. AE count rate in these figures occurs from initial loading and denotes dispersion. This shows that smaller fracture before microfracture occurs at every moment in internal specimen. Here, AE stress point in Figure 6 (a) is 14.37 N/mm2 . It is considered that this point is approximately equal with foregoing Dilatancy stress point. Therefore, it is also possible to consider microfracture by AE on Ca-concrete because the relationships between AE stress point and Dilatancy stress point usually have high relativity. In addition, when compression stress of Ca-concrete exceeds AE stress point, AE count rate in 28-day specimen does not increase rapidly and repeats a few amplitude, but it in 7-day specimen increases rapidly. From this result, when microfracture of Ca-concrete is estimated by AE,

Table 3.

Strength ratio for Ca concrete.

Test

Average strength for 28-day N/mm2

Strength ratio

Compression Tension Bending

24.81 3.07 4.66

1 0.123 0.188

it is considered that 7-day specimen shows fracture behavior which is more brittle than 28-day specimen after microfracture. 4.2.3 Ratio of compression strength and each strength Table 3 is shown about 28-day average strength for Caconcrete of compression, tension, and bending tests. Table 3 is also shown about ratio of tension strength and bending strength for compression strength. In this table, Strength ratio of tension for Caconcrete is 0.123. This is shown that Ca-concrete is slightly stronger than cement concrete for tension because strength ratio of tension for cement concrete is region of 0.07 to 0.1. In bending strength, bending strength for Caconcrete is 4.66 N/mm2 . It is possible to apply to road pavement concrete because standard of design bending strength in road pavement concrete is ruled as 4.5 N/mm2 . And strength ratio of bending for Caconcrete is 0.188. This is shown that compression strength-bending strength relation for Ca-concrete has tendency similar to cement concrete because strength ratio of bending for cement concrete is region of 0.125 to 0.2. 5

CONCLUSIONS

In this study, physical property as civil engineering material of Ca-based solidification material is investigated as paying attention to weathering. From this result of physical test, Ca-concretes are made by Cabased solidification material and the strength tests for compression, tension, and bending are conducted to consider strength properties for Ca-concretes. The knowledge obtained in this study is the following. 1. From result of density test leaving from 0-day to 56-day, density of Ca-based solidification material is hard to deteriorate than slag cement. Therefore, density of Ca-based solidification material especially has no influence of weathering. 2. From the relationship between bending strength of mortar and leaving period, bending strength of

slag cement goes down with elapsing of leaving period. On the other hand, bending strength of Ca-based solidification material does not go down with elapsing of leaving period. Therefore, Cabased solidification material is considered as material that strength deterioration by weathering does not occur easily if it is left in air. 3. The behavior under compression loading for Caconcrete shows tendency similar to cement concrete because its behavior shows non-linear behavior. In addition, it can presume that compression strength of Ca-concrete and compression strength of AE concrete have equivalent strength under same mixture condition, so that Ca-concrete will be effective material on strength as new civil engineering material. 4. Dilatancy method and AE count rate method are effective to microfracture estimation for Caconcrete. 5. In Ca-concrete, compression strength-tension strength relation is slightly stronger than cement concrete and compression strength-bending strength relation has tendency similar to cement concrete. In addition, bending strength is possible to apply to road pavement concrete because standard of design bending strength in road pavement concrete is ruled as 4.5 N/mm2 . REFERENCES Advanced Construction Technology Center. 2001. Handbook of river revetment method by porous concrete. Tokyo: Sankaidou. Akida, et al. 1996. Epiphytic condition of seaweed to concrete surface by different of surface treatment. Journal of Coastal Engineering in Japanese, 43: 1246–1250. Japan Concrete Institute. 1998. Concrete engineering. Vol. 36, No. 3: Gihodo Syuppan. Japan Conservation Engineers Co., Ltd., 2002. Proposal method by Ca-based solidification material. Japan. Kawano, K. et al. 2000. New concrete engineering. Tokyo: Asakura Shoten. Kobayashi, K. 2006. New concrete engineering. Tokyo: Morikita Syuppan. Ohsumi, M. 2003. Tale of eco cement. Tokyo: Japanese Standards Association. Paterson, S.M. 1978. Experimental Rock Deformation—The Brittle Field. Berlin, Heidelberg, New York: SpringerVerlag. Tanaka, et al. 2005, Development and application of porous concrete which purpose creation of underwater forest. Proc. of Japan Society of Civil Engineers Chugoku regional branch office in Japanese, 57: 430–431. Yoshida, et al. 2006. Study on creation of underwater forest— application of porous concrete. Concrete Engineering series in Japanese, 72: 164–169.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Correlations between filler type and the self compacted concrete properties M. Gheorghe, N. Saca & L. Radu Technical University of Civil Engineering of Bucharest, Romania

ABSTRACT: The work aim was to study the influence on limestone filler (L)—as reference—and volcanic tuff filler (VT), and other main compositional parameters on the self compacted concrete (SCC) properties referring to workability and to compression strengths time-evolution. The SCC workability defined by the two rheological parameters, a low yield stress was carried out by SCC slump flowability and the plastic viscosity was evaluated as SCC flow resistance measured by flow time, T500 . The relative consistence of the SCC precursory cement—filler paste, the aggregate particle size distribution (PSD) and the fine particles/coarse particles volume ratios were the main parameters of SCC mix design. The limestone powder was played a beneficial role on the fresh SCC mobility and on the earlier-compression strengths. The volcanic tuff powder, a veritable pozzolana, has had favorable to later-compression and split tensile strengths development.

1

achieved by rheological properties of fresh concrete, namely low yield stress value associated with a properly flow resistance defined by plastic viscosity. The yield stress point and plastic viscosity, the principal SCC rheological characteristics, were correlated with powders volume and paste volume (which control plastic viscosity) and with superplasticizers—SP (for yield stress value decreasing). The fillers adjust plastic viscosity, increase segregation resistance, modify matrix porosity and pores distribution and, as following, concrete humidity which controls shrinkage and creep strains (Bosiljkov 2003). Also, the pozzolanic filler by pozzolanic reaction contributes to the strength structure development and to perform a better resistance to chemical attack (Teoreanu 1975; Gheorghe et al. 2006; CEEX-National Research Programme-Technical Reports since 2006–2007). A pioneer, remarkable study about hydraulic activity of trass, as powder state of volcanic tuff, was carried out by famous Romanian scientist Ionescu Bujor PhD, (Bujor 1923).

INTRODUCTION

Self compacted concrete (SCC) is a fluid concrete that is compacted at maximum possible density under its own weight and it is self finished. The SCC main characteristic is workability, measured by its flowability, viscosity, passing ability and segregation resistance. Self compacting concrete has generated a great interest since its initial development in Japan (Okamura & Ozawa 1995; Okamura 1997; Okamura & Ouchi 2003). The advantages of SCC include its high ability to flowing and filling of densely reinforced or geometrically complicated structural elements, without segregation or excessive bleeding, reduced labor costs and diminished noise level due to vibrators lack during placement, as according to European Guidelines for Self compacting Concrete (EFNARC, European Guidelines for Self compacting Concrete 2002, 2005, 2006). The SCC may be included into green concrete area, due to its specific advantages—energy saving and noise pollution diminishing (Assie et al. 2007; Khatib 2007). The SCC producing needs increased costs of materials (e.g. mineral powders additions and chemical admixtures), and also needs a good technical expertise for mix proportioning, processing, and its target workability and strengths values achievement (Domone 2007). The SCC mix design seems to approach into various formulas, but it is describing onto common base referring to the fresh concrete rheological properties and to particle size modeling starting with the packing of all aggregates at a minimal interparticle voids to fill with optimal volume of paste (Okamura & Ouchi 2003; Yurugi et al. 2000; Saak et al. 2001; Su et al. 2001; Chao-Wei et al. 2001). The ability to fill up the formwork and encapsulate of reinforcing bars through the own weight only is

2 2.1

EXPERIMENTAL Materials

The material selection—cement, fillers, aggregates and superplasticizer (SP) admixture was the first sequence of the research program. The cements (EN 197-1), CEM I 42, 5R, and CEM I 52, 5R with specific area of 3300 cm2 , and, respectively of 4088 cm2 /g, were used. The limestone filler (L) and the volcanic tuff filler (VT) were used as powder additions of SCC. The filler L was a commercial product with a calcite content of

441

98.5%. The filler VT was obtained by grinding waste recovered from processing of parental rock of PersaniBrasov, Romania. This filler may be considered a secondary raw material recovered for self compacted concrete technology. A HITACHI S2600N scanning electron microscope (SEM) was used to image L and VT fillers surface morphology and EDAX microprobe study of element chemical speciations. Differential thermal analyse of fillers were made using a MOM derivatograph. The volcanic tuff mineralogical origin was described by zheolithes and tainted feldspars and its pozzolanic activity index experimentally determined was of 90.5% The L and VT fillers other characteristics are listed in Table 1. The shape and surface texture, are main parameter that controlling fresh SCC mobility. The morphology and surface characteristics were investigated by SEM and can be observed in SEM imagines of Figures 1 and 2. The L filler was characterized by more rounded and smoothly surface than VT filler. Also, the K, Ca silicoaluminate zeolithe compound preponderance—probably philipsithe (Ca, K2 [Al3 Si5 O16 ]·7H2 O as main pozzolanic mineral, (as according to Brana et al. 1986) of volcanic tuff is suggested by EDAX spectrum showed in Figure 3. The physical characteristics of the total aggregate are listed in Table 2. Investigation by thermal analysis, according to TG and DTG effects was showed a 33% weight loss into temperature area of 680 . . . 900◦ C of limestone filler, as CO2 value closed of 44% the theoretic one. The volcanic tuff, as zeolithic minerale, was characterized by thel zheolitic water loss of 8.83% total, corresponding to endothermic dehydration at 119 and 192◦ C. 2.2

Figure 1. Morphology of volcanic tuff particles SEM imagine × 1000.

Figure 2. Morphology of limestone particles SEM imagine ×1000.

SCC mix design and proportioning

At the beginning the SCC mix proportion design was conducted based on the target values for fresh SCC workability and the 28 days—compression strengths. The SCC mix design was sequential. Firstly, it was carried out the VW /VP for a zero flowing of the paste, measured as slump-flow area for different proportioning of mix cement-filler. There were prepared pastes of cement + filler + sand 0125 with different VW /VP ratios. The paste consistence was Figure 3. Table 1.

The volcanic tuff EDAX spectrum.

The characteristics of used fillers. Main oxide compounds, %

Filler

CaO

SiO2

Al2 O3

Density, kg/m3

Particles with D < 0,063 mm%

Limestone Volcanic tuff

51.4 3.5

0.6 62

0.4 13.5

2678.4 2028.2

62.5 70.1

tested by slump-flow test and measuring the flow area. The relative consistence as measure of flow ability was calculated with the equation: p =



442

Dm D0

2 −1

(1)

Table 2. The aggregate physical characteristics.

1.5

River coarse aggregate 4/16 mm

Crushed stone 4/8 mm

Volumic mass, kg/m3 Packed density, kg/m3 Water absorption, %

2600 1702 nd

2665 1665 4.6

2674 1651 4.8

1.3 W/P ratio

Characteristic

River sand

1.1 0.9 βp 0.7

y = 0.1099x + 0.821 R 2 = 0.9512

0.5 0

2.3

2 3 Relative slump-flow, Γp

4

5

Figure 4. Determination of βp of paste group with cement: L powder ratio = 1 : 0.11. 1.4

W/P ratio

1.2 1 p 0.8

y = 0.0893x + 0.901 R2 = 0.9457

0.6 0

1

2

3

4

Figure 5. Determination of βp of paste group with cement: L filler ratio = 1 : 0.2. 1.5 1.3 W/P ratio

where Dm = the average value of slump-flow area, mm ; D0 = the conic vessel diameter. In Figs. 4–6 are plotted the graphics of the VW /VP influence on the cement-filler paste relative consistence. The βp values extrapoled were varied as function of filet type L or VT and cement : filler (L or VT) ratio. The βp values obtained vary in the 0.82 . . . 0, 99 areas. The limestone filler increasing from 110 to 160 kg/m3 (corresponding to cement : filler L ratios of 1:0.11 and respectively 1:0.2) was determined W/VP values of 0,82 and 0,90 respectively. The volcanic tuff filler showed a greater water demand than limestone filler, concretized by a βp value of 0,99 at the same cement : filler ratio of 1:0.2. The pastes with volcanic tuff needed bigger water content, due to the higher specific surface particles greater content. The βp value was used mainly for the evaluation of the water demand for new batches of powders (cement + fillers). The SP admixture compatibility with the cementfiller powder was carried out by slump—flow area measuring of pastes with different admixture dosages and keeping constant VW /VP ratio. The final self compacting concrete design was carried out by trial mixes. The optimal aggregate particle size distribution (PSD) and as subsidiary, sand/coarse aggregate volume ratio were established by workability successive tests. Also, the SCC mixes were carried out with Dmax = 8 mm crushed aggregate (mixes L1, V1, L4 and V4), with Dmax = 16 mm river coarse aggregate (mixes L2, V2, L3 and V3), and with relative high powder volumes, from 190 to 216 L. The high powder volumes were correlated with the very workable and segregation stable SCC, and with an accelerated hardening rate required by densely reinforced precast elements.

1

1.1 βp

0.9

y = 0.074x + 0.9924 R2 = 0.9722

0.7 0.5 0

1

2

3

4

5

6

Relative slump flow, Γp

Figure 6. Determination of βp of paste group with cement: VT filler ratio = 1:0.2.

Concrete mixes and testing

In summary, the SCC mixes with cement dosages of 420 . . . 500 kg/m3 area, and W/C ratios from 0.38 to 0.46—as correlating to proposed compression strength class—have been prepared. Thus, the four SCC pairs have been prepared, with L filler (code

443

L1, L2, L3 and L4) and, respectively, with VT filler (code V1, V2, V3 and V4), listed in table 3. The SCC sample s were prepared into laboratory concrete mixer of (50 liters volume) The cubic samples, 15 cm edge were prepared and were cured in tape water up to compression strength term (2, 28, 56, 180 and 360 days) testing and 28 days—splitting tensile strength testing according to concrete standard procedures. The SCC workability was tested as according to EFNARC European Guidelines for Self compacting Concrete 2002, 2005. Segregation stability index was evaluated (according to Saak et al. 2001). The target values of

SCC workability were established as; SF2/SF3 consistence class (680 . . . 780 mm, slump-flow); VS2 viscosity class (as T500 > 2 s); and PA2 class L-box passing ability with 3 bars (H2/H1 ≥ 0.8). 3

experimental data it can be appreciated the limestone filler contribution to increasing SCC mobility. The volcanic tuff filler has had an unfavorable influence on SCC mobility and was needed a SP admixture surplus of 0.2 . . . 0.4%, for SCC workability criteria fulfilling. The superplasticizer and VW /VP ratio were fulfilled a decisive role on the SCC flowability, viscosity and segregation stability. The hardened SCC properties—density, compression strengths and splite tensile strength are listed in Table 4. The 360 days-compression strengths development was continuously ascendant. There are note a different increase rate as function of SCC compositional parameters variation. The structural SCC precast elements technology is requiring a great hardening rate, as an important technical and economical parameter. This technological aspect was cause to

RESULTS AND COMMENTS

The fresh SCC properties are listed in table 3. The entrained air has values up to 2.8% due superplasticizer (SP) admixture supplementary function as air entrains too. Slump flow area was frequently around 700 . . . 740 mm, except L4/V4 mixes with a greater SP amount. Also aggregate particle size contributed to modifying viscosity, as T500 flow time, e.g. sample pair L1-V1 and sample pair L4 and V4, with the greatest powder volumes. According to the Table 3.

Mix proportions and fresh properties of self compacted concrete mixes.

PARAMETER

L1

V1

L2

V2

L3

V3

L4

V4

Cement CEM I 42.5R, Cement CEM I 52.5R, kg/m3 Limestone powder, kg/m3 Volcanic tuff powder, kg/m3 River sand, kg/m3 Crushed stone, Dmax = 8 mm, kg/m3 Coarse aggregate, Dmax = 16 mm, kg/m3 Superplasticizer, % Water, L/m3

476 0 110 0 933 621 0 0.7 200

480 0 0 110 992 534 0 0.7 200

470 0 128 0 965 0 643 1 192

462 0 0 125 926 0 617 1 200

420 0 136 0 993 0 662 1 194

427 0 0 133 941 0 627 1.4 194

0 487 90 0 1025 684 0 1.7 185

0 500 0 80 1081 533 0 1.9 185

Water/Cement, W/C Ratio Powder , VP *, L/m3 Water/powder, W/P weight ratio Water/powder, Vw /Vp volumic ratio

0.42 216 0.33 0.93

0.42 216 0.32 0.93

0.41 202 0.31 0.95

0.43 206 0.32 0.97

0.46 190 0.33 1.02

0.45 208 0.33 0.93

0.38 203 0.31 0.91

0.37 214 0.30 0.86

Entrained air, % Unit weight, kg/m3 Slump-flow area, mm Viscosity, T500 , sec. L box, H2/H1, % Stability index

2.1 2341 700 6 0.85 0

3 2312 680 9 0,80 0

1.5 2330 730 4 0.85 0

1.6 2326 700 7 0.90 0

2.7 2362 740 3 0.92 0

2.8 2334 720 7 0.92 0

2.1 2334 760 5 0.92 0

2.8 2356 800 9 0.95 1

kg/m3

*VP = Vcem + Vfille r + Vsand < 125 mm . Table 4.

Hardened self compacted concrete characteristics.

SCC code

Volumic mass, ρa , kg/m3

L1 V1 L2 V2 L3 V3 L4 V4

2415 2361 2431 2380 2384 2349 2382 2395

Compression strength, MPa fc2

fc28

fc56

fc90

fc180

fc360

Split tensile strength, MPa ftd28

42.4 36.6 47.5 39.6 37.2 38.2 50.4 46.6

54.0 52.5 56.5 49.5 49.0 52.5 65.1 67.5

67.0 64.0 64.0 51.0 53.5 58.0 69.2 77.0

71.5 70.5 78.0 69.5 68.5 69.2 85.8 100.0

83.0 85.6 81.1 74.6 73.4 80.2 90.2 105.3

85.1 89.1 83.5 84.1 75.2 81.0 92.1 107.2

5.9 7.6 5.3 5.4 4.8 7.4 8.4 9.3

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etude the filler influence on the early term compression strength. The filler type has influenced the earlier and longer term-compression strengths, aspect which is showed graphically in Figures 7 and 8. The 28 dayscompression strength (fc28 ), varied from 49.5 to 67.5 MPa, in correlation with the W/P ratios, and with the cement (type and dosage). The 28 days-split tensile strength (ftd28 ), obtained testing values varied from 5.3 to 9.3 MPa. It was noted that the volcanic tuff SCC split tensile strength was constantly higher than those of the limestone filler SCC. This phenomenon could be explained by microcracks diminishing due their interruptions with the hydrosilicates precipitated by pozzolanic reaction. The new C-S-H gelic structures diminish microcracks and pores of transition zone at interface, the most vulnerable at split tensile break (Gheorghe et al. 2006). The limestone filler contribution to earlier compression strengths was better than the volcanic tuff filler one, for all the SCC samples and this process

0.9 0.8

fc2/fc28 Ratio

0.7 0.6

have been accentuated by the use of CEM 52, 5R cement, comparably to CEM 42, 5R cement use. The volcanic tuff filler was appropriated to later compression strengths better development, as described by f360 /fc2 ratio which was included in the 2.1 . . .. 2.4 values area greater than those of the limestone filler SCC included in the 1.8 . . .. 2 values area. The limestone filler particles contribution to early mechanical strengths development may be argued as an effect of the crystallization Ca(OH)2 nucleation sites (Dhir et al. 2007; Dumitrescu et al. 2007), as following of the potential calcite reactivity (Rahhal & Talero 2005), determined by the mineralogical origin. The volcanic tuff powder is preponderantly, composed of the zheolites (aluminosilicate minerals), with vitreous structure, conformed to the mineralogical structure of the parental rock. It is known that the vitreous aluminosilicate structure is the cause of volcanic tuff the pozzolanic activity. In this studied case the pozzolanic index experimentally determined has had a 90.5% value that suggested a good hydraulic activity. Consequently, the VT filler hydraulic activity was visible, in comparison with the reference SCC with L filler, and may explain the later age-compression strengths better development. To summarize, the limestone filler had contributed to early term-compression strengths, and volcanic tuff, as veritable pozzolana, has contributed to the later term-compression strength.

0.5 0.4 0.3

4

CONCLUSIONS

0.2 0.1 0 L1

V1

L2

V2

L3

V3

L4

V4

Figure 7. The filler influence on the SCC hardening rate, measured as fc2 /fc28 ratio; L1. . . L4 -SCC with limestone filler and V1. . . V4 -SCC with volcanic tuff filler. 3

fc360/fc2 Ratio

2.5 2 1.5 1 0.5 0 L1

V1

L2

V2

L3

V3

L4

V4

Figure 8. The volcanic tuff influence on the longer term compression strength of SCC, measured as fc360 /fc2 ratio. L 1 . . . L4-SCC with limestone filler and V 1 . . . V4-SCC with volcanic tuff filler.

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The work aim was to study the influence on the limestone filler (L)—as reference—and volcanic tuff filler (VT) on the self compacted concrete main properties referring to workability and compression strengths time-evolution. The volcanic tuff filler was a secondary raw material obtained by milling of waste from parental rock processing as dimensional construction products. The SCC mix design was based on the rheological properties of the SCC precursor pastes for optimizing of the VW /VP ratio of the SP type and dosage, with the fresh SCC performance criteria. The SCC workability was decisively influenced by the water/powder ratios, cement type and dosages. As according to the experimental data it can be appreciated that L filler was beneficial for the fresh state SCC increased mobility. The VT filler addition to SCC was, generally, determined by the HRWR admixture increase for reaching of the SCC accepted workability criteria. The mechanical strengths values and their development up to 360 days age were significantly controlled by main compositional parameters namely W/P ratio, cement class and filler type (L or VT). The limestone filler contribution to earlier compression strengths was significant and this process has been accentuated by the use of CEM 52, 5R. The volcanic tuff filler contributed to a good

compression strength development, and it didn’t significantly diminish early strength values, being appropriated to later compression strengths development, aspect described by fc360 /fc2 ratio which was included in the 2.1 . . . 2.4 values area greater than those of the limestone filler BAC included in the 1.7 . . . . 2 values area. The volcanic tuff SCC split tensile strength was constantly higher than those of the limestone filler SCC. This phenomenon could be explained by diminish of microcracks due their interruptions with the hydrosilicates precipitated by pozzolanic reaction of volcanic tuff. ACKNOWLEDGMENTS We gratefully acknowledge the financial support of PNCD2 contract no. 32-133/2008. REFERENCES Assie, S., Escadeillas, G. & Walter, V., 2007, Estimates of self compacting concrete ‘‘potential’’ durability, Construction and Building Materials, 21 (10), 1909–1918. Bosiljkov Bokan, V., 2003, SCC mixes with poorly graded aggregate and high volume of limestone filler, Cement and Concrete Research, 33(9), 1279–1286. Brana, V., Avramescu, C. & Calugaru, I., 1986, Nonmetalic mineral substances. Technical Publishing House, Bucharest. CEEX-National Research-Development ProgrammeRomania: Project- Innovative solution of self compacted concrete microstructure optimization for performed precast elements SICOBET, Contract no. 96/2006, Technical Reports of Technical University of Civil Engineering of Bucharest, since 2006–2007. Chao-Wei, T., Tsong, Y., Chao-Shun, C. & Kuan-Hun, C., 2001, Optimizing mixture Proportions for flowable Highperformance concrete via Rheological Testes, ACI Materials Journal, 98 (6), 493–501. Dhir, R.K., Limbachiya, M.C., McCarthy, M.J. & Chaipanich, A., 2007, Evaluation of Portland limestone cements for use in concrete construction, Materials and Structures, 40(5), 459–466.

Domone, P.L., 2007, A review of the hardened mechanical properties of self-compacting concrete, Cement and Concrete Composites, 29 (1), 1–12 Dumitrescu, C., Menicu, M. & Voicu, G., 2007, Performant, ecological and economical hydraulic binder, Romanian Journal of Materials, 37 (4), 291–298. EFNARC (European Federation of Producers and Applicators of Specialist Products for Structure), 2005, The European Guidelines for Self-Compacting Concrete/ Specification, Production and Use. EFNARC, 2002, The European Guidelines for self compacting Concrete. EFNARC, EFCA, 2006, Guidelines for Viscosity Modifying Admixtures for self compacted Concrete. Gheorghe M. &reescu E. & Voinitchi, D, 2006, Aspects regarding the durability of the concretes based on high blast furnace slag content cement, Romanian Journal of Materials, 36 (1): 29–42. Ionescu Bujor, D., 1923, Technical and Petrografical study of trass rock from Bocsa—a new method on chemical determination of hydraulic value of trass, Romanian Academy, Edited by National Culture Foundation, Bucharest. Khatib, J.M. 2007, Performance of self-compacting concrete containing fly ash, Construction and Building Materials, 22 (9): 1963–1970. Lemaire, G., Escadeillas, G. & Ringot, E., 2005. Evaluating concrete surface using an image analysis process. Construction and Building Materials, 19(8), 604–611. Okamura, H. & Ozawa, K., 1995, Mix design for self compacting concrete, Concrete Library, of JSCE, 25 (6): 107–120. Okamura, H., 1997, Self Compacting High performance concrete, Concrete International, 19 (7), 50–54. Okamura, H., Ouchi, M., Self compacted concrete, 2003, Journal of Advanced Concrete Technology, 1(1), 5–15. Rahhal, V. & Talero, R. Early hydration of Portland cement with cristalline mineral additions, Cement and Concrete Research, 2005, 35(7), 1285. Saak, A.W., Jennings, H. M., Shah, S. & P., 2001, New Methodology for Designing Self Compacting Concrete, ACI Materials Journal, 98(6): 429–437. Su, N. Hsu, K.C. & Chai, H.V., 2001, A simple mix design method for self compacting concrete, Cement and Concrete Research, 31(12), 1799–1806. Teoreanu, I., 1975, The fundamentals of the binding materials technology, Technical Publishing House, Bucharest. Yurugi, M., Sakata, N., Iwai, M. & Mandsakay, G., 1993, Proc. Intern. Conf. Concrete 2000, Dundee, 579–586.

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Delphi study on Portland cement concrete specifications of ITD H. Sadid, V. Miyyapuram & R. Wabrek Idaho State University, Pocatello, Idaho, USA

ABSTRACT: Portland cement concrete (PCC) has been the most important building material for over 180 years, and, from an empirical perspective, its behavior is well understood. However, on a chemical basis, PCC is a complex material whose mechanisms and interaction are not fully comprehended. This study employs the Delphi survey technique to gauge the degree of consensus among a group of subject matter experts regarding the Idaho Transportation Department (ITD) Portland cement concrete (PCC) specifications and requirements in ITD projects. Two topics have been considered: the risk of materials failing to meet ITD-PCC specification and the consequences associated with materials failing to meet ITD specifications. This article describes how the Delphi technique was employed to evaluate ITD-PCC specifications and, from an expert perspective, gauges the degree of risk involved with different material acceptance requirements for PCC. 1 1.1

INTRODUCTION Concrete and significance of PCC specifications

Concrete is one of the oldest and most widely used construction materials. It is comprised mainly of cement, aggregate, water, and admixtures. It was first utilized by Egyptians and then by Romans around 5600 and 2500 B.C. respectively. The hydraulic Portland cement used today was first patented by Joseph Aspdin of Leeds in 1824 and was named Portland cement because of it resembles Portland stone found in Dorset, England. Portland cement concrete (PCC) is widely used in the construction of different structures including buildings, bridges, and pavement. The performance of PCC is highly dependant on the properties of its constituents, and PCC can be optimized by proportioning and using admixtures for enhancement. The hydration and the chemical reactions during the curing process are quite complex, however, from the empirical perspective, concrete’s behavior is well understood. The Idaho Department of Transportation (ITD) requires different methods in testing for material acceptance for use in concrete production in its projects. ITD specifies rigorous testing for some material but only a visual inspection or supplier’s certification for others. The subject of this study involves two basic questions: Why does ITD require different acceptance criteria for different materials, and what is the risk involved with each of these criteria? This study gauges the degree of consensus among a group of subject matter experts relative to the risks involved with material acceptance criteria and the

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consequences of concrete performance if the material fails some of the ITD acceptance criteria. Identifying the risks involved with each requirement for material acceptance will help the ITD in its decision making process involving specifications and testing requirements for PCC. 1.2

Delphi method

The Delphi method is a systematic, interactive, iterative process for consensus-building among a group of experts who are anonymous to each other. It recognizes the value of expert opinion, experience, and intuition and allows the use of expert opinions when limited information is available and full scientific knowledge is lacking (Linestone & Turloff 2002). The Delphi method was developed by Dalkey and Helmer of the Rand Corporation in the 1950’s to predict the probability, frequency, and intensity of possible enemy attacks (Dalkey & Helmer 1963). This method was originally designed to have a very narrow scope but over the years the applications of the method have broadened to include a wide range of methodologies. In addition to the method’s original intent of prediction, the Delphi method has been used to define objectives and as the basis for evaluating participant practices. The Delphi method has been employed in a variety of areas including government, medical, environmental and social studies, as well as business and industrial research (Linestone & Turloff 2002). The Delphi method is conducted anonymously to encourage a true debate among the participants, independent of personality. It does so by eliminating the force of oratory and pedagogy. The method is facilitated by a panel of researchers who refine questions

and develop a series of sequential questionnaires. The opinions of participant are shared without mentioning the source. This allows the participants to revise their decisions after being exposed to the opinions of the other subject matter experts (Linestone & Turloff 2002). The modes of communications of Delphi method have varied drastically since its inception. The conventional/classical Delphi was conducted by faceto-face interaction in a conference room or through letters. In the modern world, the communication process in the Delphi method is done by telephone, video conference, and/or e-mail. The method of using email as the main source of communication between the facilitator and the participants is known as ‘‘e-Delphi’’ (Lindqvist & Nordanger 2008). 2

DELPHI PROCESS

In the present study, ‘‘e-Delphi’’ was employed to assess the ITD-PCC specifications which play a major role in the selection of materials, construction of structures, and also in maintaining quality and durability to the structures. A group of fifty Subject Matter Experts (SMEs) were identified among the ITD engineers, contractors, concrete suppliers, and scientists. Of the fifty identified SMEs, twenty-two volunteered to participate in the study. Fifty percent of the study group consisted of ITD employees with the other half representing suppliers, contractors, and scientists. The Delphi study participants were provided with three documents that explained the method of study, the strategy, a summary of ITD-PCC specifications, and a comparison of ITD specifications for AlkaliSilica reaction (ASR) with those of five other states with similar problems. Members of the study group were informed that the outcome of this study would benefit all parties involved and may improve the quality and durability of concrete, while reducing the cost of concrete production. In this study, a total of two rounds were sufficient to gauge the degree of consensus among the participants and understand the issues involving ITD-PCC specifications from the ITD and non-ITD perspectives. 3 3.1

RESULTS Round one

In round one of this study, four questions were posed to generate the initial discussions among the participants. The participants were provided with the ITDPCC specifications and were asked to respond to these questions in relation to the following two queries: 1. What is the risk of the material failing to meet ITDPCC specification?

2. What is the consequence of that material failing to meet ITD-PCC specification? The four questions were as follows: 1. Based on your knowledge, experience, and/or observations, and in reference to the summary of the ITD-PCC specifications for Portland cement concrete, is there any ITD-PCC specification that does not closely relate to the field performance and needs modification(s)? Please list the specification, provide a brief comment about it, and make suggestions for improvement, if you have any. 2. In some of the ITD-PCC specifications, there are limitations set forth for material acceptance including aggregate, cement, supplementary cementations material (SCM), and concrete as a final product. Some examples of the limitations include: maximum alkali in cement 0.6%, loss of less than 12% in sodium sulfate soundness test, maximum CaO in fly ash 11%, maximum expansion of 0.1% in an ASTM C1260 test, and many other limitations as shown in the ITD-PCC specification summary. What does each of these limitations mean to you? In your opinion, is there any correlation between these specifications and field performance of concrete? 3. In some areas of Idaho (mainly along the Snake River basin), the aggregate is known to be reactive and to promote Alkali-Silica reaction (ASR), resulting in premature deterioration of concrete. Please make comments on the ITD-PCC specifications in dealing with ASR problems, and in your opinion, what works the best in mitigating ASR in concrete, considering fly ash, Silica fume, lithium solution, combination and/or other Supplementary Cementations Material (SCM) used for mitigation? Are there any other methods to mitigate ASR in concrete? In your opinion, are there any other factor(s) contributing to the premature deterioration of PCC and what is the remedy for prevention? 4. In the ITD-PCC specifications, there are two tables depicting ITD specifications for the concrete mix recipe with and without fly ash. Please make any comments you may have regarding ITD mix recipes. Based on the collection and evaluation of all responses in the first round, nine summary comments were formulated. The summary comments are as follows: Comment 1: There were multiple recommendations to replace the ITD’s prescriptive based approach to concrete mix design with a performance based approach or a combination of prescriptive and performance based approaches when durability is a prime concern. Currently, ITD uses a recipe type approach, in which one recipe fits for all aggregate sources and concrete used in different locations. Suppliers/ contractors are often more knowledgeable regarding their aggregates than the specification authors. The

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suppliers use their years of experience in designing mixes with a specified performance and durability. As such, allowing the suppliers/contractors to design the mix would improve the quality of concrete while reducing the cost of production. Changing ITD’s prescriptive specifications for concrete to performance specifications or to a combination of performance/prescriptive specifications eliminates some of the tests required for approval, leading to cost reductions without compromising the quality and durability of concrete. Comment 2: Slump tests do not directly relate to field performance. However, there is a need to measure consistency of concrete. There were suggestions that it would be more appropriate to measure unit weight of concrete as a measure of consistency. This could be done during an air content test (AASHTO 152/ASTM C231) by weighing the material in the air pot and calculating the unit weight. Comment 3: Several participants indicated that the current ITD aggregate gradation specifications need to be reviewed and modified. There were calls for designing a denser concrete. Comment 4: There were comments about the use of advanced methods to better control the quality of concrete during production and placement. These advanced techniques include: a) a maturity test using a thermometer, which records time and temperature for freshly mixed concrete and b) an air void analysis. Comment 5: Currently, ITD specifies the maximum slump for concrete used in different projects. If the slump does not meet the specifications, generally suppliers add water to control workability. This could have an adverse effect on the concrete’s performance. ITD should employ additives to control concrete workability. Comment 6: Currently ITD concrete specifications require 660 lbs of cement per yard of concrete for 4,500 psi compressive strength. This is excessive. Excessive cement generates more heat of hydration promoting shrinkage and cracking. This impacts the stability of concrete by providing more paste, and it also promotes ASR by increasing the alkali content in the concrete. In addition, ITD requires a maximum water-cement ratio of 0.44 and maximum slump of 2 in. for concrete used in pavement projects. Comment 7: AASHTO M 295 and ASTM C 618 limit the maximum percentage of Loss On Ignition (LOI) in fly ash to 5% and 6% respectively. The ITD current specifications limit LOI in fly ash to a maximum of 1.5%. Comment 8: Currently, ITD requires an AASHTO T 303 (ASTM C1260) test for identifying potentially deleterious aggregate. Comment 9: A participant noted that a minimum 4% limit for air content could be inadequate to provide freeze/thaw durability for some mixes. For mixes with large nominal maximum aggregate size (1.5 in. or

larger), 4% minimum air content might be adequate, but for mixes with 1 in. or smaller nominal maximum size, the aggregate would require more paste, and 4% minimum air content might be inadequate. ACI 301 in table 4.2.2.4 recommends the use of 6% +/− 1.5% for severe freeze/thaw environments such as those that exist in Idaho.

3.2

Round two

For specificity, the nine summary comments were subdivided into 16 statements/questions. These were sent to the participants for the second round of the study. The scale used for these questions is listed below (Linestone & Turloff 2002). – Certain – low risk of being wrong – decision based upon this will not be wrong because of this ‘‘fact’’ – most inferences drawn from this will be true – Reliable – some risk of being wrong – willing to make a decision based on this but recognizing some chance of error – some incorrect inferences can be drawn – Risky – substantial risk of being wrong – not willing to make a decision based on this alone – many incorrect inferences can be drawn – Unreliable – great risk of being wrong – of no use as a decision maker The round two statements were: Statement 1: Performance based mix design reduces the risk of failure to provide satisfactory performance. Statement 2: A combination of performance/ prescriptive based approaches reduces the risk of failure to provide satisfactory performance. Statement 3: Performance based mix design reduces the cost of concrete production. Statement 4: A combination of performance/ prescriptive based approach reduces the cost of concrete production. Statement 5: What is the chance of failure for concrete to meet the specified performance and durability criteria if ITD adopts a performance based approach to concrete mix design and eliminates some of the required material acceptance tests, thereby allowing the suppliers/contractors to draw on their knowledge and experience in designing a mix recipe to meet the specified performance and durability criteria.

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Statement 6: What is the chance of failure for concrete to meet the specified performance and durability criteria if ITD adopts a combination of performance/prescriptive based approach to concrete mix design and eliminates some of the required material acceptance tests, thereby allowing the suppliers/contractors to draw on their knowledge and experience in designing a mix recipe to meet the specified performance and durability criteria. Statement 7: Using the unit weight of freshly mixed concrete as a measure of concrete consistency is more reliable than using a slump test for the same purpose. Statement 8: (Question) How would you rate the current ITD Aggregate Gradation Specifications? Statement 9: An improved aggregate gradation improves the workability and stability of concrete. Statement 10: Adopting the advanced techniques (named in Comment 4) to measure early strength, heat of hydration, consistency, and air content of concrete provides more reliable information about the quality of concrete. Statement 11: The use of super plasticizers and additives to control workability of concrete would improve the quality of its performance. Statement 12: Reducing cement content of concrete, requiring maximum water/cement, and controlling workability by improving aggregate gradation and the use of super plasticizers and admixtures would improve the quality and performance of concrete. Statement 13: Reducing cement content of concrete, requiring maximum water/cement, and controlling workability by improving aggregate gradation and the use of super plasticizers and admixtures reduces the cost of concrete production. Statement 14: Increasing the LOI limit to a maximum of 5% would not significantly impact the quality of concrete performance.

Table 1.

Figure 1.

Graph of the responses of round two.

Statement 15: (Question) what is the risk of having ASR problems for an aggregate passing the AASHTO T 303 (ASTM C1260) test? Statement 16: Requiring an ASTM C 1293 test in addition to AASHTO T 303 (ASTM C1260) to identify reactive aggregates would reduce the risk of having ASR problems. 4

CONCLUSIONS

Based on the survey results, the participants favor a combinatin of the performance/prescriptive based approach to PCC mix design. The majority of the participants felt that ITD needs to modify its aggregate gradation, and that there is a need to reduce the amount of Portland cement content in ITD’s concrete specifications. In addition, the participants felt that ITD needs to specify a maximum water/cement and use admixtures to control the workability of the concrete. Also, the participant recommended the use of advanced techniques to better control the quality of the concrete during production and the placement.

Summary of responses of round two.

Question no

Certain

Reliable

Risky

Unreliable

1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16

0.00 33.33 6.67 13.33 0.00 13.33 13.33 0.00 14.29 21.43 13.33 30.77 7.14 7.14 0.00 26.67

73.33 53.33 60.00 60.00 46.67 46.67 26.67 64.29 57.14 71.43 73.33 53.85 35.71 50.00 57.14 33.33

20 13.33 33.33 20.00 53.33 40.00 40.00 35.71 21.43 0.00 13.33 15.38 57.14 35.71 28.57 33.33

6.67 0.00 0.00 6.67 0.00 0.00 20.00 0.00 7.14 7.14 0.00 0.00 0.00 7.14 14.29 6.67

REFERENCES Dalkey, N.C. and Helmer, O. 1963. An Experimental Applica tion of the Delphi Method to the Use of Experts. Management Science 9 (3): 458–467. Linestone, H. and Turloff, M. 2002. The Delphi Method: Techniques and Applications. London, UK: AddisonWesley. Lindqvist, P. and Nordanger, U.K. 2008. (Mis-?) using the E-Delphi method. An Attempt to Articulate Practical Knowledge of Teaching. Scientific Journal Intl, Journal of Research Methods & Methodological Issues, V. 2, Issue 1.

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Effects of bamboo material on strength characteristics of calcium-based mortar H. Kawamura, K. Hashimoto & A. Shimabukuro Tokuyama College of Technology, Syunan, Japan

ABSTRACT: In this study, the calcium-based mortar using two kinds of bamboo materials as the admixture material, bamboo chips and thin strips of bamboo, are proposed to find effective utilization of bamboo. The effect of bamboo materials is investigated by getting the strength characteristic for the specimens of calcium-based mortar. As a result, the strength of calcium-based mortar mixed bamboo chips was higher than the strength of calcium-based mortar without bamboo materials. However, for calcium-based mortar to mix thin strips of bamboo, we could not find out the strength increase. Additionally, it became clear that the mortar specimen mixed bamboo materials had ductile fracture. 1

INTRODUCTION

In the construction industry of the world, concrete has become indispensable material now. Concrete has affected the modernization in various countries and the industrial recovery. However, cement which composes concrete, has the disadvantage in environment. Because cement has a strong alkalinity, there is a bad character for the natural world in no small way and the cement-based concrete which becomes the wastes is not recycled easily. On the other hand, the bamboo influences the environment now too. Recently, the bamboos forest was neglected and its area was spreading. As a result, the bamboo forest invades the man-made forest of the Japanese cedar and the cypress. Therefore, in the construction industry, the effective usage of the bamboo material has become of major interest lately. With the above mentioned points as background, the calcium-based solidification material and the bamboo material are taken up to consider the preservation of natural environment in this study. Because the calcium-based solidification material is lower alkalinity than cement, the usage of it is ecofriendly for animals and plants. In addition, when the calcium-based solidification material is disused, it may not have some adverse effect on environment. Therefore, in this study, calcium-based solidification material and bamboo material are examined to evaluate the effectiveness as the construction materials. 2

CALCIUM-BASED SOLIDIFICATION MATERIAL

The characteristics of the calcium-based solidification material are shown the following (Japan Conservation Engineers Co 2002).

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1. Natural soil is the base, and because of the brown based color, the impression obtained through a material harmonizes with environment. 2. The efflorescence phenomenon which is the fault of the cement based solidification materials, scarcely occur. 3. Because a main raw material is natural soil, it is no pollution. 4. When it weathers and deteriorates, the recycling as the reproduction spraying material is possible. 5. Because pH in the installation is approximately neutral, it makes easy the invasion of plants such as the moss.

3

ON BAMBOO

A growth power of the bamboo is strong, and the underground stem expands horizontally as if it trails near the surface of the earth. The bamboo is flexible in comparison with wood, it is tough and hard. The inside of bamboo becomes a hollow. Because bamboo can split thin in the fiber direction, it is easy to process it. With the characteristic of such a bamboo, the usage of bamboo were considered for building material, furniture, cooking utensils, a musical instrument, a tea ceremony tool set, a toy and so on. Bamboo has been made use for a living. However, in late years, because of the price decrease of the domestic bamboo shoot, the decrease of the demand for bamboo product, and the age advance of the producer, bamboo forest is neglected. As a result, an area of bamboo forest spreads, and various problems are taking place. These problems include the following. 1. Bamboo spreads through the farmland and takes nourishment of the soil such as fields.

2. Bamboo expands in tree planting, takes nourishment mutually for a conifer and kills a cedar and a cypress. 3. The tree which is shorter than bamboo becomes the duration lack of sunshine. 4. When bamboo forest is neglected, an underground stem is killed, and the landslide is generated. Therefore, the maintenance of bamboo forest and the effective utilization of bamboo become important.

4 4.1

Thin strips of bamboo.

Figure 2.

Bamboo chips.

Mixture

In this study, we must consider the test-piece to examine strength characteristic of calcium-based mortar which is mixed bamboo material. However, there is not a rule for the design of the mixture in calciumbased mortar. Therefore, in this study, the mixture calculation of calcium-based mortar is mixed bamboo material, was conducted based on a design method for the mixture of cement mortar (Japan Society of Civil Engineers 2007).The materials density which is necessary for a mixture calculation are 3.11 g/cm3 for calcium-based solidification materials, 2.64 g/cm3 for Toyoura standard sand, and 0.6 g/cm3 for thin strips of bamboo and bamboo chips. Here, the mixture ratio of the mortar using the weight ratio are 2 for calciumbased solidification materials, 5.8 for standard sand, and 0.2 for bamboo with the standard of 1 for water. The required amount is shown in Table 1. In addition, thin strips of bamboo and the bamboo chips which we used in this study, are shown in Figure 1 and Figure 2. 4.2

Figure 1.

STUDY SUMMARY

Compression test

The compression test is performed using cylindrical specimen. This examination is prescribed for ‘‘Test Methods for Compressive Strength of Mortar or Cement Paste Using Circular Column Test Pieces (JSCE-G 505)’’ and ‘‘Methods to Make Circular Column Test Pieces for Compressive Strength of Mortar or Cement Paste (JSCE-G 506)’’ (Japan Society of Civil Engineers 2002). The cylindrical specimen for Table 1.

Without bamboo With bamboo

Ca [kg/l]

S [kg/l]

0,255

0.511

1.532

0.240

0.479

1.390

σc =

P π(d/2)2

(1)

where, σc : compressive strength (MPa), P: maximum load (N), d: specimen diameter (mm).

Mix proportion of calcium-based mortar. W [kg/l]

test must be set between an upper and a bottom loading platen without the eccentric loading. Using this table of mixture proportion, six cylindrical specimens which are necessary for a compressive strength experiment and a tensile strength experiment are prepared, and three bending specimens are prepared. After the water curing of four weeks, we conducted the compressive strength experiments, the bending strength experiments, and the tensile strength experiments. The cylindrical specimen for test is loaded using loading platen so that the increase of the compressive displacement becomes 0.5 × 10−2 mm/min. Compressive strength σc was calculated by using an expression (1).

B [kg/l]

0.048

where W = water; Ca = calcium-based solidification materials; S = Toyoura standard sand; B = bamboo.

4.3

Tensile test

Three cylindrical specimens are prepared for each mixture to conduct the tensile strength experiments based on ‘‘Method of test for splitting tensile strength of concrete (JIS A 1113)’’ Japan Society of Civil Engineers 2002). All specimens are loaded in loading rate 0.5 × 10−2 mm/min.

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The bank of the test-piece and compression side for pressurization is cleaned. And the test-piece is installed so that a joint part of the mould does not contact with a pressurization version. Tensile strength σt is calculated in expression 2. 2P π d

tension strength(MPa)

σt =

2

(2)

where, σt : tensile strength (MPa), P: maximum load (N), d: specimen diameter (mm2 ), : length of the specimen(mm).

1.5

1

0.5 tension strength average 0 without bamboo

Figure 3.

4.4 Bending strength

The results of compressive test.

compressive strength (MPa)

15

10

5 compressive strength average 0 without bamboo

(3)

Figure 4.

where, σb : bending strength (MPa), P: maximum load (N), L: span length (mm), a: the length of sides for square section (mm).

bamboo chips

thin strips of bamboo

The results of tensile test.

5 bending strength (MPa)

5

thin strips of bamboo

20

Three specimens for each mixture are prepared to conduct the bending strength experiments based on the strength test of cement in ‘‘Physical testing methods for cement (JIS R 5201)’’ (Japan Society of Civil Engineers 2007). The loading method is 3-point bending. The loading span is 100 mm and the loading rate is 0.1 mm/min. The bending strength σb was calculated in an expression (3). 3 PL σb = 2 a3

bamboo chips

EXPERIMENT RESULT AND CONSIDERATION

5.1 Influence of bamboo material on each strength

4 3 2 1

bending strength average

0

A compressive, tensile and bending strength test results of calcium-based mortar which is mixed bamboo material are shown in Figure 3, Figure 5 respectively. In Figure 3, it is shown that the compressive strength becomes high in the case using bamboo chips. However, compressive strength of specimens mixed thin strips of bamboo becomes low. As for this result, the thin strips of bamboo is stick state(cf. Figure 1), but the bamboo chips is fiber state(cf. Figure 2). Therefore, it is considered that the force of adhesive for the bamboo chips to mortar is smaller than the one for thin strips of bamboo mortar. Figure 4 is the tensile test result. From this figure, by using bamboo chips, tensile strength becomes high. However, the tensile strength of specimens used thin strips of bamboo hardly change in comparison with the strength of specimen which is not used bamboo material. It is considered that bamboo chips combine effectively with mortar of low tensile strength. Figure 5 is the result of bending test. There is no significant difference between the bending strength of

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without bamboo

Figure 5.

bamboo chips

thin strips of bamboo

The results of bending test.

specimens without bamboo material and the bending strength of specimens mixed bamboo chips. In the case of tensile strength and compressive strength, it was able to estimate the strength increase by bamboo chips mixture. However, a different tendency is showed for the bending strength. In other words, the increase of the bending strength by the mixture of the bamboo chips cannot be anticipated. In addition, the compressive strength and the bending strength fall by mixture of the thin strips of bamboo. 5.2

The relations between compressive stress and longitudinal strain

The typical relations between the compressive stress and the longitudinal strain for each mixture are shown

compression stress (MPa)

20 23.9GPa 15

1

10

5

0 0

500

1000 1500 2000 logitudinal strain (× 10-6)

2500

3000

Figure 6. The relations between compressive stress and longitudinal strain for calcium-based mortar without bamboo material.

20 compression stress (MPa)

in Figure 6∼Figure 8. The numerical value in a figure is the elastic modulus. This modulus is defined by slope of the range with the longitudinal strain, 0∼100 × 10−6 . Figure 6 is the compressive stress—the longitudinal strain curve of calcium-based mortar without bamboo material. For the specimen without bamboo material, a lot of longitudinal strain does not generate in initial stage of loading. The early stress-strain curve of the specimen without bamboo material has the steep slope (23.9 GPa) in comparison with the initial slope for the specimen mixed the bamboo material (Figure 7, Figure 8). It shows non-linear behavior in the neighborhood of the peak stress. In addition, its fracture behavior is brittle. Figure 7 is the compressive stress—the longitudinal strain curve of calcium-based mortar mixed bamboo chips. The early stress-strain curve of the specimen using the bamboo chips has the gentle slope (20.4 GPa) in comparison with the initial slope for the specimen without the bamboo material. From this, it is recognized that the mortar material mixed bamboo chips has

1 10

0

compression stress (MPa)

1 10

5

0 0

500

1000 1500 2000 -6 logitudinal strain (×10 )

2500

3000

Figure 7. The relations between compressive stress and longitudinal strain for calcium-based mortar mixed bamboo chips.

500

1000 1500 2000 -6 logitudinal strain (× 10 )

2500

3000

Figure 8. The relations between compressive stress and longitudinal strain for calcium-based mortar mixed thin strips of bamboo.

the property easy to deform. In addition, the relation between the compressive stress and the longitudinal strain in the neighborhood of the peak stress shows non-linear behavior. As for the post-failure, it is known that the relation between the compressive stress and the longitudinal strain has the gentle slope and its behavior is ductile. This behavior is influenced by the fiber of the bamboo. Figure 8 is the compressive stress—the longitudinal strain curve of calcium-based mortar mixed thin strips of bamboo. In a similar behavior to the specimen using the bamboo chips, the early stress-strain curve has the gentle slope (14.6 GPa). Because the value of the initial elastic modulus is smaller than one of bamboo chips mortar, it is known that this specimen has the property with the greatest deformation. With the above mentioned points, it is recognized that calcium-based mortar shows the ductile behavior by mixing bamboo after the peak stress. 5.3

20.4GPa

5

0

20

15

14.6GPa

15

Relations between compressive stress and volumetric strain

Dilatancy is considered to be the microscopic fracture occurring previous to the macroscopic fractute (Paterson 1986). Therefore, it is considered that the microscopic fracture is occurring in the specimen if an origin of the dilatancy is obtained. Figure 9 shows the relations of compressive stress—volumetric strain. The dilatancy is defined as the inelastic increase of the volume (Paterson 1986). It can be considered that the origin of the dilatancy is the point where linear slope changed to nonlinear slope (the numerical value in the figure). In Figure 9, the stress on the origin of the dilatancy for the specimen mixed bamboo chips shows the greatest value. Moreover, the stress on the origin of the dilatancy for the specimen mixed thin strips of bamboo shows the smallest value.

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the smallest value for the specimen mixed thin strips of bamboo. This tendency is similar to the evaluation using volumetric strain. However, the stress on an origin of the dilatancy using volumetric strain shows the greater value than AE evaluation stress. Therefore, for the consideration on the microscopic fracture of calcium-based mortar, the evaluation by the origin of the dilatancy using volumetric strain is more effective than the evaluation using AE count rate.

20

compression stress (MPa)

13.5(MPa) bamboo chips 15 10.2(MPa) without bamboo 10

5

6.4(MPa) thin strips of bamboo

0 -3000 -2500 -2000 -1500 -1000 -500

0

500

1000

volumetric strain (×10-6)

6

Figure 9. The relations between compressive stress and volumetric strain for calcium-based mortar.

compression stress (MPa)

20 14.1(MPa) bamboo chips

15 13.1(MPa) without bamboo

10 9.1(MPa) thin strips of bamboo

5

0 0

50

100

150 200 250 AE (/sec)

300

350

400

Figure 10. The relations between compressive stress and AE for calcium-based mortar.

5.4

Relations between compressive stress and count rate of acoustic emission

As another technique to examine microscopic fracture, the method to evaluate by acoustic emission is used in addition to the above-mentioned dilatancy evaluation. The relation between the compressive stress and the count rate of acoustic emission (AE count rate) for the specimen without bamboo, the specimen mixed bamboo chips and the specimen mixed thin strips of bamboo, are shown in Figure 10. Figure 10 is associated deeply with Figure 9. The numerical value in the figure is the stress of the points where the curvature of the relation between the compressive stress and AE count rate (AE evaluation stress) begins to change. AE evaluation stress is defined as the stress which microscopic fracture start. This is similar to the origin of the dilatancy. For all experiments, acoustic emission was observed from the initial stage of the loading. In addition, all experiments have the same tendency that the great number of AE count rate is observed in the neighborhood of the peak stress. AE evaluation stress has the greatest value for the specimen mixed bamboo chips, the second value for the specimen without bamboo and

455

CONCLUSIONS

In this study, calcium-based mortar using calciumbased solidification materials was treated. The influence of the bamboo on the strength characteristics was investigated. Moreover, the availability of the bamboo was examined. As a result, the following became clear. 1. For the compressive strength and the tensile strength, the strength increase is able to be anticipated by mixing bamboo chips. 2. For the bending strength of calcium-based mortar, it is considered that the bamboo material does not have an influence on the strength very much. 3. By the relations of compressive stress—longitudinal strain, calcium-based mortar mixed bamboo material is easy to deform in comparison with calcium-based mortar without bamboo. In addition, calcium-based mortar has the ductile property by mixing bamboo. 4. By the relations of compressive stress—volumetric strain and the relations of compressive stress—AE count rate, It is known that the stress that is generated the microscopic fracture gives the greatest value for the specimen mixed bamboo chip. Calcium-based mortar mixed thin strips of bamboo dose not show the effect of the strength increase. 5. Because the stress on an origin of the dilatancy using volumetric strain shows the greater value than AE evaluation stress, the evaluation by the origin of the dilatancy using volumetric strain is more effective than the evaluation using AE count rate.

REFERENCES Japan Conservation Engineers Co., Ltd, 2002. Proposal method by calcium-based solidification material. Japan. Japan Society of Civil Engineers 2002. Standard Specification for Concrete Structures-2002, Test Methods and Specifications, Tokyo: Japan Society of Civil Engineers. Japan Society of Civil Engineers. 2007. Guideline for Experiment on Materials of Civil Works, 2007 Edition, Tokyo: Japan Society of Civil Engineers. Paterson, S.M. 1986. Experimental Rock Deformation—The Brittle Field, Tokyo: Kokon Shoten.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Effects of remediation and hauling on the air void stability of self-consolidating concrete N. Ghafoori & M. Barfield University of Nevada, Las Vegas, USA

ABSTRACT: Self-consolidating concrete (SCC) is a highly flowable construction material that commonly experiences high slump flow loss due to long hauling times. The main objective of this study was to compare two different techniques of remediating slump flow loss, overdosing and retempering, and their effect on the air void characteristics of SCC. One 635 mm (25 in.) slump flow mixture was subjected to hauling times between 20 and 90 minutes and subsequently remediated back to the target slump flow. The air void characteristics (specific surface and spacing factor) were measured on the fresh concrete using an Air Void Analyzer. The two forms of remediation investigated were able to produce SCC mixtures that met the target air void characteristics for adequate freeze-thaw durability and target fresh properties required for self-consolidating concrete. Both forms of remediation necessitated an increased dosage of high range water reducing admixture with increasing hauling time. 1

INTRODUCTION

Self-consolidating concrete (SCC) is a highperformance building material that has the ability to flow and consolidate under its own weight. SCC is different from conventional concrete in that it does not require mechanical consolidation (such as vibration) due to its high flow ability, typically contains a higher percentage of fine materials, and utilizes admixtures such as a high range water reducer (HRWR) and/or a viscosity modifying admixture (VMA). The departure from conventional concrete in mixture design and fresh flow properties results in mixtures that are highly sensitive to changes in the environment. Field applications of self-consolidating concrete frequently necessitate long hauling times between batching and placing, sometimes up to 90 minutes. Hauling time shall be defined as the elapsed time from the first water and cement contact to the final placement or testing of the concrete. The effects of hauling have been studied extensively, and have been shown to cause slump loss, or a reduction in fluidity and workability of concrete with time (Plante et al. 1989; Khayat & Assaad 2002). Slump loss is mainly caused by the chemical hydration of cement and the physical coagulation of cement particles (Hattori & Izumi 1998). In addition to slump loss, past research indicates that the total air content of a conventional concrete mixture typically decreases by 1–2% with hauling time, but can increase in high-slump conventional concrete mixtures (Kosmatka et al. 2002). Air-entrained concrete, which is produced by incorporating an air-entraining admixture (AEA), must

457

exhibit certain air void characteristics in order to have durability in a freezing environment. The air voids must be sufficiently small, or have a specific surface, α, of at least 25 mm2 /mm3 (635 in2 /in3 ), and be spaced closely to allow the water within the concrete to expand when freezing, which correlates to a spacing factor, L, of less than 200 μm (0.0079 in.) (Powers 1965). Air void stability refers to the resistance to decrease in specific surface and increase in spacing factor with hauling time and/or remediation. The effects of hauling time such as slump loss can be counteracted using various methods of remediation. For this investigation, two common remediation techniques were utilized, both of which involved incorporating more or less admixtures at a specified time. The first remediation technique was to overdose or under-dose the initial admixture amounts to achieve the desired fresh properties at the end of a stipulated hauling time. This shall henceforth be referred to as overdosing or remediation A. The second form of remediation employed in this study involved retempering a SCC mixture after it had been hauled (immediately before placement). The initial SCC mixture design used for retempering exhibited certain fresh properties (i.e. slump flow, air content, etc.) at 10 minutes, and then after agitating for a predetermined hauling time, more admixtures were added to achieve the same fresh properties. This shall be referred to as retempering or remediation B. Retempering with additional HRWR has been shown to damage the air void characteristics of conventional and self-consolidating concrete, but usually does not decrease the air content (Plante et al. 1989;

390 kg/m3 (657 lb/yd3 ) cement, 78 kg/m3 (131 lb/yd3 ) fly ash, and 196 kg/m3 (330 lb/yd3 ) water. The admixtures utilized included a high range water reducer (HRWR), a viscosity modifying admixture (VMA) and an air-entraining admixture (AEA). Admixtures were obtained from one admixture manufacturer commonly available in the United States. A polycarboxylate-acid HRWR, naphthalene sulfonate and welan gum VMA, and tall oil and glycol ether AEA was utilized for this investigation.

Khayat & Assaad 2002). The addition of a HRWR is normally linked with the increased fluidity of concrete. Thus, air void coalescence is facilitated, which has been found to degrade the air void characteristics (Plante et al. 1989). 2

RESEARCH SIGNIFICANCE

This research is important for the production of selfconsolidating concrete in cold regions. Repeated frost cycles can cause permanent damage and deterioration to concrete foundations, structures and roadways. The high flow ability of SCC can result in an unstable air void system—one which ultimately does not provide adequate freezing and thawing durability for the hardened concrete. Remediation of SCC mixtures is likely in practical applications, and knowledge of its effect on air void stability is important. 3

3.2

EXPERIMENTAL PROCEDURE

In order to study the effects of the two forms of remediation, one 635 mm (25 in.) slump flow concrete mixture was developed utilizing one admixture manufacturer. In completing remediation A, the concrete admixtures were overdosed or under-dosed to meet the target fresh properties at hauling times of 20, 30, 40, 50, 60, 70, 80 and 90 minutes after first cement and water contact. For remediation B, the self-consolidating concrete was retempered with admixtures after 20, 40, 60 and 80 minutes of hauling time to meet the target fresh properties. 3.1

Raw materials

The aggregates were obtained from a local quarry in Southern Nevada. The coarse aggregate had a nominal maximum size of 16 mm and was required to pass the #7 gradation limits defined by ASTM C 33. The fine aggregate also met ASTM C 33 gradation requirements and had a fineness modulus of 3.0. The coarse-to-fine aggregate ratio was set at 1.083, determined using the optimum volumetric density of the combined aggregate gradation. Fifty-two percent, or 864 kg/m3 , of the total aggregate was coarse, and the remaining 48% was fine aggregate, or 795 kg/m3 . ASTM C 150 Type V cement and ASTM C 618 Class F fly ash were used due to the high occurrence of sulfates in the region. Class F fly ash was added at 20% by weight of cement, following typical regional practices, and in order to provide the trial self-consolidating concretes with sufficient cementitious materials. The water-to-cementitious-materials ratio remained fixed at 0.40, which is the maximum value allowed by ACI 318-05 for concrete exposed to a severe frost environment. All mixtures contained

Mixing and testing procedures

The concrete was produced in a horizontal pan mixer with 0.0283 m3 capacity. For the mixing sequence, first the coarse aggregate, 1/3 of the mixing water and the AEA was added. Following two minutes of mixing, the fine aggregate and 1/3 of the water was incorporated and mixed for another two minutes. Finally, the cement, fly ash and remaining water was added. After mixing for three minutes, the HRWR and VMA was introduced and allowed to mix for an additional three minutes. At this point, the concrete was rested and mixed for two minutes each. The elapsed time of the total mixing sequence was 14 minutes, or 10 minutes following the first cement and water contact. After the mixing sequence, the mixer was switched to the agitation speed of 7.25 rpm until the desired hauling time was met (20, 30, 40, 50, 60, 70, 80 or 90 minutes after first cement and water contact). For remediation A, the mixing speed was increased to 14.5 rpm for the final minute of hauling time and then tested. For remediation B, supplemental admixtures were incorporated and mixed at 14.5 rpm for three additional minutes prior to testing. Four fresh properties of the SCC were tested to assess flow performance: 1) unconfined workability, 2) rate of flow ability, 3) passing ability, and 4) resistance to dynamic segregation. The unconfined workability was tested by measuring the slump flow (described in ASTM C 1611) and was required to be within 27 mm (1 in.) of the target slump flow. The rate of flow ability was determined by measuring the T50 (also outlined in ASTM C 1611), which is the elapsed time from when the slump cone is lifted to when the concrete reaches a 50 cm (19.7 in.) mark on the testing plate. For this study, the T50 was required to be between 2 and 5 seconds. The passing ability was determined using the J-Ring, which is outlined in ASTM C 1621. The J-Ring measurement was done in conjunction with the slump flow, and was required to be within 51 mm (2 in.) of the slump flow to indicate adequate passing ability of a SCC mixture. The stability, or resistance to dynamic segregation, of the concretes was evaluated based on the Visual Stability Index (VSI), as delineated in ASTM C 1611. The VSI essentially rates the stability of a concrete mixture on a scale of 0 to 3 (highly stable to highly unstable),

458

based on the visual appearance of the slump flow. For this investigation, the VSI was required to be a 0 or 1 (highly stable or stable). The total air content, required to be 6 ± 1%, was determined volumetrically on the fresh concrete using a roll-a-meter in accordance with ASTM C 173. The air void characteristics were determined using an Air Void Analyzer (AVA), which determines the air void specific surface and spacing factor on fresh samples of concrete. The method by which the AVA determines the air void characteristics is based on buoyancy principles, in that large bubbles rise faster in water than small bubbles. A small sample of the concrete mortar is injected into a column of water, the bubbles are released from the mortar as they exist in the concrete, and the rate of rise of the air voids is measured by a balance. The results of the Air Void Analyzer are correlated to match those determined on hardened concrete using ASTM C 457 within a 95% confidence limit, and present an adequate assessment of the size and spacing of the air voids in concrete (Aarre 1998; Magura 1996).

Table 1. Remediation cementitious materials).

admixture

dosages

(ml/kg

Hauling time (min.)

HRWR

VMA

AEA

Remediation A 10 (635-H) 20 30 40 50 60 70 80 90

2.02 2.15 2.28 2.35 2.41 2.44 2.48 2.51 2.54

0.26 0.26 0.26 0.26 0.26 0.26 0.26 0.26 0.26

0.72 0.72 0.72 0.72 0.65 0.59 0.52 0.46 0.39

Remediation B 10 (635-H) 20 40 60 80

2.02 +0.13 +0.20 +0.33 +0.46

0.26 0 0 0 0

0.72 0 0 0 0

Note: ‘+’ indicates supplementary dosage added at specified hauling time.

4

RESULTS AND DISCUSSION Table 2. of SCC.

4.1 Effects of hauling time Prior to determining the effects of remediation, the effects of hauling time were investigated to quantify the change in fresh properties of the SCC mixture to which remediation would later be applied. The trial mixture developed exhibited a slump flow of 635 mm at 10 minutes after the initial cement and water contact. As such, the admixture proportions can be seen in Table 1 (under the hauling time of 10 minutes). This mixture shall be designated 635-H with the ‘‘H’’ indicating ‘‘hauled’’ for the 635 mm slump flow mixture. Hauling time did indeed significantly change the flow properties of the SCC mixtures to the extent that they were no longer self-consolidating, as seen in Table 2. Slump flow losses were measured at 37% after 90 minutes of hauling time. The T50 rate of flow ability decreased to the point where it could not be recorded, less than 50 cm. Throughout hauling, the volumetric air content increased by 51%. In conjunction, the air void characteristics improved by 21% and 41% for the specific surfaces and spacing factor, respectively, as seen in Table 2. The type of AEA utilized in this study and the dramatic slump flow loss contributed to the increasing air content and improving air void characteristics. The AEA utilized was a wood-derived acid salt containing tall oil, which has been shown to increase the air content with hauling time (Kosmatka et al. 2002). The decreased slump flow caused large pockets of air to become entrapped throughout the mixture when it was tested, contributing to the increased air content. When the fluidity decreased with hauling time, the AEA was

459

Effects of hauling time on the fresh properties 635-H

Hauling time (min.)

Slump flow (mm)

T50 (sec.)

Air content (%)

α (mm−1 )

L (μm)

10 20 30 40 50 60 70 80 90

646 610 591 572 551 502 483 438 416

2.01 2.25 2.41 2.45 2.65 3.99 − − −

6.30 6.50 6.75 7.25 7.75 8.00 8.25 9.00 9.50

37.0 41.8 43.9 45.3 46.2 46.5 47.0 47.0 44.8

145 113 106 104 101 91 86 84 85

Note: A T50 time could not be recorded for slump flows less than 500 mm.

able to maximize its potential and entrain more small air voids, as the increasingly viscous paste imparted a cushioning effect to increase air void stability (Du & Folliard 2005). 4.2

Effects of remediation

Determination of the admixture dosages for both forms of remediation was accomplished through trialand-error. For remediation A, the admixtures were overdosed or under-dosed initially to achieve the target fresh properties at hauling times of 20, 30, 40, 50, 60, 70, 80 and 90 minutes, as seen in Table 1.

For remediation B, mixture 635-H was subjected to hauling times of 20, 40, 60 and 80 minutes and retempered with supplementary admixtures, whose dosages can also be seen in Table 1. In Table 1, a ‘‘+’’ indicates the supplemental admixture dosage added at the specified hauling time. The intervals between retempering hauling times were 20 minutes, as opposed to the 10 minute intervals for remediation A, because the admixture dosages needed for retempering were too diminutive for practical application at smaller time intervals. 4.2.1 HRWR and VMA dosages The HRWR dosages necessary for both forms of remediation are significantly related to the magnitude of slump flow loss, as documented in Table 2. For remediation A, the HRWR was overdosed an average of 0.07 ml/kg per 10 minutes of hauling time. Similarly, remediation B necessitated comparable dosages of HRWR, although used 0.07 ml/kg less HRWR, on average, than remediation A. Remediation with VMA was not necessary at any hauling time.

Table 3.

Fresh properties of remediated mixtures.

Hauling time (min.)

Slump flow (mm)

J-Ring (mm)

T50 (sec.)

Air content (%)

Remediation A 10 645 20 610 30 648 40 654 50 648 60 641 70 648 80 622 90 622

610 591 598 654 616 616 629 597 584

1.99 2.47 2.28 2.22 2.41 2.78 2.10 2.29 3.18

6.30 6.75 6.50 7.00 6.75 6.50 6.50 6.25 6.00

Remediation B 10 645 20 641 40 648 60 660 80 648

610 625 616 629 616

1.99 2.54 1.53 1.44 1.59

6.30 6.80 7.30 7.80 8.30

50 70 Hauling Time (minutes)

90

11

4.2.3 Air content In contrast to losing slump flow with hauling time, the air content increased with hauling time, as seen in Figure 1. As a result, the AEA was under-dosed for the retempered mixture to counteract the natural tendency of the mixtures to generate air. After 40 minutes of hauling, the AEA was reduced by 0.07 ml/kg for every

Remediation A

10

Remediation B Air Content (%)

4.2.2 Flow properties The fresh properties of the mixture remediated during hauling time using remediation A can be seen in Table 3. The average passing ability (measured by the difference between the slump flow and J-Ring) increased from the initial mixing time by 10 mm. The remediated mixtures were highly stable, as indicated by the VSI, which was equal to zero for all hauling times. The agitation and decreased effectiveness of the HRWR with time generally increased the stability, homogeneity and resistance to bleeding of the mixtures. Mechanisms such as particle grinding, chemical hydration of cement and coagulation of particles occurred with the continual agitation, which increased the surface area (or fineness) of the particles. Although the fineness increased the viscosity of the mixture, the additional HRWR was able to maintain or restore the desired slump flow level. The fresh properties of the retempered mixtures can also be seen in Table 3. For the retempered mixtures, the passing ability obtained was adequate, but the flow of the mixtures was typically not viscous enough to attain a T50 time between 2 and 5 seconds. The HRWR increased the flow of the mixtures but decreased the stability to the point that, at 40 and 60 minutes, slight bleeding occurred on the surface of the SCC.

No Remediation

9 8 7 6 5 10

Figure 1. niques.

30

Air content comparison of remediation tech-

10 minute increment. At 90 minutes, the AEA dosages dropped to 0.39 ml/kg. It is evident that much of the increase in air content with hauling time was due to slump loss, which resulted in increased entrapped air. Retempering the trial mixture after hauling time yielded acceptable flow characteristics, but typically the air content could not be maintained at 6 ± 1%. The procedures set did not account for increasing air content with hauling time; therefore, the initial AEA admixture dosage had to be utilized to follow true retempering methodology. The slump flow was able to be retempered by adding supplementary doses of HRWR, but the air content could not be reduced by the simple addition of more admixtures. In the field, courses of action that eliminate air voids would

460

4.2.4 Air void characteristics The Air Void Analyzer results of remediation presented herein are the average of two samples taken from the same batch of SCC. The results previously presented for hauling time (in Table 2) are typically the average of five samples from two batches; therefore, the air void characteristics for remediation are not as precise, and generally contain more scattered data. However, the remediation results do indicate if a mixture departs significantly from the initial air void characteristics (measured at 10 minutes), or if the mixture is no longer within the acceptable range of air void characteristics needed for sufficient frost durability. The air void characteristics of the remediated mixtures can be seen in Figures 2 and 3. The air void systems measured for all mixtures remained within the acceptable range of specific surface >25 mm−1 and spacing factor 1920 kg/m3 ) have good strength and durability, but lower permeability. However, mixtures with low unit weights (1600 kg/m3 ) have excellent permeability, but

low strength and durability (Kevern 2006). Resulting in-situ density of PCPC is a function of the mixture proportions, workability of the mixture at the time of finishing, and the compaction/finishing methods. The narrow range of w/c and generally low water content, along with the more specialized admixtures often result in highly variable mixture workability and consistency. Many times pervious concrete is workable and relatively fluid at the beginning of discharge and then becomes stiffer as the load progresses. Tempering a mixture (adding water) is often performed mid-load to help improve workability and to allow complete material discharge. Unfortunately, tempering results in variable w/c, workability, density, and hardened concrete properties within a particular load and throughout the placement. Areas of low density generally have poor durability creating pockets of loose material and rutting. An important aspect to improving durability is a consistent product throughout a site.

2 2.1

TRADITIONAL CONSTRUCTION TECHNIQUES Pervious concrete placement

Typical fresh pervious concrete mixtures are stiff, no slump mixtures (Tennis et al. 2004). The concrete has little to no flow and must be discharged either directly from a ready-mixed concrete truck or using a belttype placer. The void space makes pumping PCPC impossible. Large areas are placed using strip construction of formed concrete sections. Typically half of the strips are placed and allowed to cure for 7-days before removing the forms and placing the alternate sections. Figure 1 shows discharge directly from two concrete trucks on-grade. Due to the stiffness of the mixture, the discharge chutes must be adjusted for maximum angle and are limited to one extension section. The limited

Figure 1.

Discharge from ready-mix truck.

Figure 2.

Belt placing equipment.

lateral flow requires concrete discharged from a truck close to the final location. When truck access to a site is limited, belt-type placing equipment can be used. Figure 2 shows a section being placed with belt-placer. Using placement technique and sufficiently workable mixtures, 31 m3 (40 cy) per hour can be placed of 150 mm (6-inch) thick pavement. Placing concrete with both methods involves piles of concrete raked level with the forms. The piles tend to have higher density due to the weight of material and force of discharge, while the material raked in-between has lower density. The variable mixture workability combined with the variability of manual placement results in unpredictable porosity across most placements.

2.2

Compaction and finishing

Once the pervious concrete has been placed on-grade, the fresh concrete is raked roughly level with the top of the forms. Two techniques are most common for compacting and finishing. One, short riser strips are placed on top of the forms and the concrete is roughly leveled. Then a truss-screed strikes off the concrete. The riser strips are removed and a weighted roller is used to compact the concrete to the final form height. The riser strips range from 13 to 25 mm (0.5 to 1.0 in.) and provide 8 to 17% additional compaction, localized near the surface (Haselbach and Freeman 2006). Figure 3 shows concrete being finished with a trussscreed and Figure 4 shows compaction with a weighed roller after removing the riser strips. Roller-screeds are also commonly used on pervious concrete to provide compaction and finishing in one operation. Fresh concrete is roughly leveled to 25 mm above the forms (without a riser strip). The roller-screed consists of a weighted steel tube rotating opposite to the direction of placement. Figure 5 shows roller-screed finishing operations. The roller-screed

474

Since the w/c is very low, immediate application (within 10 minutes of finishing) of the plastic is necessary to ensure proper curing and prevent drying of the surface cement paste. The plastic is left in place for at least 7-days. Especially using the weighted roller method, the concrete may be exposed for too long leading to excessive raveling.

3

Figure 3.

Truss-screed.

Figure 4.

Compaction with a weighted roller.

Figure 5.

Roller-screed finishing.

IMPROVED PERVIOUS CONCRETE CONSTRUCTION TECHNIQUES

Both of the common placement techniques involve manual labor to assist discharging the concrete, raking the concrete to a final location, and either pulling the roller-screed, or pushing the weighted roller. Consequently, new and modifications to existing equipment are improving pavement production rates and placement consistency. As early as 15 years ago, some PCPC placements were performed using existing mechanized equipment. Figure 6 shows the construction of a pervious concrete parking area with asphalt placing equipment. Further evolving the roller-screed placements are several available types of powered roller-screeds. The power-screed contains drive roller tubes that operate directly on the forms, propelling the leading rotating screed tube. The power-screed in Figure 7 requires only one operator, freeing labor for other tasks. Recent interest has also prompted the development of new equipment for placing pervious concrete. Optional equipment has been developed for laserguided screeds. New pervious concrete head attachments contain an auger and spinning tubes similar to bridge-deck paving equipment, along with a sprayer for applying a curing agent (Figure 8) (Somero 2009). Several types of slip-form pavers have also been developed to improve placement. Figure 9 shows optional equipment for standard curb and gutter machines.

provides some compaction while creating the final surface. Shortly after the concrete has been compacted and finished a mist of water or curing agent may be applied and then the concrete is covered with plastic sheeting.

475

Figure 6.

Placing with an asphalt paver.

Figure 7.

Power roller-screed.

Figure 9.

Figure 10.

Slip-form paver (Gomaco).

Slip-form paving (Evolution).

It has been observed that inadequately cured pervious concrete pavements have poor surface durability and excessive raveling of aggregate particles. To improve curing, surface-applied and internal curing techniques are being developed to prevent moisture loss. Soybean oil applied to fresh concrete before covering with plastic has shown to improve strength and surface durability (Kevern 2008b).

Figure 8.

Pervious concrete laser screed (Somero).

Vibrators in the hopper allow continuous flow of concrete, while a tamper bar and vibrating pan produce a smooth surface (Gomaco 2006). Another type of available slip-form paver is shown in Figure 10, a modified vibrating grout-box. In one particular project alone, labor cost saved by using the slipform paver paid for the more specialized equipment (Evolution 2008).

4

IMPROVED PERVIOUS CONCRETE MIXTURES

The objectives for improving pervious concrete mixtures are to: – Create a robust mixture which is consistently producible and with consistent fresh properties. – Create a highly workable mixture which improves production rate and reduces labor.

476

4.1

5

Mixture proportioning

The Portland Cement Association (PCA) is currently developing a standardized mixture proportion methodology for pervious concrete based on research performed throughout the U.S. and the global community. The new procedure recognizes that an unlimited number of possible mixture exist, but will provide guidance to produce quality mixtures using locally available materials and techniques. Since workability and compaction characteristics are related to the layer of cementitious paste separating the coarse aggregate particles, cement content is adjusted for the surface area calculated from the aggregate gradation and the aggregate shape. Angular particles have a higher surface area than rounded particles, requiring more paste to achieve the correct paste thickness. Fine aggregate has been shown to thicken the paste layer surrounding the coarse aggregate, creating greater particle contact area and increasing strength. The new procedure optimizes the combined aggregate gradation to ensure a correct balance between permeability and durability. The release of this new design procedure is scheduled for early 2009 and promises to allow all concrete producers the ability to make quality pervious concrete. 4.2

Admixtures

A NEW DIRECTION: SELF-CONSOLIDATING SLIPFORM PERVIOUS CONCRETE FOR OVERLAY APPLICATIONS

While pervious concrete has been used in Europe and Japan in surface-course applications for a number of years, wearing surface applications in the U.S. have been limited to a few low-volume roads. Recent interest in developing quieter/safer pavements by the Federal Highway Administration (FHWA) initiated a project to evaluate a pervious concrete wearing course overlay at Iowa State University. This unique challenge required high strength and high porosity for durability requirements, a highly workable mixture for consistent placement and uniform noise generation, and good bonding with the existing concrete substrate. A rigorous testing program was developed and involved testing over 1000 mixtures to determine the effect on workability, strength development, bonding, and durability of: – Aggregate type, – Aggregate gradation, – Cement content, – Cementitious makeup, – Water-to-cement ratio, – Fiber type and amount, and – Various combinations of admixtures.

Producing traditional concrete at w/c around 0.30 requires special admixtures and considerations, pervious concrete is no different. New polycarboxylate water reducing agents allow high workability at such low water contents. One issue inherent to high-range water reducing admixtures is limited useful workability. Improved pervious concrete mixtures utilize hydration stabilizing admixtures to improve admixture effectiveness and prevent any premature hydration. Viscosity modifying admixtures (VMAs) are used extensively in self-consolidating concrete (SCC) and provide moisture retention and lubrication when used in pervious concrete. European and Australian experience with wearing course pervious concretes have shown latex polymers improve workability as well as strength and durability (Beeldens et al. 1997; Vorobieff 2005). A few types of latexbased admixtures designed specifically for pervious concrete are now available in the U.S. and are particularly beneficial when used to improve pasteto-aggregate bonding or for pervious concrete repair. Several companies also have available admixture packages specifically for pervious concrete. These packages often contain a combination of water reducing, set controlling, air entraining, latex, and viscosity modifying admixtures. Like the rapidly advancing filed of SCC, admixtures are important for quality PCPC.

477

Since a specified porosity and unit weight can be achieved either by a highly workable mixture or by applying additional compaction energy, a workability method was developed using low-pressure gyratory compaction which characterizes both the initial self-consolidating ability and the resistance of the mixture to additional compaction. Mixtures were designed to possess high initial workability to facilitate rapid discharge and movement through the paving equipment, but also to have a high resistance to additional

Figure 11.

Laboratory paver.

research will allow standardized, high durability placements including both parking areas and roadways. REFERENCES

Figure 12.

Laboratory slip-form overlay.

compaction to maintain porosity and for edge-holding ability (Kevern et al. 2008c). The best performing mixtures contained a narrowlygraded granite coarse aggregate containing an additional 10% fine aggregate, 24% cementitious binder by aggregate mass, w/c of 0.29, cellulose and polypropylene fibers, a latex admixture, high-range water reducer, air entrainment, and a hydration stabilizing admixture. Figure 11 shows the laboratory-scale overlay paving device. The results of a paving trial are shown in Figure 12 with uniform compaction in the corners and good edge-holding ability. 6

CONCLUSIONS

There is wide-spread interest in pervious concrete to reduce and treat stormwater. Additional benefits include noise reduction and skid resistance for roadway applications. In the past pervious concrete placements have been manual labor intensive with mixtures and placement techniques that yielded highly variable performance. Pervious concrete construction is evolving with specialized equipment for rapid mechanized placement and better understanding of mixture proportioning to tailor mixture properties to placement techniques. The recent advances and continued

Bax, N., van Duerzen, A. and Molenaar, A. 2007. ‘‘New Technique for Rapid Construction and Rehabilitation of Concrete Pavements,’’ Proceedings of the International Conference on Optimizing Paving Concrete Mixtures and Accelerated Concrete Pavement Construction and Rehabilitation, Federal Highway Administration (FHWA), Atlanta, GA, pp. 283–293. Beeldens, A., Van Gemert, D., De Winne, E., Caestecker, C. and Van Messern, M. 1997. ‘‘Development of porous polymer cement concrete for highway pavements in Belgium,’’ Proceedings of the 2nd East Asia symposium on polymers in concrete EASPIC, Koriyama, pp. 121–129. Evolution Paving, Erickson, S. 2008. Email correspondence, 11/20/2008. Haselbach, L. and Freeman, R. 2006. ‘‘Vertical Porosity Distributions in Pervious Concrete Pavement,’’ ACI Materials Journal, Vol. 103, No. 6. Kevern, J.T., Schaefer, V.R., Wang, K. and Suleiman, M.T. 2008a. ‘‘Pervious Concrete Mixture Proportions for Improved Freeze-Thaw Durability,’’ Journal of. ASTM International, Vol. 5, No. 2. Kevern, J.T. 2008b. Advancement of Pervious Concrete Durability, Ph.D. Dissertation, Ames, IA: Iowa State University. Kevern, J.T., Wang, K. and Schaefer, V.R. 2008c. ‘‘SelfConsolidating Pervious Concrete.’’ Third North American Conference on the Design and Use of Self-Consolidating Concrete (SCC2008), Center for Advanced CementBased Materials at Northwestern University, CD-ROM. Kevern, J.T. 2006. Mix Design Determination for Freezethaw Resistant Portland Cement Pervious Concrete, Master’s Thesis, Ames, IA: Iowa State University. Gomaco Inc., Klein, K. 2006. Personal meeting. Somero Enterprises 2009. Pervious Concrete White Paper, available on-line at www.Somero.com, accessed 01-08-09. Tennis, P.D., Leming, M.L. and Akers, D.J. 2004. ‘‘Pervious Concrete Pavements.’’ EB302, Portland Cement Association, Skokie, Illinois, and National Ready Mixed Concrete Association, Silver Spring, Maryland. United States Government—Federal Register notice, 2004. Effluent Limitations Guidelines and New Source Performance Standards for the Construction and Development Category. Vol. 69, Number 80. Vorobieff, G. and Habir, E. 2005. ‘‘No Fines Concrete Research Project Overview of Stages 1 and 2.’’ Interim Report R04-A018 of the pavements section, technology and technical services branch of the Australian Operations and Services Directorate.

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Experimental research on regional confined concrete columns under compression X.M. Cao, J.C. Xiao & Z.H. Huang Guizhou University, Guiyang, P.R. China

T.J. Ren Guiyang Design Institute Co. Ltd, Guiyang, P.R. China

ABSTRACT: Regional confined concrete (RCC) is a newly developed concept (patent) based on the research of normal confined concrete (NCC). With special reinforcement arrangement, a RCC element is divided into several regions and each region has its own longitudinal reinforcements and stirrups that form a skeleton, together with concrete inside, works as an individual small column. Meanwhile, regions connected by means of stirrups and adjoining concrete work as a whole column. Six columns, which consist of one type of normal reinforcement configuration(NCC-1) and two type of regional confined configurations(RCC-1 and RCC-2), with concrete strength 27 and 36 MPa, 180 mm square and 1400 mm long were tested under monotonically increasing concentric and eccentric compression. As expected, columns failed with ductile manner, due to sufficient confinement, no matter how large of the eccentricity was. Mechanical property of RCC was discussed and an analysis model was proposed. Some design recommendations were also made. 1

INTRODUCTION

Typical reinforcement configuration of regional confined concrete (RCC) columns is shown in figure 1. The configuration of regional confined concrete tends to divide the column into several individual slender components (Figure 1, area I), and each component has its own longitudinal bars and stirrups. Components work together as an integral column by means of adjoining parts (Figure 1, area II). Based on the concept of ‘regional confinement’, several comparing experimental investigations have been conducted during the past years, which included beams under moment, beams under shear forces and short columns under concentric loads (Cao 2008). The variations in the experiments include reinforcement configurations, concrete strength and ratio of longitudinal reinforcement and stirrup. Test results show that regional confinement not only improves strength and ductility of element (the effectiveness of regional confinement is nearly as 150% as that of normal confinement) but changes failure mode which will help to promote the anti-seismic capacity. With test results, model of regional confined concrete was established. The mechanic property of regional confined concrete was also analyzed according to the reinforcement configuration and phenomena of failure. As part of the investigation, experimental research on regional confined concrete columns under

479

Figure 1. column.

Typical configuration of confined concrete

concentric and eccentric compression was carried out. Six specimens were tested in which three were under concentric compression and three under eccentric compression.

2

RESEARCH SIGNIFICANCE

As mentioned before, regional confined concrete has its own characteristic compared with normal confined concrete, so it is necessary to fully understand its property under each circumstance.

3

SPECIMENS

Three types of configurations were taken, which included two types of RCC, namely RCC-1 and RCC-2 and

one type of normal confined concrete, namely NCC-1 (as shown in Figure 2). With square cross section 180 mm × 180 mm, the specimens are 1400 mm high, each ends of specimens are strengthened to guarantee the failure regions are around the middle of specimens (as shown in Figure 3). Concrete strength of 27 MPa and 30 MPa at the date of tests was employed in the tests for concentric compression and eccentric compression respectively. The yielding stresses fy and ultimate stresses fu for reinforcement and stirrups are 446 MPa, 414 MPa and 611 MPa, 672 MPa respectively. Details of specimens are listed in the Table 1. In the case of concentric y

60

50

10 10 60

10 60 50

10 60

10

40

60

10

NCC-1

4.1 10 60

60

40

RCC-1

Concentric tests

10

180

180

180

Figure 2.

TEST PROCEDURE

180

180

60

60

10

10 60

180

10 10

4

Tests were carried out in structural laboratory of Guizhou University by using a 5000 KN compression machine (shown in Figure 4). The loading speed was 50 KN/min.

y

14

10

y

tests, the variations were diameter of bars and space of stirrups to ensure all specimens have relatively equal ratio of longitudinal reinforcement and stirrup, twelve diameter 12 mm bars for NCC-1 and sixteen diameter 10 mm bars for RCC-1 and RCC-2 were chosen, and stirrup spaces S were 50 mm, 55 mm and 60 mm for NCC column and RCC columns respectively. In the case of eccentric tests, the variations were longitudinal bar’s diameter and ratio of stirrups (as shown in Table 1). The eccentricity for eccentric tests is 60 mm.

RCC-2

Configuration of specimens.

In the beginning of tests, all the three columns kept elastic until section average strains were around 0.002. Then load kept increase steadily and covers spalled. RCC-1c and RCC-2c arrived their ultimate load carrying capacity when section average strains were around 0.0058 while NCC-1c strain was around 0.011 (as shown in Figure 5). No descending curves were drawn because some gages were ruined. Since it is difficult to apply axial load strictly, columns bent from the very beginning when loads applied. The relationships of load to lateral deformation were drawn in the Figure 6. 4.2

Eccentric tests

In the process of eccentric tests, different performances were observed with the change of reinforcement’s configuration. For RCC-1e, the longitudinal reinforcements on the tensile side were pulled in the beginning, but compressed after load arrived to a. Concentric column Figure 3. Table 1.

b. Eccentric column

Specimens. Details of specimens.

Specimen

Space of stirrup s (mm)

Concrete strength fc (MPa)

NCC-1c* RCC-1c RCC-2c NCC-1e* RCC-1e RCC-2e

50 60 55 50 50 50

27

36

Ratio of reinforcement ρb

Ratio of stirrups ρv

0.042 0.039 0.039 0.042 0.039 0.039

0.027 0.025 0.026 0.027 0.028 0.031

* c stands for concentric load and e stands for eccentric load.

a. Concentric column Figure 4.

480

Test setup.

b. Eccentric column

Figure 5.

Relationship of load-strain.

Figure 6.

Relationship of load-deformation.

Figure 7.

Relationship of load-strain.

around 700 KN while corresponding reinforcements of NCC-1e and RCC-2e kept tensile constantly (see Figure 7). The relationship of concrete strains on the side of compression to loads were linear around 0.0027∼0.004 and the strains were large than 0.01 corresponding to peak load. Additionally, all the specimens developed lateral deformations when loaded and arrived peak loads when the lateral deformations of columns were around 5 mm (see Figure 8). It was noticed that, after peak loads, the carrying capacity of NCC-1s degraded more gently than RCC

481

Figure 8.

Relationship of load-deformation.

Figure 9.

Failure phenomenon of concentric columns.

Figure 10.

Failure phenomenon of eccentric columns.

ones due to the larger volumes of longitudinal bars arranged on tensile side of columns and all specimens were failure in ductile manner no matter they were loaded concentrically or eccentrically (see Figures 9,10).

5 5.1

ANALYSIS AND DISCUSSION Confinement distributions

The load carrying capacities and failure modes of RCC members can be explained with the distribution of stresses of confinement. Generally, the confined stresses exacted on concrete along the sides of column produced by dilatation of

column under axial load are not equally distributed due to arch action. Simply dividing the cross sections into several different regions according to the quantity of stirrups surround this region (see Figure 11); here the contribution of longitudinal bars is not taken into account. Region Is are classified into well confinement regions. Assume the dilatation is equal over the section, then the stresses distribution along the side of column in direction y can be drown as in Figures 12 and 13, where σy stands for the stress against column produced by stirrup, subscript 1, 2 and 3 stand for stirrups no. 1, 2 and 3, and power N stands for NCC column and R for RCC column (Cao 2008). From the Figure 12 and Figure 13 it can be seen that the stress distribution of RCC-1 is well-distributed than others. The well-confined regions are in the centre of section for NCC-1 and in four corners for RCC-1 and RCC-2. In the figures, character T stands for the tension produced in the stirrups.

5.2 Concentric loaded columns For concentric loaded well-confined short columns, load carrying capacity can be dully developed. The load carrying capacity can be predicted by Equation 1 (Hu 2002): Pc = 0.9(As fy + Ac fcc )

III

II

I

I 180

II

(1)

I

NCC

Figure 11.

1

II

RCC-1

RCC-2

Classify of confinement regions. y

= N

y3

+

R

N

R

y

y

y

=

=

=

N

R

y2

+

R

y2

N

y2

R

+

+

y1

R

y1

y1

y

arch action

2

1

arch action

2

N

N

N

T1

T1

N

T2

NCC-1

Figure 13. section. Table 2.

x

1

1

T2

arch action

2

R

R R

T11+T2 TT

R 1

R R

T1+T2

RCC-1

R R

R

T11+T2 TT

R

R R

T1+T2

1

RCC-2

Confined stress distributions of middle part of

Load carrying capacity under concentric loads. Pcal

Ptest

Specimens

fc

As

fcc

KN

KN

Pcal − Ptest Pcal

NCC-1c RCC-1c RCC-2c

27 27 27

1356.5 1256 1256

36.92 36.13 36.28

1395.2 1336.5 1340

1380 1360 1200

0.0108 −0.018 0.104

Here strength of confined concrete √ fcc = (1 + 1.09λv 1 − s/D)fc f λv = ρv fyvc —characteristic factors of stirrup; S—space of stirrups; D—width of columns. Results of calculation and test are listed in the Table 2. Comparing Figures 5 and 6 with Table 2, it can be seen that even the ultimate carrying capacities of tests matched well with calculation, and the peak loads are almost equal, but the relationships of load and deformation are different. RCC-1 and RCC-2 are more rigid in the beginning. Since the NCC-1 had more longitudinal bars along the tensile side, it was more ductile than RCC-1 and RCC-2 after peak load. By comparing Pcal with Ptest , it can be seen that with Equation 1, the load carrying capacity of column RCC-2 was overestimated.

N

y2

y1

N

R

y1

y

1

arch action

2

1

2

arch action

1

II 3

x N

N

T2+T3

N

T1

N

T1

NCC-1

Figure 12. section.

Eccentric loaded columns

y2

arch action

2

5.3

R1

+

N

N

T2+T3

R

T1

R

T1

R

T1

RCC-1

R

T1

R

R

T2

T2

RCC-2

Confined stress distributions of upper side of

For eccentric load, the load carrying capacity can be predicated by using Equation 2-1 to 2-3; and the model of confined concrete is used as shown in Figure 14 (Cao 2008). Sections are classified with intensity of confinement, as shown in Figure 15, where b1 is well confined and b2 is weakly confined. Here, the ultimate strain εcu is taken as 0.007 for well confined regions,

482

σc

N

f cc

N

e

h0 x

well-confined

As

e

A's f y

s

As

A's f y

s

c

s

c

c cu

fc

x

0. 8x

weakly-confined

Strain

Figure 16. o

0.0027

Figure 15.

Table 3.

NCC-1e RCC-1e RCC-2e

b1 b2

RCC-2

Load carrying capacity under eccentric loads.

36 36 36

0.17 0.17 0.19

43.5 811 644 44.5 808 664 43.5 796 665

840 930 870

Ncal − Ntest Ncal −0.036 −0.151 −0.093

Sections classification.

For purpose of simplification, equivalent rectangular stress distribution is taken to replace the actual stress distribution. The height of rectangular stress block is taken as 0.8x for ultimate load (Figure 16). Then Equations 4-1 and 4-2 can be expressed as:

when εc < εc0  σcI = fcc

Strain and balance condition.

Fcc Ncal N0.003 Ntest fc Specimen MPa λv = ρv MPa KN KN KN

b1

b1 b2 b1

b2 b1

RCC-1

NCC-1

Equivalent

ε

Relationship of stress and strain relationship.

b2

Figure 14.

0.007

Compressive stress

 2  εc εc 2.226 − 1.226 εc0 εc0

(2-1)

when εc0 < εc < εcu0

N = As fy − As σs + 0.8x(b1 fcc + b2 fc )

σcI = fcc

Ne = As fy (h0 − as )

(2-2)

For weakly confined regions When gεc < εc0   2  εc εc σcII = fc 2.226 − 1.226 εc0 εc0

 x + (b1 fcc + b2 fc )0.8x h0 − 0.8 2

(3-1)

When εc0 < εc < εcu0 σcII = fc

(3-2)

For equilibrium in Figure 16 N = As fy − As σs +

x (b1 σc1 + b2 σc2 )dx (4-1) 0

Ne = As fy (h0 − as ) x  x (b1 σc + b2 σc2 )dx + h0 − 2

(4-2)

0

And for deformation harmony (GB50010-2002)   0.8h0 σs = εcu −1 E (4-3) x

(5-1)

(5-2)

Combined with Equation 4-3, load carrying capacities Ncal are calculated (see Table 3). For simplicity, no second order moments are taken into account. For peak loads, fu are used instead of fy due to relatively large deformation. Analysis results when εcu = 0.003 (as specified in ACI code) are also listed in Table 3 to demonstrate the efficiency of confinement. From the Table 3, it can be seen that the results of calculation by using Equations 4 and 5 are underestimated compared with test results. The values of Ntest and Ncal are much higher than N0.003 when the ultimate strains of 0.003 are taken. Furthermore, the regional confined columns are sturdier than normal confined columns. An interesting phenomenon was observed in the test of RCC-2e: different from the results of analysis, the longitudinal bars in tensile side were pulled in the beginning when loaded, but compressed with the increase of load (referenced to Figure 7), stresses in stirrups of criteria section were higher than that of relevant NCC-1e and RCC-1e. It indicated that the mechanical property of RCC-2 should be different from that of NCC-1and RCC-1 (the analysis should be done in elsewhere).

483

6

CONCLUSIONS

1. Regional confined concrete changes the way of confinement; it transfers the well-confined region from center of normal confined concrete to corners, and makes the confinement stresses welldistributed and more efficient. 2. Regional confined concrete are more useful to eccentric loaded columns due to the way of confinement. 3. Well confined concrete can develop great deformation, no matter what are normal confined or regional confined. 4. Both of the NCC columns and RCC columns failed with ductile manner, due to sufficient confinement, no matter how large of the eccentricity is. 5. Model of confined concrete and analysis approach of load carrying capacity are recommended, in which elastic strain 0.0027 and ultimate strain 0.007 are adopt, analysis results accordingly are reasonable. ACKNOWLEDGEMENTS The investigation described here was carried out at Guizhou University with support from the Science and

Technology Bureau of Guizhou Province. The authors would like to thank Luo Xiaoyi, Wang Xiaobin and Yang Xuerui for their help. REFERENCES Cao, X.M. & Yang, L.L. 2008. Mechanical property analysis of regional confined square concrete column under axial load. The Second International Forum on Advances in Structural Engineering, Oct. 10–11, Dalian, P.R. China. Hu, H.T. 2002. The calculation of load-carrying capacity of short confined concrete columns with composite stirrups. Building Structure 32:4,12,13 and 51. Beijing, PR. China. Cao, X.M. & Yang, X.R. 2008. Experimental research on regional confined concrete columns under eccentric load. Proceedings of the tenth national conference on concrete structure theory and application, Dalian, P.R. China. Building Code Requirements for Structural Concrete (ACI318-05) and Commentary (ACI 315R–05) Code for Design of Concrete Structures (GB 50010-2002), China, 2002, Beijing. Cao, X.M. 2008. ‘Mechanical Properties of Regional confined rectangular concrete columns under axial loads’. Journal of Chongqing Jianzhu University. 30: 3, 83–86, Chongqing, PR. China.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Experimental study on dry-shrinkage of lightweight cement mortar T. Watanabe Osaka City University, Osaka, Japan

A. Mori Nagoya Institute of Technology, Nagoya, Japan

ABSTRACT: This paper presents an examination of the influence of multiple coat application and one-side drying on shrinkage strain during the drying phase of lightweight cement mortar incorporating a calcium carbonate foam filler made from ethylene-vinyl acetate resin. The examination about lightweight cement mortar with four kinds of mortar flow was carried out under the standardized atmosphere during 28 days. The authors found that the application of multiple coats reduced shrinkage strain during drying and provided additional benefits to finish surface conditions in terms of warp. In addition, the movement characteristics of the mortar are considerably different from those of the mortar with ordinary aggregates. It is expected that this quantitative examination of shrinkage will provide useful background material for persons studying the external forces that cause cracking and for designs aimed at preventing debonding. 1

INTRODUCTION

A finish application of cement mortar is an essential part of wet finishing of walls. In order to prevent cracking and flaking from the finish layer of the wall, it is important to comprehend not just its material properties, such as its drying shrinkage strain and mechanical strength, but also working conditions such as the thickness of the applied layers and elapsed time between applications. The difficulties related to plastering and other tasks can be eased considerably by using lightweight materials. This has promoted the adoption of lightweight cement mortars for wet finishing, including those that incorporate foam building materials such as styrols and calcium carbonate. More progress will be necessary before these materials are trusted sufficiently to be widely adopted. Technical references must be prepared that set forth their strength, durability and other characteristics, and quality standards must be developed for assessing their tendency to crack and debonding. As a part of a research program on the adhesion between the concrete substrate and the finish plastering, and on prevention of cracking, one of the authors has carried out a series of quantitative experiments on length changes (drying related shrinkage) in ordinary cement mortars used in finish applications (Nakano et al. 2002, 2003). These experiments were designed to reproduce real conditions occurring during plastering. In addition to shrinkage and expansion due to one-sided drying, deformation behavior during warpage was measured and the influences of the components of the plastering and elapsed

485

time between applications were observed. The focus of this report is on lightweight cement mortar containing ethylene vinyl acetate resin-based calcium carbonate foam flakes used for finish plastering. The authors have assessed the influence of multiple plaster applications and one-sided drying on drying shrinkage strain by means of the same methods used in previous research on observing drying related shrinkage.

2 2.1

EXPERIMENTAL METHOD Materials used and mix proportions

Table 1 shows the materials used in this experiment. The material types and mix proportions utilized are standard in the material manufacturing industry for pre-mixed mortar supplies. They were mixed with the typical objectives of workability in plastering, determining setting time, adhesion and post-hardening toughness. Table 2 presents the physical characteristics of the cement used while Table 3 presents the distribution of sizes in the lightweight aggregate. The four mixes given in Table 4 were prepared. The additive proportions of the cement lightweight aggregate were held constant and the water flow value was set at 150, 160, 170 and 180. 2.2

Manufacture of test specimens

All test mixes were plastered in two layers, to ensure that the application conditions would match actual use conditions. The second layer was applied three days

Table 1.

Materials used.

Cement

160

Standard Portland cement

Lightweight aggregate Additives

Water

Upper: drying

Upper strain gage

8

Ethylene vinyl acetate resin/calcium carbonate foam flakes Vinylon fiber, paper pulp, thickening agent, dispersed material, emulsion, hardening accelerator, etc. *Types controlled by a manufacturer of pre-mixed cement Tap water

7 Lower strain gage

(Unit:mm)

Top layer Side: sealed Bottom: sealed

Bottom layer

Figure 1. Attachment of gages and sealing. Table 2.

Physical properties of ordinary Portland cement. Setting

Specific Initial surf. setting Density area Water time (h-m) (g/cm3 ) (cm2 /g) (%)

Final setting time (h-m)

28-day compressive strength (N/mm2 )

3.16

3–26

62.10

3350

27.70

2–23

Table 3. Grain size distribution of the lightweight aggregate. Main constituents

Bulk density

Grain size distribution

Ethylene vinyl acetate resin/ CaCO3 foam flakes

0.88 (g/cc)

all pass

1 mm

2 mm

2.8 mm

14%

49.4%

34.7%

2%

Table 4.

Mix proportions of light cement mortar.

Flow Cement Ltwt. Additives Water H2 O/Cement value (g) agg. (g) (g) (g) (%) 150 160 170 180

1112

75

56.7 362 (total 380 for all) 398 415

32.6 34.2 35.8 37.3

after the first in order to allow the lower layer sufficient time to dry. A frame (40 × 40 × 160 mm) was used as the mold. The bottom of the mold was fabricated out of urethane blocks and a 7-mm-thick bottom layer of plaster was established. Three days later, the thickness of the underlying urethane blocks was adjusted and an 8 mm top layer of plaster was then applied to create a total plaster thickness of 15 mm. Six specimens were made for each mix, three of which were used for strain measurements and three of which were used for

measurements of mass changes. All the experiments were conducted in a room with a constant controlled ambient temperature of 20◦ C and a humidity level of 60%. The bottoms and sides of all specimens were covered with a polyethylene sheet the day after each application to ensure that evaporation took place only through the top surface. Polyester strain gages were attached to the upper and lower faces of the specimens, as shown in Figure 1. 2.3

Method for measuring drying shrinkage strain and mass change rate

The experiments were conducted in a room controlled to a constant ambient temperature of 20◦ C and humidity level of 60%. The strain measured on the day after the final application of plaster was employed as the standard for drying shrinkage strain measurements, and the specimens were allowed to age for a total of 28 days after the final application. The following equations were used to calculate the shrinkage and warpage of laminated specimens. Shrinkage is expressed in positive values of strain. Positive values for warpage indicate that the tips of the specimen bent upward toward the dry side, while negative values indicate relative downward motion of the tips. εa = (ε1 + ε2 )/2,

εb = (ε1 − ε2 )/2

(1)

where εa : Shrinkage of laminated specimens (×10−6 ), εb : Warpage (×10−6 ), ε1 : Strain at the upper face of the plaster (×10−6 ), ε2 : Strain at the lower face of the plaster (×10−6 ). The mass was measured using the same standard as the drying shrinkage strain. On the day after the final plastering application, the bottom and side faces of each specimen were sealed to restrict evaporation to the upper face, and the measured mass obtained at that time was used as the standard. The mass was then measured once per day over the following 28 days. The observed changes from the initial mass were expressed in terms of a percentage of the initial mass.

486

Drying shrinkage strain (10-6)

2000

Flow value 170-I

Least squares approximation

(a) Flow :150

1500

500 150-1 150-2 150-3

0

0

Measured value

7 14 21 Drying time (days)

28

0

7 14 21 Drying time (days)

28

Figure 2. Typical drying time-drying shrinkage strain curve (Laminated specimen).

Drying shrinkage strain (10-6)

2000

Drying time (log days)

EXPERIMENTAL RESULTS

(d) Flow :180

(c) Flow :170

1500 1000 500

180-1 180-2 180-3

170-1 170-2 170-3

0 -500 0

3.1

160-1 160-2 160-3

-500

= 489 l og (day) + 309, R 2 = 0.975

3

(b) Flow :160

1000

7 14 21 Drying time (days)

28

0

7 14 21 Drying time (days)

28

Figure 3. Relationships between drying time and drying shrinkage strain (Upper face).

Prediction of 28-day strain in interrupted measurements

3.2

Drying shrinkage strain (10-6)

2000 1500 1000

(b) Flow :160

(a) Flow :150

500 160-1 160-2 160-3

150-1 150-2 150-3

0 -500 0

7 14 21 Drying time (days)

28

0

7 14 21 Drying time (days)

28

2000

Drying shrinkage strain (10-6)

The strains in the specimens formed under flow values of 170 and 180 could only be measured through the 15th day due to an electrical problem in the experimental apparatus. As a result, the authors have attempted to estimate the drying shrinkage strain from the 15th through the 28th day of curing. The estimates obtained were based on earlier experiments conducted by the authors on long-term observations of drying shrinkage strain during which extruded cement specimens were observed during repeated cycles of drying and re-moistening. The drying shrinkage strain measured during this process showed high correlations with the logarithm of number of wet-dry cycles and the relationships could be approximated with a straight line. Figure 2 shows a typical estimate. Because these two records show a high correlations, this relationships was used to estimate the drying shrinkage strain in the present experiment, the results of which are considered below.

(d) Flow :180

1500 1000

(c) Flow :170

500 170-1 170-2 170-3

0

180-1 180-2 180-3

-500 0

7 14 21 Drying time (days)

28

0

7 14 21 Drying time (days)

28

Figure 4. Relationships between drying time and drying shrinkage strain (Lower face).

faces all showed greater compressive strains than those found on the upper face.

Drying shrinkage strain

Figure 3 shows the relationships between drying time and drying shrinkage strain at the upper face of the plaster for all four flows. The means of the three strain values at the age of 28 days were 836 × 10−6 , 928 × 10−6 , 1096 × 10−6 and 1362 × 10−6 in the samples with flow values of 150, 160, 170 and 180, respectively. The strain tended to grow with the increase of the flow number. Figure 4 shows how the strain on the lower face of the plaster varied with drying. As shown above, the following were the averages of the strains measured in the three samples each for flows of 150, 160, 170 and 180, respectively: 1855 × 10−6 , 1896 × 10−6 , 1641 × 10−6 , and 1498 × 10−6 . Except for the plaster applied with a flow value of 180, these lower

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3.3

Mass change fraction

Figure 5 shows how the mass change fraction varied with drying. There was a rapid reduction in the mass over the first several days immediately following application of the cement mortar in all of the flow value groups, after which the process reversed, with the specimens actually regaining mass. The specimens with the slowest reductions stopped losing mass after approximately seven days. This halt occurred earlier in the specimens with higher flow value. The reversal in mass loss probably resulted from the following mechanism. Once drying had progressed for seven

Mass change fraction (%)

0

Table 5.

160-1 160-2 160-3

150-1 150-2 150-3

-0.2

Measured physical parameters of specimens.

-0.4

(a) Flow :150

-0.6

Type

-0.8

(b) Flow :160

LightExpansion and weight contraction mortar (×10−6 )

-1 0

7 14 21 Drying time (days)

28

0

7 14 21 Drying time (days)

Parameter

28

Mass change fraction (%)

0 170-1 170-2 170-3

-0.2

180-1 180-2 180-3

Warpage (×10−6 )

-0.4 -0.6 -0.8

(c) Flow :170

Mass change fraction (%)

(d) Flow :180

-1 0

7 14 21 Drying time (days)

28

0

7 14 21 Drying time (days)

28

Ordinary Expansion and mortar contraction (×10−6 ) Warpage (×10−6 ) Mass change fraction (%)

Figure 5. Relationships between drying time and mass change fraction.

4 4.1

150 160 170 180 150 160 170 180 150 160 170 180 150

1031 1033 1004 1030 −399 −391 −231 −57 −0.782 −0.767 −0.589 −0.567 175

1346 1411 1369 1430 −510 −484 −273 −68 −0.722 −0.622 −0.343 −0.266 225

−27.5 31.8 −0.150 0.096

Shrinkage (10-6)

days, hydration reactions (involving absorption of atmospheric moisture by the cement) began to overcome the reduction in mass resulting from the drying process. This increase in mass following an initial decrease has also appeared in previous research conducted by the authors on ordinary cement mortars. Please note, however, that this is only a hypothesis, and awaits further investigation in detail.

Drying period Flow value 7 days 28 days

DISCUSSIONS Expansion, contraction and warp

Table 5 shows the means of the physical parameters measured on the 7th and 28th days. Figure 6 shows how the laminated specimens of the different flow values contracted with drying. No major differences in the length change at 28 days were found between the various flow values. The shrinkage was generally in the range 1300 ∼ 1500 × 10−6 . The time lapse that occurs between the application of the bottom layer and the addition of the top layer allows the bottom layer to be drier. Next, water transferred from the lower face of the top layer to the upper face of the drier bottom layer. These transfers of water changed the substantial water content of the bottom and top layers. The transfers could exert more influence than the differences in water content of the different mix proportions. That is why no significant differences were observed between the specimens with varying flow values. Figure 7 shows the relationships between drying time and warpage of the laminated specimens. At 28 days, the tips of the specimens made for all flow values had deformed in a convex manner. Ordinarily, one would expect the specimens to warp in a concave manner, i.e., the tips moved in the direction of the

Figure 6. Relationships between drying time and shrinkage of laminated specimens. Note: Ordinary (150): Ordinary cement mortar of flow value 150.

dry side. In our previous experiments with ordinary mortar, warpage was concave in single-layer plaster bodies. The following may describe why the reversal of the warp direction occurred. After it is laid, the bottom layer has some time to begin drying out before the top layer is applied on top of it. Thus, once the fresh top layer is applied, it is possible that some of the water from this fresh layer was drawn down and absorbed into the drier bottom layer. This causes a large reduction in the per-unit water volume of the top layer. The shrinkage strain of the top layer is then less than that of the bottom layer, resulting in the reversed, convex warpage of the top layer. The fact that the convex warpage of the top layer occurred in the initial drying stage, even though drying was only allowed to occur through the upper surface of the layer, suggests that there was a considerable discrepancy between the shrinkage strains of the two layers. When considering

488

Warpage (10-6)

proper use of lightweight mortars for wet finishing, and the development of new and unique methods for the application of these materials. Additionally, it will be essential to gather and analyze more experimental data under real environmental conditions in order to guarantee the resistance of such mortars against flaking and falling. The authors believe that would be users and designers must consider that, without such preparation, it is possible that lightweight mortars will never be considered sufficiently safe for use on exterior walls.

Figure 7. Relationships between drying time and warpage of laminated specimens.

only the influence of the warpage, if this is constrained by the bottom layer, it is possible that strains causing upward warpage would tend to suppress cracking better than those causing warpage toward the seal, i.e., toward the lower side. We found that specimens with greater flow values incurred reduced warpage. In high flow specimens, this has been attributed to the diminishing differences between the per-unit amounts of water in the layers due to water absorption by the bottom layer from the top layer. However, further research will be needed in the future to clarify these qualitative observations. 4.2

Comparison with ordinary mortar

The lower part of Table 6 shows the results of an experiment conducted with ordinary cement mortar. A comparison between lightweight mortar and ordinary cement mortar showed that the shrinkage in the lightweight mortar at 28 days was 1346 × 10−6 , while that in ordinary cement mortar was about 225 × 10−6 , and that the ordinary mix was about six times as resistant to drying shrinkage strain. Both strains shared the characteristic of a convex shape, but in the lightweight mortar, the strain was −510 × 10−6 , while in ordinary cement mortar, it was about −31.8 × 10−6 . This indicates that the lightweight mortar was nearly 16 times as flexible. In other words, all the strain types measured in the lightweight mortar were much larger than those observed in the ordinary cement mortar. This is a result of the differences in shrinkage and Young’s modulus in the aggregates of the two cement mortars. The main aggregate of the lightweight mortar was ethylene vinyl acetate and calcium carbonate foam flakes, while the ordinary cement mortar contained quartz sand. The risk of magnitude of the strain is partially related to a combination of the Young’s modulus of the two layers and the strength of the interfacial adhesion, but must cause the accidents and failures of many types in the constructions. Builders and/or designers should consider taking cautionary notes regarding the

5

CONCLUSIONS

This study was designed to examine some of the improvements made to plastering processes in recent years. The object was a lightweight cement mortar created using ethylene vinyl acetate resin foam flakes as a lightweight aggregate. Mortar was laid in two layers at an interval of three days at several levels of flow value. Strain resulting from shrinkage and the change in mass were measured during drying in a constanttemperature constant-humidity laboratory and compared to the corresponding parameters for standard cement mortar. The following results were obtained: 1. The drying shrinkage strain is greater in the underlying layer of the lightweight mortar than in outer layers, causing these mortars to warp in a convex manner, i.e. their tips displaced upward, toward the dry side. 2. Discrepancies in shrinkage of laminated specimens among cement mortars with differing flow values were not found. All mix proportions showed contraction of about 1400 × 10−6 after drying for 28 days, several times the shrinkage shown by standard cement mortars. 3. The mass change fraction tended to fall as the flow value of the original mix increased. More data must be gathered and analyzed regarding the performance of lightweight mortars on exterior surfaces under experimental conditions, and specialized uses for these materials must be developed. REFERENCES

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Nakano, T., Mori, A., Baba, A. 2002. An Experimental Study of Optimization on Progress Interval Time of Cement Mortar Rendering based on Moisture Movements, Journal of finishing technology, Japan Society for Finishing Technology, Vol. 9, No. 2, 1–5. Nakano, T., Mori, A., Baba, A. 2003. Effects of Rendering Construction of Cement Mortar on Drying Shrinkage, Journal of finishing technology, Japan Society for Finishing Technology, Vol. 10, No. 3, 1–6.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Flexural behavior of high strength stone dust concrete V. Bhikshma, R. Kishore & N.H.M. Raju Department of Civil Engineering, University College of Engineering, Osmania University, Hyderabad, India

ABSTRACT: The paucity of suitable river sand for use as fine aggregate, in construction applications, and the recent construction boom has led to a dramatic increase in the price. Additionally various government agencies have put restrictions on sand quarrying to conserve this diminishing natural resources. This has prompted many engineers to look for alternate materials that are cheaper while possessing similar characteristics. One such alternative is the use of stone dust—a byproduct of crushers. This study indicates that this material can be used as partial or full replacement of river sand as fine aggregate without altering the strength, workability or setting characteristics of concrete. It presents the results of tests on cubes (150 × 150 × 150 mm) and under reinforced concrete beams (150 × 230 × 1500 mm) in order to obtain the flexural behavior of under reinforced RC beams. A total of 24 cubes and 10 beams were tested for direct compression and flexure at 28 days. 1

1.1 Use of stone dust

INTRODUCTION

Due to the recent spurt in construction activity brought on by the current economic boom, the cost of construction has been increasing by up to 15% every year, a major factor for this escalation in costs is the price of raw materials like cement, steel, timber, aggregates etc. As conventional natural resources are being depleted, the costs of these materials are increasing. Current scientific data tells us that the plasticity and hardened state properties are affected greatly by the type of aggregate used. Aggregates make up a bulk (up to 80%) of the concrete mix, their properties are crucial to the properties of the concrete. Different types of aggregates that are commonly used are natural sands and gravels, crushed rocks and manufactured aggregates. Concrete is a heterogeneous mixture of cement, fine and coarse aggregates. Crushed stones (granite and basalt, for instance) of suitable sizes as coarse aggregate, the river sand as fine aggregate are adopted in conventional concrete. Although river sand is usually available in many areas of the country, it would be economical to use locally available materials to substitute river sand as fine aggregate in making mortars and concrete. Stone dust is one among such materials available in large quantities from crusher units. Diminishing natural sand resources have increased and the efforts to identify substitutes for natural sand as a constituent of Portland cement concrete. The use of crusher stone dust in making concrete and mortar by partial/full replacement of natural river sand not only provides economy in the cost of construction but at the same time solves the problem of disposal of stone dust.

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Stone dust is useful as fine aggregate in RCC and PCC, Masonry blocks construction, Road formation, Landscaping, Precast structural elements like RCC ventilators, Jallies, Rings for wells, Flower pots and Water storage tubs, Back fill material in reinforced earth work construction, Filter media in sand filters, and Lining of drains etc. 1.2 Aims and objectives of the present study To determine flexural behavior of high strength stone dust concrete. 1.3 Scope of the present studies Use of crusher dust in varying percentages as replacement (0%, 25%, 50%, 75%, 100%) of river sand in M40 grade concrete mix. A total of 30 cubes and 10 beams were tested for direct compression and flexure. 2

LITERATURE REVIEW

Katz & Baum (2006) reported that the fine aggregates (smaller than 4.75 mm, No. 4 mesh) play a very important role in controlling the properties of fresh concrete. They help to improve the cohesiveness of fresh concrete, improve its workability, and prevent segregation and bleeding. Kumar et al. investigated the flexural behaviour of high-performance reinforced concrete beams using sandstone aggregates (2006). Hudson reported that,

‘‘concrete manufactured with a high percentage of minus 75 micron material will yield a more cohesive mix then concrete made with typical natural sand’’ (1999). The experiments conducted by Mishra at PWD Research Institution, Lucknow, on use of stone dust in cement mortars explains the influence of shape and size of fine aggregate on strength of mortars (1984). The experiments conducted by Dr. D.S. Prakash Rao and V. Giridhar Kumar (2004) on strength characteristics of concrete with stone dust as fine aggregate, draws the following conclusions-The concrete cubes with crusher dust developed about 17 % higher strength in compression, 7% more split tensile strength and 20% more flexural strength (Modulus of rupture) than the concrete cubes/beams with river sand as fine aggregate. Similarly, the RC beams with crusher dust sustained about 6% more load under two point loading and developed smaller deflections and smaller strains than the beams with river sand.

3

Table 2.

Details of the test beams.

Grade of Fine concrete aggregate M40

Designation Reinforcement of beam details

River sand(RS) SD0 RS + SD SD25 RS + SD SD50 RS + SD SD75 Stone dust(SD) SD100

Table 3.

2–6 mm φ at top 4–12 mm φ at bottom 2–legged 6 mm φ stirrups @ 150 mm

Physical properties of cement.

Physical property

Value

Normal consistency Initial setting time Final setting time Specific gravity Fineness Bulk density of cement

32% 84 minutes 300 minutes 3.149 2% 1.35 gms/cc

EXPERIMENTAL PROGRAMME Table 4. Physical properties of fine and coarse aggregate.

3.1

Materials

Ordinary Portland cement (Ultra tech cement) of 53 grade confirming to Bureau of Indian Standard is used in the present study. The stone dust is procured from locally available crushers located at Kokapet Hyderabad, Andhra Pradesh, India. The stone dust is tested for various properties like specific gravity, bulk density etc, in accordance with IS 2386-1968.Machine crushed angular granite metal from single source from locally available crusher located at Kokapet is used as coarse aggregate. Super plasticizer by trade name Conplast SP-337 manufactured at Bangalore was used as water reducing agent to achieve the required workability. It is available in brown liquid instantly dispensable in water. The test result values are presented in Tables 1–4. 3.2

Physical property

Value

Specific gravity of fine aggregate Specific gravity of fine aggregate stone dust Specific gravity of course aggregate Bulk density of fine aggregate Bulk density of course aggregate Fineness modulus of fine aggregate stone dust Fineness modulus of fine aggregate Fineness modulus of coarse aggregate

2.63

Table 5.

2.71 2.72 6.93

Slump and compaction factor values. M40 grade concrete

Casting procedure

Using the properties of aggregate and cement, concrete mix M40 grade of concrete for the total of five Table 1.

2.65 2.84 1824 kg/cum 1450 kg/cum

Mix

Slump, mm

Compaction factor

SD0 SD25 SD50 SD75 SD100

50 46 42 39 32

0.80 0.78 0.75 0.72 0.69

Mix proportions of constituent materials.

Mix M40 grade

Fine Coarse WaterStone aggre- aggre- cement Sp Cement dust gate gate ratio (ml)

SD0 SD25 SD50 SD75 SD100

1 1 1 1 1

0 0.208 0.417 0.627 0.835

0.835 0.627 0.417 0.208 0

2.6 2.6 2.6 2.6 2.6

0.4 0.4 0.4 0.4 0.4

350 350 350 350 350

mixes i.e. SD0, SD25, SD50, SD75, SD100 were tested,. The percentage replacements of fine aggregate with stone dust is from 0% to 100% (i.e. 0%, 25%, 50%, 75%, 100%) is done, and corresponding five mixes were prepared and casting is done in accordance with IS 10262-1982 and SP 23-1982. Workability and concrete compressive strength values are presented in Tables 5–6.

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Table 6.

Table 8. Test results of ultimate moment, curvature and deflections.

Strength of M40 grade concrete. Compressive strength (MPa)

Mix

SD %

7 days

28 days

SD0 SD25 SD50 SD75 SD100

0 25 50 75 100

29.25 30.58 32.58 37.44 35.87

43.67 45.0 47.85 51.22 53.33

Stone dust

Moment (kNm)

Curvature ×10−3

Deflection (mm)

SD0 SD25 SD50 SD75 SD100

56 57.2 58.4 60.8 63.2

27.0 25.16 20.12 15.57 9.89

9 8.2 7.10 6.90 6.2

Table 7. Ultimate load carrying capacity and moment carrying capacity of beams. Mix proportion

Ultimate load (Kn)

Ultimate moment (kNm)

SD0 SD25 SD50 SD75 SD100

140 142 146 152 158

56.0 57.2 58.4 60.8 63.2

3.3

Tests on reinforced concrete beams

Investigations on reinforced concrete beams with M40 grade and testing the beams for flexure to study the deflection and strain parameters at one third and mid span for two points loading have been planned. Beams are of size 150 × 230 mm(breadth × depth) in cross section and of effective span 1.20 meters for different proportions of sand and crusher dust. They were provided with lime wash and labeled. The beams cast with stone dust as fine aggregate were designated as SD0, SD25, SD50, SD75, SD100, respectively. The beams were tested on a Universal Testing Machine (2000 kN capacity) under two point loading at middle third points of the span. Dial gauge readings were recorded for every incremental load of 2 kN initially and 10 kN subsequently distributed equally over two points. Dial gauges at top measure compressive strains and those at bottom measure tensile strains. Strains and deflections were recorded with the help of the dial gauges for each increment of 2 kN initially and 10 kN subsequently load distributed over two points. Both compressive and tensile strains were measured at top and bottom fibers, respectively and the mean values were calculated. Moment–Curvature relations were developed with variation of load at one–third span and mid span for different proportions of river sand and stone dust. Test results are presented in Tables 7–8.

Figure 1. Compressive strength variation with percentage replacement of stone dust for M40 grade concrete.

Figure 2. Increase in ultimate load with percentage replacement of stone dust for M40 grade concrete.

of river sand and stone dust. The elastic deflections at initial stage of loading are found small for different proportions of river sand and stone dust when compared to river sand. 3.5

3.4

Load-deflection characteristics

Figures show the Load-deflection curves at mid span section for R.C.C beams with different proportions

493

Moment-curvature relationship

Figures show the moment-curvature relations at one third span and mid span for sand and crusher dust, respectively. Both the curves for sand as well as crusher

4

RESULTS AND DISCUSSION

From the test results, physical properties of fine aggregate, Fineness modulii for river sand and stone dust were found to be 2.72 and 2.71 respectively. It indicates that crusher dust contains high percent finer particles compared to river sand. The sieve analyses on samples of fine aggregate shows that river sand and crusher dust belongs to same grading zone in many cases. In the experimental investigation, it was found that the mortar and concrete prepared with stone dust were relatively less workable than those prepared with river sand. Figure 3. Relationship between voad vs. deflection for different percentages of stone dust.

Figure 4. Relationships between moments vs. curvature for different percentages of stone dust.

dust will follow a bi-linear relation with change in slope at the moment of first crack. The curvature varies linearly with moment in all stages of loading, also in the plastic state except change of slope that occurs at the moment of first crack and the moment at first yield of reinforcement. 3.6

Cracking pattern and crack widths

The crack pattern in R.C.C beams with sand is shown in Figures and for beams with crusher dust. The cracking pattern is more distinctive and clear for reinforced concrete with crusher dust compared to conventional concrete with river sand. The maximum crack width for beams 1.5 mm with crusher dust appeared at mid span. The flexural shear cracks began at mid span at a minimum load of 2.5 kN and propagated vertically up to the neutral axis. Web shear cracks began from middle-third span and propagated diagonally towards the centre of span up to the neutral axis. The average crack width is 0.17 mm and the maximum deflection was occurred at mid span at ultimate crushing load of 60 kN.

5

CONCLUSIONS

– The stone dust as replacement for natural sand enhances the strength of concrete mix. The rough profile of stone dust provides good interlocking and bond between ultra fine particles of cement paste. – The concrete is less permeable and durable than conventional concrete with river sand. – The compressive strength increased significantly up to about 20 percent for concrete with crusher dust compared to conventional concrete. – Deflections were smaller in reinforced concrete beams with stone dust compared to conventional concrete beams. The flexural behavior of reinforced concrete beams with crusher dust was similar to that for conventional concrete at higher loads as well. – Reinforced concrete with crushed stone dust is cheaper by about 8 percent compared to conventional reinforced concrete. – The stone dust can be used as fine aggregate in reinforced concrete beams for strength as well as economy besides the reduced impact on environment. REFERENCES Katz, A. & Baum, H. (2006). ACI Materials Journal, Nov/Dec. Effect of High levels of Fines Content on Concrete Properties. Kumar, P.S., Mannan, M.A., Kurian, V.J. & Achuytha, H. (2006). Building and Environment, Volume 42, Issue 7, Investigation on the flexural behaviour of high-performance reinforced concrete beams using sandstone aggregates. Hudson, B. (1999). ‘Concrete Workability with High Fines Content Sands’, Quarry, V. 7, pp. 22–25. Misra, V.N. (1984). Use of stone dust from crushers in cement mortars, The Indian Concrete Journal, V. 58, No. 8. Prakash Rao, D.S. & Giridhar Kumar, V. (2004). ‘‘Investigations on concrete with stone crusher dust as fine aggregate’’, The Indian Concrete Journal, M.E thesis submitted to Osmania University, Hyderabad, India.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Hemp: Rediscovered raw building material F. Khestl VŠB—Technical University of Ostrava, Ostrava, Czech Republic

ABSTRACT: Hemp is also utilized in building industry, apart from other industrial branches, nowadays especially as raw material suitable for the manufacture of heat and sound insulation. Hemp utilization in structural cement-bonded particleboards is yet another option. The board is formed with cement matrix and filler material made of hemp scutch. Commercial use of hemp in various building applications can be often seen in the world but not with cement as the matrix. Produced materials are bonded mainly with lime therefore their function is often a filling one. This paper offers complete overview about hemp utilization in building industry and also the latest knowledge of about using hemp in cement-bonded particleboards for structural purposes. 1

INTRODUCTION

Building industry is a rapidly developing business nowadays. It utilizes modern materials with high requirements not only on physical-mechanical properties but also on physical-chemical as well as heattechnical, acoustic and other properties. Last but not least, it is demanded that these materials, and engineering construction made from them, meet high aesthetic requirements too. Today, we have a lot of experience with production of these cement-bonded particleboards; nevertheless, like on every material and product, there are also higher and specific requirements. An interesting option how to modify building materials, or to market new materials, is to apply fast-renewable raw materials. Their use yields technical as well as economic and environmental benefits. These fast-renewable raw materials also include technical hemp that represents an excellent progressive raw material for the manufacture of building materials. 2 2.1

HEMP General information

Hemp comes from Asia but it is grown all around the world presently. From the botanical point of view, hemp is a member of the most developed plant family in the world. This plant can be grown under any climatic and soil conditions on Earth, including infertile soils. Hemp is far the most effective renewable natural source. Hemp has become notoriously known, especially so during the last century, due to its contents of psychotropic substances—THC. For this reason, growing hemp has been banned or very strictly limited in many countries. In fact, the species is versatile and has a wide variety of possible uses. Its seeds, stalks, flowers and oils can be used in wide range

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of industry, like agriculture, automotive industry, cosmetics, building industry, furniture, paper, textile, food, recycling, etc. It is excellent alternative to wooden material for building material; moreover, in some aspects its qualities even surpass the wood characteristics. Generally, three basic variants of hemp are distinguished: – Cannabis Indica—the plants are low (up to 1.2 m) and very dense; their stalk is not so fibrous; they are heavily bifurcated and their foliage is very dense. The content of psychotropic substances is very high. – Cannabis Ruderalis—wild hemp: The plants are very low (0,5 m); they have very thin and slightly fibrous stalks; the foliage is not very dense, the leaves are relatively large; the content of psychotropic substances is low to medium. – Cannabis Sativa—technical hemp: mostly dioecious annual plant, Cannabaceae family, grows very high (up to 4.0 m). The plants are rarely bifurcated and they have thin foliage with the leaves forking to 3 to 9 thin sprouts 7 to 15 cm long. The stalk is 3 to 30 mm thick and it has 13.5 to 19.5 per cent of fibers in the bast layer that increase the stalk strength. The stalk is soft and succulent during the first growth stage and it lignifies from the bottom later. The stalks are firm and their inner bark is fibrous. The content of psychotropic substances is very low (up to 0.2 per cent) and they are not frequently present. Three forms of Cannabis Sativa are distinguished: – Northern: 0.6 to 0.8 m tall on average, mature within 60–70 days. The yield of stalks and seed is low; the seeds are tiny. It is not grown in the Czech Republic. – Southern (vegetative type): 3–4 m tall. Matures in 130 to 180 days. The yield of stalks is high; the yield of seed is low. The fibers are long and fine.

– Transitional type: 1.50–2.50 m tall. It combines the properties of the previous two forms. It matures in 90–120 days. The yield of fibers as well as seed is good.

hempen materials and new progressive hemp building materials development have now very good chance to make a break through on the building materials market and be more visible.

2.2

2.3

History of hemp

As early as the 7th to 8th centuries, hemp fiber was utilized, which ranks it among the first woven fibers in the world. Even today it is considered one of the most quality fibers ever. Hemp was brought to Europe in about the 3rd millennium B.C. and it became one of the most frequently used agricultural crops for a long time in the course of several centuries (the end of the 19th century). For centuries and even millennia, hemp has been used in the manufacture of all kinds of textiles, sailcloth, ropes as well as paper products. Practically of paints, varnishes and oil for lightning were manufactured using hemp seed oil. Later hemp became to be replaced with cereal crops but prior that, it seed (containing a high amount of plant proteins) was one of the main sources of human sustenance. However, the hemp variants containing high amount of THC became to be frequently abused during the 20th century, which has gradually led to overall ban of growing hemp in the majority of countries in the world. It seems that various bans and restrictions have erased hemp from peoples’ minds, especially so in the building industry. Cultivation, following harvesting, processing and utilization of hemp in the last century had not only problems with law restrictions. Hemp plant couldn’t compete to the price with mass production of hollow fibers and flax fibers. The biggest problem was, in particularly, hemp harvesting and post-processing automation. Hand processing of hemp was very arduous which increased more and more price of final hemp products. Development and production of appropriate machinery required a lot of money. Not enough funding led to gradual falloff in hemp cultivation even in countries that had hemp growing not strictly prohibited, as was the case with the then Czechoslovakia. These facts have led to unsystematic and very sporadic utilization of this progressive plant. More intensive development of hemp cultivation for technical purposes in the Czech Republic falls into the years after 1st January 1999 when Act no. 167/98 on addictive substances came to force. This act governs poppy and hemp growing. Section 24a of the act bans growing species and variants of hemp (Cannabis family) that may contain more than 0.3 per cent of THC group substances (tetrahydrocanabinols) and, at the same time, Section 29 requires reporting duty for persons growing hemp on total areas larger than 100 m2 . The raise of modern techniques and innovations and principally modern thinking about environment at the end of 20th century opened up new ways of hemp as a building raw material comeback. Classic

Hemp cultivation

Hemp is a quality and fast-renewable raw material that has been known for thousands of years and that has been, and still is, cultivated all over the world in almost any soil, even highly water soaked. Their root structure can even enrich and consolidate the soil. By specific character of stem structure; growth of hemp can pull from the soil of harmful substances such as heavy metals. In specific conditions can be hemp used for reclamation of old ecological burdens. Another advantage of hemp, with regard to production cost, is its growth rate—up to 50 cm per week, which is faster than all weed and therefore no pesticides are necessary. There is not so great need of dressing too, about 150 kg per hectare, which is very little. For the cultivation of hemp in the Czech Republic was registered in 1999 two following varieties: Juso—11, which is a monoecious cannabis originating in Ukraine, involving a length of 220 to 300 cm and Polish monoecious hemp variety called Beniko with stem from 250 to 300 cm of length. Both varieties belong to a group of cannabis transitional and south type. The varieties meet the criteria of the EU for content of THC for hemp cultivation as agricultural crop. Yield of dry matter is from 8.5 to 10.5 t.ha−1 ; fiber content in the stem 24–28%; the total possible yield fibers 2.1 to 3.0 t.ha−1 . According to (Tošovská & Buchtová 2008), in 2008 is registered in the Czech Republic only variety Beniko, but varieties of Bialobrzeskie from Poland, Monoica and Fibrol from Hungary are tested in state laboratories as well as other varieties. With newly approved varieties for the year 2008/09 there are more than 40 types which cultivation is permitted under the joint catalog of the EU. Presently, there is lack of hemp in Europe; the demand is becoming to exceed the supply. Current price of low-class and impure hemp is about 1,500 CZK.t−1 on the home markets, the price of high-quality scotch is about 2,000 CZK.t−1 and it can reach up to 4,000 CZK.t−1 in most extreme cases. Such high price is caused by the lack of hemp on the markets and its unsystematic planting. It is only a matter of time that hemp will be included among commonly grown crops; farmers will have necessary machinery and hemp prices will stabilize. 2.4

Hemp in the building industry

Cannabis sativa is completely usable crop. The building industry use, in particular, the whole of its stem. Other parts of this crop are used in other industries; the roots may be used as biomass in incinerators. Leaves

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Figure 1. Hemp in the building industry (photo is borrowed from www.ekooko.cz).

and roots in most cases are leaved on the field; their decomposition is the natural complement of nutrients in the soil. Hemp stalk is used in the following forms: The whole stem—the stem as a whole, or chopped for longer parts, either, including the fiber or as peeled stem. (Fig. 1 (1)) Hemp fibers (Fig. 1 (2)) Processed hemp stem—scutch in various fractions (Fig. 1 (3) (4) and (5)) Hemp is very old plant, known for thousands of years, that has been, and still is, cultivated almost all over the world. Interesting properties predetermined this plant for countless number of household and industrial uses. From the beginnings, hemp crop is also well known as a good building raw material. In the past, people exploited its high stability, good physicalmechanical properties and its humidity and wetness good resistance in various ways of construction applications. Peeled hemp was used for roofing. The whole hempen stalks or chips were used especially as reinforcement of earthen structures and into abodes, i.e. raw earthen bricks. Hemp fibers, in the form of cords and ropes, were used for securing wooden scaffolds and other structures. Hemp fibers also served well as heat insulation, e.g. between wooden beams. Hemp is a practical, cheap building material that has excellent heat as well as sound insulating properties, it has satisfactory strength and excellent diffusion properties enabling optimal passage of humidity and ensuring optimal healthy climate in rooms. It has long durability, it is incombustible and it does not contain any proteins and so it is uneatable for rodents, termites and insects, which is an absolute protection against any attack by pests; it does not suffer from blight; it repels and drains water; it is light, durable, affordable; it grows very rapidly. The resulting product does not contain any substances harmful to the environment and it can be recycled very easily. It also has an ability to trap atmospheric heat and radiate it later so buildings made of hemp have very favorable parameters from the energy point of view. Hemp is also utilized in building industry, apart from other industrial branches, nowadays especially as raw material suitable for the manufacture of heat and sound insulation. These kinds of insulation can

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be in the form of both insulating boards and mats, mostly interconnected by pressure or with added polymeric or hydraulically active substances, or loose poured or sprayed insulating material. Similarly, thin unwoven textiles serve as excellent sound insulation between building partitions or to dampen steps under wooden or laminated floors.This plant is an excellent alternative of wood for building structures that surpasses wood in some respects. Wooden material from hemp stalks offers itself for the manufacture of compressed boards, but also to produce granular mixtures for plasters, floors or in the manufacture of shaped bricks for permanent shuttering. Hemp utilization in structural cement-bonded particleboards is yet another option. The board is formed with cement matrix and filler material made of hemp scutch. Commercial use of hemp in various building applications can be often seen in the world but not with cement as the matrix. Produced materials are bonded mainly with lime therefore their function is often a filling one. This article will deal with using hemp in cement-bonded particleboards for structural purposes.

3 3.1

HEMP UTILIZATION IN CEMENT BONDED PARTICLEBOARDS Hemp utilization in cement matrix

For hemp utilization as a filler to cement bonded particleboards is very important interaction with bonding component which is cement in this case. The hemp may not impact hydration of cement components which secures compactness of this composite. There is no unified method to prescribe hemp usability as a filler to cement bonded boards. There was made own method in this purpose. Methodology and testing results were described and published in (Khestl & Bydžovský 2007). The same samples that were used in (Khestl & Bydžovský 2007) were subjected to differential thermal analysis (DTA) and RTG analysis for final elimination of any possible harmful reactions of hemp and parts of cement matrix. Results from these test showed, that no harmful interactions in cement hydration were established. 3.2

Development of hemp based cement bonded particleboards

The development of hemp based cement bonded particleboards has undergone several related stages from early stages of laboratory experiments to machineproduced hemp based cement bonded particleboards. For the purpose of the construction boards production can be used stem chopped into scutch which is currently used for its calorific value, in particular, for the production of briquettes. As a filler of cement bonded particleboards are commonly used wooden

particles (Figs. 5–6). Modification of filler was made step by step by several fractions of hemp (type A to type D) that were obtained from different processing centers. A visual comparison of each hemp fraction and wood particles is shown in the following photographs. (Figs. 7–12). Observed results from previous tests indicate that we can compare test samples from

Figure 2.

each hemp fraction obtained from different processing centers. Visual comparisons of wood-cement particleboards with hemp-cement ones is graphically shown in the following photographs. The following photos and their descriptions are chronologically ordered; the first six photos present samples made during the early stages of the development of boards by means of a screw press. In case sample series (Figs. 13–18), the press as well as the procedure were improved; it is obvious that board surface layer has improved and so have the mechanical and physical properties. In the cooperation with producer of particleboards were mixed and pressed full cement bonded particleboards with treated hemp as a filler. The slabs were conditioned as well as common cement bonded particleboards in drier and then sawed up to testing samples.

DTA, reference sample with wood particles.

Figures 5–6. cleboards.

Wood particles used in cement bonded parti-

Figures 7–8. Untreated hemp chips, type A (left) and grounded and treated hemp chips, type C (right). Figure 3.

DTA, reference sample with hemp particles.

Figure 4. Hemp in the cement matrix, in fact, line in the bottom right corner shows 1 mm.

Figures 9–10. type B.

Grounded and treated hemp particles,

Figures 11–12.

Grounded and treated hemp chips, type D.

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Figures 13–14. First series of hemp based cement bonded particleboard samples (untreated, type A and treated hemp, type B).

Figure 19.

Composition by volume.

Figures 15–16. Samples of hemp based cement bonded particleboards (treated hemp, type B + C)—without surface treatment. Figures 20–21. Hemp based cement bonded particleboards in fresh condition and final product hemp particleboards.

both sides of coarse fraction of slab mixture. This is the reason of fine surface of final products. Figures 17–18. Samples of hemp based cement bonded particleboards (treated hemp, type B + C)—without and with surface treatment. Table 1. Material strength characteristics and their comparison. Particleboards Parameter

Hemp

Values

Average

Bulk density [kg/m3 ] Tensile strength [N/mm2 ] Modulus of elasticity [N/mm2 ]

1,285 11.9 7,330

Wood

ˇ CSN EN 634-2 Minimal

1,350 11.5 6,800

1,000 9.0 4,500

On these samples were in the labs determined their physical and mechanical properties (Table 1). 3.3

Composition

Hemp based cement bonded particleboards which were made are from treated hemp particles, cement, water and hydration admixtures. Fine fraction lay on

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4

CONCLUSIONS

Strength values (28 days, standard methodology for cement based materials) of tested samples made from hemp particleboards were about 40% smaller than values shown in Table 1. Through all previous experiment conclusions it is interesting finding. One of the theories is that hempen particles absorb too much mixing water; that slows cement hydration. This behavior will be tested further. Fact is that strength values of two months old samples were much better as shows Table 1. Currently, it can be said, with certain limitations, that the results of physical-mechanical properties of hemp based cement bonded particleboards are, as compared with the properties of wood-cement particleboards, on a very similar level. It can be expected that hemp based cement bonded particleboards will show better heat insulating and acoustic properties due to hemp material characteristics. Fast renewability, profitability, 100-percent utilization and other excellent properties of hemp make it a progressive raw material for the production of building materials in the 21st century that can be a perfect alternative to building materials made of wood.

REFERENCES

Agarwal, B.D. & Broutman, L.J. 1987. Fibre Composite materials. Praha: SNTL. 1st ed. 296 p. Tošovská, M. & Buchtová, I. 2008. Situaˇcní a výhledová zpráva len a konopí 2008. Praha: Department of agriculture. [online]. ISBN 978-80-7084-695-7, Available at www.mze.cz

Khestl, F. & Bydžovský, J. 2007. Hemp utilization in cementbonded particle boards. ISEC-4. London: Taylor & Francis/ Balkema. p. 595–598. ISBN 978-0-415-45755-2. Bydžovský, J. & Khestl, F. 2008. Technical Hemp as an Alternative Material—Processing and Utilization in Building Industry. 14 CCIA. Havana, Cuba, CUJAE. pp. 80–84. ISBN 978-959-261-281-5.

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Influence of admixtures on performance of roller compacted concrete P. Hafiz Islamic Azad University, Dubai, UAE

A.R. Khaloo Sharif University of Technology, Tehran, Iran

ABSTRACT: The combination of two construction methods of concrete and earth fill dams have resulted in the emergence of Roller Compacted Concrete (RCC) method. The primary goal of the present study is to investigate the influence of admixtures on performance of RCC. All samples have been cast according to ASTM specifications. Influence of type and dosage of admixtures on compressive behavior, vebe time and specific gravity are tested. An analytical model has been developed to predict the stress-strain curve of RCC behavior, considering the type and percentage of admixtures. 1

INTRODUCTION

One billion people will be added to the world’s population in the next 10 years, while today, many of under developed countries have some problems of lack of water. It is important to have a fast and less expensive method of dam construction. The first RCC dam was put into operation in Japan in 1981, and since then till 2004, about 475 dams of this kind have been constructed in 32 countries (Dunstan 1999). Concrete, compacted with roller, known as ‘‘Roller Compacted Concrete’’, is a kind of concrete with a low water and cement content. In fact, combining the construction methods of two types of concrete and earth fill dams has resulted in the emergence of RCC method. About 38% of the dams of 15 meters height in the world (except in China) were concrete dams till 1950. This proportion reached to 25% from 1951 to 1977 and 16.5% during 1982. Indeed, although the costs of earth dams’ construction are lower, according to what experience shows, concrete dams are safer than earth dams (Dunstan 1994; Hansen 1991). Researchers were looking for a new kind of dam with the safety like concrete dams and ease of construction like earth dams. Therefore RCC dam construction method was introduced in 1960 to 1970. This idea was put into use in Italy’s ‘‘Alpe Gera’’ dam with 172 m in height in 1964. The main difference between that dam and today’s RCC dams is that, Italian engineers did not use roller to compact the concrete (Hansen 1991). Certainly, using RCC in reconstruction of Pakistan’s ‘‘Tarbela’’ dam has had an important role in developing RCC dams. High pace of material production in this project has been so far the highest rate in the world. In 1981, the first RCC dam was constructed in Japan.

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The goal of this research was to study the influence of several admixtures on performance of RCC. Fifty-six 6 in × 12 in cylinder RCC samples have been constructed and tested using fresh concrete and compressive behavior as criteria. Many properties of admixtures are provided for customers by producers; however some of their influences may remain unknown. So, in order to clarify the use of admixtures for obtaining desirable results, it is required to carry out some tests on the performance of concrete (Neville & Brooks 1990).

2

EXPERIMENTAL PROGRAM

In this research, a fixed mix design has been used to compare the influence of different admixtures. Sand and coarse aggregate used in RCC mix design were crushed stone and type II cement was provided by Tehran cement factory. A common mix design with 250 kg of cement and 110 kg of water per one cubic meter of concrete was selected (w/c = 0.44). The admixtures used are superplasticizer with strong water reduction ability, high performance plasticizer, and superplasticizer mixed with retarder. The admixtures were added to RCC mix design with two different percentages of 0.5 and 1% of cement weight. The gradations of aggregates were selected according to ‘‘U.S Army’’ Standard EM 1110-2-2006 and are shown in Tables 1 and 2 (US Army Corps of Engineering Manual 2000). Three standard tests are carried out on RCC samples (ASTM C1176 1994; ASTM C1170 1994): – Workability Test: Since RCC is a concrete without slump, this test is performed to identify the

3

Table 1. Coarse aggregate grading. Aggregate size mm

Percent finer∗ percent

Coarse aggregate percent

19–25 12.5–19 9.5–12.5 4.75–9.5

85 55 35 0

15 30 20 35

US army corps of engineers.

Table 2. Fine aggregate grading. Sieve no. mm

Percent finer∗ percent

Selected passing percent

4 8 16 30 50 100 200

95–100 75–95 55–80 35–60 24–40 12–28 6–18

100 85 70 50 30 20 8

∗

US army corps of engineers.

Table 3.

Vebe time, specific gravity and RCC compressive strength are the parameters on which, the influence of admixtures will be described. Vebe time: Vebe time indicates the workability of RCC. It is the time in which the concrete paste moves up around superload during the vibration of vebe table [7]. Figure 1 shows the influence of various admixtures on vebe time of RCC. It has been observed that using admixtures, generally results in reduction of vebe time. Vebe time was considerably increased in case of using 0.5% superplasticizer, which may have caused by non-uniform concentration of solid particles in the admixture. It can be also seen that vebe time of samples, having 1% of admixture, is reduced more than those of having 0.5% admixture, as was expected. Figure 1 shows that using retarder can usefully make a delay in RCC initial setting (for example in hot environments). Specific gravity: An increase in specific gravity of all samples due to admixtures is demonstrated in Figure 2. The increase in specific gravity ranges from 0.3 to 4.7 percent. Compressive behavior: All the admixtures used in this study, because of having the essence of plasticizing

Experimental program.

Type of admixture

No. of Percentage samples of cement weight 7 days 28 days

Superplastisizer Superplastisizer Plastisizer Plastisizer Superplastisizer + Retarder Superplastisizer + Retarder Admixture-free

0.5 1.0 0.5 1.0 0.5 1.0 0.0

4 4 4 4 4 4 4

60 60

50

4 4 4 4 4 4 4

Vebe time (sec)

∗

EXPERIMENTAL RESULTS

40 30 30 25

10

0 1

A universal testing machine with strain control capability is used to obtain the complete compressive stress-strain curve of RCC. The loading rate was 0.1 mm/sec. So far no research result has been reported on the σ − ε curve of RCC in the literature. Table 3 presents the experimental program.

20 13

non addative

Figure 1.

2

3

4

5

6

superplastisizer 0.5%

superplastisizer 1.0%

plastisizer 0.5%

plastisizer 1%

retarder 0.5%

7 retarder 1.0%

Vebe time of samples.

2700 2674

2632 2619 2610 specific gravity,kg/m^3

workability of RCC by using a standard cylinder mold filled with RCC and screwed on a vibrating table called vebe table, under an exerted superload. – Specific Gravity Test: After performing the previous test, this test is carried out on wet concrete according to standard method. – Compressive strength test: The 48 main and 8 pilot RCC samples have been tested under compression at the ages of 7 and 28 days.

25 20

20

2600

2588 2560

2553

2500

2400 1

2

admixture-free

Figure 2.

502

superplastisizer 0.5%

3 superplastisizer 1.0%

4 plastisizer 0.5%

5

6

plastisizer 1%

retarder 0.5%

Specific gravity of samples.

7 retarder 1.0%

18

25

+34.23%

+29.37%

20

+30.79% +7.34%

0.00 %

max =

14

0

15.58 MPa

= 0.0070

12

-1.98%

15

stress,MPa

Compressive strength MPa

16

+40.08%

10

10 1/2

max=

7.79 MPa

50

= 0.0113

8 6

5

4 2

0 1

admixture-free

Figure 3. 7 days.

3

superplasticizer 0.5%

5

superplasticizer 1.0%

7

9

11

13

plasticizer 0.5%

plasticizer 1.0%

retarder 0.5%

retarder 1.0%

0 0

Average compressive strength of samples at

+19.59%

+13.10%

0.00 %

stress,MPa

Compressive strength MPa

25

+20.76%

0.006

0.008

0.01

0.012

-5.94%

20 15 10

sample1 sample2 model

26 24 22 20 18 16 14 12 10 8 6 4 2 0

5

0.016

0

0.002

0.004

f c = 20.6 ×

0.006

0.018

0.02

0.022

0.024

0.008

0.01

0.012

113 5800 2 1 240 + 25000

0.014

0.016

0.018

2

0.02

strain

0 1

admi xtu re-free

Figure 4. 28 days.

3

5

supe rplasticizer supe rplasticizer 0.5% 1.0%

7

9

11

13

plasticizer 0.5%

plasticizer 1.0%

reta rder 0.5%

reta rder 1.0%

Average compressive strength of samples at

Figure 6. 28 day stress-strain curve of admixture free sample and analytical model .

35

ANALYTICAL MODEL FOR PREDICTING STRESS-STRAIN CURVE

As explained in previous part the complete stressstrain diagram of RCC samples are obtained to make a better understanding of RCC behavior. Figure 5 presents a typical stress-strain curve. The typical curve shows that a high range of strain is achieved during the test. The strain at peak strength is 0.007 and that

503

2 150 1 176 + 8000 2

25

stress,MPa

may result in liquidity of cement paste and cause better movement and consequently better adhesion between aggregates and cement paste. Therefore, before performing the tests, it was expected that admixtures increase compressive strength of RCC, as obtained results approved. All samples containing admixture experienced increase in strength; except samples with retarder in which the strength gain of 7 and 28 days was delayed. The compressive strength of similar samples has been averaged and is compared in figures 3 and 4. The maximum increases of 40.1% and 34.2% are obtained with superplastisizer. The result of retarders in Figures 3 and 4 show that retarders have worked properly in decreasing the rate of RCC setting.

f c = 31.7 ×

sample1 sample2 model

30

4

0.014

Figure 5. Typical 7 day stress-strain curve of a sample containing retarder.

+38.76% +22.40%

0.004

strain

35 30

0.002

20 15 10 5 0 0

0.002

0.004

0.006

0.008

0.01

0.012

strain

Figure 7. 28 day stress-strain curve of 1% plasticizer sample and analytical model.

at 50% of peak strength in the descending part of the curve is about 0.011. The RCC exhibits a relatively ductile behavior. Stress-strain curve of RCC at 28-day age is modeled using Sargin formulation as shown below [6]:  A(ε ε0 ) + (B − 1)(ε/ε0 )2 fc   (1) = f ‘c 1 + (A − 2)(ε ε0 ) + B(ε ε0 )2 

where fc is compressive strength and ε0 is strain corre sponding to fc . The formulation is calibrated using experimental results. With regard to the type and percentage of

stress,MPa

f c = 20.8 ×

24 22 20 18 16 14 12 10 8 6 4 2 0

sample1 sample2 model

0

0.002

0.004

0.006

0.008

0.01

125 9010 2 1 172 + 12600

0.012

0.014

2

0.016

strain

stress,MPa

Figure 8. 28 day stress-strain curve of 1% retarder sample and analytical model. f c = 27.6 ×

30 28 26 24 22 20 18 16 14 12 10 8 6 4 2 0

sample1 sample2 model

0

0.002

0.004

0.006

0.008

155 20600 2 1 172 + 6280 2

2. Simultaneous use of retarder and superplasticizer increases concrete workability and delays concrete hardening, which is the main function of using retarder. 3. Specific gravity of all of the samples has been increased due to the presence of admixtures. Superplasticizer has the maximum specific gravity increase (4.8%). 4. All admixtures increased the strength at both 7 and 28 day compared to admixture-free samples except for the samples containing retarder admixture. Admixtures have more influence on 7-day strength than 28-day strength. 5. Admixtures can be used in RCC in order to reduce the vebe time and provide enough time for placing concrete especially in dams. 6. Generally the admixtures do not change the strain corresponding to maximum stress. 7. Initial slope of stress-strain curve of all samples with admixture is almost equal to samples without admixture, which indicates little influence of admixtures on ‘‘elastic modulus’’ of RCC.

0.01

strain

Figure 9. 28 day stress-strain curve of 1% superplastisizer sample and analytical model.

admixtures, RCC behavior is predicted using equation 1 and averaging of similar samples. It must be mentioned that there were 4 samples of RCC for each type and percentage of admixture. Two of these 4 samples (named as sample 1 and sample 2 in the legends of the curves) were selected to approximate the analytical model of each case. The model predicts the stress-strain curves in both ascending and descending parts satisfactorily (Figures 6 to 9). 5

CONCLUSIONS

Based on the results of experimental and analytical study presented in this paper the following conclusions are reached:

REFERENCES ASTM, 1994. Standard Practice for Making Roller Compacted Concrete in Cylinder Molds using a Vibrating Table, Designation:C 1176–91, USA. ASTM, 1994. Standard Test Methods for Determining Consistency and Density of Roller Compacted Concrete using a Vibrating Table, Designation: C1170–91, USA. Dunstan, M.R.H. 1999. Recent development in RCC dams, hydropower & dams, 6(1): 40–45. Dunstan, M.R.H. 1994. The state—of—art of Roller compacted concrete dams, Hydropower & dams: 44–45. Hansen, K.D. 1991. Roller compacted concrete: 1–83. Mc-Graw Hill. Neville, A.M. & Brooks, J.J. 1990. Scientific and Technical publications, Longman. Sargin, 1971. Stress-strain Relationship for concrete and the Analysis of structural Concrete Sections, Study No. 4, University of Waterloo, Ontario, Canada. U.S. Army Corps of Engineers Manual. 2000, Roller Compacted Concrete Engineering and Design, No. 1110-22006 Washington, DC.

1. Suitable use of admixtures reduces vebe time of RCC (increases the workability).

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Investigation of the effect of aggregate on the performance of permeable concrete C. Lian School of Natural and Built Environments, University of South Australia, Adelaide, South Australia, Australia

Y. Zhuge University of Southern Queensland, Brisbane, Queensland, Australia

ABSTRACT: Presently natural resources are increasingly consumed due to rapid urbanization, so that various strategies are being investigated by engineers to protect and restore natural ecosystems all over the world. Permeable pavement, due to its high porosity and permeability, is considered as an alternative to traditional impervious hard pavements for the sake of controlling stormwater in an economical and friendly environmental way. Concrete as a construction material has been used in pavement surfaces since 1865 and during the past 20 years, permeable concrete developed into a subset of the broader family of pervious pavements. It is normally made of single sized aggregate bound together by Portland cement. Considering about influences of aggregate properties in the concrete compositions, different aggregate types and sizes were tried and their effects on the compressive strength and permeability of permeable concrete were investigated. The optimum aggregate for pervious concrete is consequently recommended. 1 1.1

INTRODUCTION Permeable pavement

Presently natural resources are increasingly consumed due to rapid urbanization and thereafter human construction activities, so that various strategies are being investigated by engineers to protect and restore natural ecosystems all over the world. Permeable (porous/ pervious) pavement is termed as comprising materials that facilitate storm water infiltrate and transfer to the underlying subsoil (ARMCANZ & ANZECC 2000). With sub-structure which stores water underground temporarily, it is called permeable pavement system. Instead of installing rainfall detention ponds or soakaways, this new system is more cost effective compared to the traditional impervious pavement. Meanwhile, it has been acknowledged by many researchers that permeable pavement system is capable of reducing the sediments and contaminants for lessening the pollutant loads on stormwater, thus it is considered as an economic and environmental-friendly construction as a part of city drainage system. In Australia, permeable pavement has been utilized as a potential tool of Water Sensitive Urban Design (WSUD) to manage natural water. From 1994 the University of New South Wales (UNSW) started to research into permeable concrete paving and more recently the University of South Australia (UniSA) is also involved. However, the previous studies conducted both in UNSW and UniSA mainly concentrated on

505

water quality and pollution control through permeable pavements and, only the properties of basecourse materials in permeable pavement system and segmental paving have been studied. There is still a gap of optimizing the surface materials for permeable pavements. 1.2

Permeable concrete pavement

The materials used for permeable pavement are classified into nine categories (Ferguson 2005): porous aggregate, porous turf, plastic geocells, open-jointed paving blocks, open-celled paving grids, porous concrete, porous asphalt, soft paving materials, and decks. Concrete has been used in pavement surfaces since 1865, when dense concrete street pavements were first experimentally installed in Scotland (Croney 1997). Porous concrete was first used in pavements during World War II. As a subset of the broader family of permeable pavements, porous concrete is also referred to as permeable concrete, enhanced porosity concrete, or Portland cement pervious pavement. It is normally made of single–sized aggregate bound together by Portland cement, physically and chemically identical to dense concrete (Ferguson 2005). Permeable concrete is relatively porous, providing by the omission of fine aggregates (Scholz & Grabowiecki 2007) and filled most of volume with coarse aggregate, thus, porous concrete obtains more voids in the structure leading to good water infiltration and air exchange rates. Permeable concrete typically has a void content of 15–25% compared to 3–5% for

conventional pavements according to the Interlocking Concrete Pavement Institute (ICPI 2007). Comparing with porous asphalt, permeable concrete exhibits some advantages in pavement projects. For instance, porous concrete has a better capacity of keeping high porosity in hot weather (ICPI 2007), which is more suitable for Australia’s climate. Nonetheless, the compressive strength and flexural strength are sufficient for low volume traffic areas but not for heavy traffic loading roads (Ferguson 2005). Currently they are mainly used in carparks, footpaths and bicycle trails. This study aims to improve the strength of porous concrete without losing permeability so that it could be adoptable for supporting higher volume traffic. 2

LITERATURE REVIEW

It has been generally accepted that the strength of concrete is influenced by many factors, such as the amount and type of cement, aggregate, water to cement ratio, chemical additives and curing conditions. From the view of composite structure, Larrard & Belloc (1997) pointed out the strength of concrete was indeed determined by the properties of mortar, coarse aggregate and the interface. For normal concrete, previous researches have revealed the effects of aggregates on strength with different aggregate type, size and gradation. However, these conclusions for normal concrete cannot be simply extended to permeable concrete, since porous concrete typically does not contain fine aggregate to fill the voids, only relying on cement paste to bond graded coarse aggregate together. Research of pervious concrete has ever been conducted at Tennessee Technological University (Crouch et al. 2007). It is indicated that the compressive strength, effective void content and permeability are largely dependent upon the aggregate. Crouch et al. (2007) stated that not only the size of aggregate, but also the gradation and amount of aggregate could affect the compressive strength and static modulus of elasticity on pervious Portland cement concrete. Meininger (1988) used different aggregate sizes (10 mm and 19 mm) in nonfine concrete study and the results showed that larger aggregate sizes would result in lower compressive strength, which corresponded with the results found from Yang & Jing (2003). It claimed the decrease of aggregate size led to higher pervious concrete strength, resulting from the increase of the interface strength between the aggregate and cement paste (Yang & Jing 2003). Ghafoori and Dutta (1995) also set up the relationship between gravimetric air content and permeability and porosity in no-fines concrete. However, in Australia there has been no published research that reveals the effect of aggregates on the structural performance of pervious concrete. The objective of this paper is to investigate the effect of aggregate on the performance of pervious concrete using locally available materials.

3 3.1

EXPERIMENTAL INVESTIGATION Materials

3.1.1 Cement Normal Portland cement from local supplier was used in each mix design. It exceeds the minimum specification given in AS (Australian Standard) 3972-1997. 3.1.2 Aggregate Different types of aggregates exhibit different strength, permeability and geometry stability due to different mineral composition, grain sizes, types of formation, texture and location of the aggregates source. Coarse aggregate is mainly used as a primary ingredient in making the previous concrete. Fine aggregates were not added to the mixture in this research. According to Krezel (2006), crushed igneous rocks are more preferable as coarse aggregate for concrete due to their higher strength. However, since the availability of igneous rock in Australia is becoming scarce Krezel (2006), this research diverted to the crushed sedimentary and metamorphic rocks. Three types of coarse aggregate were obtained from local quarry: quartzite, dolomite and limestone. Dolomite was a sedimentary carbonate rock, composed of the mineral dolomite, also contained impurities such as calcite, quartz and feldspar. Dolomite formed in groups of rhombohedral crystals with curved, saddle-like faces. Limestone was also sedimentary rock. Although some limestones were nearly pure calcite, there were often varying amounts of clay, silt and sand. Quartzite was a dense, hard metamorphic rock. The quartzites obtained from local quarry were red due to a large amount of iron oxide. The geology and mechanical properties of aggregate source were tested and given in Table 1. The proportions of all sample mixtures were designed at aggregate to cement ratio of 4.5 and water to cement ratio of 0.36. 3.2

Sample preparation and testing methods

3.2.1 Sieving The preparation of standard concrete test specimens is based on Australian Standards and Guidelines. For mix proportioning purposes, all of the raw 10 mm Table 1. Engineering properties of aggregates. Mean Flakiness water index absorption Aggregate % %

Los Angeles abrasion Dry value strength % KN

Type A Type B Type C

27 15 38

21 35 15

2.8 0.8 0.3

163 225 74

Type A: Quartzite; Type B: Dolomite; Type C: Limestone.

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aggregates from quarries were sieved and separated into different groups using standard sieves. Specific gradations were then obtained by recombining small fractions of separated aggregates. The mixed grading of each batch was shown in Table 2. 3.2.2 Casting and compaction Before the mixing, aggregates were washed using tap water and dried in oven for one day to clean the silt or crusher dust, in case they prevent the development of good bond between aggregate and cement paste in concrete mixture. A total of 8 cylinders with 100 mm diameter and 200 mm height were cast for each batch to explore the compressive strength and two steel beam moulds were cast when testing the flexure strength. The compaction method for making porous concrete is one of the most influential factors in the sample preparation. Two compaction methods have been assessed in previous research (Zhuge 2008), one was using compaction hammer and the other was using vibration table. While the hammer compaction packed the aggregate particles together more tightly, the density of porous concrete samples increased with the loss of permeability. As the impaction strength of a falling hammer was so strong to crush the weak aggregate and create weak layers, the vibration method seemed to be more suitable for majority of aggregates, such as limestone and dolomite. However, for the sake of achieving the maximum cohesion between aggregate particles, a combined compaction method was attempted, that was, not only applied the standard rodding compaction method, but also incorporated a static compactor in the consequent vibrating procedure. The frequency of vibration table was controlled at 75 Hz. This method allowed the coarse aggregate not deformed under compacting whilst increase the contact surface and alignment of aggregate particles,

Table 2.

Aggregate size distribution.

Sieve size (mm)

16

Mix number

Passing percentage by mass (%)

Type A A1 A2 A3 Type B B1 B2 B3 Type C C1 C2 C3

13.2

9.5

6.7

4.75

2.36

1.18

100 100 100

100 100 100

100 100 90

0 30 30

0 0 0

0 0 0

0 0 0

100 100 100

100 100 100

100 100 90

0 30 30

0 0 0

0 0 0

0 0 0

100 100 100

100 100 100

100 100 90

0 30 30

0 0 0

0 0 0

Figure 1.

Permeability test rig.

which was believed a substantial aspect to enhancing the strength of porous concrete. 3.2.3 Testing For compression test, the casted cylinders were demoulded after 24 hours, labeled and weighted. Then the samples were cured in a lime bath at 23 ± 2◦ C, according to AS 1012.8.1-2000. For each batch, two samples were prepared in permeability testing and others were for compression, three tested at 7 days and 28 days respectively. Sulphur caps were placed on the ends of samples before loading process. The unconfined compressive strengths (UCS) of specimens using different type of aggregates were determined in lab according to AS1012.9-1999. For flexure test, the moduli of rupture were determined in lab according to AS 1012.11-2000. For the permeability measurement, test apparatus was improved based on previous research. A cylindrical plastic pipe with inline steel wire and adjustable steel tie fasteners rendered the tubing device tighter to hold up the water leak from the sides of samples (see Figure 1). Permeability as a unique ability for water to penetrate through the porous concrete was expressed in millimetres per second (mm/s). Since the porous concrete generally own a much higher permeability compared to the normal dense concrete, the permeability test method for the latter one were not still suitable and valid for testing the porous concrete accurately. Thus, the falling head test method was used to determine the permeability of the all the samples and the operation was similar to the falling head test for soil, which complied with AS 1289.6.7.2-2001.

4 4.1

RESULTS AND DISCUSSIONS Compressive strength

The average compressive strength of porous concrete specimens made with quartzite, dolomite and limestone aggregates at 7 and 28 days were illustrated in Table 3.

0 0 0

507

As it is showed in Table 3, with the identical singlesized aggregate (Group 1), quartzite porous concrete A1 developed the compressive strength of 11.6 MPa and 11.8 Mpa at 7 and 28 days respectively. Dolomite B1 yielded 15.0 MPa and 15.8 MPa and Limestone C1 reached 14.3 MPa and 15.5 MPa. When extending the aggregate size fraction into 4.75 mm as described in Table 2, the compressive strength for quartzite and dolomite concrete were both increased (A2 and B2) except for limestone concrete which was slightly decreased (C2). The porous concrete made with dolomite produced the highest compressive strength among the three types of aggregates. This type of aggregate was further investigated with the size grading varying from 13.2 mm to 4.75 mm (B3). However, the results indicated that the dolomite concrete with this aggregate gradation (B3) presented a lower strength than that of B1 and B2.

4.2 Flexural strength The average flexural strength (modulus of rupture) of porous concrete specimens made with quartzite, dolomite and limestone aggregates at 7 and 28 days were shown in Table 4. With the same aggregate grading, dolomite porous concrete yielded the highest flexural strength compared to quartzite and limestone. As indicated in Table 4, it was 1.7 MPa and 1.9 MPa at 7 and 28 days curing time respectively. In addition, the flexural strength of dolomite B2 was 70% higher than B3 at 7 days and was 60% higher than B3 at 28 days which was similar to the results for compressive strength. Table 3.

Compressive strength at different ate. Compressive strength (MPa)

Curing time (days)

Quartzite

Dolomite

Limestone

A1

A2

B1

B2

B3

C1

C2

7 28

11.6 11.8

13.0 15.5

15.0 15.8

16.0 19.0

14.3 15.5

14.3 15.5

13.5 14.0

Table 4.

Flexural strength at different age.

Table 5. Permeability of porous concrete made with different aggregates at 28 day curing time. Permeability (mm/s) Quartzite

Dolomite

A1

A2

B1

B2

B3

C1

C2

27.47

13.67

19.87

8.51

14.78

13.27

15.99

4.3

Curing time (days)

Quartzite

Dolomite

Limestone

A3

B2

B3

C3

7 28

1.5 1.6

2.9 3.0

1.7 1.9

1.5 1.5

Permeability

The permeability measurement was conducted after 28 days curing time. The average permeability of porous concrete specimens made with quartzite, dolomite and limestone aggregates were given in Table 5. Three types of aggregates all showed a satisfied permeability, thus there should be a space for the future research to enhancing the strength of porous concrete made with them, because it reflected there were still enough pore voids exiting at this stage. 4.4

Effect of aggregate type

The results indicated that the type of coarse aggregate used in making porous concrete would influence the strength of porous concrete even though the aggregates were in the same size and gradation. This can be attributed to the different particle shape and texture of different aggregate, as it was shown in Figure 2. Because describing the shape of aggregate cannot only rely on vision, the flakiness index of aggregate was conducted according to AS1141.15-1999. As it was shown in Table 1, the dolomite was most flaky and limestone was the least flaky one. It could be regarded as a reason why the aggregate strength of limestone was nearly 30% lower than that of dolomite, but the compressive strength reached around 95% of dolomite (C1 versus B1) without the influence of aggregate size. It was estimated that the more flaky aggregate particles tended to be oriented in one plane under compaction force, which adversely affected the contact area between aggregate and cement, so that the more flaky aggregate did not bond with cement as well as the more rounded aggregate, such as limestone. 4.5

Flexural strength (MPa)

Limestone

Effect of aggregate strength

Comparing the porous concrete samples made with these aggregates, it can be observed that the higher strength of aggregate will result in a higher strength of porous concrete and this effect is the same regardless of whether the porous concrete is under compression or flexure. It is understandable that the strength of porous concrete cannot immensely exceed that of the major part of aggregate particles contained in. Higher

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the highest permeability compared to dolomite and limestone porous concrete. 4.7

(a) Quartzite

(b) Dolomite

(c) Limestone

Figure 2. Comparison of different aggregate.

strength aggregate, such as dolomite tended to sustain the higher stress than the lower strength aggregate, such as limestone. This property could be used to select aggregates to produce high strength porous concrete. 4.6

Effect of aggregate size and gradation

For a certain type of aggregate, take dolomite as an example, the immerged proportion of smaller size aggregate produced the higher strength of porous concrete. It can be seen from Table 3 that when changing from a single sized grading (B1) to a grading varying from 9.5 mm to 4.75 mm (B2), the compressive strength of porous concrete increased from 15.0 MPa to 16.0 MPa at 7 days and from 15.8 MPa to 19.0 MPa at 28 days. However, when larger sized aggregate was used (B3), although it showed a better gradation, the flexural strength of porous concrete decreased from 2.9 MPa to 1.7 Mpa at 7 days and from 3.0 MPa to 1.9 Mpa at 28 days when the maximum aggregate size increased from 9.5 mm (B2) to 13.2 mm (B3). It seemed the flexural strength of porous concrete was more affected than the compressive strength; although the extent of this size influence were not equal for compressive strength and flexural strength, it can be concluded that smaller aggregate size will result in a higher compressive strength and flexural strength, which was consistent with the research of Meininger (1988) and Marolf et al. (2004). Based on the results of permeability (see Table 5), it can be found that the smaller aggregate size will lead to a lower permeability of porous concrete except for that made with limestone. With the same aggregate size gradation, quartzite porous concrete obtained

509

Failure mode and bonding

It was observed that the majority of failures for porous concrete samples intensively took place in the hardened cement paste or the interface between cement and aggregate (see Fig. 3). The fractures through the aggregates were less than the kind of former two; this failure was determined by the strength of aggregate. More fractured aggregate particles appeared in the porous concrete made with limestone than that with dolomite or quartzite. However, concrete as a three phase composite material at a microscopic scale included mortar matrix, aggregate and the interfacial transition zone between the two. Although the interfacial transition zone was smaller in proportion compared to mortar matrix and aggregate, its characters influenced the mechanical behaviour of concrete significantly and it was normally regarded as the weakest link in concrete (Prokopski & Halbiniak 2000). On this hand, the porous concrete seemed to perform the same as normal concrete, which corresponded with the research of Bentur (1990). Bentur (1990) also believed there were two weak faces in the interfacial transition zone, the aggregate contact layer and matrix contact layer. On the other hand, there was a little difference between porous concrete and the normal concrete in the mode of fracture. For normal concrete, Zaitsev (1983) pointed out the separation crack occurred first due to the shrinkage of cement matrix and then along the interface of the aggregate and cement paste. Whereas the more fractures developed in the interfacial zone of porous concrete in this study, it could be certified that without fine aggregate, such as sand and any chemical admixture, the bond strength of aggregate and cement in porous concrete was not adequate at this stage and thereafter it became a controlling factor in improving the strength of porous concrete.

Figure 3.

Cracked samples of porous concrete.

4.8

Effect of other engineering properties of aggregate

Besides what have been mentioned above, the results in Table 1 also suggested dolomite was more resistant to abrasion for porous concrete, this character should be considered when the porous concrete is expected to use as pavement material in road construction. In addition, despite the quartzite showed a lower flakiness index and a better permeability than dolomite as an aggregate in this research, the clay contamination and impurities such as a large amount of iron oxide covered on the surface of quartzite cannot be omitted, for the purpose of gaining a good development of bond in porous concrete. 5

CONCLUSIONS

The laboratory testing has been carried out to explore the optimum type of aggregate for porous concrete using Australian local quarries. Three most common types of aggregate were applied and the effects of their properties were compared. Along with the study conducted on aggregate size distribution, it can be concluded that the grading of aggregate also need to be controlled in order to achieve the best strength of porous concrete. The preliminary testing results indicated that dolomite might be the proper type of aggregate for porous concrete as a permeable pavement material. REFERENCES ARMCANZ & ANZECC: Australian and New Zealand Environment and Conservation Council and Agriculture and Resource Management Council of Australia and New Zealand 2000. AS 3972-1997: Portland and blended cements. AS1012.9-1999: Methods of testing concrete—Determination of the compressive strength of concrete specimens. AS1141.15-1999: Methods for sampling and testing aggregates—Flakiness index. AS 1012.8.1-2000: Methods of testing concrete—Method of making and curing concrete—Compression and indirect tensile test specimens.

AS 1012.11-2000: Methods of testing concrete— Determination of the modulus of rupture. AS 1289.6.7.2-2001: Methods of testing soils for engineering purposes—Soil strength and consolidation tests— Determination of permeability of a soil—Falling head method for a remoulded specimen. Bentur, A. 1990. Microstructure interfacial effects and micromechanics of cementitious composites. Ceramic Trans 16: 523–549. Crouch, P.E. et al. 2007. Aggregate Effects on Pervious Portland cement Concrete Static Modulus of Elasticity. Journal of materials in civil engineering 19(7): 561–568. Croney, P. & Croney, D. 1997. The design and performance of road pavements. New York: 508. Ferguson, B.K. 2005. Porous pavements. USA: CRC. Ghafoori, N. & Dutta, S. 1995. Laboratory investigation of compacted no-fines concrete for paving materials. Journal of materials in civil engineering 7(3): 183–191. ICPI. 2007. Segmeatal concrete pavement resources for design professionals. Krezel, Z.A. 2006. Recycled aggregate concrete acoustic barrier. Swinburne University of Technology: http://adt.lib. swin.edu.au/uploads/approved/adt-VSWT20060821.154 340/publication Larrard, F. & Belloc, A. 1997. Influence of aggregate on the compressive strength of normal and high-strength concrete. ACI materials journal 94(5): 26–27. Meininger, C. 1988. No-fines Pervious Concrete for paving. Concrete International 10(8): 20–27. Marolf, A. et al. 2004. Influence of aggregate size and gradation on the acoustic absorption of enhanced Porosity Concrete: http://cobweb.ecn.purdue.edu/∼concrete/wei sis/publications/journal/Rj017.pdf. Prokopski, G. & Halbiniak, J. 2000. Interfacial transition zone in cementitious materials. Cement and Concrete Research 30: 579–83. Scholz, M. & Grabowiecki, P. 2007. Review of permeable pavement systems. Building and Environment 42(11): 3830–3836. Yang, J. & Jiang, G. 2003. Experimental study on properties of pervious concrete pavement materials. Cement and Concrete Research 33(3): 381–386. Zhuge, Y. 2008. Comparing the performance of recycled and quarry aggregate and their effect on the strength of permeable concrete. 20th Australasian Conference on the Mechanics of Structures and Materials, Toowoomba, Dec. 215–221. Zaitsev, Y.B. 1983. Crack propagation in a composite material. Fracture Mechanics of Concrete 8: 251–299.

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Investigations on flexural behavior of high strength manufactured sand concrete V. Bhikshma, R. Kishore & C.V. Raghu Pathi Department of Civil Engineering, University College of Engineering, Osmania University, Hyderabad, India

ABSTRACT: Sand is basic concrete making construction material required in large quantities. Hence, in the present scenario, it is necessary to find the most suitable substitute for sand, easy to produce and has all the required qualities for use in concrete. Manufactured sand is one among such materials to replace river sand, which can be used as an alternative fine aggregate in mortars and concretes. To attain the set out objectives of the present investigation, M50 grade concrete has been considered. Strength properties such as cube compressive strength and flexural strength of beams, and load carrying capacity, moment carrying capacity, behavior of strains in compression as well as tension fibers and cracking patterns have been studied for the grade of concrete. In this paper a total of 15 cube specimens 150 × 150 × 150 mm and 10 beam specimens of size 1500 × 150 × 230 mm were cast for testing. The results have been compared for the specimens made with natural fine aggregate. 1

INTRODUCTION

Construction has become technology oriented with new and better materials and techniques being developed. In this quest, better materials and techniques in construction have gained a strong foothold and found wide applications. However, considerable research is still to be done in the field of building materials to reduce building costs further and improved structural performance. Concrete is one of the versatile construction materials widely used by the industry. Continuous research efforts have established concrete as a versatile material. Conventionally concrete is a mix of cement, fine and coarse aggregates. The use of manufactured sand in concrete is gaining momentum these days. The present experimental investigations have been made on concrete using manufactured sand as fine aggregate and observed the effects of crushed manufactured sand on strength properties of concrete. 1.1 Aims and objectives 1. To examine the workability of different percentages of manufactured sand concrete and compare with natural sand concrete. 2. To investigate the performance of different percentages of manufactured sand concrete beams in terms of its load carrying capacity and moment carrying capacity and compare with corresponding natural sand concrete beams. 1.2 Scope of present investigation Conventional concrete and M50 grade of concrete from 25%, 50%, 75% & 100% of manufactured Sand

have been considered for casting concrete cubes and beams. Also 30 concrete cubes and 10 reinforced beams have been considered for M50 grade of concrete

2

LITERATURE REVIEW

Hudson reported that Concrete Manufactured with a high percentage of minus 75 micron material will yield a more cohesive mix than concrete made with typical natural sand. Water absorption according to Misra V.N., the percentage of water required to produce mortar of same consistency is high for Robo sand as compared to river sand of same grading and same mix proportions. Srinivasa Rao P. has found that as percentage of stone dust increases the workability decreases in each grade of concrete to compensate the decrease in workability, some quantity of water and cement were added to get normal workability. The percentage of increase in water is in the range of 5% to 7%. Bhanuprabha, observed that the percentage of weight for M20, M25 and M30 grade manufactured sand concrete increased in 5% H2 S04 and 5% Na2 S03 acid compared to plain concrete and found to be as −30.3%, −24.4%, −22.9%; and −5.3%, −2.2%, −1.25% respectively. The negative sign indicates less reduction in weight loss that means the concrete is slightly more durable to sulphuric acid attack and sulphate attack when compared to river sand. Srinivasa Rao P., and Giridhar V. have observed that the concrete prepared using crusher stone dust was found to be relatively less workable than those compared with river sand. Hudson B.P. has reported that there is considerable increase in the compressive strength

511

when concrete is made of higher amount of minus 75micron material. Comparing with natural sand, Srinivasa Rao.P has observed an increase in compressive strength by 15% for M20 and M30 grades of concrete and by 12–13% for M40 grade of concrete when the concrete is made of stone dust. Giridhar V. has observed that, for the concrete made with crusher dust, there is an increase of 6% strength split tension and an increase of 20% strength in flexural tensile tension at 28 days for M20 grade design mix. Dinesh Khare has reported that flexural tensile stress of the concrete increases as percentage of Robo sand increases. 3 3.1

Table 1.

M50

Materials Cement Sand Manufactured sand Coarse aggregate Water/Cement ratio Sp (ml)

EXPERIMENTAL PROGRAMME

Table 2.

Concrete mixing

All the materials required for making the concrete were weighed as per the required proportion and kept ready for use before the mixing started. Cement, Fine aggregate and coarse aggregate were mixed thoroughly in the dry state. The water is added finally to the dry mixture. Superplasticiser (SP 337) were mixed in the water for M50 conventional and M50 grade using different proportions of Robo Sand. For M50 grade with varying percentages of Manufactured sand (0%, 25%, 50%, 75% and 100%.) of concrete, the cross sections of the beams adopted were 150 mm in width and 230 mm in depth. The length of the beam was 1500 mm. The mix proportions adopted are presented in Table 1. 3.3

R.S. 0%

R.S. 25%

R.S. 50%

R.S. 75%

R.S. 100%

1 0.90

1 0.675

1 0.45

1 0.225

1 –

–

0.225

0.45

0.675

0.90

2.27

2.27

2.27

2.27

2.27

0.30 350

0.30 350

0.30 350

0.30 350

0.30 350

Materials used

Ordinary Portland cement of 53 grade confirming to Bureau of Indian Standard is used in the present study. Steel For longitudinal tensile reinforcement, 12 mm and 10 mm dia tor steel and 6 mm dia mild steel were used for stirrups. The locally available river sand passing through IS sieve no: 480 was used as fine aggregate. Manufactured sand (‘Robo Sand’) used for the present study was procured from M/S Robo Silicon Pvt. Ltd. from their plant near Hyderabad Andhra Pradesh India. Test results are presented in Tables 2–4. 3.2

Mix proportions of concrete.

Physical properties of cement.

Tests conducted

Value obtained

Normal consistency Initial setting time Final setting time Specific gravity of cement Compressive strength 7 days with natural sand Compressive strength 28 days with natural sand Compressive strength 7 days with manufactured sand Compressive strength 28 days with manufactured sand Fineness Bulk density of cement

32% 85 minutes 300 minutes 3.15

Table 3.

41.2 N/mm2 58.523 N/mm2 44 N/mm2 61.365 N/mm2 2% 1.35 gm/cc

Properties of steel.

Diameter

Proof stress

Percentage elongation

Tor steel 12 mm dia Tor steel 10 mm dia Mild steel 6 mm dia

445 460 434 458 263 275

28 25 23

Compressive strength

Compressive strength of the various strength of concrete, the determination of compressive strength has received a large amount of attention because the concrete is primarily meant to withstand compressive stresses. Generally, cubes are used to determine the compressive strength. In the present study the cubes of 150 mm × 150 mm × 150 mm size are used. 3.4 Loading arrangements and testing procedure The beams were mounted on the wing table of 200 tons universal testing machine in North-South direction.

The beams were simply supported on steel rollers of 38 mm in diameter that were kept in specially prepared moulds. The specimens were tested under two-point transverse loading with simply supported span of 1200 mm. An iron roller was kept between the beam and UTM, at the two points of contact at a distance of 400 mm, so that the load transferred was like two point loads. After fixing the deflection and strain dial gauges to the beam specimens at the specified points, uniform load was applied.

512

Table 4. Properties of fine and coarse aggregates. Tests conducted Specific gravity of fine aggregate (Natural river sand) Specific gravity of fine aggregate (Manufactured sand) Specific gravity of coarse aggregate Compacted bulk density of fine aggregate (sand) Loose bulk density of fine aggregate (sand) Compacted bulk density of fine aggregate (Manufactured Sand) Loose bulk density of fine aggregate(Manufactured Sand) Compacted bulk density of coarse aggregate Loose bulk density of coarse aggregate Fineness modulus of fine aggregate (sand) Fineness modulus of fine aggregate (manufactured sand) Fineness modulus of coarse aggregate

4 4.1

Value obtained 2.53 2.65 2.75 17.59 kN/m3 17.03 kN/m3 18.36 kN/m3 17.03 kN/m3

Figure 1. Compressive strength variation with percentage replacement of Robo Sand for M50 grade concrete.

15.40 kN/m3 14.13 kN/m3 2.68 2.701 6.251

TEST RESULTS AND DISCUSSIONS Reinforced concrete beams

The experimental investigation has been carried out to study the behavior and strength of reinforced concrete beams with manufactured sand and natural sand as fine aggregate. For this purpose two beams each of concrete with manufactured sand and natural sand for each grade of concrete are tested under two point loading for pure bending and corresponding test results have been shown in Table 3. 4.2

Figure 2. Relationship between load vs deflection for M50 grade concrete.

Load vs deflection relation

The performances of reinforced concrete members can be reviewed from load vs. deflection curves. All the specimen curves are presented in Figures 2–3. 4.3

Compressive strength of concrete using manufactured sand as fine aggregate

It is observed from the Table that the seven days and 28 days cube compressive strength of concrete made with manufactured sand improved the cube compressive strength by 6.89%, 10.76%, 17.24% and 20.68% for M50 grade varying ratios respectively than the concrete made with natural sand. 4.4

Strength of manufactured sand concrete beams

Ultimate strength of reinforced concrete beams in terms of load carrying capacity and moment carrying

513

Figure 3. Relationship between load vs deflection for 100% Robo concrete.

capacity: It is noticed that the ultimate strength of manufactured sand concrete beams are slightly higher than those compared with conventional concrete beams, which is around 3 to 12% for M50 grade varying ratios of concrete respectively.

Table 5.

Table 9. Test results of ultimate moment, curvature and deflections.

Workability of concrete. Workability

Type of fine aggregate Robo sand (0%) Robo sand (25%) Robo sand (50%) Robo sand (75%) Robo sand (100%)

Table 6.

Slump (mm)

Compaction factor

55 48 44 41 38

0.87 0.83 0.80 0.75 0.72

Designation of beam specimens.

Beam designation

Percentage of reinforcement

No. of beam specimens

RS(0%) RS(25%) RS(50%) RS(75%) RS(100%)

2.04% 2.04% 2.04% 2.04% 2.04%

2 2 2 2 2

Details of reinforcement.

Beam designation

Reinforcement

RS 0% (M50 grade natural sand concrete)

4 no’s 12tor +2 no’s 10tor Main Reinforcement and 2L-6 mm mild @ 150 mm c/c for stirrups 4 no’s 12tor +2 no’s 10tor Main Reinforcement and 2L-6 φ 150 mm c/c for stirrups

RS 25%, RS 50%, RS 75%, RS 100% M50 grade manufactured sand concrete)

Table 8.

Compressive strength of concrete. Compressive strength (MPa)

4.5

Ultimate moment (kNm)

Curvature ×10−5

Deflection (mm)

R.S. 0% R.S. 25% R.S. 50% R.S. 75% R.S. 100%

165 170 175 180

64 66 68 70 72

30.23 11.16 10.12 9.57 8.98

4.96 5.6 5.55 6.12 6.12

studied. It was observed that the deflection response was slightly less for the same loading conditions when compared with natural sand concrete beams. All test results are presented in Tables 5–9. 5

M50 Grade, Letter R: Robo, S: Sand. Table 7.

Mix pro.

Ultimate load (kN)

M50

7 Days

28 Days

R.S. 0% R.S. 25% R.S. 50% R.S. 75% R.S. 100%

40 45 48 49 51

58 62 65 68 70

Structural behavior manufactured sand concrete beams

The load vs. deflection of manufactured sand concrete beams for a particular grade of concrete and percentage of reinforcement, right up to the failure, have been

CONCLUSIONS

– Workability of the M50 grade manufactured sand concrete observed to be 30% less compared to the conventional concrete. – The compressive strength of M50 grade concrete with varying percentages of (0%, 25%, 50%, 75%, and 100%) manufactured concrete improved the strengths by 6.89%, 10.76%, 17.24%, 20.68%, respectively. – The load carrying capacity and Moment carrying capacity of the RC beams of manufactured sand concrete obtained 3 to 12% when compared to conventional concrete. – Manufactured sand can be substituted in making structural grade concrete, as it is giving satisfactory performance.

REFERENCES Bhanuprabha., ‘‘Studies on use of manufactured sand as Fine Aggregate’’ M. Tech dissertation, submitted to JNTU, Hyderabad, 2003, India. Dinesh Khare., ‘‘Marvelous properties of Stone Crusher dust: A Waste bye-product of stone crushers,’’ National conference on Advances of construction material, 2002, Hamirpur (H.P.), India. Pages 189 to 195. Giridhar, V., ‘‘Strength characteristics of concrete using crusher stone dust as fine aggregate’’, 63rd Annual General meeting, 23rd December 2000, Hyderabad. Pages 11 to 15. Hudson, B.P., ‘‘Manufactured sand for concrete,’’ ICJ, August 1999. Misra, V.H., ‘‘Use of Stone dust from crusher in cement and sand Mortar’’ ICJ, August 1984. Srinivasa Rao, P., Seshagiri Rao, M.V. and Sravana., ‘‘Effect of crusher stone dust on some properties of concrete’’, National conference on advances in construction materials, 2002, Hamirpur. (H.P. India pages 196 to 201).

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Moisture permeability and sorption-desorption isotherms of some porous building materials R. Miniotaite Kaunas University of Technology, Kaunas, Lithuania

ABSTRACT: The durability of surface layers of enclosures (outside walls of buildings) is highly influenced by stresses that occur in the plane of contact between finishing materials and that of the enclosure. Variations of sorption moisture in the surface layer of enclosures result in deformations which have to be evaluated together with sorption-desorption processes in the construction and expressed in specific moisture-caused deformation of construction material. The dependence of the basic building materials of enclosure constructions upon change in material moisture with variation of relative humidity of the air was investigated by way of experiments. 1

2

INTRODUCTION

Adsorption-desorption processes taking place in construction materials basically indicate the nature of mechanical, physical and molecular bond between material and water or its vapour. In designing and construction practice, it is customary to relate adsorptiondesorption calculations with the air-dry materials. Usually, the permitted moisture content, estimated strength, thermal conductivity, swelling, shrinkage and other physical and mechanical values are indicated within the limits of material hygroscopic moisture (EN ISO 12571 2000; Cerny et al. 1996; Lentinen 1996). The following stages of material humidity state are predetermined by adsorption-desorption processes: – material moisture content when the walls of capillaries and pores of the material are covered with water molecules; – material moisture content when vapour is prevailing in pores and capillaries; – material moisture content at the beginning and during intense development of capillary condensation, i.e. such air conditions are shown when water and not vapour prevails in the capillaries of the material maintained. According to the physical data on the above stages, one can judge about the ratio of moistening and drying velocities, destruction of the material in the area of microcapillaries and micro-cracks, softening of the material as well as swelling-shrinkage (moisturecaused) deformations and ‘‘fatigue’’ of the material (Freitas et al. 1996; Carmeliet & Roels 2001; Bomberg et al. 2001).

515

INVESTIGATION METHOD

The data on adsorption-desorption processes found in literature are fragmentary, not exhaustive and characterize only the limit or typical values of the above mentioned properties of some traditional materials (Ramos & Freitas 2006; Hedenblad 1993; Carmeliet & Roels 1996). Standard methodics intended for determination of adsorption-desorption were not exhaustive. They were insufficient for investigation of adsorption-desorption processes and therefore had to be essentially improved in the course of investigation. The main point of the developed adsorptiondesorption method is carrying out of experiments in three stages (Miniotaite 1999a, b, 1998, 2004). At the first stage, sorption isotherm of the investigative material is drawn at all 7 points of ambient air humidity φ. At the second stage, a reverse action is applied, i.e. desorption isotherm of the same specimen is drawn. At the third stage, desorption process is at some moment discontinued; coming back to sorption, physical coordinates of such moment are fixed: relative humidity (RH) φ of ambient air and specimen’s moisture content u. When adsorption-desorption experiment at 20◦ C temperature is completed, the investigated materials are placed in a vessel and dried again; analogous experiment is carried out at 5◦ C. It was found in literature only approximate data on sorption moisture content of materials in case of temperatures below zero, even though surface layers of walls are under action of the environment of temperatures below zero for a long period of time (on the

average, 100.3 days per year in Lithuania). With the data on adsorption-desorption processes at 20◦ C and 5◦ C, rather precise recalculation of sorption moisture content at low temperatures is possible on the basis of Clausius-Clapeyron’s equation: d ln p L =− 2 dT RT

(1)

The integral of the above equation within the temperature range from T1 to T2 when L = const, permits to calculate the pressure p1 : ln p1 (T1 ) = ln p2 (T2 ) −

L R



1 1 − T1 T2

 (2)

where: T = temperature, K; T1 < T2 ; L is heat of phases change, J/mol. In general case it consists of heat of evaporation, ice melting heat and adsorption heat. Ice melting heat is used only for temperatures below 0◦ C. R is universal gas constant, J/(K · mol); p is saturated vapour pressure, Pa; p1 < p2 . Sorption moisture content u are calculated of p values and relative humidity φ values. Adsorption-desorption investigations of the materials were carried out at 20◦ C and 5◦ C according to the improved methods of sorption investigation. Then sorption isotherms were recalculated in isotherms at 10◦ C, 0◦ C and −10◦ C. 3

RESULTS AND ANALYSIS

Graphic and summarized passports-cards of adsorption-desorption investigations of all tested materials are made up indicating a structural group of the material corresponding to similarity of moisturecaused deformations, the name of material and the main adsorption-desorption parameters. The passports-cards constitute a respective data bank on adsorption-desorption of construction material. A typical example of such a passport-card is given in Figure 1. It should be noted that at temperatures below zero significant difference between experimental results and literary data is observed. According to literary sources, material moisture content u increases with drop in temperature whereas according to experimental results, at high values of RH of the air and temperature below 0◦ C, material moisture content u is substantially less than that of isotherms at temperatures above 0◦ C. Generalizing the nature of changes in adsorptiondesorption isotherms it was determined that no distinct limits exist between individual forms of material humidity state—one humidity form gradually transfers to another and therefore consecutively expressed isotherms of adsorption and desorption possess a

Figure 1. Adsorption-desorption isotherms of lime cement plaster, ρ = 1700 kg/m3 : experimental,  = 20◦ C; calculated,  = 10◦ C, 0◦ C, −10◦ C.

flexible ‘‘S’’—shaped view. Each material possesses only for it characteristic such sorption curve. More detailed examination of the curves showed that triple moisture-material link exists. The length corresponding to ambient air RH φ from 0% to ∼10% is of a slowing rise. In this case, water vapour molecules cover the walls of material pores and capillaries initially with the layer of monomolecular thickness and then with the layer of multi-molecular thickness until a water (aqueous) film is formed. Forces of molecular attraction act between the material and vapour molecules, therefore, wet films acquire some properties of a solid body: they do not move or freeze easily and would not evaporate at standard drying temperature  = 105◦ C. The length corresponding to ambient air RH φ from (10–12)% to 45% is still of a slowing rise. Thus adsorbed, vapour molecules maintain thermodynamic equilibrium, move easily. If the pressure of water vapour in capillaries of the material is higher than ambient vapour pressure, they get evaporated at 105◦ C. The above processes also take place in the length of a rising straight line from 45% to 55%. The length corresponding to ambient air RH φ from 55% to 100% is of a quickened rise, i.e. material moisture content growth picks up speed. In this sorption area the thermodynamic equilibrium of vapour molecules is destroyed, because a part of vapour is condensed in capillaries and capillary condensation starts. In the

516

length corresponding to ambient air RH φ from 58% to (75–80)% the above action is slow; in that wherein humidity exceeds 80% the process is intense or very intense. At this time, a part of the moisture contained in the material is of a liquid aggregative state. The above must be taken into consideration when calculating moisture content and heat exchange in the walls. Thermal conductivity and moisture-caused deformations of the material significantly increase in this sorption area. The investigations indicated that in case of many materials initial point of intense capillary condensation was rather distinct, however, in case of some materials capillary condensation increase is approximately proportional to the increase in ambient RH (φ, %). Material moisture content values (u, %) of the end of intense capillary condensation and on the contrary—of the beginning of intense evaporation (drying) practically coincide irrespective of hysteresis size and ambient temperature with exception of negative (below zero) temperatures. 3.1 Group of fine grained structure articles The above group consists of cement plaster, lime cement plaster, masonry mortar, silicate brick articles. The value of maximum sorption moisture content (umax ) is up to 6% (silicate brick). Figure 1 shows adsorption-desorption isotherms of lime cement plaster. The maximum adsorption-desorption hysteresis u up to 120% in comparison with sorption isotherm. It corresponds to the range of RH of air φ = (75–80)% (fast moistening, slow drying). The increase of material moisture content from temperature is up to 0.30 mass % when temperature drops from 20◦ C to −10◦ C; RH φ = (40–60)%. 3.2

Group of coarse grained structures articles

The above group consists of concrete and expandedclay concrete articles. The bonding substance is cement. The value of maximum sorption moisture content (umax ) is up to 8% (expanded-clay concrete). Figure 2 shows adsorption-desorption isotherms of concrete. The maximum adsorption-desorption hysteresis u up to 95% in comparison with sorption isotherm. It corresponds to the range of RH of air φ = (40–70)% (fast moistening, slow drying). The increase of material moisture content from temperature is up to 0.50 mass % when temperature drops from 20◦ C to −10◦ C; RH φ = (60–75)%. 3.3

Figure 2. Adsorption-desorption isotherms of concrete, ρ = 2400 kg/m3 : experimental,  = 20◦ C; calculated,  = 10◦ C, 0◦ C, −10◦ C.

bonding material has some specific features. When cement is used as a bonding material, hysteresis of adsorption and desorption is 2 times larger than in case of lime use. The most favourable air humidity (φ) environment for the hysteresis of cement articles is 75%; in case of lime articles 40–60%. The above phenomenon is also observed in case of the articles of a different structure containing lime: use of lime decreases the hysteresis and the most favourable environment for its formation is more dry (φ ≤ 60%) environment. Summary of the investigation results: – maximum sorption moisture value (umax )—up to 20%; – maximum adsorption-desorption hysteresis in case of cement sub-group is in 75–80% RH (φ) environment—uφ up to +80%; in case of lime sub-group—40–60% φ in the environment— uφ —from +9 to +20% (fast drying); – the influence of temperature in the environment of u —40–80% RH (φ) is practically the same and does not exceed +12% (insignificant). Figure 3 shows adsorption-desorption isotherms of porous concrete. 3.4

Group of articles of porous structure

Group of ceramics

The value of maximum sorption moisture content (umax ) up to 2.5% (Fig. 4).

This group consists of sub-groups of porous concrete and porous silicate articles. The use of cement and lime

517

Figure 5. Adsorption-desorption isotherms of spruce, ρ = 500 kg/m3 : experimental,  = 20◦ C; calculated,  = 10◦ C, 0◦ C, −10◦ C.

Figure 3. Adsorption-desorption isotherms of porous concrete, ρ = 600 kg/m3 : experimental,  = 20◦ C; calculated,  = 10◦ C, 0◦ C, −10◦ C.

The increase of material moisture content from temperature is up to 0.14 mass % when temperature drops from 20◦ C to −10◦ C; RH φ = (50–65)%. 3.5

Group of organic structure articles

This group consists of wood articles. The value of maximum sorption moisture content (umax ) up to 28% (Fig. 5). The maximum adsorption-desorption hysteresis u up to 25% in comparison with sorption isotherm. It corresponds to the range of RH of air φ = (30–80)% (fast moistening, slow drying). The increase of material moisture content from temperature is up to 1.8 mass % when temperature drops from 20◦ C to −10◦ C; RH φ = (30–80)%. 4

Figure 4. Adsorption-desorption isotherms of ceramics, ρ = 1700 kg/m3 : experimental,  = 20◦ C; calculated,  = 10◦ C, 0◦ C, −10◦ C.

The maximum adsorption-desorption hysteresis u up to 225% in comparison with sorption isotherm. It corresponds to the range of RH of air φ = (60–75)% (fast moistening, slow drying).

CONCLUSIONS

It was proved by experimental investigations that characteristic points of sorption isotherms indicate the beginning or the end of changes in the essential physical links of material and moisture. Specific parameters of the above changes—the beginning and the end of the process, intensity and linear moisture-caused deformation of the material, moisture content—are characteristic of each structural group of construction materials. The highest hysteresis in materials of fine structure and porous concrete exists before the beginning of an intense process of capillary condensation at relative humidity of the ambient air φ = (70–80)%. In case of

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articles of fine-grained structure containing (30–35)% and more lime in its bonding material, the mark of the highest hysteresis shifts to the left by (10–15)%; in case of coarse-grained materials—by (15–20)% and in case of materials of organic nature by (40–50)%. Temperature influence upon adsorption-desorption process in case of all materials is the highest at relative humidity of the air close to humidity of the beginning of capillary condensation. The absolute value of temperature influence depends on the nature of the framework of material. The above influence is highest in case coarse-grained, porous and organic materials (30–35)%. In all cases, sorption moisture increases with decrease in temperature. At high relative humidity values and temperature below 0◦ C, material moisture u is lower than in isotherms at temperatures above zero. The beginning of an intense capillary condensation can be expressed rather clearly, or start gradually and continue in a wide range of ambient humidity. The end of an intense capillary condensation or the beginning of intense capillary evaporation does not depend upon the size of hysteresis or temperature. REFERENCES Bomberg, M.; Haghighat, F.; Grunewald, J. & Plagge, R. 2001. Capillary transition point as a material characteristic for HAM models. In Proc. 4th Int. conf. on IAQ, ventilation and energy conservation in buildings, 1: 755–762. Carmeliet, J. & Roels, S. 1996. Moisture transfer and durability of open porous media. Proc. of 4th Symposium on Building Physics in the Nordic Countries, 2: 587–594. Finland. Carmeliet, J. & Roels, S. 2001. Determination of the isothermal moisture transport properties of porous building materials, Journal of Thermal Envelope & Building Science 24: 183–210.

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Cerny, R.; Drchalova, J.; Hoskova, S. & Toman, J. 1996. Inverse problems of moisture transport in porous materials. In Proc. of Second ECCOMAS Conf. on numerical methods in engineering: 664–670. The Netherlands. EN ISO 12571. 2000. Building materials. Determination of hygroscopic sorption curves. Freitas, V.P.; Abrantes, V. & Crausse, P. 1996. Moisture migration in building walls—analysis of the interface phenomena. In Building and Environment, 31(2): 99–108. Hedenblad, G. 1993. Moisture permeability of some porous materials. In Proc. of 3rd Symposium on Building Physics in the Nordic Countries, 2: 697–702. Copenhagen. Lentinen, T. 1996. Capillary moisture transfer in combined porous building materials. In Proc. 4th Symposium on Building Physics in the Nordic Countries, 2: 483–490. Finland. Miniotaite, R. 1998. The durability of finish layers of external walls of buildings. In Proc. Conference on the subject of Construction and Architecture: 248–253. Kaunas: Technology. Miniotaite, R. 1999a. Compatibility of finishing layer and external surface of buildings’ walls from the standpoint of durability. Doctoral Dissertation, Kaunas: Technology. Miniotaite, R. 1999b. Compatibility of finishing layer and external surface of buildings’ walls from the standpoint of durability. Summary of the Thesis for a doctor’s degree. Kaunas: Technology. Miniotaite, R. 2004. Hygric properties of several building materials. In Proc. The 8 International conference Modern Building Materials, Structures and Techniques: 114–119. Vilnius: Technika. Ramos, N. & Freitas, V. 2006. Evaluation strategy of finishing materials contribution to the hygroscopic inertia of a room. Research in Building Physics and Building engineering—Proceedings of the Third International Building Physics Conference, Concordia University, Montreal, Canada: 543–548.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Nested ANOVA model applied to evaluate variability of ready-mixed concrete production C. Videla School of Engineering, Pontificia Universidad Católica de Chile, Santiago, Chile

C. Imbarack Lafarge Centre de Recherche (LCR), France

ABSTRACT: Statistical Nested ANOVA Model was applied to evaluate different components of compressive strength variability (variance) of ready-mixed concrete production. Three components of the total variation were assessed: ‘‘between-batch’’, ‘‘within-batch’’ and ‘‘within-test’’. Dry-batch production method was used, and 22 truck-mixers were controlled during two months. It was concluded that the main source of concrete strength variability was the ‘‘within-batch’’ component (51.5% under normal delivery conditions without the addition of water on site). The ‘‘between-batch’’ variability resulted to be the second contributor to the strength variation (41.1%). The ‘‘within-test’’ variability (attributable to the control laboratory) was the smallest contributor (7.4%). The results clearly show that the main cause of concrete strength variability was a deficient concrete homogenization within a batch during the truck mixing process. Therefore, modifications of the mixing process can be implemented to improve concrete uniformity and the same analysis could be used to assess the outcome. 1

INTRODUCTION

The ready-mix concrete industry has based its control and quality assurance systems on the historical evolution of these principles. Thus, it is reasonable to expect that elements of statistical control of processes and continuous improvement will be shortly included, looking for the total quality management standard. In this context it is required that the ready-mixed concrete producers evaluate the uniformity of their processes and products, to efficiently guarantee the fulfillment of the specifications. At the moment, there are methods to evaluate the uniformity of a concrete production, describing the quality of the mix (ACI 2002; ASTM 1997) and/or the efficiency of the mixer equipment (Charonnat 1996; Day 1999). Also, there are methods that evaluate the variability of the production and of the laboratory using the strength of the concrete as a control variable (Dewar & Anderson 1992; Ferraris 2001). However, the available methods do not allow evaluating simultaneously the variability between production units (between-batch), within the units of production (within-batch) and within concrete test specimens (within-test or laboratory). In order to determine these components of the variability, the statistical method of Hierarchic Analysis of Variance can be used (Montgomery 1984), adapting the model to the production of ready-mixed concrete. The application of the method, under an appropriate design of experiment, allows the ready-mixed concrete producer to quantify each component of the variability

521

of its production and therefore to generate actions to improve its efficiency and effectiveness. The method can also be used to investigate the controlled effect of one or more factors on the uniformity of the production or to quantify the magnitude of the improvements introduced to the system.

2

RESEARCH SIGNIFICANCE

The study validated the applicability of the Nested ANOVA Model as a useful statistical technique to quantify the components of the strength variance of ready-mixed concrete, and as a tool for determining potential industrial continuous improvements.

3

OBJECTIVE AND SCOPE

The main objective of the study was to apply the statistical Nested ANOVA Model, as a procedure to quantify the variance components of the strength of concrete in an actual Chilean industrial production, according to Equation 1: 2 σtotal = σ12 + σ22 + σ32

(1)

2 where σtotal = total variance; σ12 = ‘‘between-batch’’ variance; σ22 = ‘‘within-batch’’ variance; σ32 = ‘‘within-test’’ variance.

The ‘‘between-batch’’ variance represents the variability between production units and is due to variations of the properties of raw materials, batching and mixing process, slump adjustment, weather conditions, etc. It also includes the same factors affecting ‘‘within-batch’’ variance. The ‘‘within-batch’’ variance represents the variability inside the batch. The variability between samples taken from de same batch is affected mainly by the loading and mixing processes (mixing time, mixer rotation speed, loading materials sequence, etc.). The ‘‘within-test’’ variance represents the variability of the measurements; in this case, the variation between strength tests results of concrete specimens of the same sample. 4

BACKGROUND

In an industrial ready-mixed concrete production, each concrete batch delivered to the customer according to his requirements is considered as a basic manufacture unit. Typically, these batches are dispatched in a mixer truck or other means of transportation in a volume between 2 and 7 m3 . In concrete technology the uniformity concept has traditionally been associated to the homogeneity obtained inside each produced concrete batch. However, when it is required to evaluate the time-dependent uniformity, the concept extends to the variation of successive produced batches. Nevertheless, there is no current analytical method allowing to simultaneously evaluating the variance components as shown in Equation 1. When the study of the withinbatch variance is of interest, the simplest way of evaluation is through the comparison of some concrete properties determined to different samples taken from the same batch. These concrete properties are considered as quality indicators. For instance, NCh 1789 (INN 1986) indicates to compare the fresh density, the workability (slump), the strength at 7 days, the gravel percentage, the mortar density without air, and the air content of two concrete samples taken from the same batch. The acceptable variation range of these properties is specified. This procedure has been standardized for example by ACI (2002), ASTM (1997), and INN (1986). In addition, there are practical methods to compare the strength and other concrete properties, between samples taken from a same batch; and, also between companion specimens (cubes or cylinders) from the same sample (Dewar & Anderson 1992). The concept of uniformity (homogenity) of the mix is also used to define the efficiency of mixing process (Ferraris 2001). However, it is argue that the comparison of concrete properties from two or more samples coming from same batch has the disadvantage of being an indirect method, since the homogenity of the mixed is not directly shown, but it is assumed that a potential non-homogenity will affect the evaluated properties. Furthermore, it is possible that the

chosen measuring methods could not be sufficiently sensitive to detect local changes in the composition of the concrete. According to Charonnat (1997) a direct indicator of the homogenity reached during the mixing is to measure the composition of each sample, determining the distribution of the solids components, such as aggregates, mineral additions and cement paste. Neville (1996) also states that empirical results indicate that measurement of the cement content is the best evaluation of the uniformity reached by a mixture. Nevertheless, for the customer who receives the ready-mixed concrete on the job site, the uniformity is given by the properties of quality (performance) of the concrete (mainly strength) and it is not necessarily of his interest to verify the efficiency of the mixing by means of internal parameters; but it is of interest to the producer. Charonnat (1997) proposed a method to classify the efficiency of the mixing of concrete (mixing equipment or process) according to the composition of the mix. To determine the efficiency, and then to assign a category, the use of the variation coefficient of the results of four measuring parameters is proposed. On the other hand, Ferraris (2001) proposed a method, called the ‘‘Hybrid’’ method, which combines the evaluation by composition and by performance. In this later method the following properties are selected: distribution of the cement content and of the fine and coarse aggregate in the mixture, variation of the strength and of the consistency with time (measured by the slump test). Because of the many factors affecting the mixing performance, they must be taken into account to carry out an efficiency evaluation. Some of them are: mixing time, rotation rate, type and characteristics of the mixing equipment, wear of the blades, batching methodology or loading sequence and concrete volume with respect to the optimum capacity of the mixer (Charonnat 1997; Ferraris 2001; Day 1998; Dewar & Anderson 1992). The different so far analyzed methods deal with the ‘‘within-batch’’ uniformity of ready-mixed concrete. However, all of them lack of a statistical background allowing interpreting the results based on a significance level. Furthermore, the reviewed methods can not be directly applied for comparing the ‘‘between-batch’’ and the ‘‘within-batch’’ uniformities. Therefore ready-mixed concrete producers can not determine what part of the total variability is due to the dispersion within the batch, between-batch and the testing (laboratory). Then, the different methods that have been presented to assess the uniformity of a batch of concrete are not suitable to evaluate the evolution of the uniformity of concrete production over time. Generally, the ‘‘between-batch’’ uniformity is evaluated calculating the standard deviation of the ready-mixed concrete strength at a certain age. Other

522

properties, such as slump, are only used as acceptance criteria and are associated to uniformity only in special cases.The factors affecting the between-batch uniformity are the same affecting within-batch uniformity, because batches are successive replicas of the same process. Additionally there is time-dependant and seasonal factors, like: quality of materials, production processes (people, equipments) weather conditions, etc. ACI 214R-02 (ACI 2002) recommends a method that allows evaluating the uniformity of concrete production decomposed as the variance of the producer and the laboratory variability, as given by Equation 2: 2 σtotal = σ12 + σ22

B, are similar but non-identical for different levels of the other factor, for example A. Such arrangement is known as nested design of B in A. For the present analysis the levels 1 and 2 of the ‘‘sample’’ factor are similar but non-identical for each ‘‘batch’’ factor. Thus, a nested arrangement of the sample factor within each batch is obtained. The same occurs for the different test specimens, which are nested under each particular sample. Then, the variability of the average value of the batches represents the ‘‘between-batch’’ standard deviation, the variability between samples of a same batch constitutes the ‘‘within-batch’’ standard deviation and the dispersion between companion test specimens corresponds to the ‘‘laboratory’’ standard deviation. This kind of statistical model is suitable for the analysis of concrete production, because it recognize that the total variation comes from ‘‘eachbatch’’ and ‘‘successive-batches’’, and also allows considering the source of testing (variation of laboratory), matching with Equation 1. Figure 1 show the arrangement of batches, samples, and tests specimens, according to a nested model. To estimate the ‘‘within-batch’’ variation, at least two samples of concrete should be obtained from each batch (each truck). Also, at least two test specimens (recommended three cubes or cylinder specimens) should be cast from each concrete sample to estimate ‘‘within-test’’ variation. Due to the destructive nature of the compressive strength tests one measurement or result correspond to the strength of one specimen; then, the test variation represents the dispersion between companion specimens from the same concrete sample. The linear statistical model defined in the previous scheme, is the three stages hierarchic model expressed by Equation 3:

(2)

2 where, σtotal = total variance of the population of strength results; σ12 = producer variance or betweenbatch variability as a result of the material’s variation and production processes (load, mixed, etc.); σ22 = laboratory variance due to the variability between companion test specimens caused by variations in sampling, preparation, compaction, transporting, curing and testing of test specimens. The weakness of Equation 2 is that the ‘‘production variability’’ is not split between its components, and then the variability due to mixing process is hidden. Also, this method does not correspond to a proper variance analysis and it is based on statistical approximations. Popovics (1998) and R.C.B. (1995) show some results about concrete strength variance analysis, but no reference to statistical method was found.

5

STATISTICAL METHOD: NESTED ANOVA

The mathematical statistical model that is necessary to solve to determine the components of Equation 1 is the Hierarchic Anova Model (or Nested) of three levels, that is schematically shown in Figure 1. The method is very well described by Montgomery (1984); for this reason in this section a very brief explanation will be provided and the reader is directed to this publication for further information. In certain multi-factor experiments the levels of a factor, for example factor

Figure 1. design.

Outline of sampling arrangement in nested

Yijkl = μ + τi + βj(i) + γk(ij) + ε(ijk)l

(3)

with i = 1, 2, . . ., a; j = 1, 2, . . ., b; k = 1, 2, . . ., c; l = 1, 2, . . ., n and where μ = corresponds to the global average; τi = the effect of the i level of the batch factor; βj(i) = the effect of the j level of the withinbatch factor hierarchic or nested under each bath level; γk(ij) = the effect of the k-th hierarchized test specimen under each within-batch factor level j nested in batch i; ε(ijk)l = error term. Because of the destructive nature of the strength tests providing only one observation (result) by tested specimen, to calculate the Nested Anova Model of three stages, this must be transformed to a similar model but of two stages. This transformed model allows performing the homogeneity of variance test between groups and to have a degree of freedom of the error term different to zero. The solution of the model is obtained by the Least Squares Method and the ‘‘Anova Table’’ is determined to analyze the statistical significance of each factor.

523

The hypotheses and suitability of the model are verified by the same procedures used for any analysis of variance model (Montgomery 1984). The basic hypotheses on the population of results are the normal distribution of the response variable and the equality of the variance. The suitability of the model is verified by the analysis of the residues, that basically must distribute N (0, 1) (for standardized residues) and present a random pattern when compared to the factors and the estimated value of the response variable. In addition, it must be considered that the data must be randomly determined. Solving the model will allow to determine the statistical significance of the factors, the standard deviation (or variance) contributed to the total standard deviation by each one of the factors and the relative proportion of each factor on the total variability. Several statistical software packages are available to solve the statistical model. In this case, SPSS® 11 was used.

Table 1.

Raw material

Amount

Unit

Cement Sand-1 (0/5) Sand-2 (0/10) Gravel (5/20) Total water WRDA

390 709 117 980 212 1.445

kg/m3 kg/m3 kg/m3 kg/m3 kg/m3 l/m3

Slump target

70 ± 20

mm

discharge. Tests of fresh and hardened concrete were performed: slump, fresh and hardened density, air content, and strength (three cubes at 28 days). Only the strength test results are reported and analyzed in this document.

7 6

EXPERIMENTAL PROGRAM

The experimental program was planned considering the sampling of an actual ready-mixed concrete production. The following recommendations with respect to the variables that must be considered were followed (Imbarack 2004): 1. The concrete type (product) was fixed. The mixture proportions and components materials were constant, even the inherent variability of raw materials. This is very important because the statistical analysis is based on strength variation, and no variation was induced changing the mix dosage of the concrete. The amount of water could be variable, since in practice is adjusted based on the concrete slump, within limits specified and/or accepted by the contractor. 2. All concrete was produced (batched, mixed, etc.) and delivered from same production facility (same production conditions over time) to the same job site (pre-cast industrial facility). 3. Mixer trucks were randomly selected from the concrete producer fleet, and no special production intervention was carried out. 4. Sampling of the concrete at the job site was performed by the same operator during all experience. Also, the same accredited laboratory carried out all the testing. The design of the concrete mixture is presented in Table 1, corresponding to a normal concrete having a 45 MPa specified 28 days concrete cube strength. The sampling period was extended for two months, totaling 22 batches (trucks). For each sampled concrete batch two samples were taken, one at the 20% and the other at the 80% of the total volume during

Concrete mixture proportions.

RESULTS AND ANALYSIS

As was previously explained, at least two concrete samples should be taken to perform the analysis of the statistical model. For this research two samples were taken during the discharge of concrete, corresponding to 80% and 20% of the total volume. During the experimental program three cases occurred with respect to slump adjustment: 1. Slump adjustment type 0. This was the normal case when the slump satisfied the specify value and it was accepted to pour the concrete without any addition of water. 2. Slump adjustment type 1. In this case the adjustment of the slump with water was done before the discharge of the concrete. The concrete was properly re-mixed before discharge. 3. Slump adjustment type 2. Due to delays during concreting operation, the adjustment of the slump with the addition of water was carried out during the discharge and between the sampling stages. Then, the second concrete sample was affected and its water content was higher. All the samples taken and the test results obtained were considered valid since the concrete was accepted and placed without difficulty. Figure 2 shows the variation of the concrete compressive cube strength of the sampled batches, identifying samples 1 and 2, and the type of slump adjustment carried out on site. Each point on the graph represents the average strength of three specimens. It can be clearly observed that there is dispersion between samples of the same batch and between batches. As a consequence of the slump adjustment cases, all statistical analysis were performed considering the data grouped in three stages or data groups. The first

524

Concrete compressive strength at 28 days (MPa)

Table 2. Anova table for nested design, applied to three calculation groups.

65 60

Source of Group variation

55

Degrees of Mean freedom squares

F value

Sign

50 45 40 35

Sample #1: black color Sample #2: white color : Slump adjustment type 0 : Slump adjustment type 1 : Slump adjustment type 2

1st

Constant 1 Between-batch 13 Within-batch 14 Within-test 56

235479.24 43.34 17.17 0.78

5432.94 2,524 22,025 –

0.000 0.049 0.000 –

2nd

Constant 1 Between-batch 17 Within-batch 18 Within-test 72

299094.19 36.26 21.33 0.65

8249.39 1.700 32.942 –

0.000 0.137 0.000 –

3rd

Constant 1 Between-batch 21 Within-batch 22 Within-test 88

368893.03 31.27 20.29 0.59−

1798.69 1.541 34.602 –

0.000 0.161 0.000 –

30 0 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16 17 18 19 20 21 22 Batch # (time)

Figure 2. Concrete compressive average cube strength of samples in each batch.

stage considered data of samples with slump adjustment type 0; the second stage added the data with slump adjustment type 1; and lastly the third stage incorporated previous data plus data corresponding to slump adjustment type 2. The Nested ANOVA Model was solved according to statistical methods (Montgomery 1984); in particular, a standardized residual analysis was performed. The hypothesis of the model with respect to the dependent variable was verified, for each calculation stage and the results of normality test K-S indicate that the results come from a normal distribution. The model was considered suitable to the objective, and a detailed analysis can be found in Imbarack (2004). Table 2 presents the calculated Anova for each one of the stages. This table clearly shows that the ‘‘within-batch’’ variability resulted to be significant in all stages (sig. Al2 O3 > Fe2 O3 . The Al2 O3

and Fe2 O3 are replaced by their SiO2 equivalents to combine one ingredient of SiO2(eq) (Malvar & Lenke 2006), as shown in Equation 2.

Table 2. Crushed aggregate gradation for mortar bars. Sieve size

CaOeq = CaO + 0.905 ∗ Na2 Oeq + 1.391 ∗ MgO + 0.70 ∗ SO3

(1)

SiO2eq = SiO2 + 0.589 ∗ Al2 O3 + 0.376 ∗ Fe2 O3 (2)

2

Passing

Retained on

Mass(%)

4.75 mm (No. 4) 2.36 mm (No. 8) 1.18 mm (No. 16) 600 μm (No. 30) 300 μm (No. 50)

2.36 mm (No. 8) 1.18 mm (No. 16) 600 μm (No. 30) 300 μm (No. 50) 150 μm (No. 100)

10 25 25 25 15

RESEARCH SIGNIFICANCE Table 3.

The amount of fly ash required to suppress ASR varies depending on the chemical compositions of fly ash and aggregate. This study provides an insight in selecting minimum Class F fly ash dosages that are effective to suppress ASR expansion of reactive aggregates at different immersion ages.

3

EXPERIMENTAL PROGRAM

The raw materials utilized in this study consisted of reactive aggregates, ASTM Type V Portland cement, and Class F fly ash. The aggregates were obtained from three different quarries within the State of Nevada; two from the Southern Nevada (SN-1 and SN-2), and one from the Northern Nevada (NN-1). The compositions of Portland cement and Class F fly ash used in this investigation are shown in Table 1. The test specimens were prepared based on the proportions of the ASTM C 1260, which included crushed coarse aggregate with gradation as shown in Table 2. The mortar bars contained water-to-cementitious materials ratio (by weight) of 0.47, and the graded aggregates to total cementitious materials ratio of 2.15. The absorption and moisture content of the aggregates were accounted for in determining the actual mixing water content of the mixture. The mixing procedures were conducted in according to ASTM C 305. The mortars were molded within a total mixing time of not more than 2.25 minutes. After 24 hours of moist curing, the specimens were demolded, and initial readings were taken before immersing in tap water at 80˚C for 24 hours for which the zero readings were recorded. Afterward, the mortar bars were submerged in 1N NaOH in an Table 1. Chemical compositions of Portland cement and Class F fly ash. SCMs SiO2 Al2 O3 Fe2 O3 CaO MgO Na2 O∗ SO3 LOI Cement 21 3.6 3.4 63.1 4.7 0.42 2.6 1.3 Fly Ash 57.8 21.7 5.1 7.4 – 0.30 – 0.2 ∗ Na O = 2 eq

Na2 O + 0.658 ∗ K2 O.

Mixture constituents of mortar bars.

Agg. ID

Fly ash

Gradded agg.a

Absorption (%)

Cementa (gm)

Class F fly asha

SN-1

0 15 20 25 30

2310 2310 2310 2310 2310

2.18

1026.7 872.7 821.3 777.0 718.7

0.0 154.0 205.4 256.7 308.0

SN-2

0 15 20 25 30

2310 2310 2310 2310 2310

3.48

1026.7 872.7 821.3 777.0 718.7

0.0 154.0 205.4 256.7 308.0

NN-1

0 15 20 25 30

2310 2310 2310 2310 2310

3.12

1026.7 872.7 821.3 777.0 718.7

0.0 154.0 205.4 256.7 308.0

a

Weights were calculated on 7-bar batch size.

air-tight plastic container held in an oven maintaining the temperature of 80˚C. Subsequent readings were taken at the immersion ages of 3, 6, 10, 14 days and, thereafter, one reading per week until the 98-day immersion age was reached.

4 4.1

RESULTS AND DISCUSSIONS ASR expansion of various Class F fly ash dosages

The results of the study clearly revealed the influence of the selected Class F fly ash content in reducing the adverse effect of three reactive aggregates. A typical ASR expansion as related to the immersion ages and fly ash contents is shown in Figure 1. The 14-day expansion of SN-2 aggregate (0.247%) was reduced to 0.094, 0.056, 0.041 and 0.031% when a portion of Portland cement was replaced by 15, 20, 25 and 30%, respectively, with Class F fly ash. For the same aggregate, the expansion progression of 15% fly ash at the

534

immersion ages of 28, 56 and 98 days were 2.3, 5.5 and 9.7 times the expansion at 14 days, respectively. Using 0, 15, 20, 25 and 30% fly ash in the mortar bars of SN-2 aggregate, the 98-day expansion was increased by 5.6, 9.7, 14.3, 16.6 and 17.3 times the 14-day expansion, respectively. The characteristics of the remaining two aggregates (SN-1 and NN-1) also followed a similar pattern to that of the SN-2 aggregate. The influence of four dosages of Class F fly ash to reduce the ASR-related expansion of the trial reactive aggregates was also expressed in terms of the Reduction in Expansion (RIE) of the untreated mortars (having no fly ash content) at various immersion ages. A typical reduction in expansion as related to immersion ages and fly ash contents for SN-2 aggregate is illustrated in Figure 2. It can be seen that the RIE increased with increases in the amount of fly ash content for all immersion ages. For each of the trial fly ash content, the RIE at the immersion age of 6

days was highest and then gradually decreased with increases in the test duration. As seen in Figure 3 and with increases in fly ash from 15 to 30% with increments of 5%, the 14-day expansion was reduced by 61.9, 77.3, 83.4 and 87.4% for SN-1 aggregate, 67.7, 80.6, 87.6 and 92.3% for SN-2, and 71.9, 82.7, 89.9 and 95.1% for NN-1 aggregate, respectively. The reduction in expansion was also a function of immersion age which varied at different fly ash dosages. Figure 4 illustrates the 98-day RIE for each selected aggregates for different fly ash contents. As can be seen, when 15, 20, 25 and 30% of Portland cement were replaced Class F fly ash, the 98-day expansion was reduced by 34.0, 41.9, 50.8 and 61.1% for SN-1 aggregate, 37.3, 47.7, 59.3 and 67.2 for SN-2 aggregate, and 33.6, 44.0, 58.5 and 68.7% for NN-1 aggregate, respectively. The ASR-related cracks on the surface of the mortar bars were also examined. Figure 5 shows the influence of various fly ash dosages on the ASR-induced cracks

0% 25%

1.0

15% 30%

100

20%

Reductionin Expansion (%)\

Linear Expansion (%)

1.2

0.8 0.6 0.4 0.2 0.0 0

14

28 42 56 70 84 Immersion Age (Days)

90 80

SN-1 SN-2 NN-1

70 60 50 40 30

98

15

20

25

30

Fly Ash Content (%) Figure 1. Influence of various fly ash dosages on ASR expansion of SN-2 aggregate.

Figure 3.

100

90

15% FA

20% FA

25% FA

30% FA

Reductionin Expansion (%)

100 Reductionin Expansion (%)\

Reduction in expansion at 14 days.

80 70 60 50 40 30 20 0

14

28

42 56 70 84 Immersion Age (Days)

98

90

SN-1 SN-2

80

NN-1

70 60 50 40 30

112

15

Figure 2. The reduction in expansion of SN-2 aggregate containing various fly ash dosages and immersion ages.

535

Figure 4.

20 25 Fly Ash Content (%)

Reduction in expansion at 98 days.

30

of the mortar bars containing NN-1 aggregate. Using 20% fly ash in the mortar bars of NN-1 aggregate, the severe cracks of the control specimen (having no fly ash content) were significantly reduced, and the cracks were completely mitigated when 30% of Portland cement was replaced with the Class F fly ash (Fig. 5c). Table 4 documents the levels of alkali-silica reactivity of the selected aggregates based on the suggested expansion criteria. As can be seen, all three 14-, 28- and 56-day expansion limits resulted in the required amount of 15 and 25% fly ash contents for the SN-1 and NN-1 aggregates, respectively. For the SN2 aggregate, the 14-day expansion limit required a minimum of 20% by weight replacement of Portland cement by Class F fly ash. For the same aggregate, only 15% fly ash dosage was sufficient to meet the extended failure criteria of 0.33% at 28 days and 0.48% at 56 days. While the alkali-silica reaction

depends on numerous factors, it is recommended to use a minimum amount of Class F fly ash that is sufficient to keep ASR expansion to below 0.10% at 14 days (ASTM C 1260), 0.33% at 28 days and 0.48% at 56 days (Hooton 1991; Rogers & Hooton 1993). 4.2

Analytical approach to determine effective fly ash dosages

This section deals with the required minimum fly ash dosages in relation with the ASR expansion and the CaO content of the trial mortars. The ingredients of CaO, Na2 Oeq , MgO, and SO3 of the SCMs of the mortar bars were replaced by their CaO molar equivalents (CaOeq ) based on Equation 1. The equivalent CaO contents of mixtures containing five different dosages of Class F fly ash as a partial replacement of Portland cement is shown in Table 5. The percent fly ash content (FA) vs. the equivalent CaO content of the mixture showed a perfect linear correlation (with a R2 value of 1) as documented in Equation 3. FA(%) = −1.567 ∗ (CaOeq ) + 112.57 Table 5. bars.

Figure 5. Table 4.

Class F Fly ash dosage and CaOeq of the mortar

Cement (%)

Fly ash (%)

CaOeq x (%)

100 85 80 75 70

0 15 20 25 30

71.838 62.265 59.075 55.884 52.693

x

Influence of fly ash dosages on ASR cracks.

(3)



(Percent of SCMs by weight) * (CaOeq of SCMs).

Minimum dosage of Class F fly ash to suppress ASR expansion at different failure criteria of ASTM C 1260.

14-Dayx

28-Dayy

56-Dayy

Aggregate ID

Fly ash content (%)

ASR at (0.10%)

ASR at (0.33%)

ASR at (0.48%)

SN-1

0 15 20–30

R Inno. Inno.

R Inno. Inno.

SN-2

0 15 20–30

R R Inno.

NN-1

0 15-20 25 30

R R Inno. Inno.

Minimum dosage to control ASR (%) 14-Day

28-Day

56-Day

R Inno. Inno.

15

15

15

R Inno. Inno.

R Inno. Inno.

20

15

15

R R Inno. Inno.

R R Inno. Inno.

25

25

25

Inno. = Innocuous (Not Reactive), R = Reactive. x Based on ASTM C 1260. y Proposed by Hooton 1991, and Rogers & Hooton 1993.

536

The equivalent CaO content and the expansion of the corresponding fly ash content revealed a linear correlation for all trial aggregates as presented in Equation 4. The values of coefficient m and c, as well as the coefficient of multiple determination (R2 ), at various immersion ages of the three aggregate sources are shown in Table 6. The R2 values shown in Table 6 indicated a good fit between the CaOeq of the mixture and the corresponding ASR expansion. CaOeq = m ∗ E + c

In order to determine the optimum Class F fly ash content in controlling alkali-silica reactivity of each trial aggregate, the CaOeq in Equation 3 was replaced by Equation 4 and the value of E in Equation 4 was substituted by EFC (the expansion at or below the failure limits of the ASTM C 1260), as seen in Equation 11.

(4)

where: E is the expansion of mortar bars with various fly ash dosages (0, 15, 20, 25 and 30%). In addition, strong correlations existed between the slope and intercept of the regression lines and the expansions of the mortar bars at various immersion ages. The slope and intercept of the regression lines of the trial aggregates was expressed as the function of the expansion of the untreated mortar bars (having no fly ash content) as illustrated in Equations 5 and 6 for 14 days, Equations 7 and 8 for 28 days, and Equations 9 and 10 for 56 days, respectively.

FA(%) = −1.567(m ∗ EFC + c) + 112.56

Substituting the values of m and c at the immersion ages of 14, 28 and 56 days in Equation 11, the minimum fly ash content required to suppress ASR expansion at 14, 28 and 56 days can be evaluated by Equations 12, 13 and 14, respectively. FA14-d (%) = 143.28 ∗ E14d -FC ∗ E14-d − 147.82 ∗ E14d -FC − 4.10 ∗ E14-d + 30.33

− 86.94 ∗ E28d -FC − 1.79 ∗ E28-d + 30.64

(5)

c14-d = 2.618 ∗ E14-d + 52.483

(6)

m28-d = −29.443 ∗ E28-d + 55.479

(7)

− 67.23 ∗ E56d -FC

c28-d = 1.143 ∗ E28-d + 52.280

(8)

− 0.64 ∗ E56-d + 36.14

m56-d = −16.897 ∗ E56-d + 42.902

(9)

Table 6. Coefficients of the regression line of CaOeq of SCMs vs. ASR expansion of mortar bars.

FA56-d (%) = 26.48 ∗ E56d -FC ∗ E56-d

28

56

(14)

where: E14d -FC , E28d -FC and E56d -FC are the expansions at or below the failure limits of ASTM C 1260 Table 7. Experimental and analytical fly ash dosages to control ASR below 0.10% at 14 days, 0.33% at 28 days and 0.48% at 56 days.

Immersion Age Agg. (Days) ID 14

Agg. ID

m

c

R2

SN-1 SN-2 NN-1 SN-1 SN-2 NN-1 SN-1 SN-2 NN-1

79.83 44.67 21.80 41.88 37.36 13.91 29.41 30.65 12.58

52.86 53.95 54.54 52.47 51.22 52.99 49.10 49.07 49.50

0.93 0.93 0.92 0.96 0.97 0.93 0.96 0.99 0.98

28 14

(13)

(10)

where: mx-d and cx-d are the slope and intercept of the linear regression line of CaOeq of SCMs vs. expansions at x days, respectively. Ex-d is the control expansion at x days. x is the immersion age in days.

Age (Days)

(12)

FA28-d (%) = 46.14 ∗ E28d -FC ∗ E28-d

m14-d = −91.439 ∗ E14-d + 94.332

c56-d = 0.406 ∗ E56-d + 48.776

(11)

537

56

SN-1 SN-2 NN-1 SN-1 SN-2 NN-1 SN-1 SN-2 NN-1

a

Experimental Control fly ash a dose Exp. (%) (%)

Exp. Suppressed (tob,c ) (%)

Analytical Fly Ash dose (%)

0.247 0.418 0.83 0.475 0.570 1.414 0.799 0.725 1.794

0.094 0.081 0.084 0.218 0.261 0.244 0.250 0.409 0.352

18.7 21.5 24.5 17.1 15.3 24.4 24.1 16.0 28.0

15 20 25 15 15 25 20 15 30

Mortars with no fly ash. ash. below the failure limits of the ASTM C 1260.

b Mortars with fly c Expansion at or

at 14, 28 and 56 days, respectively. E14-d , E28-d and E56-d are the control expansions at 14, 28 and 56 days, respectively. Finally, the expansion of the untreated mortar bars and the failure limits of 0.1% at 14 days, 0.33% at 28 days, and 0.48% at 56 days were substituted in Equations 12, 13, and 14, respectively, to obtain the required minimum fly ash content at the above-mentioned immersion ages. Table 7 represents the minimum experimental and analytical fly ash dosages required to arrest excessive expansion of reactive aggregates. As can be seen, there is a good agreement between the experimental and analytical results for all three aggregate sources used in this investigation. Furthermore, the results demonstrate that the required amount of fly ash to control ASR-related expansion is a function of SCMs chemical composition and mineralogical background of aggregate. 5

CONCLUSIONS

Based on the results of the study presented herein, the following conclusions can be made: 1. The ASR expansion of the selected mortar bars decreased with increases in the fly ash content. The reduction in expansion (RIE) was highest at the early age of immersion (usually at 3 or 6 days), and then gradually decreased with an increase in the test duration. Additionally, the results revealed that the RIE decreased rapidly, with an increase in immersion age, at lower fly ash content (15%) than that obtained for the mortar bars made with a higher fly ash dosage (30%). 2. For the most part, the minimum experimental fly ash dosages sufficiently effective to suppress the alkali-silica reactivity of the trial aggregates showed a good correlation among the three failure criteria at the immersion ages of 14, 28 and 56 days. It is anticipated that the aggregate not exceeding the three failure limits is also capable to perform in innocuous behavior in the field. 3. A good agreement existed between the minimum required experimental and analytical fly ash contents. The proposed analytical method may be used to select the minimum amount of fly ash that is capable limiting the ASR expansion of reactive aggregate below the suggested failure criteria

given for different immersion ages. However, since alkali-silica reactivity of aggregate depends on numerous factors, each aggregate needs to be tested before it is used in Portland cement concrete. REFERENCES Bleszynski, R.F. and Thomas, M.D.A. 1998. Microstructural studies of alkali-silica reaction in fly ash concrete immersed in alkaline solutions. Advanced Cement Based Materials. Vol. 7, pp. 66–78. Brough, A.R., Dobson, C.M., Richardson, I.G. and Groves, G.W. 1996. Alkali activation of reactive silicas in cements: in situ 29Si MAS NMR studies of the kinetics of silicate polymerization. Journal of Materials Science, Vol. 31, pp. 3365–3373. Hooton, R.D. 1991. New aggregates alkali-reactivity test methods. Ministry of Transportation, Ontario, Research Report MAT-91-14. Hudec, P. and Banahene, N. 1993. Chemical treatment and additives for controlling alkali reactivity. Cement and Concrete Composites, Vol. 15, pp. 21–26. Lee, C. 1989. Active alkalis in cement-fly ash paste. Proceedings: Eighth InternationalConference on Alkali-Aggregate Reactions. Kyoto, Japan. pp. 223–228. Malvar, J. and Lenke, L.R. 2006. Efficiency of fly ash in mitigating alkali-silica reaction based on chemical composition. ACI Materials Journal, Vol. 103, No. 5, pp. 319–326. Malvar, J., Cline, G.D., Burke, D., Rollings, R., Sherman, T. and Greene, J. 2001. Alkali-silica reaction mitigation state-of-the-art. Technical report No. TR-2195-SHR, Naval Facilities Engineering Service Center, Washington Navy Yard, DC, 40 p. McKeen, R.G., Lenke, L.R. and Pallachulla, K.K. 1998. Mitigation of alkali-silica reactivity in New Mexico. Work performed for New Mexico State Highway and Transportation Departmen. Materials Research Center, ATR Institute, University of New Mexico, USA. Rogers, C.A. and Hooton, R.D., ‘‘Reduction in mortar and concrete expansion with reactive aggregates due to alkali leaching,’’ Cement, Concrete and Aggregates, CCAGDP, Vol. 13, No. 1, Pp. 42–49, 1993. Shehata, M.H. and Thomas, M.D.A. 2000. The effect of fly ash composition on the expansion of concrete due to alkali-silica reaction. Cement and Concrete Research. Vol. 30, pp. 1063–1072. Tauma, E.W., Fowler, D.W. and Carrasquillo, R.L. 2001. Alkali-silica reaction in Portland cement concrete: testing methods alternatives. International Center for Aggregates Research (ICAR). Research Report ICAR-301-1f.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Overdosing remediation of plastic SCC exposed to combined hauling time and temperature N. Ghafoori & H. Diawara University of Nevada, Las Vegas, USA

ABSTRACT: The present study was intended to revert, by mean of overdosing remediation, the adverse effect of combined hauling time and temperature on freshly-mixed self-consolidating concrete (SCC). Temperatures of 43 and −0.5◦ C, to simulate hot and cold weathers, respectively, were used to evaluate the unconfined workability (slump flow), flow rate (T50 ), and dynamic stability (VSI) of the matrices transported for 60 minutes. Mixing and hauling condition of T21H10, i.e. laboratory room temperature of 21◦ C and transportation duration of 10 minutes, was adopted as the control condition. Polycarboxylate-based high range water-reducing admixture (HRWRA) and viscosity modifying admixture (VMA) were used to produce the selected matrices with slump flow of 635 ± 25 mm, VSI of 0 (highly stable concrete) and T50 of 2 to 5 sec. The test results revealed that, under harsh conditions, self-consolidating concrete with suitable fresh performance can be achieved once remediated by the overdosing technique. 1

2

INTRODUCTION

Nowadays, ready-mixed plants are predominantly used in concrete industry for the manufacture of concrete, due to their suitability for close quality control, use in congested sites, and use of agitator trucks to prevent segregation and maintain workability (Neville & Brooks 1987). In general, lengthy hauling time and extreme temperature (due to excessive hot or cold weather) can adversely affect the fresh and hardened properties of concrete by accelerating or retarding the rate of moisture loss and the rate of cement hydration. The progressive use of new generation of plasticizers and superplasticizers in concrete industry has facilitated the control of slump loss caused by harsh mixing, and extreme weather and hauling conditions. Several methods of remediation have been proposed by several investigations. Cooling or heating the concretes raw materials exposed to severe cold or hot weather, making concrete with a higher initial slump than required (overdosing), retempering with water or superplasticizer upon arrival of the concrete in a job site, and others can be noted (Kosmatka et al. 2002; Neville & Brooks 1987; Mehta et al. 2002). For the purpose of this study, remediation by way of overdosing was adopted. This technique consisted of using a higher initial amount of admixtures than usually required under the control condition in order to compensate for the increased yield stress and decreased plastic viscosity created by the combined hauling time and temperature.

RESEARCH OBJECTIVES

The study presented herein was intended to remediate, by the mean of overdosing, the adverse effect of combined hauling time and temperature on freshlymixed self-consolidating concrete. Laboratory room temperatures of 43 and −0.5◦ C to simulate hot and cold weathers, respectively, were used to evaluate the freshly-mixed self-consolidating concretes transported for 60 minutes. Under the above selected temperatures, the unconfined workability (slump flow), flow rate (T50 ), dynamic segregation resistance (VSI), and passing ability of the remediated SCCs were assessed at the end of the hauling time, and compared to the equivalent fresh properties obtained at the control condition, i.e.: fresh matrix hauled for 10 minutes at normal room temperature of 21◦ C.

539

3 3.1

EXPERIMENTAL PROGRAMS Raw materials

The cementitious materials used in all mixtures consisted of ASTM C150 Type V Portland cement and ASTM C 618 class F fly ash. Their physico-chemical characteristics are shown in Table 1. The fine aggregate used had bulk and saturated surface dry specific gravity, absorption, and fineness modulus of 2.75 and 2.78, 0.8%, and 3.0, respectively. The coarse aggregate had a nominal maximum size equal to 12.50 mm and

Table 1. Chemical and physical properties of Portland cement and fly ash.

Table 2.

Chemical composition

Portland cement (%)

Fly ash (%)

SiO2 AL2 O3 Fe2 O3 CaO MgO SO3 Na2 O equivalent K2 O C2 S C3 S C3 A Loss on Ignition Insoluble residue Fineness Blaine, cm2 /g

20.64 3.4 3.4 63.5 4.7 2.4 0.46 – 9 66 4 1.2 0.14 3810

58.9 20.5 5.6 7.5

VMA2 HR1 (ml/100 (ml/100 kg) kg) CA/FA3

0.4 – – – – – 0.3 – –

complied with ASTM C 33 size number 7. Its bulk and saturated surface dry specific gravity, absorption, and dry rodded unit weight were 2.77 and 2.79, 0.6%, 1634 kg/m3 , respectively. Other concrete constituents were tap water, polycarboxylate-based high-range water reducing admixture (HRWRA) and viscosity modifying admixture (VMA) complying with the ASTM C 494 Type F requirements.

3.2 Mixture proportion All matrices were prepared with a constant water-tocementitious materials ratio of 0.4, a uniform cement factor of 391 kg/m3 , and a constant amount of fly ash representing 20% of the cement weight. In proportioning the aggregates content, particular attention was given to the coarse-to-fine aggregate ratio due to its critical role in generating sufficient amount of mortar for the selected self-consolidating concretes. The ASTM C 29 was used to determine the compacted bulk unit weight and the calculated void content using different ratios of the combined coarse and fine aggregates. The optimum volumetric coarse-to-fine aggregate ratio, utilized in the proportioning of the concrete constituents, was found at 0.52/0.48. The quantities of coarse and fine aggregates used in the matrices were 923 and 850 kg/m3 , respectively. The generated paste and mortar fractions and the total volume of coarse aggregate are presented in Table 2. The optimum (minimum) dosages of the high range water-reducing admixture (HRWRA) and viscosity modifying admixture (VMA) used at the control condition (hauling time of 10 minutes and temperature of 21◦ C) are also tabulated in Table 2. These dosages were obtained by evaluating the consistency and stability of concrete using different trial batches until a

209

Mixture proportioning.

26

Volume Paste Mortar coarse fraction fraction aggre. (%) (%) (%)

0.52/0.48 35

66

33

1

High range water reducing admixture, 2 Viscosity modifying admixture, 3 Coarse-to-fine aggregate ratio.

satisfactory slump flow of 635 ± 25 mm; T50 of 2 to 5 sec, and a visual stability index of 0 were attained. 3.3

Mixing, sampling and testing

An electric counter-current pan mixer with a capacity of 0.028 m3 was used to blend concrete components. In simulating the influence of hauling time on the fresh SCCs, a realistic concrete mixing tool with changeable velocity was needed. An environmental chamber to simulate hot or cold temperature conditions was built around the mixing apparatus. The walls, roof and floor of the room were made with plywood and insulated with polystyrene foam to maintain a uniform temperature throughout the experiments. The hot temperature was generated by a heater while a cooling unit was used to produce the cold temperatures. A temperature regulator, which was connected to the heating and cooling units and assisted by multiple probes, maintained the target temperature within ±2◦ C margins. A separate control unit also monitored and recorded the relative humidity of the environmental chamber. The mixing sequence consisted of blending the coarse aggregate with 1/3 of the mixing water for two minutes, followed by the fine aggregate with 1/3 of the mixing water for another two minutes, and the cementitious materials with the remaining 1/3 of the mixing water for three minutes. Finally, the HRWRA and VMA were added and blending of the matrix continued for an additional three minutes, followed by a two-minute rest and resumption of mixing for two additional minutes. From that point on the mixing speed (14.5 rpm) was changed to an agitating speed (7.25 rpm) until the desired hauling time was achieved. The hauling time was defined as the elapsed time between the first contact of water and cementitious materials to the beginning of concrete discharge. At the completion of each mixing and hauling, the fresh self-consolidating concretes were evaluated for unconfined workability (measured by the slump flow), the flow rate or viscosity by inference (evaluated by the T50 time), the dynamic segregation resistance (assessed by the visual stability index (VSI)), and J-ring passing ability (determined by the J-ring test); in accordance with the ASTM C 1611 and C 1621.

540

DISCUSSION OF RESULTS

The combined hauling time and temperature affected the fresh performance of self-consolidating concretes in the form of decrease in unconfined workability, and gain in flow rate or viscosity per inference. The dynamic stabilities of the fresh concretes remained unchanged. Table 3 presents the changes in the fresh performance of self-consolidating concrete as affected by the selected construction-related variables. An overview of slump flow loss and the involved mechanism of action is necessary before proceeding with the discussion on remediation. Several studies have established and reported the fundamental mechanism of slump flow loss of concrete during its hauling (Kosmatka et al. 2002, Jolicoeur and Simard 1998, Flatt et al. 1997). It involves mainly the additional fines brought to the concrete mortar by the grinding of aggregates and cement particles, the growth of the cement hydration products, and the competitive adsorption between the superplasticizer and the sulfate ions (SO2− 4 ) on the cement hydrated products throughout the hauling time3 . Diawara (2008) used the ultra violet visible (UVVis) test to find that the free concentration of polycarboxylate-based HRWRA-cement-water solution changed with material temperatures. The concentration was relatively uniform at temperatures ranging from 14 to 36◦ C and beyond that range, it decreased as the temperature moved toward the extreme cold and hot temperatures. These increases in the solution led to an augmentation of the adsorbed amount of the PCHRWRA carboxylic group (COO-) on cement grains, favoring further increase or decrease in electrostatic repulsion and steric hindrance forces.

Table 4. Required admixtures dosages for overdosing remediation at selected transportation times and temperatures. Mixing condition

HR1 (ml/100 kg)

VMA2 (ml/100 kg)

T43H10 T43H60 T21H10∗ T21H60 T-0.5H10 T-0.5H60

307 386 209 287 209 248

39 52 26 26 26 26

4.1 Optimum HRWRA dosage for the overdosing remediation Table 4 shows the optimum overdosed amount of HRWRA and VMA needed to remediate the adverse effect of the combined hauling time and temperature

Table 3. Fresh properties of hauled SCCs at hot and cold temperatures. Temp. (◦ C) 43.00 21.00 −0.50

Measured slump flow (mm) 480 651∗ 673

T50 (sec) ∗∗

2.79 2.50

VSI

Slump flow loss (mm)

0 0 0

−171 0 22

∗

Reference matrix. T50 could not be measured since the spread of matrix was less than the recommended 508 mm. ∗∗

541

Note: TxHy stands for combined temperature ‘‘x’’ and hauling time ‘‘H’’. ∗ Control condition, 1 High range water reducing admixture, 2 Viscosity modifying admixture. 200 Change in admixture dosage (ml/100 kg)

4

160

HR

VMA

120 80 40 0 -40 T-0.5H60

T21H60

T43H60

Mixing condition

Figure 1.

Admixture dosages at TxHy vs. T21H10.

on plastic self-consolidating concrete. Figure 1 compares the required overdosed admixtures at the selected combined hauling time and temperatures (TxHy) to those of the control condition (T21H10). As it can be seen in Table 4, the HRWRA dosages necessary for the overdosing remediation were unaffected by cold temperature, but increased with hauling time. In comparing to the equivalent control dosage, the optimized amount of HRWRA for the overdosing remediation of 635 mm slump flow SCCs increased on average by 19% after 60 minutes of hauling. The demand in superplasticizer for the overdosing remediation increased with increases in hot temperature. When the trial self-consolidating concretes were transported for 60 minutes, the remediation of the selected 635 mm slump flow matrix necessitated an increase in the HRWRA dosage of 85% when compared to the required superplasticizer dosages obtained under the adopted control condition. In order to revert the adverse effects of the combined hauling time and extreme temperatures on plastic self-consolidating concrete, an overdosing remediation method was used. The increase in demand for HRWRA amount was basically due to the decrease in the ratio Ads/SSAm (adsorption amount of admixture

per specific surface area of concrete mortar). The idea behind this remediation technique was to achieve by trial and error the optimum admixture dosage so that (Ads/SSAm) at the end of the mixing and hauling time became equal or similar to (Adso /SSAmo ) at the control condition T21H10. It is translated mathematically through Equations 1 and 2 given below:

In comparing to the equivalent control VMA dosage, the 635 mm self- consolidating concrete did not required any change in its initial VMA dosage in attaining the target fresh properties, when temperatures were between −0.5 and 21◦ C. However, when the temperature was elevated to 43◦ C the demand in VMA increased by 100%.

– At the control condition T21H10: 

Ads SSAm

 = control

Adso SSAmo

4.3

The fresh properties of the remediated freshly-mixed self-consolidating concretes were evaluated for unconfined workability, flow rate, dynamic stability and passing ability using the slump flow, T50 , VSI, and J-ring tests, respectively. The results are tabulated in Table 5. In comparing the slump flow values at the selected combined hauling time and extreme temperatures conditions to that measured at the control condition, the overdosed remediated fresh SCCs displayed an insignificant difference of less than 1%. The remediated matrices were within the target slump flow of 635 ± 25 mm, T50 time between 2 and 5 seconds, VSI of 0 (highly stable), and allowable J-ring value of 25 to 50 mm.

(1)

– At the target condition TxHy: 

Ads SSAm

 =

Adshaul + Adstemp + (wo + wt )temp SSAmhaul + SSAptemp    A

Adshaul+temp + SSAmhaul + SSAptemp   

Fresh properties of remediated matrices

(2)

B A

Characterizes the slump flow loss

B

Characterizes the slump flow restoration

4.4

where Adshaul = Adsorption amount of admixture induced by hauling time; Adstemp = Adsorption amount of admixture induced by temperature; SSAmhaul = Specific surface area of concrete mortar generated by hauling temperature; wo = Contribution of the mixing water at the control temperature; and wt = Contribution of the mixing water at the target temperature (could be ±). The term  Adshaul+temp corresponds to the increase in adsorption amount of HRWRA generated by the additional amount of superplasticizer with respect to the control dosage. This supplementary adsorption was required to compensate for the increased in specific surface area (SSAmhaul + SSAptemp ) and the variation in aggregate’s water content, (wo + wt )temp . The additional amount of adsorption enhanced electrostatic and steric hindrance repulsive forces between cement particles to meet the target fresh properties of the self-consolidating concrete at the completion of mixing and hauling in extreme temperatures. 4.2

Optimum VMA requirement for the overdosing remediation

The optimum dosage of viscosity modifying admixture required for the overdosing remediation are displayed in Table 4. Figure 1 presents also the increase in VMA optimum dosages as related to the selected hauling time and temperatures.

Prediction of optimized overdosed amount of HRWRA and VMA

The dosages requirement of admixtures for the overdosing remediation of slump flow loss induced by the combined hauling time and temperature were predicted using statistical modeling. The most suitable predictive relationships between the optimized dosages of HRWRA or VMA, as dependent variables, and the target slump flow, hauling time and hot or cold temperatures, as independent variables, are as follows: – In hot temperature HRoverd hot = exp(0.004367ht + 0.015477th + 0.00234SF + 3.483835) Or

(3)

HRoverd hot = exp(aht + bt + cSF + d)

VMAoverd hot = exp(0.00624ht + 0.04027th + 0.00719SF − 2.75206)

(4)

Table 5. Fresh properties of overdosed SCC hauled for 60 minutes under hot and cold temperatures. Temperature (◦ C)

Measured slump flow (mm)

T50 (sec.)

VSI

J ring value (mm)

43.00 21.00 −0.50

648 641 648

2.03 2.38 3.05

0 0 0

44 32 13

542

Or

VMAoverd hot = exp (aht + bt + cSF + d)

– In cold temperature HRoverd cold = exp(0.002839ht − 0.002410th + 0.002353SF + 3.83327) Or

(5)

HRoverd cold = exp (aht + bt + cSF + d)

VMAoverd cold = exp(0.004252ht − 9x10−17 tc + 0.005169SF − 0.330418)

(6)

Or VMAoverd cold = exp (aht + bt + cSF + d) where: HRoverd hot and HRoverd cold = Required dosages of HRWRA for the overdosing remediation of slump flow loss in hot and cold temperatures (ml/100 kg); VMAoverd hot and VMAoverd cold = Required dosages of VMA for the overdosing remediation of slump flow loss in hot and cold temperatures (ml/100 kg), respectively; SF = Target slump flow (mm), with 635 mm ≤ SF ≤ 711 mm ± 25 mm; th = Hot temperature (◦ C), with 21◦ C ≤ tt ≤ 43◦ C ± 2◦ C; tc = Cold temperature (◦ C), with −0.5◦ C ≤ tt ≤ 14◦ C ± 2◦ C. The predictive equations were tested for accuracy using R2 (the coefficient of multiple determination) and S (average standard deviation). R2 values of 92, 93, 98, and 88% were recorded for equations 3, 4, 5, and 6, respectively. For the same equations, S values of 18.82, 6.29, 4.42, and 2.61 ml/100 kg were found. The produced R2 and S values were indicative of good relationship between the dependent and independent variables. Correlations between the data predicted from the regression equations and the actual test results were evaluated using F and T tests. For all four generated predictive equations the calculated Prob(t) = 0 for all coefficients of equation 3, 4, and 5. In Equation 6, Prob(t) = 1.0 for the coefficient b, indicates that the predicted slump flow loss was independent from the selected cold temperatures. Prob(F) = 0 for all four predictive equation. In general, the F and T tests results indicated that the hauling time, the temperature and target slump flow had similar influence on the predictive HRWRA and VMA optimum dosages. The predictive equations yielded percentage errors ranging from 1 to 10% for the HRWRA, and 1 to 13% for the VMA, confirming a good relationship between actual and calculated admixture dosages.

5

hauling time under normal room temperature of 21◦ C). The equivalent change of slump flow in cold temperature was marginal and within the tolerance margin. The overdosing method of remediation was successful in mitigating the adverse effect of combined hauling time and temperature on plastic self-consolidating concrete. Fresh matrices with suitable unconfined workability, plastic viscosity, dynamic stability, and passing ability were achieved once remediated by the selected technique. No evidence of segregation or bleeding in slump flow was observed in any of the remediated self-consolidating concretes, indicating that a stable matrix condition was achieved through the adopted overdosing method. The statistical equations to predict the required optimum overdosed admixtures (HRWRA and VMA) amount to achieve the desired flow ability and stability under prolonged transportation time and extreme temperatures showed significant relationships between the dependent and independent variables.

CONCLUSIONS

Self-consolidating concrete manufactured and hauled for 60 minutes in elevated temperatures of 43◦ C experienced about 43% reduction in slump flow when compared to the equivalent concretes produced at the control temperature of T21H10 (i.e., 10-minute

543

ACKNOWLEDGEMENTS The authors would like to acknowledge the financial support of the Nevada Department of Transportation, Grant number P 077-06-803. Thanks are also extended to a number of admixture manufacturers and concrete suppliers who contributed materials used in this investigation. Their names are withheld to avoid any concern of commercialization or private concern.

REFERENCES American Society for Testing and Materials, ‘‘Standard Specification for Portland Cement,’’ (ASTM C 150), Annual Book of ASTM Standards, Vol. 4.01, 2004, pp. 150–157. American Society for Testing and Materials, ‘‘Standard Test Method for Coal Fly Ash and Raw or Calcined Natural Pozzolan for Use as a Mineral Admixture in Concrete,’’ (ASTM C 618), Annual Book of ASTM Standards, Vol. 4.02, 2004, pp. 319–312. American Society for Testing and Materials, ‘‘Standard Specification for Concrete Aggregates,’’ (ASTM C 33), Annual Book of ASTM Standards, Vol. 4.02, 2004, pp. 10–16. American Society for Testing and Materials, ‘‘Standard Specification for Chemical Admixture for Concrete,’’ (ASTM C 494), Annual Book of ASTM Standards, Vol. 4.02, 2004, pp. 271–279. American Society for Testing and Materials, ‘‘Standard Test Method for Bulk Density (‘‘Unit Weight’’) and Voids in Aggregate,’’ (ASTM C 29) Vol. 4.02, 2004, pp. 1–4. American Society for Testing and Materials, ‘‘Standard Test Method for Slump Flow of Self-Consolidating Concrete,’’ (ASTM C 1611) Annual Book of ASTM Standards, Vol. 4.02, 2005, 36–41.

American Society for Testing and Materials, ‘‘Standard Test Method for Passing Ability of Self-Consolidating Concrete by J-Ring,’’ (ASTM C 1621) Annual Book of ASTM Standards, Vol. 4.02, 2005, 42–45. Diawara, H., ‘‘Parametric Study of Self-Consolidating Concrete,’’ Doctoral dissertation, under the supervision of N., Ghafoori, University of Nevada, Las Vegas (UNLV), USA, 2008, 370 p.

Neville, A.M., and Brooks, J.J., ‘‘Concrete Technology,’’ Longman Scientific and Technical Publisher, 1987, 438 p. Kosmatka, S.H., Kerkhoff, B., and Panarese, W.C. 2002. Design and Control of Concrete Mixtures. 14th Edition, Portland Cement Association, Skokie, Illinois: 358 p. Mehta, P.K., and Monteiro Paulo, J.M., ‘‘Concrete-Structure, Properties, and Materials,’’ Prentice-Hall, Inc., Englewood Cliffs, NJ, 2002, pp. 450–548.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Research into the optimum level of rock-derived micro-fine particles in sand for concrete T. Kaya & K. Hashimoto Kotobuki Eng. and Mfg. Co, Ltd. (KEMCO), Kure, Hiroshima, Japan

H. Yamamoto Graduate School for International Development and Cooperation, Hiroshima University, Hiroshima, Japan

ABSTRACT: Since natural sand for concrete is running short globally, due to the increasing influence of environment protection legislation, and because accessible reserves are dwindling, it is obliged to improve the quality of manufactured sand, and use it in greater quantities as a substitute for the natural product. However, producing well-shaped and properly graded sand also generates micro-fine particles as an inevitable by-product. In this study, the specific size range which has the greatest detrimental effect to concrete has been successfully discovered, and hence a method of using micro-fines most effectively can be proposed.

1

INTRODUCTION

The Japanese standard JIS 5005 restricts the maximum permissible content of 150 μm or finer to less than 15%, and 75 μm or finer to less than 7% (by decantation method). Excessive content particularly of 75 μm or less is considered to have a detrimental effect to the properties of concrete. It is understood that some EU standards have already eased or deregulated the allowances of micro-fine particles contained in the sand. In practice, however, end-users seem still to avoid the inclusion of micro-fines, unconvinced that such material can be beneficial in the concrete mix. Currently, therefore, almost all manufactured sand producers eliminate substantial volumes of fine particles from the final product, thereby reducing sand recovery, and leading to additional cost in disposal of the waste. Two methods were used to identify the optimum inclusion of micro-fine particles (herein referred to as Original Micro-Fine, or OMF) which is a by-product of dry-crushed sand (herein referred to as Base Sand, or BSa) in manufactured sand for concrete. The first approach was simply to increase the volume of microfines in their original form, herein referred to as the Original Fine Grading Distribution (OFG). The second was to classify OMF into coarse and fine fractions and reintroduce only the coarser fraction (herein referred to as the Modified Fine Grading Distribution, or MFG) to BSa in increasing blending ratios. Several trial concrete mixes were performed to compare the effect of varying gradation and proportion of the reintroduced fines on the concrete. In the case of OFG, increasing the proportion of OMF proved detrimental

545

to the quality of the concrete, including a reduction in workability (Kaya 2002). With MFG, however, it was possible to avoid negative effects on the fresh properties and other key qualities of concrete, while even substantial volumes of OMF were added (Kaya 2008). The specific size range of particles within the micro-fines that has an undesirable effect to the concrete was identified. 2

TEST METHOD

A bulk sample of dry manufactured sand was made using a dry and grain shape improving sand-making process, designed to give suitable particle size distribution and shape for use as concrete sand. Ten samples were constructed as follows: One plain BSa, four gradations of MFG added to BSa, and five gradations of OFG added to BSa. For the four samples of MFG, OMF was classified into four different coarse fractions, +120 μm, +75 μm, +40 μm and +20 μm (herein referred to as Modified Micro-Fine, or MMF) and added back to BSa respectively. These ratios were equivalent to 3.75%, 9.75%, 12.83% and 18.0% with the volume of BSa being taken as 100%. For the five samples of OFG, corresponding percentages were reintroduced, 3.75%, 9.75%, 12.83% and 18.0%, together with a final sample including the original proportion of micro-fines, equivalent to 25% (20%/80%). These samples were used as fine aggregate for trial concrete mixes. The same unit water content was used for all samples, providing comparison of slump and compressive strength results. Figure 1 shows the

Pattern

Original fine grading distribution : OFG

Modified fine grading distribution : MFG

Feed

Current situation

Dry and grain shape improving sand making system (80%)

(20%)

Base sand : BSa

Original micro fine : OMF OMF

BSa OMF

How to OFG and MFG

BSa

OMF

Coarse : MMF Separator

Distributere

OMF

Fine : MMF-S

OFG

MFG

Figure 1. The production process for fine aggregate for the test. Table 1.

Test sample number, code and blending ratio. Fine aggregate

Pattern

Test No.

BSa

1 

Pattern OFG

3  3  4  5  6 

Pattern MFG

7  8  9  10 

Blending ratio (%)

Fine particle

Code

*Surface area (cm2 /g)

Code

Remarks

BSa

172

−

−

3.75 9.75 12.83 18 25

V4B V10B V13B V18B V25B

236 329 373 442 526

OMF 329 373 442 526

90% of 150 μm passed

3.75 9.75 12.83 18

S4B-120 S10B-75 S13B-40 S18B-20

184 203 221 284

MMF-120 MMF-75 MMF-40 MMF-20

+120 μm is taken from OMF by air separate. +75 μm is taken from OMF by air separate. +40 μm is taken from OMF by air separate. +20 μm is taken from OMF by air separate.

0

*Surface area are found out by air permeability method.

production process for fine aggregate, and Table 1 shows the test sample number, code and blending ratios.

3

4

TEST RESULTS AND ANALYSIS

Table 4 is a summary of the tests. Slump and compressive strength are compared for corresponding blending ratios of OFG and MFG.

TEST MATERIAL

Table 2 shows the specified mix for the 10-trial mix. For fair and clear comparison, AE water reduction agent was used and same unit water content was applied. Table 3 shows the properties of the materials used in the tests. Figure 2 shows the size distribution of BSa and an envelope which is regulated by JIS A5005. Figure 3 shows the size distributions of four kinds of MMF samples and OMF.

4.1

Slump

Figure 4 shows how slump varied as blending ratio increased. It can be seen that slump for OFG constantly decreased. However, that for MFG was maintained down to 13% blend, dropped slowly as far as 18% bend, and fell drastically beyond 18%. The drastic fall after 18% is considered to be caused by fine particles of less than 20 μm. Beyond that size,

546

Table 2.

Specified mix for concrete tests. Specified mix Fine aggregate

Admixture

S/a (%)

Water content (kg/m3 )

Cement content (kg/m3 )

Corse aggregate (kg/m3 )

BSa (kg/m3 )

NO. 70 C×(%)

202 C×(%)

Pattern

Test No.

Fine aggregate No.

BSa

1 

BSa

50

48

192

384

869

804

0

0.25

0.002

Pattern OFG

2  3  4  5  6 

V4B V10B V13B V18B V25B

50 50 50 50 50

47.1 45.7 45.1 44.3 42.9

192 192 192 192 192

384 384 384 384 384

885 907 918 933 954

759.5 697 668.3 628 574.4

28.5 68 85.7 113 143.6

0.25 0.25 0.25 0.25 0.25

0.002 0.0026 0.0035 0.0057 0.008

Pattern MFG

7  8  9  10 

S4B-120 S10B-75 S13B-40 S18B-20

50 50 50 50

47.3 45.9 45.2 44.3

192 192 192 192

384 384 384 384

883 905 916 933

762.4 698.9 670.9 628

28.6 68.1 86.1 113

0.25 0.25 0.25 0.25

0.002 0.002 0.0022 0.0047

W/C (%)

Fine particle

For fair and clear comparison, AE water reduction agent was used and constant unit water content was applied.

Fine agg.

Coarse agg.

Properties of the material used in the tests.

5

Cement

Ordinary Portland cement

Admixture

AE water reducing agent

Sand type Density Water absorption Solid ratio Aggregate type Density Water absorption Solid ratio

Manufactured from sandstone 2.61 g/cm3 saturated surface dry 1.05% 56.8% Crushed sandstone 20/5 mm 2.61 g/cm3 saturated surface dry 1.05% 59.7% (JIS A1104:2006)

CUMULATIVE PASSING (%)

4 3 2 1 0

1

10

100

1000

Size (μ )

Figure 3.

The size distributions of MMF and OMF samples.

the total surface area of MFG is considerably smaller than that of OFG in each corresponding blending ratio. Figure 6 shows the correlation between slump and surface area. Both OFG and MFG have similar curves, the greater the surface area, the less the slump.

100 90

The size distributions

Table 3.

JIS (upper)

80 70 60

B

50

a

4.2

40 30

Figure 7 shows the correlation between 28-day compressive strength and blending ratio. Figure 8 shows the correlation between 28-day compressive strength and surface area. It can be seen that OFG and MFG have similar compressive strength to BSa regardless of increased blending ratio or surface area.

JIS(lower)

20 10 0 0.10

1.00

Compressive strength

10.00

PARTICLE SIZE (mm)

Figure 2. The size distribution of base sand (BSa) and the JIS A5005 envelope.

5 particles have enormous effect on the total surface area and greatly increase cohesiveness of fresh concrete. Figure 5 shows the correlation between blending ratio of fine particles and surface area. It can be noted that

547

CONCLUSIONS

This study was intended to find the optimum use of micro fine particles generated while making sand from crushed rock, using slump and 28-day compressive

Table 4.

Summary of the test.

25

15

BSa V4B V10B V13B V18B V25B

10 5

S4B-120 S10B-75 S13B-40 S18B-20

5

400

S18B-20

OFG

300 200

MFG

100 0

0 0

S10B-75 S13B-40

V13B V18B V25B

2

Surface area(cm /g)

MFG

OFG

S4B-120

V10B

500

20 Slump (cm)

BSa V4B

600

10

15

20

0

25

5

10

15

20

25

Blending ratio (%)

Blending ratio (%)

Figure 4.

Variation of slump as blending ratio is increased.

Figure 5. The correlation between blending ratio and surface area.

548

strength as key parameters, by varying the blending ratios of unmodified and modified fines, OFG and MFG. The following is a summary of this study:

25

Slump (cm)

20 15 10 5

1. Micro fines with all of the −20 μm fraction eliminated can be used as part of the fine aggregate without detrimental effect on the properties of fresh concrete. 2. As long as blending ratio is the same, a suitably modified fines gradation is better than the original fines gradation for the properties of fresh concrete, including maintaining favourable slump. Reduced slump is related to increased surface area of fine aggregate, for both original and modified fines gradation. 3. Reduced compressive strength was not evident, regardless of increased blending ratio of original and modified fines gradation. 4. Slump in concrete using the +120 μm, 75 μm, and 40 μm fractions of modified fines gradation was not markedly reduced. A slight reduction of slump in the concrete using +20 μm was found with a slight increase of cohesiveness. However, drastic reduction of slump was found in the concrete using the −20 μm.

MFG

BSa V4B

S4B-120

V10B V13B

S10B-75 S13B-40

V18B V25B

S18B-20

OFG

0 150

250

350

450

550

2

Surface area (cm /g)

Figure 6. The correlation between slump and surface area (Both OFG and MFG have similar curves, the greater the surface area, the less the slump).

60 OFG

Compressive strength (N)

50

MFG 40 BSa

30 20 10

V4B

S4B-120

V10B

S10B-75

V13B

S13B-40

V18B

S18B-20

REFERENCES

V25B

0

0

5

10

15

20

25

Blending ratio (%)

Figure 7. Correlation between 28-day compressive strength and blending ratio.

Compressive strength (N)

60 OFG

50

MFG

40 BSa

30 20 10

V4B

S4B-120

V10B

S10B-75

V13B

S13B-40

V18B

S18B-20

V25B

0 150

250

350

450

550

Surface area (cm2/g)

Figure 8. Correlation between 28-day compressive strength and surface area.

549

Kaya, T. 2002. Research on the effect on properties of fresh concrete of adding a substantial volume of fine particles to well shaped manufactured sand. 54th Meeting for Reading Research Papers at Civil Engineering Institute, Chugoku, Japan Volume 22: pp. 551–552. (in Japanese) Kaya, T. 2008. Research on beneficial use of fine rock particles in concrete. Journal of Japan Institute of Aggregate Technology No. 140: pp. 84–89. (in Japanese)

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Retempering remediation of transported SCC under extreme temperatures H. Diawara & N. Ghafoori University of Nevada, Las Vegas, USA

ABSTRACT: The investigation presented herein is intended to study the retempering technique of remediation to mitigate the influence of combined transportation time and extreme hot and cold temperatures on plastic selfconsolidating concrete. This method of remediation consisted of adjusting admixture dosage at the end of hauling to produce a matrix with similar fresh properties to those of the control concrete prepared at the temperature of 21◦ C and transportation duration of 10 minutes. The trial matrices were prepared with a constant waterto-cementitious materials ratio, a uniform cementitious materials (cement and fly ash) content, and a constant coarse-to-fine aggregate ratio that provided the optimum aggregate gradation. Polycarboxylate-based high range water-reducing admixtures (HRWRA) and viscosity modifying admixtures (VMA) were used to produce selfconsolidating concretes with slump flow of 635 ± 25 mm. The test results revealed that the retempering method of remediation was successful in mitigating the adverse effect of combined hauling time and temperature on the fresh properties of self-consolidating concrete. 1

INTRODUCTION

Self-consolidating concrete requires special attention in mixture proportioning, mixing, transporting, placing, and curing, particularly in extreme weather condition. Mixing at a high speed or for a long period of time (about one or more hours); and high temperature due to excessive heat of hydration in mass concreting, and/or the use of hot materials can result in slump loss (Kosmatka et al. 2002; Neville & Brooks 1987; Mehta et al. 2002). ACI Committee 305 defines hot weather as any combination of high air temperature, low relative humidity, and high wind velocity tending to impair the quality of fresh and hardened concrete or otherwise resulting in abnormal properties. The hot weather condition is transferred to the concrete through the concrete ingredients. High temperature of fresh concrete than normal results in a fast hydration of cement leading to an accelerated rate of setting and a lower long term strength and hardened properties. Admissible hot concrete should have a temperature between 29◦ C to 32◦ C in the time of its placement (Kosmatka et al. 2002). ACI Committee 306 defines cold weather as a period when for more than 3 successive days the average daily air temperature drops below 5◦ C and stays below 10◦ C for more than one-half of any 24-hour period. As the temperature of concrete decreases, the rate of setting and hardening, and the development of strength decrease progressively until the freezing point is reached (Kosmatka et al. 2002). In general, slump flow loss occurs when the free concrete’s mixing water is absorbed by the hydration

551

reactions, adsorbed on the surfaces of cement hydrated products, or by evaporation (Kosmatka et al. 2002; Neville & Brooks 1987). It is a normal phenomenon which is related to the intrinsic nature of concrete. The slump flow loss can lead to an unusual rate of stiffening in fresh concrete and cause loss of entrained air, strength and durability; difficulty in pumping and placing; and excessive effort in placement and finishing operation (Neville & Brooks 1987). To overcome the slump flow loss, several remediation methods are generally practiced. Overdosing the admixture amount in attaining the target slump flow at job site or retempering with admixture instead of water are the preferred methods. Retempering with admixture instead of water is more suitable, since the use of water can induce side effects on the fresh properties and serviceability of the hardened concrete. The remediation, by mean of retempering, of the plastic self-consolidating concrete as affected by the combined transportation time and temperature is presented in the subsequent sections. 2

RESEARCH OBJECTIVE

The research investigation presented herein was intended to study the influence of combined transportation time and hot and cold temperatures on the fresh performance of self-consolidating concretes made with 635 mm slump flow. Transported self-consolidating concrete under extreme temperatures were retempered to produce fresh properties similar to those obtain for the reference concrete Additionally, the test results of the retempering remediation, presented herein, and

those reported in the companion investigation dealing with the overdosing remediation (Ghafoori & Diawara 2009) were compared. 3

EXPERIMENTAL PROGRAM

The cementitious materials used in all mixtures consisted of ASTM C 150 Type V Portland cement and ASTM C 618 class F fly ash. The Type V Portland cement had a Blaine fineness of 423 m2 /kg and the following percentages of the chemical constituents: SiO2 = 20.1%, Al2 O3 = 4.0%, Fe2 O3 = 3.6%, CaO = 63.5%, MgO = 2.8%, SO3 = 2.9%, C3 A = 4%, C3 S = 58%, C2 S = 14%, Na2 O equivalent = 0.57%, loss on ignition = 2.3%, and insoluble residue = 0.44%. The fly ash had the followings chemical composition: SiO2 = 58.2%, Al2 O3 = 17.4%, Fe2 O3 = 4.8%, CaO = 7.9%, SO3 = 0.6%, moisture content = 0.0%, and loss on ignition = 4.2%. The fine aggregate met the ASTM C 33 requirements. Its bulk and saturated surface dry specific gravity, absorption, and fineness modulus were 2.75 and 2.78, 0.8%, and 3.0, respectively. The coarse aggregate had a nominal maximum size equal to 12.50 mm and complied with ASTM C 33 size number 7. Its bulk and saturated surface dry specific gravity, absorption, and dry rodded unit weight were 2.77 and 2.79, 0.6%, 1634 kg/m3 , respectively. Other concrete constituents were tap water, polycarboxylate-based high-range water reducing admixture and viscosity modifying admixture complying with the ASTM C 494 Type F requirements. All matrices were prepared with a constant waterto-cementitious materials ratio of 0.4 (corresponding to an actual water content of 190 kg/m3 ); a uniform cement factor of 390 kg/m3 ; a fly ash amount representing 20% of the cement weight (78 kg/m3 ); and a coarse and fine aggregates contents of 923 and 850 kg/m3 , respectively. In proportioning the aggregates content, particular attention was given to the coarse-to-fine aggregate ratio due to its critical role in generating sufficient amount of mortar for the selected self-consolidating concretes. The ASTM C 29 was used to determine the compacted bulk unit weight and the corresponding void content using different ratios of the combined coarse and fine aggregates. The optimum volumetric coarse-to-fine aggregate ratio, utilized in the proportioning of the concrete constituents, was found at 0.52/0.48. The quantities of the HRWRA and VMA were 209 and 26 ml/100 kg of the cementitious materials content, respectively. These amounts of chemical admixtures represented the optimum (minimum) dosages and were obtained by evaluating the consistency and stability of concrete using different trial batches until a satisfactory slump flow of 635 ± 25 mm and a visual stability index of 0 were attained.

Laboratory trial mixtures were used to produce the required self-consolidating concretes. The influence of hauling time on the fresh SCCs was simulated by mean of a concrete mixing tool with changeable velocity. An environmental chamber to replicate hot-weather conditions was built around the mixing apparatus. The walls, roof and floor of the room were made with plywood and insulated with polystyrene foam to maintain a uniform temperature throughout the experiments. The hot temperatures were generated by a heater. A temperature-regulator, which was connected to the heating unit and assisted by multiple probes, maintained the target temperature within ±2◦ C margin. A separate control unit also monitored and recorded the relative humidity of the environmental chamber. Prior to actual mixing, the concrete’s dry ingredients were stored in the environmental chamber for 24 hours or until they reached the target temperatures. The mixing water was kept at a constant temperature of 21 ± 2◦ C to avoid any interference with the rate of cement hydration. The HRWRA and the VMA were also kept at the normal laboratory conditions as recommended by the manufacturer. The mixing sequence consisted of: (1) adding the coarse aggregate with 1/3 of the water and mixing for two minutes, (2) adding the fine aggregate and 1/3 of the water and mixing for two minutes, (3) adding the cementitious materials with the remaining 1/3 of the water and mixing for three more minutes, and (4) adding admixtures and mixing for three additional minutes, followed by a two-minute rest and the resumption of the mixing for two more minutes. Immediately upon completion of the mixing sequence, the fresh self-consolidating concretes were evaluated for the unconfined workability (measured by the slump flow), the flow rate or viscosity by inference (evaluated by T50 time), and the dynamic segregation resistance (evaluated by VSI) in accordance with the ASTM C 1611. 4

DISCUSSION OF RESULTS

The term ‘‘TxHy’’ that is used throughout this presentation indicates the mixing and environmental conditions. ‘‘Tx’’ stands for the temperature ‘‘x’’, and ‘‘Hy’’ represents the hauling time ‘‘y.’’ Fresh self-consolidating concrete manufactured and hauled in extreme temperatures experienced slump flow losses in hot conditions, and gains in cold environments when compared to the equivalent concretes produced at the control temperature of T21H10 (i.e., 10-minute hauling time under normal room temperature of 21◦ C) (Diawara 2008). The changes in fresh properties of plastic self-consolidating concretes can be explained through the increase or decrease in adsorption amount of admixture per specific surface

552

4.1

HRWRA requirement for retempering remediation

As illustrated in Table 1 and Figure 1, the total HRWRA dosage required for the retempering remediation remained unaffected by the cold and normal room temperatures when compared to the control dosage at T21H10 condition. However, it increased when the hauling time increased or when the mixing temperature was elevated from 21◦ C to 43◦ C.

Required HRWRA dosage (ml/100 kg)

area of concrete mortar (Ads/SSAm), the contribution of aggregate’s moisture content, and partial evaporation of mixing water. The contribution of the hauling time to the increase in SSAm was attributed to the grinding of aggregate and cement particles, and the growth of cement hydrated products. The influence of temperature was explained through the growth of cement hydrated products (Diawara 2008). In an attempt to mitigate the negative influence of the combined hauling time and extreme temperatures on plastic self-consolidating concrete, and to allow the freshly-mixed self-consolidating concrete to play its role of highly flowable and stable concrete, the retempering remediation method was studied. It consisted of adding additional admixtures just before placement of concrete in order to restore the desired workability and stability. The choice of the initial dosage of admixture preceding the retempering of the fresh matrix at the completion of the mixing and hauling (prior to the placement) is critical for the mixture adequacy and economy. Several scenarios were explored and HRWRA and VMA optimum dosages presented in Table 1 were used as initial dosages. These initial dosages were found effective to revert the negative effect of temperature on plastic self-consolidating concrete (Diawara, 2008).

T-0.5H60

Initial dosage (ml/100kg)

Retempered dosage (ml/100 kg)

Mixing condition

HRWRA1

VMA2

HRWRA

VMA

T21H10∗ T43H60 T21H60 T-0.5H60

209 307 209 209

26 39 26 26

0 19.6 19.6 19.6

0 0 0 0

Note: TxHy stands for combined temperature ‘‘x’’ and hauling time ‘‘H’’. ∗ Control condition; 1 High range water reducing admixture; 2 Viscosity modifying admixture.

553

T21H60

T43H60

Mixing condition

Figure 1. Optimum dosage of HRWRA for the retempering remediation of slump flow loss due to the combined hauling time and temperature.

In cold environment, when the 635 mm self-consolidating concretes was hauled for 60 minutes, the increase in HRWRA dosage with respect to the equivalent control dosage was only 9.4%. The same matrix necessitated up to 56.3% increase in HRWRA dosage when the mixing environment was changed to the elevated temperature of 43◦ C. In comparing to the initial dosage, the retempering remediation at the mixing condition TxH60 required only an additional HRWRA dosage of 19.6 ml/100 kg at the end of the hauling time. This relative small increase in the dosage requirement of the superplasticizer was mainly credited to the use of the initial admixture amount which was sufficient to eliminate the adverse effect of temperature. The additional amount of admixture was primarily useful to mitigate the adverse effects of hauling time.

4.2 Table 1. Required admixtures dosages for retempering remediation at selected construction-related variables conditions.

400 380 360 340 320 300 280 260 240 220 200

VMA requirement for retempering remediation

The optimum dosage requirements of the viscosity modifying admixture used to remediate the adverse influence of combined hauling time and extreme temperature on fresh self-consolidating concrete are also shown in Table 1. Similarly to the case of the HRWRA, the demand in VMA for the 635 mm SCC remained unchanged in cold temperatures and increased in elevated temperatures. However, despite the additional demand of HRWRA amount with the increase in transportation time as discussed in the previous section, the optimum dosage of VMA necessary for the retempering remediation was unaffected by hauling time. The additional fines brought to the matrix during the hauling enriched the concrete paste. Consequently, the specific surface area of the mortar, SSAm, increased resulting in a higher viscosity (T50 time between 2 and 5 seconds) and VSI (0 or 1).

4.3

Prediction of optimized retempering amount of admixture

errors less than 1%, confirming a very strong relationship between the actual and the calculated admixture dosages.

This section covers only HRWRA since the required retempering dosage of VMA is constant. The most suitable predictive equations of the optimized dosages of HRWRA as a function of transportation time, extreme hot and cold temperatures, target slump flow, and initial HRWRA dosages, were determined using a 95% confidence level. The relationships are as follows:

4.4

The current section presents a comparison of the fresh properties of the remediated self-consolidating concretes. The designed optimum HRWRA and VMA dosages of the retempering and overdosing remediation methods are also covered.

– In extreme hot temperature of (43◦ C): HRretemp hot = 0.32698ht − 0.06571SF + 0.13577HRinitial

(1)

Or HRretemp hot = aht + bSF + cHRinitial – In extreme cold temperature of (−0.5◦ C): HRretemp cold = 0.32698ht − 0.10408SF + 0.13590HRinitial

(2)

Or HRretemp cold = aht + bSF + cHRinitial where: HRretemphot and HRretempcold = High range water-reducing admixture retempering dosages in hot and cold temperatures, respectively, (ml/100 kg); HRinitial = High range water-reducing admixture initial dosage (ml/100 kg); SF = Target slump flow in hot temperature condition (mm), with 635 mm ± 25 mm; ht = Hauling time (min), with 20 min ≤ ht ≤ 80 min. The predictive equations were tested for accuracy using R2 (the coefficient of multiple determination) and S (average standard deviation). Correlations between the data predicted from the regression equations and the actual test results were evaluated using F and T tests. The following results were found:

4.4.1 Fresh properties of remediated matrices The measured results are tabulated in Table 2 for the overdosing method and Table 3 for the retempering approach. They show that all remediated selfconsolidating concretes were within the target slump flows of 635 ± 25 mm, T50 time between 2 and 5 seconds, VSI of 0 (highly stable), and allowable J ring value of 25 to 50 mm. This is an indication that, under combined prolonged hauling time and extreme temperature conditions, self-consolidating concrete with suitable unconfined workability, plastic viscosity, dynamic stability, and passing ability can be pro duced by both overdosing and retempering techniques of remediation. 4.4.2 Admixture dosages Figure 2 compare the overdosed and retempered amount of HRWRA required for the remediations of slump flow losses at the selected combination of hauling time and extreme temperatures. In comparing to the overdosing method, the retempering remediation of the 635 mm self-consolidating concretes, hauled for Table 2. Fresh properties of SCC remediated by mean of overdosing method.

– For Equation 1: R2 = 99%, S = 0.0074 ml/100 kg Prob(t) = 0.0000, 0.0000, and 0.0000 for a, b, and c, respectively; Prob(F) = 0.0000.

Mixing condition

Measured slump flow (mm)

T50 (sec.)

VSI

J ring value (mm)

T43H60 T21H60 T-0.5H60

648 641 648

2.03 2.38 3.05

0 0 0

44 32 13

– For Equation 2: R2 = 99 %, S = 0.0074 ml/100 kg (0.00011 oz/cwt) Prob(t) = 0.0000, 0.0000, and 0.0000 for a, b, and c, respectively; Prob(F) = 0.0000. The regression equations produced R2 and S values indicative of very strong relationship between the dependent variable (retempered dosage of HRWRA) and the independent variables (hauling time, hot and cold temperatures, target slump flow, and initial dosages of HRWRA). The F and T tests results indicated that all selected independent variables had similar influence on the predictive retempered dosages of HRWRA. The predictive equations yielded percentage

Comparison of overdosing and retempering remediations

Table 3. Fresh properties of SCC remediated by mean of retempering method. Mixing condition

Measured slump flow (mm)

T50 (sec.)

VSI

J ring value (mm)

T43H60 T21H60 T-0.5H60

654 644 638

1.97 2.18 2.81

0 0 0

44 25 3

554

HRWRA dosage (ml/100 kg)/

450.0 400.0 350.0 300.0 250.0 200.0 150.0 100.0

Retempering Remediation

50.0

Overdosing Remediation

0.0 T-0.5H60

T21H60

T43H60

Mixing condition

Figure 2. Comparison of total HRWRA dosages for overdosing and retempering remediations.

VMA dosage (ml/100 kg)/

60.0 50.0 40.0 30.0

ACKNOWLEDGEMENTS

20.0 10.0

Retempering Remediation Overdosing Remediation

0.0 T-0.5H60

T21H60

T43H60

Mixing condition

Figure 3. Comparison of total VMA dosages for overdosing and retempering remediations.

60 minutes under −0.5, 21, and 43◦ C required about 10, 20, and 15% less admixture amount, respectively. The optimum VMA dosages of the retempered selfconsolidating concretes were also lower than their counter parts obtained by the overdosing remediation method. As depicted in Figure 3, when the temperature of the raw materials and the mixing environment were set at 21◦ C or −0.5◦ C, the 635 mm slump flow matrices did not require any adjustment under both overdosing and retempering conditions. However, as the temperature increased to 43◦ C, the same retempered mixture experienced 25% decreases in VMA dosage at the end of 60 minutes hauling time when compared to that of equivalent overdosed self-consolidating concretes. 5

losses in hot conditions and gains in cold environments when compared to the equivalent concretes produced at the control temperature of 21◦ C and the transportation time of 10 minutes. – The retempering method of remediation was successful in mitigating the adverse effect of combined hauling time and temperature on self-consolidating concrete. Fresh matrices with desired unconfined workability, plastic viscosity, dynamic stability, and passing ability were achieved once remediated by the retempering methods of remediation. – In general, the retempering remediation required less HRWRA and VMA than the overdosing remediation. – The statistical equations to predict the required optimum retempered HRWRA amount to achieve the desired flow ability and stability under the combined hauling time and extreme temperatures showed significant relationships between the dependent and independent variables.

CONCLUSIONS

Based on the results of the study, the following conclusions can be drawn. – In general, the self-consolidating concrete hauled in extreme temperatures experienced slump flow

555

The authors would like to acknowledge the financial support of the Nevada Department of Transportation, Grant number P 077-06-803. Thanks are also extended to a number of admixture manufacturers and concrete suppliers who contributed materials used in this investigation. Their names are withheld to avoid any concern of commercialization or private concern.

REFERENCES Kosmatka, S.H., Kerkhoff, B., and Panarese, W.C. 2002. Design and Control of Concrete Mixtures. 14th Edition, Portland Cement Association, Skokie, Illinois: p. 358. Neville, A.M., and Brooks, J.J., ‘‘Concrete Technology,’’ Longman Scientific and Technical Publisher, 1987, p. 438. Mehta, P.K., and Monteiro Paulo, J.M., ‘‘Concrete-Structure, Properties, and Materials,’’ Prentice-Hall, Inc., Englewood Cliffs, NJ, 2002, pp. 450–548. ACI Committee 305, ACI 305.R-88, Hot Weather Concreting, American Concrete Institute, Detroit, Michigan, 1988. ACI Committee 306, ACI 306.R-88, Cold Weather Concreting, American Concrete Institute, Detroit, Michigan, 1988. Ghafoori, N., and Diawara, H., ‘‘Overdosing remediation of plastic SCC exposed to combined hauling time and temperature,’’ Proceedings of the 5th International Structural Engineering and Construction Conference (ISEC-5), 21–27 September, 2009, Las Vegas, USA. American Society for Testing and Materials, ‘‘Standard Specification for Portland Cement,’’ (ASTM C 150), Annual Book of ASTM Standards, Vol. 4.01, 2004, pp. 150–157.

American Society for Testing and Materials, ‘‘Standard Test Method for Coal Fly Ash and Raw or Calcined Natural Pozzolan for Use as a Mineral Admixture in Concrete,’’ (ASTM C 618), Annual Book of ASTM Standards, Vol. 4.02, 2004, pp. 319–312. American Society for Testing and Materials, ‘‘Standard Specification for Concrete Aggregates,’’ (ASTM C 33), Annual Book of ASTM Standards, Vol. 4.02, 2004, pp. 10–16. American Society for Testing and Materials, ‘‘Standard Specification for Chemical Admixture for Concrete,’’ (ASTM C 494), Annual Book of ASTM Standards, Vol. 4.02, 2004, pp. 271–279. American Society for Testing and Materials, ‘‘Standard Test Method for Bulk Density (‘‘Unit Weight’’) and Voids in Aggregate,’’ (ASTM C 29) Vol. 4.02, 2004, pp. 1–4.

American Society for Testing and Materials, ‘‘Standard Test Method for Slump Flow of Self-Consolidating Concrete,’’ (ASTM C 1611) Annual Book of ASTM Standards, Vol. 4.02, 2005, 36–41. American Society for Testing and Materials, ‘‘Standard Test Method for Passing Ability of Self-Consolidating Concrete by J-Ring,’’ (ASTM C 1621) Annual Book of ASTM Standards, Vol. 4.02, 2005, 42–45. Diawara, H., ‘‘Parametric Study of Self-Consolidating Concrete,’’ Doctoral Dissertation, under the supervision of N. Ghafoori, University of Nevada, Las Vegas (UNLV), USA, 2008, p. 370.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Strength property of concrete using recycled aggregate and high-volume fly ash T. Ishiyama The University of Kitakyushu Graduate School, Fukuoka, Japan

K. Takasu & Y. Matsufuji The University of Kitakyushu, Fukuoka, Japan

ABSTRACT: Taking advantage of strength development and high durability property of concrete using a high volume of fly ash as a part of the fine aggregate, we experimented on strength property that was changed the water-binder ratio and the kind of fly ash in order to make low quality recycled aggregates available as a structural concrete. The conclusions are as follows. When we used concrete using a high volume of fly ash as a part of the fine aggregate, compressive strength, split tensile strength and flexural strength increased until 91 days even though recycled aggregates were used. The more fly ash, compression strength grew although recycled aggregates were used. Static elastic modulus increased when the age passed, in this experiment, while static elastic modulus didn’t increase even if fly ash content per unit volume of concrete is increased. 1

INTRODUCTION

The exhaust of large construction and demolition waste is expected in Japan because construction which led to high economic growth between 1955 and 1973 is beginning to face longevity. It will have much effect on the economic society whether we treat it as construction wastes or unused resources. It is necessary to treat as unused resources in resources recycling society. The construction industry in Japan is working on the recycling of construction waste in this situation. In 2000, there was 85 million tons of construction waste from wood, dirt, asphalt concrete mass, concrete mass and others. 72 million tons of this, or 81%, was recycled. Especially, 96% of concrete mass was recycled1). However, considering the fact of recycling, most of the recycling of concrete masses is used for the road board material. It is necessary to take out the aggregate from concrete masses, and use it for concrete again as a recycled aggregate. The research of the recycled aggregate is being studied by researchers, and the recycled aggregate was standardized by JIS now. About 11 million tons of fly ash was generated by the thermal power plants in Japan in 2006. And the amount is expected to increase in the future. A part of the generated fly ash is used as a cement raw material but many others were buried. Moreover, the securement of securing landfills becomes more difficult every year. The authors had made mixtures that change not cement but fine aggregate into fly ash. It was clarified to show that concrete using a high volume fly ash as a part of the fine aggregate has higher strength and durability properties than mixture with

557

no fly ash concrete. Taking advantage of the strength development and high durability properties of concrete using a high volume of fly ash as a part of the fine aggregate, we experimented on the strength property that was changed in the water-binder ratio and the kind of fly ash in order to make low quality recycled aggregates available as a structural concrete. 2 2.1

EXPERIMENTS Materials

Table 1 shows the materials and Table 2 shows the foreign matters ratio of the recycled coarse aggregate. The recycled coarse aggregate contained about 11% impurities by mass ratio. The recycled coarse aggregate in this study was standardized recycled aggregate class Middle-quality by JIS (Recycled aggregate concrete using recycled aggregate class M). The recycled fine aggregate in this study was standardized Recycled aggregate class Low-quality by JIS (Recycled aggregate concrete using recycled aggregate class L). Table 3 shows physicality of fly ash. We used two kinds of fly ash that is standardized by JIS A 6201 ‘‘Fly ash for concrete’’. The ignition loss and the specific surface area of FA1 were smaller than that of F2. 2.2

Mix proportion of concrete

Table 4 shows the mix proportion of concrete. The unit water content and the unit cement content were fixed 185 kg/m3 and 285 kg/m3 that were the upper limit of JASS5. The aggregate was three combinations.

Table 1.

Materials.

Term

Type

Density (g/cm3 )

Absorption (%)

Solid content (%)

Fineness modulus (%)

C W S

Ordinary portland cement City water Sea sand Recycled Crushed andsite Recycled

3.16 1.00 2.54 (oven-dry) 2.2 (oven-dry) 2.81 (oven-dry) 2.41 (oven-dry)

– – 1.55 4.43 0.79 3.90

– – 60.3 67.6 59.4 61.6

– – 2.7 2.9 7.0 6.8

G

Table 2.

Foreign matters ratio.

Recycled coarse aggregate Table 3.

Recycled coarse aggregate (%)

Tile (%)

Mortar (%)

Glass (%)

Other (%)

88.74

0.45

2.16

0.05

8.6

Physical properties of fly ash.

Symbol

Density (g/cm3 )

Ignition loss (%)

FA1 FA2

2.25 2.21

1.4 2.22

Specific surface area (cm2 /g)

Total moisture content (%)

3410 4200

0.07 0.1

(1) sea sand and coarse aggregate, (2) sea sand and recycled coarse aggregate, (3) recycled fine aggregate and recycled coarse aggregate. The type of fly ash was tree kinds that were no mixture, 244 kg/m3 mixture of two kinds and 455 kg/m3 mixture of F1. The mix proportion is 12 kinds in total. The flowability of fresh concrete was that slump was 18 ± 2.5 cm when unit fly ash contents was 0 and 244 kg/m3 , slump flow is 50 cm–70 cm when unit fly ash contents is 455 kg/m3 . The range of amount of air was 4.5 ± 1.0%. 2.3

Test methods

Table 5 shows the methods. Each examination item and method was conducted according to the material age and curing in the table. Manufacture method of concrete specimen was conformed to JIS A 1132. 3 3.1

EXPERIMENTAL RESULTS AND DISCUSSIONS Compressive strength

Figure 1 shows the change of compression strength when FA1 was used. When we used concrete using

a high volume of fly ash as a part of the fine aggregate, compressive strength increased within a 91 day period in spite of the kind of the aggregate. The more fly ash, compression strength increased although recycled aggregates were used. Compression strength of all mix proportion increased when fly ash was mixed, compression strength of concrete using the recycled aggregate was smaller than that of concrete using the normal aggregate. The compression strength of concrete using the recycled aggregate was smaller than that of concrete using the normal aggregate at 91 days. Compression strength of concrete using the recycled aggregate was higher than that of concrete using the normal aggregate on W/B = 35% at 28 days, so the influence of the aggregate quality was hard to appear. There is no difference of compression strength of RN concrete and RR concrete cause of using a high volume of fly ash as a part of the fine aggregate. The increase rate of compression strength of concrete using a high volume of fly ash as a part of the fine aggregate in a period between 91 and 180 days was higher than that of no fly ash concrete. Figure 2 shows the relation between the unit coal ash content and compression strength in a period between 3 and 91 days when FA1 was used. In all the material ages, compression strength has increased with the unit coal ash content. This increase of compression strength appeared on the both concrete using the normal aggregate and concrete using the recycled aggregate. As for the no coal ash concrete, compression strength of concrete using the normal aggregate is smaller than that of concrete using the normal aggregate. Compression strength of concrete using the recycled aggregate and a high volume of fly ash as a part of the fine aggregate could conform to that of no fly ash concrete using the normal aggregate. Figure 3 shows the change of compression strength when FA1 and FA2 were used. By the kind of the fly ash, compression strength of the NN concrete was different in the same material age. However, compression strength differential of RN concrete become small in the same material age by the kind of the fly ash, and that of the RR concrete disappeared almost. From here onwards, the recycled aggregate was hard to appear compressive strength differential in this study in even if a different kind of fly ash were used.

558

Table 4.

Mix proportion of concrete.

Mixture

Unit content (kg/m3 )

Chemical admixture (C + FA) * %

W

AE agent 0.005 0.005 0.005 0.001 0.008 0.01 0.025 0.025 0.014 0.05 0.05 0.05

Type of aggregate Additive

Fine

Coarse

Symbol

No mixture

Normal Normal Recycled Normal Normal Recycled Normal Normal Recycled Normal Normal Recycled

Normal Recycled Recycled Normal Recycled Recycled Normal Recycled Recycled Normal Recycled Recycled

NN-65 RN-65 RR-65 NN1-35 RN1-35 RR1-35 NN2-35 RN2-35 RR2-35 NN1-25 RN1-25 RR1-25

FA1

FA2

FA1

W/B (%)

0

65

35

185

Age (day)

Compressive strength Static elastic modulus Split tensile strength Flexural strength

JIS A 1108 JIS A 1149 JIS A 1113 JIS A 1106

3, 7, 28, 91

G

841 841 771 561 561 514 561 561 514 318 318 292

974 863 863 974 863 863 974 863 863 974 863 863

0.35 0.35 0.35 1.05 0.6 0.6 0.5 0.5 0.43 1.5 1.5 1.5

NN1 (7 days) RN1 (7 days) RR1 (7 days) NN1 (3 days) RN1 (3 days) RR1 (3 days)

2

20◦ C water 7, 28, 91 28, 91

NN-65 RN-65 RR-65 NN1-35 RN1-35 RR1-35 NN1-25 RN1-25 RR1-25

2

S

High range AE waterreducer

NN1 (91 days) RN1 (91 days) RR1 (91 days) NN1 (28 days) RN1 (28 days) RR1 (28 days)

60

70

Compressive strength (N/mm )

244

455

70

Test method Curing

50

285

25

Test methods.

60

FA

65

Seletion

50 40 30 20 10 0

40 30

0

100 200 300 400 3 Unit coal ash content (kg/m )

500

Figure 2. Relation between the unit coal ash content and compression strength.

20 10 0

Figure 1.

3.2

C

Compressive strength (N/mm )

Table 5.

W/C (%)

0

20

40 60 Age (days)

80

100

Change of compressive strength.

Static elastic modulus

Figure 4 shows the change of the static elastic modulus. The static elastic modulus increased with material age in all mixture. After 7 days, NN series was the largest static elastic modulus regardless of a volume

of fly ash as a part of the fine aggregate. It became clear that static elastic modulus was influenced the aggregate quality. Compared to the RR series and the RN series, the static elastic modulus of concrete using a high volume of fly ash was higher than that of no fly ash concrete. Figure 5 shows the relation between the unit fly ash content and the static elastic modulus. In case of using a high volume of fly ash as a part of the fine aggregate, static elastic modulus increased roughly within a 91 day period even though recycled aggregates were used. In the case of W/B = 35% and W/B = 25%, the static elastic modulus was almost the same value within a 91 day period regardless of the aggregate

559

4

60

Static elastic modules ( 104 N/mm2)

Compressive strength (N/mm2)

50 40 30 20 10 0

NN1–35 NN2–35 NN–65

0

20

40 60 Age (days)

80

3.5

3

2.5

2

100

1.5

(a) Normal fine + Normal coarse

0

40 60 Age (days)

80

100

40

4 30 20 10 0

RN1–35 RN2–35 RN–65

0

20

40 60 Age (days)

80

100

(b) Normal fine + Recycled coarse 60 50 Compressive strength (N/mm2)

20

Figure 4. Change of static elastic modules and the static elastic modules.

50

Static elastic modules ( 104 N/mm2)

Compressive strength (N/mm2)

60

3.5 3 2.5 2

1

40

NN1 (91 days) RN1 (91 days) RR1 (91 days) NN1 (28 days) RN1 (28 days) RR1 (28 days)

1.5

0

RR (3 days)

100 200 300 400 Unit fly ash content (kg/m3)

500

Figure 5. Relation between Unit fly ash content and static elastic modules strength.

30 20 10 0

RR1–35 RR2–35 RR–65

0

20

40 60 Age (days)

80

100

(c) Recycled fine + Recycled coarse

Figure 3. used.

Change of compressive when FA1 and FA2 were

quality. Moreover, the static elastic modulus of the same material age was a big value in order of the NN series, the RN series and the RR series. Therefore, it was able to be confirmed that static elastic modulus is influenced by the aggregate quality than the amount of fly ash.

3.3

Relation between compression strength and static elastic modulus

Figure 6 shows the relation between compression strength and the static elastic modulus of concrete using the recycled aggregate and a high volume of fly ash. The relation between compression strength and the static elastic modulus of concrete using the normal aggregate and a high volume of fly ash as a fine aggregate was that static elastic modulus was higher than compression strength with reference to the expression of the NEW-RC (36 N/mm2 or more) and the criteria expression of AIJ (36 N/mm2 or less) when γ = 2.1 (weight of unit volume of concrete), K1 = 1.0 (influence with the aggregate) and K2 = 1.1. Concrete using the recycled aggregate and a high volume of fly ash as a part of fine aggregate distributed smaller than

560

5

3.5 Split tensile strength (N/mm2)

Static elastic modules ( 104 N/mm2)

4

3 2.5 2

NN1 RN1 RR1 NN-65 NEW-RC equation

1.5 1

Standard equation

0

Figure 6.

0

10

3

2 NN1 (91 days) RN1 (91 days) RR1 (91 days) NN1 (28 days) RN1 (28 days) RR1 (28 days) NN1 (7 days) RN1 (7 days) RR1 (7 days)

1 of AIJ

NEW-RC equation ( =2.1,K1=1,K2=1.1) Standard equation of AIJ ( =2.1)

0.5

4

0

20 30 40 50 60 2 Compressive strength (N/mm )

0

100 200 300 400 Unit coal ash content (kg/m3)

500

70

Figure 8. Relation between the unit fly ash content and split tensile strength.

Relation between compression strength. Split tensile strength(N/mm2)

6 Split tensile strength (N/mm2)

5

4

3

2

1

0

0

Figure 7.

20

40 60 Age (days)

80

Carrasquillo4)

4 Noguchi and Tomosawa

5)

3 2 1

6)

Karino

0

0

NN1 RN1 RR1 NN 65

20 40 60 80 100 Compressive strength (N/mm2)

120

Figure 9. Relation between compression strength and split tensile strength.

100

Change of split tensile strength.

concrete using the normal aggregate. Based on the expression of the NEW-RC and the criteria expression of AIJ, values of concrete using the recycled aggregate and a high volume of fly ash was distributed on the safety side.

fine aggregate increased with the material age within a 91 day period though recycled aggregates were used. It is recognized that split tensile strength of concrete using the recycled aggregate after a 28 day period did not increase remarkably like compressive strength even if the mixed quantity of fly ash increased. 3.5

3.4

5

Split tensile strength

Figure 7 shows the change of split tensile strength. Other than RR1-25, Split tensile strength increased with the material age. Split tensile strength of RR did not increase in a period between 28 and 91 days. In spite of the aggregate variety, split tensile strength of concrete using a high volume of fly ash was higher than that of NN-65 that is no fly ash in the material age within a 91 day period. Figure 8 shows the relation between the unit fly ash content and split tensile strength. Split tensile strength of concrete using a high volume of fly ash as a part of

561

Relation between compressive strength and split tensile strength

Figure 9 shows the relation between compression strength and split tensile strength. Split tensile strength was within the range of 1/15∼1/8 of compression strength regardless of the unit coal ash content, the aggregate quality or the material age. Those values was an equal value to the regression expression by the research in the past. It had been understood that the relation between compression strength and split tensile strength of concrete using recycled aggregate and a high volume of fly ash showed a similar tend to that of no fly ash concrete using the normal aggregate.

4

8.5

2

Flexural strength (N/mm )

8 7.5 7 6.5 6 5.5 5 4.5 20

Figure 10.

30

40

50 60 70 Age (days)

80

90

100

Change of flexural strength.

8.5

2

Flexural strength (N/mm )

8 7.5 7 6.5 6 5.5

NN1 (91 days) RN1 (91 days) RR1 (91 days) NN1 (28 days) RN1 (28 days) RR1 (28 days)

5

CONCLUSIONS

1. As for concrete using a high volume of fly ash as a part of the fine aggregate, compressive strength increased until 91 days even if recycled aggregates were used. As fly ash content per unit volume of concrete increased, compression strength grew though recycled aggregates were used. It is thought that concrete became elaborateness because sea sand was substituted for a fine powder like fly ash. 2. Static elastic modulus of concrete using the recycled aggregate and a high volume of fly ash as a fine aggregate increased when the age passed, in this experiment. However, static elastic modulus of concrete using the recycled aggregate and a high volume of fly ash as a fine aggregate were smaller than that of concrete using the fine aggregate. 3. The relation between compressive strength and static elastic modulus of concrete using the recycled aggregate and a high volume of fly ash as a part of the fine aggregate was distributed from the NEW-RC equation (in Japan). (4) Split tensile strength and flexural strength of concrete using recycled aggregates and fly ash as a part of fine aggregate increased until 91 days, but they did not increase remarkably like compressive strength even if fly ash content per unit volume of concrete increased. From now on, it is necessary to accumulate the experimental data and to examine this tendency. ACKNOWLEDGMENTS

Figure 11. Relation between the unit fly ash content and flexural strength.

The authors acknowledge the assistance in this work provided by Mr. E. Mikura, Mr. T. Hayashida and Mr. T. Kitamura. Provision of materials by Kyoboshi Co. is gratefully acknowledged. This research was financially supported by Grant-in-Aid for Scientific Research (Young Scientists B) No. 19760395.

3.6

REFERENCES

4.5

0

100

200

300

400

500

3

Unit coal ash content (kg/m )

Flexural strength

Figure 10 shows change of flexural strength. In all mixture, flexural strength at 91 days was higher than flexural strength at 28 days. The flexural strength could be expected to increase with the material age.Figure 11 shows the relation between the unit fly ash content and flexural strength. As for concrete using a high volume of fly ash, it was clarified that flexural strength increased even if recycled aggregates were used within a 91 day period. Flexural strength of W/B = 35% was higher than that of W/B = 25% in the RR series. We found that flexural strength did not increase remarkably like compressive strength even if unit coal ash content increased.

Matsufuji, Y., et al. 2001. Study on Mixed of Concrete Using a high volume of fly ash as a part of the Fine Aggregate. Journal of Concrete Engineering 12 (2): 51–60. Architectural Institute of Japan. 2005. High-strength Concrete Guide for Construction (Scheme). Exposition. Tokyo: AIJ. Carrasquillo., et al. 1981. Properties of High Strength Concrete Object to Short-Term Load. Journal of ACI (78)3: 171–178. Noguchi, T. and Tomosawa, F. 1995. Relationship between Compressive Strength Concrete. Journal of Structul and Construction Engineering 4: 1–6. Karino, H. 1940. Study on tension of Concrete. Journal of Architectural Institute 4: 1–6.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Strength, sorptivity and carbonation of geopolymer concrete A.A. Adam Tadulako University, Palu, Indonesia

T.C.K. Molyneaux & I. Patnaikuni RMIT University, Melbourne, Victoria, Australia

D.W. Law Herriot Watt University, Edinburgh, Scotland, UK

ABSTRACT: Several studies have shown that alkali activated binders can achieve similar strengths to both ordinary Portland cement (OPC) and blended cements. However, to date little research has been undertaken on their durability properties. This study investigated the influence of activators concentration and alkali modulus on strength, sorptivity and carbonation of alkali activated slag (AAS) and fly ash (FA) based geopolymer concrete. The same tests were also conducted on blended concrete with 30%, 50%, and 70% OPC replacement with ground granulated blast-furnace slag (GGBS) along with control concrete. Results indicate that the alkali modulus has a major effect on sorptivity of both AAS and geopolymer, however no significant effects of the alkali modulus on carbonation was observed on AAS concrete. The phenolphthalein indicator gave no clear indication between carbonated and non-carbonated area in geopolymer specimens. The sorptivity of blended concrete reduced but the carbonation increased as the replacement level increased. 1

INTRODUCTION

It is widely known that the production of Portland cement consumes high energy and contributes large quantities of CO2 to the atmosphere. However, at present Portland cement is still the main binder in concrete construction and the search for more environmentally friendly materials is essential. One possible alternative is the use of alkali-activated binder using industrial by-products containing silicate materials (Philleo 1989). The most common industrial by-products used as binder materials are fly ash (FA) and ground granulated blast furnace slag (GGBS). GGBS has been widely used as a cement replacement material due to its latent hydraulic properties, while fly ash has been used as a pozzolanic material to enhance physical, chemical and mechanical properties of cements and concretes. GGBS is a latent hydraulic material which can react directly with water, but requires an alkali activator. In concrete, this is the Ca(OH)2 released from the hydration of Portland cement. While FA is a pozzolanic material which reacts with Ca(OH)2 from Portland cement hydration forming calcium silicate hydrate (C-S-H) as the hydration product of Portland cement. Thus, when used with Portland cement, GGBS or FA will not start to react until some Portland cement hydration has taken place. This delay, causes blended

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Portland cements to develop strength more slowly at early ages than Portland cement alone. Recent research has shown that it is possible to use fly ash or slag as a sole binder in mortar by activating them with an alkali component, such as; caustic alkalis, silicate salts, and non silicate salts of weak acids (Talling and Branstetr 1989). There are two models of alkali activation. Activation by low to mild alkali of a material containing primarily silicate and calcium will produce calcium silicate hydrate gel (C-S-H), similar to that formed in Portland cements, but with a lower Ca/Si ratio (Bakharev and Patnaikuni 1997). The second mechanism involves the activation of material containing primarily silicate and aluminates using a highly alkaline solution. This reaction will form an inorganic binder through a polymerization process (Xu 2002). The term ‘‘Geopolymeric’’ is used to characterize this reaction from the previous, and accordingly, the name ‘‘geopolymer’’ has been adopted for this type of binder (Davidovits 1991). The geopolmeric reaction differentiates geopolymer from other types of alkali activated materials (such as; alkali activated slag/fly ash since the product is polymer rather than a C-S-H gel. In order to compare the strength, sorptivity, and carbonation of blended OPC-GGBS, alkali activated slag (AAS) and geopolymer concrete; some concrete specimens were prepared with a range of OPC and

GGBS ratios of (0, 30, 50 and 70% GGBS), while others prepared with GGBS and fly ash activated by alkaline solution with different alkali modulus.

2 2.1

MATERIALS Cementitious materials

The GGBS supplied conformed to AS 3582.2-2001. A Scanning Electron Microscope (SEM) image of the GGBS is shown in Fig. 1. The fly ash was a class F fly ash from Gladstone power station conforming to AS 3582.1-1998. A SEM image of the FA is shown in Figure 2. The OPC used in this investigation was general purpose (GP) cement. The chemical analysis of these materials is given in Table 1. 2.2

Alkaline activators

A Grade D sodium silicate solution (Na2 SiO3 ) of 1.53 g/cc density with an alkali modulus (AM) = 2 (Na2 O = 14.7% and SiO2 = 29.4%) was supplied

Table 1.

Component

Cement

Slag

Fly ash

SiO2 Al2 O3 Fe2 O3 CaO MgO K2 O Na2 O TiO2 P2 O5 Mn2 O3 SO3 S2− Cl

19.9 4.62 3.97 64.27 1.73 0.57 0.15 0.23

33.45 13.46 0.31 41.74 5.99 0.29 0.16 0.84 0.12 0.40 2.74 0.58 0.01

49.45 29.61 10.72 3.47 1.3 0.54 0.31 1.76 0.53 0.17 0.27 0.21 0.001

SEM image of GGBS used in this study.

Figure 2.

SEM image of Fly ash used in this study.

0.06 2.56

by PQ Australia. Sodium hydroxide solution (NaOH) was prepared by dissolving sodium hydroxide pellets in deionised water.

3

Figure 1.

Composition of cementitious materials (%).

MIX PROPORTIONS AND TEST SPECIMENS

A w/b ratio of 0.5 was used to prepare all blended GGBS-OPC and control concrete. Table 2 shows the mix design of control and blended GGBS-OPC concrete. The proportions of GGBS were 30%, 50%, and 70% of the total binder A water/solid ratio of 0.45 was used for AAS and 0.29 for geopolymer concrete. In the case of AAS and FA based geopolymer concrete, the amount of water in the mix was the sum of water contained in the sodium silicate, sodium hydroxide and added water. The amount of solid is the sum of GGBS or FA, the solid in the Na2 SiO3 solution, and the NaOH pellets. The detailed mixes of the AAS and FA based geopolymer concrete are shown in Tables 3 and 4. Liquid sodium silicate and sodium hydroxide were blended in different proportions (Table 5), providing an alkali modulus (AM) in solution (mass ratio of SiO2 to Na2 O) ranging from 0.75 to 1.25. The alkali concentrations (Percentage of Na2 O by mass of binder) in the solution, were 5% for AAS and 7.5% for FA based geopolymer concrete. The mixing was performed using a 120-liter mixer, the mix was then poured into 100 mm diameter × 200 mm high cylinder moulds and vibrated for 1 minute. The blended GGBS-OPC, control, and AAS concrete specimens were demoulded after 24 hours followed by water curing at 20◦ C for 6 days and then exposed to the laboratory environment (26◦ C and 40%RH) prior to testing. The geopolymer specimens did not achieve structural integrity at room temperature. As such the curing regime was 24 hours at room temperature, followed 24 hours at 80◦ C (covered with clingfilm). The specimens were then allowed to cool in the mould

564

Table 2.

Details of the blended concrete mixes (kg/m3 ). Binder

Aggregate

Mix

OPC

GGBS

Sand

7 mm

10 mm

Water

CTL S30 S50 S70

428 296 210 125

– 127 210 293

784 784 784 784

346 346 346 346

693 693 693 693

222 220 219 217

Table 3.

4.2

Alkaline sol.

GGBS Sand 7 mm 10 mm Na2 SiO3 NaOH Water

AAS50.75 419 AAS51.00 415 AAS51.25 412

Table 4.

I = A + St 1/2 784 346

693

53

56

137

784 346

693

71

46

136

784 346

693

87

33

135

Alkaline sol.

FA Sand 7 mm 10 mm Na2 SiO3 NaOH Water

G7.5-0.75 476 784 346 G7.5-1.00 467 784 346 G7.5-1.25 461 784 346

Table 5.

693 693 693

90 119 147

95 75 56

40 38 36

Proportion of alkaline activators. Binder (%)

Alkaline solution

Mix

GGBS

FA

Na2 O/binder (%)

AM (SiO2 /Na2 O)

AAS5-0.75 AAS5-1.00 AAS5-1.25 G7.5-0.75 G7.5-1.00 G7.5-1.25

100 100 100 – – –

– – – 100 100 100

5 5 5 7.5 7.5 7.5

0.75 1.00 1.25 0.75 1.00 1.25

4.1

4.3 Depth of carbonation test For depth of carbonation test, the 100 mm diameter × 200 mm high concrete cylinder were cut into three parts. In order to keep carbonation direction in the radial direction, top and bottom of each specimen were coated with epoxy. The specimens were transferred to a specially designed chamber to accelerate the carbonation process. The chamber was supplied with carbon dioxide (CO2 ) to maintain a CO2 level of 20%. A saturated NaCl solution was used to maintain the humidity level between 75%–80%. At weekly intervals specimens were split and then sprayed with a phenolphthalein indicator, as prescribed by RILEM (1994). An average carbonation depth was then taken from the cross-sectioned slices.

5 5.1

before being demoulded. The specimens were then left in the laboratory environment until testing.

4

(1)

where I is the cumulative absorbed volume after time t per unit area of inflow surface, I = Dw/ar, Dw being the increase in weight, a the cross-sectional area and r the density of water.

Details of the FA based geopolymer mixes (kg/m3 ). Aggregate

Mix

Sorptivity test

The Sorptivity tests were undertaken in accordance with DIN 52617. The sides of the specimens were coated with epoxy to allow free water movement only through the bottom face (unidirectional flow). The results were plotted against the square root of the time to obtain a slope of the best fit straight line. According to Hall (1989), the penetration of water under capillary action can be modeled by:

Details of the AAS mixes (kg/m3 ). Aggregate

Mix

regime with a loading rate of 20 MPa/min. Three to five cylinders were tested for each data point. The specimens were tested at 7, 28 and 90 days after casting.

TEST PROGRAM Compressive strength test

Compressive strength measurements of concretes were performed on an MTS machine under a load control

RESULTS AND DISCUSSIONS Comparison of strength

The strength of blended, AAS, and FA based geopolymer concrete are shown in Tables 6, 7, and 8 respectively. In general, the 28-days compressive strengths of the AAS and FA geopolymer concretes are comparable with that of 100% OPC concrete and blended OPC-GGBS concretes as shown in Figures 3, 4 and 5. It should be noted that heat curing was applied to the FA geopolymer concrete to achieve structural integrity. Heat curing in general will result in increased early strengths. As such a comparison of the 28 and 90 days strengths will give a better assessment of the comparable strengths, than the 7 days data.

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Compressive strength of blended concretes. Cementitious (%)

Strength (MPa)

Mix

OPC

GGBS

7d

28d

90d

CTL S30 S50 S70

100 70 50 30

0 30 50 70

36 32 29 26

50 47 47 36

57 50 53 43

Table 7.

Compressive strength (MPa)

Table 6.

Compressive strength of AAS concretes.

60

7 days

90 days

40 30 20 10 0 AAS5-0.75

AAS5-1

AAS5-1.25

Strength (MPa)

Na2 O (%)

AM (SiO2 /Na2 O)

7d

28d

90d

AAS5-0.75 AAS5-1.00 AAS5-1.25

5 5 5

0.75 1.00 1.25

25 36 42

33 44 46

37 43 47

Figure 4.

compressive strength (MPa)

Mix

Table 8.

28 days

50

Compressive strength of Geoplymer concretes. Strength (MPa)

Mix

Na2 O (% )

AM (SiO2 /Na2 O)

7d

28d

90d

G7.5-0.75 G7.5-1.00 G7.5-1.25

7.5 7.5 7.5

0.75 1.00 1.25

39 50 52

44 43 57

44 54 57

Strength development of AAS concrete. 7 days

28 days

90 days

60 50 40 30 20 10 0

Figure 5.

G7.5-0.75

G7.5-1

G7.5-1.25

Strength development of geopolymer concrete.

Compressive strength (MPa)

70 7 days

60

28 days

90 days

50 40 30 20 10 0 CTL

Figure 3. concrete.

S30

S50

S70

Strength development of blended GGBS-OPC

The AM of the activator has a significant influence on the strength of AAS and FA-based geopolymer concrete up to AM = 1, beyond this level the influence reduced. The strength of AM = 1.25 geopolymer was slightly higher than that of AM = 1, by contrast the strength of AM = 1.25 AAS was slightly lower than that of AM = 1 AAS concrete specimens. In comparison, the blended OPC-GGBS developed strength slowly at an early age, and decreased

in strength as the level of replacement increased. At 28 days age, the strength of 30% and 50% blended OPC-GGBS concretes were constant, but the strength reduced at 70% replacement. At 90 days the strength of the 50% blended OPC-GGBS concrete was highest, with the 70% again the lowest. The 100% OPC control concrete displayed a higher strength than the blended concretes at 7, 28 and 90 days. It is expected that the blended concretes will exhibit higher strengths as the time increases. It should be noted that the water curing only applied for 7 days, this delayed the strength development of blended as the hydration of slag is more sensitive to water curing than those for OPC concrete. The hydration of slag requires Ca(OH)2 from Portland cement hydration, and it will not start until the hydration of OPC has taken place. For FA based geopolymer concrete most of the strength were gained by 7 days and no further increase in strength was observed up to 28 days, this was attributed to the heat curing. As for the AAS concrete, the alkali modulus of the activator has a significant influence on the strength of the FA-based geopolymer concrete up to an AM = 1. Beyond this limit the influence was marginal. Increasing the alkali modulus in these examples resulted in

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an increase in soluble silicates and consequently an increase in the reaction rate (a higher concentration of reactants induces a higher reaction rate). Overall the FA based geopolymer displayed higher strengths than the AAS concrete specimens 5.2

There was a large reduction in sorptivity of both AAS and geopolymer concrete as the alkali modulus increased from 0.75 to 1.00. However only a small reduction in sorptivity was observed as the alkali modulus increased from 1.00 to 1.25 for FA based geopolymer concrete, and no further reduction observed for AAS concrete. The results show that there was an optimum alkali modulus for both AAS and geopolymer concrete, in this case the value was AM = 1. Increasing the alkali modulus, which also means increasing the SiO2 content in the system will reduced the porosity of both AAS and FA-based geopolymer concrete. The FA geopolymer concrete specimens display a significantly lower value than both the AAS concrete and the control concrete. As the level of GGBS replacement increases the sorptivity becomes comparable with the FA based geopolymer.

Sorptivity

1/2

Sorptivity (mm/min )

The results of the sorptivity tests are presented in Figures 6–8. The results for blended GGBS-OPC shows that the reduction of sorptivity was proportional to the level of GGBS replacement. 0.18 0.16

56 days

90 days

0.14 0.12 0.10 0.08

5.3

0.06 0.04 0.02 0.00 CTL

Figure 6.

S30

S50

S70

Sorptivity of blended GGBS-OPC concrete.

1/2

Sorptivity (mm/min )

0.30 56 days

90 days

0.25 0.20

Depth of carbonation

A summary of the results for the accelerated carbonation at 20◦ C and 75%–80% RH is shown in Figure 9. It can be seen that the rate of carbonation was higher when the level of GGBS replacement increased. The result shows that in blended OPC-GGBS, the carbonation rate is not primarily influenced by the porosity of the concrete, as can be observed in OPC concrete. Carbonation in normal concrete is regarded as a reaction that takes place between carbon dioxide (CO2 ) and Ca(OH)2 . This reaction, that takes place in aqueous solution, which can be written as (Bertolini et al. 2004):

0.15

CO2 + Ca(OH)2

0.10

H2 O, NaOH

−→

0.00 AAS5-0.75

AAS5-1

AAS5-1.25 CTL AAS5-0.75

Depth of carbonation (mm)

1/2

Sorptivity (mm/min )

56 days

0.10

S30 AAS5-1

S50 AAS5-1.25

S70

40

Sorptivity of AAS concrete.

0.12 90 days

0.08 0.06 0.04 0.02

35 30 25 20 15 10 5 0 0

0.00 G7.5-0.75

Figure 8.

(2)

Regardless of the alkali modulus, the depth of carbonation of alkali activated slag was even higher than

0.05

Figure 7.

CaCO3 + H2 O

G7.5-1

1

2

3

4

5

6

7

8

9

week

G7.5-1.25

Figure 9. Depth of carbonation of blended OPC-GGBS and AAS concrete at 20% CO2 (20◦ C and 75–80% RH).

Sorptivity of FA-based geopolymer concrete.

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– There was a large reduction in sorptivity of both AAS and geopolymer concrete as the alkali modulus increased from 0.75 to 1.00. – The carbonation mechanism in AAS concrete is different from that of OPC concrete since the reactant is C-S-H whereas it is Ca(OH)2 in OPC. In blended OPC-GGBS concrete the mechanisms appears to be a combination of both. – The phenolphthalein gave no clear indication between carbonated and non-carbonated area in geopolymer specimens.

that of blended OPC-GGBS, and about three times that of control concrete. Comparing the rate of carbonation in OPC, blended GGBS-OPC, and AAS concrete, as seen in Figure 9, the carbonation of OPC was almost constant after week 6, whereas AAS continue to carbonate at a relatively high rate. The carbonation of blended OPC-GGBS also continues to occur although at a slower rate compared to the AAS concrete. In blended OPC-GGBS concrete, after initial carbonation, the product of this reaction fills up the pores resulting in a higher density of the matrix and further slowing the diffusion of CO2 . Since the matrix of the AAS concrete contains marginal volumes of Ca(OH)2 , the main carbonation reactant must be coming from another source. According to Peter et al. (2008), other constituents in the concrete can also carbonate, particularly calcium-silicate hydrates (C-S-H). The latter is believed to be the main carbonation reactant in AAS concrete, whereas in blended OPC-GGBS concrete it is the combination of the two. The phenolphthalein indicator gave no clear border between colour and colourless area for geopolymer as seen in control, blended GGBS-OPC and AAS concrete. Therefore it was not possible to measure the carbonation depth. This was attributed to the polymeric type reaction of the geopolymer, which did not produce either the C-S-H gel or the Ca(OH)2 . Further testing is being undertaken on the pH and carbonation of the geopolymer concretes.

6

REFERENCES

CONCLUSIONS

– The blended OPC-GGBS concrete gains strength more slowly than the OPC concrete for the same water cement ratio. – The early strength of 1.0 AM and above for FAbased geopolymer concrete was considerably higher than that of OPC concrete and similar for 28-days strength. This is attributed to the heat curing for the FA geopolymer concrete – Increasing the alkali modulus (AM) up to 1 enhances strength but further increases of AM have minimal impact on the strength of FA based geopolymer concrete and reduction on strength for AAS concrete.

Bakharev, T. and Patnaikuni, I., 1997. Microstructure and durability of alkali activated cementitious pastes. In: K.C.G. Ong (Editor), The Fifth International Conference on Structural Failure, Durability and Retrofitting. Singapore Concrete Institute, Singapore, p. 200. Bertolini, L., Elsener, B., Pedeferri, P. and Polder, R., 2004. Corrosion of Steel in Concrete. WILEY-VCH Verlag GmbH & Co. KGaA, Weinheim, p. 392. Davidovits, J., 1991. GEOPOLYMERS: INORGANIC POLYMERIC NEW MATERIALS. Journal of Thermal Analysis, 37: 1633–1656. Hall, C., 1989. Water sorptivity of mortars and concretes: a review. Magazine of Concrete Research 41(147): 51–61. Peter, M.A., Muntean, A., Meier, S.A. and Böhm, M., 2008. Competition of several carbonation reactions in concrete: A parametric study. Cement and Concrete Research, 38(12): 1385–1393. Philleo, R.E., 1989. Slag or other supplementary materials? In: V.M. Malhotra (Editor), Third international conference on the use of fly ash, silica fume, slag and natural pozzolan in concrete. American Concrete Institute, Trondheim, Norway, pp. 1197–1208. RILEM, 1994. CPC 18. Measurement of hardened concrete carbonation depth, 1988. In: RILEM (Editor), RILEM Recommendations for the Testing and Use of Constructions Materials. E & FN SPON, pp. 56–58. Talling, B. and Brandstetr, J., 1989. Present State and Future of Alkali-Activated Slag Concretes. In: V.M. Malhotra (Editor), third international conference on fly ash, silica fume, slag, and natural pozzolans in concrete. Publication SP; 114. American Concrete Institute, Trondheim, Norway, pp. 1519–1545. Xu, H., 2002. Geopolymerisation of Aluminosilicate Minerals, The University of Melbourne, Melbourne.

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Suitability of some Ghanaian mineral admixtures for masonry mortar formulation M. Bediako & E. Atiemo Building and Road Research Institute, Kumasi, Ashanti Region, Ghana

S.K.Y. Gawu & A.A. Adjaottor Kwame Nkrumah University of Science and Technology, Kumasi, Ashanti Region, Ghana

ABSTRACT: The suitability of masonry mortar for various constructional applications is dependent on some vital engineering properties and production cost. In a majority of masonry formulations, ordinary Portland cement (OPC) is the principal binding agent. However, the current trend of cement cost in Ghana has rendered masonry mortar formulation quite expensive. In this paper clay pozzolana produced from cost efficient local technology and limestone powder was used as admixtures in cement. Physical and chemical properties of the mineral admixtures were analyzed. The particle sizes of the materials were also investigated. Binary and ternary pastes and mortars were formulated using some percentages of pozzolana, limestone and pozzolana-limestone to replace part of the expensive ordinary Portland cement. Water demand and setting time tests were determined on the binder paste whilst compressive strength test was performed on mortars cured in water for 7 and 28 days. Test results indicated that ASTM type M and S mortars could be formulated from binary and ternary mortar mixes. 1

INTRODUCTION

The utilization of mineral admixtures as cement replacement materials is now widely adopted in USA, Europe and part of Asia for the production of mortars for masonry works like plastering, rendering and jointing of bricks, blocks and stones (Saad et al. 1982; Malhotra et al. 1995). Examples of such mineral admixture commonly used are natural resources like pozzolans and limestone filler, heat-activated additions such as clay and metakaolin and industrial byproducts like fly ash, blast-furnace and steel-making slag, silica fume, rice husk ash (Mehta & Monteriro, 1987; Kim et al. 2007). Many authors have found that the use of mineral admixtures in Portland cement for mortar formulation is very beneficial in areas like cost reduction, improved mechanical properties, reduction in heat evolution, decreased permeability, increased chemical resistance and reduction in gas emission that contribute to green-house effect (Carrasco et al. 2005; Tagnit-Hamou, 2003; Kadri et al. 2002). In most developing countries like Ghana, the technology of using mineral admixtures in mortar formulation is not well known. This is because Portland cement mortars are the most widely used traditional mortar for masonry works. However, Portland cement cost is among the limiting constraints for most builders and engineers in terms of construction cost. In 2005,

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Portland cement sold for $5.50/50 kg bag and currently in 2009 is sold at $10.50 which represents about 91% increase in cost. It is estimated that by 2010 cement price/50 kg will be between $18.00 and $25.00. The annual price increase of Portland cement is attributed to the importation of clinker and gypsum, the main ingredients for cement production from Europe and Asia. Clinker and gypsum importation is said to cost the country about $200 million annually (Annon, 2007). The need to develop alternative materials to reduce the cost of Portland cement mortar formulation in the country is thus necessary. Clay and limestone minerals are found in abundance in Ghana and could be processed into cementitious materials used in mortar formulation. It is estimated that about 1392 and 215 million tonnes of clay and limestone respectively exist untapped (Kesse, 1985). Previous studies by Atiemo (2005) has identified and evaluated some clays in the country that are suitable for clay pozzolana formulation. Information regarding the use of limestone for masonry mortars is very rare. In this study clay pozzolana produced from Mankranso clay in Ashanti region and limestone from Orterkpolu in the Eastern region were used to formulate binary and ternary mortars. The aim was to produce suitable and alternative blended cement mortar for cost efficient masonry works.

2 2.1

MATERIALS AND METHODS Materials

Ordinary Portland cement (CEM I 42.5N) produced by Ghana cement works (Ghacem) that conformed to EN 197-1 and labeled OPC was used. Clay sample from Mankranso in the Ashanti Region of Ghana was used to produce the pozzolana (MCP) through a calcination process whereas limestone from Orterkpolu (OL) in the Eastern region of Ghana was crushed in a jaw crusher and milled for utilization. Tables 1 and 2 represent the physical and chemical properties of the powder mineral specimen. The particle size analysis of the powder samples done by the sedimentation method in accordance with the BS 1377 produced a graph as shown in Figure 1. The mean particle size of MCP and OL were 30 μm and 32 μm respectively which values were higher than that of OPC which was 4 μm. The specific gravity values of MCP and OL are close being 2.58 and 2.56 respectively whilst that of OPC was much higher with a value of 3.14 as shown in Table 1.

Table 1.

Figure 1.

Particle size distribution of OPC, MCP and OL.

Figure 2.

X-ray diffraction analysis of MCP.

Figure 3.

X-ray diffraction analysis of OL.

Physical properties of OPC, MCP and OL.

Property

OPC

MCP

OL

Clay (%) Silt (%) Sand (%) Gravel (%) Colour Specific gravity Blaine fineness (m2 /kg) Mean particle size (μm) % passing 75 μm

− − − − grey 3.14 338 4 92

15 79 6 − brown 2.58 410 30 99.6

0 89 11 − white 2.56 420 32 98

Table 2. Chemical composition (wt %) of OPC, MCP, OL and mineralogical composition of OPC. Composition (%) Chemical SiO2 Al2 O3 Fe2 O3 CaO MgO K2 O Na2 O SO3 LOI Mineralogical C3 S C2 S C3 A C4 AF

OPC

MCP

OL

19.7 5.0 3.16 63.03 1.75 0.16 0.2 2.8 2.58

61.89 13.51 5.84 0.21 1.74 1.07 0.14 0.14 10.0

17.65 3.45 1.56 49.57 2.11 0.78 0.3 0.3 23.43

59.6 12.6 7.86 9.49

− − − −

− − − −

It could be deduced from the chemical properties shown in Table 2 that MCP contained 81% of SiO2 + Al2 O3 + Fe2 O3 , 0.14% of SO3 and 10% of LOI which can be classified as a class N pozzolan according to ASTM C618. Orterkpolu limestone (OL) contained 88.5% CaCO3 in calcite form which was calculated from CaO content illustrated in Table 2. Figures 2 and 3 show the X-ray diffraction (XRD) patterns of MCP and OL. According to Figure 2, MCP contains mostly kaolinite and quartz whereas Figure 3 shows the presence of calcite, dolomite and quartz in OL.

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2.2

Methods

2.2.1 Clay pozzolana production The production process for Mankranso pozzolana involved clay winning and drying, dried clay and palm kernel shells milling with a hammer mill, mixing of dry clay and palm kernel shells in a mixer, formation of nodules, nodules drying, calcination of dried nodules in a locally fabricated shaft kiln at 700–800◦ C and finally milling the calcined nodules using a hammer mill. Figure 4 shows the production flow diagram for pozzolana production (Atiemo, 2005) whilst Figure 5 shows the locally fabricated shaft kiln. The production process for OL involved limestone digging and drying, limestone crushing in a jaw crusher, milling of the crushed limestone in a hammer mill into fine powder and finally sieving the powder through a 75 μm sieve size. The undersize was used for the study. 2.2.2 Specimen preparation, casting, curing and testing Binary binder pastes and mortars were prepared by using 10–40% OL or MCP with the remainder being

Table 3.

Mix design for binary and ternary mixes.

Mix

OPC (wt %)

MCP (wt %)

OL (wt %)

L1 L2 L3 L4 L5 L6 P1 P2 P3 P4 P5 P6 T1 T2 T3 T4

90 80 75 70 65 60 90 80 75 70 65 60 70 70 70 50

− − − − − − 10 20 25 30 35 40 10 20 15 30

10 20 25 30 35 40 − − − − − − 20 10 15 20

ordinary portland cement (OPC). The ternary paste and mortars were also prepared using MCP and OL to replace up to 50% OPC. Table 3 shows the mix design for binary and ternary mixes. A 1:3 binder to sand ratio was used for mortar preparation. The water to binder (w/b) ratio for OPC-L was 0.4 whilst 0.5 was used for OPC-MCP and the ternary blend made of OPC-MCP-OL. The mortar was cast into 75 mm metallic cube moulds. Water demand and both initial and final setting times were determined on the binder paste by the vicat apparatus according to EN 197-1 standard. Compressive strength determination was done on an average of 3 mortar specimens at 7 and 28 days of standard curing. This was determined in accordance with ASTM C109 standard.

Figure 4. Production process flow diagram for Mankranso clay pozzolana (Atiemo, 2005).

3

Figure 5.

RESULTS AND DISCUSSIONS

Table 4 indicates the mechanical properties of the binary and ternary paste and mortars. It can be deduced from Table 4 that on addition of between 10 and 40% MCP to OPC, the water demand increased between 10.7% and 42.9% when compared with the control (C). OL addition within the same range increased the water demand between 3.6% and 7% compared to the control. Again there was a slight increase in water demand when % OL content was increased above 30%. For the ternary mixture, deductions from Table 4 showed that water demand increased by 17.6 to 35.7% compared to the control mortar. The results also showed that incorporating Orterkpolu limestone (OL) from 10 to 40%, clay pozzolana (MCP) from 10 to 40% or both at 30 to 50% mineral addition had a relatively high water absorption capability as compared

Locally fabricated brick shaft kiln.

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Table 4. Water demand, setting times and compressive strength of binary and ternary paste and mortar. Setting time (mins)

Compressive strength (MPa)

Mix

Water demand

Initial

Final

7d

28d

C P1 P2 P3 P4 P5 P6 L1 L2 L3 L4 L5 L6 T1 T2 T3 T4

0.28 0.31 0.33 0.35 0.36 0.38 0.4 0.29 0.29 0.29 0.29 0.3 0.3 0.36 0.33 0.35 0.38

83 93 98 89 127 201 214 141 134 140 138 136 136 216 205 209 223

231 221 268 294 310 298 295 174 172 202 183 192 192 259 247 259 228

24.1 16.3 17.6 18.1 17.6 16.7 12.1 22.8 18.6 20.6 13.3 11.2 11.7 17.3 16.3 16.2 10.3

26.0 21.7 20.3 18.5 18.0 17.9 16.2 23.1 20.4 23.2 19.2 17.8 11.7 25.6 24.3 21.3 16.3

C = Control.

to the plain cement paste. This was in conformity with the investigations done by Ahmad and Shaikh (1992). The results of the setting times as shown in Table 4 indicated that between 10 and 25% MCP content, initial setting time got nearer to the control OPC paste. However, further MCP addition delayed the initial setting time of the binder paste up to 40%. The final setting time occured faster at 10% MCP content than the control OPC paste. However further addition of MCP up to 40% also delayed the final setting of the binder paste. The ternary mixture which contained both MCP and OL up to 30% and 50% generally caused a delay in both initial and final setting times as compared to the control paste. Meanwhile OPC replacement at 30% MCP and OL content showed that at 10% MCP and 20% OL, both the initial and final setting times delayed compared to 20% MCP and 10% OL batch. The decrease in final setting time of OL replacement in the mix was consistent with work done by Helal (2002). He explained that, the observation was mainly due to the formation of increased amounts of calcium carboaluminate hydrates, which have a high rate of formation during the early stages of the hydration process. Heikal et al. (2000) explained that limestone addition enhances the formation of Ca(OH)2 at early ages because it provides nucleating sites for its growth. Irrasar et al. (2000) also reported that this effect accelerates the cement hydration process at the early stages. The % MCP replacement levels which caused retardation in the setting times were similar to the work done by Brooks et al. (2000). This was

also evident in 30% MCP and 20% OL batch ternary mix where there was more quantity of pozzolana. Chindaprasirt et al. (2005) reported that generally delayed setting times, particularly initial setting time could be beneficial in hot tropical climates since early stiffening of mortar in hot conditions can lead to cracking and delaminating of masonry mortar. Figures 6–8 and Table 5 indicate the compressive strength figures of OPC-OL, OPC-MCP and OPC-MCP-OL and general recommendations for masonry mortar selection

Figure 6. 28 days compressive strength of OPC-OL masonry mortar compared to ASTM standard.

Figure 7. 28 days compressive strength of OPC-MCP masonry mortar compared to ASTM standard.

Figure 8. 28 days compressive strength of OPC-MCP-OL masonry mortar compared to ASTM standard.

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Table 5. General recommendation on masonry mortar type selection by ASTM C270. Masonry mortar type Location

28 day compressive strength (MPa)

Building segment

Type M Exterior above Load bearing grade walls interior load/non load Bearing walls Type S

Table 6. mortar.

20

4

Exterior at or foundation walls, 14.5 below grade retaining walls, manholes, sewers, pavements, patio, parapet walls

1:3 mortar mix

Cost ($)

Plain

25%

OL 40%

30%

35%

Admixture cost 0.00 20.26 32.42 24.34 28.34 OPC cost, $ 162.09 121.57 97.26 113.41 105.41 Sand cost, $ 33.53 31.44 30.18 31.02 30.60 Total cost, $ 195.63 173.27 159.86 168.77 164.35

respectively. Figure 6 shows that 10 to 30% OL content satisfied type M mortars whilst 35% OL satisfied mortar type S. In Figure 7, 10% and 20% MCP was good for type M mortars whilst 25, 30, 35 and 40% MCP satisfied type S mortars. Figure 8 also illustrated that a ternary blend containing T1, T2 and T3 could produce a type M mortar whilst T4 produced a type S mortar. Binary mortars that contained 30% OL or 20% MCP and 40% OL or 40% MCP best satisfied type M and S mortars respectively. Again in a ternary mixture system containing OL and MCP, up to 30% of the mineral admixtures could be suitable for a type M mortar. 3.1

CONCLUSIONS

Based on the studies the following conclusions were drawn

Cost analysis of 1:3 mortar/m3 of plain and blended

MCP

From Table 6, it is indicated that for a type M mortar containing either 25% MCP or 30% OL could make a savings of 11.43 and 13.7% over plain mortar whilst type S mortar formulation containing either 40% MCP or 35% OL could also make a savings of 18.3 and 16% over plain mortar respectively.

Economic analysis of plain and blended mortars

As already illustrated in Figures 6 and 7, the optimum mix for a type M mortar was at either 30% OL or 25% MCP whereas that for a type S mortar was at 35% OL or 40% MCP respectively. Table 6 represents the cost analysis of the formulation using 1:3 binder to sand ratio of plain and the optimum percentages for MCP and OL mortar mixes. The cost analysis was based on the following assumptions: – Cost of OPC per 50 kg = $8.60 – Cost of MCP or OL = $4.31

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– Mankranso clay was suitable to produce a class N pozzolan according to ASTM C618 – Orterkpolu limestone contained 88.5% CaCO3 as calcite – At 1:3 binder to sand ratio, binary mixture containing up to 30% Orterkpolu limestone (OL) or 25% Mankranso clay pozzolana (MCP) produced type M mortars whereas up to 40% of OL or MCP produced a type S mortar. – Addition of OL, MCP or both had a relatively high water absorption capability compared to the control. – Mankranso clay pozzolana (MCP) blends between 10 and 40% had a higher absorption capability than OL blends. – Final setting time of binary mixture containing OL up to 40% occurred faster than MCP mixture. – In a ternary mixture containing MCP and OL, increasing OL up to 20% caused a faster setting of the paste. – Utilization of Orterkpolu limestone or Mankranso clay pozzolana as mineral admixtures in a ternary or binary mixture has economic benefits when used for masonry mortar formulation.

REFERENCES Ahmed, S.Y. & Shakh, Z. 1992. Portland-pozzolana cement from bagasse: In: Lime and other alternative cements, Ne-ville Hill (ed): 172–179. Atiemo, E. 2005. Production of pozzolana from some local clays-prospects for application in housing construction, Journal of the Building & Road Research Institute 9(1&2): 34–37. Annon, 2007. Minerals in Ghana. A report of the Ministry of Trade and Industry, Ghana. ASTM C109-77. 1979. Compressive strength of hydraulic cement mortars (using 2—in or 50 mm cube specimen)1 . ASTM standards, PA. ASTM C618-03. 2003. Standard specification for coal fly ash and raw or calcined natural pozzolana for use in concrete, ASTM international, PA. British Standard Institution. 2000. Composition, specifications and conformity criteria for common cements, BS EN 197-1(Part 1), BSI London.

Brooks, J.J., Megat Johari, M.A. & Mazloom, M. 2000. Effects of admixtures on the setting times of high strength concrete, Cement and Concrete Composites 22: 293–301. Carrasco, M.F., Menendez, G., Bonavetti, V. & Irassar, E.F. 2005. Strength optimization of ‘‘tailor-made cement’’ with limestone filler and blast furnace slag, Cement & Concrete. Research 35: 1324–1331. Chindaprasirt, P., Buapa N. & Cao, H.T. 2005. Mixed cement containing fly ash for masonry and plastering work, Construction & Building Material 19: 612–618. Heikal, M., El-Didamony, H. & Morsy, M.S. 2000. Limestone-filled pozzolanic cement, Cement. & Concrete Research. 30: 1827–1834. Helal, M.A. 2002. Effect of curing time on the physicomechanical characteristics of the hardened cement pastes containing limestone, Cement & Concrete Research. 32(3): 447–450. Kadri, E.-H. & Duval, R. 2002. Effect of ultrafine particles on heat of hydration of cement mortars, ACI Mat. J 99(11): 138–142.

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Kesse, G.O. 1985. The mineral and rock resources of Ghana, Netherlands, Balkema. Irassar, E.F., Gonzalez, M. & Rahhal, V. 2000. Sulphate resistance of type V cements with limestone filler and natural pozzolana, Cement & Concrete Composites 22: 361–368. Kim, H.-S., Lee, S.-H. & Moon, H.-Y. 2007. Strength properties and durability aspects of high strength concrete using Korean metakaolin, Construction & Building Materials. 21: 1229–1237. Malhotra, V.M. & Hemmings, R.T. 1995. Blended cements in North America—A review, Cement & Concrete Composite: 23–35. Mehta, P.K. & Monteiro, P.J.M. 1993. Concrete structure, Properties and Materials, Prentice hall, New Jersey. Saad, M.N.A., de Andrade, W.P. & Paulon, V.A. 1982. Properties of mass concrete containing active pozzolana made from clay, Conc. Int: 59–65. Tagnit-Hamou, A., Pertove, N. & Luke, K. 2003. Properties of concrete containing diatomaceous earth, ACI Mat. J 100(1): 73–8.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Ultra light-weight self consolidating concrete M. Hubertova Diploma Engineering Lias Vintirov, LSM (Light-weight Building Material) k.s., Czech Republic

R. Hela Diploma Engineering Brno University of Technology, Institute of Technology of Building Materials and Components, Brno, Czech Republic

ABSTRACT: Technology of compact self consolidating concrete is widely used in common production nowadays. However, properties of highly fluidizated concrete are becoming more utilized for light-weight self consolidating concrete, with volume weight 1500–1850 kg/m3 , and the most recent ultra light-weight concrete with volume weight as low as 800–1400 kg/m3 . The paper describes methods of designing ultra light-weight self consolidating concrete with light-weight expanded clay aggregate. It also observes upon problems of maintaining rheological properties for longer periods (90 minutes min.) with light-weight aggregate, which has very high water-absorbing capacity. This type of concrete is interesting not only for its capacity to lighten building structures but also for its good compressive strength or excellent heat insulating properties of ultra light-weight concrete. The paper also states practical experience gained at different constructions.

1

50 N/mm2 are widely known now. Some examples from the past two years:

INTRODUCTION

Application of lightweight self consolidating concrete (LWSCC) with expanded clay aggregate (trade name Liapor) are used more frequently in Europe in recent time. Using porous aggregate for high strength concretes might be surprising considering importance of strength of aggregate for strength of high-strength concrete. Lightweight aggregate (LWA) is porous and not very strong. Nevertheless, drop of volume weight of concrete with strength of 40–50 N/mm2 below 1800 kg/m3 can represent certain cost saving due to reduction of total construction weight. Thanks to favorable physical properties, low volume weight and relatively high strength combined with good workability, low noise emission and reduction of consumed work in the course of placing, there is a wide range of application for LWSCC, in particular in the field of precast elements and reconstruction of old buildings, where extra load would be undesirable. LWSCC was for the first time in the Czech Republic applied in 2005. This enabled a comparison of properties of fresh and hardened concrete mixed in laboratory and the same formula mixed in-situ in the mixing plant. The comparison unambiguously proved that it is possible to produce and place LWSCC in practice, in spite of certain differences and high sensitivity of proportioning, batching and mixing (Hubertova & Hela 2007). Applications of LWSCC with volume weight 1500 to 1800 kg/m3 and compressive strength up to

– reconstruction of an old building from 15th century—lightening of three-floor building by using LWSC for floor slabs—concrete placed by pumping without negative vibrations – precast elements—production of bleachers benches for sports stadium Eden Prague and multi-functional hall in Carlsbad (direct-finish concrete) (Hubertova 2007) – precast elements—production of balcony banister for renovation of prefabricated buildings (minimal load necessary) – three-dimensional prefabricated parts—manufacture of car-shelters and bathroom cells for flats and hotels – Architects have supported the use of light-weight concrete for face or direct finish purposes in recent time. Several buildings were constructed form direct finish thermo-insulating concrete in Switzerland and Germany. The advantage of direct finish concrete is its high architectonic value. Monolithic single-ply bearing constructions cast from concrete have especially high durability, since there are no plaster coats and covering (which brings more cost saving of work and material). The conception of light-weight concrete also saves costs of thermoinsulation (no need for sandwich-type structures). One of the most interesting and extraordinary applications was the use of thermo-insulating LWSCC with

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dry volume weight below 1000 kg/m3 for construction of a family house. A Swiss architect Patrick Gartmann designed and built a family house near the town of Chur. The main idea was to cast a house from concrete, with free formation, massive and homogeneous. Material used for the structure was light-weight concrete with volume weight 1100 kg/m3 and compressive strength 11 MPa and thermal conductivity coefficient λ = 0,3 W/mK (other information Liapornews 2005, 2007, 2008). This very successful project was the initial idea of development of mix design of concrete of similar or higher performance from raw materials available on the Czech market. The main requirement was to design a structural material for perimeter structures in the form of direct finish thermo-insulating light-weight concrete. Thickness of load bearing perimeter walls should not exceed 450 mm. Concrete for all structures was intended to be direct finish. Delivery of concrete to the site was required to be in the form of ready mix. 2

Composition of mix design.

m [kg] on 1 m3 LWA 4–8 mm LWA 1–4 mm Cement I 42,5 R Admixture (slag) Superplasticizer Foaming agent Water Volume of air pores

0,88 m3 0,37 m3 380 kg 120 kg 3l 2l 140 l 20% vol.

MIX DESIGN, FRESH CONCRETE, APPLICATION

Material used for the house of architect Gartmann in Switzerland was lightweight concrete with heat conductivity λ = 0,3 W/mK. Because standards in the Czech Republic are more rigorous, we had to design concrete with heat conductivity coefficient λ smaller than 0.25 W/mK. We had to solve a problem of designing very lightweight concrete with strength qualities of structural concrete, self consolidating, direct finish and with good thermo insulating properties. Basic requirements are summed up in following items: – – – – –

Table 1.

Figure 1. Measuring of rheological properties at the building site (Slump Flow test).

thermo insulating properties structural strength at least 6 N/mm2 visual concrete without needs of surface finishing self consolidating properties supply by ready mix

We designed and tested a mix design both in laboratory conditions and in-situ with given type of formwork (PERI) and release agent in climatic conditions expected in the time of building family houses (summer weather, 25–30◦ C). Composition of mix design is stated in Table 1. We used pre-wetted LWA for manufacture of fresh concrete for the reason of minimizing fluctuation of rheological properties, which can be caused by uncontrolled water absorbing capacity of LWA during mixing and placing and to keep the cement matrix constantly frothed. Slump flow value of fresh concrete after mixing was 65 cm (see Figure 1). Considering low volume weight and in spite of prewetting LWA we assumed, that the concrete will not be pumpable (high water absorbing capacity of LWA under high pressure during pumping, difficulties in

Figure 2. Pouring fresh concrete from truck mixer into the skip and placing into formwork.

keeping constant structure of concrete), therefore concrete was transported from truck mixer into formwork in the skip (see Figure 2).

3

CONCRETE

At the design stage we mixed several mix designs with dry volume weight from 850 kg/m3 to 1100 kg/m3 .

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Table 2. Measured values of heat conductivity coefficient. λ[W · m−1 · K−1 ]

Figure 3. weight.

1 2 3 4 5 6 7 8 9 10 average

Dependency of compressive strength on volume

0,2655 0,2243 0,2265 0,2235 0,2355 0,2458 0,2541 0,2344 0,2587 0,2611 0,2429

16 14

Days (-)

12 10 8 6 4 2 0 0

20

40

60

80

100

120

140

160

180

200

Compressive strength (N.mm-1)

Figure 4. Development of tensile bending strength in time up to 180 days from mixing.

Figure 6. Possible defects caused by uncontrolled foaming of fresh concrete.

14 12 days

10 8 6 4 2 0 0

50

100

150

200

dynamic and static elasticity modulus (GPa) watter environment laboratory environment static elasticity modulus - watter environment

Figure 5. Development of dynamic and static elasticity modules in time.

Relation between these volume weights and compressive strength is stated in Figure 4. Typical compressive strength of light weight concrete (measured in the course of tests) is 8 MPa with dry volume weight 970 kg/m3 . Figure 4 shows development of compressive strength in time. Figure 5 shows development of dynamic elasticity modulus (determined with non-destructive pulse ultra sonic method) of test samples stored in water environment and samples stored in laboratory conditions (20◦ C, low relative humidity). Values are compared to values of static elasticity modulus. Table 2 states measured

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values of heat conductivity coefficient measured at test specimens sampled in the course of casting concrete into formwork. Static elasticity modulus of this concrete is 5.8 GPa. In the course of experimental casting we encountered a problem of keeping constant volume of frothing of cement matrix. As we mentioned above, we frothed the mix at 20% by volume. Figure 6 shows possible defects of over frothing of cement matrix. In this case the limit was exceeded by 11% by volume. Fresh concrete was very unstable and homogeneity in formwork was disrupted. The layer of mastic cement did not form at the boundary with the formwork, either. For these reasons we closely ob-served correct dosage of not only frothing admixture but also mixing water and pre-wetted LWA during placing. We measured values of volume weight of fresh concrete very carefully not only after mixing but also after transport and during casting.

4

PROPERTIES OF CONCRETE STORED IN CORROSIVE ENVIRONMENT

This type of concrete was studied also with respect to durability in corrosive environment. Samples were

Table 3.

Test results. Volume weight

Environment

28 days [kg/m3 ]

180 days [kg/m3 ]

Rc,cu 180 days [N/mm2 ]

CO2 −1 CO2 −2 CO2 −3 SO2− 4 −1 SO2− 4 −2 SO2− 4 −3 Mg2+ − 1 Mg2+ − 2 Mg2+ − 3

1060 1055 1082 1064 1060 1103 1058 1064 1093

976 978 937 1092 1062 1068 1059 1112 1072

11,4 12,7 12,5 9,7 11,6 13,1 13,1 10,9 13,7

Average

Environment

Compressive Volume strength weight [%] [%]

−102

12,2

−12,95 −9,57

−2

11,5

−18,18 −0,16

9

12,6

−10,36

0,86

Note: Variation of strength is considered with respect to referential test samples stored in water for 180 days, variation of volume weight is considered with respect to volume weight at the age of 28 and 180 days.

Table 4.

View of finished wall.

Variation

DiffVolume erence weight volume 180 days weight 180 days [kg/m3 ] [kg/m3 ] [N/mm2 ]

CO2 − 1 CO2 − 2 1082 CO2 − 3 SO2− 4 −1 SO2− 4 − 2 1074 SO2− 4 −3 Mg2+ − 1 Mg2+ − 2 1081 Mg2+ − 3

Figure 7.

Results of X-ray diffraction analysis.

Environment Minerals contained Portlandit, ortoklas,ettringit, β køemen, kalcit, β cristobalit, biotit, vaterit CO2 2) Portlandite, orthoclase, etringit, β silica, limestone, β cristobalite, biotit, vaterit SO2− 4 1) Portlandite, orthoclase, etringit, ß silica, limestone, ß cristobalite, biotit SO2− 4 2) Portlandite, orthoclase, etringit, ß silica, limestone, ß cristobalite, biotit, gypstone Mg2+ 1) Portlandite, orthoclase, etringit, ß silica, limestone, ß cristobalite, biotit Mg2+ 2) Portlandite, orthoclase, etringit, ß silica, limestone, ß cristobalite, biotit, chrysotile

pH

CO2 1)

stored in selected types of corrosive environment for the period 180 days, in particular solution of sulphates of concentration 34600 mg/l, solution of magnesia of concentration 10000 mg/l and water with corrosive carbon dioxide of concentration 50000 mg/l. Compressive strength and value of pH were determined and X-ray diffraction analysis made after 180 days. Tables 3 and 4 show the results. Changes of volume weights after 180 days of storage samples in water with corrosive carbon dioxide were about −9.57%, changes in solutions of sulphates and magnesia did not exceed 1%. Loss of 9.75% weight in water environment with CO2 is explained by change of portlandite Ca(OH)2 to calcium carbonate CaCO3 and then to calcium hydrogen carbonate, which is easily soluble in water and washes out of concrete. This phenomenon is apparent form X-ray analysis. Changes of compressive strength were −10.36% in magnesia solutions, 12.95% in water with corrosive carbon dioxide and −18.18% in sulphates. This concrete is not resistant to environment class XF4 or XA3 (according to European standard EN 206-1). It can be used for structures protected from penetration of corrosive chemicals.

5

8,93 11,07 10,93

11,06 8,89

11,42

Note: 1) sample from the surface, 2) sample from the depth of 20 mm.

CONCLUSIONS

Experimental placing proved positive utilization of super light weight insulating concrete. Frequency of test of both fresh and hardened concrete was higher during all the time of placing. On the basis of these tests we can confirm that it is possible to manufacture super-light self consolidating concrete of defined properties with no major variations, however, only on condition of increased attention in all production steps. The developed mix-design is now applied for construction of family houses. The advantage of this concrete is the possibility of use of secondary raw materials as active additives, namely fly ash or micronized slag and ultra fine

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additives like microsilica, metakaolin and micronized lime stone. Admixtures used are polycarboxyl based superplasticizers and foaming agents. The way of mixing of this concrete is special: lightweight aggregate is recommended to be pre-wetted either in the mixing device by batching LWA and water (and adding the rest of materials after a while) or before the process of mixing (in the store or similar). Pre-wetted LWA gives higher stability of rheological behavior of fresh concrete and foaming of cement matrix is easier to control. Ultra light-weight self consolidating concrete can be placed with skip and crane; however, it can-not be pumped. The period of workability is between 60 and 90 minutes depending on environmental conditions (namely weather). Requirements of application of this material are the same as with common concrete including the necessity of curing. ACKNOWLEDGEMENT The part of this outcome has been achieved with the financial support of project GA 103/07/076 ‘‘Development of lightweight concrete for wide constructional application and study of it’s durability’’ and with the financial support of the Ministry of Industry and Trade of the Czech Republic, project No. MPO FI-IM5/016, ‘‘Development of light-weight high performance concrete for monolithic constructions and for precast elements’’.

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REFERENCES Hubertova, M. 2007 Lightweight self compacting concrete used in SK Slavia Prague stadium. BFT Betonwerk + Fertigteil-Technik; Concrete plant + precast Technology. 2007. 2007(12). pp. 4–13. ISSN 0373–4331. Hubertova, M.; Hela, R. 2007 Development and experimental study on the properties of lightweight self compacting concrete. In 5th RILEM Symposium on Self-Compacting Concrete. 1. Ghent, Belgium. 2007. Hubertova, M.; Hela, R. 2007 The Effect of Metakaolin and Silica fume on the Properties of Lightweight Self-Consolidating Concrete. In Ninth CANMET/ACI International Conference on Recent Advances in Concrete Technology. 1. Warsaw, American concrete institute. 2007. pp. 35–48. ISBN 0-87031-235-9. Hela, R., Hubertova, M., Kleˇcka,T. 2006 Lightweight Self Compacting Concrete (LWSCC) in the Ready Mix, The 6th International Symposium on Cement and Concrete, CANMET/ACI International Symposium on Concrete Technology for Suistanable Development, XI’ANChina 2006, str.1381–1388, ISBN 7-119-02249-0. EFNARC, ‘Specification and Guidelines for SelfCompacting Concrete’, Surrey United Kindom 2002, ISBN 0-9539733-4-4, www.efnarc.org. Liapornews 2/2005 Liapor GmbH Pautzfeld; www.liapor.com. Liapornews 1/2008 Liapor GmbH Pautzfeld; www.liapor.com. Liapornews 3/2007 Lias Vintíøov, Lehký stavební materiál k.s.; www.liapor.cz.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Wood use in Type I and II (noncombustible) construction D.G. Bueche Hoover Treated Wood Products, Thomson, Georgia, USA

ABSTRACT: As with many products, the building code regulates the use of wood in construction. Two broad categories separate materials: combustible and noncombustible. Codes limit the application of combustible materials on the basis of fire and life safety. The question then is, ‘‘Are there options available in the 2009 International Building Code for using wood in structural applications in lieu of noncombustible materials?’’ Fire Retardant Treated Wood (FRTW) provides that option. The 2009 IBC recognize FRTW for many applications where a noncombustible material is mandated. A few applications allow FRTW in lieu of 1-hour ratings. 1

INTERNATIONAL BUILDING CODE

In 1994 the three existing national building code organizations created the Internal Code Council (ICC), a nonprofit organization dedicated to developing a single set of comprehensive and coordinated national model construction codes. The goal of the ICC was to have a family of national codes available by the year 2000, and ICC met that goal. In 1996 work began on the creation of the International Building Code (IBC). It was developed primarily from the provisions of the three nationally recognized model building codes: the National Building Code (NBC), Standard Building Code (SBC), and Uniform Building Code (UBC). The IBC is a comprehensive code and is the coordinating document for the suite of International codes. In certain instances, the IBC provisions are identical to those of the three model codes. In other instances, the provisions are a modification of requirements from one or more of the three regional codes. Some provisions are entirely new and unique to the IBC.

2

FIRE SAFETY

Fire safety is the reduction of the potential for harm to life as a result of fire in buildings. Although the potential for being killed or injured in a fire cannot be completely eliminated, fire safety in a building can be achieved through proven building design features intended to minimize the risk of harm to people from fire to the greatest extent possible. Designing a building to ensure minimal risk or to meet a prescribed level of safety from fire is more complex than just the simple consideration of what building materials will be used in construction of the building. Many factors must be considered including the use of the building, the number of occupants, how

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easily they can exit the building in case of a fire and how a fire can be contained. The IBC only regulates those elements which are part of the building construction. The building contents found in any building are typically not regulated by the IBC but in some cases are regulated by the fire codes. The classification of buildings or parts of buildings according to their intended use accounts for: – the quantity and type of combustible materials likely to be present (potential fire load); – the number of persons likely to be exposed to the threat of fire; – the area of the building; – the height of the building. This classification is the starting point in determining which fire safety requirements apply to a particular building. Classification dictates: – the type of building construction; – the level of fire protection; – the degree of structural protection against fire spread between parts of a building that are used for different purposes. Even materials that do not sustain fire do not guarantee the safety of a structure. Steel, for instance, quickly loses its strength when heated and its yield point decreases significantly as it absorbs heat, endangering the stability of the structure. An unprotected, conventional steel joist system will fail in less than 10 minutes under standard laboratory fire exposure test methods, while a conventional wood joist floor system can last up to 15 minutes. Even reinforced concrete is not immune to fire. Though concrete structures have rarely collapsed, concrete will spall under elevated temperatures, exposing the steel reinforcement and weakening structural members.

It is generally recognized then, that there is really no such thing as a fireproof building. Fires can occur in any type of structure. The severity of a fire, however, is contingent on the ability of a construction to: – confine the fire; – limit its effects on the supporting structure; – control the spread of smoke and gases. To varying degrees, any type of construction can be designed as a system, that is, a combination of construction assemblies, to limit the effects of fire. This allows occupants sufficient time to escape the building and for firefighters to safely reach the seat of the fire. Occupant safety also depends on other parameters such as detection and exit paths, and the use of automatic fire suppression systems such as sprinklers. These concepts form the basis of the IBC. 3

FIRE RETARDANT TREATED WOOD

Wood loses its strength in a different way than metals. In the early stages of a fire, wood’s strength is increased because of a reduction in moisture content. Wood is a good insulator and does not transfer the heat on its surfaces to its core very quickly. While it may be burning or charring on its surface, its interior will be relatively cool for a long time. All this increases the length of time wood and fire retardant treated wood will retain its integrity—time to get the people out of the building, time to get the firemen to the building, and time to extinguish the fire. Fire retardant treated wood has the added advantage of maintaining structural integrity even longer because it chars at a slower rate than untreated wood is consumed. In addition, fire retardant treated wood will not spread the fire from one portion of a building to another, and it will extinguish itself once the ignition source is removed. Wood is principally composed of cellulose, hemicellulose, and lignin, all of which change their physical and chemical characteristics by oxidation and chemical decomposition. This phenomenon is called combustion or burning. It destroys the structure of the wood so it will not support a load by reducing it to a small amount (1% or less) of a mineral substance called ash. The kindling temperature of wood (500◦ F) is the temperature above which the wood will ignite spontaneously. At this point, as a result of chemical decomposition, the wood contributes a heat of its own, and the temperature rises even higher. Combustible gases and smoke are given off, and the wood begins to char. Because wood in itself is a good insulator, the high temperature at the surface is not readily transmitted to its interior. This insulating quality and the moisture always present in the wood results in a slow destruction, especially of large timbers. How quickly

the material burns, depends on its size, shape, the air circulation, and the control of radiation. Therefore, if the conduction of heat in the wood can be controlled to prevent the temperature of the wood from exceeding its kindling temperature, its rate of destruction can be greatly reduced. This is the key to Fire Retardant Pressure Treated Wood. Research has shown that certain ingredients, when added to the wood, are able to insulate its surfaces so that its temperature remains below the kindling temperature for an extended period of time no matter how hot the heat source might become. Among the ingredients used for this purpose are the acid salts of sulfates and phosphates, borates, and boric acid. All fire retardant treatments are water-soluble so water is used as the vehicle for carrying the treatments into the wood. The only effective method of application is by the Pressure Treatment Process. After pressure impregnation, most of the moisture is removed until the treated wood has a moisture content of no more than 19% for lumber and 15% for plywood. Fire retardant treatments do not necessarily prevent wood from being destroyed by fire, but they are the necessary ingredient that, when added to wood, slow down the decomposition to such an extent that the wood structurally out performs most other building materials during actual fire conditions. When temperatures reach a point slightly below the kindling point, the chemicals react with each other. Nonflammable gases and water vapor are formed and released at a slow persistent rate which envelope the wood fibers insulating them from temperatures that cause the wood to decompose. The inflammable gases and tars are reduced and an insulating char forms on the surface of the wood, further slowing down the process of decomposition. Because of the greatly reduced rate of decomposition or burning, the structural integrity of the wood is preserved for a long period of time, smoke and toxic fumes are greatly reduced, and when the heat source is removed, the wood ceases to decompose and the spread of fire by the wood is eliminated. In Section 2303.2, the IBC defines fire-retardanttreated wood as ‘‘any wood product which, when impregnated with chemicals by a pressure process or other means during manufacture, shall have, when tested in accordance with ASTM E 84, a listed flame spread index of 25 or less and show no evidence of significant progressive combustion when the test is continued for an additional 20-minute period. In addition, the flame front shall not progress more than 10.5 feet (3200 mm) beyond the centerline of the burners at any time during the test.’’ This definition is a performance specification. Unlike the specifications for wood preservatives, fire retardant treated wood is specified on the basis of performance and not retention. The ‘‘Flame Spread’’ index is a measure of the surface burning characteristics of

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a building material when compared to the relative surface burning characteristics of cement board (rated at 0) and untreated select red oak flooring (rated at 100). The index is determined by relative performance in a 25-foot long fire test tunnel furnace under controlled conditions of draft and temperature. In the ASTM E-84 tunnel test, a gas jet is located near one end of the tunnel. Without a test specimen present in the tunnel, the ignition flame from the gas jet extends down the tunnel for a distance of 4½ feet from the burners. After a test specimen is placed in the tunnel and exposed to the ignition flame for a period of 10 minutes, the spread of the flame is measured from the fire end of the tunnel. Keep in mind that the standard flame spread test is only for 10 minutes. The opinion of researchers was that this method only demonstrated delayed ignition and gave little indication of non-combustibility. To remedy this, the test period for fire retardant treated wood was extended another 20 minutes to 30 minutes duration. If the specimen shows no evidence of additional spread of flame beyond the limit attained in 10 minutes, then the specimen could be said to show no significant progressive combustion. Fire retardant treated wood must meet this additional test standard to be used for structural applications. 4

TYPES OF CONSTRUCTION

Similar to the previous codes, construction type provisions are set out in Chapter 6 of the IBC. The type of construction is determined by two factors: 1) whether the materials used in the structural frame are combustible or noncombustible, and 2) the fire resistance of building elements. Noncombustible materials can have very little fire resistance (for instance, steel framing when subjected to a hot fire can fail quickly), and so combustibility and fire resistance are separate determining factors in type of construction. In the IBC there are five types of construction, and they are summarized in Table 1. Types III through V are primarily wood frame construction; Type III is wood frame with noncombustible or fire-retardant treated wood exterior walls, Type IV is heavy timber, and Type V is generally thought of as wood frame. The ‘‘A’’ designation in these construction types means the building elements for the most part are required to be of one-hour rated construction. The ‘‘B’’ designation means that no fire resistance rating is required (referred to as ‘‘unprotected’’ wood frame construction). Even in the noncombustible construction types (Types I and II), many nonstructural elements of the building, such as floor coverings, windows and doors, and interior finishes, can be wood. Permitted combustible building elements in noncombustible buildings are conveniently listed in Section 603 of the

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Table 1.

Types of construction in the IBC.

Type

IBC description

Noncombustible

Type I (A & B) Type II (A) Type II (B)

Mixed noncombustible and combustible Type III (A) including frame and heavy timber (HT) Type III (B) Type IV (HT) Combustible—traditional wood frame Type V (A) Type V (B)

Table 2.

Allowable uses of FRTW in the IBC.

Type

Building assembly

I and II

Nonbearing partitions where the required fire-resistance rating is 2 hours or less (Section 603.1.25.1) Nonbearing exterior walls where no fire rating is required (Section 603.1.25.2) Roof construction including girders, trusses, framing and decking (Section 603.1.25.3)

III and IV

Permitted within exterior wall assemblies of a 2-hour rating or less (Sections 602.3 and 602.4)

V

Use of FRTW is unrestricted

IBC. This list includes structural elements that are constructed of fire retardant treated wood (FRTW). FRTW is not considered noncombustible, but can often be used in place of noncombustible materials. For instance, FRTW can be used in place of noncombustible materials in exterior walls of Type III and IV buildings, and in roof structures of low-rise buildings of Types I and II construction. Table 2 summarizes where fire-retardant treated wood is permitted to be used in lieu of noncombustible materials. In Section 602.3, the IBC defines Type III Construction as being ‘‘that type in which the exterior walls are of noncombustible materials and the interior building elements are of any material permitted by the code.’’ The section goes on to say that fire-retardant treated wood is permitted in exterior wall assemblies in lieu of noncombustible materials when the rating of the wall is required to be 2-hours or less. Therefore, for many of the most common occupancies, buildings constructed entirely of wood can be just as large and as high as noncombustible buildings. Table 3 shows that buildings of IBC Type IIIB in many occupancies, may be just as large as buildings of Type IIB (noncombustible unprotected). Buildings of Type IIIB, for the occupancies shown, may be entirely of wood if FRTW is used in the exterior walls.

Table 3. Comparison of IBC Type IIB and IIIB construction. IBC Table 503 allowable area (sq ft)

IBC Table 503 allowable height (stories/feet)

IBC Occupancy

Type IIB

Type IIIB

Type IIB

Type IIIB

A-3 B E M R-1 R-2 S-2

9,500 23,000 14,500 12,500 16,000 16,000 26,000

9,500 19,000 14,500 12,500 16,000 16,000 26,000

2/55 3/55 2/55 2/55 4/55 4/55 3/55

2/55 3/55 2/55 2/55 4/55 4/55 3/55

The IBC permits sprinklered buildings with National Fire Protection Association NFPA 13 systems to contain one additional story and be increased in height 20 feet. Residential buildings with NFPA 13R systems may be increased one story and 20 feet in height and are not subject to the total building area limit of a three story building (13R is only appropriate up to four stories above grade plane). In the IBC, a rated wall in accordance with Table 602 can be used to separate a building into two smaller areas, neither of which exceeds threshold values that require installation of sprinklers. This is not considered a fire wall separating buildings. It is a fire separation assembly, separating the building into fire areas. 5

FIRE RESISTANCE

Fire retardant treated wood has a surface burning classification and, by itself, does not have a resistance rating in hours any greater than untreated wood. Fire ratings in hours are assigned to wall, floor, and roof deck assemblies, following testing in accordance with ASTM E 119. References such as the Underwriters Laboratories ‘‘Fire Resistance Directory’’ specifically point out that FRT wood may be substituted for untreated wood in any related assembly. FRTW can be used as a component of such assemblies in structures where the code does not permit the use of untreated wood. Descriptions of fire resistance rated assemblies incorporating structural lumber are listed in IBC Table 720.1(2) as well several publications referenced by the IBC including:

wall or partition assembly (WP 3605) that has wood  studs covered by 5/8 Type X gypsum board with specified nailing and positioning of the panels. This assembly could be used for interior, non-bearing partitions, requiring a one hour rating in a noncombustible structure if the studs were FRTW. In a similar manner, by substituting FRTW for untreated wood, other one and two hour wall and ceiling assemblies can be used in noncombustible type buildings. The IBC also permit use of ceiling assemblies with the top membrane omitted where only unused attic space is above. The IBC permits asymmetric testing for fire resistance rating (testing from the inside only) where the distance to the property line is at least 5 feet. If sprinklering is not used for H&A increases, it is permitted to reduce fire resistive requirements by one hour for all construction elements except exterior walls.

6

CASE STUDIES

Figure 1 shows a 1.2 million square foot warehouse that was developed for multiple tenants and features a hybrid panelized roof system utilizing fire retardant treated wood. The hybrid roof system consists of 4 ft × 8 ft fire retardant treated plywood on 2x and 3x fire retardant treated sawn lumber subpurlins. The primary framing consists of steel bar joist spaced 8 ft on center with steel girder trusses as the main structural members. This system uses panelized units assembled on the ground and then lifted into position at the roof level, where the steel bar joists are welded or bolted to the primary steel girder trusses. The free edge of the wood decking for each panelized unit is nailed to the framing edge of the previously placed unit. Pre-framed panel ends attached to the main steel trusses complete the assembly. In a fire retardant treated wood roof system, the panelized wood sections speed the erection process

– Fire Resistance Directory, published by Underwriters Laboratories; – Fire Resistance Design Manual, published by the Gypsum Association. As way of example, the Gypsum Association’s ‘‘Fire Resistance Design Manual’’ contains a one hour

Figure 1.

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FRTW panelized roof system.

Figure 2.

Apartment complex utilizing FRTW studs.

Figure 3.

The Orchard shopping mall.

and add strength, dimensional stability, and high diaphragm capacity to the roof. The ability to preframe large roof panelized units reduce cost, cuts construction time, and enhances job site safety since fewer man-hours are spent on the roof. Panelized roof systems are one of the safest systems to erect because most of the work is accomplished on the ground during the fabrication of the large pre-framed roof panels. Once the large panels are lifted into position at the proper roof elevation, only one or two workers are required on the roof to complete the final purlin attachments and diaphragm nailing.

585

Another advantage to the hybrid panelized roof system is the speed at which it can be constructed. In this case, the entire 1,200,000 square-foot roof was erected in only 5 weeks with minimal overtime required. An experienced 4 man crew can erect 25,000 square feet of roof per day. The apartment complex in Figure 2 contains 500,000 square feet of residential space, 40,000 square feet of retail space, and a 350,000 square foot parking garage. The exterior bearing walls are constructed with fire retardant treated wood studs. The interior framing is untreated wood. The two story parking garage consists of a onestory enclosed parking garage and a one-story open garage. The parking garages are not considered in determining the maximum number of stories allowed in the building under the IBC when constructed of Type I construction and a three hour occupancy separation is maintained between the parking garage and the residential occupancy. The IBC allows the base area to be doubled and a one-story height increase when a NFPA 13 sprinkler system is installed. Figure 3 is The Orchard at Westminster, designed to be reminiscent of the Main Streets of small towns in the first half of the 20th century. It is an openair, entertainment and lifestyle center that comprises a million square feet of exclusive, outdoor, fashionoriented retail including big boxes, department stores and smaller, upscale retailers. The development also includes 500 housing units and office space. The name was selected to reflect the rich agricultural heritage of the area, which was home to some of the largest apple orchards in Colorado. It was designed under the IBC to be a Type II noncombustible structure of concrete and steel. The IBC allows fire retardant treated to be used in the roofs and nonbearing walls of noncombustible types of construction. Fire retardant treated plywood was used as roof sheathing over light gauge steel roof trusses providing an easy to nail surface for the 3-dimensional architectural composite shingles. REFERENCE 2009 International Building Code. International Code Council, Inc. Country Club Hills, IL.

Composite materials

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Computational models for textile reinforced concrete structures W. Graf, M. Kaliske, A. Hoffmann, J.-U. Sickert & F. Steinigen Institute for Structural Analysis, TU Dresden, Germany

ABSTRACT: This paper is a contribution towards computational modeling of textile reinforced concrete (TRC) structures. Beside new TRC structures the focus is set on using TRC for strengthening of existing RC structures. Structural behavior is realistically analyzed by means of an extended layer model with specific kinematics the so-called multi-reference-plane model (MRM), considering the complete load and modification process of the structure as well as uncertain time dependent parameters. Uncertain load-displacement dependencies are computed for a RC barrel vault strengthened by TRC layers and for a TRC pedestrian bridge.

1

TEXTILE-REINFORCED CONCRETE

Textile-reinforced concrete (TRC) is a new composite material consisting of textile fabrics in a fine-grained concrete matrix, see Figure 1. The textile fabrics are made of filament yarns (so-called rovings) which are connected with the aid of stitching yarn. The filament yarn is a bundle comprised of a large number (400 . . . 2.000) of single filaments. Filaments may be made of different materials, e.g. alkali-resistant glass (ARG) or carbon fibers (CF). In the paper, attention is focused on TRC comprised of carbon filaments. Textile reinforcement is not only applied in new structures but may also be used for the additional strengthening of existing structures. The use of textiles with rovings aligned with the principal stress direction of the composite is more effective than using the same amount of fiber material in the form of randomly distributed short fibers, as is the case in fiber-reinforced concrete. The design and production of textile-reinforced structures require a reliable prediction of load-bearing behavior with the aid of suitable numerical models. The strengthening of RC structures by means of TRC results in a multilayer composite. The loadbearing capacity of the TRC and its material properties as well as the bond behavior between the old construction and the strengthening layer must therefore be modeled as realistically as possible. The paper focuses on computer models and the application of these models to various examples investigated in a research project undertaken within the framework of the German Collaborative Research Center (SFB) 528. The investigations demonstrate that structural responses are highly dependent on the spatial and temporal variations of the uncertainty of material and geometric data. For this reason, the present paper concentrates on the generalized uncertainty modeling of material and geometric parameters.

589

The application of TRC for strengthening purposes implies a new strengthening technology for damaged steel-reinforced concrete structures, which are normally referred to as old constructions. Finegrained concrete layers reinforced with textile fabrics are thereby applied to the exposed surfaces of old constructions.

2

REMARKS ON UNCERTAINTY

The realistic analysis of textile reinforced concrete structures requires reliable (input) data beside the suitably matched computational model summarized in Section 3. As a rule, the data and the model possess uncertainty. According to this, the geometrical, material and loading data required for structural analysis of RC structures with textile strengthening are also more or less characterized by data uncertainty, see Möller et al. 2006. It is necessary to take into account this uncertainty appropriately. The following mathematical models are capable to describe uncertainty of the strengthened structure: – Randomness assessed by probability density function (Fig. 2a) – Fuzziness assessed by membership function (Fig. 2b) – Fuzzy randomness assessed by fuzzy probability density function (Fig. 2c), whereas fuzziness and randomness are considered as special cases of the general model fuzzy randomness (Möller & Beer 2004). The choice of the model depends on the available data. Textile reinforced concrete structures show data uncertainty of different characteristic. If an event (regarding its occurrence), as a random result of a test, may be observed as a crisp value on an almost

Figure 1. a)

Figure 2.

Textile reinforced concrete (TRC). b)

c)

Mathematical models of uncertainty.

unlimited number of occasions under constant boundary conditions, this represents a stochastic uncertainty. The uncertainty characteristic randomness is assigned to this stochastic uncertainty. If the boundary conditions are (apparently) subject to arbitrary fluctuations, a comprehensive system overview is lacking, the number of observations are only available to a limited extend, or the sample elements are of doubtful accuracy (non-precise), an information deficit exists. The outcome of this is a gap between the mathematical quality requirements of data if using stochastic methods and the real available non-precise data. The data do not fully satisfy real-valued probability laws. In fact, the data may be quantified by imprecise probability, see e.g. Viertl 1996. The impreciseness results from epistemic uncertainty in terms of non-precise recognition of data or statistical inference (determining stochastic input parameters, such as expected values, variances and probability distribution functions). Here, this epistemic uncertainty is described by the uncertainty characteristic fuzziness and is mathematically quantified on the basis of the fuzzy set theory. The uncertainty consisting of randomness and fuzziness is summarized in the characteristic fuzzy randomness. The fuzzy random data are assessed with the aid of the uncertain measure fuzzy probability. The fuzzy probability is defined as fuzzy set of real-valued probabilities. If data only show random properties, fuzziness is quantified by zero, i.e. a real-valued random variable results. Non-precise data without random properties are quantified by fuzzy values. Data uncertain (e.g. for material parameters, loading or boundary conditions) may be characterized by fuzzy random fluctuations, which depend on external

conditions. External conditions include, for example, time τ , the spatial coordinates θ = {θ1 , θ2 , θ3 }, air pressure or temperature, which are lumped together in the parameter vector t = {τ , θ, . . .}. The varying uncertainty of parameters (dependent of arbitrary arguments t) is quantified using fuzzy random functions. In the case of exclusive time-dependency, a fuzzy random process is created. Fuzzy random processes are included as special case in the following definition of fuzzy random functions. Based on the theories of fuzzy stochastic (Möller & Beer 2004), fuzzy sets (Zimmermann 1992), and stochastic (Schenk & Schuëller 2005), fuzzy random functions X˜ (t) are defined as a set of discrete fuzzy random variables    (1) X˜ (t) = X˜ t = X˜ (t)∀t t ∈ T For fuzzy random functions, fuzzy probability distribution functions F˜ t (x) and fuzzy probability density functions f˜t (x) may be generated. The functions F˜ t (x) and f˜t (x) are fuzzy functions (Möller & Beer 2004). A fuzzy function x˜ (t) is the result of the uncertain mapping x˜ : T→F(X ˜ )

(2)

of the parameter space T onto the set of all fuzzy values F(X ) in X .For a parametric repre fuzzy  functions,  sentation x˜ t = x s˜, t with the aid of fuzzy bunch parameters s˜ can be applied advantageously (Sickert 2005). The quantification of fuzziness by fuzzy bunch parameters s˜ leads to the description of the fuzzy functions Ft (˜s, x) and ft (˜s, x) as assessed bunches of real-valued functions Ft (s, x) and ft (s, x) which are specified for all s ∈ s˜ . Uncertain load and modification processes of structures are described as fuzzy random processes. These fuzzy random processes represent the input of the structural analysis which is contained within the numerical approach introduced in Section 3. As result the structural stress state and damage state are computed also as fuzzy random process. Because of that mapping of fuzzy random input processes onto ˜ fuzzy random result processes Z(t),the approach is referred to as fuzzy stochastic analysis. The computational model and the numerical solution strategy are described in Section 3. 3

COMPUTATIONAL MODEL

The numerical solution is based on a time-discretization of the uncertain processes  into  a set of fuzzy random variables X˜ t X˜ t = X˜ t . Thereby, fuzziness is quantified   by means of fuzzy bunch parameters X˜ t = X t s˜ .

590

Figure 3.

Multi-reference plane model with three reference planes.

In order   to determine fuzzy random result processes Z˜ t , a three loop analysis algorithm was developed. In the outer loop, a fuzzy structural analysis is performed in order to map the fuzziness of  the bunch parameters s˜ onto the fuzziness of Z˜ t . The α-level optimization using a modified evolution strategy (Möller et al. 2000) is applied for this task. Thereby, the so-called stochastic fundamental solution has to be determined repeatedly. In principle, any type of stochastic algorithm may be applied for this task. Owing to nonlinear structural behavior of the structures under consideration here that may only be stated in a nonclosed form as a point set, only simulation methods like Monte-Carlo simulation are suitable for the stochastic fundamental analysis. All these methods can be combined with response surface simulation, which is mandatory for solving problems of practical relevance. Within the stochastic fundamental analysis simulation points are determined by a FE model. The combination of that FE model with stochastic and fuzzy analysis is referred to as fuzzy stochastic finite element method (FSFEM) (Sickert 2005; Möller et al. 2006). Here, a physically nonlinear FE model is applied, which is described in the following. An extended layer model with specific kinematics a so-called multi-reference-plane model (MRM) is used to describe the load-bearing behavior of RC constructions with textile strengthening. The MRM consists of the concrete layers and the steel reinforcement layers of the old construction, the strengthening layers comprised of the inhomogeneous material textile concrete, and the interface layers (Fig. 3). This multilayer continuum has the following kinematic peculiarities. Because the modification of the concrete layer thickness is very small and can be neglected, this means that εzz = 0 holds. Furthermore, the transverse shear stresses in the concrete layers have no significant influence on the deformation, which means that εxz and εyz can be set to zero. The deformation state of the concrete layers may thus be described by Kirchhoff kinematics. The independent degrees of freedom are assigned to a reference plane which can be selected arbitrarily.

591

The very thin strengthening layers are subject to the same kinematic assumptions. Kirchhoff kinematics with a reference plane are also assigned to each strengthening layer. The independent degrees of freedom of the strengthening layer lie in the reference plane. The bond between the layers of reinforced concrete and an arbitrary strengthening layer is modeled by an interface. The interface is an immaterial layer of zero thickness. The bonding state is assessed with the help of the relative displacements vx , vy , vz between the contact surfaces. In conjunction with a bonding matrix, the relative displacements enable postulations to be made regarding delamination and shear failure. The FE discretization of the multi-reference-plane model is based on the functional of the complementary energy extended by the static transition condi+

tions pr, e − p r, e = 0 to Opr, e and by the differential +

equilibrium conditions G ·σ eel + p e − ρ e · v¨ e = 0 in V e . mh =

⎧ τ2

n ⎨  τ1 e=1

+

wc (σ eel ) + (G · σ eel + p e

⎩

Ve

 −ρ e · v¨ e )T ve dV + (σ eel )T · εe0 dV Ve

−

+r, e T

( pr, e − p

) · vr, e dO

Opr, e

−

(pr, e )T · v

Ovr, e

with

wc (σ eel )—internal

+ r, e

⎫ ⎪ ⎬ dO

⎪ ⎭

dτ

complementary

(3)

energy;

+

G—matrix of differential operators; p e —external forces in V e ; ρ e —density in V e ;¨ve —internal acceleration in V e ; ε0e —initial strain; pr, e —internal forces +

in the boundary surface Opr, e ; p r, e —external forces along the boundary surface Opr, e ; vr, e —displacements

+

of the boundary surface Opr, e ; v r, e —prescribed displacements of the boundary surface Ovr, e ; τ —time. After some transformations, the quasi-static part +

of the equilibrium conditions (G · σ eel + p e ) and the kinetic energy are visible in the mixed hybrid functional. ⎧ τ2

n ⎨   + mh = wc (σ eel ) + (G · σ eel + p e )T ve dV ⎩ τ1 e=1

Ve



+ Ve

 1 e ρ · (˙v e )T · v˙ e dV 2



+

(σ eel )T · εe0 dV Ve



−

+

(pr, e − p r, e )T · vr, e dO

Opr, e

−

+

(pr, e )T · v r, e dO

Ovr, e

⎫ ⎪ ⎬ ⎪ ⎭

dτ

(4)

The same stress shape functions, boundary displacement shape functions and element displacement shape functions are chosen for all sub-elements. The physically nonlinear analysis of reinforced concrete strengthened with textile concrete is a non-conservative problem arising from crack formation, nonlinear material behavior, bonding and damage. In order to solve this non-conservative problem, a differential load variation is considered. Based on Equation (2), the equilibrium conditions can be determined. In order to deal with the transition to incremental (finite) load steps, an iterative technique must be adopted. The following holds for incremental load steps. An evaluation of the steady-state condition requires lumping of the k+1 layered sub-elements with the k interfaces to yield the MRM-element and leads to the equation of motion KT · q + M · ¨q − R − RK = 0

(5)

and possibly damaged construction. The process of system modification is simulated numerically by the incremental execution with the MRM. In order to describe the composite structure comprised of reinforced concrete and textile strengthening, different nonlinear material laws are applied to the individual sub-layers of concrete, steel and textile. Endochronic material laws for concrete and steel are applied for general loading, unloading and cyclic loading processes, and taking into account the accumulated material damage during the load history. The endochronic material law for concrete was adapted to the fine-grained concrete. A nonlinear elastic-brittle material law is used for the textile reinforcement. Under cyclic loading, damage occurs in the strengthening layer, in the fine-grained concrete matrix and in the textile structure as well as disruption of the bond between the old concrete and the TRC layer. These forms of damage and the additional plastic deformations may be described theoretically by means of plasticity and continuum damage theory (Steinigen 2006). The bond behavior between steel and concrete (tension stiffening) and the bond behavior between roving and fine grained concrete are taken into account phenomenologically on a macroscopic scale. The rovings are homogenized in the cross section and, thus, form one or more sub-layers of the strengthening layer. The different load-bearing characteristics of the boundary and core fibers are taken into account with the aid of a strain-dependent damage function. The parameters of this function are determined by means of uniaxial tensile tests. The results of the tensile test with AR glass (Fig. 4a) and carbon (Fig. 4b) reinforcement differ in the distance between the stress-strain relationship of the TRC specimen and that of the damaged roving. In the case of AR glass, the distance can be described with the stress σTS (tension stiffening). In the case of carbon, the strain adjustment εS is introduced. The following holds in the IIb state (ε > εe )  σ = Ef · Vf ·

ε + σTS if AR glass (ε − εS ) if carbon

(7)

with K T —tangential system stiffness matrix, M — system mass matrix, dR, dRK —differential load contributions. The matrix K T and the vectors dR, and dRK are identical to the corresponding quantities of the hybrid procedure. The relevant algebraic eigenvalue problem is given as   (6) K T − ω2 M q = 0 The post-strengthening of a RC structure means a system modification, i.e. a changing of a preloaded

Figure 4. Stress-strain relationship of fiber reinforced fine-grained concrete (qualitatively) with a) AR glass and b) carbon.

592

with the Young’s modulus Ef of the roving and the volumetric fiber content Vf .

4

EXAMPLES

4.1 Barrel vault The damaged RC barrel vault of the historic tax office in Zwickau (Germany) was strengthened with carbon reinforced fine-grained concrete on both sides. The barrel vault with 8 rooflights is 10 cm thick and is braced with 11 ribs with a height of 36 cm (Figs. 5 and 6). For the physically linear and nonlinear numerical simulation of the load bearing behavior of the structure, 3230 finite MRM-elements were used with 7 sub-layers in the vault and 13 sub-layers in the ribs. The influence of the following uncertain material and geometric properties on the load bearing behavior is taken into account: concrete compressive strength as fuzzy random variable (lognormal distribution (LND), μ = 24 N/mm2 , σ˜ = 2.0, 2.2, 2.4); cross section of steel reinforcement (F(a) = exp[b · (a − a0 )], b = 10, a0 = 3.14 cm2 /m); Young’s modulus (LND, μ = 28500 N/mm2 , cov = 0.1); compressive strength (LND, μ = 76.3 N/mm2 , cov = 0.1) and tentensile strength of fine-grained concrete ft = (0.3 + c ·

Figure 5.

Figure 6.

2/3

0.1) · fc with the parameter c (normal distribution, μ = −0.86, σ = 0.08) as random variables. The uncertain structural responses are computed with the aid of the FSFEM. Thereby, the deterministic nonlinear structural FE analysis is repeatedly performed. The loading process is divided into increments so that the system of nonlinear equations can be solved iteratively. Figure 7 shows 150 trajectories of the uncertain load-displacement dependencies of node 1943 in the case of increasing the dead load. The ultimate load increased due to the strengthening by 20 percent and the variation is decreased. 4.2

Pedestrian bridge

In Figure 8, a pedestrian bridge is shown made of TRC. The bridge consists of 10 precasted segments. Ribs strengthen the boundaries of the segments. The normal thickness (3 cm) increases at the edges, at the hand rails and at the supports. Additionally, two longitudinal stiffeners are formed in the bottom. The precasted segments are glued together and prestressed with unbonded cables. From the prestressing results that the fine grained concrete matrix is precompressed for service loads. Figure 9 displays the dimensions of the bridge. The measures refer to the reference planes of the finite elements. The effective span is 8.60 m, the

Barrel vault.

FE model.

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Figure 7.

Uncertain load-displacement dependency.

Figure 8.

TRC brigde situated in Oschatz (Germany).

structure is meshed with MRM-elements with one reference plane. Each element consists of five fine grained concrete and two textile layers. In the ribs, reinforcement layers are added which are modelled by the high-grade steel reinforcement. A rigid bond is assumed between the segments. Prestressing is also considered. As explained, the load-bearing behavior of TRC is influenced by the uncertain material properties. Here, the tensile strength of the fine grained concrete is modelled as fuzzy random field. Because of the handcrafted manufacture the geometry is also uncertain. Therefore, the thickness and the width are modelled as fuzzy triangular numbers with d˜ = 8, 9, 11 cm and b˜ = 2.0, 3.5, 5.0 cm. Prestressing forces and anchorage forces of tendons as well as dead load and live load are lumped together in the deterministic loading process. The uncertain structural responses are computed again with the aid of the FSFEM. Figure 10 shows 440 trajectories of the fuzzy random load-displacement dependency. The increase of the displacement v3 of the FE node 1735 as a result of increasing live load p is displayed. The result of loading test is also drawn in order to validate the numerical results.

Figure 9.

Figure 10.

FE model.

5

CONCLUSIONS

Analyzing a structure close to reality requires to consider the complete load and modification process. The parameters of the load and modification process are generally uncertain. They may be described by fuzzy random processes. The load-bearing behavior of folded-plate RC structures with textile strengthening under loading processes can be numerically simulated with the developed multi-reference-plane model (MRM). The presented algorithms are also applicable for the structural analysis of new TRC structures under consideration of uncertainty. ACKNOWLEDGEMENT The authors gratefully acknowledge the support of the German Research Foundation (DFG).

REFERENCES Bažant, Z.P. & Shieh, C.-L. 1980. Hysteretic Fracturing Endochronic Theory for Concrete, Journal of the Engineering Mech. Division, 106, pp. 929–950, Errata in: Journal of the Engineering Mech. Division, 107, pp. 728–729. Graf, W., Hoffmann, A., Möller, B. Sickert, J.-U. & Steinigen, F. 2007. Analysis of textile reinforced concrete structures under consideration of non-traditional uncertainty models, Engineering Structures, 29, pp. 3420–3431. Möller, B., Graf, W. & Beer, M. 2000. Fuzzy structural analysis using α-level optimization, Computational Mechanics, 26, pp. 547–565. Möller, B., Graf, W. & Beer, M. 2003. Safety assessment of structures in view of Fuzzy randomness, Computers & Structures, 81, pp. 1567–1582. Möller, B. & Beer, M. 2004. Fuzzy Randomness— Uncertainty in Civil Engineering and Computational Mechanics, Berlin, Heidelberg: Springer. Möller, B., Graf, W., Sickert, J.-U. & Beer, M. 2006. Timedependent reliability of textile strengthened RC structures under consideration of fuzzy randomness, Computers & Structures, 84, pp. 585–603. . Schenk, C.A. & Schuëller, G.I. 2005. Uncertainty Assessment of Large Finite Element Systems, Berlin, Heidelberg: Springer. Sickert, J.-U. 2005. Fuzzy-Zufallsfunktionen und ihre Anwendung bei der Tragwerksanalyse und Sicherheitsbeurteilung, TU Dresden, Veröffentlichungen Institut für Statik und Dynamik der Tragwerke, Heft 9. Steinigen, F. 2006. Numerische Simulation des Tragverhaltens textilverstärkter Bauwerke, TU Dresden, Veröffentlichungen Institut für Statik und Dynamik der Tragwerke, Heft 11. Viertl, R. 1996. Statistical Methods for Non-Precise Data, CRC Press, Boca Raton, New York, London, Tokyo Zimmermann, H.-J. 1992. Fuzzy set theory and its applications, Boston, London: Kluwer.

Uncertain load-displacement dependency.

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Properties of natural fiber cement boards for building partitions Y.W. Liu Department of Civil and Water Resources Engineering, National Chiayi University, Chiayi, Taiwan

H.H. Pan Department of Civil Engineering, Kaohsiung University of Applied Sciences, Kaohsiung, Taiwan

ABSTRACT: This paper uses bamboo fibers, coconut fibers, rice-husks and sugar cane-dregs, respectively, to make natural fiber cement boards for the building partition. Experimental results show that the unit weight of natural fiber cement boards are about 1430–1630 kgf/m3 . The flexural strength of natural fiber cement boards is 80% higher than that of typical building materials, except for rice-husks cement board. The length change in the absorption test is within the range of 0.09%–0.16%, and the thermal conductivity with 0.201–0.296 kcal/m · ◦ C · hr shows a good heat-resistant capability. For B10, C10 and S10 materials after impact test, no cracks, the detachment, pinholes and the split exist on impact surface, and the indentation diameters are below 26 mm. Besides, three cement boards containing 10% natural fibers satisfy the 2nd and 3rd rank of incombustibility standard. 1

INTRODUCTION

Many cement boards have been used as building partitions for over one century (Pamel & Schwarz 1979; Schwarz et al. 1983; Schwarz & Simatupang 1984; MacVicar et al. 1999). However, the unit weight of cement boards is still high, more than 2000 kgf/m3 . In order to adapt the varieties of the functions and the space for high-rise structure, the partitions to separate building space demand to be lightweight, easy to construct fast and assembled simple. In Taiwan, cement board, calcium silicate board and gypsum board are common used as the materials of building partition. Among them, moisture content for calcium silicate board and gypsum board gets up to 80% and 75%, respectively, due to the humid climate in Taiwan. High humidity made the partition deform and warp easily in use. One of the methods to improve the deformation of building partition affected by humidity is to add some fibers into the partition board. The useless agricultural products like rice-husks, sugar cane-dregs and coconut shell are always thrown away as the waste without any considerations in Taiwan. In fact, these agricultural wastes containing some natural fibers are valuable and can utilize to improve mechanical properties of the materials. This paper selects four natural fibers collected from bamboo, coconut shell, rice-husks and sugar canedregs, respectively, to produce natural fiber cement board considered as the building partition. We discuss material properties of natural fiber cement board including water absorption, bulk density, length change

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induced by absorption of water, impact endurance, fireproof capability and heat-resistant capability. The experimental results can be used as a reference in building industry.

2 2.1

EXPERIMENTAL PROGRAM Materials

Four natural fibers, bamboo fiber, coconut fiber, ricehusks and sugar cane-dregs, were added to the cement board, a kind of natural fiber cement board (NFCB). A comparison material is the cement board without adding natural fibers inside. NFCB consists of cementitious matrix and natural fibers. The constituents of cementitious matrix include: (1) Type I Portland cement (ASTM C150); (2) slag with a specific gravity of 2.89 supplied by China Hi-Ment corporation (Taiwan); (3) river sand having a fineness modulus of 2.68, a specific gravity of 2.63, and an absorption of 2.0%; and (4) fresh water. To prepare bamboo fibers, we first cut bamboo wood into the pieces with 40 mm length, and then use the disintegrator to separate bamboo wood into the fibers, shown in Fig. 1. Sieve analysis for bamboo fibers is shown in Table 1, where dry specific gravity of 0.85, specific gravity of 0.93 with 10–12% moisture content in air and water absorption of 66% after 48 hours’ absorption test. Bamboo fibers retaining in sieve 4 are shown in Fig. 2 with 13 mm fiber length, passing through sieve 4 and retaining in sieve 8 are shown in Fig. 3 with 15 mm fiber length, and in sieve

Figure 3. Bamboo fibers pass through sieve 4 and retain in sieve 8 with 15 mm length.

Figure 1. Table 1.

Bamboo fibers before sieve analysis. Sieve analysis of bamboo fibers.

Sieve

Retaining (%)

Accumulation (%)

3/8 #4 #8 #16 #30 #50 Pan Total

0 6.0 31.1 33.9 18.0 8.8 2.2 100

0 6.0 37.1 71.0 89.0 97.8 100.0 –

Figure 4. length.

Figure 2. length.

Bamboo fibers retain in sieve 50 with 5∼15 mm

Bamboo fibers retain in sieve 4 with 13 mm

50 with 5∼15 mm fiber length are shown in Fig. 4, respectively. Only the sizes of bamboo fiber between sieve 4 and sieve 50 were chosen to manufacture the fiber/cement board here. Meanwhile, as we know bamboo fiber can retard the hydration of cement. Thereby, we need to overcome this retardant reaction in bamboo/cement boars by using following treatments. First, bamboo fibers were soaked in water, and then dried by heat. After that, bamboo fibers were also immersed in the solution with 20 to 1 of 1% organic titanium solution-to-bamboo fiber ratio by weight. Finally, bamboo fibers are ready to use after drying.

Figure 5. Sugar cane-dregs.

Because sugar cane-dregs have the same retardant hydration effect to the cement, we also need to do the same treatments as bamboo fibers before the use. Material properties of sugar cane-dregs have 68% absorption of water and specific gravity of 0.63, and shown in Figure 5.

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Table 2.

Mixture proportions of NFCB (unit: kgf/m3 ).

Material

Air Fiber∗ Water Cement Slag Sand content

Comparison B10 C10 R10 S10

0 10 10 10 10

360 360 360 360 360

432 432 432 432 432

288 288 288 288 288

797 534 534 534 534

10 10 10 10 10

* Fiber is in volume percent.

2.3 Figure 6.

Experimental method

Natural fibers are difficult to mix well with the cementitious material. The mixture method here for NFCB is conducted as follows.

Coconut fibers.

1. Weigh the constituents shown in Table 2. 2. Mix the cement and slag together in two minutes at dry condition, and then pour 75% water into the mixture, and finally blend two minutes by middle speed of the mixer. 3. Turn off the mixer, add natural fibers into the mixture material, and then blend one minute by middle speed of the mixer. 4. Turn off the mixer again, pour the remaining 25% water into the mixture, and then blend ten minutes by middle speed of the mixer.

Figure 7.

Rice-husks.

Besides, coconut fiber has 92% water content after 24 hours’ absorption test and specific gravity of 0.62, respectively, shown in Fig. 6. Fig. 7 is rice-husks with 12% absorption and specific gravity of 0.53. 2.2 Mixture proportions The mixture proportions of natural fiber cement board are shown in Table 2, where the water-to-cementitious matrix ratio is 0.5 by weight, and the cement-to-slag ratio is 1.5 by weight, or 60% cement and 40% slag, respectively. In order to compare the effect of natural fibers, we add natural fibers of 10% in volume to cement board. Totally, four kinds of NFCB represented by B10, C10, R10 and S10 shown in Table 2, are referred as the cement board containing bamboo fiber, coconut fiber, rice-husks and sugar cane-dregs, respectively. Besides, the comparison material shown in Table 2 means the cement board containing no natural fibers inside.

Each batch of materials was prepared for nine samples with the size of 50 × 50 × 50 mm for compressive test, three samples with 100 × 100 × 10 mm for the absorption test and bulk specific gravity test, three samples with 40 × 160 × 10 mm for length change induced by absorption of water, six samples with 250 × 350 × 20 mm for bending test, one with 200 × 200 × 10 mm for heat-resistant capability, one with 220 × 220 × 10 mm for fireproof capability, and one with 300 × 300 × 10 mm for impact endurance test, respectively. All samples were placed on the vibration table to shake one minute. The surface of cement boards was leveled to be smooth by the trowel, and then one hour later, a surcharge with 30 g/cm2 were loaded to confine the size of NFCB. Samples were removed from the mold 24 hours later, and placed indoor for curing and for testing.

597

3 3.1

RESULTS AND DISCUSSION Bulk density and water content

We measure bulk density and water content of NFCB in accordance with ASTM C1185, and results were shown in Table 3. In Table 3, the bulk density of the comparison material is about 1860 kgf/m3 . Obviously, the bulk density of natural fiber cement boards for

B10, C10, R10 and S10 are all lighter than that of the comparison material about 12.4%, 15.1%, 21.2%, and 23.1%, respectively. According to CNS 3802 requirements, the optimum density for fiber cement boards is claimed to 1300–1400 kgf/m3 , where CNS means Chinese National Standards. It is anticipated to reduce the bulk density of NFCB if the fibers adding to cement board are more than 10% in volume. In Table 3, the water absorption of all NFCB with the value from 11.2% to 13.1%, respectively, is higher than that of comparison material. Among them, the water content of rice-husks cement board (R10) seemly does not increase much, only 4.7% increases. 3.2

Length change due to absorption of water

Length change of cement board after the absorption test is shown in Table 4. The length change of the comparison material is only 0.01%, but the length change of NFCB is within the range of 0.09%–0.16%. Although the length change of natural fiber cement boards is higher than that of the comparison material, the value of length change for NFCB is still small as compared with calcium silicate board and gypsum board. The building partition made from NFCB containing 10% natural fibers is suitable in use. 3.3

Compressive strength and flexural strength

The specimens were tested at the material age of 28 days in compression (ASTM D1037) and bending test (ASTM C1185), and the experimental results are shown in Table 5. The compressive strength of natural fiber cement boards with the value 14.2∼23.8 N/mm2 is lower than that of the comparison material with

26.9 N/mm2 , especially the compressive strength of R10 is 14.2 N/mm2 and decrease more than 30%. This is because the intrinsic quality of natural fibers can strengthen the tensile strength but not the compression. The compressive strength of the cement board with coconut fibers (C10) is 23.8 N/mm2 , and is only 11.6% less with respect to the comparison material. On the contrary, the flexural strength of natural fiber cement boards is higher than that of the comparison material, except for the rice-husks cement board (R10) shown in Table 5. For example, the flexural strength of bamboo fiber cement board (B10) and of the comparison material is 7.48 N/mm2 and 4.15 N/mm2 , respectively. The flexural strength of B10 is almost 80% stronger. 3.4

Impact test

Cement boards were tested by impact loads in accordance with CNS9961, where the dimension of the impact specimen is of 300 mm × 300 × 10 mm, the weight of impact ball is 530 grams with the diameter of 51 mm, and the drop distance of the ball is 1400 mm, respectively. The impact results are shown in Table 6, where the number marked 1, 2, 3, and 4 is represented to the defects with the crack, the detachment, the pinholes, and the split, respectively. The comparison material has the existence of defects like cracks, the detachment, pinholes and the split on the impact surface after the impact test, shown in Fig. 8. From Table 6, only the cement board with ricehusks (R10) has similar defects like the comparison material does. Table 5.

Table 3.

Bulk density and water content of cement boards.

Material

Bulk density (kgf/m3 )

Water content (%)

Comparison B10 C10 R10 S10

1860 1630 1580 1465 1430

10.7 12.8 13.1 11.2 12.6

Material

Compressive strength

Flexural strength

Comparison B10 C10 R10 S10

26.9 22.1 23.8 14.2 19.2

4.15 7.48 6.87 3.55 4.54

Table 6. Table 4.

Strength of cement boards (N/mm2 ).

Length change of cement boards.

Material

Length change (%)

Comparison B10 C10 R10 S10

0.01 0.13 0.16 0.09 0.12

Impact test of cement boards.

Material

Indentation diameter (mm)

Impact* surface

Comparison B10 C10 R10 S10

– 10 15 – 26

1, 2, 3, 4 no defects no defects 1, 2, 3, 4 no defects

*1: crack, 2: detachment, 3: pinholes, 4: split.

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Reverse* surface small crack crack crack

After the impact test, the cement board with 10% bamboo fibers (B10) did not find the cracks, the detachment, pinholes and the split on the impact surface, and the overall indentation diameter is about 10 mm, shown in Fig. 9. Meanwhile, the reverse side of impact surface (reverse surface) displayed a convex surface with some microcracks for B10 shown in Fig. 10. For C10 material shown in Table 6, the impact surface also did not discover any defects after the impact test, and the indentation diameter is about 15 mm. The reverse surface for C10 shows a convex surface containing obvious cracks. For the cement board adding sugar cane-dregs (S10), no defects found on the impact surface, and the overall indentation diameter is about 26 mm. The reverse surface for S10 also shows a convex surface with obvious cracks.

3.5

Thermal conductivity

To be building partitions, cement boards should have heat-resistant capability. Here, we examine the heatresistant capability of cement boards by using the thermal conductivity in accordance with ASTM C518, and results are shown in Table 7. The thermal conductivity of the comparison material was measured and is 0.826 kcal/m · ◦ C · hr. For natural fiber cement boards, the values of thermal conductivities is within 0.201 and 0.296 kcal/m ·◦ C · hr. The thermal conductivity of B10 and C10 is only one fourth of the comparison material shown in Table 7. Therefore, NFCB have desirable heat-resistant capability compared with that the comparison material. 3.6

Fireproof capability

Fireproof capability is also an important index for building partitions. Cement boards were tested by incombustibility test claimed by ASTM E84 and CNS 6532, and the results are shown in Table 8. Figs. 11–12 show the heating surface and the reverse surface of B10 after the incombustibility test, respectively. Fireproof capability of B10 and C10 materials satisfies the 2nd standard of incombustibility claimed

Figure 8.

Fracture of counterpart material after impact test.

Figure 10. Table 7.

Figure 9.

Impact surface of B10 after impact test.

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Reverse surface of B10 after impact test. Thermal conductivity of cement boards.

Material

Thermal conductivity (kcal/m · ◦ C · hr)

Comparison B10 C10 R10 S10

0.826 0.201 0.217 0.266 0.296

Table 8.

Incombustibility of cement boards.

Material

Incombustibility rank

Comparison B10 C10 R10 S10

2nd standard 2nd standard 2nd standard – 3rd standard

bamboo, the coconut shell, rice-husks and sugar canedregs, respectively, to make natural fiber cement board as the material of building partitions. In this research, the volume fraction of natural fiber added to cement board is 10%. The mixture method of making natural fiber cement boards is also presented. Experimental results show that most of material properties for cement boards containing bamboo fiber, coconut fiber, and sugar cane-dregs are better than those of the comparison cement board (without natural fibers), including the incombustibility satisfied the national standards. Although the bulk density of natural fiber cement boards we made is about 1430–1630 kgf/m3 , higher than 1400 kgf/m3 claimed by the Code, we can afford to add natural fibers more than 10% volume fraction to lower the weight of natural fiber cement boards in future. It seems that we can use those three natural fibers to make the natural fiber cement board used as building partitions in the building industry. ACKNOWLEDGMENTS

Figure 11. test.

Heating surface of B10 after incombustibility

The authors would like to thank the Taiwan National Science Council under NSC 97-2221-E-415-011MY2, and Architecture and Building Research Institute, Ministry of the Interior in Taiwan, for the financial supports. REFERENCES

Figure 12. test.

Reverse surface of B10 after incombustibility

by CNS 3802, and is the same as that of the comparison material. For the S10 material, it reaches the 3rd standard of incombustibility. However, R10 material failed under the incombustibility test. 4

Kossatz, G., Lempfer, K. &. Sattler H. 1983. Woodbased panels with inorganic binders. FESY Annual Report 1982/1983: 98–110 WKI-Mitteilung Nr.364, Braunschweig. MacVicar, R., Matuana, L.M. & Balatineez, J.J. 1999. Aging mechanisms in cellulose fiber reinforced cement composites. Cement and concrete Composites 21: 189–196. Pamel, H. & Schwarz, H.-G. 1979. Technologie und verfahrens-technik zemetgebundener spanplatten. Holz als roh-und werkstoff 37: 195–202. Schwarz, H.-G. & Simatupang, M.H. 1984. Eignung des buchenholzes zur herstellung zementgebundener holzwerkstoffe. Holz als roh- und werkstoff 42: 265–270.

CONCLUSIONS

This paper attempted to make use of natural fibers collected from useless agricultural products such as the

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Studies on glass fiber reinforced concrete composites – strength and behavior B.L.P. Swami Department of Civil Engineering, Vasavi College of Engineering, Hyderabad, India

A.K. Asthana Department of Civil Engineering, JNTU College of Engineering, Hyderabad, India

U. Masood Department of Civil Engineering, Deccan College of Engineering & Technology, Hyderabad, India

ABSTRACT: It has been established that discrete metallic (steel) fibers when added in certain percentage to the concrete improve the strength properties as well as crack resistance and ductility. Glass fibers in cement mortar have been tried in applications like architectural features, panel walls, tunnel lining etc. The present paper outlines the experimental investigation conducted on the use of glass fiber with structural concrete. CEMFIL Anti Crack High Dispersion, Alkali resistant glass fiber of diameter 14 micron, having an aspect ratio of 855 was employed in percentages, varying from 0.2 to 1.5 in concrete and the properties of this FRC (Fiber Reinforced Concrete) like compressive strength, tensile strength, and flexural strength were studied. In addition by employing steel fiber along with glass fiber in concrete, the properties of Mixed Fiber Reinforced Concrete (MFRC) were also studied. Conclusions are drawn on properties like strength, ductility and crack resistance of structural concrete. 1

INTRODUCTION

Alkali Resistant Glass Fiber is a recent introduction in making fibrous concrete. Glass fiber which is originally used in conjunction with cement was found to be affected by alkaline condition of cement. Therefore Cem-Fil (2002) alkali resistant glass fiber has been developed and used. GRCA (2006). The paper presents an experimental investigation on the use of Cem-Fil ARC14 306 HD and dual mixed fibers—The alkaline resistant glass fibers and monofilament steel fibers, with locally available materials. Experiments have been carried out by several authors using fibers of glass, carbon, asbestos, polypropylene etc. Heurik et al. (2004) has outlined the classification and structural applications of composite materials. Majumdar et al. (1991) has worked on the development of fiber reinforced cements. Sivakumar et al. (2007) has studied the properties of FRC using high percentage dosages of hybrid fiber like steel, glass and polypropylene. High strength concretes were produced by using the above fibers. From the various investigations studied it is clear that non metallic fibers like glass and polypropylene can be tried in FRC to derive benefits like more ductility, impact resistance, crack resistance etc. In the present experimental investigations, strength properties in compression, split tension and flexure

601

using glass fiber with aspect ratio of 857:1 with various volume percentages of fibers of 0.20. 0.25, 0.30, 0.50, 1.00 and 1.50 are studied. Investigation is also carried out by mixing glass fibers with steel fibers. The total volume percentages of mixed fibers at 0.50, 0.75.and 1.00 are adopted. From the total mixed fiber content, glass fiber at percentages 0, 25, 50 and 75 was used as replacement to steel fiber. The strength properties of mixed fiber reinforced concrete are also studied. The optimum strength properties have been arrived.

2

EXPERIMENTAL INVESTIGATION

The details of materials used in the present experimental investigation are as follows. 2.1

Cement

OPC of 53 grade having specific gravity of 2.9 and fineness of 2800 cm2 /gm is used. 2.2

Coarse aggregate

Machine crushed well graded angular granite aggregate of nominal size 20 mm from local source are used.

The specific gravity is 2.6 with flakiness index of 4.35 percent and elongation index of 3.64 percentage. It is free from impurities such as dust, clay and organic matter. 2.3

Fine aggregate

River sand locally available is used. The specific gravity is 2.4. 2.4

Glass fiber

Cem-Fil ARC14 306 HD glass fiber is used. The properties are shown in Table 1. 2.5

Water

Locally available potable water is used. 2.7

Concrete mix

The M20 grade of concrete and quantities used per cubic meter are shown in Table 2. The water cement ratio has been fixed depending upon the compaction factor test, keeping medium workability. 2.8

RESULTS AND DISCUSSIONS

3.1

Workability

For fiber glass percentages of 0.2, 0.25, 0.3 by volume, the workability is found to be nearly 0.88 by compaction factor. For 0.5 percent it is 0.85 compared to 0.9 compaction factor for plain concrete mixes. The water Cement ratio taken is 0.5. Higher percentages of fiber beyond 0.5 percent and up to 1.5 percent require super plasticizer. At 2 percent fiber content balling has occurred and mix was not in a workable condition with a W/C ratio of 0.5. The replacement of Steel fiber with Glass fiber by 0, 25, 50 and 100 percentages from total fiber content of 0.5, 0.75, and 1.0 by volume, the workability was affected marginally.

Steel fiber

Monofilament steel fiber of 1 mm diameter & aspect ratio 55 is used. 2.6

3

Mixing

After mixing carefully in a pan mixer, the mix was cast in moulds. For each percentage of fiber sufficient number of cubes, cylinders and flexural beams were cast for testing at the ages of 7d and 28d.

Table 2.

Elastic modulus (GPa)

Tensile strength (MPa)

Density (micron)

Length (mm)

No. of fibers (million/kg)

AR-Glass 2.6

73

1700

14

12

212

Compressive strength

From Table 3, it is observed that with increase in fiber percentage, the compressive strength also increases with age. At the age of 7 days with 1.5 percentage fiber the compressive strength is 16.36 percent in excess over the strength of reference mix and for 28 days it is 17.49 percent in excess of reference mix. As the percentage replacement of steel fiber by glass fiber is increased, the compressive strength decreases, but on the overall it is more than that of control concrete as can be seen from the Table 4. Table 3. Compressive Strength of GFRC with various percentages of glass fibers.

Glass Fiber (%) 0.00 0.20 0.25 0.30 0.50 1.00 1.50

Properties of glass fiber Cem-Fil ARC14 306 HD.

Density (t/m3 )

Fibers

Table 1.

3.2

Compressive strength (N/mm2 ) 7d

28d

Percentage increase over the reference mix 28d

34.48 36.10 36.78 38.54 38.75 38.89 40.15

46.30 49.80 51.00 54.00 54.07 54.17 54.40

7.50 10.15 16.60 16.80 17.00 17.49

Table 4. Typical compressive strength results for 0.75 percent total fiber content at 28 days. Glass fiber (%)

Steel fiber (%)

Ultimate compressive load (KN)

Compressive strength (N/mm2 )

Increase in strength (%)

0 25 50 75 100 0

100 75 50 25 0 0

637.0 625.6 612.8 594.5 572.7 463.0

63.70 62.56 61.28 59.45 57.27 46.30

27.32 26.00 24.45 22.12 19.15 0.00

Materials required for 1 m3 of concrete.

Grade

Cement (kg)

Fine aggregate (kg)

M20

425

682

Coarse aggregate (kg)

Water cement ratio

1277.76

0.5

602

3.3

Split tensile strength

45.56 percent in excess over the strength of reference mix. The variation of flexural strength at 28 days with various percentages of glass fiber of 0, 25, 50, 100 percent by volume used as replacement for steel fiber in total fiber content of 0, 0.5, 0.75. 1.0 percentages were studied and results for typical 0.75 total fiber content are presented in Table 8. It is observed that as the percentage of total fiber content is increased, the flexural strength also increases. As the percentage replacement of steel fiber by glass fiber is increased and steel fiber percentage is decreased, the flexural strength goes on decreasing.

From Table 5, it is observed that with increase in fiber percentage, the split tensile strength also increases with age. At the age of 7 days with 1.5 percentage fiber the split tensile strength is 44.48 percent in excess over strength of reference mix and for 28 days it is 65.45 percent in excess. The variation of split tensile strength at the end of 28 days with various percentages of glass fibers of 0, 25, 50 and 100 percentages by volume used as replacement of Steel fibers in total fiber content of 0, 0.5, 0.75 and 1.0 percentages were studied and results for typical 0.75 total fiber content are presented in Table 6. It is observed that as the percentage of total dual fiber content (steel fiber and glass fiber) is increased, the split tensile strength also increases. It is also observed that as the percentage replacement of steel fiber by glass fiber is increased and steel fiber percentage decreases, the split tensile strength goes on decreasing. 3.4

3.5

Beam specimens of M20 Mix with various percentages of fibers have been tested for flexural strength under two point loading as per the standard specifications. The flexural specimens tested have exhibited ductility characteristics. At the failure load a diagonal crack has appeared in between the loading points and the specimens have not failed suddenly. The failure is not brittle and is entirely different from that of plain concrete, where failure is brittle. The ductility characteristics exhibited by the specimens are due to the introduction of fiber in the mix.

Flexural strength

From Table 7, it is observed that with increase in fiber percentage, the flexural strength also increases with age. At the age of 7 days with 1.5 percentage fiber the flexural strength is 84.21 percent in excess over the strength of reference mix and for 28 days it is

Table 7. Flexural strength of GFRC with various percentages of glass fibers.

Table 5. Split tensile strength of GFRC with various percentages of glass fibers. Split tensile strength (N/mm2 ) Glass fiber (%)

7d

28d

0.00 0.20 0.25 0.30 0.50 1.00 1.50

2.45 2.67 2.86 3.10 3.13 3.18 3.54

3.30 3.80 3.89 4.30 4.35 4.60 5.46

Ductility characteristics

Percentage increase over the reference mix 28d

Glass fiber (%) 0.00 0.20 0.25 0.30 0.50 1.00 1.50

15.15 17.88 30.30 32.00 39.39 65.45

Flexural strength (N/mm2 ) 7d

28d

Percentage increase over the reference mix 28d

3.8 4.5 4.8 5.1 5.50 6.25 7.00

5.75 6.30 6.80 7.10 7.36 7.62 8.37

9.56 18.26 23.48 28.00 32.52 45.56

Table 6. Typical split tensile strength results for 0.75 percentage total fiber content at 28 days.

Table 8. Typical flexural strength results for 0. 75 Percentage total fiber content at 28 days.

Glass fiber (%)

Steel fiber (%)

Ultimate split tensile load (KN)

Split tensile strength (N/mm2 )

Increase in strength (%)

Glass fiber (%)

Steel fiber (%)

Ultimate flexural load (KN)

Flexural strength (N/mm2 )

Increase of strength (%)

0 25 50 75 100 0

100 75 50 25 0 0

342.84 350.62 338.61 322.35 304.67 251.66

5.08 4.96 4.79 4.56 4.31 3.56

29.92 28.30 25.70 21.93 17.41 0.00

0 25 50 75 100 0

100 75 50 25 0 0

15.62 15.76 15.52 14.96 14.12 11.52

7.95 7.88 7.76 7.48 7.06 5.76

27.55 26.90 25.77 22.99 18.42 0.00

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3.6

Cracking characteristics

Observation of specimens during Split tensile strength test shows a single crack occurring at failure along diameter of cross section without any appearance of longitudinal crack. This can indicate that glass fiber contributes to crack resistance. It is observed that failure has taken place gradually with the formation of cracks. In the case of plain concrete specimens the failure is sudden and brittle. Hence it is established that the presence of fibers in the matrix has contributed towards arresting sudden crack formation.

6.

7. 4

CONCLUSIONS

Based on the present experimental investigation conducted and the analysis of test results, the following conclusions are drawn. 1. Higher percentages of Glass fibers from 1.0 percentage affect the workability of concrete, and may require the use of super plasticizers (workability agents) to maintain the workability. For the nominal M20 mix with a water cement ratio of 0.5, the workability of concrete is only marginally affected even with a total fiber content of 1.0 percent by volume. Steel fiber of 1 mm diameter and length of 55 mm having an aspect ratio of 55 can be satisfactorily mixed along with glass fiber having an aspect ratio of nearly 857, to increase the strength and other characteristics. 2. The compressive strength of ‘‘Cemfil Anti crack HD’’ fiber concrete is found to be maximum at 1.5 percentage of fiber. With this percentage there is an increase of 17.49 percent for M20 Grade mix at 28 days. 3. The Split Tensile strength of ‘‘Cemfil Anti crack HD’’ fiber concrete is found to be maximum at 1.5 percent of fiber mix and there is an increase of 65.45 percent for M20 Grade mix at 28 days. 4. The Flexure strength of ‘‘Cemfil Anti crack HD’’ fiber concrete is also found to be maximum at 1.5 percent of fiber, and there is an increase of 45.56 percent for M20 Grade mix at 28 days. 5. The compressive strength of dual fiber concrete is maximum at 1.0 percent total fiber content of steel. With this percentage there is an increase of 29.03 percent at 28 days compared to plain concrete. With a total of 1.0 percent glass fiber by volume the increase of compressive strength at 28 days is 21.53 percent compared to plain concrete. There is substantial increase in the compressive strength for mixed fiber combination As the percentage of steel fiber is reduced and glass fiber is increased, the

8.

9.

compressive strength is getting reduced compared to that of 100 percent steel fiber in the matrix. The split tensile strength of dual fiber concrete is maximum at 1.0 percent total steel fiber content. With this percentage there is an increase of 31 percent at 28 days compared to plain concrete. With a total of 1.0 percent glass fiber by volume the increase of split tensile strength at 28 days is 22.77 percent compared to plain concrete. As the percentage of steel fiber is reduced and glass fiber is increase, the split tensile strength is getting reduced compared to that of 100 percent steel fiber in the matrix. The flexural strength of dual fiber concrete is found to be maximum at 1.0 percent total steel fiber content. With this percentage there is an increase of 31 percent at 28 days compared to plain concrete. With a total of 1.0 percent glass fiber by volume the increase of flexural strength at 28 days is 21.74 percent compared to plain concrete. The ductility characteristics have improved with the addition of glass fibers. The failure is gradual compared to that of brittle failure of plain concrete. The Ductility characteristics improved by adding Steel fibers also. Cracks can be controlled by introducing glass fibers. Cracks have occurred and propagated gradually till the final failure. This phenomenon is true with all the percentages of glass fiber. Glass fiber also helps in controlling the shrinkage cracks.

Compared to metallic fibers like steel, alkali resistant glass fiber gives corrosion free concrete. By judiciously combining Glass fiber with Steel fiber, optimum FRC possessing required strength and other properties can be produced. REFERENCES Heurik Strang, Victor C. Li. 2004. Classification of Fiber Reinforced Cementitious Materials for Structural Applications 6th Rilem symposium on FRC, 20–22 Sep. Varenna, Italy, pp. 197–218. Majumdar, A.J., Laws, V. 1991. Building Research Establishment book on Glass fibre reinforced cement 2nd ed. Oxford; Boston: BSP Professional. Saint Gobain Vetrotex, Cem – Fil. 2002. Why Alkaline Resistant Glass Fibers. In Technical data sheets. www.cemfil.com Sivakumar, A. and Santhanam Manu. 2007. Mechanical Properties of High Strength Concrete Reinforced with Metallic and Non-Metallic Fibers. Cement and Concrete Composites (29) pp. 603–608. GRCA 2nd edition. 2006. Specification for the manufacture, curing and testing of GRC products.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Use of bamboo composites as structural members in building construction T.H. Nguyen & T. Shehab California State University, Long Beach, CA, USA

A. Nowroozi Tajann – Engineering & Construction, Inc., CA, USA

ABSTRACT: This paper presents a study that was conducted to investigate the potential of using bamboo composites as structural members in building construction. First, to develop basic information about the flexural performance of the bamboo composite, different bamboo composite specimens with different layer configurations were fabricated. The specimens then were tested in three-point bending to failure. Results showed that the bamboo laminates with reinforcing pins were the strongest and the stiffest of the specimens tested.

1

to develop basic information about the flexural performance of the bamboo composite, different bamboo composite specimens with different layer configurations were fabricated: one group of specimens had reinforcing pins inserted into the bamboo laminate and another one without pins. The specimens then were tested in three-point bending to failure. The following sections of this paper are organized as follows: first, the fabrication process of bamboo specimens is explained; then the flexure tests and results are presented and discussed; and finally the conclusion section summarizes the project outcomes.

INTRODUCTION

On a global scale, the fact that forests are being over logged and severely destroyed is attracting worldwide concern for the global environment. Materials that offer alternatives to traditional wood products, such as bamboo, must be considered. In effect, bamboo is one of the fastest renewable plants with about 1250 species (Austin & Ueda 1972; ABS 2008) and a maturity cycle of 3–4 years, thus making it a huge and highly attractive natural resource compared to forest hardwoods. In the construction industry, bamboo has been increasingly adopted as an environmentally acceptable building material for construction projects in many countries over the world because of its super properties like high strength to weight ratio, high tensile strength and other factors like low cost, easy availability and harmless to the environment during service. Recently, many bamboo-based products (e.g. bamboo boards, bamboo mats, bamboo ceiling tiles, and bamboo flooring boards) have been manufactured using laminate technology (Bansal & Zoolagud 2002; INBR 2001). The process involves basic engineering of the bamboo before the manufacture of the specified building materials. Such bamboo-based products, however, were not designed to be used as structural members (e.g. beams, columns, and trusses). These manufactured products have very limited and in some cases no structural capacity. As results, a number of researchers (Ghavami 1995; Yao & Li 2003; Li et al. 2001) have recently investigated different engineering methods for fabricating bamboo composites in attempt to maximize the structural strength of the products. This paper presents a study that was conducted to investigate the potential of using bamboo composites as structural members in building construction. First,

2 2.1

FABRICATION OF SPECIMEN Preparing materials

Materials required for fabricating the composite specimen include bamboo laminates and adhesive. 2.1.1 Bamboo laminates Bamboo laminates with dimensions of 400 × 25 × 15 mm were supplied by a bamboo flooring manufacturer. These bamboo laminates were made from slivers milled out from the bamboo culm. The bamboo strips were then dried and treated for anti-fungal before being subjected to surface and edge gluing. The slivers were arranged systematically and subjected to a hydraulic hot press (Temp ∼150◦ C and Pressure ∼17 kg/cm2 ) to make them into panels. The pressed laminate panels were then put through trimming, sanding and grooving machines to give a pre-finish shape. It was noticed that the bamboo strips were processed by super heat steam, which not only killing moth inside but also carbonizing the strips. Thus, the bamboo laminates will be able to

605

resist moth, mildew, and insect diseases and will be more durable than natural bamboo. It has been reported that mechanical properties of bamboo vary from species to species and within a species the difference in mechanical properties depends on the location of the bamboo sample in the stem (Janssen 1981). The bottom area of the culm of a bamboo pole is usually stronger than that at the top. Ages of the culm and moisture content are also the factors that determine the mechanical properties. In general, the strength of a bamboo reaches maximum when the bamboo is fully mature. In addition, the inner fiber is weaker in flexure than the outer fiber; therefore, the bamboo laminates in the specimen were layout such that the outer fiber of bamboo is placed at top and bottom faces of the specimen in order to take advantage of their maximum strength. Table 1 shows the mechanical properties of bamboo laminates and teak wood. It was found from the Table that the bamboo laminates (using species of bambusa bambose), in overall, have higher structural strength than teak wood. 2.1.2 Adhesive The adhesive material used for binding bamboo laminates of the specimen was a biobased product, named KR Bond. This is a water-based polymer-isocyanate adhesive. It has been found that formaldehyde-free KR Bond does not contribute harmful Volatile Organic Compounds (VOCs) to the indoor atmosphere (Smith 1987). 2.2

Specimen

The fabrication procedure of the specimen can be summarized as follows: – Arrange the bamboo laminates 400 × 25 × 15 mm obtained from a manufacturer (Refer to Subsection

Pin 10 cm 25 mm

45 mm

Bamboo laminates Figure 1.

Property

Unit

Bamboo laminate (Bambusa bambose)

Teak wood @12% M.C

Density R E S C W Face Edge

kg/cm3 N/mm2 N/mm2 N/mm2 N/mm2

715 75 10,970 11 72

604 94 11,720 9 52

N N

4,540 3,624

3,900 2,881

M.C. = moisture content; R = modulus of rupture; E = modulus of elasticity; S = block shear strength; C = compressive strength; W = screw withdrawal strength.

The specimen with bamboo pins.

2.1.1 Bamboo laminates for the fabrication of these bamboo laminates) in 3 layers to make a specimen with a dimension of 400 mm length, 25 mm thickness, and 45 mm width. – Clean the surface of the bamboo laminates and apply adhesive bonding between the laminates using a hot press. – Insert a bamboo pin having a diameter of 5 mm into the specimen at spacing of 10 cm on center. The construction and dimensions of the bamboo specimen are shown in Figure 1. Additionally, there are two groups of specimen that were fabricated for this experiment program: one includes five specimens with pins and the other has five specimens without pins. These pins are made of bamboo and have a diameter of 5 mm. It is anticipated that the insertion of the bamboo pin would improve the bonding strength of the bamboo laminates.

3 3.1

Table 1. Compare mechanical properties of bamboo laminates and teak wood.

400 mm

FLEXURE TESTS AND RESULTS Flexure tests

The flexure tests (or three-point bending tests) were conducted in accordance to ASTM (American Standards of Testing and Materials) D3043-87, ‘‘Standard Test Methods for Testing Structural Panels in Flexure.’’ A SATEC (Model 22 EMF) universal testing machine was used to perform the three-point bending test. The system is a closed loop computer controlled load frame with a high speed data acquisition system. The machine has a maximum loading capacity of 100 kN at room temperature. The flexural load was set at a stroke rate of 0.5 mm/min. The span and the width for the specimens were 300 mm and 45 mm respectively. The midpoint deflection was determined by the linear variable differential transducer. The bending test setup is shown in Figure 2.

606

Figure 4.

Figure 2.

addition, Figure 3 shows that the residual flexural stress of both specimens was about 10 MPa when the midspan deflections in both curves reached 25 mm. The failure mode of the flexural test is shown in Figure 4. It can be seen that the specimen failed by the weak tensile strength of bamboo laminates (i.e. the bottom face of the specimen) and due to the high bond strength of adhesive between bamboo laminates, there was no interfacial delamination shown after the failure. After the peak stress, due to the failure of the bamboo laminates, the flexural stress significantly dropped with the increase in deflection.

Three-point bending test setup.

120 100 Flexural Stress (MPa)

Failure mode.

80 60 40 20

4

CONCLUSIONS

0 -20

0

5

10

15

20

25

30

Midpoint deflection (mm)

Figure 3. Stress-deflection curve a: Specimen with pin b: Specimen without pin.

3.2

Results

The test results are presented in Figure 3 showing the flexural curves of stress—deflection at the mid-point of the span. The curve denoted ‘a’ shows the flexural test results of the specimen with pins and the curve denoted ‘b’ is for the specimen without pins. As seen in Figure 3, the flexural behaviors of the bamboo laminates with and without pins are different. For both bamboo laminate specimens, up to the stress of 20 MPa, it showed a linear behavior. It is found that the gradient of the two curves slightly change with an increase in stress. The rigidity of the specimen with pins is obviously larger than that without pins as the former started fracturing at a higher ultimate flexural stress (about 105 MPa) compared to the ultimate flexural stress of the latter (about 70 MPa). This is an expected strength improvement resulted from the insertion of bamboo pins into the bamboo laminate specimen. It is also noticed that under the same applied load, the deflection of the specimen without pins is larger than that of the specimen with pin. In

607

Bamboo is a cheap and sustainable building material that is abundantly available in some countries such as Vietnam, China, and India. Due to its advantages, bamboo has been engineered and used to fabricate a variety of composite materials for building construction. The engineered bamboo composites available in the market; however, were used for making furniture, floor covering, or decorative elements and not designed to carry structural loads. This research project investigated the potential of using bamboo laminates used as structural members in construction projects. Two groups of specimens were fabricated for testing: one includes bamboo laminates with reinforcing pins and the other comprises of specimens without pins. The results of this investigation show that the strength of the bamboo laminates with reinforcing pins can be improved up to 105 MPa. In summary, if the bamboo composites are properly engineered and fabricated, the product’s structural strength would be maximized and can be used to replace wooden structural members in building construction. ACKNOWLEDGEMENTS The authors would like to acknowledge the support from the bamboo flooring manufacturer (Plyboo, Inc.) by providing bamboo materials for this experiment program.

REFERENCES Austin, R. & Ueda, K. 1972. Bamboo, New York, Weather Hill Publishing. ABS 2008. American Bamboo Society at http://www.bamboo. org/About.html (accessed on October 30, 2008) Bansal, A. & Zoolagud S.S. 2002. Bamboo Composites: Material of the Future, Journal of Bamboo and Rattan, 1(2), 119–130. (available at www.vsppub.com) Ghavami, K. 1995. Ultimate load behavior of bambooreinforced lightweight concrete beams. Cem. Concr. Compos., 17(4), 281–288. Janssen, J.J.A. 1981. Bamboo in Building Structures, Ph.D dissertation at the Department of Technical Sciences at the Eindhoven University of Technology, Netherlands.

INBR 2001. International Network for Bamboo and Rattan, at http://www.panasia.org.sg/inbar/bamboo.htm (accessed on October 12, 2008). Li, Z., Liu, C.P., & Yu, T. 2001. Laminate of Reformed Bamboo and extruded Fiber-Reinforced Cementious Plate. Journal of Materials in Civil Engineering, 14(5), October 1, ASCE, 359–365. Smith, W.S., 1987. Properties of Constituent materials, in Engineered Materials Handbook, Vol. 1, Composites, ASM International. Yao, W. & Li, Z. 2003. Flexural behavior of bamboo–fiberreinforced mortar laminates. Cement and Concrete Research 33 (2003) 15–19, Elsevier Science.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Young’s modulus of newly mixed cementitious extrusion-molded materials T. Watanabe Osaka City University, Osaka, Japan

A. Mori Nagoya Institute of Technology, Nagoya, Japan

ABSTRACT: In this paper, a method for predicting Young’s modulus of cementitious materials produced by extrusion moulding with several kinds of new fiber additives is proposed and they are compared with a standard material containing chrysotile. The volumes of the new additives were equal to that of the added chrysotile, and the variation in flexural strength was monitored by observing changes to water volume ratios. The relationship of flexural Young’s modulus to the mortar mix was identified for each new additive. The new expression for estimating the flexural Young’s modulus of these extrusion-molded materials is formed by the equations for estimating the values of dispersed particle composites. The validity of this equation was then examined. It is expected that the information gained in this research will aid designers in selecting materials and designing mixes in order to obtain a desired Young’s flexural modulus. 1

INTRODUCTION

Cementitious products formed by extrusion-molding can be designed with thin cross sections and exhibit superior flexural strength and durability to those of other cementitious products. These capabilities have brought them into wide use for exterior building panels and residential siding and other exterior construction materials. Furthermore, in a new and innovative formwork method, the thin cementitious panels have gained recent attention in Japan as more economical permanent forms which are pieces of steel-reinforced concrete components such as stay-in-place formworks. In case of use for these components, its flexural strength and Young’s modulus are critical parameters. A noted advantage of extrusion-molded materials is their generally high first-crack flexural strength, which often permits the cross section of a component to be determined by the allowed deformation. However, this requires the designer have access to a method for predicting the Young’s flexural modulus of the materials while designing a cross section. Unfortunately, there have been almost no practically oriented studies of this property. An additional complicating factor is the longstanding use of chrysotile fibers as additives in these products. These fibers been blamed for adverse effects on human health, and the industry has been anticipating the development of a practical replacement. Currently, aramid carbon polypropylene and other new fibers have gained popularity as such replacements. We have performed flexural tests on specimens

609

containing various new-material fiber additives; the volumes of these additives were the same as that of chrysotile, while the water-content ratio was varied. Thus, the authors clarified the relation between the mix proportion and the Young’s modulus when various types of new-material fibers were used. The knowledge obtained in this study is useful for determining the mix proportion design and the selection of materials for an extrusion-molded material in order to obtain a Young’s modulus suitable for the allowed deformation. Since the common cross-sectional shape of extrusion-molded materials used as construction materials resembles a hollow thin plate, flexural Young’s modulus were obtained in the present study for a thinplate-shaped cross-section so that the basic requirements for determining the flexural performances of the products can be obtained.

2 2.1

EXPERIMENTAL METHOD Materials used

The materials used in this study were ordinary Portland cement as the binding material, Toyoura silica sand used in ‘‘Testing method for strength of cement’’ of the old JIS R 5201 (1992) as the aggregate, tap water, and methylcellulose as the thickening agent. In addition to chrysotile, which is normally used in extrusion molding, other fibrous additives such as sepiolite, aramid, carbon, and polypropylene were used as new-material fibers. Tables 1 through 3 show the physical properties

Table 1.

Physical properties of ordinary Portland cement.

Compounding

Specific surface Water Setting time and Density area content Initial Finish (h-m) (h-m) (g/cm3 ) (cm2 /g) (%)

Compressive strength (28d) (MPa)

3.16

41.6

3190

28.3

2–36

3–45

Mixing Kneading

Kneader

Extrusion

Extruder

Curing

Figure 1.

Table 2. Physical properties of fine aggregates. Water Specific Specific Bulk Angle of absorp- surface gravity density repose tion area (cm2 /g) (g/cm3 ) (g/cm3 ) (degree) (%)

Kinds

Toyoura silica sand 2.66

1.56

0.59

0.6

256

Physical properties of fiber.

Kinds Chrysotile Sepiolite Aramid Carbon Polypropylene

Table 4.

Density Diam(g/cm3 ) eter

Aspect Tensile Young’s Ratio strength modulus (–) (MPa) (GPa)

2.74 2.30 1.45 1.77

7 0.2–5 μm 5 μm 7 μm

− 5–200 >500 430

0.909

0.4 mm

15

1863 – 3100 3300 490

70–140 – 110 240 11.8

Extrusion process.

Production method

The production method is shown in Figure 1. The materials were first compounded with an Omni mixer (30 L), and then kneaded using a one-axis kneader with a screw diameter of 100 mm. Then, the kneaded material, i.e., the mortar, was extruded with an extruder with a die opening having dimensions 12.0 × 60.0 mm and a barrel I.D. of 50.0 mm (die cross-sectional area/barrel cross-sectional area = drawing ratio: 36.7%), which has a vacuum chamber between the pug mill section and the barrel section. The screw rotation speed in the pug mill and barrel sections was set at a constant value of 11.0 rpm.

Mix proportion of mortars used in the experiment.

S/C

Water content ratio (–)

Watercement ratio (–)

Methylcellulose content (–)

Fiber additive content (–)

1.0

0.15–0.27

0.25–0.65

0.01

0.05

of the ordinary Portland cement, fine aggregates, and fibrous additives, respectively. 2.2

20°C in water for 28 days

the admixture, regardless of the water-content ratio and the type of fibrous additive material. The additive amount of chrysotile was set at 5 mass% of the total mortar mass, regardless of the water-content ratio, and that of all the other fibers were set at the additive amount (4 vol%) of an equal volume of chrysotile. 2.3

Table 3.

Raw materials (Cement, Fine aggregates, Water, Visco additives, Fiber additives) Omni mixer

Mix proportions

The mix proportions of the mortars are shown in Table 4. The ratio of the cement mass to the aggregate mass (sand-cement ratio: S/C hereafter) was maintained at a constant value of 1.00, and the water-content ratio (a given value of water content divided by the sum of a cement mass, an aggregate mass, and a water content) was varied in the entire extrusion-moldable range of 0.12–0.27. The additive amount of the thickening agent was set at 1% of the total mortar amount without

2.4

Flexural test method

A long extruded compact with a cross-section of 12.0 × 60.0 mm and a length of 550 mm was covered with a vinyl sheet so that water did not dissipate and was seal-cured for one day in a room maintained at 20◦ C. The compact was additionally cured for 27 days in water maintained at 20◦ C. The cured compact was cut into 150-mm-long specimens for the flexural strength measurement, and the excess water on their surfaces was wiped with a waste cloth. The specimens with dried surfaces were subject to flexural strength testing. The flexural test was performed with a span of 120 mm, displacement control at 1.0 mm/min, and loading at the exact center of the span. The extrusion-molded materials are generally considered to be anisotropic, and the measured flexural strengths and deflections at the loading points are in the direction of the extrusion molding. Flexural Young’s modulus were obtained from the secant lines for 1/3 stresses of flexural strengths. The flexural Young’s modulus were obtained from the slopes of the secant lines for 1/3 of the flexural strengths.

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Measurement of the mix proportions and densities during the molding process

The surface-dry and absolute-dry densities were experimentally measured, and the water-cement ratios and air contents were calculated by those densities and mix proportions. The surface-dry density of each mortar was obtained from the mix proportions and physical properties of each material by assuming that no air was present in the material. The water-cement ratio at this point is described as the designed water-cement ratio. The surface-dry densities of the compacts before cement setting and immediately after extrusion molding were obtained from the measurements of the masses and volumes. In addition, after the compacts had been dried using a dryer with an agitator maintained at 105◦ C until the masses became constant, the masses were measured again and the absolute-dry densities were obtained. The water-content ratios in the mortars immediately after the molding were calculated from differences between the above mentioned two masses. The water-cement ratio obtained based on the water-content ratio at this point is termed as the water-cement ratio after molding. The air content was obtained from the difference between the surface-dry density from the mix proportions and that after molding. Furthermore, the water-cement ratio obtained from the sum of (1) air in the compact converted to an equivalent volume of water and (2) the actual amount of water mixed was regarded as the effective value, and its reciprocal is termed as the effective cement-water ratio. Immediately prior to the flexural test, the surfacedry density was obtained by measuring the mass and volume of each specimen.

3

EXPERIMENTAL RESULTS

which are dependent on the type of fibrous additive material, are lower than the designed water-content ratios by 0–4%. This was caused by the evaporation of water in the mortars during kneading. The variation in the water-content ratios due to differences in the types of fibers is very small in this case. Figure 3 shows the relation between the designed water-cement ratio and the effective water-cement ratio after extrusion. The effective water-cement ratios are larger than the designed water-cement ratios in one case and smaller in the other case. In the case of using aramid, larger values of the effective water-cement ratio were obtained because air remained in the sample even after vacuum deaeration during molding. The average air content was 2.3% in this case. When sepiolite was used, the water-cement ratios were smaller due to the partial dissipation of water during the molding process and more importantly, the efficient elimination of air by vacuum deaeration. These observations indicate that the mixed air content is dependent on the type of fibrous material. 3.2

Relation between the water-cement ratio and the Young’s modulus

Figure 4 shows the relation between the designed water-cement ratio and the Young’s modulus. As shown in the figure, the cementitious composite materials show a wide range of Young’s modulus, 28–5 GPa, in the designed water-cement ratio range between 0.25 0.8

Effective water cement ratio

2.5

0.6

0.4

0

3.1 Changes in the mix proportion

0

Figure 2 shows the relation between the designed water-content ratio and the water-content ratio after molding. The water-content ratios after molding,

0.2

0.4

0.6

0.8

Designed water cement ratio (−)

Figure 3. Relationships between designed water cement ratio and effective one. 35

30 Chrysotile Sepiolite Carbon Aramid Polypropylene

25 20

Young’s modulus (GPa)

Water content ratio after extrusion moulding (%)

Chrysotile Sepiolite Aramid Carbon Polypropylene

0.2

15 10 5

Sepiolite Chrysotile Carbon Aramid Polypropylene

30 25 20 15 10 5 0

0 0

5

10

15

20

25

30

0

Designed water contentratio (%)

0.2

0.4

0.6

0.8

1.0

Designed water-cement ratio (−)

Figure 2. Relationships between designed water content ratio and that after extrusion moulding.

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Figure 4. Relationships between designed water cement ratio and Young’s modulus.

Young’s modulus (GPa)

30 25 20 15

Chrysotile Sepiolite Aramid Carbon Polypropylene

Y = 0.564 * D4.33

10 5 0 0

0.5

1

1.5

2

2.5

3

Surface dry density D (g/cm3)

Figure 7. Relationship between surface dry density & Young’s modulus. Flexural strength (MPa)

and 0.65. All the fibrous additive materials show a uniform decrease in the Young’s modulus with an increase in this range of the designed water-cement ratio. The Young’s modulus for same designed watercement ratios significantly differ with the type of the fibrous additive material and decrease in the order sepiolite > polypropylene and chrysotile > aramid > carbon in terms of the magnitude of Young’s modulus. The relation between the designed cement-water ratio and the Young’s modulus, and that between the effective cement-water ratio and the Young’s modulus are shown in Figures 5 and 6, respectively. In either case, for a given cement-fine aggregate ratio, the Young’s modulus shows an approximately proportional increase with the cement-water ratio. However, for the same cement-water ratio, the difference in the Young’s modulus is smaller for the plot of the Young modulus against the effective cement-water ratio than for the plot of the Young’s modulus against the designed cement-water ratio.

30 Chrysotile

25

Sepiolite Aramid

20

Carbon Polypropylene

15 10 5 0 0.0

1.0

2.0

3.0

4.0

5.0

Effective cement water ratio ( )

3.3 Relation between the surface-dry density and the Young’s modulus Figure 7 shows the relation between the surfacedry density and the flexural Young’s modulus of the specimens. All the materials show an increase in the Young’s modulus with the surface-dry density, and the increase rate is also directly proportional to the surface-dry density. In addition, the relation between the Young’s modulus and the surface-dry density, as shown by an equation in the figure, can be expressed

Figure 8. Relationship between effective cement-water ratio and flexural strength.

by an approximately same regression curve even for different fibrous materials. This indicates that the Young’s modulus of an extrusion-molded material can be roughly predicted from its surface-dry density when it is measured.

Young s modulus (GPa)

3.4 35 Chrysotile Sepiolite Aramid Carbon Polypropylene

30 25 20 15 10 5 0 0

1

2

3

4

5

Figure 5. Relationships between designed cement-water ratio and Young’s modulus. Young s modulus (GPa)

35 30 25

Relation between the effective cement-water ratio and the flexural strength

Figure 8 shows the relation between the effective cement-water ratio and the flexural strength. In the range of effective cement-water ratios used in the present experiment, the flexural strength increases uniformly with the effective cement-water ratio, regardless of the type of fibrous additive material. However, the increase rate tends to be small for the effective cement-water ratios beyond approximately 3.0. Based on the difference in the type of the fibrous additive material, the flexural strengths decrease in the order sepiolite and chrysotile > polypropylene and aramid > carbon.

Chrysotile Sepiolite Aramid Carbon Polypropylene

20

4

DISCUSSIONS

15 10

4.1

5 00

1

2

3

4

5

Effective cement water ratio (–)

Figure 6. Relationships between effective cement-water ratio and Young’s modulus.

Prediction of the Young’s modulus of matrices of composite materials

For the design of the Young’s modulus of cementitious materials produced by extrusion molding, it is convenient to treat them as particle-dispersed systems of composite materials and to obtain the Young’s modulus

612

of the composite materials from the Young’s modulus and the composition ratios of the particles and matrices. Meanwhile, regarding the Young’s modulus of formwork-molded concrete, the effectiveness of the Hashin-Hansen equation shown by Eq. (1) proposed for the self-consistent approximation model has been confirmed (Hansen 1965; Kishitani & Baba 1975).

relation can be expressed by T.C. Powers’ equation (Powers 1948) shown in Eq. 2. The cavity ratio here is the sum of the gel cavity, capillary cavity, and cavity due to air entrainment, and can be calculated by Eq. 3. This equation was provided on the assumption that when 100 g of cement completely hydrates, it reacts with 22.7 g of water, and the water volume reduces to 0.75 times that of the initial value. The degree of hydration of cement was experimentally obtained on the 28th day depending on the respective water-cement ratio, in accordance with JIS R 5202 (1995).

Ec /Em = [n + 1 + (n − 1) · Va ] (1)

where, Ec : Young’s modulus of a composite material (the matrix and the aggregate), Em : Young’s modulus of the matrix (cement paste and admixtures) of the composite material, n: ratio of the Young’s modulus of the aggregate to that of the matrix, and Va : volume concentration ratio of the aggregate. Therefore, the experimental results Ec and Eq. 1 were utilized in plotting the Young’s modulus Em of the matrices as a function of the effective cementwater ratios, as shown in Figure 9. The Young’s modulus of Toyoura silica sand used in this study is set at 77.5 GPa. The Young’s modulus Em for fibers other than carbon were approximately the same for the same effective cement-water ratio. A cross-sectional observation of a carbon-containing composite after molding revealed the presence of fiber balls, indicating a failure in uniform mixing. Figure 10 shows the relation between the surfacedry densities of the matrices and the Young’s modulus Em . The surface-dry densities are obtained from the mix proportions. The surface-dry densities of the matrices can be expressed by an approximately same regression curve, as shown by an equation in the figure, regardless of the type of fibrous additive material, similar to the case of the extrusion-molded materials.

Ep = E0 (1 − Vp )3

(2)

Vp = (We − 0.170α)/(We + 0.316)

(3)

where Ep : Young’s modulus of cement paste, Vp : cavity ratio of cement paste, E0 : Young’s modulus for a zero-cavity ratio, We : effective water-cement ratio, α: degree of hydration of the cement. Figure 11 shows the calculation result. The calculated Young’s modulus Em of the matrices for any fibrous additive material are approximately proportional to (1 − Vp )3 . The influence of the difference in the type of fibrous additive material on the Young’s modulus for the same (1 − Vp ) is small, and the relation can be expressed by an approximately same regression curve. The value of E0 , which is an experimental constant, is 46.8 GPa obtained through calculations by a least-square method. Although the

4.2 Relation between the Young’s modulus and the cavity ratio of matrices of composite materials

35 30

Young s modulus (GPa)

/[n + 1 − (n − 1) · Va ]

15 10 5 0

Young’s modulus (GPa)

Young s modulus (GPa)

8 6 4 2 0 1

2 3 4 Effective cement-water ratio ( )

1 1.5 2 Surface dry density D (g/cm3)

2.5

3

35

Chrysotile Sepiolite Aramid Carbon Polypropylene

0

0.5

Figure 10. Relationship between surface dry density and Young’s modulus Em of matrices.

14

10

Em=1.37 D2.84

20

0

The Young’s modulus of a normal cement paste is well known to be closely related to its cavity ratio, and the

12

Chrysotile Sepiolite Aramid Carbon Polypropylene

25

Chrysotile Sepiolite Aramid Carbon Polypropylene

30 25 20 15 10 5 0 0

5

Figure 9. Relationship between effective cement-water ratio and Young’s modulus Em of matrices.

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VP )3

Em

0.2

0.4

0.6

0.8

1.0

Figure 11. Relationship between (1 − Vp ) and Young’s modulus Em of matrices.

application of Eq. 3, which was experimentally obtained for a cement paste without admixtures is, strictly speaking, considered to be inappropriate because fibrous additive materials and thickening agents are commonly included in cementitious extrusion-molded materials, it was confirmed that the equation can be applied to extrusion-molded materials by using the Young’s modulus for a zero-cavity ratio obtained in the present experiment. Therefore, it can be concluded that the Young’s modulus of a cementitious extrusionmolded composite material can be obtained from the law expressed by Eq. 1 on the matrix and the aggregate of the Young’s modulus of a particle-dispersed system of composite materials and the law expressed by Eq. 2 on the Young’s modulus of cement paste.

are first obtained and introduced in Eq. 3. The value of Vp obtained here is substituted in Eq. 2, and Ep is obtained, regardless of the type of fibrous additive material. Then, the Young’s modulus of an extrusionmolded material can be obtained from Eq. 1, the Young’s modulus (Ea ) of the aggregate, and volume (Va ) of the aggregate. Figure 13 shows the relation between the experimentally obtained Young’s modulus of the extrusion-molded materials and the values calculated by the proposed method. The two values are approximately the same, and therefore, the present calculation method can predict Young’s modulus with a good accuracy.

5 4.3

Proposal of a prediction method for Young’s modulus

Figure 12 shows the calculation flow for the prediction of a Young’s modulus. For obtaining the cavity ratio of a matrix, the predicted values of α and We Start Effective water-cement ratio and degree of hydration 1. Predicting Vp E0 2. Calculating Young s modulus of matrices by Cube Law Em = E 0 ⋅ Vp 3 Ea (Young s modulus of aggregates) and volumetric ratio (Va) 3. Calculating Young s modulus (Ec) of cementitious materials produced by extrusion moulding by Hashin-Hansen equation End

Figure 12. Method of predicting Young’s modulus of cementitious materials produced by extrusion moulding.

CONCLUSIONS

Flexural tests were performed with experimental factors depending on the type of fibrous additive material and mix proportions for the clarification of the relation between the Young’s modulus and the material and that between the Young’s modulus and the mix proportion of new-material-fiber-mixed cementitious extrusion-molded materials. The following information was obtained, and a prediction method for the Young’s modulus of extrusion-molded materials was proposed based on this information. 1. The Young’s modulus are strongly related to the surface-dry densities. 2. The relation between the flexural strength and the Young’s modulus can be evaluated by an approximately unique correlation in the range employed in the present experiment. However, the relation sometimes varies with the type of fibrous additive material. 3. The cube law of T.C. Powers can be applied to the Young’s modulus of matrices as a basis for the prediction of the Young’s modulus of a particledispersed system of composite materials, and the relation can be uniquely defined, regardless of the type of fibrous additive material.

Calcurated value (GPa)

30 Chrysotile Sepiolite Aramid Carbon Polypropylene

25

REFERENCES

20 15 10 5 0 0

5

10

15

20

25

30

Experimental value (GPa)

Figure 13. Relationship between experimental values of Young’s modulus of cementitious materials produced by extrusion moulding and calculated ones.

Hansen, T.C. 1965. Influence of Aggregate and Voids on Modulus of Elasticity of Concrete, Cement Mortar, and Cement Past, Journal of ACI, pp. 193–216. JIS R 5201. 1992. Physical testing methods for cement, Japanese Standards Association. JIS R 5202. 1995. Method for chemical analysis of Portland cement, Japanese Standards Association. Kishitani, K. & Baba, A. 1975. Drying shrinkage mechanism of building materials, Cement and Concrete, No. 346, 30–39. Powers, T.C. & Brownyard, T.L. 1948. Studies of the physical properties of hardened Portland cement pastes, J. of ACI.

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Construction methods

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Active pier underpinning of Jin-bin light rail bridge in Tianjin J. Bu School of civil Engineering, Shijiazhuang Railway Institute, Hebei, China

N. Sun China Academy of Railway Sciences, China

S. Huang Tianjin Municipal Engineering Design and Research Institute, Tianjin, China

ABSTRACT: The eight-lane Central Road is in Jin-bin new area of Tianjin city, two ways traffic, and will pass under Jin-bin light rail bridge as a tunnel. Two piers of the bridge will be replaced by a new hyperstatic system, made of a PRC lattice girder (including two beams in longitudinal direction and five beams in transverse direction) and nine piles (distributed in three rows). This underpinning engineering is the key of the Central Road project. The underpinning load is 1400 tons, and the lattice girder weighs 1850 tons. The displacement of the two replaced piers and the adjacent two must be controlled strictly in the construction because the light rail bridge can’t be strengthened and the train can’t be off-the-line. In order to ensure that the trains run safely on the Jin-bin light rail, besides limiting the speed of the train, we designed a plan to monitoring many aspects in the construction process on real time. 1

PROJECT OVERVIEW

The eight-lane Central Road (Fig. 1) is in Jin-bin new area of Tianjin city, two ways traffic, and will pass under Jin-bin light rail bridge as a tunnel. In order to make the new constructed road has good performance, two piers (A339 and A340) of the light rail bridge need to be replaced, the light rail bridge can’t be strengthened and the train can’t be off-the-line. A339 and A340 both are located at transition curve region with 0.6 percent longitudinal. The left curve radius is 446.16 m, curve length 282.26 m, transition curve length 130 m, the right curve radius is 450 m, curve length 273.57 m, transition curve length 120 m. The track is non-fragments track, the welded rail, with elastic support block, lock temperature 27◦ , CHN60 rail. The train moving on the light rail makes up of two motor cars and two trailers, maximum wheel load is 12 t, and the peak speed is 80 km/h. A339 and A340 are brake pier and connecting pier of the continuous beam bridge, respectively. The underpinning load of the two piers is about 1400 t, considering temperature stress and off-track power, etc. The cross section of the replaced piers of the light rail bridge is 2.2 × 1.4 m, the sizes of bearing platform are 6.2 × 5.6 × 2.0 m. Each bearing platform has 8 bored piles, and their diameter and length are 0.8 m and 44 m, respectively. The PRC lattice girder is composed of includes two beams in longitudinal direction and five beams

Figure 1.

Central road before construction.

in transverse direction, considering its lateral stability and weight. The cross sections and span length of the longitudinal beams are 1.8 m × 2.0 m and 2 × 18.75 m. The cross sections of the transverse beams are 3.0 m × 2.0 m, for 1# and 3# , 3.5 m × 2.0 m for 2# , 3.5 m × 2.0 m for 4# –5# . The roof and bottom board thickness is 0.3 cm. The span length of the transverse beams is 8.4 m. Figure 2 gives the details of the underpinning structure system. The lattice girder itself weighs 1850 tons, and its concrete type is C60. The nine new piles are bored piles, the diameter and length are, for 1# –3# and 7# –9# , 1.8 m and 70 m, for 4# –6# , 2.0 m and 90 m, respectively, and the concrete type is C35. Because the soft soil is too deep, the piles cannot be embedded in bedrock, so they all are friction piles.

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Light rail bridge

A339 Pier

A340 Pier

5# beam #

1 beam Lattice girder

1# New piles 2#

Figure 2.

4# beam

2# beam

3# beam 7#

4# #

The cut piles 5 # 6 and bearing 3# platform

8#

9#

Underpinning structure system.

Figure 3.

Lifting and setting the reinforcement cage.

Figure 4.

Cutting grooves.

Figure 5.

Planting steel bar in the pier.

The PRC lattice girder and the 9 new piles formed a complex underpinning structure system to replace the adjacent two piers. Design criterion scheduled time of the project is 100 years. Seismic peak ground acceleration is 0.15 g, and the structure security rating is first-level.

2

CONSTRUCTION TECHNOLOGIES

The flow of the active replacement construction include seven steps, constructing the nine new piles, constructing the two longitudinal beams and 1# –3# transverse beams of the lattice girder on the full hall holder, constructing 4# –5# transverse beams after jointing with the two replaced piers, hoisting jack (arranged on the pile top) loading, cutting the two replaced piers and transforming the load from piers to the new piles, digging foundation pit, constructing tunnel box, adjusting hoisting jack and mixing the lattice girder and nine piles together. The key processes are explanted as follows. 2.1 New piles construction There is only about 7 meters space for piles construction. Based on the experience of test piles, we made use of the forward cycle drilling machine to drill holes, made use of air-lift reverse circulation to clear holes in the new piles construction. Each section of reinforcement cage is about 4 m long (Fig. 3), and the reinforcement cage connected by straight thread. 2.2 Constructing the jointed point between pier and beam Grooves were cut around the two replaced piers where the region joined with 4# and 5# beams, the distance between grooves is 15 cm, depth is 25 mm (Fig. 4). The steel bars, their diameter is 32 mm, were planted in the old concrete of the piers deep in 0.5 cm (Fig. 5). The distance between steel bars is 15 cm. Bidirectional

pre-stress were imposed on the joined point. The longitudinal tendons are strands, and the transverse is precision rolling twisted bar. 2.3

Constructing the transverse restraints of the new piles

In order to prevent the lattice girder to move along the transverse and longitudinal direction under the train dynamic load and others, we set transverse restraints on the top of all nine new piles. The transverse restraints are made of steel tube(diameter 0.21 mm, thickness 10 mm) set on the bottom of the lattice girder and concrete-filled steel tube columns (steel tube diameter 0.21 mm, thickness 10 mm, concrete type C50) set on the top of the new piles, detailed in

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Figures 6 and 7. We set reinforced mesh in concrete corresponding to two ends of the transverse restraints to prevent the stress concentration. 2.4

Hoisting jack loading

Three 200 t hoisting jacks as a group were placed on top of the piles with 1.8 m diameter, including angle 120◦ , and five 200 t hoisting jacks as a group on top of the piles with diameter 2.0 m, including angle of 72◦ . Hydraulic system of the hoisting jacks is controlled by computer (Fig. 8). Then, we can complete automatic displacement synchronization, control force and displacement, realize operation atresia in the process of loading. For the purpose of making the piles to complete part of the settlement, the upper underpinning load was divided into 10 grades (20%, 40%, 50%, 60%, 70%, 80%, 85%, 90%, 95% and 100%), and each graded load is distributed to nine piles according to the simulated results (Yin & Ke 2008) from finite element simulation analysis. When hoisting jack loading, the information of the underpinning structure system and the light rail bridge, such as stress of the lattice girder, settlement of the new piles and displacement of the replaced piers top, etc., can be obtained and showed by the real-time monitor system. And the information guide when and how to load the next grade.

1 Hoisting jack; 2 Electron Dial gauge; 3 Cut-off valve; 4 Stress sensor; 5 Intake pipe; 6 Oil return pipe; 7 Signal Line; 8 System Terminal (Siemens S7-300); 9 Divider; 10 Cylinder; 11 Hydraulic pressure pumping station;12 Computer.

Figure 8.

Control principle of the hoisting jack lifting.

Figure 9.

Preparing cutting the piers.

2.5

Figure 6.

Settings on the top of new piles.

Figure 7.

Bottom template of the beams above the new piles.

Cutting the replaced piers

The two replaced piers were cut by two sets of highintensity diamond chainsaw, typed LP/TS32, at the same time. The parameters of the cutting equipment are described as following, power 35 kW, cutting speed 1.0 m2 /h, rotational speed 1500∼2800 turns/min, driving pressure 26 Mpa. Unloading about 15% of the upper underpinning load by operating the hoisting jack system, it is ensure that there is smaller compressive stress in the cut region. If the cut region is in tension, the pier will move up after cutting and can not be controlled. Of course, the ideal cut-off state is zero stress, but it cannot be realized in the practice. The cutting sections keep space 15 cm to the bottom of 4# and 5# beams of the lattice girder. We set twelve hoisting jacks around the two replaced piers to prevent the upper structure to drop down suddenly (Figs. 9–10). The two replaced piers were cut keeping pace, the monitoring system worked at the same time, and the emergency plan were made active. After cutting, the twelve hoisting jacks around the two replaced piers unloaded and the thirty-three hoisting jacks on the new piles top loaded at the same time, realized the load transformation (Figs. 11–12).

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After the underpinning load transformation, the tunnel below the bridge was constructed by cut-andcover method. 2.6

Figure 10.

Cutting the replaced piers.

Figure 11.

The replaced pier after cutting.

Figure 12.

After the load transformation.

Mixing the piles and beams

When the pile settlement reached steady-state (the settlement is less than 0.1 mm for three consecutive months), according the local temperature changes throughout the year we chose an appropriate temperature after analyzing the latter temperature stress to mix the new piles and the corresponding beams. After strength, stability, and compression deformation calculation, we decided to adopt 33 steel columns, type Q345, diameter 0.2 m, replace the 33 jacks by one-to-one (Fig. 13). The detailed mixing construction can be divided six steps, and they are installing the steel columns, operating the hoisting jack system to adjust the elevation of the lattice girder and the two replaced piers, welding the transverse restraints, connecting the steel columns on the same pile as a whole, connecting the reserved steel bar (Fig. 14), setting up the template and irrigating concrete. The vibration-free and micro-expansive concrete was used in the concrete construction, and its type was C40 (Figs. 15–16).

Figure 14.

Figure 13. column.

Connecting the reserved steel bar.

Hoisting jack have been replaced by steel Figure 15. Setting up the template and irrigating concrete.

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3.2

Figure 16.

The new-constructed concrete in maintenance.

Figure 17.

The central road after opening.

Monitor contents include stress of the rail bottom induced passing train and underpinning construction, safety performance of the passing train (train axle lateral force, train derailment factor and reduction rate of train wheel load), elevation changes of the three bridge spans affected by construction and the vertical and horizontal dynamic acceleration and amplitude at their middle span point, the vertical, horizontal and longitudinal displacements, and horizontal and longitudinal angle of the two replaced piers and the adjacent two, the elevation changes and stress of the lattice girder, the vertical displacement and stress of the new piles, and the stress and relative slip of the pier-girder connection point, etc. Parts of the monitor contents are control indexes in the construction, such as the vertical displacement of the new piles and the replaced two piers, etc, and the other parts are used in the analysis. There are 347 measurement points in the monitoring plan, and they constitute multi-level, all-round monitoring system, combining static and dynamic, analysis and control indexes, manual measurement and automatic transmission acquisition, real-time testing (Fig. 20) and regular testing together. 3.3

Figure 17 shows the central road for opening, passing under the light railway bridge after underpinning construction.

3 3.1

Monitor contents

Monitor implementation and results

From the beginning to the end of the underpinning construction, the monitoring had been carried out almost 11 months and it provided reliable information for the

MONITORING SYSTEM Wireless data transmission system

For ease of the long-term monitor, we decided to use the wireless data transmission system. Making use of the wireless data transmission system not only may facilitate real-time monitor, reduce the electric cable and make the constructing disturbance to be small. Moreover, the system maintenance is convenient also. The data wireless transmission system is composed of certain wireless sensing unit, and the wireless sensing unit is a circuit board connecting a group of sensors. The wireless sensing unit has realized the data acquisition, the memory and the wireless transmission integrated design. The power supplied to the circuit board by the solar energy charge type 12 volts lithium battery group system. This is not only may reduce to the outside power source’s dependence, but also advantageous to the environmental protection. The circuit board is 9 cm × 6 cm. The Wireless data transmission system, wireless sensing unit and solar panel are showed in Figures 18 and 19, respectively.

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Figure 18.

Wireless data transmission system schematic.

Figure 19.

Wireless sensing unit and solar panel.

limited space. From Table 1, we can clearly find that the train operation is normal, and the status of the underpinning structure system is safe under the underpinning construction, and the listed results agree the calculated ones (Yin & Ke 2008) well. 4

Figure 20.

This project is a new sample in civil engineering, and its successful implementation will provide more wide views for the design of transport facilities. The following observations can be concluded from above sections:

Real-time monitoring system interface.

1. Two piers of light rail bridge ware replaced by a new hyperstatic system, made of a lattice girder (including two beams in longitudinal direction and five beams in transverse direction) and nine new piles (distributed in three rows). This may provided a good example for future underpinning project. 2. At the joint point between pier and beam, using cutting groove, planting steel bar in the old concrete and bidirectional pre-stress technologies can guarantee the effect of load transmission. 3. The implementation of real-time monitoring is very important in the key construction process, since the structure system is time-varying in the construction (Bu, et al. 2009). 4. Mixing the new piles and beams must choice a suitable temperature (considering the temperature effect in the hyperstatic system) and a series reasonable approach. 5. The careful design, the creative construction methods, the monitor in the entire process and the close co-operation of all aspects are the advantageous guarantee for the project success implementation.

Table 1. Measured maximum values of the control indexes obtained in the construction process.

No. 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16

Monitoring contents Stress of the rail bottom Train axle lateral force Train derailment factor Reduction rate of train wheel load Vertical displacement of the replaced piers top Vertical displacement difference of the replaced piers top Longitudinal angle of the replaced piers top Transverse angle of the replaced piers top Transverse displacement of the piers top Longitudinal displacement of the piers top Press stress of the lattice girder Elevation difference of the lattice girder Settlement of the new pile Settlement difference of the new pile Stress of the new pile Relative slip of the pierbeam connection point**

Measured values Max.

Warning value*

55.44 MPa 20.73 MPa 0.31

363 MPa 32 MPa 0.70

0.42

0.60

1.15 mm

3.0 mm

0.7 mm

3.0 mm

55

3

25

3

0.59 mm

5 mm

3.16 mm

10 mm

−14.94 MPa

−22 MPa

3.58 mm

5 mm

2.26 mm

/

1.5 mm −2.04 MPa

2 mm −13 MPa

0 mm

0.5 mm

* The warning value can be found in reference Yang et al. (2008). ** Not considering the elastic deformation.

construction decision-making. The measured maximum values of the control indexes obtained in the construction process can be found in Table 1. Only the main indexes are listed in the table because of the

CONCLUSIONS

ACKNOWLEDGEMENT The Project Supported by National Natural Science Foundation of China (No. 50878134) and Natural Science Foundation of Hebei Province (No. E2006000394). REFERENCES Jianqing Bu, Ning Sun and Yiqing Guo. (2009) Analysis on time-varying Process of active pier underpinning structure system for Jinbin light railway. Railway Standard Design, In press, in Chinese. Fuzeng Yang, Jianqing Bu, Zaitian Ke, et al. (2008) Informationization construction of Jinbin light railway pier active underpinning in Tianjin. Urban Roads Bridges & Flood Control, (7):82–86, in Chinese. Jin Yin and Zaitian Ke. (2008) Displacement control of hyperstatic structure system of active pile foundation underpinning. Railway Engineering, (11):24–26, in Chinese.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

CFRP liner quality control for repair of prestressed concrete cylinder pipe A. Allan Fibrwrap Construction Inc., Oswego, Illinois, USA

H. Carr Fibrwrap Construction Inc., Ontario, California, USA

ABSTRACT: Presented in the paper are the most recent methods developed for the safe and reliable installation of carbon fiber liners. Discussion of mobilization concerns, material requirements and related verification testing, critical paths for installation, and inspection methods for both during and post installation are provided. The installation of carbon fiber liners has many critical steps. Development of a usable quality control standard for material selection, design, installation, and inspection is critical for a successful installation. 1

OVERVIEW

The use of fiber reinforced polymers (FRPs) to strengthen concrete dates back to the early 1980’s. There are now thousands of successfully completed projects around the world. Over 200 million square feet of FRPs have been installed by various installers around the globe. The applications are varied and the markets are numerous: 1. 2. 3. 4. 5. 6.

Transportation Commercial Industrial Waterfront Structures Water Transmission and Tanks Blast Protection

Elements strengthened with FRPs include slabs, beams, columns, pier caps, walls, piles and pipes. The substrates can be either concrete, masonry, brick or wood. Currently research is being done to bond FRPs to strengthen steel. Work in area of FRPs began in the US in the 1980’s with studies conducted by the University of California San Diego for the Department of California Transportation (Caltrans) for the purpose of retrofitting highway bridge columns. Caltrans was looking for a more efficient method to increase the ductility of their bridge columns to allow them to survive a seismic event. The traditional technology involved installing a metal can around the columns and injecting the annulus with a cementatious grout. This solution was not practical and changed the aesthetics of the structures. After various tests and after surviving several earthquakes, the technology evolved and gained acceptance. The applications then moved from the

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transportation market to the general building industry following the guidelines of American Concrete Institute (ACI 2002). During the mid 1990’s this technology entered the pipeline industry for the retrofit of Prestressed Concrete Cylinder Pipe (PCCP) with broken wires. Back in the early days some water transmission lines were built using PCCP. These cylinder pipes consisted of concrete pipes that were manufactured using prestressing wires that imparted compression on the concrete and resisted the internal water pressures. We are well aware of the potential for corrosion when using reinforced concrete. Some interesting statistics regarding the use of PCCP: 1. Almost 30 million meters of PCCP was manufactured between 1940 and 2006. 2. A database of PCCP failures contains 592 independent entries, representing three categories of failure in 35 states and the District of Columbia. 3. Within 50 years of installation, one rupture and 66 other failures for every 70,000 meters of pipe. 4. A significant increased failure rate occurred for pipe installed between 1971 and 1979 because of the type of prestresssing wire used during this period. 5. Half of the recorded catastrophic leaks and breaks involved pipe manufactured between 1971 and 1979. Therefore, a huge potential was recognized in this water transmission market. However there were no standards or specifications for this new application and no quality control standard directly related to this application was available. Initial methods drew from the column wrapping methods developed by Caltrans, International Code Council (ICC) and those outlined in ICC AC 125. However, as this application became

more popular both the Fyfe and Fibrwrap Construction companies made large advancements in materials, application, and quality control methods. In the year 2000, the Metropolitan Water District of Southern California (MWD) took a serious interest in carbon fiber liners for the retrofit of its 263 kilometers of PCCP and implemented a testing and review program for CFRP (carbon fiber reinforced polymer). 2

QUALITY CONTROL DOCUMENTS

As the development of FRP’s for use in repair and strengthening of structures advanced in the early 1990’s, there became an industry requirement for quality control and quality assurance standards and techniques. This requirement resulted in many studies and collaborations between manufacturers, universities, building code approval authorities and public agencies to create inspection and testing criteria. At the forefront of this was the California Department of Transportation, who performed the initial testing and qualifications of various FRP systems. Caltrans developed and set forth criteria for manufacturers to qualify their FRP systems for use on public Department of Transportation projects, specifically to retrofit bridge columns for seismic deficiencies. These criteria included large scale structural testing of columns, durability testing of the materials, requirements for manufacturer quality control procedures and requirements for inspection of the installation procedure. The cooperative research and criteria set forth by Caltrans led the way for other approval agencies to establish quality control and quality assurance documents. Criteria were established by ICC AC 125 and AC 178 (International Code Council, 2003) inspection criteria, and ACI (the American Concrete Institute) authored the ACI 440 guideline. There were various other criteria and documents established for the Quality Assurance (QA) and Quality Control (QC) of FRP systems; however these procedures were slow to be adopted and standardized which led to project specific requirements, typically developed by the Engineer of Record in conjunction with manufacturer specifications. In the mid to late 1990’s research was performed to strengthen PCCP by applying FRP internally to resist both internal pressure and external loading conditions due to deficiencies caused by the failure of the pre-stressing wire. Commercial applications soon followed, and once again, the FRP industry was in need of a Quality Control/Quality Assurance standard for the strengthening of PCCP and other pipeline structures with FRP’s. Some of the early QA/QC and other FRP approval documents, such as ICC AC125, ACI 440 and Caltrans, were utilized to pre select and approve manufacturers’ systems and set forth project specific quality control documents. This approach fell short of

an industry standard and many agencies were skeptical of using FRP’s for lack of QA/QC standards, performance history and proof testing. MWD implemented a research program to address these issues in 2000. The program consisted of large scale structural testing, durability testing, manufacturer approval and qualification, quality control/quality assurance standards, inspection criteria and training and FRP design considerations. This program was completed in late 2002 and MWD began to use FRP to strengthen a portion of their PCCP system. This Pre-Qualification, QA/QC program (MWD, 2004) has proven to be very thorough and has since been adopted by other agencies and FRP installation companies. 3

INSTALLATION PROCEDURE

The field installation of FRP composite systems must be performed correctly in order to make certain that the goals the engineer has designed for are met. In the pipe, the surfaces are prepared for bonding by means of abrasive blasting using grit or a water-blasting system to achieve 1.6 mm minimum amplitude. All contact surfaces are then cleaned and dried in preparation for the composite to be applied. Using a roller, one prime coat of the manufacturer’s epoxy is applied and allowed to cure for a minimum of one hour. Any uneven surfaces left from the blasting are filled in with the manufacturer’s thickened epoxy using a trowel. The epoxy matrix is prepared by combining components at a ratio specified by the system manufacturer. The components of epoxy are mixed with a mechanical mixer until uniformly mixed. The dry fabric is saturated with the saturate epoxy using a saturator machine. The machine monitors the epoxy to fiber ratio. The saturated fiber is layed up by hand inside the pipe. The FRP applied to the wall of the pipe is troweled smooth to remove any air or excess epoxy behind the fabric. Thickened epoxy is troweled on in between layers to ensure strong interlaminate bond. The number of layers applied, butt splice width, end joint details, and overlaps are all done according to the project shop drawings. All seams and edges are coved with thickened epoxy. After proper cure times are met, the final epoxy coat is applied by roller or trowel and then cured. 4

INSPECTION METHODS

The Metropolitan Water District of Southern California together with Fibrwrap Construction, Inc. developed a standard for field inspection and quality control on FRP strengthening for pipes. The methods of inspection appear in all of MWD’s standard specifications for FRP. Each step in the installation process has a check to ensure that it was performed correctly.

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In order to keep track of materials used, installed, and submitted for testing, the Contractor is required to keep and maintain a quality control log. In this log the contractor will record the lot numbers of the fabric and epoxies used on the project, calculated fabric to epoxy weight ratios, test samples made, and locations of the material (i.e. lot numbers corresponding to layer number installed in the pipe). This data should be recorded in the log book every shift and should include dates and times respectively. This log is the contractor’s means to re-create the quality control measures taken in case there is any dispute. The owner’s inspector should keep a similar log so that any inconsistencies can be addressed. The first onsite test to be performed will address the environment inside the pipe where the layup is to be performed. The surface temperature of the substrate to which the FRP is to be applied must fall between 1.5 and 38 degrees Celsius. If the surface temperature falls outside of this range, then corrective measures should be taken to control the atmosphere inside the pipe. Surface preparation is critical to the performance of the FRP as a strengthening material. It is important to ensure a proper surface preparation so that the FRP material will bond to the substrate and transfer the loads properly through the bond interface as designed for in the engineering stages of the project. The American Society for Testing and Materials (ASTM) D4541 Standard Test Method for Pull-Off Strength of Coatings Using Portable Adhesion Testers cannot be performed in a pipe because there is no flat surface to perform the test on. Therefore the inspection of the surface preparation must be visual. The inspector is looking for a roughened surface with minimum 1.6 mm amplitude. A visual clue that the surface preparation is completed accordingly is the absence of any smooth surface or other laitance still remaining. In most cases the surface preparation will expose aggregate of the concrete lining. Figure 1 shows a concretesurface that has an acceptable surface preparation next to a concrete surface that has not yet been prepared. The surface must also be free from fins or sharp edges that could create voids behind the fiber.

Figure 1.

Surface preparation to pipe wall.

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Following the surface preparation, the substrate must be dry before the prime coat of epoxy is applied. If the prime coat of epoxy is specially formulated for a wet surface prime, then the substrate may still be damp when applied. If the epoxy has not been formulated this way, then proper measures should be taken to ensure that the substrate is dry before application. The simplest, yet still effective, way to test for moisture is to press four layers of tissue paper against the substrate with the thumb and hold for approximately 10 seconds. If the tissue paper is dry and shows no signs of wetness, then the pipe surface is ready to be primed. The tissue test should be performed at different locations, a minimum of 10 locations for a spool of 6 meter length. If moisture is found on the tissue, then the pipe will need further heat and dehumidification. Space heaters can also be used to dry isolated locations that still have moisture. 4.1

Material

All epoxies prepared for priming, saturation, and final coats must be monitored for correct mix ratios. The epoxies used onsite for the FRP strengthening are two part epoxies consisting of a part A and part B. The epoxies should come from the manufacturer in preweighed containers that are have the correct ratio for the part B to be poured directly into the part A and mixed, without further measurement. If the epoxies are not sent pre-metered by the manufacturer, rather in bulk 220 liter drums, then the buckets should be weighed on a calibrated scale onsite to achieve the proper ratios as set by the material manufacturer’s quality control manual, prior to the mixing of the two parts. The weighing of the units should be recorded and noted in both the contractor’s field log and the inspector’s report. It is also important that all materials that are brought to the site are checked that they have not exceeded their shelf life. The fiber that arrives on site at the project must first be inspected to confirm that the fiber sent to the project is in acceptable condition. The ‘‘curl’’ test is used to verify proper material weave. The roll of fiber is rolled out and the edges are examined. The edges should not curl up on the fiber if the material is to be determined acceptable. The fiber should also be examined with attention that the tows are well stitched together and not pulling apart. Fiber saturation is controlled by the saturated fabric test. The saturator’s rollers are set at the correct spacing per the manufacturer’s recommendations for the material being used. The saturated fabric test is performed to ensure that the gap in the saturator has been set properly so that the proper epoxy to fabric ratio is obtained. A test piece of fabric, two feet in length, unsaturated, is cut and weighed to the nearest one hundredth of a pound on a calibrated scale on site, as seen in Figure 2.

The test piece is then run through the saturator, seen in the background of figure 2. The now saturated test piece is placed on the scale and weighed. The epoxy to fabric ratio is then calculated and should be within ±5 percent of the manufacturer’s written recommendation. This test should be performed at the start of each wrapping shift. If the epoxy to fiber ratio does not fall within the allotted variance, then the saturator is to be re-gapped and the weight test run again until the proper ratio is obtained. Periodic visual inspection should take place during the course of the fiber liner installation. The inspector should look for the orientation of the fiber with respect to the pipe wall. The pieces oriented in the circumferential direction should not vary more than 12.7 mm over 305 mm or 5 degrees from the aligned axis. The inspector should also note that the end details are correctly done, and that thickened epoxy is being applied between the layers. The visual inspection also incorporates checking for imperfections in the materials such as rips, voids or air bubbles behind the material. The gaps between consecutive bands should not exceed 12.7 mm width in the fabric’s transverse joint. The inspector should also note that the proper overlaps, as detailed in the project drawings, are being achieved. Throughout the layup process the inspector should check the materials being sent to the strengthening location and the application process.

use a mobile ultrasonic inspection system, like one pictured below in Figure 3. The ultrasonic inspection unit is scanned over the surface of the fiber liner, and the user looks for variances in the readings to pinpoint defects. If voids or de-laminations are found from the test then repairs are recommended. Defects less than 75 mm in diameter should be injected with resin or backfilled for large defects, greater than 75 mm in diameter. The engineer should recommend an alternate repair depending on what type of defect has occurred and where in the composite liner it has occurred. Arguably the most important quality control measure is the laboratory testing performed following the completion of the installation. During the course of the project, test panels are prepared on site. These panels measure 305 mm by 305 mm or 305 mm by 610 mm and consist of 1 or 2 layers thick, dependent on the type of material installed. Two samples are made for each wrapping shift, one sample is given to the inspector, and one sample is retained by the

4.2 Testing Following the installation and cure, the fiber liner should be checked for defects, bubbles, de-laminations, or fabric tears. The inspector has a couple of methods for examining the surface. First, a hard tool such as a quarter or a marble can be hit against the installed composite. The inspector should scan over the surface of the pipe at various locations listening for hollow spots in the liner. A hollow sound implies a void behind the material. A more advanced method of this test is to

Figure 2.

Weighing the saturated material.

Figure 3.

Inspection for ‘‘bubble trouble’’.

Figure 4.

ASTM D 3039 Tensile test.

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contractor. A predetermined percentage of the samples made are submitted to an approved testing laboratory. The ASTM D 3039 Standard Test Method for Tensile Properties of Polymer Matrix Composite Materials is performed on the samples that were made onsite as seen in Figure 4. This test verifies that the mechanical properties of the material applied in the field meet those properties used in the design of the pipe strengthening.

5

PLANNING & PREQUALIFICATION

There are many quality control measures that can be taken before the project ever begins. Material qualification, contractor pre-qualification, stringent specifications, detailed drawings, and a shutdown plan are all keys to preparing for a successful project and eliminating problems ahead of time. Manufacturer, material and installer prequalification are essential to a quality FRP installation. The project specification should address the years of experience for both the manufacturer and the installation contractor. A minimum of five years experience is typical. The specification should also detail the number of specific projects with references. For PCCP repairs, it is typical to require a minimum of ten documented successful installations for the installation company and a minimum of five key personnel with the same experience. These references should be reviewed as part of the Manufacturer/Installer prequalification process. Materials pre-qualification should consist of a durability report for the proposed system performed by an independent accredited testing facility. The report should reflect a minimum of 10,000 hours of accelerated durability under comparable conditions in which the material may be applied. In addition the materials should have been in service for a period of at least five years. Large scale proof testing of the proposed

material, such as internal water pressure testing and external de-load testing, as shown in Figure 5, should be a requirement for pre-qualification. All of the materials shall have NSF 61 certification and be approved for potable water applications. 6

The coordination between the Municipality and the Contractor that they have hired is essential to completing the job successfully. The owner must feel confident in the ability of the contractor that they have hired to understand the correct measures that must be taken to ensure a safe project and correct installation. Likewise the contractor relies on the owner to take the necessary precautions to ensure the contractor’s safety and facilitation of the work. The Metropolitan Water District of Southern California’s Foothill Feeder required spot repairs at two different locations in the line. Carbon fiber lining was chosen as the best repair method for the spot repairs. MWD then negotiated the project with Fibrwrap Construction, Inc. a certified installer of Fyfe Company materials. Both Fibrwrap Construction, Inc. and Fyfe Company are pre-approved contractor and material supplier, respectively for MWD. The Foothill Feeder Pipeline that required repair was a 5.1 meter diameter pipe, the largest that MWD had strengthened using FRP composites. There were many factors that posed challenges for the project. The repair locations were 365 meters and 732 meters from access points into the line. Because of the sizeof the pipe and the distance to access points, new safety standards had to be applied. These conditions had not been encountered on previously performed carbon fiber liner repairs. MWD feeds water to many neighboring districts. The shutdown of the Foothill Feeder Pipeline caused protest amongst many of the other agencies and utilities. MWD was forced to take these opinions to heart and limit the length of the shutdown period in order to satisfy the needs of the other distributors. Any contractor hired to perform the carbon fiber installation must have sufficient experience to perform these repairs without failure. The pre-qualification, QA/QC program allowed MWD the confidence to set a very aggressive short shutdown duration for such a logistically complicated job. 7

Figure 5.

CASE STUDY OF MWD FOOTHILL FEEDER

CONCLUSIONS

There are many steps involved in ensuring that the FRP work is done correctly and to the satisfaction and approval of the client. Including the QA and QC program details in the specifications is one to way to ensure a quality installation. As was the situation

External Load testing at the MWD facility.

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with MWD, the shutdown window had to be a short as possible. This would generally be the case with all owners of active water transmission systems. If sufficient consideration is given to access, dew atering planning, permitting, design coordination, mobilization, contract requirements, and field support, then the project should go as planned.

2003. Interim Criteria for Inspection and Verification of Concrete and Reinforced and Unreinforced Masonry Strengthening Using Fiber-Reinforced Masonry Polymer (FRP) Composite Systems (AC 178). International Code Council. 2004. Fiber Reinforced Polymer Composite Material Lining (Section 09980). The Metropolitan Water District of Southern California.

REFERENCES 2002. Design and Construction of Externally Bonded FRP Systems for Strengthening Concrete Structures (ACI 440.2 R-02). American Concrete Institute. 1997. AENC (DOC 8.11.97). California Department of Transportation.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Configuration, evaluation and selection tool (CET) for tunnel construction methods B. Schaiter & G. Girmscheid Institute for Construction Engineering and Management, ETH Zurich, Switzerland

ABSTRACT: Today the selection of the appropriate construction method for a specific tunnel project is mainly based on the individual, subjective experience of the personnel involved in the planning and work preparation departments. Out of this reason construction companies are demanding tools to support a rational approach for the selection of construction methods to optimize their costs. The scientific objective of this research project is to provide decision-makers with a Configuration, Evaluation and Selection Tool (CET) that enable them to perform their construction method selection in full awareness of the impacts of the same on the overall construction process. This means abandoning the former practice of analyzing individual sub processes—largely in isolation due to a lack of appropriate tools—in favor of the complete cross linking of the sub processes and, as such, the integrated analysis of the technical and economic impacts of individual decisions form each work cycle. 1

INTRODUCTION

Tunnel execution contracts are in the German speaking countries under very intensive price competition. A selective study by the authors could clearly reveal that – Tunnel projects with similar geotechnical and geometric boundary conditions have been executed with different construction methods and equipment configurations. – Execution companies achieved very different financial results between profit and loss an such projects. Further observations revealed that the selection of construction methods in design offices and particular in construction companies are only subject to the experience of the work preparation departments and the execution representatives. In discussion with some major leading construction companies it was stipulated that beside individual experiences a systematic tool should be developed to support the design and the bit phase as well as the execution preparation phase. This tool should support in a rational manner the evaluation and selection of the most appropriate construction method and the related optimized process with the equipment and auxiliary configuration. Normally for the calculation/construction process the value, the quality and the details of the infrastructure is stipulated. In the case of execution of the now finally designed infrastructure in regard to performance and quality, only the economical minimal principle is valid. That means the construction company must strive for the most appropriate method which fulfills the economical minimal principle.

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Out of this reason construction companies are demanding tools to support a rational approach to support and guide the experts by the selection of the most appropriate/economical construction method for the specific project.

2

STATE OF RESEARCH

Following in-depth market analysis focusing on the approaches to dealing with strategic decision-making issues during the bidding and work preparation phases, intensive literature research was conducted with the aim of finding information about automated decisionmaking models. Fundamental methodical approaches were found in the works by Girmscheid (2004b), Mawdesley (2002) and Motzko (1990). Since the works conducted so far only reveal elementary principles, or do not focus on the overall problem as a whole, a basic specification had to be drawn up for a first model construction which then served as the basis for setting up the Configuration, Evaluation and Selection Tool (CET) for Tunnel Construction Methods. So far this tool has been developed to the stage where it can simulate the impact of the selection of various tunnel construction methods in the different homogeneous longitudinal sections of a tunnel on the construction time and the associated costs.

3

RESEARCH METHODOLOGY

The scientific framework of the presented research project is embedded in the hermeneutic science

program (HSP) to understand, interpret and construct new socio-technical realities. Within the HSP Glaserfeld (1998) developed the constructivist research paradigm. Glaserfeld stipulates that constructivist models must be viable, valid and reliable to fulfill the intended target-means-relation. The presented Configuration, Evaluation and Selection Tool (CES-Tool) for tunnel construction methods with the intended target means relationship is constructed as an actional, generically-deductive model (tool) by following the constructivist research paradigm (Piaget 1973, Glaserfeld 1998, Girmscheid 2007). The scientific quality is achieved by triangulation (Yin 1994) due to:

The module 1 of the CES-Tool is structured as follows: 1. Input data: – Geological and hydrological classification for the different homogeneous longitudinal sections of the tunnel. – Geometrical cross section and longitudinal Cartesian data (inclination, length, curves etc) geological and hydrological classification for the different homogeneous longitudinal sections of the tunnel. 2. Tunneling construction methods 2.1 Tunneling excavation concepts

– viability of the actional, generic-deductive model, – validation through a theoretical framework and – reliability through testing the intended impact (target-means-relation).

• Cross section excavation concepts – full face – partial face

The Configuration, Evaluation and Selection Tool (CES-Tool) for tunnel construction is viable constructed as an actional, generic-deductively structured tool according to the intended target-means-relation. For the validation purpose the CES-Tool is theoretically-deductively structured by underlying the principles of the system theory (Bertalanffy 1969; Boulding 1956). Reliability was achieved due to calibration and realization test on executed projects. According to Yin (2003) the above shown triangulation concept fulfills the scientific requirements. 4

• Tunnel drive concepts – – – –

2.2 Tunnel support and lining methods – – – – –

Shotcrete Anchors Arches Full round lining Segmental lining

2.3 Tunnel logistic concepts

THE STRUCTURE OF THE CES-TOOL

The development of the CES-Tool for the selection of appropriate construction methods and the associated equipment configurations is divided fundamentally into two principal tasks. The first task focuses on modeling the structures, with the aim being to collate the geometric constraints into a building infrastructure model that can be processed systematically. The second task focuses on the definition of the interactions and interdependencies among and between the individual partial construction processes. The only means of ensuring with a high degree of probability that the partial construction processes and their interferences have been correctly defined and, in consequence, can be correctly incorporated into the construction process model, is by involving experienced experts. The CES-Tool for tunnel construction is structured in two modules. In module 1 the input data are collected out of geometrical and geotechnical data as well as resource data for the different construction methods (excavation methods, support and lining methods and logistic concepts). In module 2 the given data are processed automatically in a cycle time and cost estimation program.

Drill and blast Mechanical excavation by road header TBM (Tunnel boring machine) Special methods

– Mucking concepts – Construction material transport concepts The construction methods with their possible combination of excavation concepts (face excavation and drive concepts) and their various support and lining methods as well as the logistic concepts can be processed in a number of different overall construction processes. For each sequence and concept of construction resource use performance parameter are determined for the estimation of time and cost which will be determined in the following module 2 of the CES-Tool. The module 2 of the CES-Tool for tunnel construction (Fig. 1) is structured as follows: 1. Performance calculation: The CES-Tool includes a semi-automated cycle time an schedule program. The cycle times for the different stages of tunnel construction will be calculated according to Girmscheid (2004c) for the excavation cycles, loading and transport cycles, supporting and lining cycles. The interactions of the cycles and stages of construction are considered within the homogeneous

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geological longitudinal sections as well as under the different driving conditions. The total time of the interactive cycles as well as over the different homogeneous longitudinal sections is automatically aggregated. 2. Cost estimation: The CES-Tool includes a automated cost estimation program. Based on the input of resources like man power, equipment, materials etc and the results of the cycle time calculation of the performance calculation, the costs can be calculated. To do so additional cost input parameters according the country specific conditions can be determined, like labor wages, equipment costs, material costs etc. The method of calculation is based on recommendations by Girmscheid (2004c) and complies largely with the guidelines issued by the Swiss Association of Construction Entrepreneurs (Schweizer Baumeisterverband, SBV). 3. Risk analysis: The risk analysis will take the risks involved in the execution of construction projects into account, as defined by Girmscheid & Busch (2004) by: – Entering the main influence parameters as the expected value plus upper and lower limits. – Performing a Monte Carlo Simulation on this combination of values. The ensuing result is a bandwidth of the possible costs combined with the degree of likelihood of their occurring.

5

PRESENTATION OF THE CES-TOOL FOR TUNNEL CONSTRUCTION Figure 1.

5.1

CES-Tool for tunnel construction.

Systematic

Figure 1 presents in schematic form the systematic approach to perform the evaluation and selection of tunnel construction methods for a conventional soft ground excavation (i.e., using a tunnel excavator and rock support with shotcrete, anchors and steel arches). In addition to the example shown here, decisionmaking models for drill and blast excavation and TBM (Tunnel Boring Machine) drives in hard rock have also been developed. The utilization requirements of the tunnel determine the boundary conditions for the geometric design of the tunnel structure and the geological, geotechnical and hydrological boundary conditions. Once the most accurate possible understanding of these influencing boundary conditions have been gained, the engineer is responsible for developing possible construction methods based on geotechnical calculations and subsequently developed tunneling concepts which have to be are optimized for construction. A thorough analysis that incorporates all

631

possible tunneling methods is crucial in this phase. This is generally thwarted at present by the lack of appropriate aids that would allow an easy systematic assessment of the economic impacts of the project specific construction methods. The alternatives which have thus been identified must be subjected to a detailed analysis in the next step. The CES-Tool described in Section 5.2 is used to perform this analysis. This CES-Tool permits an iterative solution to be identified by varying the aforementioned influencing factors and other action factors until they produce the best possible construction method combination within the maximum performance at minimal cost. The CES-Tool requires the following input variables: – Utilization requirements (influencing factors) – Geology/Hydrology (influencing factors) – Human resources (action factors)

– Materials (action factors) – Inventory (action factors) The CES-Tool produces the following output variables: – – – – – –

Excavation performance Construction time Quantities of materials Wage costs Cost of materials Inventory costs

In order to select from the various possible excavation alternatives, the best excavation concept must be chosen based on the comparison of the respective output variables. Obviously in practice this decision can only be made by incorporating risk analyses, even if this paper does not specifically address this issue for capacity reasons. The conceptual findings from the comparisons and, above all, the lessons learned from executing construction projects must be documented and analyzed. These findings must be incorporated into subsequent excavation analyses as new action and/or influencing factors. Time constraints often mean that this iterative learning process is neglected in construction practice, although it is precisely this process that allows a company to substantially improve its ability to compete. 5.2

Computer program

The CES-Tool has been written to allow an initial construction method analysis. The objective of making the program suitable for general application produced the following requirements:

diagrams’’ and ‘‘Performance, quantities and costs’’ supply the output data/results. Figure 2 represents the core part of the CES-Tool and basically performs a classic calculation of performance. The program can perform calculations for 12 different execution combinations (excavation methods, supporting and lining concepts as well as logistic concepts) for each project. Since the performance calculation incorporates the boundary conditions (human resource and equipment concept), it can be performed by entering just a few input values. The calculation is structured in the following steps: excavation, supporting the vault, supporting the top heading invert, and additional works (only partially visible in the section of the table as shown). The upper section of the performance calculation table also displays all performance values for calculating the individual subcycles. These can be easily varied to see how individual changes to subcycles directly impact the overall result. The results of the performance calculation are displayed graphically as cycle diagrams (Fig. 3) for the individual excavation methods. Cycle diagrams serve to control and optimize work flows. They also immediately show whether individual works have been incorrectly assessed or even forgotten during planning. Based on the performance calculations for the individual excavation methods for a project and the ensuing construction program, the program enables a simple and rapid estimation of the total project cost, in addition to the detailed partial costs. As such, the decision maker is provided with clear information on the entire cost structure of the calculated alternatives, which greatly simplifies an assessment of

– The program must be well structured, clearly understandable and designed for simple application by the user. – Its field of application should be as generalized as possible: Option for drill and blast excavation in hard rock and soft ground excavation by means of tunnel excavator; one or more bores in use simultaneously; various layer modes; ability to cover all common cycles; application of all commonly used supporting techniques; deployment of all commonly used equipment. – The program must have sufficient calculation capacity. The CES-Tool is structured in four levels, as shown by the schematic diagram in Fig. 1. The input cells in the spreadsheets ‘‘Deployment of resources’’ and ‘‘Performance calculation’’ are marked in color. Some of them already contain examples of data that can be either adopted or overwritten. All of the cells that are not color coded do not need to be addressed by the user. The values in these cells are generated automatically by stored formulae. The spreadsheets ‘‘Cycle

Figure 2.

632

Performance calculation.

the individual alternatives. Fig. 4 shows an excerpt of the excavation costs per tunnel meter for various excavation classes of the tunnel with different homogeneous longitudinal sections, together with the total excavation costs. Furthermore, the program provides a detailed breakdown of the labor, material and equipment costs. Figure 5 shows the excavation performance in the various excavation areas, together with specific performance indicators.

Figure 6. Detailed comparison of excavation costs for different excavation concepts.

Figure 7.

5.3

Figure 3.

Cycle diagram.

Figure 4.

Excavation costs.

Performance—cost analysis.

Comparison of alternative construction methods

In addition to the pure investment costs, construction time is enormously important when comparing alternatives, especially in the case of infrastructure projects, such as tunnels. This program therefore displays not only the costs (Fig. 6), but also the excavation performance relative to the costs (Fig. 7). The program can provide graphic illustrations of alternative comparisons for any number of calculated alternatives. The tables shown in Figs. 6 and 7 are linked to the digital performance calculation in such a way that any changes to the input data (input spreadsheet ‘‘Resources’’ are automatically reflected in the charts. This permits the specific analysis of cause and effect relationships—e.g.: What changes occur to the excavation performance and costs in alternative 1 in Figure 7 if the distance between the steel arches used for supporting the tunnel is increased from 1.0 m to 1.2 m? The model on which the program is based allows all time and cost impacts on an alternative to be taken into consideration. By cross linking the sub cycles of the tunnel construction, it also shows all of the impacts on the other sub cycles. 6

Figure 5. Excavation performance and specific performance indicators.

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CONCLUSIONS

The Configuration, Evaluation and Selection Tool (CET) developed in this research project demonstrates

how detailed analyses for tunnel excavation concepts can be performed fast, systematically and economically. The tool enables planners and entrepreneurs to perform a rational in-depth analysis within the— usually short—periods available for preparing a bit. In particular, and equally for the first time, planners are given a means of directly ascertaining the economic relevance of their planning decisions while still in the initial project phases, which in turn enables them to realize more cost-efficient overall concepts. Only the systematic consideration of all costeffective decisions in a very early project stage enables the selection of the appropriate construction method and later the target-oriented optimization of the overallproject. REFERENCES Bertalanffy, L. 1968. General System Theory: Foundations, Development, Applications. New York: George Braziller. Boulding, K. 1956. General System Theory: The Skeleton of Science. In: General Systems, 1. Jg., S. 11–17. Girmscheid, G. 2004a. Forschungsmethodik in den Baubetriebswissenschaften. Eigenverlag des IBB an der ETH Zürich, Zürich.

Girmscheid, G. 2004b. Ausführungs- und Angebotsmanagement— Leitfaden für Bauunternehmen, Springer Verlag, Berlin (GER) 2004. Girmscheid, G. 2004c. Kostenkalkulation und Preisbildung in Bauunternehmen, Springer Verlag, Berlin (GER). Glasersfeld, E.V. 1998. Radikaler Konstruktivismus: Ideen, Ergebnisse, Probleme. 2. Aufl., Surhkamp, Frankfurt a.M. Mawdesley, M.J. et al. 2002. A model for the automated generation of earthwork planning activities. Mayring, P. 2002. Einführung in die quantitative Sozialforschung: eine Anleitung zu qualitativem Denken. 5. Aufl., Beltz, Weinheim. Motzco, C. 1990. Ein Verfahren zur ganzheitlichen Erfassung und rechnergestützten Einsatzplanung von Schalungssystemen, VDI Verlag GmbH, Düsseldorf, 1990. Piaget, J. 1973. Erkenntnistheorie der Wissenschaften vom Menschen: die Wissenschaften vom Menschen und ihre Stellung im Wissenschaftssystem. Ullstein, Frankfurt/M. Weber, M. 1992. Gesammelte Aufsätze zur Wissenschaftslehre. Mohr, Tübingen. Yin, R. 2003. Case Study Research—Design and Methods. 3rd ed., Sage Publications, Thousand Oaks.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Formwork specific, process orientated geometrical-path-velocitytime-model (GPVT-model) M. Kersting & G. Girmscheid ETH Zurich, Institute for Construction Engineering and Management, Zurich, Switzerland

ABSTRACT: The critical path method and path-time analysis, conventionally used for work preparation for formwork activities, only use average activity times per square meter for stripping and erecting formwork as planning parameter without considering the specific geometrical and structural project conditions. Considerable uncertainty therefore surrounds activity times and resource usage and sometimes results in serious financial deviations. To resolve these uncertainties ETH Zurich has developed a new geometrical-path-velocity-time-model (GPVT-model) in cooperation with the scaffolding/formwork manufacturer DOKA. This GPVT-model determines the realistic and optimized path-velocity-time-relation of all major works related to formwork activities under consideration of the real geometrical and structural dimensions of the building based on data empirically collected on different sites. The optimal crew size and the realistic activity duration are then determined as input for the newly developed formwork-selection-process-model. Construction companies can now select the optimal formwork system for the specific project and optimize their manpower resources and minimize their operation costs. 1

INTRODUCTION

The unique character of each new building raises the question of which formwork system to choose during every site preparation planning process. To date, only rough cost and time estimation methods have been available to identify the optimal formwork system. Therefore personal preferences influence the decision making process instead of analytical work simulations. This leads in most cases to suboptimal solutions for the specific construction project. For these rough time and cost estimations the construction process is usually split into several main processes. The time consumption per main process is calculated based upon general performance factors (Zentralverband des Deutschen Baugewerbes e.V. 2001) such as ‘‘duration for stripping and erecting the slab formwork per square meter’’ or ‘‘duration for pouring concrete per square meter’’. These factors are gained from the project controlling results of previous projects and do not reflect any geometrical and structural properties of the respective project. 2

RESEARCH METHODOLOGY

The scientific framework of the presented work is embedded in the hermeneutic science program (HSP) to understand, interpret and construct new sociotechnical realities. Within the HSP, Glasersfeld (1998) developed the constructivist research paradigm. Glasersfeld stipulates that constructivist models must

635

be viable and have to fulfill the intended target-means relation. The structure of this GPVT-model is actional, generically-deductive as a target-means relationship using the constructivist research paradigm (Girmscheid 2007b; Glasersfeld 1998; Piaget 1973). The scientific quality is achieved by triangulation (Yin 1994) due to: – Viability of the generic-deductive model – Validation through a theoretical framework – Reliabilitation through testing the intended impact (target-means relation) The presented GPVT-Model is viably constructed as an actional, generic-deductively structured model according to the target-means-relations. For the validation purpose the GPVT-Model is theoretically-deductively structured using the principles of system theory (Bertalanffy 1969; Boulding 1956). Reliabilitation will be achieved using realization tests to check if alternative target relations exist under equal means. According to Yin (1994) the above sketch triangulation concept fulfills the scientific quality requirements. 3

SYSTEM BOUNDARIES

The design types of new building constructions vary enormously. It is impossible to develop a model from scratch that will satisfy all possible constraints for all

varieties of design types and simultaneously reflect all possible influencing factors. Therefore the GPVT-model is focusing only on multi-story office buildings, which represent a large proportion of building construction. In the GPVT-model, the following restrictions have been set:

Formwork-selection-process-model (FSP-model) Formwork systems System A

System B

System not possible Part I

Test: Applicability

Geometrical-path-velocity-time-model (GPVT-model) INPUT: Geometrical and structural properties Geometrical-path-velocitiy-time-analysis

In a later step, the GPVT-model can be extended to buildings with other floor layouts (i.e., not rectangular).

OUTPUT: Basic-elementary-process duration

x

a k LE k ;

Tk

y v

Logistics-interaction-model (LogIn-model)

Part II

ALLOCATION OF THE GPVT-MODEL IN THE FORMWORK-SELECTIONPROCESS-MODEL (FSP-MODEL)

INPUT: Resource availability Team size determination and manpower allocation OUTPUT: Activity duration per level i

ATi

AT k ,m k

T k ,m

m

k

m

Variation of the team size no

1 nAK ,k ,m

1 AK , k , m

Test: Suitable cycle duration yes

System not possible

INPUT: Preparation

Wh ,all

INPUT: Post processing

Logistic-interaction-analysis

All team sizes unsuccesful

The developed GPVT-model is the first part of the formwork-selection-process-model (FSP-model; see Figure 1). Before applying the geometrical-path-velocitytime-analysis it is essential to test the applicability of the formwork systems in question in regard to the technical and construction process boundary conditions using specific knock-out criteria. Such knock-out criteria could be the size of the façade openings of the building, which do not allow the table formworks to be simply shifted out of the building. After selection of the appropriate formwork system the geometrical-path-velocity-time-analysis can be conducted with the geometrical and structural properties as input data. The results of part I are the basicelementary-process (BEP) durations of the different sequences of works without considering the team size and the parallel performance of the works by several teams. The results of this analysis will be transferred to the logistics-interactions-analysis (part II of the FSPmodel). In this analysis the activity duration will be calculated per floor based on the chosen/determined team size and team number as well as equipment assignment. Following incorporation of the preparation and post processing of the used equipment, the total activity duration will be summarized over all floors and compared with the total timeframe allowed for the construction project. In the next step, both the consumption of wage hours and the allocation of construction equipment are combined in order to determine the demand for manpower and equipment. These formwork specific cost calculations enable the identification of the optimum system from all potential formwork systems.

no

System ...

yes

– Multi-story office buildings with standard floor plan – Standard floor plans are rectangular – Use of prefabricated columns

4

System n

OUTPUT: Activity duration for all levels

ATall

ATi i

Variation of the team size no

Consumption of wage hours AT k , m n AK ,k ,m i

k

Test: Maximum project duration yes Consumption of construction equipment CEqu ,all CEqu ,k

m

i

k

Calculation (economic minimization principle) Identification of the optimum formwork system

Figure 1.

Placement in the FSP-model.

Both the geometrical-path-velocity-time-analysis and the logistic-interactions-analysis have to be performed for both slab and wall formworks. This paper only addresses the implications for the slab formworks.

636

Table 1.

Classification of work steps.

Formwork stripping

Example

Construction processes

• Construction of an office building • Structural works • Building pit • Excavation • Wall erection • Slab erection • Formwork stripping • Formwork shifting/lifting • Formwork erecting • Armoring • Concreting • Hooking the formwork table to the crane • Lifting • Turning the crane/Moving the crane trolley • Lowering • Positioning the formwork table

Main processes Module processes Elementary processes

Sequences of work

5

DEVELOPMENT OF THE ELEMENTARY PROCESSES

The process orientated approach to the investigated problem requires a distinct classification of the different work steps according to Girmscheid (2007a) (see Table 1). Initial investigations focus on the level of the sequences of work. Once these sequences of work have been completely described, the geometricalpath-velocity-time-relations can be aggregated on the elementary process level for subsequent use in the logistics-interaction-model.

6

DEVELOPMENT OF THE MATHEMATICAL FUNCTIONS

(1)

637

veli

Le

vel

Erecting of floor props

i

Le

i-1 vel

Li

Le

Shifting/Lifting SL1: Li SL2: Li -Li+1 Formwork erecting L i+1 Placing the reinforcement Li+1

Figure 2.

Concreting Li+1

Hardening Li+1

Process chain in cycle i.

During cycle i, the formwork is stripped then shifted on level i and finally lifted from level i to level i + 1. On level i + 1, still in cycle i, the formwork is erected so that the reinforcement can be placed prior to commencing concreting (Figure 2). The next cycle i + 1 starts after the concrete has hardened. The following nomenclature has been developed to allow clear classification of the different basic elementary process (BEP) durations (abbreviated as T ) and the relevant elements:

Wall/Slab,Formwork System

TElementary process,Cycle

(2)

The superscript index classifies whether it is valid for walls or slabs and what formwork system is described with this BEP-duration. The subscript index shows which elementary process is surveyed and in which elementary process this occurs in the analyzed cycle.

6.1

At first it is necessary to distinguish between the different performance definitions which exist according to Girmscheid (2005) in order the reflect the big set of influencing factors to the performance. Starting with the theoretical performance QTheo it is possible to calculate the effective performance QEff under consideration of factor k1 , reflecting the age and the condition of the formwork, of factor k2 for the geometrically dependent effects (space availability and complexity), and finally of factor k3 for operationally and organizationally dependent effects. The effective performance QEff can now determined: QEff = QTheo · k1 · k2 · k3

+1

Li

Level

Focus on formwork shifting/lifting

Figure 2 shows the whole process chain for the production of one structural concrete floor with the elementary processes. The elementary process of formwork shifting/lifting is split into two interacting sequences (Shifting/Lifting1: SL1 and Shifting/Lifting2: SL2). The first sequence (SL1) describes the horizontal shifting on level i, usually supported by equipment like shifting trolleys. The dashed line in Figure 3 shows this shifting sequence (SL1) of formwork tables. Only the inward positioned tables (lighter colored in Figure 3 will be shifted using a table trolley. The second sequence (SL2) describes the lifting from level i to level i + 1 as a combined vertical and horizontal movement, operated by the crane (solid line in Figure 3).

t Fix

FTj −1

t Prep

t Fix

FTj +1

FTj

FTj + 2

t Prep

Figure 3.

6.2

Formwork shifting and lifting to the next level.

Only the inward positioned formwork tables have to be shifted to the level edge. Nevertheless the formwork tables positioned on the level edge have to be considered in regard to the preparation works. The process time in Equation 3 is composed on the one hand of performance factors related to the number of formwork tables (inward and edge) and on the other hand of the distance covered with the shifting trolley (only inward):

+

 j(inward)

, SFT vSL 1

Path-time graph for SL1 (shifting trolley). Crane activities for formwork tables.

Abbreviation

Activity

tL1

• • • • •

THo tFi

• • • • •

tT tM tL0

  S,FT ,Edge S,FT ,Edge · nS,FT ,Edge + aPrep + aFix 

ΔxFTj +2,e + ΔyFTj +2,e

, SFT vSL 1

Table 2.

  S,FT S,FT ,Inward ,Inward · nS,FT ,Inward TSL + aS,FT 1,Cyci = aPrep Fix



ΔxFTj ,l + ΔyFTj ,l

, SFT vSL 1

Figure 4.

Horizontal shifting of formwork tables (SL1)



ΔxFTj ,e + ΔyFTj ,e

xFTj,e + yFTj,e + xFTj,l + yFTj,l

tPo



Lifting from i to i + 1 Lowering from i + 1 to i Insert the crane hook Remove the crane hook Fixing the formwork table to the crane hook Turning the crane Move the crane trolley Lifting on level i + 1 Lowering on level i + 1 Positioning the formwork table at i + 1 and releasing from the crane

S,FT vSL 1

(3)

6.3

The performance factors used in Equation 3 describe the time consumed in preparing and fixing the formwork tables to the crane hook. Preparation in this context involves the following work sequence: positioning of the table trolley under the formwork table, loosening the formwork table from the concrete slab and lowering the formwork table with the table trolley in order to shift the formwork table below the other tables. Following horizontal shifting the formwork table is briefly parked at the edge of the floor in order to attach it to the crane hook. The preparation of the formwork tables on the edge differs slightly in that the table trolley is not shifted horizontally. In Figure 4 the path-time-graph for SL1 displays the movements and activities related to the different positions (inward and edge). The dashed lines show the transfer point where the interaction with SL2 takes place. The workers of SL1 finish with formwork table FTj and fix it to the crane hook. From there on FTj continues with the crane in SL2. In SL1 the workers deal with the next formwork table FTj+1 , in this example with a formwork table positioned at the level edge, trolley shifting is therefore not needed.

Vertical and horizontal lifting of formwork tables by crane (SL2)

Equation 4 consists of the different sequences of work elements (abbreviated with t) of the activities related to the crane lifting (see Table 2): S,FT TSL 2,Cyci =



(tL1 + tHo + tFi + tHo + tL1

j

+ Max(tT ,FTj ; tM ,FTj ) + tL0 + tPo + tHo + tL0 + Max(tM ,FTj ; tT ,FTj )) (4) Most crane activities appear twice in Equation 4 as both paths (loaded and empty crane hook) are considered. The crane hook follows these points for each formwork table (subscripted with j):

638

– – – –

Fork insert point (point 1) Start position of formwork table at level i (point 2) Back to point 1 Target position of formwork table at level i + 1 (point 3) – Fork release point (point 4)

Positioning and erecting the formwork: tL 0 t Po t Ho tL 0

   S,TM S,Floor + TErec,Cyc = aS,TM − AS,Core aS,TM Erec · Ai+1 i+1 Erec,η · Zη i

FTj −1

η

FTj +1

FTj

(8)

Placing of reinforcement: t L1

Figure 5.

S TRein,Cyc = aSRein · mS i

t L1

t Ho t Fi t Ho

tT , FTj resp. t M ,FT j

t M ,FT j resp.tT ,FT j

Concreting:

Path-time graph for SL2 (crane).

In Figure 5 these points (with the point-sequence per table 1-2-1-3-4-1) are combined with the crane activities from Table 2 for the formwork table FTj . At point 2 the transfer A (from SL1) can be found again. Transfer B shows the interaction to the next elementary process of positioning and erecting. Transfer A is of great importance as this results in a strong interrelation between the two elementary processes SL1 and SL2. This illustration also shows the changing load status of the crane hook between empty and loaded. 6.4 Consolidation of the BEP-durations In addition to the geometrically related BEP-durations, the other BEP-durations mainly depend on conventional performance factors. All BEP-durations are consolidated in Equations 5 to 10: Stripping the formwork: S,TM TStr,Cyc = i

 η

S,TM ,FP aS,TM Str,η × Zη + aStr,Cyci

  × AS,Floor − AS,Core i i

(5)

Shifting and Lifting 1:   S,FT S,FT ,Inward ,Inward · nS,FT ,Inward TSL + aS,FT 1,Cyci = aPrep Fix   S,FT ,Edge S,FT ,Edge · nS,FT ,Edge + aPrep + aFix +

   xFTj,e + yFTj,e + xFTj,l + yFTj,l

j(inward)

S,FT vSL 1

(6) Shifting and Lifting 2: S,FT TSL 2,Cyci =

   dS  S,Floor S,Core S S · = A − A + a TConc,Cyc Post i+1 i+1 S i QConc (10)

7

 (tL1 + tHo + tFi + tHo + tL1

COMPLETION OF THE CALCULATIONS

With the results from Equations 5 to 10 all necessary information for the LogIn-model (part II of the FSPmodel in Figure 1) has been developed. Applying and then optimizing the team sizes under consideration of parallel performance of the works by several teams will then lead to the total project duration including the total consumption of wage hours and construction equipment.

8 8.1

IN-SITU RECORDINGS Identification of the performance factors

Since not all of the performance factors used in Equations 5 to 9 are available in literature (Zentralverband des Deutschen Baugewerbes e.V. 2001), in-situ recordings have been performed on different Swiss construction sites. Values have been determined for both the average shifting velocity (vSL1 ) and the process times for the crane activities (as in Table 2). 8.2





(9)

Preparation of the recordings

A setting plan for the formwork tables is usually established before the start of construction in order to determine the number and position of the formwork tables. These setting plans can therefore be used during the recording process to accurately record the movements of the formwork table trolleys. These records are captured by a digital video camera in order to analyze all simultaneous sequences of work retrospectively.

j

8.3

+ Max(tT ,FTj ; tM ,FTj ) + tL0 + tPo + tHo + tL0 + Max(tM ,FTj ; tT ,FTj )) (7)

639

Results of the recordings

The surveyed construction site used formwork tables measuring 5.00 m by 2.50 m. The average velocity (vSL1 ) has been determined at 12 m/min.

9

COMPARISON OF MODEL CALCULATION AND IN-SITU RECORDS

The calculations based on the values gained from the in-situ records have been compared with the real process times on the construction site. The discrepancy between the two values has been very small. Further recordings will have to be performed in order to obtain values for the wall formwork activities as well. 10

Abbr. Meaning

Unit

a A AT C d L k1

i.e. [min/m2 ] [m2 ] [min] [CHF] [m] [–]

k2

CONCLUSIONS

The developed process orientated geometrical-pathvelocity-time-model considers the basic work sequences as time related work steps. Much more reliable time and cost estimations are possible by incorporating the geometrical and structural parameters of the building project and the specific work sequences. The conducted in-situ recordings confirm the developed model. The results from these estimations allow also a secure decision to be made in favor of one of the potential formwork systems.

k3 m n Q T t v W Z

Performance factor Area Activity duration Costs Thickness Level Factor reflecting age and condition Factor reflecting geometrical effects Factor reflecting operational and organizational effects Weight Number of formwork tables Output Basic-elementary-process (BEP) duration Sequences of work elements Velocity Wage Special formwork

Index Meaning REFERENCES Bertalanffy, L.v. (1969). General system theory; foundations, development, applications. New York: G. Braziller. Boulding, K.E. (1956). General Systems Theory-The Skeleton of Science. Management Science 2(3): 197–208. Girmscheid, G. (2005). Leistungsermittlungshandbuch für Baumaschinen und Bauprozesse. Berlin; Heidelberg; Zürich: Springer; Vdf, Hochsch.-Verl. an der ETH. Girmscheid, G. (2007a). Bauproduktionsprozesse des Tiefund Hochbaus. Zürich: Eigenverlag des IBB an der ETH Zürich. Girmscheid, G. (2007b). Forschungsmethodik in den Baubetriebswissenschaften. Zürich: Eigenverlag des IBB an der ETH Zürich Glasersfeld, E.v. (1998). Radikaler Konstruktivismus Ideen, Ergebnisse, Probleme. Frankfurt a.M.: Suhrkamp. Piaget, J. (1973). Erkenntnistheorie der Wissenschaften vom Menschen die Wissenschaften vom Menschen und ihre Stellung im Wissenschaftssystem. Frankfurt/M: Ullstein. Yin, R.K. (1994). Case study research: design and methods. Thousand Oaks: Sage Publications. Zentralverband des Deutschen Baugewerbes e.V. (2001). Arbeitszeit Richtwerte Hochbau. Neu-Isenburg: Zeittechnik Verlag GmbH.

ABBREVIATIONS Abbr. Meaning x y

Distance covered in x-direction Distance covered in y-direction

Unit [m] [m]

Conc Cyc e Edge Equ Erec Fix

Concreting Cycle Empty Fork/Hook Position of the formwork table Equipment Erection of the formwork Fixing the formwork tables to the crane hook Floor Floor of the building FP Floor prop FT Formwork tables i Index for the floors Inward Position of the formwork table j Index for formwork tables k Index for the different elementary processes l Loaded Fork/Hook m Index for parts of the elementary processes Pos Positioning of floor props Post Postprocessing Prep Preparation of formwork tables Rein Placing of reinforcement S Slab SL1 Shifting/Lifting 1: horizontal with trolley SL2 Shifting/Lifting 2: vertical/ horizontal with the crane Str Stripping the formwork η Index for special formwork

640

[–] [–] [–] [kg] [–] [m3 /min] [min] [min] [m/min] [CHF] [–]

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Open building manufacture systems: A new era for collaboration? M.D. Sharp CIRIA, London, UK

J.S. Goulding University of Salford, Greater Manchester, UK

ABSTRACT: The UK construction industry has been the focus of concern regarding its skills provision. This has created a series of skill gaps and produced a general shortfall of ‘qualified’ workers. This paper reports on findings from a research project funded by the European Commission, to establish current attitudes towards education and training regarding Open Building Manufacturing (OBM) systems—a somewhat ‘new’ paradigm. An attitudinal survey ‘questionnaire’ approach was used to ascertain gaps in knowledge that could undermine the effective future use of OBM. Findings note that whilst fragmentation still exists, attitudes have significantly changed over the past 15 years regarding the overall appreciation of the need to address future skill sets for the successful uptake of OBM systems within the UK context (in order to ensure that mistakes from the past are not repeated). A collaborative approach between industry and academia is suggested as a possible way forward. 1 1.1

INTRODUCTION Overview

Over the last 50 years, research has been carried out to identify the means through which the construction industry could be improved. However, little has been achieved in this respect (Koskela 2000). Several factors could well be levelled as tangible excuses, not least skills shortages (McNair and Flynn 2006), and the need for a diametric change in culture (RIBA 2005). For example, the fragmented nature of the construction industry (Emmerson 1962; Banwell 1964; Latham 1994) has often been cited as a primary factor that has adversely affected performance and productivity. In this respect, contemporary ‘change’ initiatives have tried to improve performance by focussing on time, quality or cost elements; and, Kagioglou et al. (2001) noted that the majority of problems in the construction industry were more often than not process related, and not product related. Furthermore, the ‘Rethinking Construction’ initiative (Egan 1998) highlighted a specific ‘crisis’ in training, as the proportion of trainees in the workforce appeared to have declined by half since the 1970’s, which created increased concern about skills shortages in the industry. From a United Kingdom (UK) perspective, traditional building concepts and procurement approaches have been identified as targets for manufacturing mechanisms (Egan 1998). This, combined with a re luctance to innovate and an industry steeped in risk aversion, has led to a market that is ostensibly slow in introducing appropriate technologies to deal with the problems associated with better quality, lower

641

cost and more efficient construction processes. The uptake of this technology has therefore been slow and parochial, as designers have tended to stay loyal to ‘traditional’, tried and tested technologies—their ‘comfort zone’ (Goulding et al. 2007). It is therefore assumed that Egan (1998) intended to address these issues by advocating ‘standardisation’, ‘manufacturing’ and adopting a ‘holistic approach’ to construction. 1.2

Research aim and objectives

This paper gauges the level of industry interest (and relevance) for the development of industrial training packages to address the skills shortage within the construction industry with specific reference to ‘standardisation’ and ‘manufacturing’. This research focuses specifically on modern methods of construction, particularly the concept of Offsite Manufacturing (OSM) processes and their practical application to exhume a modern built environment fit for social, economic and environmental legacy. 2 2.1

OFFSITE MANUFACTURING Background

Despite various initiatives to identify broad paradigms of the nature of the behaviour of the construction industry (Sharp et al. 2005), it ostensibly still remains fragmented and slow to introduce appropriate technologies that could potentially increase efficiencies and productivities by reducing costs and time. Previous studies, including the Levene Efficiency Scrutiny

(Cabinet Office Efficiency Unit 1995), highlighted the inefficiencies of traditional industry methods; whereas, Yeomans (2000), noted the importance of performance issues. Modern methods of construction (as defined by the House Builders Federation), are those that enable an efficient product management process to provide more quality products over a shorter time period. These methods include those of OSM which is a relatively new paradigm for building production in the UK. The concept of OSM deviates from traditional ‘‘craft and resource-based construction’’ to combine ambient manufacturing methods with an open system for products and components. OSM offers a diversity of supply in the market, with a high degree of design flexibility at a lower cost. It is a combination of ambient manufacturing methods and value-driven business processes appropriately supported by information communication technology (ICT) to provide affordable, customised and flexible (configurable on demand) buildings. This is purported to improve the quality of life and provide better value to the customer through a diverse range of ‘‘plug and fix’’ modules and components and related services offered by knowledge-driven small/medium enterprises (SME). In this respect, various initiatives have already been funded to promote and demonstrate these benefits (ManuBuild 2009). The application of OSM in the UK construction industry has been the focus of considerable attention over the past few years, particularly within the housing market (Sharp et al. 2005). In an attempt to promote technology and alternative methods of construction, the Housing Forum produced a number of reports to inform industry stakeholders of the benefits of OSM and its practical application. Despite this, past UK experience of non-standard house construction and the inadequacy of current levels of training/skills amongst the workforce were identified as major obstacles that needed to be addressed if OSM was to have a major impact on the UK house building industry. This is further highlighted by the fact that construction output is expected to increase with the announcement of a £4.7 bn government programme to renew or rebuild every secondary school in the country. In this respect, this will be the largest single project, dwarfing developments for the 2012 Olympics; and it would appear that the shortage of an appropriately skilled workforce is about the only thing to stop the increased output. From a European Union (EU) perspective, the EU construction sector is one of the largest industrial employers (2.3 million enterprises) with a total of 11.8 million employed. Furthermore, jobs held by 26 million workers in the EU are dependant (directly or indirectly) on construction. However, despite the fact that 96% of construction enterprises in Europe are SMEs (with fewer than 20 operatives); the construction sector is of significant importance to the European economy with a Gross Domestic product (GDP)

contribution of 9.8% and European employment with an overall employment rate of 7.1% of the European workforce (Business Watch 2005). Furthermore, the gap between the different EU project results and industry practices has been identified as one of the reasons for the industry’s reluctance to uptake new technologies and concepts (Rezugi and Zarli 2006). Hence training and education is purported to play an important role in communicating and demonstrating technological solutions and benefits to stakeholders. However, the ‘typical’ educational model has often been criticised for providing general instruction to students, with the anticipation that the prospective employer would be responsible for delivering on-thejob training. This is characterised as the ‘skills gap’. In the UK, there has been an overall increase in the productivity of the construction industry over the years. Despite this, there is evidence that the industry is still slow to adapt to change, and this is reflected in the productivity gap between the construction sector and other industries (DFEE 2000). The Egan Report (1998) recognised the need for workforce training and continued professional development for improvements in construction performance. The perceived skills gap between the construction sector and other industries is further confirmation of the need for improvement in the knowledge and skills base of all construction industry stakeholders (see Table 1). Table 1. Expected skills gap between the construction sector and other industries within the UK from 2001–2003 (DFEE 2000). Mining Construction All and and specialist indusquarrying contracting tries Difficulties in meeting customer service objectives Increased operating costs Delays in developing new products or services Loss of business or orders to competitors Withdraw certain products or services altogether Difficulties introducing technological change Difficulties in meeting required quality standards Difficulties introducing new working practices

642

21

40

33

6

38

32

4

25

20

5

24

19

−

22

12

10

21

18

10

20

27

5

16

21

In addition, Hurst et al. (2007) conducted a survey of some of the largest UK construction companies to establish the level of staff development that currently exists. The results indicated that only 37.14% had a comprehensive implementation of training programmes within their organisations, with 5.71% of companies stating a complete absence of training and education policies. These figures reflect the significant need for improvement in continued professional development (CPD) of industry professionals. Other research studies have identified a number of industry concerns with respect to the adoption of OSM as a preferred method of construction (Sharp and Goulding 2008), particularly the lack of skills within the workforce to effectively project manage offsite manufactured projects (both during design and during the construction stages), therefore compromising design and construction. 3

RESEARCH METHODOLOGY

This research examines the requirements for and the implementation of training for OSM within the construction sector. The research methodological approach adopted commenced with an initial review of the core research material and literature pertaining to the general methods of learning per se, the direct links to application (from an OSM skills requirement perspective), and change. This seminal literature was distilled and augmented into several core skill gap areas reflecting the ‘needs and gaps’. These areas were then piloted with domain experts for relevance and appropriateness, and then separated into 14 questions. From a data collection perspective, an online survey was considered an appropriate tool for capturing feedback data, as bespoke database of 104 organisations, together with an Industrial Training Network was available as a test bed for collecting this. Thus, the survey took the form of questionnaire approach, using 14 core questions as a means of capturing respondents’ responses, and 1 general ‘catch-all’ question in order to capture additional tacit information from respondents. The ‘essence’ of these 14 questions can be seen in Table 2. The rationale for this survey approach was to capture and identify the areas of interest within OSM that would be topical and relevant to stakeholders. The target group included all members of the Construction Productivity Network (CPN) which is a network that is managed by a member-based research and information organisation in the UK (CIRIA 2009). In this respect, CIRIA provides a platform for construction stakeholders to share experiences in order to improve industry performance and promote bestpractice. Currently, there are 104 members of the CPN and include leading contractors, clients, designers, suppliers and research bodies with an interest in

643

Table 2.

Questionnaire construct.

No.

Question

1 2 3 4 5 6 7 8 9 10 11 12 13 14 15

SMART components Space/Flexibility Assembly Design specification Value-driven processes Benchmarking Organisational models Service models Business models New manufacturing technologies Portable/mobile factories Logistics—design to manufacture Logistics—manufacture to assembly Logistics—assembly to completion General ‘catch all’

innovative construction methods. In addition to this, an Industrial Training Network was established to invite professionals outside of the CPN to take part in the survey. 4

RESULTS AND DISCUSSION

The survey was sent out to all members of the CPN (104 organisations), and an overall response rate of 38% was achieved. The link through the Industrial Training Network generated a further 163 responses, which therefore provided a broader survey sample. Table 2 presents a list of particular aspects of OSM that were compiled and included in the survey questionnaire to determine their relevance to industry if they were to be integrated into a training package. From the results, all 14 aspects of OSM were considered to be highly relevant to industry—see Figure 1. The most interest was gauged in the development and implementation of a service model, addressing the issues of facilities management to include adaptability, recyclability and re-use. However, it is interesting to note that the design of integration to the use of space was not considered as ‘relevant’, nor was the need to take new technologies in to account when identifying industrial training needs of the sector. Whilst this perceived lack of relevance could not be overtly identified, especially as there has been a long standing debate (Thompson, 2007) over the importance of design and new technologies to volumetric building, the resonance of this is a critical aspect which will need further investigation. Other results demonstrate that respondents’ identified that serviceability, site assembly and seamless logistics support for both manufacture and site assembly were highest on their list of training needs. These issues have been identified as part of the need to integrate supply chains throughout the construction sector

Figure 1.

OSM relevance to industry—survey results.

(Love et al. 2004). Furthermore, it would appear that there is an opportunity for the sector to advance past its fragmentation to unite in delivering an augmented industry that delivers on its promise to be economically viable, environmentally friendly, and to be flexible enough to meet the demands of an emerging society. 5

CONCLUSIONS

This report concludes that there is still an overall need to improve construction industry performance. Whilst the Levene Efficiency Scrutiny (1995) recognised the inefficiencies of traditional building methods, and the Egan Report (1998) highlighted the need for workforce training and continued professional development to achieve an increase in construction industry performance; it is also apparent that OSM relevance to industry still needs investigation in order to fully understand industry’s somewhat entrenched position. In this context, it is imperative that industry stakeholders are engaged and consulted in the development of the industrial training packages in order to provide knowledge that is relevant in order to address the ‘skills gap’. By addressing these identified factors holistically, the sector will then be in a better position to move forward in delivering a process ‘fit for purpose’—specifically, what it was designed to do, thereby embracing the key operands of being: ‘‘flexible’’, ‘‘socially dynamic’’, ‘‘economically efficient’’, ‘‘environmentally acceptable’’ etc. This paradigm would however naturally require the concatenation of similar studies of this nature in order to create a longitudinal analysis of all the interceding variables. REFERENCES Banwell Sir H. (1964)., The Placing and Management of Building Contract., The Banwell Report., HMSO, London.

Business Watch, The European e-Business Market Watch (2005)., Sector Report No. 08-II, ICT and Electronic Business in the Construction Industry, IT adoption and e-business activity in 2005, European Commission, Enterprise and Industry Directorate General. Cabinet Office Efficiency Unit (1995)., Construction Procurement by Government: An Efficiency Office Scrutiny (the Levene Report). London: The Stationery Office. CIRIA (2009)., www.ciria.org DFEE (2000)., An Assessment of Skill Needs in Construction and Related Industries., Department for Education and Employment Publications, Nottingham. http://www.dfee. gov.uk. Egan Sir J., (1998)., Rethinking Construction: The Report of the Construction Task Force., Blackwell Science Ltd. Emerson H., (1962)., Survey of Problems before the Construction Industries, H.M.S.O., London. Goulding J.S., Sexton, M.G., Zhang, X., Kagioglou, M., Aouad, G., and Barrett, P. (2007)., Technology Adoption: Breaking Down Barriers Using a Virtual Reality Design Support Tool for Hybrid Concrete, Journal of Construction Management and Economics, Vol. 25, No. 12, pp. 1239–1250. Hurst A., Hodgkinson M., and Mutch A. (2007)., The Implementation of Continuing Professional Development for Construction Managers., Proceedings of the 7th International Postgraduate Conference, Salford Centre for Research and Innovation, SCRI, pp. 294–305. Kagioglou M., Cooper R., and Aouad G., (2001)., Performance Management in Construction: A Conceptual Framework, Journal of Construction Management and Economics, Vol. 19, No. 1, pp. 85–95. Koskela L. (2000)., An Exploration towards a Production Theory and its Application to Construction., Esppo. Technical Research Centre of Finland, VTT Publications PhD: 296. Latham Sir M. (1994)., Constructing The Team,. HMSO, London. Love P., Irani Z., and Edwards D. (2004)., A Seamless Supply Chain Management Model for Construction., International Journal of Supply Chain Management, Emerald. ManuBuild (2009)., www.manubuild.net McNair S., and Flynn M. (2006)., Managing an Ageing Workforce in Construction: A Report for Employers., Produced for the Department for Work and Pensions by the Centre for Research into the Older Workforce. Rezugi Y., and Zarli A. (2006)., Paving the Way to the Vision of Digital Construction: A Strategic Roadmap., Journal of Construction Engineering Management, 132 (7) pp. 767–776. Sharp M., Jones K., and Clarke R. (2005)., An Investigation into Inefficiency in the Condition Based Maintenance Process., In: Khosrowshahi, F. (Ed.), 21st Annual ARCOM Conference, 7–9 September 2005, SOAS, University of London., Association of Researchers in Construction Management, Vol. 1, pp. 385–92. Thompson S. (2007)., Creating Places: Sustainable Communities using an Open Building Manufacturing Approach, Transformation of the Industry—Open Building Manufacturing, ManuBuild. Yeomans D. (2000)., The Characteristics of Traditional Construction., Proceedings of the International Conference on the Seismic Performance of Traditional Buildings, Istanbul, Turkey, Nov. 16–18, 2000.

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Precast ferrocement barrel shell planks as low cost roof S.F. Ahmad Saudi Oger Ltd, Riyadh, Saudi Arabia

ABSTRACT: Ferrocement Barrel Shell Plank was designed as segmental element, which is very easy to cast, cure and then manually erect over low cost concrete blocks used as bearing walls. This innovative roof and floor element is very competitive in cost too. The finished roofs, of a typical low cost housing unit measuring a floor area of about 75 m2 (about 800 ft2 ), were cheaper by at least 25% as compared with any other roof constructed using the conventional building materials and techniques. A model low cost one-room housing unit and later a full-scale low cost housing unit was built for the purpose of demonstration. This paper discusses in detail the process of conception, design & production, cost economics, load tests and later construction of the model demonstration low cost housing units using this innovative technique. In Pakistan there is a huge scope of investments in low cost housing sector. 1

AN OVERVIEW OF HOUSING ISSUE

Providing affordable housing to low-income people has remained a challenge in Pakistan since long. Although some efforts have been made in the past, both in the public and private sectors, to mitigate the same yet the issue seems to be still out of hand. The phenomenal increase in the cost of land together with rising prices of building materials and increased labor cost, all of them are in fact making the goal of the low-income housing unattainable. In Pakistan historically three basic strategies were adopted to address the housing needs of low-income group. They are outlined as follows (Salim 1996). 1.1

Housing colonies

In this arrangement, the Government normally undertook the construction of large housing colonies, which were complete with all services. These were built in the outskirts of the cities and the inner city squatters were moved into such settlements. In order to finance such ambitious schemes, the State had to raise loans from national and international donor agencies at low interest rates. Beneficiaries were expected to repay the cost over a period of fifteen to twenty years. The strategy couldn’t work any longer as the squatters started to beeline, the payback was not up to the mark, and as such the donors were no longer forthcoming.

because of the exorbitant costs of such developed lands, which were beyond the reach and capacity of the low-income groups. As a result such schemes, namely the Metroville project in Karachi, remained either unoccupied for a long time or were purchased and occupied by the speculators or the middle class buyers for investment purposes. 1.3

The current strategy, in place since past two decades at least, is that the State itself has now turned into a developer. This essentially means that the projects are announced before they are physically commenced and the public is invited to apply for plots (piece of land) with an advance down payment and pay the rest in east installments, as the construction works on the plot proceeds. This form of development, though ostensibly meant for the low-income groups, in effect does not really cater for them or reach them because of the following reasons. 1. They want the land immediately and can not wait for many years 2. They do not know how to fulfill formalities and deal with the State offices. 3. The cost of building system is very high.

2 1.2

Site & services schemes

Due to failure of the strategy of building housing colonies the planners then aimed at selling lands, developed by official agencies, to the lower income groups. Such a strategy also could not make any headway

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State turns developer

HISTORY OF LOW INCOME PROJECTS

In private sector the low cost housing concept was first introduced by A-Azam Limited, which had launched the low cost housing society in Karachi during the early 60s and 70s. It had built very low cost flats (G + 4) in both Karachi and Hyderabad, the two major cities

in the south of Pakistan. Such types of accommodations had the merit that they were really affordable by low-income groups. During the 70s yet another enterprise called Rukunuddin Construction Company, following Al-Azam’s footsteps, introduced its own low cost projects. This company introduced low-cost prefabricated housing units, in addition to their lowpriced regular apartment projects. The quality and safety of their units, however, were generally viewed as a bit inferior. Maymar, yet another construction company, sprang up and entered the burgeoning housing market in the same period as Rukunuddin’s. They focused on residential flats, apartments and housing units. Unlike Rukunudin’s, they were probably the best ever built and thus were successful too. As demand for better housing facilities emerged and increased, Maymar then moved up market leaving behind lowcost housing. Entered Cellrock Company and Abidi Ltd in the early 80s. They focused on low-cost prefabricated elements; Cellrock Company into light weight concrete prefabricated panels while Abdi Ltd into normal weight concrete prefabricated panels. The costs of such projects were high and beyond reach of the low income groups and more over, the quality were also questionable. As a result, such projects could not hit the ground as well. In Karachi, both Al-Azam Ltd and Maymar were able to survive and hence evolve primarily owing to the patronage from a local government institution called KDA (Karachi Development Authority). Sensing the demand and need for low-cost housing units, KDA established a wing by the name of Public Housing Scheme in the early 80s and thus entrusted them both with the task of making new low cost development schemes for the city of Karachi. Surjani Town project was thus envisaged and cheaper lands were allotted to these two builders first and later on to many more others. This also saw allotments of such lands to many non-professional builders via corrupt officials. The corrupt and non-professional builders thus contributed in damaging this good scheme by building low quality units. The Public Housing Scheme was in the meanwhile hit hard by the collapse of the utility companies who failed the challenges of providing them the utility connections in time. Thus the last hope of low-income people getting good and cheaper housing, with safety of their investments, was also dashed with the closure of this wing of Public Housing Scheme. All said and done, the one more important point, which was one of the major reasons of failure of this scheme, was the lack of innovations in both the materials and the design of the structural systems used in these schemes. These old and new builders involved in Surjani Town Scheme never consulted any building research institutes to help them with new and innovative designs so as to reduce the cost and hence attract the buyers. They continued to adopt the age-old designs and as such there was nothing new in them to attract the buyers.

National Building Research Institute (NBRI)—a building research organization under the aegis of Council for Works and Housing Research (CWHR) (of ministry of science and technology) cognizant to the requirements of the construction industry, thus took up the challenge of conducting researches into alternative and cheap building materials and building systems. This was so as to provide the necessary technical supports and assistances to these builders in producing cheaper and safer housing units. Several research programs were thus undertaken to meet this objective. Since the load-bearing walls and the roof elements contributed towards major elements of the costs, in a typical housing unit, main researches were thus focused in these two fields alone. Precast Ferrocement Barrel Shell Plank as Low Cost Roofs is one such proud research accomplishments of NBRI (Akhtar 1990). 3

FERROCEMENT BARREL SHELL PLANK

Ferrocement is a well-known and time-tested building material. It is basically a composite material wherein cement, sand and water are mixed in appropriate proportions and is often reinforced with wire mesh or steel fibers with a reduced spacing. Reduced spacing helps yield uniform force dispersion and thus increases its strengths. Besides strengths, it also offers great economy owing to thin sections. Ferrocement Barrel Shell Plank (FBSP) is a segmental element, which is developed at NBRI to be used as roofs and floors in low-cost housing projects. The principal purpose of the same was to reduce cost, as roofs/floors contribute a major cost in a typical low cost house. 3.1

Description of FBSP

It is a simple technique that has been developed at NBRI/CWHR as an alternative structural roof/floor element to replace high cost conventional systems. This is for the purpose of reducing the cost of a low cost housing unit and thus makes it affordable to the low-income groups. This is a thin-shell Ferrocement element, which has a thickness of 16 mm only in the main body while the ends, and a rib in the mid-span is thickened to 50 mm. This is done for the purpose of increasing stiffness. The width in cross-section is 375 mm while the polar height (height of the crown from the base) is 200 mm. The FBSP with this cross-section can safely take a span of 3 m to 4 m, which is normally the span, in shorter directions, of a typical low-cost housing unit. The weight of one such segmental element is about 95 kg, which can easily be lifted by two persons. The finished roof of such geometry of FBSP can easily sustain a design live load of 2.5 kN/m2 .

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3.2

Construction technique

The construction technique is very simple. These are cast in a simple mould/form that can be constructed out of wood and GI-sheets as shown in Figure 1. They can also be made out of simple mud/clay by a semi-skilled laborer, even in villages. They can be made at the jobsite on a long-line bench made from mud/clay. The mud/clay formwork needs to be lined with polyethylene sheets so that the mud/clay form does not absorb water from the mortar. First of all, the mould/form surface is applied with some oil. The wire-mesh is then carefully laid on the surface of the mould/form. After this preparation is over, cement-sand mortar is prepared in a ratio of 1:3, with appropriate amount of water used. The mortar is then carefully applied over the wire-mesh to the desired thickness of 16 mm. In Figure 1 below is shown a typical mould/form to be used for casting of a single plank of FBSP. After mortar application is over, the entire formwork is then wrapped with hessian cloth and then allowed to remain wet for curing purpose. After three days of curing, the FBSP is then manually lifted as shown in Figure 2 and stacked over bricks/blocks at two ends to allow for further curing. 3.3

Figure 2.

FBSP being lifted from formwork after curing.

Figure 3.

Men standing over FBSP after load testing.

Load testing of FBSP

After the curing period is over, a load test arrangement is normally erected. In this case first the segmental elements of FBSP are lifted manually and placed side by side on a wall arrangement built specially for the purpose. The top of FBSP is then filled with screed in order to have a level top. Under the midspan of the FBSP, a strain gauge is fitted. The top of FBSP is then gradually loaded with either stone aggregates or concrete blocks. This amounts to 1.5 times the design Dead and Live loads to be superimposed during its service life. The total loadings thus

computed are loaded in three incremental stages and after each stage of loading, the induced deflections of the system are recorded. Subsequently while unloading, the same is also done in three stages and rebound recorded accordingly. The total induced deflection is then compared with the theoretically allowable deflection. In all cases, the induced deflections were much lesser than what was the allowable value. The load test arrangements can be seen in Figures 3 and 4. Three individual FBSP were tested while two large test setups were also tested. The data from the load test are given as follows:

Figure 1.

– Span of FBSP tested = 3.5 m – Test loads = 1.5 × (Dead + Live loads) – Allowable deflection = L/360 = 3.5 m × 1000/360 = 9.72 mm – Total deflection recorded = 4 mm.

A wood-GI steel formwork for FBSP.

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Table 1.

Material break-up and cost of FBSP system.

Using FBSP Cement:Sand = 1:3

Using Concrete Solid Slab

Item

COST BREAKUP FOR EACH FBSP (3.5 m)

Cost of Roof/floor System Per sq. ft/sq. m

Cost of Roof/floor System Per sq. ft/sq. m

Cement Sand Wire mesh

Rs. 63.00 Rs. 9.00 Rs. 69.00

Rs. 21.00 Per sq. ft/ Rs. 226.00 Per sq. m

Rs. 30.00 Per sq. ft/ Rs. 326 Per sq. m

Note: Rft = Running feet; Rm = Running meter.

Figure 4.

A finished load testing arrangement for FBSP.

Figure 6. built. Figure 5. FBSP.

3.4

The full-scale model demonstration house being

A single room model house constructed using

Cost economics

The cost economics of this innovative structural system is quite favorable and compares well with any other conventional systems in use. The most common roof/floor system used in low-cost housing unit is a solid slab of thickness varying from 100 mm to 120 mm, depending upon the spans, of course. This innovative structural system employing FBSP was established to be cheaper by at least 25% compared to its nearest competitor system. The material break up together with their costs per single unit of FBSP is presented in a tabular form in Table 1. This is based on the prevailing market prices

at the time of research on the same. The cost is in Pak Rupees (Rupees or Rs is the local currency) and the conversion rate against US dollar at that time was: 1 US $ = 42 Pak Rs. 3.5

Applicability

Ferrocement Barrel Shell Plank (FBSP) is easy to cast, light in weight and quick & fast to erect. Even laborers with minimal skill can learn to cast and erect the same. They are cost effective also and offer a reasonable amount of heat insulation. As such, they are very suitable for use as structural roof and floor elements in typical low-income housing units. Small investors and small-scale contractors can set up their own casting yards and sell them on commercial basis also and thus earn a decent livelihood.

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3.6

Demonstration housing unit

4

In order to disseminate this research and to demonstrate its effective application in a low cost housing unit, first a single room demonstration house was built inside the office compound of NBRI in Karachi. Figure 5 shows a front view of this single room house with FBSP as the roof element. Later on, a full-scale low cost housing unit was also constructed on a land measuring a floor area of 75 m2 . This was constructed on a donated land free of cost with an idea of demonstration to the inhabitants of the area about the efficacy and efficiency of the FBSP units as roof elements. The under construction low cost demonstration housing unit is being shown in Figure 6 while the finished unit is shown in Figure 7.

CONCLUSIONS AND RECOMMENDATIONS

Ferrocement Barrel Shell Plank is efficient roof/floor element with lots of advantages that make it a viable option to be used in low cost housing units. It is in fact an answer to the increasing cost of roofs and floors that are off setting the cost of a low cost housing unit. It is simple to cast and even a low skilled laborer in a village can easily do it. It is easy to erect on load bearing walls. Moreover, it is cost effective and is cheaper by 25% as compared with any other conventional type of roofs. It can be produced by small investors on a commercial scale also and sold to the small contractors and builders of low cost houses or directly to the owners of the house, like concrete blocks and clay bricks are sold. In Pakistan, millions of low-income groups do not have house of their own and there is a huge potential in investment in low cost housing sector. ACKNOWLEDGMENTS Acknowledgements are due to Dr. Ataullah Maher, the Director General of NBRI for his continuous guidance and assistance towards completion of this research project. REFERENCES

Figure 7.

Akhtar, H.K. 1990. House building by low-income families. OPP Manual., Karachi, Pakistan. Hendry, A.W. 1981. Structural brickwork. McMillan press, London, U.K. Oktay, U. 1981. Construction of lower cost housing. John Wiley and Sons, London, U.K. Perween, R. 1990. Low cost housing model. OPP manual, Karachi, Pakistan. Salim, A. 1996. Low cost housing program of Orangi Pilot Project, Karachi, Pakistan.

The finished model house using FBSP as roofs.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Tall building boom – now bust? I.R. Skelton & D. Bouchlaghem CICE, Department of Civil and Building Engineering, Loughborough University, UK

P. Demian Department of Civil and Building Engineering, Loughborough University, UK

C. Anumba Pennsylvania State University, USA

ABSTRACT: From early 2006 up to the freeze induced by the worlds faltering financial markets during the first quarter of 2008, Britain experienced demand for tall buildings of an unprecedented high level—in London alone, ten tall buildings have started, or are due to start on site between first quarter 2007 to fourth quarter 2008. This is directly comparable in size to America’s Manhattan Island skyscraper boom of the 1920’s. The objectives of this paper are: Firstly, to investigate the evolution of the UK tall building construction and determine the reasons behind its growth at previously unprecedented rates; Secondly, to create a definition of the UK tall building and compare it to the international tall building stage; Thirdly, to analyse the differing types of demand and define these sub sectors of UK tall building market; Finally, to calculate the size and value of this specialist construction market, forecast its growth potential and model it against the Skyscraper Index. 1 1.1

INTRODUCTION Key findings

The key findings of the research undertaken for this paper are: The tall building form is not a passing design trend in the UK, it is here to stay and is currently backed from the upper echelons of Central Government down to popular public opinion; The UK tall building is defined in this research as twenty stories plus, due primarily to the change in building methodology required. However, this only equates to mid-rise on the international tall building stage; Several new tall building clusters are being encouraged in London by Central Government, the existing London Plan, CABE and by unsatisfied demand from the office, mixed use and residential sectors; The three biggest threats to the continued growth of the UK tall building market are the current US-led economic slump, UNESCO’s recent pressure to stop tall buildings being constructed close to British heritage sites and the new London Mayor’s potential reversal of the current protall building stance, expressing disapproval of tall buildings that block historic views; There are four types of UK tall buildings, driven by four distinct areas of demand, creating four sub sectors of the market: the ‘fat’ office tower (18% of market demand); the ‘thin’ or ‘iconic’ office tower (36%); the mixed use tower (18%) and the residential tower (28%). In the fourth quarter of 2007, London had thirty nine tall buildings potentially reaching site in the next

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three to five years, the South East had eight and the balance of the UK had thirty. The total estimated net trade cost of which is £9.77 billion, directly equivalent to the construction budget for the 2012 Olympics; The average gestation and build period for a UK tall building is eight years, made up of an average of five years preconstruction and three years build period; The average view of the many independent economic forecasts considered in this research prior to December 2007 for the years 2008 to 2012, was one of sustained growth for both the commercial and residential drivers of the tall building market, with an increasing focus on mixed use towers as the most efficient and sustainable tall building format. This positive economic outlook has subsequently been tempered throughout 2008 by the US led economic slump, now reducing the UK’s commercial and residential markets growth potential; The UK tall building boom overlaid with 2006–2007 financial market buoyancy and the current economic uncertainty are a mirror of the model conditions presented in the Skyscraper Index (Lawrence 1999), a tool used to forecast the economic downside of building tall. This infamous index historically demonstrates that tall building construction follows the peak of a country’s economic cycle and is followed by a significant economic slump. 2

EVOLUTION OF THE TALL BUILDING

The skylines of many world cities are defined and punctuated by tall buildings. The drivers for such

dominant skylines range from land scarcity and social needs, high real estate values, commercial opportunity and corporate demand, through to metropolitan signposting (CTBUH 2006). This obsession with the current form of tall, slender buildings can be traced to the Italian patrician families who created the 11th Century skyline of San Gimignano by building seventy tower-houses, some fifty metres tall, as symbols of their wealth and power. This was most famously followed in the late 19th Century with the Manhattan skyline. This obsession with building tall continues to spread worldwide and is forecast in this research to grow into the future. Even after the events of 9/11, the tall building format is very much here to stay. The development of increasingly sophisticated construction materials and technologies has driven the evolution of the modern tall building throughout the 20th and early 21st century. The resulting buildings have reflected this evolution in height, but the form has generally adhered to one of two design philosophies; straight up or stepped, due to light-protecting planning laws. However, the new ‘iconic’ breed of innovatively designed tall buildings brings unprecedented challenges to the developers, designers and not least, the builders: commercial feasibility must be achieved; technological obstacles must be overcome; cutting edge design must be converted into a built reality; safety of its builders and occupants must be ensured; risk of cost and programme overrun must be minimized—tall challenges for the tall building industry to surmount. 2.1

The rise and rise of tall buildings in London

Britain’s experience of tall buildings has been blighted by post-war regeneration during the 1950’s to 1970’s which produced a large number of local authority housing towers and brutalist office towers between ten and thirty storeys high. The high profile failure of many of these post war experiments was due to weak design, detailing and construction, leading to a general rejection of the high rise form in the 1980’s. There were a few exceptions to this rule in the first generation of UK tall buildings, the most successful of which have now achieved listed building status, including Centrepoint (Grade II), BT Tower (Grade II) and the Trellick Tower (Grade II*). Over the last ten years, general interest in tall buildings has risen to new heights, both in the commercial and residential sectors. This is evident on the supply side, especially in London, where a pro-tall building stance is notable in: The London Plan (GLA 2004); the number of planning proposals submitted for tall buildings; the granting of planning consent to new proposals such as Heron Tower, The Shard in Southwark, 122 Leadenhall Building, 20 Fenchurch street, DIFA (Bishopsgate) Tower and Columbus Tower; the recent completion of numerous tall buildings in

various London locations such as Paddington, The West End, The City (London’s Financial District) and Canary Wharf and the successful refurbishment of first generation tall buildings including Tower 42 and City Point. In response to this favourable profile, signature architects are now scrambling to design tall buildings (Strelitz 2005). This high profile is also reflected in the demand side. There is now a strong City ambition to build high. This has grown from Foster and Partners formbreaking design for the Swiss Re building, 30 St Mary’s Axe, which won support from Commission for Architecture and the Built Environment, English Heritage and the City, all of whom were keen to secure bespoke headquarters for major commercial institutions (Morrell 2006). The continuing high profile success of Swiss Re’s 30 St Mary’s Axe undoubtedly led to increasing demand for more commercial towers (Linklaters 2002). The Heron Inquiry followed, forcing the evolution of city policy, developing the concept of an ‘Eastern Cluster’of tall buildings in the city, not affected by St Paul’s Cathedral Heights, Grid, Strategic Viewing Corridors (GLA 2001), or Conservation Areas. The forthcoming 50 storey 122 Leadenhall Building currently being built by Bovis Lend Lease, will become the focal point of this new tall building cluster. The rising profile of London as a ‘World City’ (LPAC 1998) over the past decade, allied to the refocus of the planning system for high density developments and brown field schemes, has assisted this growth in building tall. London, now seen as the de facto capital of Europe, is consolidating its position as a world leading financial centre, second only in trade value to Frankfurt. The new London Mayor believes London is now challenging Tokyo and New York as their only global competitors. Planning policies laid down by the previous Mayor underpinned this vision, permitting the provision of world class office space and infrastructure and encourages ‘London to continue to reach for the skies’ (Livingstone 2001). Research undertaken for this paper shows the current suite of tall buildings being constructed in the heart of the City, (122 Leadenhall, The Shard of Glass, Heron Tower, 20 Fenchurch Street, The DIFA/ Bishopsgate Tower and the Broadgate Tower) were all commissioned due to the threat of London Docklands on the City’s position in the mid 1990’s. The City responded by relaxing plot ratios to encourage development. The reaction to a ten year old threat is now finally hitting the streets, even though the threat is long gone as Canary Wharf has been 90% full for the last three years. 3

DEFINING A UK TALL BUILDING

The second objective of this paper is to define a UK tall building and compare it to the international

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tall building stage. This research has defined the UK tall building as between twenty to sixty stories, circa seventy to three hundred meters (depending on whether the building has commercial or residential floor to floor heights). A generally accepted definition of a tall building in town planning terms is one which stands above the prevailing skyline. A good construction definition has been determined as a building which has technical and design differentiation from its neighbours. Even with modern build methods, above twenty stories a building becomes technically distinct in its structure, services, vertical circulation, life safety and cost. Therefore, above twenty stories they deserve a different classification—the UK tall building. Research undertaken for this report shows that of the seventy seven UK tall buildings currently proposed, almost 70% by value are in London. This report therefore focuses on London, but also considers the South East and the balance of the UK (‘Other Regions’). 3.1

London high rise = global mid rise

London’s skyline is predominantly low rise with distinct pockets of medium to high rise, allowing space for the St Paul’s Cathedral sight lines to strategic London viewpoints (Planning Act, GLA 2004). A tall building in the London context is considered by this research to be twenty to sixty storeys. At the top end of the London scale are the proposed London Bridge Tower and The Bishopsgate Tower in excess of three hundred meters high, containing eighty plus storeys. The lower end of the London scale is dictated by the need at this height for technological changes to the way buildings are constructed, utilising tall building techniques as opposed to low rise construction techniques. New York’s Manhattan Island is widely recognised as one of London’s main competitors for the status of financial centre of the world. Its skyline, by comparison to London’s, is predominantly medium rise with widespread pockets of high rise. A tall building (skyscraper) here is deemed to be thirty to one hundred plus stories (although local fire codes change at fifteen storeys). In the last seven years America’s appetite for commercial tall buildings has cooled, but residential demand remains strong and international developments are beginning to influence corporate decision-making in New York, especially regarding sustainable design (Fuerst 2007). The future of the skyscraper seems assured in New York City, even after the soul-searching in Manhattan after the loss of nearly three thousand lives in the Twin Towers disaster. Tokyo is the second main competitor to London for the status of financial centre of the world. The Asian skylines of central Tokyo, Hong Kong and Shanghai, in comparison to London’s, are predominantly high rise with isolated pockets of low rise on the peripheries. Asia is regarded as the natural environment of the very

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tall building; the format makes sense as the density and the urban infrastructure make it the logical way to occupy land. High density is a historically accepted norm in much of Asia. Many towers are simultaneously going up in Hong Kong, Guangzhou, and the other high-growth Asian cities. The tallest, densest buildings are rising over rail stations, with airport access (Willis 2007). With the massive Chinese population rapidly industrialising in a modern version of Victorian Britain, there is a boom in tall buildings on an unprecedented scale. China currently has around sixty, three hundred metre plus buildings, or ‘supertalls’, at some level of development. This research found that the UK tall building is therefore defined on the international stage by its more conservative height, individual architectural approach to the internal and external form, its response to its non-regular site footprint and the heritage of the surrounding city-scape. These local factors generally result in high quality, individualistic tall building designs demanding high quality building solutions. Modularity and repetition are not seen as UK tall building traits, resulting in a more costly tall building solution. 4

WHO’S DRIVING DEMAND FOR LONDON TALL BUILDINGS?

This research has determined there are four distinct occupiers of tall buildings in London, driving demand for four different types, or sub sectors of tall buildings: The first tall building sub sector is represented by large corporations wishing to relocate in a single building, requiring a ‘fat’ tower with large floor plates of 3000 m2 gross and up to 50000 m2 total area. This is usually a planned amalgamation of various operations, aimed at creating synergies and savings between business units, whilst reducing facilities management costs. Examples include HSBC, Citygroup, Barclays and most recently, JP Morgan, who will move to Canary Wharf as it offers the right mix of floor plate, quality of space, size and critical mass of complimentary businesses. These types of offices are now achieving an average rent of £70 per sq ft across the seven London fringes (EGi 2007). The second tall building sub sector is represented by small international companies demanding a prestige location in a multi-tenanted, thin or ‘iconic’ building. Their floor plate requirement is between 1000–2000 m2 gross. They value prestige, high quality, shared facilities and opportunities for interrelations with neighbouring businesses. The demand of this type of occupier is shown by low vacancy rates and high rental yields for these iconic buildings. These types of offices are regularly achieving £100 per sq ft across London, a new record set during the fourth quarter of 2006 (EGi 2007).

forms almost 70% of the current UK tall building market by value, but also considers other areas of the UK.

London's Proposed Tall Buildings

18% 28% Office (Fat) Office (Iconic) Mixed Use Residential 18%

Figure 1.

36%

London’s Proposed Tall Buildings, Q4 2007.

A third, emerging tall building sub sector is the mixed use tower, incorporating a mix of residential, retail, office and possibly hotel and leisure space. This tall building form shows signs of increasing its tall building market share as they are inherently efficient with higher densities, complimentary structural requirements and potential heating and cooling shares between different occupiers systems. They are generally located over, or close to public transport hubs and are generating an image of being an efficient and sustainable tall building solution. The fourth tall building sub sector is the tall residential market, rapidly growing in popularity due to high potential returns on investment. The renaissance of residential tall buildings is due to increasing house prices outstripping build cost inflation along with the rising profile of ‘city living’. This has lead to a relatively new phenomenon of a price premium relative to the height of the residential development. Research undertaken for this paper show the fourth quarter 2007 tall building market sub sector split for London is: 54% commercial towers (18% fat and 36% iconic); 28% residential towers and 18% mixed use towers, shown graphically in Figure 1 below: The sum of these four tall building sub sectors show that London’s demand for tall buildings was at an unprecedented level in December 2007. The UK construction industry now waits to see the impact of the US led economic slump throughout 2009. 5

METHODOLOGY OF THE MARKET ANALYSIS

The market analysis undertaken for this paper created a unique snapshot of the UK tall building market in the fourth quarter of 2007. It captured the market’s mood, categorised the types of demand, determined the market’s current value and was then used to forecast its growth. The full picture of the tall building market presented in this analysis was built up from a blend of new data generated during this research, live information gained from industry recognised expert sources via targeted interviews and questionnaires, plus in house theoretical and practical construction market knowledge. The analysis concentrates on London as it

5.1

Cataloguing current UK tall buildings

A catalogue of all proposed UK tall buildings was compiled to determine the size of the current tall building market in the UK, the type of tall building, the proposed height above ground, the feasibility of actually being built and also captured construction cost information if available. This catalogue of tall buildings was then filtered to include only those tall buildings deemed to be feasible of reaching site within three to five years and excluded ‘visionary’ tall buildings with a low likelihood of being built. The catalogue contains seventy seven proposed UK tall buildings, which have been broken down geographically into ‘London’ (which has thirty nine), the ‘South East’ (which has eight) and the ‘Balance of the UK’ (which has thirty) and then sorted into the previously determined four tall building sub sectors: Commercial (Fat), Commercial (Iconic), Mixed Use and Residential for each geographic area. This breakdown is shown in Figure 2 below. 5.2

Calculating current UK tall building market value

Definitive or reliable construction cost information can rarely be found for the majority of projects due to the extremely confidential nature of finance for tall buildings, therefore, the shell and core construction costs of five Bovis Lend Lease tall building projects were selected. These tall building projects were selected on the basis of being a London project that would feasibly enter the construction phase within the next three to five years, having had the cost plan checked for robustness during 2007, the projects consisting of competitively tendered packages under a construction management contract arrangement and the type of projects proportionately representing the four previously determined UK tall building market sub sectors. The project details of the five selected tall buildings are withheld due to the confidential nature of the project information. These figures were supplemented by published details for a sixth tower, The London Bridge Tower, or Shard. Demand For UK Tall Buildings

15 Office (Fat)

10 Number of Tall

Office (Iconic)

Buildings Proposed 5

Mixed Use Residential

0 London

South East

Balance of UK

Geographical Market Area

Figure 2.

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Demand for UK Tall Buildings, Q4 2007.

An average of the six sample tall building net trade shell and core construction costs per meter height above ground was taken. This representative ‘average’ tall building shell and core construction cost was then multiplied by the cumulative height of the filtered catalogued of tall buildings for each geographical location in the UK, which was then factored utilising published tall building cost location factors (London and the South East cost index 1, the balance of the UK an average cost index of 0.92) (Watts 2007). This results in the calculation of the total value of the UK tall building construction market. (NTC = Net Trade Cost, in this calculation is defined as a tall building shell and core construction cost, excluding demolition and enabling works, external works, incoming services and fitting out, developer’s professional and statutory fees, taxation, insurances, finance charges, disposal costs and design and construction related professional fees, VAT and any site abnormalities). Tower 1, NTC £253 m/222 m height = £1.14 m/m Tower 2, NTC £181 m/161 m height = £1.12 m/m Tower 3, NTC £80 m/100 m height = £0.80 m/m Tower 4, NTC £257 m/160 m height = £1.60 m/m Tower 5, NTC £55 m/140 m height = £0.39 m/m Tower 6, The Shard NTC £350 m/310 m height = £1.13 m/m Average construction (NTC) cost per m height = £1.03 m/m height Therefore, total construction value for tall buildings in London and the South East = cumulative height of catalogued tall buildings × average NTC/m height = (6968 m London + 422 m South East) × £1,030,000/m × 1 (Location Factor London & SE) = £7,611,700,000 Using the same method of calculation, the total construction value of tall buildings in the balance of the UK (Location Factor 0.92) = 2277 m × £1,030,000/ m × 0.92 = £2,157,700,000 Therefore, the total construction value for the current UK tall building market = £9,770,000,000 To determine the potential UK tall building construction value per year, the average gestation period of a UK tall building needed to be determined. By analysis of the 20 most recently awarded UK tall buildings for construction up until the fourth quarter of 2007, the average UK tall building gestation period (from project planning proposal date to completion of construction date) has been calculated as eight years. This consists of an average period of five years for preconstruction (from project planning proposal date to start on site date) and three years for construction (from start on site date to completion of construction date). Due in part to the current uncertain economic climate, it is not certain when any building on the tall building list will progress from planning and preconstruction into the construction phase and hence

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generate potential construction spend for that year, so an equal spread over the average gestation period of eight years must be assumed. This gives an average annual construction spend (construction market value) of £1,221,200,000 for the UK tall building market from 2007 until 2014 inclusive. 6

DISCUSSION

This forecast construction value for the UK’s tall building market is of a scale directly comparable to the latest Government declared construction budget for the 2012 Olympics of £9.325 billion pounds over seven years (2006–2012), but if nurtured, could be a sustainable value source, not a one off event. It is recognised that the London Olympic win has increased delivery pressures on the current swath of tall buildings. 2012 has become an artificial deadline for a large number of major projects, which will cause consolidation of work (Thompson 2006). There is concern that the simultaneous start of construction of a significant number of these tall building projects, running concurrently with the Olympics, will overheat the construction market, creating local shortages of skilled labour and materials and force prices up by factors of up to 20% for steel reinforcement and concrete. Davis Langdon’s Market Forecast Report for 2007 summarises that the top and bottom of London’s construction market is polarising, whereby large projects such as tall buildings are suffering from greater inflationary pressures, while smaller schemes retain a more competitive edge (Davis Langdon 2007). This forecast of rising construction costs has not noticeably dampened the demand for the UK tall building, but this may be due to its effect being swamped early in 2008 by the US’s economic uncertainty and risk of global recession, the effects of which are now being seen in the UK tall building market by the stalling of some speculative developments and a requirement for higher pre-let percentages prior to construction start. Bringing this forecast up to date with recent economic developments as of fourth quarter 2008, the amount of commercial development in the UK has fallen to its lowest level in five years (Savills 2008) and is directly attributed to tighter bank lending conditions, deteriorating market sentiment and weaker growth prospects for the global economy. This has affected three of the four tall building markets (fat office, iconic office and mixed use), the residential sector being separately affected by the current UK housing market stagnation caused by a lack of liquidity in the mortgage market. The tall building boom presented here, overlaid with 2006–2007 financial market buoyancy and the current economic uncertainty are a mirror of the model conditions presented in the Skyscraper Index

(Lawrence 1999), a tool used to forecast the potential economic downside of building tall. When London’s tall building market conditions determined in this research are overlaid, there is an almost perfect match. This infamous index historically demonstrated that tall building construction follows the peak of a country’s economic cycle and is followed by a significant economic slump (Thornton 2005). This index was previously thought to be unable to predict the UK tall building market as it was based on American economic cycles and whilst it’s logic stood for Americas historic cycles, it unsuccessfully predicted the last two American economic slumps (the last of which was 9/11 driven), due to changing investment criteria and expectations since the index was conceived in 1999. However, this research shows that the recent history of London’s tall buildings shows strict correlation to the Skyscraper Index. London’s office market suffered downturns in 1974, 1982, 1990 and 2002. The two most recent falls were marked by the construction of London’s best-known skyscrapers: the Canary Wharf Tower in 1991 and 30 St Mary Axe in 2003. As previously proposed, the average gestation period (from proposal to completion) for a UK tall building is eight years and each economic cycle lasts for some ten years. This makes it virtually impossible to get the timing right on tall buildings (Damesick 2008). It is apparent from this research that the current suite of tall buildings being constructed in the heart of the City (122 Leadenhall, The London Bridge Tower, Heron Tower, 20 Fenchurch Street, the Bishopsgate Tower/Pinnacle and the Broadgate Tower) were all commissioned in the mid to late 1990’s and are due for completion between 2008 and 2011. If the Skyscraper Index is to be believed, the current uncertainty in the UK economy will degenerate into a full blown recession as these tall buildings near completion over the next few years. 7

CONCLUSIONS

Tower cranes across the London’s skyline have traditionally been a highly visible measure of health of the construction industry as well as an accepted indicator of the strength of the UK economy as a whole. If the view from the City to St Paul’s is unblemished by Wolff ’s, Liebherrs and Potains, then a slump is on the horizon. (Morby 2007). If this indicator is to be believed, then the London tall building market is thriving as almost thirty tower cranes were counted on 1st January 2009 from London’s St Pauls. At the other end of the forecast spectrum, if the Skycraper Index is to be believed, then both UK tall buildings and whole economic outlook is dire. Arguably, one of the world’s most enduring famous tall buildings, the Empire State Building closely followed the Index’s prediction. On completion, it was nicknamed the Empty State Building due to its low occupancy

rates until after World War II. Sales publicity for the building claimed the feeling of looking out from its viewing gallery was better than air travel. It was not publicised that the viewing platform was only built because the office space could not be sold (Garcia 2006). Will the UK’s tall building momentum stall, will London’s skyline soon be host to a myriad of skyhigh, empty viewing galleries, or will the UK’s new tall buildings continue ever upwards, unbending in the current economic storm?

REFERENCES CTBUH June 2006 Chicago International Conference. Thinking outside the Box—Tapered, Tilted and Twisted Towers. Damesick, P. 2008. CB Richard Ellis Research website. Davis Langdon, 2007. Market Forecast for 2007, Published 31st October 2007, Building Magazine. London. EGi’s LONDON OFFICE DATABASE. Market Analysis, Issue 4, Oct–Dec 2007. Estates Gazette. Fuerst, F. 2007. Manhattan’s Financial Firms, Three Years After. Research Associate at City University of New York’s Graduate Centre. Russell Sage Foundation. Garcia, J M. 2006. The Skyscraper Boom—Better Than Flying. The Economist Print Edition, Dubai, London and New York. June 1st 2006. GLA, Greater London Authority, 2001. Interim Strategic Planning Guidance on Tall Buildings, Strategic Views and the Skyline in London. GLA, Greater London Authority, 2004. The London Plan. Lawrence, A. 1999. The Curse Bites: The Skyscraper Index Strikes. Property Report, Dresdner, Kleinwort, Benson Research (March). Linklaters, 2002. Tall Office Buildings in London, Guidance on Planning, pp. 10–11. BCO 2002. Livingstone, K. 2001. Foreword to: Interim Strategic Planning Guidance on Tall Buildings, Strategic Views and the Skyline in London. GLA, Greater London Authority, 2001. LPAC, London Planning Advisory Committee, 1998. High Buildings and Strategic Views. A Study for LPAC by Building Design Partnership, London Property Research, London Research Centre and Ziona Strelitz Associates. Morby, A. 2007. Confidence is Soaring. Construction Forecast 2007, Construction News, Emap Construct Ltd. Morrell, P. 2006. The Economics of the Unusual. Director, Davis Langdon & Seah International. CTBUH Talking Tall Conference, 2006. Savills and NTC Research, 2008. Commercial Development Activity Index. Property Week June 2008. Strelitz, Ziona. 2005. Tall Buildings, A Strategic Design Guide. The British Council for Offices & RIBA Publishing. ISBN 1 85946 168 9. Thompson, 2006. Director, SKM Stadium. Stadiums, New Civil Engineer, Sept. 2006. Thornton, M. 2005 Skyscrapers and Business Cycles 2005. The Quarterly Journal of Austrian Economics, V8, No. 1 (Spring 2005). Watts, S. 2007. Tall Buildings—Cost Model, April 2007. Davis Langdon, London. Willis, C. 2007. Skyscrapers and Skylines in New York and Chicago. Founder and Director of The Skyscraper Museum. <www.skuscraper.org>.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

The state-of-the-art of building tall I.R. Skelton & D. Bouchlaghem CICE, Department of Civil and Building Engineering, Loughborough University, UK

P. Demian Department of Civil and Building Engineering, Loughborough University, UK

C. Anumba Architectural Engineering, Pennsylvania State University, USA

ABSTRACT: Following on paper titled ‘Tall Building Boom-Now Bust?’ which concluded that Britain’s recent demand for tall buildings was of an unprecedented high level, directly comparable in size to Americas Manhattan Island skyscraper boom of the 1920’s, but that the construction market was ultimately heading for a recession, this second paper determines the global state-of-the-art of building tall buildings. This has been achieved by designing a questionnaire which captures the most pressing issues of the tall building process, targeting the questionnaire at the most active specialist tall building professionals around the globe, then delivering these questionnaires face to face, resulting in an 80% response rate. The results give great insight to the consensus of professional opinion across the globe and across the specialist sectors of the industry. This paper investigates five key areas: current state-of-the-art of the international tall building industry; build process of a tall building; tall building principal contractor key attributes; ‘wins’ and ‘losses’ inherent with building tall and new techniques from overseas or other industries. 1 1.1

INTRODUCTION Preceding research

The paper by the same author titled ‘Tall Building Boom—Now Bust?’ established that Britain experienced demand for tall buildings of an unprecedented high level from late 2006 to the financial freeze of late 2008. During this period in London alone, ten tall buildings have started, or were due to start on site. This is directly comparable in size to America’s Manhattan Island skyscraper boom of the 1920’s. The first paper investigated the evolution of the UK tall building and determined the reasons behind its growth at previously unprecedented rates; it created a definition of the UK tall building of twenty stories and above, which compares to the international tall building stage as mid-rise; it determined the average gestation and build period for a UK tall building as eight years, made up of an average of five years pre-construction and three years build period; it deter-mined the three biggest threats to the continued growth of the UK tall building market; it analyzed the differing types of demand, defining four distinct sub sectors and calculated the size and value of UK tall building market as £9.77 billion, directly equivalent to the construction budget for the 2012 Olympics; it concluded by forecasting the UK tall building market growth potential and modeled this against the Skyscraper Index (Lawrence 1999),

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resulting in an almost perfect match (this infamous index historically demonstrates that tall building construction follows the peak of a country’s economic cycle and is followed by a significant economic slump). This second paper follows on from this work and sets out to establish the state-of-the-art of the international tall building industry, concentrating on the four main geographical areas of Europe, UAE, USA and Asia Pacific. This has been achieved by: firstly, undertaking a series of interviews and pilot questionnaires with targeted tall building industry specialist, then utilizing these findings to design a questionnaire which tackles the most pressing issues of the tall building process; secondly, by targeting the questionnaire at the most active specialist tall building professionals from each key discipline around the globe; thirdly, by delivering the questionnaires face to face, thereby gaining an 80% response rate and over 150 valid responses; fourthly, by analyzing the responses across all geographical areas and industry disciplines, giving insight to the professional opinions ranging from Dubai to London, Shanghai to Chicago, Sydney to Tokyo, Holland to Vietnam. This paper will be followed by a more in-depth analysis of the results, correlating industry specialist sector per geographical region, drawing out contrasts and trends between specialism and geographical location in the global tall building industry and isolating areas of global innovation in tall building construction that

could be beneficially applied to the UK tall building industry. The objective of this paper is to investigate five key areas of the global tall building industry: The current state-of-the-art of the international tall building industry; The international build process of a tall building; The tall building principal contractors key features; Wins and losses inherent with past tall building projects; New techniques from overseas or other industries. The analyzed results lead to some surprising conclusions, but offer a clearly signposted way ahead for the innovative construction of tall buildings. 1.2

Pilot interviews and questionnaire development

The research for this paper initially involved understanding the specific issues associated with building tall, firstly on a UK basis, then expanding this to a global view. This stage commenced with a literature review, followed by targeted structured interviews held with the four most prolific tall building principal contractors in the UK. These interviews gave shape, direction and provided specialist insight to the tall building process, risks, experienced ‘project losses’ and some innovative ‘project wins’, plus signposted some areas demanding further development. This stage was followed by a series of pi-lot questionnaires, tested on academic and professional colleagues, each version being further refined and tailored to capture the most pressing issues of the tall building process. This ultimately led to the design of the ‘State of the Art of Building Tall Questionnaire’, issued by hand at Dubai and London tall building conferences and on the American Council on Tall Buildings and Urban Habitat website: , as featured in the global CTBUH Tall Building Newsletter, May 2008. The final questionnaire design captured qualitative and quantitative data, aimed at building a comprehensive picture of the global tall building industry. The respondents targeted for the questionnaire were the most active and high profile specialist tall building professionals around the globe, all attending or presenting at the Council of Tall Buildings and Urban Habitat (CTBUH) 8th World Congress, held in March 2008 in Dubai and the New Civil Engineer’s ‘Engineering Tall Buildings September 2008 Conference’, held in London. The results gained from over 150 questionnaire responses are presented and discussed in this paper. 2

THE QUESTIONNAIRE

Questionnaire responses were gained from five tall building industry sectors, representing a cross

section of specialists in the global tall building industry: – The tall building End User or Client; – The tall building Investor or Developer; – The tall building Design Team Member or Consultant; – The tall building Specialist Contractor or Supplier; – The tall building Principal Contractor. A minimum of five and maximum of ten responses from each of the five specialist sectors were gained for each of the four geographical areas considered, resulting in a good representation of the global tall building industry. The questionnaire was split into six sections, the first five sections addressing each of the five tall building key areas and the sixth capturing respondent’s professional details, including current tall building project type, name and their industry specialist sector. The analysed responses to each section are presented below. 2.1

Section 1. International tall building industry—current state-of-the-art

This section set out to establish an overview of the tall building industry across the globe and the key issues inherent in building tall buildings. The results showed that the majority of respondents from all specialist sectors and locations believe that: The international construction industry is not keeping pace with the latest, cutting edge design developments in tall buildings; The UK construction industry is not keeping pace with overseas construction industry developments; The UAE has the most innovative construction industry, followed by China, the USA, Japan, Australia and the UK (joint), then Korea; The global demand for tall buildings will continue to grow; The ‘iconic’ tall building form will take over from the more traditional, rectilinear form; The tall building format provides a sustainable future for the growing global population; The sustainability or ‘green image’ of a tall building is growing in importance; The sustainability of the construction process of a tall building is not as important as that of the finished building; Safety is of paramount importance in the construction of tall buildings; Falls from height are recognized as a large contributor to health and safety incidents in the construction of tall buildings; A more innovative build approach should be sought to minimize falls from heights during the build process.

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2.2

Section 2. The build process of a tall building

This section investigated the build process of a tall building, wherein respondents rated fourteen risks inherent with a tall building project. ‘Principal contractor staff experience’, ‘inclement weather (winding-off tower cranes)’, ‘specialist trade procurement’ and ‘defects completion and handover for progressive occupation’ were consistently ranked the highest risk. These were followed by ‘logistical problems (man and material access, hoist/crane strategies)’, ‘superstructure cycle times/speed of erection’, ‘façade installation’, ‘services installation/commissioning’ risks. The next series of risks were ‘lift installation/builders use/commissioning’, ‘roof/waterproofing/cleaning/specialist architectural features’, ‘shell & core interface with fit-out works’. The lowest rated risks for the tall build process were perceived as ‘demolition of existing building/site clearance’, ‘ground conditions/foundations’, followed by ‘substructure construction’. This section also investigated the respondent’s desire for innovation in tall building construction and their experience of structural frame build speeds, a critical-path activity of every tall building. It concluded that the majority of respondents would strongly embrace and promote innovative construction approach on their tall building project, over a tried and tested construction technique (the example given was the potential use of an innovative crane accessory reducing the effect of wind on material lifts). It also found that the majority of respondents believe a typical tall building concrete frame can be built one floor to the next floor (floor cycle time) averaging 2–4 days. The majority of respondents also believe a typical tall building steel frame can be built with an average piece rate (number of pieces of structural steel erected per crane per day) of 16–20 pieces. 2.3

Construction Management is currently the preferred procurement route for a tall building principal contractor and this form will continue to grow in favor; Previous tall building experience is critical in the selection process of a principal contractor for a tall building project. This section also showed that Construction Management and Two Stage Lump Sum forms of Contract were the two most widely used forms to enter in contract with the tall building principal contractor on the respondents ‘live’ tall building projects. Respondents were then asked to rate the importance of nine inherent tall building project risks previously disseminated from the structured interviews held with the four most prolific tall building principal contractors in the UK. The results showed that the majority of respondents believe that ‘securing finance’, ‘construction program surety’ and ‘cost control/certainty’ were the three highest risks. These were followed by ‘the design process meeting expectation’, ‘securing tenant pre-lets’, ‘build quality’ and ‘construction safety’. The lowest risks were seen as ‘declining demand for tall buildings’ and ‘regulatory and statutory requirements’. Respondents were then asked to rate the importance of principal contractor key attributes that they would consider in selecting the principal contractor for their tall building project. The most important attribute was the ‘provision of an experienced tall building team’. This was followed by ‘lowest cost’, ‘innovative build approach’, ‘history of program certainty’, ‘logistics management efficiency’, ‘procurement expertise’ and ‘local knowledge and experience’. Mid-rated attributes included ‘history of cost certainty’, ‘design management ability’ and ‘value management ability’. Lower ranked attributes included ‘safety record’, ‘established supply chain’, ‘political connections’ and ‘rank or position held in the construction industry’. The least important attribute was the ‘ability to offer project funding’.

Section 3. Tall building principle contractors

This section investigated the tall building principal contractor, wherein respondents rated statements regarding experiences of procuring a tall building project principal contractor, the perceived inherent benefits and most desired attributes. The results showed that the majority of respondents believe that: Tall building principal contractors offer a poor level of safety analysis and value analysis (build ability) of the design at preconstruction stage; Involving the principal contractor at an early stage in the tall building design does assists in delivering value, safety, program and cost certainty; Procurement route options are severely restricted on tall buildings due to the limited number of high quality, capable principal contractors;

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2.4

Section 4. Wins and losses inherent with building tall

This section investigated respondent’s experience of tall building project ‘wins’ or ‘losses’. ‘Wins’ were defined as things that were done well on a tall building project that significantly contributed to the success of the construction process. ‘Losses’ were defined as things that were not done well on a tall building project that negatively contributed to the construction process. This section was not completed in 20% of the responses. However, of the 80% completed, the qualitative responses were highly varied, relating to management techniques or systems, technological advances such as innovative material or methods, plus design related wins and losses. However, the majority

of both wins and losses related to the perceived skills of the tall building project team. The most repeated tall building project win was regarding a high quality construction and management team, with tall building experience from around the globe. The second most repeated win was the early involvement of key trade contractors or specialist suppliers, positively influencing the cost, buildability and program surety. Good and consistent team communication on project issues such as cost, program and design drivers was also a recurring theme. Five recurring types of innovative construction methods were also captured, including slipform advances, tower-crane/hoist advances, concrete related advances and delivery phasing or staging related advances. The majority of these technological wins came from respondents across the five specialist sectors who were directly involved with super-tall towers in the UAE. The most repeated tall building project loss was regarding a perceived low quality construction and management team, lacking tall building experience and skills. Noted weaknesses or specific areas where mistakes had been made included: poor management of the design team; poor trade contractor and supplier procurement; underestimating cost (inadequate budget), design complexity and program; lack of understanding of efficient construction methods and techniques (relying on trade contractor knowledge, rather than in-house expertise). 2.5

Figure 1. Leadenhall Building, London. Bovis Lend Lease bottom-up demolition.

Section 5. New techniques from overseas or other industries

This section investigated new or innovative techniques or practices witnessed by the respondents, which could be adopted in the construction process of a tall building project. These ideas could be either from overseas construction methods, other industry practices, or simply areas where the traditional building approach seems outdated and in need of a fresh approach. It was completed by 80% of respondents, whose observations covered a wide range of topics. They covered aspects from each project phase from detailed design development, through construction to completion, handover and occupancy of the tall building. A selection of the most radical and potentially most beneficial from each project phase include: Design development—A Dubai mixed use tall building utilized early specialist input to influence the design to incorporate a structural ‘jump start’ at Level 8. This allowed the construction works for this section of building to run early, in parallel with the lower levels; Construction and completion—the Leadenhall Building in London developed a ‘bottomup’ demolition of the existing building to allow an early start on excavation and substructure construction. (see Fig 1 and 2); Handover and occupancy—An Australian residential tall building in Melbourne

Figure 2. Leadenhall Building, completion—currently on hold.

London.

View on

developed a phased completion strategy accepted by the statutory authorities, allowing early sectional completion, occupation and easing project cash flow.

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2.6

Section 6. Respondents details

This section captured the respondent’s professional specialist sector, or discipline, within the tall building industry, categorized as: the tall building End User or Client; Investor or Developer; Design Team Member or Consultant; Specialist Contractor or Supplier; and lastly the tall building Principal Contractor. This section also captured the type of tall building project the respondent was currently involved with, showing that the majority of respondents were working on ‘Commercial/Office’, followed by ‘Residential’, then ‘Mixed Use’ tall buildings. It also captured the respondent’s geographical location, their organization/company, and the name of their current tall building project. Although the last two sets of information remain confidential, it is relevant to note that responses were gained from specialist involved with the majority of the current set of iconic tall and super tall buildings currently under design and construction in USA, UAE, London, Paris, Italy, Vietnam, Korea, Japan, Australia and across China. This information will allow further analysis of the responses to be undertaken, investigating the correlation of each industry specialist sector across each geographical region, drawing out contrasts and trends between specialism and geographical location in the global tall building industry and isolating areas of global innovation in tall building construction that could be beneficially applied to the UK tall building industry. 3

DISCUSSION

The analysis undertaken for this paper created a unique snapshot of the global state-of-the-art of the tall building industry over the first to third quarters of 2008. It captured the industry’s buoyant mood and strong belief in continual growth in demand for tall buildings, especially for iconic tall buildings and its unexpected thirst for innovation in the build process over tried-andtested approaches. It reflects the industry’s growing desire for sustainability in tall buildings, if not in the construction process itself. A high level of appreciation of safety risks associated with building tall was common across all industry sectors and recognition of falls from heights as a primary cause of incidents on tall buildings. It also shows that the industry’s leading practitioners believe that the construction industry is not keeping pace with cutting edge designs for tall buildings. This may be reflecting a frustration on the Design Team, Consultants and Client’s perspectives that their iconic designs cannot be constructed as cheaply or quickly as the more traditional rectilinear designs for tall buildings. From a UK perspective, it highlighted some surprising results as the UK was deemed not to be keeping

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up with overseas construction industry developments and was ranked as joint sixth out of seven countries for an innovative approach to construction. This shows the industry as a whole and particularly the UK needs to increase the level of innovation in the tall building construction process. The risk rated the highest in the tall building process was the provision of experienced principal contractor staff, showing the majority of the industry feel they are under-resourced with skilled, experienced tall building professionals. This was mirrored by responses in the principal contractor section, where it was strongly felt that procurement route options were restricted due to the limited number of high quality, capable tall building principal contractors globally. This theme was also reflected by the top rated principal contractor attribute being ‘provision of an experienced tall building team’. Additionally, the most common tall building ‘win’ was related to a high quality construction and management team, and most common ‘loss’ was related to a poor quality construction and management team, lacking tall building experience and skills. The recuringtheme of the responses throughout each section of the questionnaire point to an overheating tall building construction market during the first three quarters of 2008, with insufficient skilled resources to cover the unprecedented demand for tall buildings. It is interesting to note that the declining demand for tall buildings was seem as the lowest of nine tall building risks across all industry sectors and geographic locations. Clearly, in the first to third quarters of 2008, the industry specialists did not foresee the Skyscraper Index (Lawrence 1999) about to bite. (this infamous index historically demonstrates that tall building construction follows the peak of a country’s economic cycle and is followed by a significant economic slump). 4

CONCLUSIONS

This paper has satisfied the objective of investigating five key areas of the global tall building industry, across four main geographical areas of Europe, UAE, USA and Asia Pacific: It has established the current state-of-the-art of the international tall building industry; It has captured key features of, and ranks perceived risk in the international build process of a tall building; It rates the desired tall building principal contractor’s key attributes; It has captured ‘wins’ and ‘losses’ inherent with past tall building projects; It has captured new ideas and techniques from overseas and other industries, potentially bringing benefit to the UK tall building industry. This initial analysis of the results lead to some surprising conclusions, but offers a clearly signposted

way ahead for the innovative construction of tall buildings. This paper will be followed up by a more in-depth analysis of the results, correlating industry specialist sector per geographical region, drawing out contrasts between industry specialism and geographical location in the global tall building industry. This further research will also focus on isolating areas of global innovation in tall building construction that could be beneficially applied to the UK tall building industry along with areas clearly needing further innovation to improve the current state-of-the-art.

REFERENCES CTBUH Tall Building Newsletter, May 2008. http://newsletter. ctbuh.org/newsletter/08-05ctbuhnewsletter.html Lawrence, A. 1999. The Curse Bites: The Skyscraper Index Strikes. Property Report, Dresdner, Kleinwort, Benson Research (March). Skelton, I. 2009. Tall Building Boom—Now Bust? ISEC-5, September 2009, University of Nevada, Las Vegas.

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Construction management

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

An anatomy of speculative claims in construction H.Y. Pang & S.O. Cheung Department of Building and Construction, City University of Hong Kong

ABSTRACT: Protracted resolution of construction dispute has been considered as epidemic. Notwithstanding that significant efforts have been directed to prevent it from happening, disputes regularly crop out in almost every construction project. Most disputes are contractual because entitlement must have a contractual base. For the same reason, contractual disputes may not be that disruptive as a reasonably prepared contract should be able to deal with all phases of a dispute cycle. In this paper, a more disruptive type of construction dispute is discussed. It is proposed that speculative claims are more damaging as these do not fall squarely within the contract governance and often energized by opportunism and fuelled with conflict. This paper outlines this conceptualization by drawing on the well developed concepts of contract incompleteness, opportunism and affective conflict. 1

2

INTRODUCTION

In construction, much of the research on disputes has focused on the prevention, management, causation and resolution of contractual disputes. This paper aims to further the understanding of construction dispute by discussing one of its root causes—speculation. Speculative claims are seldom defined, although its happening seems quite common. This study aims to conceptualize speculative claims that fall outside the governance of contract provision. Complexity and contingencies, preclude a construction project be tendered with a fully completed contract document. As such, changes and variations during construction stage are unavoidable. In addition, self-interest has been identified as one of the motivating forces behind commercial transactions. Unrestrained self-interest seeking behavior can easily be exercised opportunistically. Furthermore, Bac (1993) observed that opportunistic behavior is also linked with contract incompleteness. Construction project involves cooperation of participants of different professional background. Conflicts may arise within or between team members due to divergence in core values both ex ante and ex post. Affective conflict in fact is a perception of interpersonal incompatibilities among project team members, including feeling tension, friction and personality clash. It is proposed that speculative claims in construction can be conceptualized as having three core components: 1) contract incompleteness, 2) opportunistic behavior and 3) affective conflict among project team members. The conceptual underpinnings are founded primary in transaction cost analysis.

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2.1

SPECULATIVE CLAIMS Transaction cost analysis

Transaction cost analysis was originally developed to explain behavior of firms. Coase (1984) explicitly views a firm as a governance structure and proposes that the costs of conducting economic exchange in a market may exceed the costs of organizing the exchange within a firm. Therefore, transaction costs are the costs of running the system and include such ex ante costs in preparing and negotiating contracts and ex post costs of monitoring and enforcing agreements. Over the past three decades, Williamson (1975, 1985, 1996) has extended the concept of transaction cost analysis. He has felicitously referred to Knight’s (1965) works about human attributes and risk-bearing; considered Coase’s (1984) general argument about cognitive and self-interestedness. He suggests that transaction cost include both the direct costs of managing relationships and the possible opportunity cost of making inferior governance decisions; works out two main assumptions of human behavior (bounded rationality and opportunism) and two key dimensions of transactions (asset specificity and uncertainty). Recently, considerations of transaction cost analysis are also applied in construction industry (Winch 1989; Yates 1998; Walker & Chau 1999; Mitropoulos & Howell 2001). Mitropoulos & Howell (2001) advocated that construction projects are long-term transactions with high degree of uncertainty and complexity; it is impossible to resolve every detail, foresee every contingency and fully address every circumstance with respective

contract provisions at the time of contract. This is conceptually be described by Simon (1972) as bounded rationality. The contracts are thus invariably incomplete (Williamson 1975). Such contract incompleteness will form the consequential ex post performance problems, if the contracting parties seek to behave opportunistically. Mitropoulos & Howell (2001) further elaborated Williamson’s (1979) framework of Market Failures and identified project uncertainty, contractual problem and opportunistic behavior the driving factors of contractual disputes. This view was shared by with Yates (1998) who employed transaction cost analysis as a framework for rationalizing the nature, causes and management of conflict and dispute. 2.2

Conflict

Conflict has been defined as a process in which one party perceives that its interests are being opposed or negatively affected by another party (Wall & Callister 1995). Conflict has also been defined as awareness on the part of the parties involved of discrepancies, incompatibility (Boulding 1963). It is consistent with the view of Jehn (1995) who views conflict as a result of disagreements or inconsistencies among the members either intra-groups or inter-groups. Such conflict may occur as a result of incompatibilities or disagreements between some or all of the members of a group and its leader. In this connection, conflict is inevitable in all construction projects (Fenn et al. 1997; Kumaraswamy 1997). Conflict can obstruct decision making, hinder work implementation and affect team’s cohesion (Schweiger & Sandberg 1991) and lower workers’ satisfaction levels (Schweiger et al. 1986). Conflict also affects the organization’s economics and social image (Pondy 1967). Generally, there are two types of conflict; task conflict and affective conflict. Guetzkow & Gyr (1954) distinguished conflict related to the substance of the task that the group is performing from those that dwell on the group’s interpersonal relations. Likewise, Wall & Nolan (1986) differentiated between relationship-focused people conflicts and conflicts resulted from the substantive content of a task. By the same token, Priem & Price (1991) characterized conflict as social-emotional conflicts if interpersonal disagreements are involved. More recently, Jehn (1995) and Jehn & Mannix (2001) advocated that task conflict as a perception of disagreement among group members or individuals about the content of their decisions, idea and opinions; affective conflict as an awareness of interpersonal incompatibilities and includes affective dimensions such as tension, friction, annoyance, frustration, and irritation. In terms of consequence, affective conflict possibly produces negative emotional reactions in workers

such as anxiety, fear, mistrust, or resentment (Jehn 1995). High affective conflict indicated that workers suffer frustration, tension, and fear of being rejected by other group members (Murnighan & Conlon 1991). Affective conflict would also cause dysfunctional effect, diminish group decision commitment, decrease organizational commitment (Jehn & Mannix 2001), raise communication barriers (Baron 1991), diminish work satisfaction (Jehn 1995; Jehn & Mannix 2001), and increase stress levels (Jehn & Mannix 2001). Amason & Schweiger (1994) further described that affective conflict tends to be emotional and focuses on personal incompatibles or disputes. These affective attributes have been in turn related to the propensity to avoid, accommodate, compromise, compete or collaborate among individuals, and affect individuals’ propensity to avoid, confront and resolve claims in the uncertainties flooded construction projects. 2.3

Conceptual model of speculative claims

According to the previous proposition of the nature of construction project, the etiology of transaction cost and conflict, speculative claims has been conceptualized as shown in Fig. 1. Realistically, it is difficult to prepare a fully complete and consistent contract ex ante (contract incompleteness); limit the self-interest seeking behavior, such as the contractor concealing the discrepancy found in the drawing whereby claims can be raised (opportunistic behavior); and manage or control the incompatibility or argument between project team members (affective conflict). For instance, both of the project team members cannot control their emotion thus inhibiting a compromise. Claims out of speculation are then submitted as a result. 2.3.1 Contract incompleteness Contract is used to record an agreement. Macauley (1963) stated that contract involves two distinct elements: 1) Rational planning of the transaction with careful provision for as many future contingencies as

Figure 1.

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The conceptual model of speculative claims.

can be foreseen, 2) the existence or use of actual or potential legal sanctions to induce performance of the exchange or to compensate for non-performance. Transaction cost approach argues that the cost of contracting on an unlikely contingency may well outweigh the benefits. The notion of bounded rationality suggests that agents either have a limited ability to evaluate or foresee contingencies. While bounded rationality is hard to be formalized, Anderlini & Felli (1994) argued that bounded rationality can result in contractual incompleteness. Furthermore, Spier (1992) observed that contract incompleteness is directly related to asymmetric information. Contract incompleteness has different meaning to lawyers and economists. For the lawyers, incompleteness is viewed as obligations have not been adequately specified in the contracts. It would be described as obligationally incomplete, when the obligations of the contracting parties are not fully specified for all eventualities. Nevertheless, the economists view incompleteness as contracts that fail to fully realize the potential gains from trade in all states. It would be described as contingently incomplete, where contingencies may be insufficiently stated. From the transaction cost perspective, Williamson (1975) identified that contract is incomplete due to two main reasons. Firstly, uncertainty implies the existence of a large number of possible contingencies which may be very costly to know and specify in advance responses to these possibilities. Secondly, the degree of devotion for the complex task may be costly to measure. Therefore, self-interest seeking transactors who have ability and incentive to renege can hold up the transaction. He also further named and discussed this phenomenon as opportunistic behavior, and pointed out the hold-up potential accordingly. Recently, Maskin (2002) identified three reasons for contractual incompleteness: 1) some aspects of the state of the world may not be common knowledge or commonly observable, such as the responsibility of the contract may not be able to be ascertained; 2) some aspects of the state may be unforeseen or indescribable by the parties in advance; 3) contingent plans may be exorbitant although some certain aspects are foreseeable. Consequently, incompleteness may lead to underinvestment when contract does not specify in some future contingency and renegotiation of contract terms happened. It would create a bargaining disadvantage and the party may have incentive to extend their power or diminish other’s right during the renegotiation, such as increasing the duties of one party without increasing their right. 2.3.2 Opportunistic behavior Transaction costs analysis proposes that, because of the contracting partners behave opportunistically; high levels of asset specificity increase the costs

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of safeguarding contractual agreements. In this connection, Williamson (1975) developed a general proposition about transaction cost: 1) opportunism is a central concept in transaction cost analysis; 2) opportunism is a crucial element for studying economic which involves transaction-specific investments in human and physical capital; 3) the efficient processing of information is important; 4) the assessment of transaction cost is a comparative institutional undertaking. Opportunism arises because of imperfect control, personal interests, and incomplete contract (Leibenstein 1976). Williamson (1985) furthermore expressed opportunistic behavior goes beyond the simple self-interest seeking but include self-interest seeking with guile. He described guile as ‘‘lying, stealing, cheating and calculated efforts to mislead, distort, disguise, obfuscate, or otherwise confuse.’’ Moreover, this ‘‘strong form’’ opportunism, also termed as blatant opportunism, may manifest itself through both 1) deliberate misrepresentation of various kinds during relationship initiation and 2) various forms of violations over the course of the relationship. When opportunism manifests itself actively, shirking or evasion of obligations in the ongoing relationship happened. Also it manifests itself passively by withholding critical information (Kreps 1990), deliberately lying or misrepresenting material facts (Williamson 1985). Active and passive opportunistic behaviors come up within the existing and new circumstances respectively. According to Parkhe (1993), opportunistic behavior would arise when one firm may not abide by the terms of the agreement in order to exploit the other of short-term gains, such as withholding or distorting information, shirking or failing promises or obligations. Therefore, opportunistic behavior predictably leads to claims and disputes (Calfee & Rubin 1993). Self-interest maximization with guile mostly happens in long-run relationship (John 1984), and parties may have incentives to behave opportunistically under bounded rationality and uncertainty such as the speculative nature of construction project. 2.3.3 Affective conflict Conflict has been broadly defined as perceived incompatibilities and discrepancies among team members (Boulding 1963). Conflict is ineluctable in project teams due to the interdependence among project team members (Fenn et al. 1997; Kumaraswamy 1997). Many researchers have studied the types, sources, handling methods and management styles of conflict (Amason & Schweiger 1994; Jehn 1995, Cheung & Chuah 1999). As claims and disputes are the manifestations of underlying conflicts, a sufficient level of conflict is a triggering dispute factor (Cheung & Yiu 2006). However, they are mostly focused on the

task-related discipline and overlooked the influence of affective-related discipline. Previous studies of affective conflict suggested that where group members are having interpersonal problems and get angry with one another. The friction between them may become personality clashes. In such circumstance, they will work less effectively and produce suboptimal products (Argyris 1962). Besides, Deutsch (1969) described that affective conflicts decrease goodwill and mutual understanding, thus hindering task completion. Ratz (1977) pointed out that individuals may experience stress and pressure from conflicts that are anchored in the substance of the group’s tasks. Ross (1989) viewed affective conflict as emotional conflict which involves personal and relationship components that are characterized by friction, frustration and personal clashes within the group. This conception is in line with the view of Jehn & Chatman (2000) who further suggested that affective conflict involves personal and social issues that are not taskrelated. In fact, Simons & Peterson (2000) viewed affective conflict as the shadow of task conflict. As a consequence of the complexity and uncertainty works in construction projects, invariably construction contract is incomplete due to bounded rationality; contracting behavior is opportunistic as a result. Furthermore, conflict easily arises in project teams with members of diverse interest and personality (Jehn 1995). Jehn (1995) also found that members are psychologically distressed when there were frequent arguments about interpersonal issues among members. The interpersonal problems manifested themselves in intense dislike. Affective conflicts caused distress and animosity among members, encouraging withdrawal, and would be negatively related to group and individual performance. Kelley (1979) explained that a person who is angry or antagonistic simply loses perspective about the task to be performed. Moreover, Staw et al. (1981) suggested that the threat and anxiety associated with affective conflict also tend to inhibit people’s cognitive functioning in processing complex information, and individual performance. Baron (1991) also found that effective communication and cooperation among group members were affected by interpersonal conflicts. Jehn & Mannix (2001) found that 1) emotional conflict limits the information processing ability; 2) increasing group members’ stress and anxiety levels limits cognitive functioning; 3) affective conflict encourages antagonistic.

3

DISCUSSION

All construction contracts are unavoidably incomplete due to the impossibility of foreseeing

every events/contingencies. Likewise, the possible consequences cannot be fully specified ex ante. Contract incompleteness is one of the key conceptual underpinnings to explain the difficulty in exhausting the contractual obligations and rights of the contracting parties. In this connection, contracting parties are accorded the opportunity to maximize their own interests when unspecified eventualities materialize—a form of opportunism. Occurrence of opportunistic behavior is more likely under complex and uncertain environment as suggested by Williamson (1996) and Gibbons (1999). Nevertheless manifestations of opportunism were not addressed. Conflict is intrinsic in dynamic teams. Construction project teams are classic examples. Affective conflict germinates with value incompatibility or disagreement. It has been argued that affective conflict is negatively associated with satisfaction, commitment, psychological well-bring and affective acceptance of group decisions (Schweiger et al. 1986, Jehn & Mannix 2001, Amason & Schweiger 1994). This is because dysfunctional consequences of conflict have been primarily related to affective reactions (Medina et al. 2005 and Varela et al. 2008) and affective climate (Gamero et al. 2008). Varela et al. (2008) also pinpointed that affective conflict affects group behavior and outcomes. The risk of opportunism is thus highly related to affective conflict. In this regard, the three components of speculative claims are interrelated. Contract provision does not feature in this conceptualization as its inclusion would down play the behavior dimension of construction claim. The proposed model puts speculative claims in proper perspective as enshrined by the theoretical anchors. From the perspective of construction project management, understanding opportunistic behavior and affective conflict are essential to achieve effective settlement and prevent dispute escalating. Firstly, project managers have the responsibility to encourage open discussion in order to improve the quality of decision and enhance the acceptance of team members. Secondly, project team members should suppress affective conflict by prompt problem resolution.

4

CONCLUSION

Claims and disputes are inevitable in all construction projects. It is essential to study disputes in construction industry historically. Speculative claim is damaging but has not been adequately discussed in a conceptual context. This study presents such an attempt and proposes that speculative claims have three core components: contract incompleteness, opportunistic behavior and affective conflict. Transaction cost analysis is presented in an alternative approach

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and used to formulate the conceptualization of speculative claims in construction industry. Complexity and uncertainty preclude, construction contracts to be complete. Contingencies occur when they are not fully and clearly covered by respective contractual provisions. Competitive tendering system in construction has not been supportive either. Due to the fact that each construction project standard form of contract may not be adequate to cover the unique features of individual projects. The conceptualization founds on transaction cost analysis that explains that happening contract incompleteness and opportunism. The notion of bounded rationality is pivotal in illustrating one of the root causes of contract incompleteness. It is also suggested that speculative claims is fuelled by conflict that is emotionally-based basing on the belief that a rational and emotionally restrained person would prefer the contractual route for recovery.

ACKNOWLEDGEMENT The work described in this paper is fully supported by a HKSAR Research Grant Council Project (No. 111606).

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Deutsch, M. 1969. Conflicts: Productive and destructive. Journal of Social Issues 25: 7–41. Fenn, P., Lowe D. & Speck C. 1997. Conflict and dispute in construction. Construction Management and Economics 15: 513–518. Gamero, N., González-Romá, V. & Peiró, J.M. 2008. The influence of intra-team conflict on work teams’ affective climate: A longitudinal study. Journal of Occupational and Organizational Psychology 81: 47–67. Gibbons, R. 1999. Taking Coase seriously. Administrative Science Quarterly 44: 145–57. Guetzkow, H. & Gyr, J. 1954. An analysis of conflict in decision-making groups. Human Relations 7: 367–381. Jehn, K.A. 1995. A multimethod examination of the benefits and detriments of intragroup conflict. Administrative Science Quarterly 40: 256–282. Jehn, K.A. & Chatman, J.A. 2000. The influence of proportional and perceptual conflict composition on team performance. International Journal of Conflict Management 11: 56–73. Jehn, K.A. & Mannix, E.A. 2001. The dynamic nature of conflict: A longitudinal study of intragroup conflict and group performance. Academy of Management Journal 44(2): 238–251. John, G. 1984. An empirical investigation of some antecedents of opportunism in a marketing channel. Journal of Marketing Research 21: 278–289. Kelley, H.H. 1979. Personal relationships: Their structure and prophecies. Hillsdale, NJ: Erlbaum. Knight, F.H. 1965. Risk, uncertainty and profit. New York: Harper & Row. Kreps, D.M. 1990. A Course in Microeconomic Theory. Princeton, NJ: Princeton University Press. Kumaraswamy, M.M. 1997. Conflicts, claims and disputes in construction. Engineering, Construction and Architectural Management 4(2): 95–111. Maskin, E.S. 2002. On indescribable contingencies and incomplete contracts. European Economic Review 46: 725–733. Macauley, S. 1963. Non-contractual relations in Business: A preliminary study. American Sociological Review 28: 55–67. Medina, F.J., Munduate, L. & Dorado, M.A. 2005. Types of intragroup conflict and affective reactions. Journal of Managerial Psychology 20 (3/4): 219–230. Mitropoulos, P. & Howell, G. 2001. Model for understanding, preventing, and resolving project disputes. Journal of Construction Engineering and Management 127(3): 223–233. Murnighan, J.K. & Conlon, D.E. 1991. The Dynamics of Intense Work Groups: A Study of British String Quartets. Administrative Science Quarterly 36(2): 165–186. Leibenstein, H. 1976. Beyond economic man: A new foundation for microeconomic. Harvard University Press, Cambridge, MA. Parkhe, A. 1993. Strategic alliances structuring: A game theoretic and transaction cost examination of interfirm cooperation. Academy Management Journal 36(4): 794–829. Priem, R.L. & Price K.H. 1991. Process and outcome expectations for the dialectical inquiry, devil’s advocacy, and consensus techniques of strategic decision making. Group and Organization Management 16(2): 206–225.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Builders’ perceptions of the impact of procurement method on project quality S. Saha & M. Hardie School of Engineering, University of Western Sydney, Australia

ABSTRACT: Quality in relation to a product or service refers to the degree to which it meets the customer’s needs and expectations. It is speculated that traditional competitive bid contracting may have an adverse impact on quality in some circumstances. A survey of the perceptions of construction professionals in the Sydney metropolitan area was undertaken. The respondents demonstrated mixed views on the effectiveness of the competitive tender system for the delivery of quality project outcomes. Some support was expressed for the contention that quality of outcome can be reduced by an overly cost driven contractor selection process. In particular, when economic pressures cause bidders to reduce the time allocated and the amount of checking done for the tender process, an inaccurate and unreliable bid may win. A move to value-based rather than cost-based procurement is likely to be necessary to improve industry performance and customer satisfaction.

1

INTRODUCTION

Traditional contractor selection in the construction industry in Australia has long been managed through the process of calling for competitive cost-based tenders. Such bidding processes may also include estimated construction time as a selection component but, in general, they do not include an assessment of the contractor’s ability to deliver quality project outcomes (Cartlidge 2004; Masterman 2002). Consequently, the lowest priced tenderer may not ultimately be able to deliver the best value for the project’s end users. The traditional system attempts to manage quality either by the exclusion of all but known or ‘invited’ tenderers or by independent supervision of the contractor’s work by an architect or other expert. There are potential difficulties with both these options. In the former case of invited tenders or pre-qualification, there is a risk of a ‘closed shop’ situation developing where new contractors are locked out of the process (Fong and Choi 2000; Mahdi et al. 2002; Ng et al. 1999; Palaneeswaran & Kumaraswamy 2001). In the latter case of architect supervision there may be either a lack of expertise in ‘buildability’ issues or insufficient resources to make timely decisions on problems that arise on. As a result of these problems over several decades, the industry has seen a gradual increase in the use of other procurement methods such as Design and Construct, In-House Development, Partnering and Relationship Contracting (Chan et al. 2003). Nevertheless, the traditional competitive tender system has not disappeared. Despite evidence of the system’s short-comings owners sometimes see competitive prices as their only significant way of ensuring some market input into

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the procurement of an expensive asset. It is not possible to comparison shop for new buildings in quite the same way as it is for real estate. Competitive tendering gives the impression of fulfilling this role although it may not necessarily always deliver a successful outcome. Builders can favour competitive tendering because it allows them to specialise in the delivery of projects and avoid involvement in design iterations and project approval processes. Whatever the procurement method the relationship between cost and value remains a problematic one. In general, the competitive bidding process emphasises cost at the expense of value and the result is a problem with delivering project quality (Uher 1999; Walker and Hampson 2003; Hampson 2005).

2 2.1

BACKGROUND Cost versus value in construction projects

Client satisfaction with the outcome of a building project can be identified as having three parameters which relate to the value triangle (see Figure 1). These are: final project cost compared to the budget; timely project delivery; and quality of the built project result. The quality parameter is particularly difficult to judge in construction because the customer’s decision to buy is based on the concept of a product rather than on the demonstrated finished product as is the case in most industries. Some people have limited visual literacy and have difficulty understanding the end result from design drawings. A client who has difficulty visualising the completed project may cling to cost

Quality

Value

Time

Cost

Figure 1.

Cost—Time—Quality: The Value Triangle.

assessment as something that can be counted on. If cost overruns occur for whatever reason, the client’s satisfaction level is likely to rapidly decline, even if the cost increases are caused by the failure to accurately specify needs or to understand documentation. Of course, contractual arrangements can be used to manage potential price escalation during the project, but these arrangements sometimes result protracted legal battles when a bid has been inadequately prepared or when unforeseen issues arise during the construction period for any reason. An inexperienced owner may not be aware of the potential pitfalls inherent in the complexity of construction project delivery. The lowest tender price from several contractor bids will loom large in the assessment of such an owner because it is easily ranked in a way that delivered quality is not. Similarly, estimated construction time is easily ranked but the veracity of the estimate may be hard for an owner to judge. The competitive bidding process often places the prime contractor and the owner in adversarial positions, a situation which can lead to undesirable outcomes for project quality (Langford et al. 2007). The result is that competitive bidding can, and sometimes does, lead to the selection of incompetent contractors, excessive variations to contracts, litigation and protracted disputes. If a contractor has made a mistake in preparing his estimate or has deliberately underpriced in order to win a contract in the hope of future work, the inexperienced client will probably not detect the problem till it is too late to effectively make another contractor selection. 2.2

Identification of client needs

A further issue with the process involved in Design Bid Build procurement is the problem of clearly identifying and quantifying the client’s needs. In the traditional

system the brief making process is carried out by the owner and the designer with little or no involvement from the builder. This is not a problem for many projects but can create difficulty when the proposed building is unusual or bespoke and solutions are proposed that require new construction methods or practices. Several researchers have pointed out the critical need to get the briefing process right in order to deliver a good project result (London et al. 2005; Yu et al. 2008; Smith et al. 2008). The quantifiable aspects of formulating a brief include floor areas, spatial relationships, service requirements and performance standards. The qualitative aspects are even more problematic as they include concepts of image, aesthetics and human comfort expectations. There are few endeavours where a customer agrees to pay for something when they necessarily have limited knowledge of exactly how the final product will turn out. This disparity between the customer’s identified needs and the specific built solution provided is one of the main sources of customer dissatisfaction with buildings. Research has identified a lack of customer focus in the construction industry (Dulaimi 2004.). Many building contractors see themselves as supplying a product according to predetermined specifications and costs and they put little effort into understanding what the customer actually wants or expects from their project. This is despite the fact that reputation and ‘word of mouth’ recommendations of former clients are a valuable marketing tool which can enable builders to survive economic downturns and manage continuous work streams in a disjointed and project-based industry. Some builders see the construction industry as a service based activity which responds to client demand rather than creates demand. This is not always so, however, and some builders are very successful at developing new technical solutions which lead the market to new performance levels rather than simply responding to what the customer asks for in a standard manner. This raises the issues of how the Design Bid Build procurement can incorporate innovation and new technical development. 2.3

Innovation

Innovation as defined by Slaughter (1998) is the successful introduction of new products, processes or equipment. Traditional procurement tends to leave the realm of innovation to the designer and require the builder to simply deliver what has been specified within cost and time parameters. This denies the particular expertise of builders any significant creative input in the project delivery process. There can be non-conforming tenders where the builder makes suggestions that impact on the designed solution but these tend to mean that the main benefit of Design Bid Build

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is lost as a market price for the work may not be able to be established through price competition. Non-conforming tenders present a particular difficulty for the inexperienced client to assess. Like is not being compared with like and a judgement call has to be made on the issue of which prospective solution best suits the client’s needs. In general, procurement methods such as partnering and relationship contracting have proven to be better at allowing for the incorporation of innovation solutions than traditional competitive tender procurement (Walker & Hampson 2003; Barlow & Jashapara 1998; Kumaraswamy & Dulaimi 2001). 2.4

Adversariality

The final issue of significance with regard to traditional procurement is that it tends to lead to an adversarial relationship between the client and the builder. The client is seeking the cheapest price and the builder is trying to maximise profits. This can lead to the ‘variation’ game, where any discrepancies, errors or omissions in the documentation are used to increase return to the builder at the owner’s expense. Partnering and alliancing procurement systems seek to solve this problem by sharing both the pain and the gain in project delivery and thereby eliminating adversarial contests between the parties and substituting a ‘best for project’ criterion for resolving disagreements. This may well present a useful future path for the industry but at the moment, at least in Australia, such ‘relationship contracts’ are largely the province of big companies and government organisation procurement and have not greatly impacted on procurement at the smaller commercial and residential project level. 3

RESEARCH SIGNIFICANCE

Having identified significant issues with the operation of traditional competitive tender procurement, it was decided to find out if builders in the local area of the Sydney metropolitan region share the misgivings of researchers about the functioning of the competitive tender system of construction procurement. There are commercial, social and environmental reasons for this study. If a relationship exists between procurement method and project quality, it is beneficial for construction companies to know which method of project delivery supports greater quality, in order to ensure maximum positive effect on reputation and repeat business. This information is also important for property investors as the quality of the built property on the market impacts on their investments and return. The issue is important on a social level as everyone uses the built environment and its quality impacts on the lives of all users. It is also significant for the

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environment as construction has a major impact on resource and energy usage, and waste generation and each of these has a direct impact on the natural environment. The research questions to be addressed in this study are whether or not builders have a sense of dissatisfaction with traditional competitive tender procurement and whether or not they believe that procurement systems impact on the quality of their delivered product. 4

METHODOLOGY

A survey was created to test reactions to these issues. Fifty surveys were sent out, with twenty five going to builders, and twenty five going to property developer/ builders. Participants were identified via their reputation as established participants in the building industry in Sydney. They do not represent as randomised sample but rather a purposive selection of participants known to be engaged with the issues under consideration. As such it is not claimed that the responses represent a rigorous statistical opinion poll of industry attitudes but rather an impression of a range of attitudes from a snapshot of industry participants at a particular time in a particular place. All the survey participants were the principal of their construction or development company. They ranged between 30 and 60 years of age and they all held either university or technical college qualifications in construction. Twenty eight completed surveys were received for a response rate of 55%. Twenty one of the respondents classified themselves as ‘Builders’ and seven classified themselves as ‘Property Developer/Builders’. Some respondents came from the commercial building sector, some from residential construction and some operated in both sectors. A large majority (22 respondents or 80%) had been in business for more than ten years. 75% of survey respondents reported that competitive bidding was the main source of work for their organisation. 5

RESULTS

As shown in Table 1, all respondents reported that the quality of the building project was the most important aspect for their organisation when constructing a building project. No respondents rated time or cost as more important than quality. This is at least partly explained by the phenomenon of ‘survey compliance’, in that it would have been clear to the participants that this was both the acceptable and the desired response to the survey. Although assurance was given that survey participants would not be identified in any way, it is still likely that respondents would have consideration for the appearance that their response gives of the company and the industry. Similarly, on the question

Table 1.

Responses to questionnaire.

Question

Response

Which aspect of a project is most important to your organization when constructing a project?

a) 100%

Table 1. (Continued).

b) 0%

c) 0%

a) Total cost incurred, b) Time it takes to build; or c) Quality of the project. Does your organization mainly award sub-contracts based on price alone? Yes 0% Has your organization ever awarded a sub-contract to the lowest bidder which resulted in an unsatisfactory quality of work? Yes 35% How do you think shortening a construction schedule in order to lower the cost of the project impacts the overall quality of the project? a) 50%

No 100%

No 65%

b) 25%

c) 25%

a) No impact b) Small impact c) Large impact When pricing a project, does your organization attempt to lower its bid in order to win the tender? If yes. Has your organization ever lowered the bid significantly that resulted in a less then satisfactory profit return? In your opinion, how does awarding a job to the lowest bidder, as apposed to the more reliable bidder for sub-contracts, impact the quality of the end project?

Yes 25%

No 75%

6

Yes 0%

No 100%

a) 0%

b) 25%

Yes 100%

No 0%

Yes 25%

No 75%

a) 75%

b) 25%

of whether their organisation awards sub-contracts on the basis of price alone, all respondents answered ‘No’. Clearly there is some recognition that cost is not appropriate or sufficient as the sole criterion for determining competence to perform building tasks.

c) 75%

a) No impact b) Small impact c) Large impact In your opinion, do you think that lowering the cost also lowers the quality of the project?

Which of the following do you believe general contractors strive to achieve most when undertaking a construction project? Circle your answer: a) Quality of the product b) b. Profit Has your organization ever completed a project and come out with an unsatisfactory quality according to the client? Which organization do you think would strive to achieve higher quality in a project? a) General contractor under competitive bidding procurement method; or b) b. General contractor under an alternative procurement method, which ends with the general contractor owning the building which they have to sell.

Yes 100% No 0% (Continued)

DISCUSSION

On the question of whether their organisation had ever awarded a sub-contract to the lowest bidder which resulted in unsatisfactory work, a clear majority replied ‘no’. In hindsight it would have been advisable to follow this question with one on whether the organisation had ever awarded a sub-contract to a bidder who was not the lowest price bidder and still had unsatisfactory work resulting. This would have been stronger evidence for the lowest cost bid being closely related to unsatisfactory results. As it is we can only say that for a significant percentage of survey respondents the choice of lowest price bidder sometimes results in unsatisfactory work. There was a mixed response to the question of whether shortening of the construction schedule in order to lower costs has an impact on the quality of the delivered project. Given the previously mentioned caveat that some respondents may have been answering with a view to presenting themselves and their industry in the best light, it is interesting that a

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quarter of respondents stated that shortening the time schedule would have a large impact on the quality of the result. It is evident that the value triangle presented earlier is intrinsically understood to impact on project satisfaction rates. Nevertheless the reported mean assessment of the impact of shortening the schedule lay between ‘no impact’ and ‘small impact’. Developer/builders in the sample rated the likelihood of impact slightly higher than builders did but this was not statistically significant. Similarly university graduates rated the impact slightly higher than did technical college graduates. A larger survey is needed to confirm. whether or not this difference is significant. Respondents with long experience in the industry (over twenty years) were considerably more likely to indicate that the shortening of the schedule would result in lowering of quality. No significant difference was observed between those respondents who mainly undertake competitive bidding and those who mainly initiate their own projects. Survey respondents were asked whether their organisation ever lowers its bid in order to win a tender. Seventy five percent of respondents answered ‘no’. It may be that some respondents interpreted this question as dealing with unethical collusion or improper negotiation of tender prices. Of course, all estimators will try to get the lowest possible effective price when preparing a tender. The question was unclear about the point in the process where price lowering would take place. All the respondents who replied in the affirmative to this question also said that the lowering of their tender price did not have any impact on the profitability of the project. This suggests that they apply their own standards and priorities to the negotiation of final prices in tenders. The impact of awarding a tender to the lowest price compared to a more reliable sub-contractor was considered to be ‘large’ by three-quarters of respondents. The use of the word ‘reliable’ seems to be of significance here. Although several builders felt they could lower their own prices without impact, they were wary of sub-contractors whose pricing might not necessarily be ‘reliable’. There is an acknowledgment of the perils of lowest cost tendering as inexperience, dishonesty or incompetence may lead to sub-contractors overreaching and submitting unachievable prices. This is reinforced by the 100% affirmative response to the question of whether cost reduction results in decreased quality of projects. Three quarters of respondents rated profit as the principal goal that they were aiming to achieve in a building project, while one quarter put quality first. A slightly different emphasis was evident in response to the next question on what is generally achieved at the end of a building project with 100% of answers saying that both a quality product and a satisfactory profit would be achieved.

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One quarter of respondents admitted having had a project where the quality of the end product was unsatisfactory to the client. This rate is considerably lower than the rate of client dissatisfaction measured elsewhere. A larger study may be needed to shed light on this discrepancy. Similarly 75% of respondents declared that they saw no problem with traditional competitive tendering. This contrasts with widespread reports of unsatisfactory results from the system. When asked directly whether higher quality could be achieved via competitive tendering or through another procurement method, the 75% who work mainly in competitive tendering listed that option and the 25% who operate under other systems chose the ‘other procurement method’ response. This simply indicates that the contractors surveyed have chosen to operate in a system that suits them and does not give any objective assessment of how procurement affects project quality. 7

CONCLUSIONS

Although the literature of construction research provides ample information on the shortcomings of competitive tendering as a procurement system, the perceptions of practitioners who operate within the system may be much more mixed (Zaghloul & Hartman 2006). The builders and developer/builders surveyed for this study were surprisingly supportive of the system as it exists. While this may be partly due to the ‘inertia of current practice’, it may also be partly due to the fact that competitive tendering still answers the needs of some parts of the industry. It can continue to supply a constant stream of work for the effective operator while giving the owner some reassurance that a market tested price is being paid. Nevertheless there is a continuing likelihood in economic downturns that some contractors will win tenders based solely on price and prove incapable of delivering a quality outcome. It is likely that in much of industry the use of a truly open call for tenders is rarely done. Effectively there is always a level of vetting or pre-qualification to ensure the worse potential problems with the bidding system are avoided. This may not be done formally but rather through a greater reliance on informally invited rather than open tenders. Competitive tendering is likely to remain a part of the suite of procurement methods used for building projects despite the growth in the various forms of Design and Build or Relationship contracting. The impact of using cost as the only parameter for contractor selection, however, needs to be modified by inclusion of the other considerations which effect satisfaction with the end project. Quality of the delivered end product is not entirely dependent on cost but neither is it desirable to leave cost out of the assessment equation. The value triangle has three corners and each must be addressed in some way in

the evaluation of potential contractors for building projects. ACKNOWLEDGEMENT The authors wish to acknowledge the assistance of Professor Alan Jeary and construction student Lee Salvaggio in the preparation of this research.

REFERENCES Barlow, J. and Jashapara, A. 1998. Organizational learning and inter-firm partnering in the UK construction industry. The Learning Organization, 5(2): 86–98. Cartlidge, D.P. 2004. Procurement of Built Assets, Elsevier Butterworth-Heinemann, Oxford. Chan, A.P.C., Chan, D.W.M. and Ho, K.S.K. 2003. An empirical study of the benefits of construction partnering in Hong Kong. Construction Management & Economics, 21(5): 523–533. Dulaimi, M.F. 2004. The challenge of customer orientation in the construction industry. Construction Innovation(1), 3–5. Fong, P.S. and Choi, S.K. 2000. Final contractor selection using analytical hierarchy process. Construction Management & Economics, 18: 547–557. Hampson, K. 2005. Collaboration and Innovation in Property and Construction. Building Economist, 7. Kumaraswamy, M. and Dulaimi, M. 2001. Empowering innovative improvements through creative construction procurement. Engineering, Construction and Architectural Management, 8(5): 325–334. Langford, D., Martines, V. and Bititci, U. 2007. Best Value in Construction–Towards an Interpretation of Value from Client and Constructor Perspectives. Journal of Construction Procurement, 9(1).

London, K., Chen, J. and Bavinton, N. 2005. Adopting reflexive capability in international briefing. Facilities, 23(7): 295–318. Mahdi, I.M., Riley, M.J., Fereig, S.M. and Alex, A.P. (2002). A multicriteria approach to contractor selection. Engineering Construction & Architectural Management, 9(1): 29–37. Masterman, J.W.E. (2002). The selection of procurement systems E & FN Spon, London. Ng, T.S., Skitmore, R.M. and Smith, N.J. (1999). Decision makers’ perception in the formulation of prequalification criteria. Engineering Construction & Architectural Management, 6(2): 155–165. Palaneeswaran, E. and Kumaraswamy, M.M. 2001. Recent advances and proposed improvements in contractor prequalification methodologies. Building and Environment, 36(1): 73–87. Slaughter, E.S. 1998. Models of construction innovation. Journal of Construction Engineering and Management, 124(3): 226–231. Smith, J., Wyatt, R. and Love, P.E.D. 2008. Key decisionmaking attributes for project inception. Facilities, 26(7/8), 289–309 Uher, T.E. 1999. Partnering Performance in Australia. Journal of Construction Procurement: 5(2). Walker, D.H.T. and Hampson, K. 2003. Procurement strategies : a relationship-based approach. Blackwell Science, Oxford UK. Yu, A.T.W., Shen, Q., Kelly, J. and Hunter, K. 2008. Comparative Study of the Variables in Construction Project Briefing/Architectural Programming. Journal of Construction Engineering & Management, 134(2): 122–138. Zaghloul, R. and Hartman, F. 2003. Construction contracts: the cost of mistrust. International Journal of Project Management, 21(6): 419–424.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Business model of the prefab concrete industry – a two-dimensional cooperation network T. Rinas & G. Girmscheid Institute for Construction Engineering and Management, ETH Zurich, Switzerland

ABSTRACT: Building construction in Switzerland is characterized by small scaled construction and prefabrication enterprises and still dominated by manual production techniques. New strategies and concepts are necessary to empower the small scaled prefabrication and construction enterprises as well as local architects or engineers to benefit both from industrialization concepts and tools, and partly mass-customized production methods for client value generation. This paper presents a business model that teams small scaled prefabrication enterprises as well as local architects and engineers by combining the two cooperation dimensions of development and production, and local assembly and sales. The business model empowers Small and Medium Enterprises (SMEs) in the Swiss construction market to harvest the benefits of industrialized processes for client value creation. The small scaled prefabrication companies will be enabled to transfer manual production to a partly mass-customized production. Local architects and engineers will profit from the client value generation and developed system solutions. 1 1.1

This is owing to the following factors:

INTRODUCTION Initial situation

Industrialization in construction has been the ambitious objective of international research for many years. Enormous effort has been invested and progress made by the international research community in recent years and several approaches have been developed (Bakens 1997). But only little progress in practice has been made in Switzerland, where the construction industry and especially the prefabrication industry are characterized by small players and where ‘powerful’ local architects play a dominant role in every single construction project. Added to which, construction processes are still dominated by labor-intensive on-site manual construction production processes. Together with the Swiss prefabrication industry ETH Zurich is developing a cooperative business model to realize the potential offered by industrialization in construction and automated prefabrication processes. 1.2

– Architects and engineers largely lack knowledge of planning prefabricated concrete elements and systems. – Traditional construction companies are interested to use their own scaffolding for amortization reasons instead of prefabricated elements. – The relatively small prefabrication companies cannot cover the large investments required to transfer the production of a different range of products from manual to automated production. In consequence the level of concrete prefabrication in Switzerland is very low by European standard (Fig. 1).

Problem definition

Neither the Swiss prefabrication industry nor the Swiss construction industry are able to implement the full range of industrialization tools or concepts in the construction process for different construction products or complete building systems (slabs, walls and girders) in Switzerland.

677

Figure 1. Level of concrete prefabrication in Switzerland, Germany and Europe (Girmscheid & Kröcher 2007).

Neither international nor national research efforts have focused yet the managerial approach of a holistic generic business model, utilizing these basic elements of industrialization for the Swiss construction market and its given market conditions. Own empirical studies (Rinas et al. 2008) in Switzerland show additionally:

800

sales [1000 tons]

700 600 500 400 300 200

– Architects and engineers are convinced that prefabrication offers huge potential. – Architects and engineers are extremely interested in using prefabricated concrete elements. – Architects and engineers have a positive attitude toward prefabricated concrete elements. – Architects and engineers lack the knowledge to efficiently plan using prefabricated concrete elements and ‘‘design to production’’. – Architects expect individual design possibilities using CAM/CIM.

100 0 2000

2001

Total

2002

2003

2004

Swiss prefabrication

2005

2006

2007

2008

Foreign prefabrication

Figure 2. Total sales of prefabricated concrete elements (Switzerland and imports from other European countries).

2

STATE OF THE ART

It is state of the art within the European prefabrication industry to use platform-based or system-based mass customized production methods. Not a single prefabrication company in Switzerland uses them. As a result imports of prefabricated concrete elements from other European countries into Switzerland have significantly increased in recent years (Fig. 2). It is state of the art within the timber construction industry to set up an almost complete digital chain using computer integrated manufacturing (CIM) from the very early planning stage through to manufacturing. The prefabrication industry in Switzerland is still at the beginning using the information and communication technology (ICT) available today. It is state of the art in almost every industry having distribution and marketing concepts to sell (product) solutions. The construction industry still mainly focuses on selling labor instead of solutions. Distribution systems and sales concepts are not established in the traditional construction market. Service provision and life cycle solutions are slowly emerging and client focus was introduced just a few years ago (Girmscheid 2006).

3

As a result, 62% of architects and engineers as well as 57% of prefabrication companies would accept an external consultant or planning firm to substitute their own lack of knowledge, and make ‘‘design to production’’ and ‘‘best use of prefabricated concrete elements or systems’’ reality. Sequentially there is an essential need to transfer industrialized construction concepts and tools in the Swiss construction market by developing new business models and strategies in order to compete for the future as Prahalad & Hamel (1996) used to say. Following questions have to be answered: – How can current technological possibilities in planning and automation be adopted by the clustered Swiss prefabrication industry and the construction market to provide individual prefabricated elements using mass customization? – How can prefabrication potential be implemented as a standard and how can execution know-how be transferred to the early planning stage? – How can market penetration with prefabricated elements or system buildings be improved by integrating the traditional and regional trades players?

STATE OF RESEARCH 4

Both the international research community and the international construction industry focus enormous effort on all basic elements of industrialization in construction, such as off-site fabrication (Gibb 1999) and on-site fabrication (Bock et al. 2007), logistic issues (Boenert & Blömeke 2003), partnering and cooperation concepts (Eccles 1981, Hofmann 1999, Girmscheid 2005), potentials and options for further developing the construction industry (Girmscheid & Kröcher 2007) as well as industry-led research projects (Kazi et al. 2007).

RESEARCH METHODOLOGY

This research work is embedded in the holistic SysBau® approach (SysCon—system provider approach) of the Institute for Construction Engineering and Management at ETH Zurich which covers the life cycle oriented research fields—integrated delivery and lean construction (Girmscheid 2001). The scientific framework of the hermeneutic science program (HPS) for the socio-technical environment consists of the interpretativist research paradigm (Weber & Winckelmann 2002) and the constructivist

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research paradigm (Glasersfeld 1998). The interpretativist research paradigm is used to understand and explain the reality by conceptualize the reality in a phenomenological explanatory model. The constructivist research paradigm scientifically leads the researcher to construct logical-deductively the reality to reach the intended targets (target-means relationship). The presented business model is designed logicaldeductively as a target-means relationship by following the constructivist research paradigm (Piaget 1973, Glasersfeld 1998, Girmscheid 2007). The scientific excellence in the constructivist research paradigm is achieved by triangulation due to – viability of the generic-deductive model, – validation through a theoretical framework and – confirmation of reliability through testing the intended impact. The model is theoretically structured by the Principal agent Theory (Jensen & Meckling 1976) and the Theory of Structuration (Giddens 1985). The triangulation will be closed by the realization test to determine whether any alternative interpretation is possible in regard to the target-means relationship, resulting in a scientifically confirmed model which is mostly free of internal opportunistic behaviors.

5

OBJECTIVES OF THE TWO-DIMENSIONAL COOPERATIVE BUSINESS MODEL

The objectives of the two-dimensional cooperative business model are: First: Value creation for the client through industrializing building processes. Second: Combine competences to develop innovative and integrated elements and systems (solutions). Combine marketing competences and key accounts to distribute the solutions into local markets and increase the market penetration with prefabricated concrete elements. Third: Use current ICT and ‘‘digital chain processes’’ in the planning and production processes and a strict orientation of processes on the value creation chain to avoid waste (value engineering). Fourth: Introduce a ‘‘design to production’’— paradigm and maximize the usage of prefabricated concrete elements as a standard in the planning process. Fifth: Enable the prefabrication cooperation partners to develop industrialization and partly automated mass customization in specific product ranges which are complementary from the involved partners to the offered solutions. Sixth: Contribute toward solving the structural problems in the Swiss construction market which obviate the full use of prefabrication potentials in the construction production process.

679

6 6.1

APPROACH Strategic cooperative network approach

Strategic cooperative networks are promising means of achieving the objectives in the given clustered conditions of the construction market in Switzerland. A focal enterprise as introduced by Sydow (1992) or a ‘hub’-firm as called by Jarillo & Ricart (1987) allocates responsibilities, tasks and roles to the cooperation partners. A focal enterprise gives the network a stronger commitment among its partners. The focal enterprise also leads the network processes along the value creation chain and ensures that the processes and players are controlled. By contrast, virtual enterprises do not have this institutional character and are therefore less committed and more at risk of losing control of the value creation chain process (Maier 2002; Sydow 2006). 6.2

Concept of the two-dimensional cooperative business model

The concept for succeeding with this business model is based on four fundamental elements: First: Create client value by providing value orientated system solutions and exploiting industrialization potentials through a holistic approach of the building process chain (functional organization) and cooperative work chain (structural organization). Second: Create partner value by providing value orientated system solutions und additional client value to enable differentiation from competitors and entry into other regional markets. Third: Optimize the realization of the value creation process and combine complementary partner competences and resources at different value chain levels by means of a cooperative organization of specific but supplementary automated production and development among the partners to reduce investment cost per partner. Fourth: Multiply client relationship through a cooperative organization of distribution and assembly. 6.2.1 Create client value The most fundamental element in a sustained business is to create client value. Three main categories of client value in the construction market have been identified (Fig. 3): – guaranties – reduction in life cycle costs – increase in life cycle value The two-dimensional cooperative business model aims to decrease the life cycle cost of a building by – the holistic and integrated planning of all types of work and ‘‘design to production’’ to avoid waste and ineffective work processes,

Figure 3.

Categories of client value.

– the extensive use of industrialized and supplementary distributed automated prefabrication among the partners in a cooperative way to improve quality, accelerate the projects and reduce financing costs, – implementation of mass customization in the industrialized and automated production process of prefabricated concrete elements, – integration of all types of works in prefabricated building solutions, including HVAC with passive heat storage components and alternative energy systems to deliver optimized LC-buildings for the client, – a holistic view on both the planning and execution processes and implementation of a complete, continuous and systematic process for the value creation chain, – integration of innovative solutions developed by traditional building and prefabrication companies in early stages of the planning process and transferring execution knowledge back into the planning process (CIP feedback circle). Additionally, a shorter planning and execution process will result in earlier availability, thus increasing the potential life cycle value of the building for the client. Furthermore a structured planning and execution process will give service providers a better opportunity to sell service guaranties on price and schedule by lowering the risks of the planning and execution process. The client value then has been created by fulfilling client demands for service solutions and life cycle optimized buildings.

6.2.2 Create partner value Another fundamental element in a sustained cooperative business is to create partner value. The partner value stems from the opportunity to be a part of the cooperative provider of full service solutions for life cycle optimized buildings. Each partner himself cannot provide these service solutions alone due to a lack of system competences, technologies and resources as well as a limited risk bearing ability. The continuous improvement of the value creation chain by repetition and replication of processes and interfaces is a further advantage of the cooperative work and an important partner value creation element. Elements of Principal Agent Theory and Theory of Structuration largely repel opportunistic behavior by partners among themselves. Less opportunism between partners decreases the transaction cost for economic activities and also increases partner value. As already stated, a paradigm shift (Girmscheid 2005) in the construction market has already started. This value-price competition opens up a new level of competition for players to differentiate themselves from competitors (Fig. 4). Therefore the competition strategy (Porter 1996) of this business model is predominantly based on differentiation, although cost leadership must also be secured over the medium term. Differentiation in service provision is essential to follow the paradigm shift toward the provision of service solutions and client focus in the construction market. Differentiation comes from creating added client value which traditional players in the construction market cannot offer.

Figure 4. price.

680

The competition layer between client value and

Cost leadership comes from the incremental implementation of mass customized prefabricated concrete elements. Additional partner value is created through improving the market penetration of service provision solutions using a distribution and sales network. The regional range of the services provided by each partner is limited; none can provide a widespread distribution system. The widespread distribution of services sets the framework for industrialization in construction with local players involved as partners. 6.2.3 Realization of the value creation process The existing traditional project delivery processes need to be redefined into a strictly value creation chain orientated order. The construction production process must be value-optimized by – introducing new value-adding steps, – summarizing different value-adding steps and realizing potential synergetic effects, – increasing value creation at a specific valueadding step. Interfaces and competences must be adjusted while redefining the project processes. The development and production cooperation in the presented business model (First layer in fig. 5) will ensure that the processes and interfaces are redefined with strict orientation toward value creation. In the traditional construction process, players only focus on their own work. The first layer of this business model merges the necessary players into a cooperative network to develop and provide system solutions with maximum prefabrication. This includes a system architect, several prefabrication companies with different specialized and automated production lines, a connection specialist, a HVAC engineer and a HVAC company to provide integrated energetic elements and to develop new solutions for alternative energy supply and passive energy storage elements in the building.

Figure 5. The concept of the two-dimensional business model for the Swiss market.

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A focal enterprise and competence center organizes the project specific works, controls competence allocation (prefabrication and HVAC competences as well as service solution competences) and the competence coordination among the partners. All partners of the cooperation hold shares in the focal enterprise. The partners continue to perform their traditional performances in addition to their commitment to the cooperation. This cooperation focuses both on further developing the integrated prefabricated elements and systems and improving service provision, and on introducing efficient production methods. A process of continuous development and continuous improvement is essential for the successful transformation from traditional labor provision to value creating service provision with a maximized implementation of distributed cooperative prefabrication. This development and production cooperation enables the prefabrication industry in Switzerland to implement mass customized production. Mass customization in turn enables the cost-effective production of almost all elements independently from any series. This is mandatory for successful penetration of the Swiss market and to enable the exploration of new markets in neighboring countries. It is a precondition for competing with traditional construction methods and foreign prefabrication companies and for retaining the individual nature of architecture. A continuous improvement of solutions and the increasing efficiency of the value creation chain are necessary elements to successfully expand the business model in the second dimension of sales and assembly. 6.2.4 Multiply client relationship The second dimension (Second layer in fig. 5) links the strategic development and production cooperation to the clustered regional markets and multiplies the potential clientele. Local architects in Switzerland play the dominant role in every single construction project. Therefore local architects must become partners in the cooperation to ensure a successful sales network and local assembly. The local architects will still draft individual designs to meet the client’s building requirements but will also cooperate with the prefab network to ensure the productibility. Cooperation among local players and a development and production cooperation in one network enable direct distribution concepts from the service provider to potential clients. It significantly increases the potential for successfully penetrating the market with prefabricated elements and system solutions. Local architects benefit from the provision of system solutions and realization of the existing potential currently offered by prefabrication while still retaining their architectural freedom.

It’s mandatory to teach local architects about the future oriented alignment of the construction market and demonstrating the opportunities offered by the cooperation. The results from the development and production cooperation help to convince local players of the improvements in the planning process and execution. Local trades should also be involved in the in-situ installation of the final works. 7

CONCLUSIONS

This two-dimensional cooperative business model fulfills the requirements of a sustained cooperative business model (Girmscheid 2006). – It creates value for clients and partners. – It has a leading competition strategy to convince the client of the benefits of the service provision and to allow differentiation from competitors. – It organizes the lean realization of a value creation process with responsibilities, tasks and roles assigned to each of the cooperation partners. As such, this two-dimensional cooperative business model represents a promising holistic approach to: – implement prefabricated concrete elements and HVAC systems as an viable alternative in the construction process of individual residential buildings in Switzerland and to – exploit the currently existing potentials both for industrialization with mass customization and cost savings in the Swiss construction market. The business model adapts a cooperative organized structure to produce and to assemble prefabricated concrete elements and systems by merging essential players and competences as well as by focusing project delivery processes directly on the value creation chain. It considers especially the local conditions in the clustered market environment in Switzerland. Despite the relatively small size of the market compared with other countries, it provides Swiss companies with all the potentials offered by industrialization in construction. REFERENCES Bakens, W. 1997. International Trends in Building and Construction Research. Journal of Construction Engineering and Management 123: 102–104. Bock, T., Navon, R., Ramanathan, M. & Sunil, M.K. 2007. 24th International Symposium on Automation & Robotics in Construction (ISARC 2007) Construction Automation Group, I.I.T. Madras—Keynote Papers. I.A.A.R.C. Boenert, L. & Blömeke, M. 2003. Logistikkonzepte im Schlüsselfertigbau zur Erhöhung der Kostenführerschaft. Bauingenieur 78: 277–283.

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Conservation project management by the architectural digital photogrammetry F. Navarro, A.L. Rodríguez & V. Ávila Holguin University, Holguin, Cuba

C. Loch Santa Catarina Federal University, Florianópolis, Brasil

ABSTRACT: The conservation works to the overall project and, principally, to the depth of an intervention in conservation, which involves technical processes, scientific and historical knowledge, manual ability, sensibility, etc, with the objective of prolonging the site’s working life, emphasizing the use of tools like techniques of architectural digital photogrammetry. It is make reference to the importance of saving the images of sites with patrimonial value over these geographic areas, which would constitute an important base of information for planning a consistent conservation of the sites and would best safeguard the patrimony. The considered subject seeks to penetrate the conservation projects of sites with patrimonial value—in various states of deterioration—present in the historical centers of the principal cities of the country, as well as contemplate their surroundings, in order to make an analysis regarding the treatment given to the different activities that compose the conservation of sites.

1

PATHOLOGIES ON CONSTRUCTION

The study of pathologies on construction works aims at the detection of their deteriorated or damaged conditions, with the objective to determine their causes and the technologies that should be used to carry on amends and/or rehabilitation. Usually, these conservation actions consist of works with special characteristics, quite different from the constructive and assembling tasks that are commonly held, no matter if they were executed in another time, using different techniques and materials no longer applied, or due to intrinsic problems concerning the requirement of unusual, accessory works. All this, turns conservation actions into long-term, expensive investments that should be fulfilled only after a thorough, specialized study that guaranties the adequate accomplishment of the investment at a lower cost, with the only purpose to build a useful and lasting construction. Traditional methods to study the previously mentioned pathologies require the realization of penetration, excavation, perforation of materials, and other destructive actions that contribute to the deterioration of the building and are insufficient to overcome the inconvenient of ‘‘point by point’’ methods (Menéndez 1987). Similarly, they are not always effective to provide information about the necessary measures to cover the damages and their causes. This is why it is needed to use modern methods that allow the non- destructive, simultaneous memorization and interpretation of the

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qualitative and quantitative data of the construction platforms. The development of computing technology gave birth to the possibility of identification of a building by means of a transcription of its image into a digital model, offering extremely accurate data to execute comparative analysis at a dimensional information level. The International Council of Monuments and Sites (ICOMOS) recommended that each country becomes a photogrammetric file of its historical monuments. The implementation of digital, architectural photogrammetry along with the use of data banks having information on the monuments is, undoubtedly, one of the most significant technological tools available for the record (Altrock Von 2004). The present work offers an example of the application of these techniques to the process of conservation of cities with historical and patrimonial values.

2 2.1

DEVELOPMENT Photogrammetric techniques

The Venice Chart defines historical authenticity as an essential value to assure the objectivity of restoration works, and it promotes the importance of the precise raising of monuments at a certain period, in order to preserve aesthetic, technical, and historicalhints. This is why digital photogrammetry is so valuable since it renders quicker and more accurate results, which

allows obtaining milimetrically precise measures to store scientific information on cultural properties, and for the analysis and control of distortions in buildings. From 1950 on there is some considerable development of the electronic computer, which constituted a forward step in the application of photogrammetry technique. People in charge of safeguarding patrimonial values, who had as well moved towards more strict methods to sustain the conservation principles, took advantage of this event. This is why the International Council of Monuments and Sites (ICOMOS) organized in Paris, 1968, three years after its foundation, the First International Symposium, in honor to architectonic photogrammetry. As a result of the resolutions passed, the International Committee for Architectonic Photogrammetry (CIPA) was created in 1970, and it is nowadays a permanent working group at the International Society of Photogrammetry and Teledetection (ISPRS). In Cuba, the implementation of photography to study architectonic constructions started at the end of the 1970s by the Cuban Institute of Geodesy and Cartography (ICGC), aimed mainly at the preservation and restoration of monuments in Old Havana, Trinidad and Santiago de Cuba. The main task of photogrammetry is to establish a geometric relationship between image and object, keeping the appearance this last one had at the very moment of the shot. Once this link is properly set, it is possible to make a photogrammetric restitution, that is, to obtain a graphic or numeric representation of the photographed object. This connection can be classified into three categories: – Graphic: using geometric relations. – Analogical: using optic-mechanical components. – Analytical/digital: using a mathematical model and digital processing. Digital photogrammetry turned photogrammetric restitution into a simpler, more flexible and accessible procedure, making possible the use of multiple applications once the processing is done by minicomputers through the use of specific photogrammetry programs associated to other regular-use equipment, like scanner and different sorts of cameras. This work also describes the assumed procedures and the results obtained from the record of architectonic edifications using the digital photogrammetric software Photomodeler, which employs monorestitution and restitution techniques using multiple photographs; this program allows the procurement of various products from digital images: point coordinates, distances, 3D models, three-dimensional models with the original texture of the object, orthophotos, and it permits the exportation of results in different formats. (Details can be found on www.photomodeler.com).

2.1.1 Monorestitution Monorestitution assumes the solution of a problem through one single photograph. To achieve this, it is required to get some information about the geometric features of the object, since in a photo the coordinates of one single point may correspond to some other countless coordinates located all over the object. To solve this problem it is necessary to have some other information like parallelism, perpendicularity of the object’s edges, and identification of its axes X, Y, Z in respect to the position of the camera when photographing. To determine the scale it is needed, at least, to know one of the dimensions of the object in order to classify those selected for restitution, using a unique photo, into three different types: flat objects, flat, irregular objects, and objects where the three axes can be identified. Monorestitution also allows obtaining an object’s dimensions from less complex programs like DigiCAD 3D and RolleiMetric, which rectify the photo once the coordinates of two of the surface vortices are known. After being rectified, the photos can be exported to a design editor to sketch its geometric features. However, although this process offers some practical limitations for the only parts on the photo that can be restituted are the visible ones, it is still a low-cost technique easy to use and fast to work with. Its application is recommended for the restitution of non-existent monuments of which there is only one photograph left. 2.1.2 Restitution using multiple photographs This technique permits to photograph objects in different positions, where each part must be caught several times and with superimposed points. From the identification of homologous or common points among multiple photos, it is possible to draw the intersections for the restitution of the object. This was the first type of restitution ever used since the beginnings of the application of photogrammetry, and it antecedes as well the already mentioned methods. It was firstly implemented by means of the ‘‘point by point’’ measurement and representation graphic processes; but it had limitations since it lacked fastness and precision. Technological improvements and the increasing computing capacity favored the application of the principles and foundations of this technique—graphical method—to create programs able to solve problems analytically by means of mathematic equations. The implementation of the graphical method has regularly increased nowadays, especially for architectonic constructions, mostly due to these factors: – Employment of general-use equipment like scanner, photographic, ordinary or digital camera, and microcomputers associated to specific programs.

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– Different focal distances and diverse angles that permit a photographic study of the whole edification. – Reduction of the amount of control points. – Variety of procurable products: photorealistic models, orthophotos, computed designs that favor the integration of the files in multiple programs. – Analytical calculation of measurements and unknown parameters, for which it is unnecessary to know the parameters of the camera, its position or orientation at the very moment of taking the picture. – Possibility of error detection. Despite all the advantages this method offers, restitution through multiple photos is limited, because of the absence of stereoscopic vision, to the detection of discrete points that are easily identified on diverse images. This is not enough in many cases though, especially if the object shows geographical forms that can be clearly defined through points and lines, and irregular or curved shapes like sculptures and architectonic details.

3

3.1

SAMPLE: RECORD OF CONSTRUCTIONS LOCATED ON A BLOCK AT THE HISTORICAL CENTER OF HOLGUIN CITY

Analysis of the constructions conditions

Analysis of architectonic and constructive features in different regions of the country

Selection of Works and planning of the raising

Field work and orientation

Restitution

Conservation proposal

Figure 1.

Constructions selected.

H A

Methodology

G

The carried on research should offer some information about a series of basic aspects. Some of these are:

B

– Characteristics of injuries in the objects under investigation. C D E F – Proportions, distribution, and manifestation of injuries. – Causes of the injuries. Figure 2. Position diagram in restitution to photograph. – Optimal state of the images for the record of each kind of damage. Selection of constructions: To choose the works it was required to take into account some factors like: type of edification, accessibility to take measures, possibilities of a photographic study from different angles. As a result, all the constructions of the block comprised by Maceo Street, between Frexes and Marti Streets, Holguin were selected as the prototype for the preservation task held at the historical center of Holguin, Cuba. To work on this, the following method was applied: The photogrammetry techniques used to explore the characteristics of the constructions were restitution by multiple photos carried on through the Photomodeler. These were the procedures followed: – Planning the photographic study: To determine the photogrammetric techniques to be used, proper equipment, and control points. – Field work and orientation: First, to photograph the edifications and to make an analysis of the control

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points. Then, to determine the positions of the photos according to the perspective center or to the system of reference of the object, that is the internal and external orientation of the object. – Restitution: To proceed with the study and to obtain all the graphical or numerical data of the photographed object, once the orientation parameters of the photo are known. – Preservation proposal: With the implementation of digital, architectural photogrammetry as a useful tool to explore the characteristics of the edifications, it was possible to design a preservation project which offered technical solutions that favored as well the financing of the work. It was also possible to optimize time and quality of the execution during the whole process, having some considerable accuracy on the study held, which allowed making a complete restitution of facades, even of very little details, at the desired scale.

4

ANALYSIS OF THE RESULTS

The assessment of the results was made by comparing 3D models resulting from the programs used and from the dimensions offered by measurements taken directly from the edifications. Most of the photos taken by sections had the proper contrast and resolution of the image, which favored exactitude— Error < 1 mm. The average on precision of the three-dimensional model that was created varies from 4 to 6 cm. 5

CONCLUSIONS

The utilization of GIS and architectural digital photogrammetry for the management of conservation

projects in the patrimonial scope constitute a tool that favors the obtaining of greater effectiveness and efficiency in the sustainability of projects, contributing to it a character both scientific and novel. REFERENCES Altrock Von, P. 2004. Aplicações da fotogrametria arquitetural digital na documentação de edificações históricas— estudo das obras do brigadeiro Joseph da Silva Paes, séc. XVIII. Tesis doctoral. UFSC. Brasil. Leão de Amorim, A. & Groetelaars, N.J. 2004. Técnicas de restituição fotogramétricas digitais aplicadas à Arquitetura: um estudo de caso. UFBA. Brazil. Menéndez, J. 1987. Desperfectos en obras de ingeniería y arquitectura. Ed. MICONS. La Habana.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Construction productivity and production rates: Developing countries C.R. Guntuk & E. Koehn Lamar University, Beaumont, Texas, USA

ABSTRACT: This paper reports on a study of factors affecting productivity among members of the construction workforce in developing countries such as India. Economic and socio-psychological factors that affect labor performance and are of increasing importance in a developing country are evaluated and discussed. Construction labor productivity is of great interest to international projects as it affects project cost and time overruns. Therefore, up-to-date cost and production data for the construction of various international projects should be readily available for estimating project planning and bidding purposes. In this investigation, the costs of construction projects in developing countries related to labor expenses are evaluated and compared to that of a developed country, USA. The findings should assist constructors working on international projects during the planning process to achieve the desired estimating, scheduling, quality, cost effectiveness, duration and updating of construction projects. 1

INTRODUCTION

2

India is located in the southern part of Asia between the Arabian Sea and the Bay of Bengal (Chandra 1990). Its natural resources include coal, iron ore, manganese, mica, bauxite, titanium ore, chromite, natural gas, diamonds, petroleum, and limestone. The country has a population of roughly 1.2 billion. Construction labor in all the big cities in India, can generally work throughout the year. In addition, there is also some consideration for safety and health. But in the rural portions of the country, laborers normally work on a farm for approximately 6 months and in construction the remainder of the year. Here, there is little concern for construction health and safety. For these and numerous other reasons, there tends to be a great difference in productivity, quality control, and project duration in various parts of the country (Rate 2007–08). The economy of developing countries such as India is a mixture of traditional village farming, agriculture, handicrafts, a relatively small number of industries, and a multitude of support services. Faster economic growth in the last two decades created an increase in per capita private consumption. Nevertheless, a large proportion of the population, perhaps as much as 50 percent, is below the poverty line and cannot afford an adequate diet. Poor nourishment causes fatigue, which may tend to reduce productivity, and increase the accident and fatality rate for workers, especially in the construction industry. This is a problem that must be addressed.

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CONSTRUCTION INDUSTRY IN INDIA

The construction industry is a major economic activity in India. Construction activities contribute annually about 10 percent to the gross national product (GNP), thus playing a major role in development plans, since most sectors of India’s economy involve construction activities (Sheer 1993). The Table 1 shows the approximate construction component in respect to various development projects throughout the country. The construction industry is one of the largest in India and employs about 50–100 million skilled and unskilled workers. Workers in the construction industry, as in most countries of the world, are hired as needed and fired on completion of the project. Due to the temporary nature of their job, workers remain idle until they find work at new construction sites. In large

Table 1.

Construction projects in India.

Development project Irrigation works, dams, canals, etc. Roads and bridges shipyards, harbors, airports, etc. Shipyards, harbors, airports, etc. Thermal power plants, steel mills, aluminum plants, etc.

Approximate construction component expressed as a percentage of total project cost 90–100 90–100 45–55 15–20

construction firms, personnel at the supervisory and managerial levels are employed on a permanent basis, but they must move from one project site to another to keep their position. Construction laborers, however, are hired on a temporary basis and lead a migratory life, working on different sites throughout the country (Sheer 1993). The economic condition of construction labor is poor because of the laborers low bargaining power, lack of unions, illiteracy, and the temporary nature of their employment. Basic amenities such as shelter, drinking water, sanitary conveniences, etc., are not adequately provided at small construction sites. In addition, there is little concern for the education of the children of migratory construction workers. Article 24 of the Indian constitution specifies that no child below the age of 14 years shall be employed to work in any factory or mine (Chandra 1990). Nevertheless, there are still a large number of child workers throughout the country, especially in construction. Construction laborers are members of the most disorganized sector of the Indian workforce, and they do not enjoy benefits that their counterparts in other industries and organized economic sectors enjoy. Construction workers have no job security and have little training. In many developed countries, such as the US, Japan, etc., there are programs to train workers for particular trades, such as mechanics, pipe fitters, plumbers, machine operators, electricians, etc. Although industrial institutes have been established in India, their contribution toward training of construction labor is inadequate (Safety 1996).

3

PRODUCTIVITY OF A PROJECT

According to the principles of management, a project may be considered to consist of a combination of four Ms: – – – –

Money (capital); Manpower (labor); Materials (resources); Machines (equipment).

The principles of management indicate that the availability of specific resources often determines the optimal manner in which they will be used (Building 1986; Koehn 1989). Since the cost of labor, especially in developed countries, may constitute a large percentage of the total project cost, increasing labor productivity is important in controlling both project cost and duration (Sharp 1997). In fact, according to some authorities, productivity may be defined as the ratio of output to input. In this investigation, output is taken as the specific units of different construction activities and input as the work-hours (labor hours) required to accomplish that work.

In numerous instances, the availability of specific resources determines the optimal manner in which they will be used. This is particularly true in developing countries. In these areas, the labor cost per hour is less expensive than of developed regions and, due to the large supply of workers, labor intensive construction methods are followed and the use of construction equipment is limited (Chandra 1990). By maintaining proper scheduling, communication (coordination), and engineering design functions, production (productivity), however, can be increased, but consideration should also be given to revising restrictive local work rules (Arditi 1985; Koehn 1989). In addition, the lack of realistic labor cost and production estimating data is a major problem in developing countries. 3.1

Comparative labor requirement factors

It is known that labor productivity (production), as well as the cost of capital, materials, and equipment, is extremely difficult to measure due to the heterogeneity of the construction industry’s products and inputs (Koch 1979). The comparative labor requirement for India was investigated in this paper. In order to accomplish this task, two principle data sources, the Rate Analysis for Construction Works of India and Building Construction Cost Data of the US were used (Rate 2008; Building 2006). The following expression can be use to determine the comparative labor requirement factor: LR1 = K × LR2

(1)

where, LR1 = labor requirements in India; LR2 = labor requirements in US; and K = comparative labor requirement factor. In this study, the labor requirement is calculated on the basis of the work hours required for 100 units of work in a typical 8-hour day. In India, the normal work day is 8 hours, from 8:00 a.m. to 5:00 p.m., with a 1-hour lunch break. For comparison, the labor requirement in the US is considered as the standard of reference. Thirty-three different activities associated with building construction, such as site preparation, earth work, concrete and masonry placement, timber, flooring, finishing, and dismantling are compared in this presentation. However, it is expected that data could be applied to other sectors of the construction industry such as highways, bridges, dams, and airports by applying various technical adjustments. The comprehensive results of the investigation are shown in Table 2. Columns 6 and 7 lists the comparative labor requirement factors in a developing country (India) for both smaller and larger firms compared with that of the US. As shown, the comparative labor requirement factor for smaller firms is generally greater than unity (1.00). This is due to the extensive use of labor for most

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Table 2.

Comparative labor requirement factors.

Developing region India Rural or Urban

Comparative labor requirement factor India/USA Rural or Urban

Developed Region USA

Small Firms

Large Firms

Small Firms

Large Firms

4.30

2.43

3.64

2.06

Man hours required

Activity Description

Unit

SITE PREPARATION Clear and Grub

m2

1.18

EARTH WORK Excvation (silty soil) Excavation (sandy soil) Leveling/Grading Earth Filling

m3 m3 m3 m3

261.29 130.65 12.70 137.70

519.2 320.9 13.11 198.6

302.6 158.6 12.17 151.8

1.99 2.46 1.03 1.44

1.16 1.2 0.96 1.1

CONCRETE Concrete (foundations) Concrete (slab/ beam) Reinforcement Form work (beam) Form work (foundation) Form work (slab)

m3 m3 N m2 m2 m2

137.40 228.50 0.20 166.60 322.89 92.20

843.6 1189 0.89 509.6 1014 275.3

265.0 363.6 0.63 273.8 494.3 137.6

6.14 5.20 4.45 3.06 3.14 2.98

1.93 1.6 3.15 1.65 1.53 1.50

MASONRY Brick Work Stone work

m3 m3

1486.80 1027.30

1709 4619

1511 3621

1.15 4.50

1.02 3.50

TIMBER Timber roof trusses Eaves Plywood partitions Plywood ceiling

m3 m2 m2 m2

798.10 7.50 40.90 48.74

9872 16.33 60.93 128.9

1768 10.92 47.33 53.34

12.37 2.18 1.49 2.65

2.21 1.46 1.16 1.10

FLOORING  2 Concrete Mosaic Brick flooring (flat) Brick on edge paving Wood flooring Parquet Net cement finishing

m2 m2 m2 m2 m2 m2 m2

22.40 181.20 180.80 249.97 33.37 86.10 28.73

61.85 529.2 177.8 240.3 83.66 144.2 29.83

26.30 255.0 159.8 224.6 52.07 101.5 25.67

2.76 2.92 0.98 0.96 2.50 1.68 1.04

1.17 1.40 0.89 0.90 1.57 1.18 0.90

FINISHING Cement plaster (stucco) Cement paints, 2 coats Acrylic paint, 2 coats Enamel paint, 2 coats Aluminum paint, 2 coats Bituminous Asphalt, 1 coat

m2 m2 m2 m2 m2 m2

179.37 17.54 27.98 22.92 86.08 12.91

96.38 22.98 63.29 55.31 114.2 24.67

82.42 19.28 33.81 27.16 97.22 20.53

0.54 1.31 2.26 2.41 1.33 1.91

0.46 1.1 1.21 1.20 1.13 1.60

DISMANTLING Reinforced concrete Plain concrete Brick masonry

m3 m3 m3

827.43 780.11 125.37

2788 1638 386.0

1121 887.4 113.7

3.37 2.10 3.08

1.34 1.34 0.91

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activities performed in developing countries. But in the case of larger firms, the comparative labor requirement factors is close to unity (1.00), which is due to the extensive use of technologically advanced equipment in firms located in cities of developing countries. The data show that smaller firms are more labor intensive than larger organizations. Also, larger firms in India are more productive. In addition to the technical lag in smaller firms, the labor wage is low in India. For example, the wage of a skilled laborer is approximately $13 to $14 US dollars and an unskilled laborer earns 4 to 5 US$ for an 8-hour work day. Therefore, proper nutrition is not affordable for laborers who work in smaller firms. They often get tired after working for a few hours, and also lacking a positive attitude towards work, tend to experience lower productivity. Due to the large supply of workers, labor intensive construction methods may often be followed, and the use of construction equipment is limited (Koehn & Regmi 1990, Punmia 1987). In addition, workers often take a high risk at the work place, which tends to increase the level of accidents and fatalities on the job site. In fact, research has shown that effective productivity tends to be below 55 percent when the average earnings of the labor force are low (Dallavia 1957). Sometimes laborers also are required to work overtime without overtime payment (The Employee 1938). Another major cause of low productivity is climate. India is relatively hot, especially in summer. This causes labor fatigue resulting in a decrease in productivity. Lack of training also contributes to a decrease in productivity. Contractors of smaller firms do not provide any training in the use of equipment, if it is available. Also, laborers may not use hard hats, goggles, hand gloves, and boots, due to their unavailability at some construction sites (Safety 1996). Here, the productivity can be increased with the use of more efficient construction equipment, following improved methods of construction, adopting safety measures, education and training, use of modern materials, and use of planning and scheduling techniques, etc. 3.2

Labor production (productivity)

The labor production (productivity) data shown in Table 2 is quite varied. For example, the results indicate that for activities requiring minimum use of tools and equipment, such as hand leveling/grading (factor = 1.03) and flat brick flooring (0.98), the comparative labor requirement in a developing country is relatively low. As the labor wage in India is approximately one-twenty-fourth (1/24) of that in the US, it can be concluded that the labor cost of these activities is much lower than that in developed countries. For activities where improved tools and equipment can be used, such as excavation (factor = 2.46) and cement paint (factor = 1.31), the comparative labor requirement factor may be relatively high. This may indicate

that, for some activities, the use of improved methods and tools could be adopted. For activities where prefabrication is possible, the labor requirement factor is very high in developing countries. As an example, timber roof trusses (12.37) and formwork (3.14) have high labor requirements. One reason for this may be that wood is not a common construction material since good quality timber is very expensive. In developing countries, finished lumber of specified sizes and lengths are not available and has to be finished at site, which requires a great deal of work hours. Activities where extensive use of equipment are possible, such as concrete (factor = 6.14), presently experience a high comparative labor requirement factor. This is because, although mixing is done mechanically, transportation and placement is done manually. But in the case of larger firms, batch plants and onsite mixers are used. For mosaic flooring (2.92), polishing machines and stone cutting equipment are already used; however, the use of air-dryers and other improved tools could possibly increase productivity. It is seen that stone work (4.5) requires more labor than brick work (1.15). This is because brick construction has a long tradition and masons tend to be very skillful. The activities listed in Table 2 are generally associated with the construction of reinforced concrete or brick masonry buildings. Steel structures are rarely used due to the high cost of steel and perceived maintenance problems. Although the labor requirement is relatively high in developing countries, labor costs are much lower than in developed countries. This indicates a tradeoff between high labor requirements and low labor costs. Nevertheless, by improved planning and optimum use of labor, a desirable productivity can be achieved. 3.3

Different activities/categories

The 33 different activities listed in Table 2 may also be subdivided into four categories as illustrated in Tables 3–6, based on the possibility of using improved tools and equipment and possible prefabrication techniques. As shown, the average labor requirement factors for smaller firms range from 3.76 for activities where extensive use of equipment is possible to 0.99 for activities requiring minimum use of tools. The average labor factors for larger firms range from 1.65 to 0.92 respectively. Assuming, based on experience at a typical building site, for smaller firm, a distribution of approximately 9 percent for activities requiring minimum usage of tools and equipment, 16 percent for activities where use of improved tools and equipment is possible, 3.5 percent for activities where pre-fabrication is possible, and 40 percent for activities where extensive use of equipment is possible, the overall comparative labor requirement factor of a developing country, such as

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Table 3. Activities requiring minimum use of tools and equipment.

Table 6. Activities where extensive use of equipment is possible.

Labor requirement factor

Labor requirement factor

Activities

Smaller firms

Larger firms

Activities

Smaller firms

Larger firms

Leveling/Grading Brick flooring (flat) Brick on edge paving Average

1.03 0.98 0.96 0.99

0.96 0.89 0.90 0.92

Stone work Concrete (foundations) Concrete (slab/beam)  2 concrete Mosaic Dismantling, RCC work Dismantling, plain concrete Dismantling, brick masonry Average

4.50 6.14 5.20 2.76 2.92 3.37 2.10 3.08 3.76

3.50 1.93 1.6 1.17 1.40 1.34 1.34 0.91 1.65

Table 4. Activities where use of improved tools and equipment is possible. Labor requirement factor Activities

Smaller firms

Larger firms

Excavation (silty soil) Excavation (sandy soil) Earth filling Brick work Parquet Net cement finishing Cement plaster (stucco) Cement paints, 2 coats Acrylic paint, 2 coats Enamel paint, 2 coats Aluminum paint, 2 coats Bituminous Asphalt, 1 coat Average

1.99 2.46 1.44 1.15 1.68 1.04 0.54 1.31 2.26 2.41 1.33 1.91 1.63

1.16 1.2 1.1 1.02 1.18 0.90 0.46 1.1 1.21 1.20 1.13 1.60 1.11

4

Table 5. Activities where prefabrication (Standard Size) is possible. Labor requirement factor Activities

Smaller firms

Larger firms

Reinforcement Form work (beam) Form work (foundation) Form work (slab) Timber roof trustees Eaves Plywood partitions Plywood ceiling Wood flooring Average

4.45 3.06 3.14 2.98 12.37 2.18 1.49 2.65 2.50 3.87

3.15 1.65 1.53 1.50 2.21 1.46 1.16 1.10 1.57 1.70

SUMMARY AND CONCLUSION

The level of construction productivity varies for different countries depending, in part, on the degree of development and industrialization. This article presents data concerning construction productivity for a developing country like India and is compared with that of a developed country such as the US. For example, the findings suggest that the overall comparative labor requirement factor for smaller firms in India is 2.56. In contrast, larger firms, which tend to be less labor intensive, experience a factor of 1.35. This indicates that there may be, overall, approximately 3 times more workers on a construction site in a typical developing country compared to that of a developed region. In summary, the data presented involving productivity and safety should be of universal interest to any contractor considering projects in developing countries. Developing countries normally exhibit low labor productivity. Utilizing the labor requirement factors in this report will facilitate the efficient management, estimating, scheduling, monitoring, and updating of resources required for construction operations. The application of this data should assist international contractors in achieving their goal of successfully completing quality construction projects in developing countries. ACKNOWLEDGEMENT The authors wish to recognize Ms. Linda Dousay for her assistance with the production activities involved in preparing this paper.

India, may be calculated as a weighted average to be equal to 2.56. In the case of larger firms the distributions are approximately 18 percent, 20 percent, 32 percent, 30 percent respectively, and the overall comparative labor requirement factor is calculated as a weighted average to be equal to 1.35.

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REFERENCES Arditi, D. 1985. Construction Productivity Improvement. Journal of Construction Engineering and Management, No. 1.

Building Construction Cost Data. Kingston, MA: R.S. Means Company, Inc. Building Construction Cost Data. 1986. R.S. Means Company, Inc. Kingston, MA. Chandra, H. 1990. Management of Construction in Developing Countries: Indian Experience. Proceedings, International Council for Building Research Symposium. CIB90, Sydney, Australia, 5, 1957: 211–224. Dallavia, L. 1957. Estimating General Construction Costs (2nd edition). New York, F.W. Dodge Corporation. Fatal Injury Act. 1955. Government of India, New Delhi, India. Hinze, J., and J.B. Russell. 1995. Analysis of fatalities recorded by OSHA. Journal of Construction Engineering and Management 121, no. 2: 209–214. Jaselskis, E. J., & G.A.R. Sauazo. 1993. Comparision of Construction Safety Codes in the US and Honduras. Journal of Construction Engineering and Management. ASCE, 119, no. 3, 560–573. Koch, J.A., & F. Moavenzadeh. 1979. Productivity and Technology in Construction. Journal of the Construction Division No. 4, 105: 351–366. Koehn, E., and D.C. Regmi. 1990. Quality in Constructed Projects: International Firms and Developing Countries. Journal of Professional Issues in Engineering 116, no. 4, 388–396.

Koehn, E., and J. Manuel. 1989. Subcontractor’s Concerns During Construction. Proceedings of Construction Congress I. ASCE. Koehn, E., R.K. Kothari, & C.S. Pan. 1995 Safety in Developing Countries: Professional and Bureaucratic Problems. Journal of Construction Engineering and Management 121, no. 3, 261–265. Koehn, E., & J. Manuel. 1989. Subcontractor’s Concerns During Construction. Proceedings of Construction Congress I. ASCE: 278–283. Punmia, B.C. 1987. Building Construction. Vol. 3. New Delhi, India: Lakshmi Publishers. Rate Analysis for Construction Works. 2007–08. Public Works Department, Government of the People’s Republic of India, New Delhi, India. Sharp, H.T. 1997. Finding Building Product and Construction Cost Data Internationally. Cost Engineering No. 1, 39: 25–39. Safety First. 1996. Safety handbook of Beximco Engineering, Ltd., Bombay, India. Schedule of Rates. 2007–08. Government of Andhra Pradesh PWD, presidency division, Mumbai, India. Sheer & Gahlot. 1993. Construction Planning and Management. Delhi, India: New Age International Publishers. The Employee Liability Act. 1938. Government of India, New Delhi, India.

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Contractors’ influence within the design process of design-build projects H. Haroglu, J. Glass & T. Thorpe Civil and Building Engineering Department, Loughborough University, UK

C. Goodchild The Concrete Centre, UK

ABSTRACT: Nowadays buildings are much more complex than ever and many diverse skills are needed to design them. During the design phase we have the greatest opportunity to influence the final cost of the constructed facility. Also the evaluation of alternative frame types which has a huge influence on the value of the building to the client is involved in design phase. The research initially examined the structural frame selection process through an industry survey. Design-build procurement approach was found to be used for about 50% of UK construction projects. This suggested the importance of the contractor’s potential influence in the decision-making process. A case study approach has therefore been adopted to examine whether the contractor influences or actually changes any specifications on a design-build project. This paper presents findings on the main contractors’ influence over the decision-making process and will therefore be of interest to designers, contractors and clients.

1

INTRODUCTION

Nowadays buildings are much more complex than ever and many diverse skills are needed to design them. During the design phase we have the greatest opportunity to influence the final cost of the constructed facility (Austin et al. 2001). A change for the better in a project is least costly whilst it is still at the design stage. Traditional approaches to the management of design are not sufficient as they have evolved more slowly than the industry and society as a whole (Gray and Hughes 2001). Nonetheless, decisions made at this stage, particularly those involving costs and speeds, are often based on subjective judgement, rules of thumb and familiarity (Idrus and Newman 2002). It is therefore essential to examine the decision making process in the design phase, so a four-year research programme initiated by Loughborough University and The Concrete Centre has investigated the key drivers and barriers in the concrete frame procurement process in the pre-construction phase as this is a generally under-researched area. The procurement process plays a significant role in project success and determines the responsibilities of project team members (Rowlinson and McDermott 1999) so there is good reason to examine its possible influence on decision making process of building projects in the design phase. The research initially investigated the structural frame selection process through a state-of-the-art literature review, semi-structured interviews and a postal questionnaire survey. One of the main findings was that the design & build (D&B)

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procurement approach was used for about 50% of UK construction projects. Therefore, we tend to make an assumption that the contractor can potentially exert significant influence in the decision-making process during the design phase. This paper presents the findings from case studies of four D&B building projects undertaken in the UK. The overall research aim was to investigate the main contractor influence on major decisions within the decision making process on a D&B project. The paper begins with a description of the background for the study which is followed by research methodology and a summary of key findings. The results are then discussed in connection with some other recent studies. Major new findings on the main contractors’ influence over the structural frame selection process are presented which will be of interest to designers, contractors and clients/customers. 2

BACKGROUND REVIEW

Design is typically defined as ‘‘the formulation of an idea and turning it into a practical reality’’ (Blockley 2005). The design concept and design process in the construction industry have been defined in many ways. For instance, Gray and Hughes (2001) described the design as mainly a personal task with the whole projects’ design becoming a combination of the motivation and expressions of many individuals. Akin (1986) stated that design is trade-off between many conflicting needs until there is a solution that enables everyone to move forward to the next aspect of the

problem. On the other hand, design process is defined by Pahl and Beitz (1988) as the intellectual attempt to meet certain demands in the best possible way. Moreover, the design process is seen as a negotiation between problem and solution through the three activities of analysis, synthesis and evaluation. The common idea behind all these ‘maps’ of the design process is that it consists of a sequence of distinct and identifiable activities which occur in some predictable and identifiably logical order (Lawson 2006). Reducing cost and time by effective design management requires appropriate design choices, monitoring design development, applying suitable management practices to activities on and off site, building relationships with subcontractors and suppliers, and more (Best and Valence 2002). Both the influential Latham (1994) and Egan (1998) reports identified that improvements designed to reduce budget and timescale and to increase quality would only be achieved if main contractors were involved sufficiently early in the design process and fully understood the needs of the Client. Hence, the rise in popularity of procurement routes and forms of contract that permit early contractor involvement (ECI) such as D&B within which contractors are involved early (and ideally appointed formally) to increase the level of supply chain integration. Although some confusion exists amongst inexperienced clients, the term D&B has almost been unanimously interpreted and defined as ‘‘An arrangement where one contracting organisation takes sole responsibility, normally on a lump sum fixed price basis, for the bespoke design and construction of a client’s project’’ (Masterman 2006). Although D&B has been used in the UK construction industry for decades, it has gained increased market share in the past ten years (Arditi and Lee, 2003). D&B arguably places more responsibility and liability on to the contractor than any form of procurement (Akintoye 1994; Peace and Bennett 2005). The key benefits include single point responsibility, availability of the contractor’s knowledge of ‘buildability’ and the standardisation of the construction process (Franks 1990; Janssens 1991; Akintoye 1994; Turner 1995). Furthermore, according to Peace and Bennett (2005), D&B projects based on a minimal statement are completed 40% faster, while those based on an outline design are completed 25% faster than projects using a traditional approach. Also, D&B projects are much more likely to be completed on time and are reportedly 15% cheaper than equivalent traditional projects. However, the D&B method also has a number of disadvantages. One of the major reported drawbacks is the poor quality of design (Franks 1990; NJJC 1995), the main reason for this being that architects seem to have less control over the design process than they would in a traditional approach (Haroglu et al. 2009). Also, the advantages of competition may not be passed onto the client when using D&B (Peace and Bennett

2005; Rawlinson 2006). The principal variants of the D&B (integrated) procurement systems are named by Masterman (2006) as: – – – –

Novated D&B; Package deals; Develop and construct; and, Turnkey.

It was discovered by Haroglu et al. (2009) from a recent survey of project managers, cost consultants and clients that D&B procurement route was the most popular form of contract used in the UK construction industry indicating how much the UK construction industry has transformed in the past decade. Originally intended for simple buildings, D&B is now used routinely to deliver complex, highquality projects, sometimes designed by signature architects (Rawlinson 2007). Also, there is a variety of tender and contractor arrangements being adopted with D&B procurement including Single-Stage (Competitive) and Negotiated Tendering, along with the more innovative Two-Stage Tendering and Partnering arrangements. As a result, effective design management is at the core of delivering a successful project (Austin 2002). With increasing D&B projects in the UK, the influence that the main contractor has during the design phase appears under-researched. Particularly it warrants consideration in terms of the various tendering arrangements adopted in Design-Build procurement routes, the size of the contractor, the client-main contractor risk relationship, the stage at which the main contractor is involved both informally and contractually, and so a case study approach has been selected to investigate these factors on a series of D&B projects in the UK. The adopted research methodology is described in detail in the next section.

3

ADOPTED APPROACH

The aim of this paper is to explore how contractor influences major decisions within the decision making process of D&B projects. To achieve this, the following research objectives were identified, in the context of D&B projects: – determine the degree and types of involvement that main contractors have within a range of D&B contracts; – identify the similarities and differences amongst various D&B projects in relation to the degree of contractor involvement; – identify what changes, if any, are typically made by main contractors during the design process and; – make recommendations on degrees of influence that contractors can/should have in the design process of building projects.

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These objectives were met by undertaking a series of company case studies with UK-based D&B contractors;‘‘Case studies are used when the researcher intends to support his/her argument by an in-depth analysis of a person, a group of persons, an organisation or a particular project’’ (Naoum 2007). Case studies are recognised as a suitable research methodology for this research as the case study is the method of choice when the phenomenon under study has not been investigated within its context (Yin, 2003; Fellows and Liu, 2003). Case studies can provide examples illustrating the influence of contractors in D&B projects and details of the drawbacks and benefits involved. Due to the diversity of company size and structure, an exploratory case study design based on multiple cases with single units of analysis has been adopted for the research in accordance with guidance offered by Yin (2003). That is, four construction contracting companies of different sizes and structures will be studied using the same case study protocol which was developed around the research hypothesis and a series of associated research questions. A shortlist of possible target companies for the case studies was produced from the results of a questionnaire survey (Haroglu et al. 2009), within which respondents had volunteered specific building projects for consideration in this stage of the research. These 23 companies represented a cross-section of UK contractors by size and the type of procurement routes used because they are small-medium and large contractors (categories which account for more than 90% (based on ranking in the construction top 100 of Construction News 2008) of construction contracts by turnover per annum, have been involved in a design & build projects recently and were willing to participate in the study. Multiple cases needed to be employed to ensure that the results presented a breadth and depth of main contractors’ involvement in design-build projects. In this study, four building projects were selected (in discussion with the sample of contractors) and four case studies undertaken using personal interviews with project team members who had been involved in choosing the structural frame material at the design stages or thereafter. The objective was to collate the views of the various project team members who were involved in the design stages or thereafter. So the interviews were held with a range of professional groups including, but not necessarily limited to; main contractor, structural engineer, architect and cost consultant to collate a broad range of perspectives covering different aspects of the contractor involvement. Interviews were held with the related project team members or their representatives on an individual basis. In all cases, a letter was sent, or telephone call made to each selected individual outlining the research and inviting them to participate. For confidentiality the parties concerned are referred to as Contractors A, B, C and D. There were significant differences in the

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contractors’ size as regards the number of employee and financial turnover. The types of building projects examined were also different, plus Contractor A and Contractor C were appointed through single-stage competitive tendering, whereas Contractors B and D were employed using two-stage tendering arrangement in the design-build projects. The four contractors together with the design-build projects they were involved are therefore deemed to present a breadth and depth of main contractors’ involvement in designbuild projects. Data were collected through interviews along with the researcher observations, documentation and records during interviews. Interviews are one of the main sources of case study information (Tellis 1997). These interviews were semi-structured using a number of key and supplementary questions. The choice of interviews as the primary source of data was determined by consideration of the scope and depth required for this case study research. Furthermore, interviews vary in their nature, they can be: structured, semi-structured and unstructured; the major differences lie in the constraints placed on the respondent and the interviewer (Fellows and Liu 2003). The structured interview does not provide sufficient scope to probe ideas further, as the questions are set quite tightly defined (Hancock, 1998). On the other hand, semi-structured interviews can yield a variety of kinds of information; even within one interview one can (Drever 1995): – Gather factual information about people’s circumstances – Collect statements of their preferences and opinions – Explore in some depth their experiences, motivations, and reasoning. In line with the overall aim of this research, the focus of the interviews was on any changes that had occurred during the design phase after both the informal and the contractual involvement of the main contractor in the project. The case study interviews lasted between 30 minutes to one hour, depending on the interviewees, were tape recorded and transcribed verbatim. The data (content) analysis involved exploring the themes and patterns revealed in the interviews in order to draw similarities and differences within and cross-case analyses. Also, the process of analysis involved continually revisiting the data and reviewing the categorisation of data until the researcher was sure that the themes and categories used to summarise and describe the findings were a truthful and accurate reflection of the data (Hancock 1998). In this study every effort was made to avoid bias that might influence the qualitative data analysis and the researcher was aware of avoiding prejudice stemming from either the researcher or the interviewee. A summary of findings is presented and discussed in the following section.

4

MAIN CASE STUDY FINDINGS

This section explains the findings of case studies conducted to examine the main contractors’ influence on a D&B project. The degree and type of involvement that each contractor had in the case study projects is described here, together with any changes made by the contractor during the design phase. Because of the extent of the data collected from the case study interviews, only the key findings are discussed and bullet-pointed here. A more detailed analysis forms the body of a forthcoming paper. 4.1

Case study A

This development project is comprised of seven buildings, six residential buildings with retail units on the ground floor and one office building. Initially five residential buildings were designed and subsequently tendered through single-stage tendering. The last two buildings (residential and office buildings) were brought on board later when Contractor A had already started on site and were procured using two-stage tendering. The project team members interviewed for this case study included; Main contractor (Contractor A), Architect, Structural Engineer and Cost Consultant. As stated by the project team members interviewed, although Contractor A claimed that they got involved at stage D of RIBA stages in the first five building projects, they were not contractually appointed until the end of stage E of RIBA Plan of Work (RIBA 2007). As a result, Contractor A prompted a few minor changes in the first fiveresidential-building project including the concrete specifications (using ggbs) and the method of construction (using slipform construction). For the final two buildings in this case, Contractor A was involved from the beginning, although not contractually; during this time, Contractor A asserted that they had a major impact in the structural frame selection process, which was corroborated by the architect and structural engineer that Contractor A did influence the frame choice of the office building as well as buildability. 4.2

Case study B

This was a laboratory building with in-situ concrete frame flat slabs, procured through two-stage tendering. The project team members interviewed for this case study were; Main contractor (Contractor B), Architect, Structural Engineer, Project Manager (External) and Client’s representative. Contractor B became involved in the project from RIBA Stage C onwards, but contractually took over after Stage E. According to the project team members interviewed Contractor B did not make any major change during the design process, rather they influenced finishes, materials used,

and gave useful advice on market prices. The Client’s representative stated that Contractor B’s involvement did help all the project team members to work together as a team, to resolve problems before the work started on site. The external project manager also believed that Contractor B had a positive influence and made an accurate market assessment. Contractor B said that their early involvement helped them get appropriate sub-contractors and thinking some of lead-in during the design process. In the words of Contractor B, ‘‘The envelope of the building was originally designed to have dual curve which was proven extremely expensive so the design team (architect, structural engineer, services engineer, etc.) chose to go for a single curve which was a lot cheaper than dual curve. We used our supply chain. We had key packages and we knew the workload’’. 4.3

Case study C

A three-storey steel frame hospital building which was procured through singlestage Design-Build. The project team members interviewed for these case studies included; Main contractor (Contractor C), Architect and Structural Engineer. Contractor C was appointed at the end of RIBA Stage E but reportedly did not have any influence in the design process until the contract was placed. Thereafter, Contractor C tried to make the developed design more buildable and economical for the client, in the words of Contractor C; ‘‘Quality is also paramount and using cheaper materials does not necessarily mean that it would undermine the quality’’. In addition, the foundation design was changed by Contractor C due to difficult ground conditions. 4.4

Case study D

This was an in-situ concrete frame residential block with retail units on the ground floor. It was procured through two-stage design-build, but competitively tendered at RIBA Stage C. The project team members interviewed for this case study were; Main contractor (Contractor D), Architect, Structural Engineer and Client. Although it used two-stage tendering, Contractor D was not involved early in this project. The client indicated that it was a straightforward building so they did not have to get the contractor involved in the early design of the project and the structural engineer stated that the majority of design had been developed by the time Contractor D was appointed. Therefore Contractor D had little influence in developing the design in this case as there was little scope for them to change anything. The client pointed out that Contractor D did however influence the sequence, method of construction and the design of the structural frame, plus they took control of the design by using their own supply chain. In addition the structural engineer claimed that

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Contractor D was informally involved during Stage D and they knew that they were the preferred contractor so they refined their costs somewhat. Interestingly the client added ‘‘I am not a fan of two-stage DesignBuild as we ended up spending more money’’ and also asserted that they would go for single-stage Design Build in the future. 5

the two-stage tendering. As Rawlinson (2006) stated that the benefits of the two-stage approach are most likely to be secured when the contractor is proactive in its engagement with the design, buildability and financial aspects of the project. In addition Mosey (2008) asserted that properly structured two stage tendering, using an early conditional contractor appointment, is the best means for clients to control projects and obtain added value from their contractors.

DISCUSSION

It is not possible to provide a full cross-case analysis, but a number of important results from the study are presented here. According to the main contractors interviewed, they get involved in a substantial number of D&B projects. This corroborates the evidence from both the findings of Haroglu et al. (2009) and others in showing that D&B procurement is extensively used in the current UK construction industry. From the study undertaken by Akintoye (1994); D&B was not favoured by the contractors as it does not put contractor’s organisation in charge of the whole project. However, there is a range of advantages of using D&B in order to enhance the implementation of projects (Rowlinson 1997; Leung, 1999). Whilst Contractor C stated that D&B in general is proactive and challenging, in the words of Contractor B, ‘‘we are very happy with Design-Build approach, particularly twostage D&B as it allows us to drive and control the design which also means that we control our own destiny’’. Contractor A added, ‘‘The advantage of D&B for the contractor is that we can manage the process effectively to alter the design if we get involved in the project in good time’’. Although it was recognized that main contractors do not generally get involved early in the design process of D&B projects, particularly in projects using single-stage tendering, they were found to be influential over the issues of buildability, programme and the materials used in all the case study buildings, plus were able to provide advice on market prices, all of which would offer cost savings to the client. Whilst the contractors’ level of influence was found to be dependent on their readiness and ability to affect the design process, it was also recognized that it is the client’s preference/motivation that determines the whether or not the main contractor can exert influence in D&B projects. Furthermore, according to the interviewees in the case study projects early contractor involvement in general was found to be hugely beneficial as the contractors have specialist knowledge in construction methods and techniques as well as market prices. Yet early contractor involvement does not happen very often in practice, as in the words of the structural engineer in the Case Study D ‘‘From a designer point of view it would be great to have contractor earlier however this is not the case for the developers’’. Nevertheless it was found that D&B contractors are generally involved early in the design process under

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6

CONCLUSIONS

This paper has shown with the help of case studies, the influence of main contractors on D&B building projects. The study provided an in-depth understanding of the degree and type of involvement main contractors have within a range of D&B building projects. It identified the similarities and differences amongst the various sizes of contractors in relation to the degree of their involvement. It also showed the benefits of early contractor involvement in general for building projects. That said, it was clear from the case study findings that early involvement allowed the contractor to offer advice on buildability, market conditions and an appropriate supply chain. With D&B contracts continuing to be popular in the UK building construction and most of the rest of the world, contractors are expected to be somewhat influential in the design process of building projects particularly in projects using two-stage tendering. Contrary to popular opinion, contractors in the case studies appeared to be fairly content with the D&B procurement route. However, D&B contractors should be well adapted to their key role and should take a proactive approach to other project team members about design, buildability and financial aspects of the projects. A series of practical recommendations can be provided for construction contracting companies to help better position themselves on the D&B projects they undertake as well as understanding the requirements of the client. The influential contractor/early contractor involvement can enable design team to produce better designs in a shorter time with reduced cost which the improvements mentioned in Latham (1994) and Egan (1998) reports could be achieved. 7

ACKNOWLEDGEMENT

The research project reported in this article was funded by the Engineering and Physical Sciences Research Council (EPSRC) along with The Concrete Centre (TCC) and this support is gratefully acknowledged. The authors are grateful to all of those involved with the project. The authors would also like to give their special thanks to all the people that had spent their valuable time to participate in this study.

REFERENCES Akin, 1986. Ö. Akin, Psychology of architectural design, London: Pion Limited. Akintoye, A. 1994. ‘‘Design and build: a survey of construction contractors’ views’’, Construction Management and Economics, Vol.12 No. 2, pp. 155–63. Arditi, D. and Lee, D.E. 2003. Assessing the corporate service quality performance of design-build contractors using quality function deployment. Construction Management & Economics. 21(2): 175–185. Austin, S., Baldwin, A., Hammond, J., Murray, M., Root, D., Thomson, D. and Thorpe, A. (2001). Design Chains: A handbook for Integrated Collaborative Design. Hampshire: Hobbs the Printers. Best, R. and Valence, G.D. 2002. Design and Construction: Building in Value. Oxford: ButterworthHeinemann. Drever, E. 1995. Using Semi-Structured Interviews in SmallScale Research: a teacher’s guide, Scottish Council for Research in Education, Glasgow: SCRE publication. Egan, J. 1998. Rethinking Construction, The report of the Construction Task Force on the scope for improving quality and efficiency in UK construction, Department of the Environment, Transport and the Regions, London. Fellows, R. and Liu, A. 2003. Research Methods for Construction. Oxford: Blackwell Science Ltd. Franks, J. 1990. Building Procurement Systems—A Guide to Building Project Management, Chartered Institute of Building, Ascot. Hancock, B. 1998. Trent Focus for Research and Development in Primary Health Care: A Introduction to Qualitative Research. Trent Focus, Nottingham University. Haroglu, H., Glass, J., Thorpe, T., Goodchild, C. and Minson, A. 2009. A study of professional perspectives on structural frame selection (Forthcoming). Idrus, A. and Newman, J. 2002. IFESS: a computer tool to aid structural engineers at the conceptual design stage. Construction Innovation, Vol. 3, 127–143. Janssens, D.E.L. 1991. Design-Build Explained, London: Macmillan. Latham, M. 1994. Constructing the team, Joint Review of Procurement and Contractual Arrangements in the UK Construction Industry, the Stationery Office, London.

Lawson, B. 2006. How designers think: the design process demystified. Oxford: Elsevier. Leung, K.S. 1999. Characteristics of design and build projects. Seminar Proc., on Design and Build Procurement System, Hong Kong, 1–10. Masterman, J.W.E. 2006. An Introduction to Building Procurement Systems, London: E&FN Spon. Mosey, D. 2008. It’s just too sad to be single, Building Magazine, 49: 47. Naoum, S. 2007. Dissertation Research and Writing for Construction Students, 2nd edition, Oxford: ButterHeinemann. NJCC. 1995. Code of Procedure for Selective Tendering for Design and Build. National Joint Consultative Committee for Building, London. Pahl, G. and Beitz, W. 1988. Engineering design: A systematic approach, London/Berlin: Design Council/SpringerVerla. Peace, S. and Bennett, J. 2005. How to use the construction industry successfully: a client guide, CIOB, Ascot, Berks. Rawlinson, S. 2006. Successful projects, Building Magazine. 6: 73–77. Rawlinson, S. 2007. Procurement Public sector projects, Building Magazine, Vol. 47, pp. 52–56. RIBA 2007. The RIBA Plan of Work Stages 2007. See . Rowlinson, S. 1997. Procurement systems: the view from Hong Kong. Proc., CIB W92 Procurement—A key to innovation, Univ. de Montreal, Montreal, 665–672. Rowlinson, S. and McDermott, P. 1999. Procurement Systems: A Guide to Best Practice in Construction, London: E & FN Spon. Tellis, W. 1997. Introduction to case study [68 paragraphs]. The Qualitative Report [On-line serial], 3 (2). Available on line: Turner, D.F. 1995. Design and Build Contract Practice, Harlow: Longman Scientific and Technical. Yin, R.K. 2003. Applications of Case Study Research, 2nd Edt., London: Sage Publications.

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Delays in the Iranian construction projects: Stakeholders and economy E. Asnaashari, A. Knight & A. Hurst Nottingham Trent University, Nottingham, United Kingdom

ABSTRACT: The construction industry has a major role in the Iranian economy by generating employment and wealth. However, delay is one of the most reoccurring problems in the Iranian construction industry and has negative impacts on project success in terms of time, cost, and quality. This paper presents one part of an investigation into the factors which cause delay in construction projects in Iran. In that study, eleven in-depth interviews were carried out with Iranian construction practitioners. Qualitative analysis of responses categorized causes of delay in five groups as: (1) Stakeholders; (2) Economy; (3) Politic; (4) Management; (5) Miscellaneous. This paper, by focusing on the two first groups, develops a comprehensive interpretation to reveal the role of stakeholders and economy in causing delay in construction projects in Iran.

1

INTRODUCTION

Project scheduling is the first challenge at the inception of most construction projects. Without a systematic planning and scheduling, managing construction projects is hardly possible. As Keane & Caletka (2008) pointed out, high level of experience and accurate time and cost estimating are necessary to plan a project effectively. However, projects’ schedules often are affected by several factors which can endanger the success of the project. One of the most important factors that seriously affect projects’ schedule is delay. Delay is ‘the time overrun either beyond completion date specified in a contract, or beyond the date that the parties agreed upon for delivery of a project’ (Assaf & Al-Hejji 2006). Delay is also defined as an ‘act or event which extends required time to perform or complete work of the contract manifests itself as additional days of work’ (Zack 2003). In spite of utilizing advanced technology and understanding of project management techniques, construction projects continue to suffer from delay. Today’s, clients are not content only with minimal cost and adequate functional performance for their projects. Interest rates fluctuations, inflation and other commercial pressures mean that, in most projects, it is more cost-effective to complete the job within the shortest possible time. Hence, a project that is not completed on time can hardly be called successful from a project management point of view. It means that delay is a risk that always threatens the success of the projects. In fact, delay is an inherent risk in construction and should be treated in a similar fashion as other risks. It can be managed, shared, minimized, or accepted but must not be ignored (Keane & Caletka 2008). Poor attention to the risk of delay may have negative impacts such as time overrun, cost overrun,

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increase in disputes, arbitration, litigation, and project abandonment (Sambasivan & Soon 2007). Delay is often the result of an unpredicted event that does not necessarily indicate that the management system is ineffective; but the way that a management team responds to delays can expose its strength or weakness in relation to delay analysis and management. For a systematic delay analysis, first the cause of delay should be identified and documented. Then, the impact of delay on the schedule should be recognized and the risk of further delay should be considered. Systematic analysis will provide valuable information about different causes of construction delay that can be referred to in the future projects. On most construction projects, there are common causes of delay that frequently happen. Identifying these causes may help practitioners to anticipate potential delays and plan to minimize their effects. Therefore, identifying delays and investigating their causes is an important part of delay analysis process. This study aims to interpret some causes of delay in the Iranian construction projects that are associated with stakeholders and economy. The result may be used in the first stage of delay analysis in different construction projects in Iran. 2

THE CONSTRUCTION INDUSTRY IN IRAN

Many researchers have confirmed that delays in construction projects are a global phenomenon. The construction industry in Iran is no exception. The Iranian construction industry, with an annual turnover of US$38.4 billion (Australian Government-Austrade 2007), is continuing to grow with an average growth of 4.40% over 2008 to 2012 (Companies and Markets 2008). However, the process of construction in Iran is

very slow and expensive. This along with other issues such as the sharp increase in the price of land, construction materials and machines, have posed a dilemma for the low-income stratum in Iran (Aftab News 2006). The problem is more crucial especially in the housing sector. From March 2004 to March 2005 total Iranian households were 15.1 million and the total numbers of dwelling units were 13.5 million (Statistical Centre of Iran (SCI) 2006). Moreover, every year there is a need for 750,000 additional units as young couples embark on married life. This shows a huge difference between demand and supply in the residential sector. In addition, Iran’s geographical position over a seismic belt necessitates the reinforcement and renovation of buildings in Iran (Australian Government-Austrade 2007). To feed this enormous demand, the Iranian government has tried to encourage construction firms to improve productivity and efficiency of their projects. In July 2008, the Iranian President, Mahmoud Ahmadi Nejad, announced that his government is looking for different ways to reduce time and costs of construction projects, especially in residential projects, by insisting on industrializing the construction of buildings (ISNA 2008). Before this announcement, the government spokesman had publicized that 30 prefabricated building factories will be open by April 2009 (Fars News Agency 2008). The government and practitioners are also looking for ways to reduce the time of construction by minimizing delays in projects. As it was explained before, identifying different causes of delays is the first step to avoid time overrun and the other negative impacts of delays in construction projects. The aim of this paper is to introduce the causes of delays in Iranian construction projects from stakeholders and economy point of view. By recognizing these causes, practitioners will be able to foresee delays in their projects and minimize their negative impacts or at least adapt themselves to their consequences. 3

PREVIOUS RESEARCH

Identifying causes of delays in the construction industry has been topic of much research in different countries. Odeh and Battaineh (2002), Kaming et al. (1997), Alaghbari et al. (2007) Sweis et al. (2008) stressed the importance of early identification of construction delays to avoid or reduce negative impacts of them on projects. Several researchers studied causes of delay in countries in the Middle East and Persian Gulf region. The results of a study in Jordan indicated that the main causes of delay in construction projects in this country relate to designers, user changes, weather, site conditions, late deliveries, economic conditions and increase in quantity (Al-Momani 2000). Mezher & Tawil (1998) explained that preparation and approval of drawings, slowness of the owner’s

decision-making process, obtaining permits or approval from different government authorities, non-availability of materials on time, and unskilled manpower are major causes of delay in Lebanese construction projects. A study in 2006 in Saudi Arabia by Assaf and Al-Hejji revealed the highest frequent factors of delay in this country as: awarding contracts to the lowest bidder, changes in orders by owners during construction, delay in payments, ineffective planning and scheduling by contractors, poor site management by contractors, shortage of labor and difficulties in financing. In the UAE, preparation and approval of drawings, slowness of the owner’s decision-making process and inadequate early planning of the project, shortage of manpower, conflict between contractors and the consultants are the major causes of delay (Faridi & El-Sayegh 2006). In Egypt, El-Razek et al. (2008) identified the most important causes of delay as financing by contractors during construction, delays in contractors’ payment by owners, design changes by owners or their agents during construction, and no utilization of professional construction. Clearly, some delay causes are common between different countries and some others not. The reason is that delay is a factor that has a close relationship with working culture, stakeholders, the government policy, economy situation and availability of resources (man, money and machine). These are concepts that often vary from one country to another. Hence, it is not so surprising that some causes of delay may be more significant or more frequent in one country in comparison to the others. Faridi and El-Sayegh (2006) had a similar interpretation when they compared causes of delay in the UAE to Saudi Arabia. Therefore, for a better result, identifying causes of delay should be done in a specific country, which is Iran in this research. An important point to rise about the literature is that the majority of research in the area of delay has been conducted using quantitative methodologies and attempted to rank causes of delay from three points of view: clients, consultants and contractors. Consequently, there is a lack of qualitative research to interpret the causes of delays from different perspectives. This paper, in contrast with other studies with similar topic, will use qualitative methodology to provide in-depth understanding of delay causes in construction projects in Iran. 4

RESEARCH METHODS

To grasp an in-depth understanding of the reasons for delays in construction projects in Iran a qualitative research strategy was undertaken that used a small, but focused and carefully selected sample. In the first step, literature associated with delay in construction were found and reviewed. The aim of the literature review was to find different causes

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of construction projects’ delay in the other countries, specifically in Persian Gulf region where Iran is located as well. Fortunately, some research was conducted in countries such as Jordan, Lebanon, Saudi Arabia, UAE and Egypt. However, because there is a limited amount of literature about construction projects’ delay in Iran, it was essential to get help from experienced people in the field of construction in the country. Eleven open-ended interviews were conducted with practitioners who are involved with day-to-day issues of the construction industry in Iran. All participants have 10 or more years of experience and are involved in residential, commercial, and road building projects in Iran. Data gathered from interviews was analyzed by using a qualitative data analysis method. First, responses were classified under five relevant categories as (1) Stakeholders; (2) Economy; (3) Politic; (4) Management; (5) Miscellaneous. Under these categories, new sub-categories were developed by progressing through the transcript of interviews. Then, a comprehensive interpretation was developed to produce well-grounded conclusions. Wherever it was suitable, the participants’ direct quotes are cited anonymously to make the interpretation more meaningful. Figure 1 illustrates the structure of the categories and sub-categories. It should be explained that the result of this study is restricted to the participants’ experiences and their viewpoints and cannot be generalized in wider contexts. In addition, the result may not be the whole reality as, in social studies like this, there may be multiple realities. However, to achieve reliable data, interviewees were selected carefully among construction practitioners who have three specifications: a) to have worked as a senior manager in a construction specialist company, b) to have 10 or more years of experience in construction, and c) to be completely familiar with culture and environment of construction in Iran. Reliable level of data saturation was achieved within 11 interviews.

Stakeholders

Causes of Delay

5

NEGATIVE IMPACTS OF DELAY IN CONSTRUCTION PROJECTS IN IRAN

The majority of interviewees confirmed that delays have negative impacts on construction projects. The first negative effect of delay is the increase in the rate of disputes. Dealing with disputes and claims is a challenge for all parties involved in the projects. Respondents expressed that dealing with disputes is expensive and time consuming and detract from the ability of a manager to successfully administrate projects. Interviewees also believed that delay may raise the cost of construction in the case of a high inflation rate when contractors should spend more money to buy construction materials and equipments. In addition, respondents emphasize that delay raises the overhead cost of projects, which includes office expenses, administrative costs, utility bills, maintenance, and repair expenses. One of the respondents, who is the senior manager of a general contractor company, emphasized the importance of time in construction by explaining a metaphor about contractor activities in Iran: ‘‘Generally, construction contracting in Iran is like a bucket with a hole at the bottom. Whenever you draw it up sooner from the well there will be more water in it.’’ Some respondents stated that delay in infrastructure and public projects is more crucial. For instance, delay in building a bridge will cause both financial and social problems for the Government. In the agriculture sector, the problem is more crucial. For example, delay in the construction of a damp causes many problems for farmers in terms of irrigation of agricultural lands that will lead to huge loss. However, few respondents thought that delay is not always negative in terms of profit. They explained that delay is a relative concept and depends on the political and economic conditions in which a project is being built.

Inflation

Economy

Contractors

Consultants Clients

Resources

Finance

Material

Categories Sub-Categories

Figure 1.

Manpower

Categories and sub-categories (emerged from the qualitative data analysis).

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Machines

‘‘Imagine that you are constructing a four-storey building. You have a choice to finish it in six months and sell it at £100,000 or finish it in 18 months and sell it at £140,000. Which one do you prefer? The second choice gives the opportunity to you to work less and earn more. So the second choice seems better.’’

‘‘We were hired by a client to build a multi-storey parking. When the steel structure was completed the clients asked us to change the functionality of the building to residential. It lasted more than one year to change the drawings, design for M&Es, change the material, prepare new contract documents, etc.’’

In fact, in a chaotic market with a high rate of inflation and huge gap between supply and demand, the above argument might be true for property developers and owners. This situation happened in the residential market between 2006 and 2008 in Iran. Yet, it should be considered that the situation of the market is always changing and, in the case of long delay, this idea can be challenged easily when the price of materials and services also increase.

Another problem is that the process of the clients’ decision-making is often slow. Selection of wrong consultants and contractors may also happen because most clients are looking for the lowest price within tenders. Therefore, contractors with low competency will get the job and this is the starting point of a chain of problems that finally causes delay for the project. Delay in payments by clients is a common problem in most projects as respondents conformed. The reason may arise from poor cost estimation or lack of different sources of finance that will be explained in Economy section. Another issue is the changing of the priority of project objectives by clients. This problem is more critical in complex projects like airport or oil platform construction where sophisticated facilities may be imported and installed into the structure. This also causes problems in allocating resources:

6

CAUSES OF DELAY IN IRANIAN CONSTRUCTION PROJECTS

The role of stakeholders and economy in construction projects is sensitive. Stakeholders, as the groups that may have different and sometimes discrepant interests in projects, are able to either avoid or contribute to delays in construction projects. The condition of the economy is also important in causing delays for projects. Recession, inflation rates, interest rates and availability of resources are concepts that seriously affect the process of execution of projects. Here, causes of delay that associate with stakeholders and economic condition will be focused.

6.1

‘‘I was involved in construction of a residential building complex. We had been involved with concrete framework of the first building that the client asked us to start excavation for constructing the foundation of another building at that complex. We did not have excavator onsite and we did not have cash to hire one as we had spent all of our money on concrete and cement. It was a terrible situation that caused a big delay in that project.’’

Stakeholders

Stakeholders are parties that are involved in construction projects and affect, or can be affected by the project. This section describes the role of the most important stakeholders (clients, consultants, and contractors) in causing delays in construction projects in Iran. Based on information obtained from the interviews, clients can cause delay in several ways. A majority of clients does not have enough knowledge about construction legislation that is enforced by the municipality or other authorities. Hence, sometimes they request something that is basically illegal. Clients also are not able to visualize their projects by looking at the drawings and plans. Therefore, when the structure is almost built they will find that this is not something that they wanted. This will lead to demolition and rework which can cause serious delay in projects. Changes in drawings are also a problem that frequently happens. This may be due to a matter of taste or even changing the functionality of the building. One of the respondents has a desperate experience of changing functionality:

According to the interviews, consultants have the least role in causing delay for the projects among other stakeholders. They may cause delay by making mistakes in architectural drawings or designing the structure. In addition, consultants by insisting on unusual architectural design or using materials and fixtures that are not easily available in Iran may cause delay in projects. Contractors and sub-contractors may cause delay as well. In recent years, many contractor companies have been established. These new companies try to get jobs by offering very low prices to the clients. In the middle of the projects, because of unreal cost estimation, they will have many problems to execute the project. Hence, they have to look for finance and that takes a long time to do in the Iranian banking system. Most of contractors try to negotiate with the clients about the final cost of the project. Yet, the process of negotiation, and preparing new contract documents is time-consuming that will lead to delay. Moreover, there is no guarantee that the new estimation that they

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produce is accurate enough to finish the job in the proposed quality. This increases the rate of disputes between clients and contractors that may cause more delay for the project. Most of the Iranian contractor firms have tall hierarchical organizational structures with high levels of bureaucracy. Traditional managers, who do not have enough knowledge about modern management concepts, usually have problems with team making and distributing duties to the team members. Hence, they have to undertake much responsibility themselves. This will put the projects in serious trouble in managers absence. In the event of a problem if the manager is not available, nobody is able to make a decision because team members have neither enough information nor enough confidence to make decisions. Slow response to small problems will lead to crisis that may cause delay to the projects.

two or three months an applicant can get the money needed. Other sources of finance are limited and expensive. International financing is very rare due to political reasons and unfamiliarity of practitioners with the process of getting them. When a typical contractor is in short of cash, he or she cannot buy resources like material at the right time. Hence, especially for items with a long lead-time, materials or equipment cannot be delivered to the site on-time and this will cause delay. Moreover, cash constraints usually lead to payments disorder. In this case, subcontractors lose their motivation and sometimes leave the project.

6.2

Shortage of construction materials is also an issue that usually causes delay. Cement and concrete are two materials which are highly critical in projects. All respondents emphasize that they are often in shortage of concrete:

Economy

High inflation rates and non-availability of resources may cause delays for construction projects. If the price of resources increases sharply and continuously, clients and contractor cannot afford to buy them at the time that they are needed. This may lead to project abandonment as the result of cash constraints. To overcome inflation, the President Deputy for Strategic Planning and Control produces price indices each year that are a series of tables which shows average price for materials, equipment, labor and services. Practitioners calculate project cost escalation using these tables. However, for some items like cement and bitumen, there is no symmetry between prices in the indices and the real inflation rate. It means that price indices cannot cover real costs of resources in the case of high inflation rate. This increases the amounts of dispute in the supply chain that will lead to delay. One of the respondents has a problem with a cement supplier: ‘‘I signed a contract to buy a large amount of cement for my project three months ago and paid for it completely. The supplier provided me with a quarter of the contract and after that, he argued because the price of cement had increased, I have to pay the difference between the current price and price of three months ago. It is not fair because I paid for that contract three months ago. So why should I pay more? Inflation forces suppliers to be not committed to their contracts.’’ When suppliers do not keep their commitments, it can be expected that the project will be delayed. Availability of resources (money, material, machine, and manpower) have a direct impact on construction projects and may make them delayed. Construction practitioners in Iran often have trouble in financing their projects. Getting loans from Iranian banks is time-consuming and there is no guarantee that after

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‘‘In one of our project the payment to subcontractor was delayed for four months. They reduced the numbers of workers on site into half. This affected our schedule badly and the project was delayed.’’

‘‘It frequently happens that our project is suspended for 20 days or more because there is no concrete in the market.’’ Projects also may be delayed because of a shortage of other types of materials such as bricks, plaster or asphalt in spring and summer. In addition, in these seasons some steel sections become rare. Hence, contractors have to use imported steel sections that may have poor quality and do not comply with national standards. Low quality materials increase disputes and will contribute to delay. In terms of machines and equipment, most respondents explained that basic construction machines are usually available and there is no delay associated with them. However, just small delays may occur due to repair and maintenance of these machines. Shortage in quantity and skill of labor has a direct impact on construction projects. In the recent years, construction projects have been in shortage of labor because the government decided to deport Afghan laborers from Iran. Afghan laborers have worked in the Iranian construction industry for many years. After the establishment of Karzai’s Government in Afghanistan, the Iranian Government has returned Afghan laborers to their country. Although a few Afghan laborers could get working permits, most of them have to work illegally in projects. Occasionally, immigration inspectors visit construction sites to catch and deport illegal Afghan laborers. This causes a shortage in the labor pool in the construction industry. When a dramatic fall in the numbers of Afghan laborers happened, Iranian laborers asked for higher wages:

‘‘When Afghan workers were deported, Iranian laborers announced that they will not work at current wages anymore. In some cases we have to double their wages and they still do no work effectively.’’ In addition to shortages, workers’ productivity is low too. Some respondents believed that Afghan laborers are more skilled and work faster and better than Iranian laborers. The majority of construction laborers is not trained properly and makes many mistakes on site that will lead to delay: ‘‘Last night, about 15 cubic meters of concrete were wasted during pouring concrete by mistake of one of the laborers. Regardless to its cost, we have to order concrete and bring a concrete truck and pump to the site again and that is timeconsuming.’’

7

CONCLUSIONS

This study was designed to investigate causes of delays in construction projects in Iran that associate with stakeholders and economy. It showed that delay is a significant problem in construction projects in Iran. Time overrun, cost overrun, and increased disputes are consequences of delay in projects. This study, by using qualitative methodology described and interpreted the main causes of delay associated with stakeholders and economy. Major causes of delay that discussed in this paper are: clients’ lack of knowledge about construction legislation and process, changes in drawings, mistakes in structural design, slow decision making by clients, unreal cost estimation by contractors, disputes in the supply chain, high rate of inflation, shortage of materials and labor, delay in payments, and cash constraints. This paper suggests that special attention to factors identified in this study will help industry practitioners in minimizing the risk of delay in projects. However, it should be considered that only distinguishing causes are not enough and effective delay analysis and management should be carried out to mitigate delay impact on project. The findings of this study enhance understanding of factors that cause delay in projects and pertinent to stakeholders and economy. However, this paper is only a base for future investigations about delays in the Iranian construction industry. It is recommended that further research be undertaken to reveal the best practice in dealing with delays.

Alaghbari, W., Razali, M., Kadir, S., Ernawat, G. (2007). ‘‘The significant factors causing delay of building construction projects in Malaysia’’ Eng Construction and Architectural Management 14(2), 192–206. Al-Momani, A.H. (2000). ‘‘Construction delay: a quantitative analysis’’ International Journal of Project Management 18 (2000), 51–59. Assaf, S.A., Al-Hejji, S. (2006). ‘‘Causes of delay in large construction projects’’ International Journal of Project Management 24(4), 349–357. Australian Government-Austrade, 2007 [online] Available at: ≤http://www.austrade.gov.au/Construction-to-Iran/ default.aspx> [Accessed February 2008]. Companies and Markets, 2008 [online] Available at: [Accessed November 2008]. El-Razek, M.E., Bassioni, H.A., Mobarak, A.M. (2008). ‘‘Causes of Delay in Building Construction Projects in Egypt’’ Journal of Construction Engineering and Management (ASCE), 831–841. Faridi, A.S., El-Sayegh, S.M. (2006). ‘‘Significant factors causing delay in the UAE construction industry’’ Construction Management and Economics 24, 1167–1176. Fars News Agency, 2008 [online] Available at: ≤http://www. farsnews.com/newstext.php?nn=8703110422> [Accessed May 2008] (in Farsi). Iranian Students News Agency, 2008 [online] Available at: [Accessed October 2008] (in Farsi). Kaming, P., Olomolaiye, P., Holt, G., Harris, F. (1997). ‘‘Factors influencing construction time and cost overruns on high-rise projects in Indonesia’’ Construction Management and Economy 15(1), 83–94. Keane, P.J., Caletka, A.F. (2008). Delay Analysis in Construction Contracts, Wiley-Blackwell, UK. Mezher, T.M., Tawil, W. (1998). ‘‘Causes of delays in the construction industry in Lebanon’’ Engineering, Construction and Architectural Management 5(3), 252–60. Odeh, A.M., Battaineh, H.T. (2002). ‘‘Causes of construction delay: traditional contracts’’ International Journal of Project Management 20(1), 67–73. Sambasivan, M., Soon, Y.W. (2007). ‘‘Causes and effects of delays in Malaysian construction industry’’ International Journal of Project Management 25(5), 517–526. Saunders, M.N.K., Lewis, Ph., Thornhill, A. (2006). Research Methods for Business Students, 4th ed., Prentice Hall. Sweis, G., Sweis, R., Abu Hammad, A., Shboul, A. (2008). ‘‘Delays in construction projects: The case of Jordan’’ International Journal of Project Management 26, 665–674. Statistical Centre of Iran (SCI), 2006 [online] Available at: [Accessed January 2008] (In Farsi). Zack, J.G. (2003). Schedule delay analysis; is there agreement? Proceedings of PMI-CPM College of Performance Spring Conference, Project Management Institute, New Orleans, USA.

REFERENCES Aftab News, 2006 [online] Available at: [Accessed February 2008] (In Farsi).

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Designing the relationship between contractor and client to partnership K. Spang University of Kassel, Kassel, Germany

ABSTRACT: In many projects all over the world the relationship between client and contractor is shaped by many disputes and not by a jointed effort for high grade project results. This paper develops a general model of a cooperative relationship between client and contractor based on results of a recent research work. This general model shall be the base for jointly shaping singular partnering models for special project conditions by client and contractors. 1

INTRODUCTION

More and more clients and contractors complain about the many perturbances and disputes in construction and plant projects. The relationship between client and contractor often is neither cooperation nor a real partnership; it is rather a fight for individual advantages and suboptimisations. Is it possible to get optimized high grade constructions, roads or plants in this way? The typical situation on the international construction and plant market has been characterized by Egan (1998), Girmscheid (2005), Ingram & Bennet (1997), Odeh & Battaineh (2002), FIEC (2006) and Spang (2007) as follows: – dissatisfied clients, – distrust between client and contractor/s, – decrease of know-how due to ‘‘lowest priceprinciple’’, – increasing expenses for claim and anti claim management, – a growing number of disputes and litigations between clients and their contractors, – delayed payments, – low rate of return and high risk of business failure. So, a central question concerning the future of construction projects is how this situation can be changed. A current research work about partnering in large infrastructure projects (Spang & Faber 2008; Spang & Faber 2007) shows significant conclusions. In the first part of this field study project engineers and managers have been asked about the relationship between client and contractors. The answers to 3 essential questions seem to be typical for the actual situation: Figure 1a: Is the relationship between client and contractor cooperative? Figure 1b: Are you comfortable with the present situation? Figure 1c: Which party is the winner of the actual situation? The answers seem to be very typical for the actual situation. The first important result of the study is that

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Figure 1a. Relationship client-contractor. The interviewees could answer: entirely yes (++), predominantly yes (+), indifferently (0), marginally (−), not at all (−−). co = contractor, cl = client (Spang & Faber 2008).

Figure 1b. Acceptance of the actual situation. (Spang & Faber 2008).

clients as well as contractors rarely feel ‘‘partnership’’ in their projects (Fig. 1a). Another important result is that contractors and clients do not feel comfortable with this situation, which is a very good precondition for a change. And the third result is that they mostly do not see one party as a winner (Fig. 1c).

Figure 1c.

Winners of the situation. (Spang & Faber 2008). Figure 2. Results of a field study about the actual situation in German public construction projects (Spang & Faber 2008).

This is probably the most important precondition for a basic change: both parties seem to be losers in the projects. As losers may not be able to deliver really good results for the project and for their own, we suppose project results often far from an optimum.

2

yet a general standard. In Germany actually the introduction of ADS is in discussion. A field study, made by Gralla & Sundermann, 2008, shows a very high agreement for a general introduction of this method in Germany. Nevertheless, there is no need to wait for legal arrangements; the contract parties may include ADS in the contract—if they want to do it! ADS give a structured process for the solution of disputes and reduce troubles and their secondary effects as well.

THE PROJECT SITUATION TODAY

The first results of the study affirmed the reports from other researchers as well as the ‘‘feeling’’ from the acting people. The need for a change of the clientcontractor-relationship seems to be obvious. The next step now has to be the exploration of the causes of the situation in particular. We have to find answers to the predominant question ‘‘Why do we not have partnership between client and contractor?’’ So the above mentioned study was continued with typical themes in construction projects as ‘‘technical specifications’’, ‘‘specification changes’’, ‘‘risks’’, ‘‘responsibility’’, ‘‘disputes’’ and ‘‘motivation’’. Figure 2 shows the results of these questions. The study results in a great number of very revealing results and explanations for the project reality:

2.3

2.4 2.1 Incentives contribute to partnership Incentives may be given to a contractor for the increase of quality or technical specifications or for the decrease of costs by alternative solutions, proposed by himself. This leads to a project optimization and results in a win-win-situation for the client as well as for the contractor. 2.2 Alternative dispute solutions (ADS) contribute to partnership The results of our study are ambiguous in this case, which may be caused by the fact that ADS actually are very seldom in Germany. In spite of a strong tendency to alternative dispute solutions in the UK and in international projects, this new approach to solve troubles and disputes between clients and contractors is not

Responsibility is not clear

Construction and plant projects are dynamic ‘‘companies’’—so responsibility combined with the speed of decision making are important success factors. There are more problems on the client side (from the contractor’s point of view) than on the contractor side (from the client’s point of view). But often vague responsibility causes troubles concerning all parties. Poor data quality contributes to disputes

Every party has its own project data. So the parties have an unequal state of data quantity and data quality. This often causes a lot of disputes. 2.5

Risks often are not fairly distributed

This is especially a contractor’s view, but even the clients admit it predominantly. Generally both sides have information about risks, which they do not communicate to each other. Furthermore, each party intends to transmit the maximum of risks to the other side. And finally risks and troubles in a project mean money and time, for one party or often for both. 2.6

Claims contribute to the increase of disputes

Claims often occur as a consequence of the change of requirements by the client or because of the poor

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quality of the technical specifications. In both cases it implies an increase of costs and an extension of time for the client, but also a risk of income for the contractor. The lack of defined processes and requirements for project changes tightens the situation and often is the opening for the disputes. 2.7

Technical specifications are often not clear

This means, that the client probably demands other technical specifications that the bidder or future contractor has calculated and has planned to build and to supply. So the contract as the basic instrument of the client-contractor-relationship is often poor and probably the cause of many disputes. 2.8

problem: everybody wants more cooperation and partnership, but the ‘‘system’’ demands contrary parties! Is there any chance to achieve a partnership between client and contractor? A simple agreement or promise to work cooperatively will probably not be successful. So it could be a combination of soft and hard facts. As soft facts we mean the goodwill, a real intention for partnership and the acceptance of a win-win-situation, which will allow advantages and positive outcomes for both sides. As hard facts we mean clear regulations and processes for the most critical project elements. Rules and processes must be defined for a collaborative relationship between contractors and client in the contract.

3

Conclusions

The before mentioned statements give useful explanations for the actual situation and its background. They show main reasons for problems and disputes between contractors and client. Hence, approaches for a real change of the management of construction projects and particularly for the relationship between client and contractors are possible. They are based on a clear request of all concerned parties and hence there are actually very good preconditions for a successful change. But even if the need of change may be obvious there is a basic theory against its realisation, the ‘‘principal agent theory’’ (Müller & Turner 2005). This theory deals with the relationship between principal (the client) and agent (contractor). Its basic theses are: – In the tendering phase the principal often may not really evaluate the agent appropriately, his objectives and his intentions. Hence, he may choose a wrong or ‘‘not appropriate’’ agent. – The principal has more information about the project out of the design phase, so he has an advantage of information in the tendering phase. – During the project realisation the agent has more information about the project than the principal. He cannot be certain, if the agent’s activities and decisions may be the right ones for him too and must take big efforts for the survey of the agent. – Most of the agent’s activities influence the results of the principal—he is depending on the agent. – The principal decides on the payment and often pretends the conditions of contract—so the agent is depending on the principal. – Their interests in the ‘‘common’’ project are not the same. The client wants maximum work for minimum costs—the contractor wants maximum remuneration with a minimum work. As a conclusion of the actual ‘‘real’’ and of the ‘‘theoretical’’ situation we seem to have an insoluble

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BASIC PRINCIPLE OF SOLUTIONS

The project processes as well as the client-con-tractorrelationship have evolved over decades. They are shaped by (bad) experiences, prejudices and the meaning that ‘‘it always was like that’’. They are also influenced by national, social and economic situations and by increasing cost necessities. Earlier attempts to change the situation didn’t change a lot. Nevertheless, some contractors seem to have realized the necessity for change. Berger (2003) reports about a European field study where managers of construction companies were asked about the future development of the market. They acknowledged the increasing importance of cooperation with clients, effective risk management, establishing new contracting models, cooperating with suppliers and establishing new operating models. Latham (1994) reports about changes in the British construction market, based on a cooperative relationship. AGC, the largest American construction trade association, notes in 2006 ‘‘making collaboration the top priority in the execution of a project’’ (AGC 2006). In several parts of the world different types of cooperation have been developed, so called ‘‘Partnering’’, ‘‘Project Partnership’’, ‘‘Alliancing’’ or ‘‘Project Execution in Partnership’’. They all have a lot of elements in common—Figure 3 shows the most typical elements of collaborative project models. The different collaborative project models vary especially in the degree of partnership, choice of elements and type of contract. 3.1

Degree of participation of the contractor

An early involvement of the contractor in the project and in the design reduces the risk of insufficient technical specifications. But at the same time there is a strong influence of the contractor on the design and less chance of an independent client orientated design. Another aspect is that an early participation of

Partnering elements international client

win -win

contractor

trust choice of partner common objectives and scopes open books risk distribution feedback / good documentation clear responsibility alternative dispute resolutions incentive systems

Project success for all project participants

Figure 3.

Basic elements for partnering (Spang 2006).

the contractor complicates the bidding competition or even excludes it. 3.2

Choice of the ‘‘partners’’

Long term collaboration facilitates the development of trust but makes transparency difficult. Public tendering makes long term collaboration very difficult but facilitates transparency and competition.

– to accept that successful projects demand winning parties and not losing parties. Build up on these basic elements, which may be established in a preamble of a collaborative contract, the presented model can serve as a basis for specific project guidelines. It consists of 9 regulations: 4.1

4.2 3.3

Type of contracting and remuneration

There is a clear relationship between the type of contracting and remuneration (lump sum, target pricing, unit pricing or cost plus fee) and the quality of technical specifications, risks, contractor’s responsibility and competence. The more balanced the two sides are, the more cooperative may the relationship be. 3.4

Conclusions

So we conclude that contractors and clients should develop specific models of partnering for their specific purposes. Central questions to be answered in preparing such models are: a) How to create trust and fairness between contractors and customers? b) How to define a solid and optimal relationship between contractors and customers? And c) how to involve customers and contractors to ensure a solid and fair change and claim strategy? 4

GENERAL MODEL FOR PARTNERING IN CONSTRUCTION AND PLANT PROJECTS

There are indispensible preconditions for the success of each partnering model: the intention of the parties – to really want partnership, – to trust each other, – to accept the win of each other and

Clear project specifications

The project owner has to define his project specifications very clearly, especially quantities, qualities, performances and services to be included in the project. He may do this by a detailed design before tendering, which leads to detailed descriptions and specifications of the works (A). He may do this by a basic design and a well defined project performance and quality ‘‘functional’’ description (B). He may do this by developing the project and contract specifications together with the later contractor in a project preliminary phase, which precedes the final execution contract. This way excludes a usual tendering process and may be appropriate in case of long term relationship between contractor and client. Due to the lack of a competition, it needs open books (C). Trust

A precondition for building up trust is, that each party is really willing partnership and gives the other party a leap of faith at the beginning of the project. Another precondition is to accept the win of the other party, because trust cannot grow up together with enviousness and fear of the partner’s advantage. Further elements of building up trust in a project are: 4.2.1 Regular project meetings Client and contractor must have regular meetings to inform each other about the project news, project progress, need of changes and need of decisions. It is recommended to have different levels of meetings, for project managers, technical specialists etc. The meetings must be regularly from the beginning of the project, the frequency depending on the projects durance and dynamics. Meetings should be at least once a month, with obligatory topics and minutes. 4.2.2 Project reviews and feedback meetings In addition to the regular project meetings (4.2.1) regular project reviews are recommended. Their purpose is to collect and evaluate actual experiences in the project in order to define best practice and lessons learned to continuously improve the project processes. These reviews should be held up jointly by client and contractor. The frequency depends on the project durance and may be quarterly or twice a year and at the end of the project (‘‘Final review’’).

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4.2.3 General transparency Clear and transparent processes as well as all activities, which serve the exchange of information and knowledge between client and contractor contributes to the development and upkeep of trust. 4.2.4 Regulations for responsibility and decisions Clear decision authority and clear responsibility for people as project managers, and for decision makers on the client’s and on the contractor’s side is important. They are needed for fast decisions, for project progress and thus for reducing the dispute potential in a project. Trust will be created, if everybody knows the decision makers and the way of decisions. Further elements are clear and transparent organization (4.6), open communication (4.4), common data systems (4.7) and fair risk handling (4.3). 4.3

Fair risk handling

Construction and plant projects generally contain a wide range of risks, which touch three main questions: 1. Are they known or identifiable? 2. What is their probability of occurrence? 3. Who will bear them in the contract? Risk handling and risk distribution between client and contractor are basic themes in the bidding and in the execution phase. There are typical contractor’s risks as machine failure, logistics or quality of people’s work. But there are risks as ground conditions, price of raw material, legal procedures or political conditions, which are hardly influenceable by the contractor, these are typical client’s risks. Thus, fair risk handling means: 4.3.1 Fair risk distribution This means risk to the party, which can manage and influence it the best. Each party should describe clearly the risks, which the other side has to carry. 4.3.2 Risk compensation The contract may contain regulations to pay defined disturbances by predefined rates or risk compensation or risk supplements are paid as necessary. 4.3.3 Clear regulations Regulations for handling new risks in the current project are to be defined. Risks that may not be recognized in advance may be treated by a risk committee (representatives of the client and of the contractor). This committee decides on behalf of the ‘‘what’’, the ‘‘who’’ and the ‘‘how’’ of solutions as well as of the payment. To facilitate the payment special rates for material, machinery and work should be predefined. 4.4

Open communication (see 4.2, trust)

All information concerning the project, the participants and the stakeholders, who may influence the

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project, shall be transported in an open and fast way to the concerned persons in the project. Regulations about the information system should be a part of the contract, of the project handbook or of the kick-offdocuments. They should be defined jointly by client and contractors. Elements of regular information may be project meetings, meetings of the steering committee or a common website. 4.5

Clear processes for project changes

Large projects cannot be planned so exactly, that no changes will be necessary during the contract time. Changes of requirements by the client, changes of project conditions or lacks at the technical specifications are in many projects the source of contract changes. Contractors often are pleased about this ‘‘claims’’, for having a basis for additional payment requests. Clients fear them, because they mean additional costs and often lead to cost overflows. Predefined processes and requirements (when, what, who) for project changes and claims relax the change situation and give transparency to processes and to the cost and time situation of the project. They reduce the dispute potential to a minimum.

4.6

Clear and transparent organization

The project organization has to be clear concerning duties and responsibilities. The organizations of the client and of the contractor (s) have to be linked in a manner, that information is clear, open and fast and allows to find decisions in the demanded time (see 4.2.). Duties and responsibilities have to be transparent for all parties. For large projects it is recommended to have a steering committee (SC) with high representatives of the client’s and the main contractor’s side. The SC has to control the basic projects targets, take decisions of high importance on the project time, cost and targets and has to decide disputes between the project people as far as possible.

4.7

Common data systems

During the contract time many data are necessary for the contractor as well as for the client (contract data, project changes, costs, milestones, risks, accounts etc.). These data are used by both sides or have a contractual impact and so it is necessary to agree about these data. If every party has its own data (collection), they are often a matter of disputes—which data are the right ones? The best way to have agreed data is to use the same data. Hence contractor und client should agree early on a common data collection, treatment and documentation. The client should prescribe his requirements in the tender documents.

4.8

Contractual incentive regulations

Generally contractors try to find the maximum advantages, the contract will allow them, but they do not aim for a project optimum in the client’s sense (see Principal-Agent-Theory). So the client must motivate the contractor for further project optimisations. The contract therefore should contain predefined incentive regulations for a continuous optimisation and improvement of the project in technical, quality and cost terms. These may be additional remuneration for very high quality (higher than standard), shortening of the completion time or reducing costs for equivalent. So the client gets an added value by having more quality for the contracted prize or a lower prize for the contracted quality. For this optimisation the contractor gets more money (in function to the more of quality) or the cost reduction will be split in a contractor’s and a client’s part (e.g. 50/50). This type of incentive after the closure of the contract has a double effect. The contractor takes an additional effort to optimize the project in order to get an additional profit and both parties will take an effort to minimize disputes, because otherwise optimisation solutions are not possible. 4.9

Contractual alternative dispute solutions

In addition to the above mentioned regulations, which will reduce the number of disputes in a project, there should be an alternative dispute regulation fixed in the contract. This regulation will contain several grades for a dispute before coming to the court. In each grade another level of representatives of the parties will be responsible for the negotiations. Each grade will have time limits too. The first grades are limited on the involved parties. If they will not jointly find a solution, the further process is managed by one or three arbiters. The decision of the arbiter should be compulsory. 5

FINAL REMARKS

Construction and plant projects are very different as well as their national and juridical preconditions. So a single partnering model cannot be appropriate, but specific models should be developed for singular conditions. But all types of partnering or alliancing contracts and of collaborative working may not be possible with changing only one thing. Based on the indispensable willing of partnering and of win-win various elements have to be respected. Risk handling, incentives, alternative dispute solutions, clear responsibility, deciding regulations, open communication—all these elements contribute to partnership and they help to build up trust. Collaborative working is a ‘‘package’’, which may not exist partly with selecting elements, appropriate only for one side. Hence, the best way to develop a partnering model is to develop it in

cooperation with client and contractors, based on this general model. In Germany guidelines for partnering in infrastructure projects, which have been developed jointly by researchers, representatives of client organizations, contractors and the ministry of transport in a research project are actually to be proved in a testing phase. REFERENCES Egan, J. (1998): Rethinking Construction. Dept. of Trade and Industry, London, 1998. Girmscheid, G. (2005): Partnerschaften und Kooperationen in der Bauwirtschaft. Bauingenieur, Band 80, Febr. 2005, pp. 103–113. Ingram, I. & Bennett, J. (1997): Book Reviews, Construction Management & Economics, Vol. 15, No. 3, May 1997, Routledge, UK. Odeh, A.M. & Battaineh, H.T. (2002): Causes of construction delay: traditional contracts. Int. Journal. Of Project Management, 20 (2002) 67–73. FIEC (2006): FIEC Answers to the consultation on ‘‘The future of the internal Market’’, 2006-6-15. Spang, K. (2007): Advanced project management in civil engineering. In: Proceedings 4th ISEC Conference, Melbourne, Australia, 2007, pp. 1467–1472. Spang, K. & Faber, S. (2008): Partnerschaftliche Projektabwicklung zwischen Auftraggeber und Auftragnehmer bei Infrastrukturprojekten. Progress Report (non published) for the Ministry of Transport and building, 2008. Spang, K. & Faber. S. (2007): Partnerschaft zwischen Auftraggeber und Auftragnehmer—die Zukunft des Bauens. in: Proceedings 3rd PM Symposium Kassel, 2007, ISBN 978-3-00-022130-9, pp. 3–28. Lihs, A. (2004): Stand, Entwicklungspotenziale und Tendenzen des Projektmanagements großer Infrastrukturprojekte im deutschsprachigen Raum. Master Thesis, Universität Kassel, Chair of Projektmanagement, April 2004. Gralla, M. & Sundermann, M. (2008): Adjudikation— außergerichtliches Streitlösungsverfahren für Baukonflikte auf gesetzlicher Basis? Bauingenieur Vol. 83, May 2008: 238–247. Müller, R. & Turner, J.R. (2005): ‘‘The Impact of PrincipalAgent Relationship and Contract Type on Communication between Project Owner and Manager’’, International Journal of Project Management, Vol. 23, No. 5, pp. 398–403. Berger (2003): Erfolgsfaktoren der Bauindustrie— Europaweite Studie. Roland Berger Strategy Consultants, Munich, 2003. Latham, M. (1994): Construction the team. Final report. Her Majesty’s Stationary Office (HMSO), London, 1994. AGC (2006): AGC Construction News, June 15, 2006. Spang, K. (2006): Innovative Projektabwicklung bei Bauprojekten—Plädoyer für einen Paradigmenwechsel. In Journal ‘‘Bauingenieur’’, Vol. 81, March 2006, pp. 117–125. Spang, K. Özcan S. (2007) (ed): Partnerschaftsmodelle bei Infrastrukturprojekten und Projekten des Großanlagenbaus. Proc. 3rd PM Symp. Kassel, 2007, ISBN 978-3-00-0221309.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Determining schedule delay causes under the Build-Operate-Transfer model in Taiwan J.B. Yang & C.C. Yang Institute of Construction Management, Chung Hua University, Hsinchu, Taiwan

ABSTRACT: Project delivery method of Build-Operate-Transfer (BOT) improves the commencement probability of public construction works through private investments, and brings the development of related industries. In Taiwan, the number of public construction works that apply the BOT model as their project delivery method is increased gradually. Although some projects select the promoter successfully, and then advance into the stages of build and operate; some projects still stay in the stages of tendering and negotiation. Some projects encounter critical problems for advancement. This study tried to find the delay causes in all stages of a BOT project. Study results reveal that the stage of negotiation and signing of concession agreement is the most essential stage, in which ‘improper contract planning’ and ‘uncertainty on political issues and government-finished items’ are the most significant delay causes on the perspective of importance and frequency, respectively.

1

INTRODUCTION

Project delivery method of Build-Operate-Transfer (BOT) improves the commencement probability of public construction works through private investments, and brings the development of related industries. BOT is a typical model of Public-Private-Partnership (PPP). In Taiwan, the number of huge public construction works that apply the BOT model as their project delivery method is increased gradually. Although some projects select the promoter successfully, and then advance into the stages of build and operate; some projects still stay in the stages of tendering and negotiation. Some projects encounter critical problems for advancement. This study tried to find the delay causes in all stages of a BOT project. This disclosed information would be beneficial to advance and execute the following BOT projects. This study used two questionnaire surveys to collect the response on delay causes from BOT project participants. After collecting the causes from the literature, this study used one questionnaire to confirm the feasibility of the causes used for BOT projects. Furthermore, this study used another questionnaire to collect the opinions on the importance and frequency of delay causes. Various statistical analysis methods, including descriptive statistics, reliability, validity and correction analysis, were used to find the characteristics of importance and frequency of all causes. Moreover, this study employed the path analysis method in Structural Equation Modeling (SEM) approach to clarify the corrections among all causes on the dimensions of importance and frequency. Essential causes in the dimensions of importance and

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frequency were ranked by the result of multiplying objective evaluation meaning value by factor loading value. Identified causes can be used to prevent delay occurrence on the following BOT projects in the future.

2 2.1

PROMOTION OF PRIVATE PARTICIPATION IN TAIWAN Act of promotion of private participation

The act for promotion of private participation in infrastructure projects, which was promulgated on Feb. 9, 2000, espouses the spirit of vigorous innovation and, from the aspect of creating benefit, establishes partnership relations between the government and the private sector (Public Construction Commission 2000). The main features of the act include the following: (1) embodying general-type legislation: the act applies alike to all types of industries, sectors, and development plans; it maintains flexibility of articles, and expands the delegation of administrative authority to the government officials implementing the projects; (2) embodying civil contract concept: the act adopts the principle of civil contracts under which rights and obligations between the government and private sector shall be stipulated in the concession agreement to reflect the partnership spirit of equal cooperation and create ‘‘win-win’’ investment conditions; (3) maximizing private participation: not only is the scope of infrastructure development in which private participation is permitted extremely broad and the

methods of participation diverse, but private entities may also carry out their own planning of proposals for participation in infrastructure projects; this allows the private sector to discover investment opportunities, and to give full rein to its creativity in planning investment projects; (4) maximizing government carefulness: In the interest of completeness, feasibility studies and preliminary planning should be conducted for all infrastructure projects which the government plans for private participation; from the viewpoint of private sector, the feasibility of private investment should be evaluated carefully and, in consideration of the special characteristics of the infrastructure project, commercial incentives should be incorporated in the formulation of the preliminary plan. 2.2

Models of private participation

There are many models allowed by private participation in Taiwan, including: – The private institution invests in the construction and operation of the infrastructure project, and upon expiration of the operation period, transfers the ownership to such project to the government. (BuildOperate-Transfer, or BOT model.) – The private institution invests in the construction of the infrastructure project and upon completion of the building, relinquishes the ownership to the government without compensation. The government then commissions the operation of the infrastructure project in question to the same private institution. Upon expiration of the operation period, the right to operate reverts back to the government. (Non-compensable Build-Transfer-Operate, or Non-compensable BTO model.) – The private institution invests in the construction of the infrastructure project and upon completion of the construction, the government acquires the ownership through the payment of the construction expenses, either by a lump sum payment or by installment payments. The government then commissions the operation of the infrastructure project in question to the same private institution. Upon expiration of the operation period, the right to operate reverts back to the government. (Compensable Build-Transfer-Operate, or Compensable BTO model.) – The government commissions the private institution, or the private institution leases from the government, existing facilities for operation after making renovations or expansions. Upon expiration of the operation period, the right to operate reverts back to the government. (Rehabilitate-OperateTransfer, or ROT model.) – The government invests in the construction of the infrastructure project and then commissions the operation thereof to the private institution. Upon

expiration of the operation period, the right to operate reverts back to the government. (Operatetransfer, or OT model.) – To support the national policy, the private institution invests in the construction of the infrastructure project and owns the ownership thereto upon completion of the construction, and then either operates the facility by itself or commissions a third party for operation. (Build-Own-Operate, or BOO model.) – Any other model as may be approved by the competent authority. 2.3

Government promotion mechanism

The Executive Yuan established the committee for the promotion of private participation in infrastructure projects at the end of 2002 in order to provide a crossministerial coordination mechanism for such projects. The committee is headed by the premier and is composed of the heads of central-government ministries and commissions. An coordination task force was also set up, with the competent authority (Public Construction Commission) handling staff operations, for the task of promoting private participation in infrastructure projects by central government ministries and local governments, and for intensified coordination in removing investment obstacles that require crossministerial cooperation. The ministries and commissions, as well as county and city governments, have

Feasibility study, preliminary plan, and preparatory work

Announcement by a public notice

Submission of application

Evaluation and selection

Negotiation and signing of concession agreement

Construction

Operation

Transfer

Figure 1. Implementation procedure of private participation projects in Taiwan.

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also set up their own private participation promotion committees to oversee coordination and resolution of problems encountered in carrying out private participation projects within their jurisdictions. 2.4

3.1

STUDY DELAY CAUSES

Stage 2

Announcement and submission of application

Stage 3

Evaluation and selection

Stage 4

Negotiation and signing of concession agreement

Stage 5

Design

Stage 6

Construction

Stage 7

Operation

Stage 8

Transfer

General delay causes

Several studies have discussed delay causes in different countries. These studies are generally grouped into two broad categories: independent and comparative studies. Most of the previous studies belong to the independent category: to identify and then to rank delay causes by using different indices in one area. For example, Odeh & Battaineh (2002) investigated 28 delay causes from large construction projects with traditional type contracts in Jordan and ranked the delay causes by a relative importance index that was calculated by a normalized weighted frequency value. The identified delay causes include (1) inadequate contractor experience, (2) financing and payments of completed works, (3) owner interference, (4) labor supply, and (5) subcontractors. The other type of study, the comparative study, usually conducts comparisons on delay causes in different countries or from different sources. For example, Majid & McCaffer (1998) thoroughly reviewed the literature concerning delay causes analysis, which included 42 sources and focused on the issue of nonexcusable delays. Majid & McCaffer compared the delay causes within different economic climates and in countries with different industrialized statuses and then gave an integrated ranking for the top 25 nonexcusable delays. The identified top five delay causes include (1) late delivery or slow mobilization, (2) damaged materials, (3) poor planning, (4) equipment breakdown and (5) improper equipment. 3.2

Feasibility study and preliminary plan

Implementation procedure

For a project delivered by any model of private participation, it follows general implementation procedures shown as Figure 1. Although different model might have different implementation procedure, some stages are all the same, i.e. (1) feasibility study, preliminary plan, and preparatory work, (2) announcement by a public notice, (3) submission of application, (4) evaluation and selection, (5) negotiation and signing of concession agreement. Detailed information about the promotion of private participation in Taiwan could be found on the web (Public Construction Commission 2009). 3

Stage 1

BOT implementation stages

For implementing a BOT project, eight stages are organized to success it (Public Construction Commission

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Figure 2.

Stages of BOT project implementation.

2004). Figure 2 shows the stages consisting of (1) feasibility study and preliminary plan; (2) announcement and submission of application; (3) evaluation and selection; (4) negotiation and signing of concession agreement; (5) design; (6) construction; (7) operation and (8) transfer. 3.3

Possible delay causes in BOT implementation

This study collected various delay causes from varied literature and organized as Table 1. Detailed reference could be found elsewhere (Yang 2008). These identified delay causes are the basis for further investigation.

4

STRUCTURAL EQUATION MODELING (SEM) METHOD

The Structural Equation Modeling (SEM) technique was originally developed by sociologists and psychologists. SEM is a multivariate technique used to estimate a series of interrelated dependent relationships simultaneously (Hair et al. 1998). SEM is regarded as an extension of standardized regression modeling to deal with poorly measured independent variables and is ideally suited for many of the research issues dealt with in construction engineering and management (Molenaar et al. 2000). Further, SEM has

Table 1.

Study delay causes in different stages.

Stage

Delay causes

(A) Feasibility study and preliminary plan

(1) Model selection for PPP; (2) Law and regulation change; (3) Shortage of professional service fee; (4) Limitation of land use regulation; (5) Urban plan change; (6) Selection of professional consultant; (7) Lack of determination of entitled government; (8) Implementation schedule change. (1) Announcement content change; (2) Low enthusiasm of private investment; (3) Improper announcement content; (4) Law and regulation change; (5) Urban plan change; (6) Low self-liquidating ratio; (7) Lack of incentive for private investment; (8) Trivial administrative procedures; (9) Rigid investment content; (10) No investment consultant; (11) Short tendering period; (12) Impractical financial feasibility. (1) Evaluation and selection committee change; (2) Hard to define objective evaluation rule; (3) Improper evaluation and selection procedure; (4) No bidder; (5) Divergent results by evaluation and selection committee; (6) No qualified bidder. (1) Design change; (2) Law and regulation change; (3) Schedule delay by administrative procedure; (4) Project debt collateral; (5) Mechanism for forced transfer; (6) Role conflict in negotiation; (7) Urban plan change; (8) Uncertainty on political issues and government-finished items; (9) Unclear definition of changeable and unchangeable items; (10) Land rental fee; (11) Debt problem; (12) Rigid land rental fee; (13) Divergence in negotiation and contract signing; (14) Improper timing for negotiation; (15) Dispute on operation duration; (16) Dispute on land usage; (17) Delayed land liberation schedule; (18) Improper contract planning. (1) Demand change on the client; (2) Wrong client’s demand recognized by project team; (3) Inexperienced project team; (4) Improper planning and schedule developed by project team; (5) Improper design caused by law and regulation change; (6) Trivial administrative procedures; (7) Improper selection of designer and consultant; (8) Uncompleted client-finished items. (1) Change orders; (2) Unexpected increased quantity; (3) Late site liberation by client; (4) Shortage of construction budget; (5) Bad weather and disaster; (6) Law and regulation change; (7) Fluctuation on resource price; (8) Shortage of materials; (9) Failed examination and inspection; (10) Failed final examination. (1) Construction schedule delay; (2) Law and regulation change; (3) Unclear operation plan; (4) Shortage of operation manpower; (5) Shortage of operation cash flow; (6) Shortage of debt capital; (7) Inadequate personnel skill; (8) Bad weather and disaster; (9) Operation disaster; (10) Resistance by residents. (1) No takeover entity; (2) Incompletion of property transfer; (3) Unclear definition of compensable and non-compensable project; (4) Indefinite property list (5) Incomplete refunded project loan; (6) Insufficient document preparation (7) Divergent transfer scope; (8) Incomplete duty from contract party.

(B) Announcement and submission of application (C) Evaluation and selection (D) Negotiation and signing of concession agreement

(E) Design

(F) Construction

(G) Operation

(H) Transfer

been a popular tool for data analysis in marketing and consumer behavior domains (Baumgartner & Homburg 1996). SEM is an advanced data analysis tool to explore data relationships quantitatively. Yang & Ou (2008) used SEM to analyze the relationships among key delay causes for construction projects delivered by traditional method. The study has proven that SEM is capable of quantifying comprehensive relationships among investigated factors.

5

RESULTS

Based on a questionnaire survey for obtaining replies from BOT project participants in Taiwan, this study employed SEM and traditional statistic method to analyze the results of schedule delay causes for BOT project. This study sent 320 questionnaires and had 195 responses.

5.1

SEM

In SEM applications, the goodness-of-fit (GOF) indices are essential tools for assessing the fitness of developed models. The popular and obtained indices from the software, LISREL version 8.72, employed in this study include: χˆ2 (chi-square), RMSR (rootmean-square residual), GFI (goodness-of-fit index), AGFI (adjusted goodness-of-fit index), NFI (Normal fit index), NNFI (Tucker-Lewis index) and CFI (Comparative fit index). The values of above indexes in this study are greater than the suggested acceptable levels. It provides evidence that the fit between the measurement model and the data is acceptable. Figures 3 and 4 show the analysis results by SEM on the dimensions of importance and frequency. From importance perspective, the stage of negotiation and signing of concession agreement (coefficient: 0.55) is the most essential stage. In the stage, the class, ‘land rental fee,’ ‘unclear definition of changeable and unchangeable items’ and ‘debt problem’ are the

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Feasibility study and preliminary plan

0.11

Announcement and submission of application

0.15

Evaluation and selection

0.15

Negotiation and signing of concession agreement

0.13

Design

0 3 0.3 0.1

0.13

0.3

1 0.55 BOT project schedule delay

0.24

Figure 3.

0.14

Construction

0.11

Operation

0.12

Transfer

0.1 0.1 5 3

0 0.2

Structural equation model based on importance perspective.

Feasibility study and preliminary plan

0.12

Announcement and submission of application

0.14

Evaluation and selection

0.16

Negotiation and signing of concession agreement

0.14

9 5 0.2 0.1

0.11

0.3

3

0.52 BOT project schedule delay

0.28

Design

.20

Figure 4.

Construction

0.11

Operation

0.11

Transfer

0.1 0.1 7 2

0 0.15

Structural equation model based on frequency perspective.

Table 2.

Top significant delay cause in different stages on importance perspective.

Stage

Delay causes

(A) Feasibility study and preliminary plan (B) Announcement and submission of application (C) Evaluation and selection (D) Negotiation and signing of concession agreement (E) Design (F) Construction (G) Operation (H) Transfer

(7) Lack of determination of entitled government (7) Lack of incentive for private investment (6) No qualified bidder (18) Improper contract planning (8) Uncompleted client-finished items (7) Fluctuation on resource price (1) Construction schedule delay (3) Unclear definition of compensable and non-compensable project

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Table 3.

Top significant delay cause in different stages on frequency perspective.

Stage

Delay causes

(A) Feasibility study and preliminary plan (B) Announcement and submission of application (C) Evaluation and selection (D) Negotiation and signing of concession agreement

(7) Lack of determination of entitled government (7) Lack of incentive for private investment (2) Hard to define objective evaluation rule (8) Uncertainty on political issues and governmentfinished items (8) Uncompleted client-finished items (7) Fluctuation on resource price (1) Construction schedule delay (2) Incompletion of property transfer

(E) Design (F) Construction (G) Operation (H) Transfer

most significant top-three delay causes. From importance perspective, the stage of negotiation and signing of concession agreement (coefficient: 0.52) is also the most essential stage. In the stage, the class, ‘uncertainty on political issues and government-finished items,’ ‘divergence in negotiation and contract signing’ and ‘improper contract planning’ are the most significant top-three delay causes. 5.2

Traditional statistical method

Excluding analysis by SEM, this study used traditional statistical method to rank all delay causes, i.e. by the result of multiplying objective evaluation meaning value by factor loading value. Tables 2 and 3 list the top significant delay causes in different stages on the perspective of importance and frequency, respectively. 6

CONCLUSIONS

Based on the analysis of SEM and traditional statistical methods, this study concluded that the stage of negotiation and signing of concession agreement is the most essential stage, in which ‘improper contract planning’ and ‘uncertainty on political issues and government-finished items’ are the most significant delay causes on the perspective of importance and frequency, respectively. Project delivery method of BOT improves the commencement probability of public construction works through private investments, and brings the development of related industries. Project delivered by BOT method is a trend for public infrastructure projects worldwide. Although the delay causes are identified from the respondent’s perspective in Taiwan, the results are valuable for other areas. Identified causes can be used to prevent their occurrence on the following BOT projects in the future. ACKNOWLEDGEMENTS

this research under Contract No. NSC96-2221-E-216027-MY2 and the Chung Hua University for providing partial funding (Funding No. CHU96-2211-E-216027-MY2) to this research. REFERENCES Baumgartner, H. & Homburg, C. 1996. Applications of structural equation modeling in marketing and consumer research: a review. International Journal of Research in Marketing, 13(2): 139–161. Hair, J.F., Anderson, R.E., Tatham, R.L. & Black, W.C. Multivariate data analysis. 5th Edition. Upper Saddle River, NJ: Prentice-Hall. 1998. Majid, M.Z.A. & McCaffer, R. 1998. Factors of non excusable delays that influence contractors’ performance. Journal of Management in Engineering 14(3): 42–49. Molenaar, K., Washington, S. & Diekmann, J. 2000. Structural equation model of construction contract dispute potential. Journal of Construction Engineering and Management 126(4): 268–277. Odeh, A.M. & Battaineh, H.T. 2002. Causes of construction delay: traditional contracts. International Journal of Project Management 20(1): 67–73. Public Construction Commission. The act for promotion of private participation in infrastructure projects. The Executive Yuan, Taiwan. 2000. Public Construction Commission. Reference manuals of schedule and quality management for implementing BOT projects. The Public Construction Commission, Executive Yuan, Taiwan. 2004. Public Construction Commission. Promotion of private participation. http://ppp.pcc.gov.tw/pcc_site/english/index. cfm, The Public Construction Commission, Executive Yuan, Taiwan. 2009. Yang, C.C. Determining schedule delay causes for private participating public construction works under the BuildOperate-Transfer model. Master thesis of Institute of Construction Management, Chung Hua University, Hsinchu, Taiwan. 2008. Yang, J.B. & Ou, S.F. 2008. Using structural equation modeling to analyze relationships among key causes of delay in construction. Canadian Journal of Civil Engineering 35(4): 321–332.

The authors would like to thank the National Science Council, Taiwan, ROC, for financially supporting

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Dey Street Tunnel: The challenges of a design build project in a congested urban setting M. Trabold AECOM, New York, USA

ABSTRACT: The Dey Street Tunnel is a Design Build project which connects two transportation hubs—the new PATH station at the site of the World Trade Center and the NYCT Fulton Street Transit Center—with a tunnel beneath Dey Street in lower Manhattan. The paper will outline the challenges of the project and show how an effective design build team—Skanska USA; AECOM (as DMJM Harris) and Weidlinger Associates—working together with the owner can develop innovative solutions that save time and money. The challenges of the Dey Street Tunnel project consisted of underpinning and excavation below two active subway stations; installation of secant pile walls adjacent to historic buildings; construction of a 300 foot (91 m) long cut and cover tunnel beneath an active city street; and related challenges working with existing structures that are over 100 years old.

1 1.1

SITE AND PROJECT OVERVIEW Site location

Dey Street is a small street running east-west located in downtown Manhattan between Church Street and Broadway. Directly to the west is the site of the former World Trade Center (WTC). The northern boundary is the Millennium Hilton Hotel, a 58-story steel structure and the office building 195 Broadway, which is 19 stories and is on New York State’s Preservation and Historic registry. To the south are a department store (21st Century) and other medium rise buildings. See Figure 1. The Dey Street Tunnel directly connects two existing Metropolitan Transit Authority—New York City Transit (MTA NYCT) subway stations—the Cortlandt Station which runs under Church Street and the Fulton Street Station which runs under Broadway. The Fulton Street Station is part of a complex of stations served by different subway lines connected by existing tunnels and mezzanines. The connections between these stations will be improved as part of the overall NYCT Fulton Street Station Complex Rehabilitation, of which the Dey Street Tunnel is a part. The overall construction cost for the Fulton Street Station Complex Rehabilitation is approximately $700 million with the Dey Street Tunnel costing $140 million. As a way to expedite the construction schedule, the Dey Street Tunnel Project was bid by MTA—NYCT as a design build contract with the Skanska/ AECOM team selected as the Contractor/Designer. The successful bid by Skanska came in almost $60 million less than the next closet bid, which allowed MTA-NYCT to realize considerable cost savings by

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Figure 1.

Site location plan.

using the Design Build delivery method. Issuing the Dey Street project in June of 2005 allowed for a considerable time saving in the overall construction schedule of the Fulton Street Transit Complex. With the next construction contract for the Fulton Street Complex being awarded in December of 2008 upon completion of final design in early 2008, the MTA had greater flexibility in scheduling the future work since the two subway stations were already underpinned under the Dey Street project. 1.2

Site geology

Generally the subsurface conditions underlying the site consisted of fill underlain by a variable soil stratigraphy consisting of fine to medium sand, silty fine sand,

stratified silt and silty fine sand, glacial till and bedrock. The excavations for the project were approximately 40 to 45 feet (12 to 14 m) below the ground surface which places them in the stratified silt and silty fine sand stratum. The local groundwater elevation was approximately 104 feet, which put the Cortlandt Street Station concourse below the water table, and the Fulton Street Station concourse above the water table.

2 2.1

UNDERPINNING OF THE SUBWAY LINES Existing NYCT subway structure

Both subway stations were built in the early 1900s. The existing transit subway structure is steel bent construction with bents spaced every 5 feet (1.5 m). In between the bents for the walls and roof are concrete jack arches. The invert consists of steel invert beams and a flat concrete slab. All concrete is unreinforced and the invert slabs rest on soil at both stations. 2.2

General orders (G.O.s)

A key aspect of maintaining the project schedule was performing the underpinning work for the existing subway in the allocated weekend outages. NYCT’s subways operate 24 hours a day, year round so in order to perform any track work, a General Order (G.O.) must be obtained from NYCT’s operations department. These general orders typically shut down a section of a particular subway line from 11 pm on Friday night until 5 am Monday morning, allowing the contractor a 55-hour window in which to perform the work. For the Dey Street project, NYCT allowed the contractor 11 consecutive weekends to perform the underpinning work for each station. The underpinning work for each station was done independently and at different times of the year. 2.3

of mini piles required to support the underpinning loads. See Figure 2 for the installation of mini piles using a low headroom rig. The original conceptual design also called for some of the piles to be socketed into rock and used as permanent support for the concourse base slab because of a concern about settlement issues in the concourse base slab and soil liquefaction during a seismic event. To alleviate these concerns, during the proposal phase the project team looked closely at the boring logs, and using local knowledge of the site, determined that settlement of the concourse mat slabs could be controlled with a ‘floating’ slab and therefore not require pile supports. Upon award, the project team undertook a soil exploration program that included standard penetration and cone penetration tests (SPT and CPT respectively) to verify the assumptions made during the proposal phase. The use of a floating slab enabled the design team to eliminate the requirement for mini piles to be used as permanent elements. Therefore mini piles were only needed for the temporary underpinning. These two changes resulted in eliminating more than half the mini piles in the conceptual design, and resulted in both a cost and time savings to the project. The original contract had approximately 480 mini piles, while the revised design had just over 200. The time savings of having to install fewer mini piles during the G.O.s allowed the contractor to do additional work during their G.O. work schedule such as the installation of the underpinning framing, which was also modified from the conceptual design. 2.4

Track invert

Another challenge was rebuilding the track invert in the allocated space as per the contract documents. Due to restrictions on the concourse’s top of slab elevation (to align with adjacent future contracts), internal clearance heights, and maintaining the existing track

Mini piles

The project team felt a key aspect to the success of the project was to perform the underpinning work the quickest way possible and to install as much permanent work in the underpinning system. During the proposal stage, the project team met numerous times to come up with a viable solution to the underpinning. Using NYCT’s conceptual design as a baseline, the project team looked at ways to improve upon it. The first item considered was the mini pile spacing. As the majority of mini piles would have to be installed from the track bed during a G.O., reducing the total number was critical. By changing the underpinning framing the project team was able to reduce the number

Figure 2. Photo of low headroom mini pile rig on cortlandt street station platform.

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alignment, limited space was available to rebuild the track invert and permanent underpinning steel. The original solution proposed in the conceptual design was to temporarily support the track on mini piles and upon completing installation of the permanent underpinning steel, rebuild the track invert slab. As part of the brainstorming meetings during the proposal phase, the project team was able to develop a concept that would use the temporary track slab as the permanent invert slab, thus allowing the new track slab to be placed before the permanent underpinning steel was installed. The underpinning system was installed as follows. The mini piles were installed 3 across the existing tracks—one at the structure centerline, and the other two just beyond the platform column lines. See Figure 3. The existing center line columns are spaced 5 feet (1.5 m) on centers, while the platform columns are spaced at 15 feet (4.5 m). The rows of mini piles were installed at 10 feet (3 m) on centers. The existing track slab was then demolished and the existing invert slab was locally demolished at the line of the mini piles to allow for the installation of heavy W14 beams transversely on top of the mini piles to support the track. Three W10 stringers spanning between the W14s were then installed under each rail to transfer the train load from the rail (through 8 inch (20 cm) wood track ties) to the W10s to the W14s to the mini piles. After installing the W14s, W10s and temporary wood ties, the remainder of the invert slab was demolished. At this time the track was supported by the mini piles, and the existing grade was still at the bottom of the now demolished invert slab. Reinforcing steel was placed between the W14s as part of the permanent invert slab and plywood was placed against the earth to act as a one sided form. Concrete was then placed to form the permanent invert slab, and on the following weekend concrete was placed to form the track slab. The W14s and W10s were left in place and encased in the new

Figure 3.

Typical underpinning at track invert level.

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concrete slabs. Part of the W10s had to be burned off when installing the permanent track ties. Being able to place the permanent invert and track slabs, before starting excavation to install the permanent underpinning steel had a major impact on the construction schedule.

3 3.1

WORKING IN AN URBAN ENVIRONMENT Secant pile wall installation

As a way to allow the excavation to proceed without having a draw down of the water table, a water tight support of excavation wall was required. The existing subway structure and adjacent buildings were built close to the existing property lines. Typically a utility corridor was left between the subway and the property lines to provide space for the utilities that could not be located between the top of the subway box and the roadway. The typical cover above the subway structure is 6 to 8 feet (2 to 2.5 m), so utilities unable to be placed there are located within this corridor. The utility corridor by the Cortland Street Station was filled with a 3 foot (1 m) wide, egg shaped sewer and an 8 foot by 3 foot (2.5 m by 1 m) electrical duct bank. At the narrowest points the distance between the existing subway structure and the basement structure of the millenium Hotel and Century 21 Department Store was 3 feet (1 m). Along the western side of the Cortlandt Street station adjacent to the World Trade Center Site was an existing concrete retaining wall built before the subway structure that extended down below the proposed depth of the concourse slab and into rock, so only three sides of the Cortlandt St. station had to be enclosed with secant piles to form the watertight bathtub. Transversely across the street 3 foot (1 m) diameter secant piles were used, and in areas of the subway structure, jet grout columns were installed from the track bed using low head room jet grout rigs. In the areas between the buildings and the subway structure due to limited space 2 foot 6 inch (75 cm) diameter secant piles were used instead of the larger secants. In order to install the secant piles, the sewer had to be temporarily flumed and re-routed through the subway structure and the telecommunication duct bank was demolished and the cables temporarily hung from the hotel’s existing basement wall. See Figure 7. Installing the secants adjacent to the Century 21 building proved most problematic. See Figure 4. Unlike at the Millennium Hotel whose above ground structure was set back from the building line, the Century 21 building was built up to, and sometimes over the building line. When allocating the space to install the secants, the project team had to consider not just

always be available for deliveries for Century 21, so work had to be staged to allow for deliveries around the clock. During installation of the secant piles in front of the store, extensive coordination between the contractor and the building owner was required to insure uninterrupted deliveries and to avoid negative impacts to the construction schedule.

3.2

Figure 4.

Secant pile rig in front of century 21 dept store.

the diameter of the secant but the space needed for the equipment and how close it could get to the existing structure. Once the secant piles were installed, they were used for support of the roadway decking as well as support of excavation and water cut off. The secant piles were installed 5 feet (1.5 m) into the till layer, and not socketed into rock. One issue that arose while installing the support of excavation walls at both the north and south end of the Cortland Street station was the interface between the secant pile wall and the jet grout wall. As mentioned earlier, the jet grout columns were installed from within the subway structure and the secant piles were installed from the street level outside the areas of the subway structure. Due to scheduling constraints with the track outages and street closing permits, the jet grout columns had to be installed prior to the installation of the secant piles. Upon completion of the bathtub and once excavation went below the water table, a breech was discovered between the jet grout columns and the secant piles. During installation, the secant piles chewed up parts of the jet grout columns causing a breech between the walls. Areas of the jet grout columns were chipped away, which resulted in water infiltration in the excavation. Since track outages were not available, all grouting to control the water infiltration had to occur from the face of the excavation. This condition did not occur at the south end, where the jet grout columns were installed after the secant piles. Working along Dey Street proved to be quite challenging. Dey Street is a narrow street with a width of 24 ft (8 m). A contract stipulation required one lane

Building vaults

Another challenge was ‘building vaults’. These vaults are extensions of building basements that extend beyond the property line and under the sidewalk or street. If the need arises as per local laws the city can ‘take back’ these vaults from the building owner. Buildings typically do not use them for storage or easily removed space, but for housing of vital building utilities like boilers and furnaces, electrical equipment, and oil tanks, thus making ‘taking back’ of these spaces difficult and time consuming. The building vaults for 195 Broadway were the most challenging on the project. The vaults extended under the sidewalk to the curb line and were 30 feet (9 m) deep (with two levels—upper and lower) below street level. They housed abandoned oil tanks, electrical and telephone equipment, and through an agreement with the adjacent hotel, staff locker rooms. These spaces were not easily reclaimed in a timely manner that would meet the construction schedule, as removing these vaults were scheduled for early in the project and were on the critical path. The contract called for removal of all the lower level vaults and the majority of the upper level vaults. Unfortunately part of the upper level vaults housed both the electrical and telephone feeds to the building, as well as the telephone switch room, which were difficult to remove and relocate in a timely matter without delaying the contract. An agreement was made with the building owner to allow the electrical and telephone feeders to remain along with the telephone switch room. Resulting in a redesign of the tunnel to allow for these vaults to remain using a ‘notch’ built into the corner of the tunnel box. For 95 feet (29 m) in length the clear height of the tunnel was reduced from 21 feet (6.5 m) to 16 feet (5 m) for a width of 10 feet (3 m). During the proposal phase, the conceptual design called for a moment connection steel frame tunnel box. The clear width was 29 feet (9 m), the clear height was 21 feet (6.5 m) and the length was 300 feet (91 m). The design team switched to a concrete reinforced tunnel box. The team believed that using concrete in this situation would allow construction to proceed faster, and due to the high cost of steel at the time, cheaper. In addition, using concrete would make accommodating any irregularities such as a notch much easier than with steel framing.

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4 4.1

WORKING WITH 100 YR OLD STRUCTURES Fulton Street Station—platform work

The Fulton Street Station under Broadway was one of the original stations built by NYCT and dates back to 1910. Part of the proposal documents were ‘as built’ drawings from the original construction. For this station the drawings indicated a 3 foot (1 m) wide by 4 foot (1.2 m) deep grade beam running the length of the platform, parallel to the tracks, which supported both the platform slab and columns supporting the roof beams. The station structure is supported on grade. The original underpinning scheme utilized these grade beams as an integral part of the underpinning system. During preparatory work for the upcoming track outages, Skanska took exploratory cores of this grade beam, and discovered it was not a concrete grade beam, but was a dry stone wall, with a cementitious facing that gave the appearance of a concrete beam from track level. Instead of the grade beams supporting the platform columns, the columns were supported by individual ‘wedding cake’ spread footings, which are typical of early subway construction. See Figure 5. Since the G.O.s for the track outages were quickly approaching and could not be re-scheduled, the project team, including the NYCT’s in house design staff came up with a solution that would allow for the underpinning of the platform, while still performing the work to underpin the track structure. During an additional 10 G.O.s the project team had to build the ‘missing’ grade beams, while still performing the track work for the track underpinning during the original track outages. The new grade beams would be used to underpin the platform columns and temporarily support the roof structure off of the newly built grade beams. The construction was done in steps.

Step 1 consisted of demolishing the platform slabs and building temporary wooden platforms. Step 2 consisted of installing the new concrete grade beam in between the existing column footings. See Figure 6. Step 3 consisted of jacking the load out of the existing columns and onto temporary columns located on top of the newly installed grade beams. Since this was done the following weekend the concrete had to support the roof loads within 5 days of placing the grade beams. Step 4 consisted of removal of the platform columns and the installation of a mini pile directly below the columns. The mini pile would be used for support for the temporary underpinning, similar to the Cortlandt Street Station. Step 5 consisted of placing the remainder of the grade beams, which was the area where the existing spread footing was located. Step 6 consisted of transferring the load from the temporary columns back onto the existing columns, placing of the platform beams and new platform slab. Since this station is an historic station, the existing nearly 100 year old round, cast iron columns had to be replaced in kind and new steel columns could not be installed, further complicating the matter, as handling of cast iron columns must be done with great care.

Figure 6. View of 4/5 fulton street platform with new concrete grade beams and existing spread footings.

Figure 5. Existing ‘wedding cake’ spread footings at 4/5 fulton street station track level.

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Figure 7.

Existing utilities adjacent to cortlandt St. Station.

These steps were done sequentially along the station platform and in conjunction with the work at track level. This work required the contractor to have multiple crews working within close proximity to each other, and required close coordination among the trades and operating engineers. To ease construction for the grade beam between the columns, the rebar cages were prefabricated during the week and then placed into the excavated area during the weekend G.O. The new grade beam was designed to span in between the mini piles to support the temporary load, and to span in between the future underpinning steel below. All this had to happen while also supporting an existing sewer and a NYCT signal duct bank under the platform. They were originally supported on grade, but to allow for the construction of the new concourse below, they had to be supported by the new platform beams. For this station, with the tight clearances at street level, the city sewer was routed below the northbound platform. Once again, this sewer had to be temporarily flumed, but this time instead of being flumed inside the tunnel, it was flumed into vacated building vaults along Broadway. 4.2 Cortlandt Street Station—platform work As part of the project the northbound Cortland Street platform was to be extended and widened approximately 100 feet (30 m) north to allow for more platform space, wider stairs and new elevators between the street surface, platform level and concourse below. This platform widening and extension was located directly in front of the Millennium Hotel. Once construction started and meetings were held with hotel management and the project team, it was determined that the project’s work would have a serious impact upon the hotel’s ability to conduct business, since the work required a secant pile rig to be in front of its main entrance for up to 4 weeks. As the downtown area was rebounding from the effects of 9/11, all parties agreed that the project as shown on the conceptual design had to be reworked. Through a series of working meetings with all affected parties, the team came up with a solution. The platform would not be extended 100 feet (90 m), but only widened for 40 feet (12 m) and the street elevator would be shifted to the other side of the street. The street stairs would be widened, but not as much as originally planned, and they would be shifted away from the hotel’s main entrance towards the intersection of Dey and Church streets. The platform to concourse elevator would remain as would the platform to concourse stair, but it was re-configured for the design change. While MTA NYCT wanted to maximize the width of the platform along the northbound platform, there were a few constraints with widening of the platform—the existing utilities and the existing building

lines. Upon award of the contract, the project team surveyed the existing stations, and dug some exploratory trenches to locate the existing buildings’ basement walls. The Hotel was set back from the building line above the street level, but below ground it was built right up to the property line. Once the surveys were complete, the project team met with the owner to determine the width of the platform. Working from the building line, space had to be allocated for both the temporary work—secant piles and temporarily hung electrical ducts—and for new work—wall structure, the sewer line and electrical duct bank. The space required for the new sewer was determined by the NYC Department of Environmental Protection (NYC DEP) and was larger than was currently provided for the existing sewer. Through several meetings with NYC DEP’s construction group, a width for the new sewer was agreed upon. Once NYC DEP requirements were satisfied, the design team looked at ways to reduce the wall structure. Typical NYCT construction is steel framing with non reinforced concrete jack arches. The typical width for this construction is 14 inches (36 cm). The design team proposed the same column spacing, but a flat slab with steel reinforcing bars instead of an arch, and used heavier, but shallower members to increase the space available. Every inch mattered to the project team and owner, as both the stairs and elevator shaft had to fit at the platform level. The width of the wall was reduced to 10 inches (25 cm). 5

CONCLUSIONS

The Dey Street Tunnel was a challenging project to design and build due to the constraints of working in an urban environment and around active subways. But an effective design build team, where all parties work together—owner, designer, and contractor— allows for innovative solutions to be developed, which result in the project being built in a cost effective and timely manner. Conducting multiple working meetings throughout all stages of the project with all concerned parties helped streamline the design, review and approval processes, and allowed for quicker resolution to design and construction issues. ACKNOWLEDGEMENTS I would like to thank the following people who helped edit this paper: Jaidev Sankar, Tom Steinhardt and Maura Kelly of AECOM; Stephen Vick of Skanska, USA.; Hsin Wu of MTA Capital Construction and everyone else who helped review this paper.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Developing a document management model for resolving contract disputes for contractor J.B. Yang & K.M. Huang Institute of Technology Management, Chung Hua University, Hsinchu, Taiwan

ABSTRACT: Disputes in public construction works appear increased during the construction stage, because the works suffer many challenges beyond a contractor’s control. If a contractor has limited ability in contract management, it is usually incapable to provide complete documents and the evidence for expenses due to the duties from other contract parties. This makes the contractor suffer from invisible damages, and diminish its project profit. This paper presents a document management model for resolving contract disputes. The model was constructed by analyzing dispute settlement procedures and steps through an actual case. The feasibility of developed model was examined by another case. Research outcomes can assist a contractor to solve increased contract disputes, by which the contractor can manage dispute-related documents proactively. 1

the contractors can manage dispute-related documents proactively.

INTRODUCTION

Disputes in public construction works appear increased during the construction stage, because the works suffer many challenges beyond a contractor’s control. For example, a contractor may incur damage due to unfair provisions (unclear provision but clarifying by the client) in a typical contract, unforeseen site condition, and even the bad weather or the disaster. Generally, a construction contract is issued by a government agency, the draft contract is initially prepared by the agency who mainly concentrates on protecting client’s own rights and interests. Therefore, a contractor usually incurs damage due to the unfair provision in the contract while a dispute appears. Above situations subsequently result in contract claims unceasingly. While disputes occur, if a contractor has limited ability in contract management, the contractor usually incurs loss. An inability contractor is unable to actually prove that the loss is due to the owner’s breach of contract, or the behavior of unite efforts by the owner. Furthermore, an inability contractor is usually unable to provide complete documents and the evidence for expenses due to the duties from other contract parties. This makes the contractor suffer from invisible damages, and diminish its project profit. This study used the technique of case study and selected a local large-scale construction company in Taiwan as the study target. This study analyzed dispute settlement procedures and steps through an actual case to establish a dispute document management model. The feasibility of developed model was examined by another case. Research outcomes can assist the contractor to solve increased contract disputes, by which

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2

RESEARCH METHODOLOGY

The case study method is a qualitative and descriptive research method, not a sampling research method (Feagin et al. 1991; Stake 1995; Yin 2003). The case study method is an ideal methodology when a holistic, in-depth investigation is required (Feagin et al. 1991). The case study method has been used in various domains, particularly in sociological investigations. This study used the case study methodology to identify the dispute settlement procedures and steps from an actual case, and then examined by another case. The cases were retrieved from a local contractor, which is one of the top-ten general contractors in Taiwan. The contractor won compensable cost from the client in these cases. Although the examined cases are not fresh or huge, the characteristics for collecting data to dispute resolution from those cases are worthy to reflect general situations in typical contracts in Taiwan. When the ability for contract management is regarded as an essential capability for a contractor increased, the information learned from Taiwan’s cases is valuable for others. 3

LITERATURE REVIEW

Al-Saggaf (1998) proposed a procedure for construction project delay analysis, consisting of (1) gathering

Table 1.

General delay analysis processes.

Author(s)

Process

Reams (1989)

Step 1: isolate in time when each alleged delay, impact, or acceleration directive occurred; Step 2: select the earliest occurring delay; Step 3: determine how previous delays and impacts have affected the delay under analysis; Step 4: determine the type of the alleged delay; Step 5: perform a schedule analysis; Step 6: interpret the results; Step 7: if warranted from Step 6, allocate time and sum adjustments; Step 8: update the Schedule; Step 9: repeat Step 3 through 8 for each alleged delay or acceleration directive. Step 1: as-planned network and classification of delays; Step 2: identify first delaying event; Step 3: identify progress at delay data; Step 4: progress update the network; Step 5: simulate the first relevant event; Step 6: consider mitigating action; Step 7: subsequent relevant events; Step 8: consider the effect of omissions; Step 9: conclusion and reassess if necessary; Step 10: collate and present results. Step 1: review the bid documents and base-line schedule; Step 2: analyze the scheduling updated; Step 3: make a site visit; Step 4: review correspondence, memos to file, and photographs; Step 5: review the request for information (RFI) file; Step 6: review the change order files; Step 7: review the project progress meeting minutes; Step 8: review the superintendent’s daily report and quality control report; Step 9: interview field personnel; Step 10: review the pay request/pay report; Step 11: prepare an as-built schedule.

Bordoli & Baldwin (1998)

Zafar (1996)

all relevant information concerning the delay; (2) analyzing collected information; (3) identifying root delay causes; (4) classifying delay type and (5) assigning responsibility. For a comprehensive delay analysis, he suggested collecting the information including bid documents, baseline schedule, progress reports, project correspondence/meeting minutes, change-order file and payment requests/reports. Baki (1999) proposed a step-by-step approach for delay claims management in construction. The developed comprehensive claims management program covers three basic claims scenarios: (1) claims prevention; (2) claims preparation; and (3) claims defense. Furthermore, the study also suggested a step-by-step approach for successful claim or defense, including following phases: (1) identify and develop claim issues; (2) factual and detailed schedule analysis; (3) work-hour cost analysis; (4) damage calculations and assessment; and (5) preparation of a claim report. Detailed steps can be found elsewhere. To help a claim analyst for collecting required information, several researchers (Reams 1989; Bordoli & Baldwin 1998; Zafar 1996) proposed various analysis processes. Table 1 shows detailed process. Although these different approaches cover different topics, they can be divided into five phases: (1) preparation phase: to collect required information including as-planned schedule, bid documents, construction daily reports etc.; (2) diagnosis phase: to identify impacted delay events for further analysis; (3) analysis phase: to calculate schedule impact according to each impacted delay event; (4) interpretation phase: to clarify schedule impact on critical path or total duration and (5) summation phase: to summarize all analysis results and to generate a comprehensive analysis report (Yang et al. 2007).

4

CASE INFORMATION

The studied company (termed company X hereinafter) is a registered ‘‘A class’’ general contractor in Taiwan. It was set up in 1960s and had the personnel numbering over 500 persons. In the construction industry in Taiwan, the studied company was ranked as one of the top ten construction firms according to their annual revenues. Recently, the project types executed by company X include housing and hitech buildings, infrastructure, and mass transit projects. Table 2 shows the case information. The project had contract duration of 900 calendar days, but finally delayed extra 836 days. The study case is one of the West Coast Expressway construction projects in Taiwan. Because the construction site was located nearby the sea, the weather condition (wind frequency and high rate) for construction is different to the tender documents described in tendering. This is the issue that pushes the company X to activate claim preparation mechanism. 4.1

SOP for dispute resolution

Company X establishes an SOP (standard operating procedures) for dispute resolution. Figure 1 shows the procedures that start with wining a construction project. The deputy project manager in the project screens tender documents to identify potential claim issues and collects documents related to claim issue incorporated with project staffs. If any claim exists, project team proposes a request to the client for possible compensation. If the client declines the request, company X will establish a dispute case controlled by project-based management mechanism to trace its

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final result. In this situation, potential claim issues are assessed by legal staffs. If the evaluation result reflects that the claim issue is essential to project cost or duration, company X will send a formal letter to the client to do further claim; otherwise, company X

Table 2.

shelves the case until project completion, but will send a format notice to the client to protect its right. In Taiwan, Public Construction Commission, Executive Yuan hosts a mediation mechanism for resolving disputes for public construction projects. In general,

Case information.

Feature

Data

Project name Client

AAA New Construction Project in West Coast Expressway BBB Project Office, Taiwan Area National Expressway Engineering Bureau, Ministry of Transportation and Communications Yuanli Town, Miaoli County $1,029,105,970 NTD (Incl. Tax) 1/1000 of contract cost per day, that is, 1,029,106 NTD/day Commenced on Apr. 1, 2002, anticipated completion date: Sep. 16, 2004, a total of 900 calendar days Delay started on Sep. 17, 2004, project actually completed on Dec. 31, 2006, delayed 836 days

Construction site Contract cost Penalty rate Contract duration Delay information

Contract start

Identify claim issue

Incorporation with project staffs for data collection

Collect documents related to claim issue during contract performance period No Propose a request to the client

Obtain compensable cost from the client

Yes Develop a contract dispute case

All documents are managed by anentitled person

Manage the dispute case by project control mechanism

Potential issues are assessed by legal staffs

Assess the importance of dispute case Important

Discuss the letter with legal staffs

Unimportant 1. Shelve the case until project completion 2. Send a format notice to the client to protect the right

Send formal letter to protect the right

Assess win probability by a lawyer

Low

High Successful

Close the case

Figure 1.

Apply mediation for the dispute case Unsuccessful Try other legal settlement: arbitration or litigation

Standard operating procedures for dispute resolution.

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Close the case

Send/receive files by project office

Correspondence units screen the files Assess its characteristic related to claim issue by a contract manager No

Yes Related to claim issue?

Classify the documents according to general purpose

Contract manager copies the files as-is

Ordinary files

Classify the documents according to claim issue

File archive by serial number

Figure 2.

File archive by quality control records

Contract manager follows up claim development until the end

Document management procedure for identifying dispute-related documents. Write a draft letter to the client for interpretation or protect the right in contract Contract manager assesses wording in contract perspective No Project manager decide to send the letter or not

Close the case

Yes Project site office dispatches a formal letter

Contract manager monitors its development

Contract manager collects related documents and correspondences

Contract manager assesses followup treatments after client’s reply

Contract manager develops dispute management schedule for monitoring

Develop a special file for management

Figure 3.

Dispute-related documents management procedure.

company X will assess the win probability for the case by a lawyer. If the win probability is low, company X will close the case, otherwise, apply mediation for dispute settlement. If the win probability is high, company X may try other legal settlement methods, i.e. arbitration or litigation while losing mediation.

4.2

Identifying dispute-related documents

Project office in study case has a standardized document management procedure (Fig. 2) for identifying dispute-related documents. Under this standardized procedure, all files sent/received by project office will be assessed by a contract manager to classify which

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Stage 1: Basic works

Document Management Model

Stage 2: Preparation works before construction

1. Organize related documents systematically. 2. Study related contract terms. 3. Collect standard operating procedures issued by the client. 4. Collect internal or external dispute cases similar to study project, and possible dispute cases. 5. Collect related laws and regulation on the prohibition against construction. 6. Collect related regulation on contract violation and penalty fees. 7. Classify related contract terms on the dispute clauses. 8. Establish correct construction log recording method by training. 9. Collect information and images on original geographic data of the site before physical construction starts. 10. Survey house fracture cases occurred around construction site 11. Establish a database for public pipeline layout. 11. Develop self-aware cognition in claim for staffs by training, i.e. the contract and risk consciousness.

Stage 3: Supervision and management works during construction

Record and keep the following data: 1. Daily report, monthly report, and weather report. 2. Construction progress report, estimation application. 3. Materials testing report, instruments testing report. 4. Construction process quality control report and record. 5. Construction drawings and self-design shop drawings. 6. Application for construction commencement and on-site inspection records. 7. Documents of construction commencement, completion, suspension, resumption, and approved plan and schedule. 8. Construction improvement records. 9. Meeting minutes, coordination records, and correspondences. 10. All other records from construction site. 11. Safety, hygiene, and environmental protection reports. 12. Accidents and disasters records. 13. Insurance, assurance, compensation, and claims records. 14. Third party assessment records. 15. Vouchers of additional expenditures. 16. Vouchers of direct and indirect costs resulted from delayed construction duration.

Stage 4: Necessary and urgent works after dispute appear

Figure 4.

1. Appoint a project manger. 2. Develop a construction team. 3. Assign a contract manager with construction and legal backgrounds.

1. Review and organize related documents prepared during previous stages. 2. Establish document management mechanism according dispute-related documents management procedure. 3. The contract manager intervenes into the dispute case to lead the case. (1) Monitor the time effectiveness of the documents, and regulations regarding the contract. (2) Read the daily report filled out on-site engineers every day. (3) Communicate with on-site engineers, the client, and supervisor closely. (4) Collect vouchers for additional cost, and ask on-site engineers to take photos of construction process. (5) Classify the cost of additional fees, especially for the costs related to dispute case. (6) Claim for compensation against the client. 4. The contract manager shall keep close interaction with the legal staffs of the company, and seek assistance when necessary. 5. The contract manager shall participate in all forms of meetings that are related to the dispute. 6. Refer to similar dispute cases to analyze root problems. 7. Propose defensive and offensive strategies, and prepare important evidences. 8. Actively analyze dispute key points. 9. Identify the matters that the intermediator (judge or arbitrator) wants to clarify. 10. Prepare complete proofs for argument, and clear supplementary information.

Document management model for dispute resolution.

document is related to dispute cases. The contract manager not just copies all related files as-is, but also classify related documents to specified dispute issue. Furthermore, the contract manager will manage all documents until the dispute case close. 4.3

Dispute-related documents management procedure

After dispute-related documents were identified, project office in study case established a document management procedure (Fig. 3), in which several situations will be monitored: (1) specifications and drawings are not clear; (2) construction obstacles onsite encountered; (3) construction progress impeded by third party occurred; (4) client-finish contract items

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are delayed; and (5) other obstacles to construction progress encountered. Figure 3 shows detailed processes. If the project manager decides to send a formal letter to client for specification and drawings interpretation or protect the right in contract, the contract manager should collect several types of documents for reference. These documents include: (1) correspondences between both contract parties; (2) contract provisions related to study claim; (3) final dispute results from nearby projects; (4) any extra cost for claim case; and (5) other information and files in favor of claim case. Furthermore, the contract manger should develop a dispute management schedule for monitoring. Based on above tasks, a special file will be developed to trace the development of claim case.

5 5.1

DOCUMENT MANAGEMENT MODEL Model description

Based on the study case, a documents management model is established. The model consists of four stages: (1) basic works for all construction projects; (2) preparation works before construction commence; (3) supervision and management works during construction; (4) necessary and urgent works after dispute appear. All detailed tasks in each stage are listed in Figure 4. In stage 1, some basic works should be finished. Among these works, to assign a contract manager with construction and legal backgrounds is the most essential task. In stage 2, before physical construction commence, several preparation works should be finished. Excluding collecting various documents and references, to develop self-aware cognition for claim for all staffs is necessary. This makes all staffs doing the right works, collecting the right data, and being on the right time. In stage 3, during construction, all project staffs should try to record and keep completed constructionrelated data and even computerized files. Figure 4 details the required data. In stage 4, if a dispute case appears, some necessary and urgent works should be finished. The contract manager needs to review and organize the documents related to dispute case during previous stages, and establishes a document management mechanism, as Figure 3 shows, to manage dispute-related documents. Other complete works can be found in Figure 4. 5.2

Model examination

Each construction project is usually recognized as a unique one; therefore, each project has its special circumstances. For verifying the feasibility of developed model to general dispute resolution, this study used another case from company X. Although most situations encountered by first case did not occur in second one. Developed model was assessed by project participants in another case. All participants agreed that the developed model provides a guiding map for collecting required documents and will be helpful for resolving construction disputes. 6

CONCLUSIONS

Contract disputes appear frequently in construction projects. Based on studying cases that the contractor won cost compensation from the client, this study provides valuable outcomes. This study established a

contract dispute document management model learned from one case and evaluated by another case. A contractor may follow the developed model during the stages of contract commencement, performing, and dispute occurred. The model can help a contractor to carry out document management at various stages more efficiently than managing by experience. Furthermore, this study presents several procedures for dispute-related document management, which are valuable for contract dispute resolution. A contractor may follow these procedures to face encountered disputes. Developed model can prevent a contractor from being in the state of inferiority because of the ineffectiveness and incomplete document management during prosecution period. Research outcomes can also assist a contractor to solve the increasing contractual disputes through developed document management model. ACKNOWLEDGEMENTS The authors would like to thank the National Science Council, Taiwan, ROC, for financially supporting this research under Contract No. NSC96-2221-E-216027-MY2 and the Chung Hua University for providing partial funding (Funding No. CHU96-2211-E-216027-MY2) to this research.

REFERENCES Al-Saggaf, H.A. 1998. The five commandments of construction project delay analysis, Cost Engineering 40(4): 37–41. Baki, M.A. 1999. Delay claims management in construction—a step-by-step approach, Cost Engineering 41(10): 36–38. Bordoli, D.W. & Baldwin, A.N. 1998. A methodology for assessing construction project delays. Construction Management and Economics 16(3): 327–337. Feagin, J., Orum, A. & Sjoberg, G. (eds.) 1991. A case for case study. University of North Carolina Press, Chapel Hill, NC. Reams, J.S. 1989. Delay analysis: a systematic approach. Cost Engineering 31(2): 12–16. Stake, R. 1995. The art of case research. Sage Publications, Newbury Park, CA. Yang, J.B., Yin, P.C. & Kao, C.K. Comparison of various delay analysis methodologies for construction projects. 4th International Structural Engineering and Construction Conference (ISEC 04), Melbourne, Australia, pp. 1395–1401, 2007. Yin, R.K. 2003. Case study research: design and methods. 3rd edition, Sage Publications, Thousand Oaks, CA. Zafar, Z.Q. 1996. Construction project delay analysis. Cost Engineering 38(3): 23–28.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Development of a decision-making model for requirements management N. Krönert & G. Girmscheid Institute for Construction Engineering and Management, ETH Zurich, Switzerland

ABSTRACT: Nowadays, investors and clients strive to bring construction projects to market as quickly as possible in order to exploit the yield potential early on. The objectives and requirements of the stakeholders must therefore be focused and captured as early as possible during the simultaneous engineering process in order to avoid subsequent amendments to the project that are either expensive or reduce the return. A concept for systematic requirements management to implement the target-oriented decisions at the appropriate stages of the project has not yet become successfully established in modern-day construction practice. This research project is collating the objectives, requirements and decisions needed for implementation at each stage of a project and combining them into one generic decision-making model for requirements management. This decision-making model forms the basis for optimizing the accelerated, target-oriented simultaneous engineering planning processes in order to ensure early implementation of the objectives in a decision-oriented requirements management process. 1

INTRODUCTION

2

Construction processes are constantly gaining in speed, from the fundamental decision in favor of the project and investment right up to completion of the same. The ‘‘time-to-market’’ principle has become commonplace, even in the construction industry. The winners are those who can provide and rent property first in response to market demand. It is therefore absolutely crucial to ensure the target-oriented structuring of processes and decisions without multiple changes to planning and calculations, some of which are even implemented during the construction of the building. Total service contractors and planners must accept this challenge, define the objectives together with the property developers, and subsequently derive the requirements structure to meet these objectives in order to enable target-compliant decisions to be made in the right phases during the ongoing construction process. A closer look at this issue reveals, however, that a requirements management system for the construction industry is virtually non-existent (Fernie et al. 2003). It was not until Girmscheid developed a generic axiomatic requirements management (Girmscheid 2007a) that a basic concept was provided for processing requirements in the construction sector. With the aid of this theory the objectives, requirements and decisions needed for implementation at each stage of a project shall be generically combined into one decision-making model for requirements management.

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RESEARCH METHODOLOGY AND OBJECTIVES

Construction management with their socio-technical challenges can be classified as a cross-section discipline that combines engineering and business management sciences and includes psychological, sociological and linguistic influences (Girmscheid 2007b). Based on these aspects, research in construction management sciences is assigned to the world of products of the human mind in Popper’s theoretical classification of three worlds (Popper 2002). The hermeneutic approach to science is aimed at changing, shaping and managing socio-technical reality. As such, the model is being developed on the basis of the radical constructivism as defined by Piaget and Ernst von Glasersfeld within the framework of the hermeneutic approach to scientific understanding. Constructivist scientific philosophy (Glasersfeld 1998) builds on the issue of how new and/or improved realities can be created by mankind and, as such, knowledge and reality generated. It is not the structure of external reality but rather reality itself that is perceived. Radical constructivism replaces the perception of truth with the quality criteria of viability, validity and reliability. Generic-deductiv methods and the theoretical development of processes drive the constructivist research approach. The presented decision-making model for requirements management (RM-DM-Model) is constructed actional generically-deductive as a target-means relationship by the constructivist research paradigm

(Girmscheid 2007b; Glasersfeld 1998; Piaget 1973). The scientific quality is achieved by triangulation (Yin 2003) due to: – Viability of the generic-deductive model – Validation through a theoretical framework – Reliabilitation through testing the intended impact (target-means relation) The presented RM-DM-Model is viably constructed as an actional, generic-deductively structured model according to the target-means-relations. For the validation purpose the RM-DM-Model is theoretically-deductively structured using the principles of system theory (Bertalanffy 1973; Boulding 1956). Reliabilitation will be achieved due to realization test to check if alternative target relation exists under equal means. According to Yin (2003) the above sketched triangulation concept fulfills the scientific quality requirements.

3

STATE OF PRACTICE AND RESEARCH

An established requirements management system is currently non-existent in the construction industry (Fernie et al. 2003). By contrast, extensive requirements process and requirements management approaches do exist for other sectors of industry. These include, above all, the software sector and the aviation, aerospace and automotive industries. Standards and comprehensive application literature exist for all of these industries. Important standards have been issued, for example, by the ‘‘Institute of Electrical and Electronics Engineers’’ (IEEE). These include the ‘‘Guide for Developing System Requirements Specifications’’ (IEEE 1233 1998) and the ‘‘Recommended Practice for Software Requirements Specifications’’ (IEEE 830 1998). A comprehensive collection of practice-oriented literature (Leffingwell et al. 2000; Pohl 2007; Robertson et al. 1999) exists alongside the standards. These requirements processes and management approaches cannot, however, be simply transferred to the construction industry since the products of the construction industry, in particular, are mainly complex and extensively custom built. As such, requirements management and the decision-making process in the construction industry must be developed to match the construction phases and content of the same for the specific construction project. The aim of requirements management in the construction industry is to develop the yield potential for the investor in line with the objectives while keeping transaction costs as low as possible. The minimum principle for transaction costs therefore necessitates phase-oriented requirements management for the

largely intangible, interactive and integrative construction process. This requirements management must focus on the objectives of the investor/property developer to enable structured decisions to be made in the right phases in order to achieve the objectives. To this end, milestones must be defined for the end of each phase to enable an assessment of whether the basis for making decisions is sufficient to allow the project to progress to the next phase. Various approaches to breaking a construction project down into phases can be chosen (Diederichs 2006; Girmscheid 2006; Halpin et al. 1998; Werner 2003). A phase model based on HOAI is commonly used in Germany. This phase model is used as a basis for further procedure in the decision-making model under development. The state of research is just as varied as the state of practice. The software sector is currently addressing specific issues aimed, above all, at improving existing processes and methods. Examples of this include the improvement of communication (Fricker et al. 2008) or the identification of stakeholders (Ballejos et al. 2008). By contrast, fundamental concepts are virtually non-existent in the construction sector. Girmscheid combines various concepts in his research to produce an integrated and life cycle oriented requirements management system (Girmscheid 2007a). Apart from this holistic perspective, only a very few research projects focus on partial aspects of requirements management (Evbuomwan et al. 2002). This research work builds on the holistic approach according to Girmscheid and delves deeper into the correlation between objectives, requirements and decisions, incorporating a phase-oriented axiomatic approach. 4

QUALITATIVE AND QUANTITATIVE REQUIREMENTS PROCESS MODEL

The holistic requirements process and decision making model (RM-DM-Model) comprises a qualitative descriptive part and a quantitative axiomatic part. The qualitative model focuses, above all, on the fundamental approach to specifying requirements and on the formation of decision-making functions based on the objectives and requirements from each project phase and their recursivity. The quantitative model addresses the interaction among the objectives, requirements and decisions based on mathematical formulation of the decision-making functions. The RM-DM-Model forms part of the overall construction process (Girmscheid 2006).The requirements management must be incorporated across the entire project duration and, as such, across all phases of the entire construction process. It can therefore be classified as a support process of the provision process of products and services.

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4.1

In addition to the methods for capturing and specifying requirements (Girmscheid 2007c; Pohl 2007; Robertson et al. 1999; Rupp 2007; Sommerville et al. 1997), the documentation and validation of requirements are, above all, extremely important (Pohl 2007; Sommerville et al. 1997). The process of generic axiomatic requirements management (RM-DM-Model) necessitates that requirements are detailed and expanded within the individual project phases to ensure an increasing level of specification. Generic spheres are therefore created based on axiomatic planning theory (Albano 1992) to enable a better structuring and interactive identification of requirements. These spheres and phases must be adapted to the characteristics of the respective project (Fig. 3).

Qualitative requirements process model

The qualitative part of requirements process and decision making model (RM-DM-Model) can basically be divided into two work packages. The first of these primarily focuses on identifying, capturing and specifying customer objectives and requirements. The second package repetitively focuses on addressing the objectives and requirements at each phase using decision-making functions. As such, the holistic RMDM-Model defines the objectives and requirements in each basic cybernetic sub-process to enable decisions to be made in respect of the aesthetic, functional, technical and qualitative characteristics of the building (Fig. 1). The main target vector is formed in the project phases on the basis of the original objective. This might, for example, contain the architectural, functional, technical, qualitative and financial sub-goals for the principle objective of erecting an office building to house 1000 employees. The requirements based on the objectives must be developed and specified if the objectives are to be converted into decisions. As the intangible objectives become increasingly defined, objectives are converted into decisions interactively in the individual phases of a construction project in a basic cybernetic sub-process that incorporates the existing interdependencies between technical, functional, qualitative and aesthetic characteristics and standards, etc., and the retrospective decisions made in parallel phases (Fig. 2). The basic cybernetic sub-processes will be repeated in each project phase detailing the requirements and continuing with the necessary decisions at each specific milestone. Fundamental quality criteria must be defined for the implementation of the project orientated requirements catalog as outlined in the ‘‘Recommended Practice for Software Requirements Specifications’’ (IEEE 830 1998). Although such a fundamental catalog (DIN 18205 04-1996) does exist in Germany in the shape of DIN 18205, it does not include any phase orientation.

Quantitative requirements process model

The qualitative process model links both the time and content factors of the target vectors with the requirements matrix in each project phase to form the decision Basic objectives (Zi)0

Identify requirements from objectives

Modification of the objectives (Zi)n to (Zi)n+1 target ( ∃Zi )n : ( Zi )n < ( Zi )n

( Zi )n

< ? ( Zitarget ) n >

Objectives (Zi)n

Obj. Achievement ( Zi )n Requirements

( A j )n

Decision

( Ek )n

Determine the level of objective achievement target ( ∀Zi )n : ( Zi ) > ( Z i )n n

( Zi )n = f ( Zi ,Ek ,zik )

Basic cybernetic sub-process for identifiying requirements and producing a basis for decisions

Customer objectives (Z i) 0

Project development

Pre -planing

( Ek )n = f ( A j,e jk )

Figure 2. Basic cybernetic sub-process for processing objectives and requirements in a decision-making in each project phase.

Design / Approval planing

Execution planing / Award phase

Execution phase / Property supervision

Project completion

Holistic requirements management and decision making process model (RM-DM-Model).

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Milestone VI

Milestone V

Milestone IV

Milestone III

Milestone II

Milestone I

Specify / document requirements (Aj)n

Validate requirements (Aj)n

Make decisions

Process Milestone

Figure 1.

Analyze requirements

Construction Management

Developer / Consultant

Phases

4.2

tn 1

Generic Spheres

Client Requirement / Objective

Functional Sphere Physical Sphere

Functional Requirement

Draft Parameters

Process Sphere

Process Variables

Client Sphere

Activities

Client Sphere

Actn-1

tn 1, n

Output : Client Result

Feedback Process System and user re-quirements ((output) and cost frame

Draft parameters:: Dimensions,, quality standards,, materials, costs

Interdependencie s in the planning and constcution process

Result: • Function • Quality/Cost • Timeframe (Output)

Figure 3. Generic spheres and interaction processes (Girmscheid 2007a).

vector. In addition, retrospective decisions from previous phases and lateral decisions from the project phase in question, both of which influence the respective decision, are also linked in terms of both content and time. As such, a requirements matrix and a retrospective and/or lateral decision matrix must be formed for each phase, or even each sub-phase of a project. A decision vector comprises decisions, such as – – – – –

selection of the building’s function selection of the system or component elements definition of design parameters and specification completion of documentation or plans ...

The respective interdependencies are linked to each other in the decision matrix. Requirements management is sub-divided into – identifying the client’s objectives – specifying and detailing the requirements for meeting the client’s objectives at the respective phase of the project – making decisions at the respective phase of the project to ensure that the objectives are met at the milestone. The purpose of the quantitative process model is to identify the critical decision path within the planning and execution processes. The GERT method (Moore et al. 1976) can be used to calculate the critical decision and flow path. This method also takes account of risk activities and likelihoods of occurrence. In the case of parallel requirements processes within the project development, planning and execution processes, the interdependent relationships and cybernetic development of the information relating to the requirements and decisions to ensure compliance with the property developer’s objectives must be incorporated into the analysis (Girmscheid 2007a). Various temporal process interdependencies exist in this respect between the requirements and the retrospective and lateral decisions at point in time t-n and the decisions made at point in time t (Fig. 4).

tn n

1, n

n-1

Bn

1, n

ta,n

=1

n

n

ta,n

t

ta,n

t

Bn

1, n

=0

n-1

tn 1, n

tn 1, n

=0

n

ta,n

ta,n 1 t te,n 1 n 1

tn 1 te,n 1

ta,n 1 n-1

Actn

Temporal Interdependency

Client objectives,, attributes in respect of the building/solution

te,n 1

n-1

t

Bn

1, n

=

tn 1,n tn 1

t

Bn

1, n

=

tn 1,n tn 1

Figure 4. Temporal interdependencies between the development of, and need for, information to enable decisions to be made and to trigger subsequent processes (Girmscheid 2007a).

Linking the interdependencies allows qualitative conclusions to be drawn from the functional requirements with regard to the necessary planning and execution process steps. The quantitative requirements process model therefore describes a specific and practice-oriented method for converting objectives into requirements and the ensuing decisions that takes all requisite interdependencies into account.

5

CONCLUSIONS

These holistic qualitative and quantitative requirements process and decision making (RM-DM-Model) models represent a fundament approach to generic axiomatic requirements management. Initial qualitative and quantitative approaches for processing requirements and objectives in respect of the necessary decisions and design parameters are presented. The RM-DM-Model can be used to develop a holistic approach that incorporates requirements management in the construction industry. This requirements management not only considers the entrepreneurial objectives (cost, quality, deadlines), but also the wishes and ideas of the client and other stakeholders.

REFERENCES Albano, L.D. (1992). An Axiomatic Approach to Performance-Based Design. Massachusetts Institute of Technology. Ballejos, L.C., and Montagna, J.M. (2008). Method for stakeholder identification in interorganizational environments. Requirements Engineering 13(4): 281–297. Bertalanffy, L.v. (1973). General system theory: foundations, development, applications. Harmondsworth: Penguin. Boulding, K.E. (1956). General Systems Theory—The Skeleton of Science. Management Science 2(3): 197–208. Diederichs, C.J. (2006). Immobilienmanagement im Lebenszyklus: Projektentwicklung, Projektmanagement, Facility Management, Immobilienbewertung. Berlin: Springer. DIN 18205. (04-1996). DIN 18205—Bedarfsplanung im Bauwesen. Deutsches Institut für Normung e.V., Berlin: Beuth.

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Evbuomwan, N.F.O., Kamara, J.M., and Anumba, C.J. (2002). Capturing client requirements in construction projects. London: Thomas Telford. Fernie, S., Green, S.D., and Weller, S.J. (2003). Dilettantes, discipline and discourse: requirements management for construction. Engineering, Construction and Architectural Management 10(5): 354–367. Fricker, S., Gorschek, T., and Glinz, M. (2008). GoalOriented Requirements Communication in New Product Development. 2nd International Workshop on Software Product Management (IWSPM’08), Barcelona, Spain. Girmscheid, G. (2006). Strategisches Bauunternehmensmanagement prozessorientiertes integriertes Management für Unternehmen in der Bauwirtschaft. Heidelberg: Springer. Girmscheid, G. (2007a). Fast Track Projects—Generisches, axiomatisches Anforderungsmanagement. Bauingenieur 82(5): 224–230. Girmscheid, G. (2007b). Forschungsmethodik in den Baubetriebswissenschaften. Zürich: Institut für Bauplanung und Baubetrieb, ETH Zürich. Girmscheid, G. (2007c). Projektabwicklung in der Bauwirtschaft Wege zur Win-Win-Situation für Auftraggeber und Auftragnehmer. Berlin: Springer. Glasersfeld, E.v. (1998). Radikaler Konstruktivismus Ideen, Ergebnisse, Probleme. Frankfurt a.M.: Suhrkamp. Halpin, D.W., and Woodhead, R.W. (1998). Construction management. New York: Wiley. IEEE 830. (1998). IEEE Recommended Practice für Software Requirements Specifications (IEEE Std 830-1998). IEEE, New York: IEEE Standards Board.

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IEEE 1233. (1998). IEEE Guide for Developing System Requirements Specifications (IEEE Std 1233-1998). IEEE, New York: IEEE Standards Board. Leffingwell, D., and Widrig, D. (2000). Managing Software Requirements—A Unified Approach. Reading, Mass. [etc.]: Addison-Wesley. Moore, L.J., and Clayton, E.R. (1976). GERT modeling and simulation fundamentals and applications. New York: Petrocelli/Charter. Piaget, J. (1973). Erkenntnistheorie der Wissenschaften vom Menschen. Frankfurt (M.): Ullstein. Pohl, K. (2007). Requirements Engineering—Grundlagen, Prinzipien, Techniken. Heidelberg: dpunkt. Popper, K.R. (2002). Logik der Forschung. Tübingen: Mohr. Robertson, S., and Robertson, J. (1999). Mastering the requirements process. London: ACM Press. Rupp, C. (2007). Requirements-Engineering und— Management—professionelle, iterative Anforderungsanalyse für die Praxis. München: Hanser. Sommerville, I., and Sawyer, P. (1997). Requirements engineering a good practice guide. Chichester [etc.]: Wiley. Werner, U. (2003). VOB : Vergabe- und Vertragsordnung für Bauleistungen; HOAI: Honorarordnung für Architekten und Ingenieure. München: Dt. Taschenbuch-Verlag. Yin, R.K. (2003). Case study research: design and methods. Thousand Oaks, Calif.: Sage.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Improving the MEP coordination process through information sharing and establishing trust T.M. Korman California Polytechnic State University, San Luis Obispo, CA, USA

ABSTRACT: The coordination of mechanical, electrical, and plumbing (MEP) systems in the design phase of building construction requires the work of several trades to locate equipment and route connecting elements for each system to avoid physical interferences, allow full system functionality, and comply with several types of criteria. The process involves sequentially overlaying and comparing drawings from multiple systems, during which representatives from each MEP trade work together to detect, and eliminate spatial and functional interferences between MEP systems. Coordination teams usually belong to different professions, ethnic groups, nationalities, and organizations, which often makes trust difficult. This research asks how the MEP coordination process can be improved using concepts of information evolution and sensitivity combined with trust theory. To explore this question a qualitative study of MEP practitioners was conducted based upon semi-structured interviews. The result of this research provides a foundation insight to improving the coordination process through information sharing and establishing trust. 1 1.1

INTRODUCTION Mechanical, electrical and plumbing coordination

The coordination of mechanical, electrical and plumbing (MEP) building systems requires the work of several design professionals and specialty contractors to design and route connecting elements for building systems to avoid physical interferences, allow full system functionality, and comply with several types of criteria. The coordination process involves sequentially overlaying and comparing drawings from multiple systems, during which representatives from each MEP trade work together to detect, and eliminate spatial and functional interferences between MEP systems. This multi-discipline effort is time-consuming, expensive and knowledge critical to the project life cycle. The result of the coordination effort is a product that meets the design intent, is constructible, and can be maintained by operations personnel; however, this product relies on the coordination team establishing trust and sharing information. This research asks how the MEP coordination process can be improved using concepts of information evolution and sensitivity combined with trust theory. 1.2

MEP coordination teams

MEP coordination teams are formed from representatives from each of the specialty contractors that have been contracted to fabricate and install the building

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systems (HVAC wet and dry, plumbing, electrical, and fire protection). They begin meeting after design and preliminary routing of all building systems (mechanical, plumbing, electrical, etc.) has been completed. The design is considered complete when engineers have sized all components (e.g., HVAC duct, pipe, conduits), completed the engineering calculations, and produced the diagrammatic drawings; however, engineers have not defined specific routing. Most specialty contractors refer to these drawings as schematic design drawings. It is the coordination teams’ task to produce a set of coordination drawings which determine the exact location of each building system components.

1.3

Trust

All organizations involve social relationships which a major component is trust. Trusts is considered to be necessary for cooperation and an important influence on project performance, especially in the MEP coordination process were members of the coordination teams usually belong to different professions, organizations, and possess a varied educational background. These factors often make trust very difficult because team members have different goals and expectations. Design professional are concerned with maintaining the building systems functionality, specialty contractors are concerned reducing the cost of fabricating and installation of the systems, and operations and maintenance personal are primarily with the ease of

serving the systems upon completion of the facility. MEP coordination team members must often use some form of technology to communicate. This can make communication more difficult and less personal. This often makes it more difficult to develop shared understandings, upon which trust can develop. Misunderstandings that can damage trust are more likely to happen and once trust is damaged it is difficult to repair. 1.4

Low Sensitivity

Slow Evolution

Figure 1.

Best and worst case scenarios.

Fast Evolution Swift & Uncritical

Swift & Critical

Low Sensitivity

High Sensitivity Slow & Uncritical

Slow & Critical

Slow Evolution

Figure 2.

Trust/task scenarios.

Fast Evolution Superficial Trust Development

Evolution and sensitivity with respect to trust

The evolution rate of trust is defined as how quickly a person decides that someone is trustworthy. If trust is essential or even helpful in the performance of the task, then fast evolution of trust is better. Trust sensitivity is the degree to which the project can be adversely affected by a poor trust decision. A poor trust decision is deciding to trust when it is not warranted or withholding trust when trust is warranted. If trust sensitivity is high it is desired that all team members are cautious before deciding whether to trust or not to trust. Thus the best case scenario is when the situation allows for fast trust evolution (trust develops quickly) and low trust sensitivity (an incorrect trust decision does not have great adverse consequences). The worst case scenario is when the situation allows for slow trust evolution (it takes a long time to establish trust) and high trust sensitivity (trust is a very important decision).

High Sensitivity Worst Case Scennario

Information sensitivity and evolution

During the course of the MEP coordination process, the sharing of information is critical. Information regarding design intent, constructability, and operations and maintenance is all shared. Evolution is that rate at which information becomes shared. Fast evolution means that preliminary information can be counted on early in a project. Additionally, during the course of the MEP coordination process, the information is constantly being changed and updated as the process continues through the coordination effort. Information regarding the systems routings and equipment locations is shared as it becomes available. Sensitivity is the term used to describe the degree to which changes in information affect the project. High sensitivity means that a minor change is the information available has the potential to cause major changes in the overall coordination effect. There are two potential applications of evolution and sensitivity with respect to trust. One is the way evolution and sensitivity relates to trust. The other is the way evolution and sensitivity applies to the performance of the particular interdependent tasks. 1.5

Fast Evolution Best Case Scennario

Working Trust Development

Low Sensitivity

High Sensitivity

Background Trust Development

Guarded Trust Development

Slow Evolution

Figure 3.

Development of different types of trust.

Thus the different quadrants can be characterized by how swift is the development of trust and how critical the trust decision is. (See Figure 2). These different scenarios are likely to produce different types of trust relationships between workgroup members. (See Figure 3).

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2

RESEARCH METHOLOGY

The impetus for the research grew out of the need to determine how to improve the overall performance of the MEP coordination teams. Preliminary research revealed that MEP coordination team members often lack the trust of each other. Reasons stated that indicated that this was most frequently due to members belonging to different professions, organizations, and possessing a varied educational background. Existing literature on trust and on improving project performance indicated that performance could be improved by increased information sharing. Prior research in the area of information sensitivity has demonstrated that the more critical the information the larger the impact on downstream activities, e.g. project scheduling and construction estimating; therefore, the earlier the team members shared information, the earlier they were in establishing trust. Additionally, prior research as demonstrated that as information evolves, that faster that information is shared among project participants there is a correlation in the quality of the final project or project. Interview questionnaires were designed to focus data collection on critical incidents which impact the MEP coordination process. Interviews were semistructured with design engineers who design MEP building systems, specialty contractors which fabrication and install MEP building systems, and operations and maintenance personnel who service MEP building systems. Additionally, interviews were conducted with project managers representing general contractors who oversee the MEP coordination process. With general contractors, those designated as MEP coordinators were selected for interviews. 3

SUMMARY OF FINDINGS

To tackle these questions, several trust theories were integrated into a model that identifies the major variables that predict interpersonal trust and describes the relationship between them. Then MEP coordination teams were observed at work. Following observations, team members were interviewed individually. Results of the interviews suggest that crossfunctional, geographically distributed workers may rely on early impressions of perceived trustworthiness when evaluating how their team members are delivering on commitments, because reliable information about actual follow-through is lacking or difficult to interpret. Consistent with this, it was determined that perceived trustworthiness, perceived follow-through and trust were relatively stable over time. Evidence that MEP coordination team members perceive one another to be less trustworthy when they are distributed as compared with collocated. It was also determined that trust is more stable in distributed

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dyads, a pair of workers—it increases less, but it also decreases less than in collocated dyads. In inter-organizational MEP coordination teams, coordination efforts viewed trust differently and as MEP coordination teams moved from a hierarchy to clan or network organization management, the coordination team needs to consider how to encourage norms of trust for these two different roles. When MEP coordination efforts employed the use of Building Information Modeling (BIM) technology, the teamwork was found to have positive as well as negative effects on trust. The geographic distribution reduced task interdependence, personal communication, perceived trustworthiness and trust. The use of the BIM model and cross-functional nature of MEP coordination teamwork increased perceived trustworthiness, supporting the theory of Swift Trust. The MEP Coordination team members were more likely to be geographically distributed and to have lower task interdependence and to have less personal communication. Also noteworthy, was that the trustor’s perceptions of the trustee’s follow-through and ability to the trustee matched the project manager’s assessment of the trustee. But trustors in distributed dyads, rated the ability of their partners lower than the MEP coordinator, who was working on behalf of the general contractor. Finally, it was determined that the trustors in dyads with discipline diversity assessed the trustee’s ability and perceived follow-through higher than those in the same discipline. Trusting improved the trustor’s process performance in terms of better flexibility, information sharing, problem solving and creativity, but had a negative effect on the trustor’s output performance. In contrast, being trusted improved the trustee’s output performance in terms of better time, cost and quality without affecting the trustee’s process performance. The quality of information had a greater impact on trust than the quantity of information. Trust was more strongly associated with the organization’s culture of openness. Also noted were some interesting differences between distributed and collocate dyads. In collocated dyads, a team member’s evaluation of another’s follow through determines the perceived trustworthiness of the other party and that predicts trust. In distributed dyads, it is the other way around. The team member uses perceived trustworthiness of the other team member to evaluate the extent to which the team member is perceived to have followed through and that predicts trust. This implies that in a distributed team, team members use their opinion of the person to assess his or her performance. This could happen because in a distributed team it is difficult to get information about a team member’s performance. In a distributed team that is also cross-functional, this difficulty could be accentuated by the inability to understand the other partner’s discipline.

4

CONCLUSIONS

Interactions between factors that influence trust and the effects that trust has on job performance and satisfaction are diverse. These diverse and cross-functional teams require trust between team members to accomplish their goals, but they also make trust difficult to achieve. Some theories propose that interpersonal trust is based upon shared social categories, roles, third party information, social rules, history of the relationship and the trustor’s disposition. Although most of these theories have been tested individually, no model combining these theories has been tested yet. These theories lead to questions such as: When these trust development factors are combined, which ones are the most influential in predicting trust in global, cross-functional teams? Are these factors consistent under all conditions? What risk factors that could lead to a failure of trust? These questions are important because as MEP coordination teams become more diverse, more distributed and tackle more complex projects, the risk of failure due to lack of trust increases. When forming MEP coordination teams, it would be useful to know if there are limits on the extent of team diversity that one team can handle.

REFERENCES Korman, Thomas, M. 2001. Integrating Multiple Products over their Life Cycle: An Investigation of Mechanical, Electrical, and Plumbing Coordination. Ph.D. Thesis, Stanford University, CA, June 2001. Tatum, C.B., and Korman, Thomas M. 2000. Coordinating Building Systems: Process and Knowledge. ASCE Journal of Architectural Engineering, Volume 6, No. 4, December 2000, pp. 116–121. Thompson, Leigh, Aranda, Eileen, and Robbins, Stephen. 2000. Tools for Teams. Pearson Custom Publishing, Boston, MA, 2000. Zolin Roxanne, Fruchter, R., and Levitt, R.E. 2003. Realism and Control? Key Characteristics of Problem-based Learning Environments as a data source for work-related studies, International Journal of Engineering Education, 19(6) 788–798 Zolin, Roxanne, Fruchter, Renate, and Levitt, Raymond E. 2000. Building, Maintaining And Repairing Trust In Global AEC Teams. International Conference on Computers in Civil and Building Engineering. Zolin, R., Hinds, P.J. and Fruchter, R. 2003. Communication, Trust and Performance: The Influence of Trust on Performance in A/E/C Crossfunctional, Geographically Distributed Work, Stanford University, CA, Center for Integrated Facility Engineering. Working Paper No. 78.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Key competences of design-build clients in the People’s Republic of China B. Xia & A.P.C. Chan Building and Real Estate Department, The Hong Kong Polytechnic University, Hong Kong, China

ABSTRACT: The design-build (DB) system has been demonstrated as an effective delivery method and gained popularity worldwide. Although there are an increasing number of clients adopting DB method in China, most of them remain inexperienced with method. The objective of this study is therefore to identify the key competences that a client or its consultant should possess to ensure the success of DB projects. Face-to-face interviews and a two-round Delphi questionnaire survey were conducted to find the following six key competences of clients, which include the (1) ability to clearly articulate project scope and objectives; (2) financial capacity for DB projects; (3) capability in contract management; (4) adequate staff or consulting team; (5) effective coordination with contractors and (6) experience with similar DB projects. This study will hopefully provide clients with measures to evaluate their DB competence and further promote their understanding of DB system in the PRC.

1

INTRODUCTION

Design-build (DB) has been demonstrated to be an effective delivery method and has gained popularity overseas in recent years. It provides owners with a number of advantages such as single-point responsibility, time saving, enhanced financial certainty, reduced disputes and increased productivity. At the some time however, the owners are also required to possess certain competences to successfully deliver DB projects. For example, the owners are required to be very clear about the project scope and expected outcomes, and provide firm performance specifications at the early stage of the project. It is generally accepted that the DB method, compared with the traditional delivery system, has new requirements for the owners. Many researchers and practitioners have discovered that the owner’s competences are critical to the success of DB projects. Songer and Molenaar (1997) indicate that owner’s competence is one of critical factors for successful public-sector DB projects. Mo and Ng (1997)’s survey shows that architects and builders in Hong Kong view client’s experience as critical to the DB project success. Molenaar and Songer (1998) relate owner’s capabilities to the success in the selection of public sector DB project. Leung (1999) and Chan (2001) find that client’s competences were important to bring successful DB project outcome. Quatman (2003) asserts that the success of DB projects depends on the owner’s abilities and attitudes. Ling (2004) uses neural network to predict performance of DB projects in Singapore and discovered that key variables affecting project performance may be attributed to both contractors and clients.

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In the construction market of the PRC, many clients lack adequate competences to successfully deliver DB projects. Currently, only less than 10 percent of construction projects are delivered in DB method (China Construction Industry Association, ‘‘CCIA’’ 2006), and most of the clients remained inexperienced with DB system. Moreover, many clients still keep traditional perspectives to the DB system. For example, the owners try to transfer most of the traditional risks such as the design errors to the DB contractors while leaving very limited room for the design-builder’s design input. This will give rise to interest conflicts and misunderstandings between each other. Therefore, this study aims to identify the key competences that the client should possess to ensure the success of DB projects in the construction market of the PRC. A literature review on the required competences of DB clients was undertaken. Then face-to-face interviews with experts in the construction industry were conducted to solicit the views from the practitioners. Finally, a two-round Delphi questionnaire survey was carried out with a group of panel experts to identify the key competences of DB clients. It is expected that this study will provide clients with measures to evaluate their DB competences and further promote their understanding of DB system in the PRC construction market. 2

LITERATURE REVIEW

Clients play important roles in contributing to the success of DB project. Several major studies have identified a list of the required competences of DB

clients. Songer & Molenaar (1997) identified 15 primary project characteristics that affect the success of public-sector DB projects. Among the 15 characteristics, the owners are required to (1) have a precise understanding of the project scope; (2) have the ability to precise define the project scope; (3) have adequate staff dedicated to the project. Besides these required competences, the owner should prefer to shift some of the traditional risk to the contractor and be willing to give up a large amount of the design input after design-builder selection. Molenaar & Songer (1998) set up a selection model for public sector DB projects. Among all the prediction variables contributing to the success of DB projects, the requirements and capabilities of owner include: (1) owner/agency experiences, and (2) owner/agency staffing. The owners are required to have DB related experiences and adequate staff to answer design and construction related questions. The owners must also convey the parameters of the project to the designbuilder in the form of a detailed contract document or a simple request for proposal. Chan (2001) conducted multiple regression analysis and found project team’s commitment, client’s competence, and contractor’s competences were important to bring about successful project outcome. The client’s competencies include: (1) a good capability of managing DB project; (2) a precise understanding of the DB project scope before it was submitted to the contractor; (3) clear articulation of end-user’s needs. Ling & Liu (2004) used artificial neural network (ANN) technique to construct the models to predict DB project performance. To ensure project success, client and his consultants should have (1) adequate staffing level to attend to contractor, (2) experience with similar construction projects, and (3) experience of DB projects. It is also recommended that the owners and their consultants well manage the projectrelated factors such as tender selection criteria, project scope, and form of contract to have higher chance of success. To the client’s required competences, many studies focus on the abilities to clearly articulate the project scope, owner’s requirements or objectives (Mo & Ng 1997; Songer & Molenaar 1997, 1998; Leung 1999; Pearson & Skues 1999; Chan et al. 2001), have the ability to manage design changes (Deakin 1999; Pearson & Skues 1999); have sufficient staff to coordinate with the other participants (Songer & Molenaar 1997, Ling & Liu 2004), and have similar DB experience in the past (Mo & Ng 1997, Molenaar & Songer 1998; Ling & Liu 2004). 3

RESEARCH METHODS

The research methods employed in this paper included: (1) structured face-to-face interviews; and (2) Delphi

questionnaire survey. The structured interviews were conducted to identify a list of required competences of DB clients in the PRC construction market, and subsequently two rounds of Delphi questionnaire survey were undertaken to assess the appropriateness of the proposed competences by rating them against their level of importance based on a ten-point Likert scales. The face-to-face interview offers the possibility of dispelling ambiguity because the interviewer will be next to the respondent as the questions are being answered. Another main advantage of using face-to-face interviews lies in the quality of the data obtained. Data and valuable experience often rely on the minds, attitudes, feelings or reactions of the respondents. Considering the fact that most of the DB contractors and clients remain inexperienced with DB system in the PRC, mail survey response may suffer from the depth of coverage of this topic in response to an open-ended question. Structured face-to-face interviews with five field experts were therefore conducted to solicit potential competences of DB clients in the light of specific conditions of the PRC construction market. A total of six competences of DB clients were finally formulated and consolidated for further analysis. A Delphi questionnaire survey was then conducted following the face-to-face interviews. The Delphi method is designed to extract the maximum amount of unbiased information from a panel of experts (Chan et al. 2001). Even if these collective judgments of experts are made up of subjective opinions, it is considered to be more reliable than individual statements, thus, more objective in its outcomes (Masini 1993). The Delphi method typically involves the selection of suitable experts, development of appropriate questions to be put to them and analysis of their answers (Cahanis 2001; Outhred 2001). The original Delphi procedures have three features: (1) anonymous response; (2) iteration and controlled feedback; and (3) statistical group responses (Adnan & Morledge 2003). The features are designed to minimize biasing affects of dominant individuals, irrelevant communications, and group pressure toward conformity. The Delphi method used in this research was composed of two rounds with 20 experts. In round 1 of the Delphi questionnaire, the respondents were asked to provide ratings against the levels of importance on each of proposed competences of DB clients, based on a ten-point Likert scale. In round 2 of the Delphi questionnaire, respondents were asked to reconsider the ratings of each competence in the light of the consolidated results from round 1. The number of rounds varies between two and seven (Rowe and Wright, 1999; Adnan and Morledge, 2003). Too many rounds would waste panel members’ time, and stopping the study too soon could yield meaningless results (Schmidt, 1997). The majority of Delphi studies have used between 15–20 respondents (Ludwig, 2001).

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4

ANALYSIS OF STRUCTURED INTERVIEWS

5.2

In the first round of the Delphi questionnaire survey, the panel experts were requested to assess the importance of each of the six short-listed competences of DB clients. A 10-point likert scale was used. Although the 1–10 ordinal scale is not as frequently used as the 1–7 or 1–5 system in Delphi research, it is much more familiar to the Chinese experts. If a score is lower than 6 point, it is commonly regarded as failing to pass the threshold of importance evaluation. Therefore, in this research, a mean score of 6 becomes a cut-off point and only the criteria with mean score of 6.0 point or above will be re-evaluated in the next round. Finally, 17 experts completed the questionnaire in late April 2008. A statistical analysis was performed on the 17 replied questionnaires; the results are shown in Table 2. Meanwhile, in order to measure the degree of agreement between the panel members on the ordered list by mean rankings, the Kendall’s Coefficient of Concordance (W ) was calculated with the aid of the SPSS software. According to the level of significance (shown in Table 2), which is less than 0.05, the null hypothesis that the respondent’s ratings within the group are unrelated to each other would have to be rejected.

The five interviewees are leading industrial practitioners in the DB fields. All of them have extensive hands-on experience in the DB market and hold senior positions in their organizations. Before conducting the Delphi survey, the interview results were sent back to all the interviewees for their verification. Finally, six key competences of DB clients in the PRC for further study were identified which include: (1) ability to develop clear project scope and objectives, (2) experience with similar DB projects, (3) sufficient staff or consulting team devoting to the DB projects, (4) ability in contract management, (5) ability to coordinate effectively with the design-builder, and (6) financial ability to provide sustained capital supply for the DB project.

5

TWO ROUNDS OF DELPHI SURVEY

5.1

Selection of expert panel

One of the most important considerations when carrying out Delphi study is the identification and selection of potential members to constitute the panel of experts (Ludwing 1997; Stone & Busby 1996). The selection of members or panelists is important because the validity of the study is directly related to this selection process. In this Delphi survey, the researchers attempted to identify panelists who meet all the following selection criteria:

5.3

Finally, 20 experts who meet the selection requirements agreed to participate in the Delphi survey. Most of the experts have sufficient experience and expertise in DB projects and hold senior positions in their organizations. Table 1 depicts the frequency of the respondent’s number of years working in the construction industry and in DB field.

Table 2.

Table 1. Respondent classification by years in construction industry and DB field. In construction field

In DB field

0−5 5−10 10−20 20+ Average

5% 30% 30% 35% 15

15% 50% 30% 5% 9

Round 2 Delphi survey: Re-assessing the ratings

In round 2 Delphi survey, the experts were asked to re-assess their ratings in the light of the consolidated results obtained in round 1. The round 2 Delphi questionnaire was distributed to the same group of panel experts by email in late April 2006. Finally, 16 experts completed the questionnaire in late May 2008. Most experts had reconsidered their ratings provided in the previous round and had made adjustments to their ratings. However, Table 3 shows that the rankings of all competences remain unchanged when compared with the consolidated results in Round 1. The Kendall’s Coefficient of Concordance (W ) for the

1. Having sufficient working experience or knowledge in the DB field, 2. Working in relevant organizations in the construction industry, 3. Holding senior positions in their organizations.

Years

Round 1 Delphi questionnaire: Ratings obtained from the experts

Results of the round 1 questionnaire survey.

The required competences of DB clients

Mean

Rank

Clear project scope and objectives Financial capacity for the projects Capacity in contract management Sufficient staff or consulting team Effective coordination with contractor Experience with similar DB projects

8.88 8.41 8.06 7.88 7.70 7.41

1 2 3 4 5 6

* Number (n) = 17. Kendall’s Coefficient of Concordance (W ) = 0.144. Level of significance = 0.032.

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Table 3.

achieving the intended procurement objectives. During the process of DB contract management, clients should carefully negotiate the terms and conditions of contracts with the DB contractor and monitor the performance of design-builder to esnure the contracted projects are provided in accordance with the specifiation and terms of the contract. In addtion, clients should also document and agree any changes that may arise during its implementation or execuation of cotnract. Although the adminstrative burden will be greatly reduced in the DB projects, clients should possess the ability to effectvely manage the DB contract to achieve the intended project objetives.

Result of round-2 questionnaire survey.

The required competence of DB clients

Mean

Rank

Clear project scope and objectives Financial capacity for the projects Capacity in contract management Adequate staff or consulting team Effective coordination with contractor Experience with similar DB projects

8.97 8.41 8.16 8.03 7.72 7.50

1 2 3 4 5 6

* Number (n) = 16. Kendall’s Coefficient of Concordance (W ) = 0.243, Level of significance = 0.002.

6.4

rankings of these variables is also provided in Table 3. The increased value of Kendall’s Coefficient of Concordance means that the agreement among the panel experts has improved.

6 6.1

DISCUSSION OF KEY COMPETENCES Clear project scope and objectives

The competence to develop clearly articulated project scope, objectives and requirements in the brief is regarded as the mostly required competence of DB clients. Quatman (2003) believes that a project’s success depends on the owner’s abilities and attitudes. If the Owner is very clear about the project’s goals, scope, and expected outcome, then the DB system will work to the owner’s benefit. Client’s objectives/requirements need to be established in advance of the procurement selection; otherwise it can be very costly if the information provided by the owner to the contractor at the outset of the design build process is erroneous (Mogibel 1999). 6.2

Financial capability

Clients play important roles in contributing to the success of DB projects. Besides having a clear definition of DB projects, clients in the PRC are particularly required to have a strong financial capacity for the project. In China, the DB system is usually applied in large and complex public projects, which require large capital scale. When clients transfer most of the risk to the contractors, they are required to provide sustained capital supply for the DB projects. Otherwise the contractor will demand higher contract price to compensate the extra risk or even decline the opportunity to bid. 6.3

Contract management ability

To clients, managing the DB contract and monitoring the performance of the DB contractor is crucial to

Sufficient staff

In DB projects, clients should have sufficient staff to prepare specifications and project definition for the bidding job. After the DB contractor is selected, clients should also have adequate staff to answer the design-related and construction-related questions. If clients do not have in-house staff for the DB projects, out-source adviser or design consultant should be employed. Otherwise, with the same firm designing and building the project, there may not be an independent party providing the necessary service to protect the owner, in particular in the immature DB market of the PRC. 6.5

Effective coordination ability

Effective communication between client and contractor is critical to the success of the design-build project (Ng & Aminah 2006). The design-build system provides clients and contractors an opportunity to interact more often and more directly than the traditional contract. To clients, it is especially important to have effective coordination with contractors to avoid misunderstanding or conflicts occurred during the process stage of the project. In DB projects, the design documents are often preliminary at the early stage and usually change over the course of the project. In addition, a lot of decisions will be made during the execution of DB projects. This requires close communication between clients and contractors. Developing and maintaining good relationships will bring additional benefits for both parties. 6.6

DB related experience

Many studies indicated that DB projects would be more likely to be successful if the clients have the similar design-build experience (Mo & Ng 1997; Molenaar & Songer 1998; Ling & Liu 2004). Even though most of the clients remain inexperienced with the DB system in the PRC, it will be much easier to implement the DB system if the clients have sufficient design ability and general construction experience in

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the past. To the inexpericed DB clients in the PRC construction market, selecitng the DB delivery system does not mean that they can simply leave all the project and responsibilty to the DB contrator. They should possess the design and/or construction experiences to ensure the smooth delivery of DB projects. Otherwise, experienced design consultants or advisers should be employed at the initial stage of the DB projects. 7

CONCLUSIONS

The DB system is developing rapidly in the construction market of the PRC. Considering that many clients are not quite familiar with this system, it is necessary to evaluate their competences before taking up the DB business; otherwise, it will increase the risk of project failure and give rise to conflicts of interest between the participants. The research findings indicate that the key competences the DB clients in the PRC should possess are: (1) clear articulation of project scope and objectives; (2) financial capacity for the projects; (3) capacity in contract management; (4) sufficient staff or consulting team; (5) effective coordination with contractor and (6) experience with design-build related projects. In identifying the key competences of DB clients, the Delphi method serves as a self-validating mechanism and provides a valuable framework for tapping expert knowledge on this field. It is worth noting that Delphi technique cannot fully eliminate the subjectivity of evaluation, however the number of panel experts in this study is considered adequate to provide reliable results. Furthermore, since the key competences of clients were identified in the PRC, further research should be conducted in other geographical locations to find out their similarities and differences for international comparisons. REFERENCES Adnan, H. & Morledge, R. 2003. Application of Delphi method on critical success factors in joint venture projects in the Malaysian construction industry, CITC-II Conference, Hong Kong, 10–12 December. Hong Kong. Chan, A.P.C, Ho, D.C.K. & Tam, C.M. 2001. Design and build project success factors: Multivariate analysis. Journal of construction engineering and management, 127(2): 93–100. Cheng, R.T.L. 1995. Design and build-contractor’s role. Proceeding on design and build projects, 232–241. Deakin, P.1999. Client’s local experience on design and build projects. Seminar Proceedings on Design and Build Procurement System, 11–15.

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Hemlin, D.1999. Contractor’s local experience on design and build projects. Seminar Proceedings on Design and Build Procurement System, 17–25. Johnson, D. & King, M.1988. Basic Forecasting Techniques. Butterworths: London. Jones, T. 1980. Options for the future: a comparative analysis of policy oriented forecasts. New York. Leung, K.S. 1999. Characteristics of design and build projects. Seminar Proceedings on Design and Build Procurement System,1–10. Ling, F.Y.Y., Chan, S.L., Chong, E. & Ee, L.P. 2004. Predicting Performance of Design-Build and Design-Bid-Build Projects. Journal of Construction Engineering and Management, ASCE, 130 (1): 75–83. Ling, F.Y.Y. & Liu, M. 2004. Using neural network to predict performance of design-build projects in Singapore. Building and Environment 39: 1263–1274. Linstone, H.A. & Turoff, M. 1975. The Delphi method: Techniques and applications. Reading, MA: Addison-Wesley. Ludwig, B. 1997, 2001. Predicting the future: Have you considered using Delphi methodology? Journal of Extension, 35(5). Masini, E. 1993. Why Future Studies? Grey Seal: London. Mo, J.K. & Ng, L.Y. 1997. Design and build procurement method in Hong Kong-An overview. Proc., CIBW92 Procurement-A Key to Innovation, Procurement Sys. Symp., 453–462. Molenaar, K.R. & Songer, A.D. 1998. Model for public sector design-build project selection. J. Constr. Engrg. Mgmt. ASCE, 124(6): 467–479. Moore, C.M. 1987. Group technique for idea building, SAGA Publications: California. Ng, W.S. & Aminah, M.Y. 2006. The success factors of design and build procurement method: a literature visit. Proceedings of the 6th Asia-Pacific Structural Engineering and Construction Conference (APSEC2006), 5–6 Sept, Kuala Lumpur, Malaysia. Pearson, M. & Skues, D. 1999. Control of projects implemented through design and build contracts. Seminar Proceedings on Design and Build Procurement System, 49–60. Quatman & Dhar 2003. The Architect’s Guide to DesignBuild Services. John Willey and sons, Inc. Hoboken: New Jersey. Rowe, G. & Wright G. 1999. The Delphi Technique as a Forecasting Tool: Issues and Analysis. International of forecasting15(4): 353–375. Schmidt, R.C. 1997. Managing Delphi survey Using Nonparametric Statistical Techniques. Decision Science 28(3): 763–774. Songer, A.D. & Molenaar, K.R. 1997. Project characteristics for successful public-sector design-build. Journal of Construction Engineering and Management, ASCE, 123(1): 34–40. Stone, F.L. & Busby, D.M. 1996. The Delphi Research Methods in family therapy. Guildford: New York.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

PPP-risk identification and allocation model: The crucial success factor for PPPs T. Pohle & G. Girmscheid Institute for Construction Engineering and Management, ETH Zurich, Switzerland

ABSTRACT: In practice, risk allocation among the participants in a PPP is a purely intuitive, habitative and opportunistic process driven by the negotiating strength of the partners. There is a lack of clear and structured decision-making criteria and methods for ensuring the cost-minimizing allocation of the risks of a PPP. The Institute for Construction Engineering and Management at ETH Zurich is developing a multi-dimensional risk identification and allocation model (PPP-RIA-Model) that incorporates both the professional competences and means available to both partners to influence the occurrence and to minimize the extent of risks, and the financial risk-bearing capacity of the risk taker when allocating the risks. This PPP-RIA-Model can be used to ensure that the risk allocation acts as an entrepreneurial incentive to stimulate improvements in efficiency on the part of the private partner while at the same time assuring that the budgetary considerations for protecting the site qualities of the public sector are taken into consideration. 1

INTRODUCTION

Around the globe, public private partnerships (PPPs) have emerged as a very widespread and successful alternative for performing public tasks. A PPP aims to release synergies by incorporating the specific expertise and economic competence of the private partner in collaboration with the public partner when performing public sector tasks. These synergies are released by ensuring the best possible allocation of risks that is commensurate with the performance capabilities of the partners. Practice and research both agree that it is the allocation of risks among the partners, in particular, that constitutes the critical success factor for the long-term economic efficiency of a PPP. In practice, however, a standardized and systematic approach to risk allocation is not recognizable. Intuitive, habitative and opportunistic aspects relating to the negotiating strength of the partners usually determine the allocation of the risks. There is a lack of clear decisionmaking criteria and methods for ensuring the cost minimizing allocation of the risks of a PPP, which would enable the public sector to optimally allocate the risks, bearing in mind the risk coverage capacity of the private partner. This paper presents a risk identification and allocation model (RIA-Model) for the performance of tasks by a PPP which is being developed by the Institute for Construction Engineering and Management at ETH Zurich in collaboration with the Swiss cities and towns of Zurich, Lucerne and Uster for the public sector. The model aims to allow municipalities

745

to identify, evaluate and assess the long-term risks of this partnership and to independently systematically optimize the allocation of the risks using clear decision-making criteria. 2

STATE OF RESEARCH

In addition to the pertinent standard literature on risk management (Girmscheid & Busch 2008a; Girmscheid & Busch 2008b; Schierenbeck 2003) numerous publications have meanwhile been written in German and other languages that focus specifically on the issue of risk management in a PPP. Pertinent publications include: (Akintoye et al. 2003; Elbing 2006; Grimsey & Lewis 2002; HM-Treasury 2004; Merna & Lamb 2003). Although most publications confirm that optimal risk analysis and allocation plays a central role in securing the economic success of the project (Akintoye et al. 2003; European-Commission 2002; Grimsey & Lewis 2002), they do not provide the tools for actually putting the aim of minimizing costs into practice. In research and practice, ‘‘optimal risk allocation’’ is generally understood as being a state where neither of the partners bears risks that could be more efficiently shouldered by the other partner, for example, because the latter has better means of influencing that risk (PPP Task Force NRW 2007). This reflects the general fundamental principle of risk allocation whereby the party that can manage the risks more cost effectively should also bear them (Akintoye et al. 2003; Boussabaine 2007; European-Commission 2002). Apart from these

general fundamental principles of risk allocation, literature does not suggest any further possible approaches to optimizing the same. Neither decisionmaking aids for risk allocation are offered, nor any specific criteria that could be applied to such allocation. The competence of the partners is the only starting point for optimizing risk allocation; both practice and research propagate the allocation of risks in line with this competence. In addition to general fundamental principles of risk allocation, Boussabaine is the only author to state that, in practice, the risk taker does not always have sufficient financial capacity to actually bear the risks allocated to it (Boussabaine 2007). As such, he is the only person, alongside Girmscheid & Busch (Girmscheid & Busch 2008a) to spotlight a further dimension that needs to taken into consideration and incorporated as an input parameter into a model for risk allocation, in addition to the professional competence of the two partners. Generally speaking, there is a lack of clear allocation criteria in literature that could be used as the basis for a cost-minimizing allocation of risks and thus secure the success of the long-term partnership for both parties. As part of the ‘‘Risk allocation in PPP maintenance of municipal street networks’’ research project at ETH Zurich, a concept is being developed to allocate risks in line with the professional competence and financial capacities of the partners in respect of their ability to influence both the occurrence and the impact of the risks, and taking account of the financial risk coverage capacity of the private partner. The aim is to provide municipalities with a tool for the cost-minimizing allocation of risks based on clear allocation criteria and thus for ensuring the optimal allocation of the risks among the partners.

conceptualize the categories for risk identification. The presented PPP-RIA-Model represents an actional generically-deductive structure as a target-means relationship using the constructivist research paradigm (Girmscheid 2007a; Glasersfeld 1998; Piaget 1973). The scientific quality is achieved by triangulation (Yin 2003) due to: – Viability of the generic-deductive model – Validation through a theoretical framework – Reliabilitation through testing the intended impact (target-means-relation). For the purpose of validation the PPP-RIA-Model is theoretically-deductively structured using the principles of system theory (Bertalanffy 1973; Boulding 1956) and risk load capacity theory (Girmscheid & Busch 2008b). Reliabilitation will be achieved using realization tests to check if alternative target relations exist under equal means. According to Yin (2003) the above sketched triangulation concept fulfills the scientific quality requirements. 4

PPP RISK IDENTIFICATION AND ALLOCATION MODEL (RIA-MODEL)

The multi-dimensional PPP-RIA-Model is comprised of four individual modules (Fig. 1) that can be used to allocate risks, once they have been identified and analyzed, based on the competences and risk coverage capacity of the partners.

RESEARCH METHODOLOGY

Constructivist research approach

According to Dilthey and Heidegger (Dilthey 1900; Heidegger 1938), philosophical hermeneutics—the anthropological basis for understanding the sociotechnical structure of the world around us—forms the scientific-philosophical platform for target-means relationships. The hermeneutic science program is underlaid with the interpretativist research paradigm (Weber 1980) and the constructivist research paradigm (Glasersfeld 1998) for the purpose of obtaining knowledge. Construction management forms part of this sociotechnical world structured by man. As such, scientific challenges in this field can be addressed using these research paradigms. The epistomology on which to base the risk identification and allocation model builds on the constructivist research program defined by Glasersfeld. An empirical study based on qualitative social research methods (Mayring 2002) was used to

• • • •

PPP risk identification PPP risk analysis PPP risk evaluation PPP risk categorization

PPP risk load dimension • risk cause related criteria • risk impact criteria minimum of risk costs

PPP risk coverage dimension based on : cash flow, profit and loss balance sheet, future prognoses, fluctuation margins (scenario from normal to extreme)

Test of PPP risk bearing -ability based on the determined risk load capacity & the primary risk control of the PPP -partners • total risk costs and analysis of the alternative risk distribution variants • multidimensional value benefit analysis

PPP-Risk Identification and Allocation Model (PPP-RIA-Model)

PPP risk identification and assessment

3

Figure 1. Risk identification and allocation model (PPP-RIA).

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4.1

Module 1: PPP risk identification and assessment

Risk identification for a specific project and/or area of responsibility is generally based on a generically logical structure that classifies and structures the risks by categories. This risk categorization must be performed for the specific project and/or area of responsibility and must contain the risk clusters specific to that same project and/or task in order to enable systematic and structured risk analysis. For purposes of risk categorization, an empirical survey was conducted to identify the typical risk clusters in municipal street maintenance. The risks were categorized in risk causes, risk groups and risk types. (Fig. 2). In future, this tool will enable municipalities to independently identify the risks that are specific to both the municipality and the respective situation. Based on the risk categorization, a risk checklist can be used to systematically identify the risks that are specific to a municipality in future. The risks that are structured with the aid of checklists (Fig. 2) can then be analyzed and evaluated. Risk assessment involves predicting the likelihood of occurrence (O) and impact (I) of the identified risks, and serves as an indicator for judging the degree to which achievement of the project objectives is jeopardized. Ra,i = Oa,i ∗ Ia,i

(1)

where Ra,i = anticipated risk costs; Oa,i = anticipated likelihood of occurrence; Ia,i = anticipated impact (damage) of the risk. Risk cluster (causes of risks)

Risk groups changes in law

Political political

changes of budgets changes in standards

contractural changes

Risk types / Single risks • • • • • • • • • • •

Contractual

partner related changes of requirements natural anthropogenic/ man made

operating Operating performance/quality

management

• • • • • • • • • • • • • • • • • • •

changes in legislative changes in government policy changes in taxation rescheduling coverage of budgets construction standard operating standard technical standard changes to the general project conditions contract formulation unforeseen technical problems or environmental impacts bankruptcy failure of the partner to perform/ provide the requisite quality changes in users needs demographic/social changes storm floods extreme winters demonstration festivals, street parades sports events opportunistic behaviour vandalism useability restrictions availability residual value bad planning contract formulation controlling restriction

The purpose of the subsequent risk classification is to sort the risks according to the urgency needed to address them. The evaluation of risks in terms of the anticipated risk costs can therefore mean that risks are weighted differently in a project. Various methods for visualizing the prioritized treatment of risks can be used in the classification process; the portfolio method and ABC analysis (Girmscheid and Busch 2008a) are both ideally suited and very widespread. Generally, only A risks (and B risks) are examined in more detail in the further risk assessment process. 4.2

Module 2: Risk load dimension—Assessment of the risks in terms of their influenceability and minimization of their impacts (‘‘Initial risk allocation by the PPP project company’’)

Risk allocation aims to generate the lowest possible risk costs in line with the economic minimum principle (Ri,opt ). This is especially the case when either the likelihood of occurrence (Oi,opt ) or the impact (Ii,opt ) or the likelihood of occurrence and impact of the respective risks can be minimized by allocating the risks to the partners in line with their ability/means to minimize these two parameters. In order to ensure cost minimization in the allocation of the risks, this module examines and analyzes the identified risks more closely to see whether they can be allocated more cheaply to either of the partners. The risks are examined to determine whether one of the two partners has the means to influence the cause of the risk and therefore to optimize the occurrence of the same (Oi,opt ) or—if it cannot influence the occurrence of the risk—at least has the means to limit the impact/consequences of the risk (Ii,opt ) (Fig. 3). The necessary measures and related costs of the same are also examined (Fig. 4). The following questions need to be answered to judge which partner can bear the costs more cheaply:

Identified risk ri

Definition of publics sector’s ability to influence Oi and I i

Definition of publics sector’s ability PR to influence Oi and I i

Definition of any risk costs associated with achieving Oi PS and I i PS

Definition of any risk costs associated with achieving Oi PSand I i PS

PS i

PS i

( OiPR ; I i PR )

(O ; I )

Calculation of public sector’s risk costs

PS PS PS Ri = ( Oi * Ii ) + costs of measures

Calculation of private partner’s risk costs

PR

Ri

= ( Oi PR* Ii PR ) + costs of measures

Comparison of both partner’s risk costs for each identified risk

Figure 2. Categorization of risks—PPP risks in the operational phase, structured in causes of risks, risk groups and risk types.

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Ri

PR Min ( RPS i ; Ri )

Figure 3.

Flow chart on the minimization of risk costs Ri .

Influenceability of the individual risk cost parameters

Risk costs (per ann.) 1 Oi

Risk types

Ii

Budget redeployment

Ri

2

Influenceability of likelihood of occurance Oi Measures (how?) Long-term

financial planning

Party (by whom?) PS

Technical standards

Incomplete/incorrect performance specifications Construction risks

Quality control/ Involvement of experts Continual risk, quality and process management

2

Influenceability of impact I i

Costs of measures

Private Party partner (by whom?) (PR)

Measures (how?) Flexible reallocation of capacitied Analysis of trends and developements in the construction history; on-going trainig

Public sector (PS)

(R Private partner (PR)

Public sector (PS)

80 % PR

bzw. R

80 % PR

X

Structural safeguards, reserve accounts

30 % PR

X

R i,min

R i,min

R i,min

X

Approval/Ban by the public sector

80 % PS

Details of size and location

20 % PS

X

Processions, festivals

Approval/Ban by the public sector

100 % PS

Details of size and location, Expenses

20 % PS

X Pr

Total Ri :

)

R i,min

X

Continual risk, quality and process management

PS

Public sector PS (R )

R i,min

X

30 % PR

Demonstrations

O: Likelihood of occurrence I: Impact R: Risk costs

PR

Private partner PR (R )

X

PR

100% PS

Force majeure risks (one-off risks, such as storms, flooding)

Risk costs per risk bearer

Risk bearer

R i,min R i,min öff

Total R resp. R :

1: Likelihood of occurrence Oi in number of incidents per year and/or one-off incidents, in percent 2: Influenceability as a percentage PR: Private partne PS: Public sector r

Figure 4. Abstract of risk allocation matrix—Minimization of risk costs in terms of their influenceability and minimization of their impacts by the partners.

– Who can influence the occurrence of the risks (likelihood of occurrence)? – Who can influence the impact of the risks to minimize the consequences? – Which minimization measures need to be taken prior to occurrence and/or impact of the risks, and what costs will this incur (costs of measures)? The proportionate risk costs of both parties are produced from the individual aggregates of the risks that each party can bear more cheaply given its means of influencing or minimizing the impact of the same. A comparison of the risk costs between the public sector and the private partner must incorporate the implemented to minimize the likelihood of occurrence and/or the impacts (Fig. 4). 

RPS opt =



   PS PS Oi,opt Oj ∗ Ij,opt ∗ Ii +

i

+

j





PS PS Ok,opt + ∗ Ik,opt

k



PR Ropt =



Mi,j,k

(2)

i.j.k



   PR OPR Oj ∗ Ij,opt i,opt ∗ Ii +

i

+



j

  PR PR Ok,opt + ∗ Ik,opt Mi,j,k

k

(3)

i.j.k

where Ropt = optimized risk costs (of the private or public partner); Oi,opt = optimized likelihood of occurrence of the risk Ri ; Ii,opt = optimized impact (damage) of the risk Ri ; Mi = costs of measures.

The aggregate of the risk costs of both partners produce the total risk costs of the PPP project. 

RPPP =



RTotal opt =



RPS opt +



RPR opt

 (4)

where RPPP = optimized total risk costs of the PPP project; RPS opt = optimized risk costs of the public sector; RPR opt = optimized risk costs of the private partner. 4.3

Module 3: Risk coverage dimension— Assessment of the private partner’s (PPP project company’s) ability to cover the risk

The purpose of module 3 is to analyze the capacity of the private partner to cover the risk. The risk coverage capacity of the private partner is assessed, in particular, under consideration of proven concepts applied by banks when evaluating the risk-bearing ability of potential borrowers for loan approval purposes. These concepts are based on a corporate rating that assesses the creditworthiness of a company. According to Basle II requirements (Gleissner & Füser 2003), corporate credit ratings are based on two different factors: an assessment of the financial capacity (quantitative evaluation criteria) and an assessment of a company’s sustainability (qualitative evaluation criteria) (Gleissner & Füser 2003; Standard & Poor’s 2006). This paper only addresses PPPs that are organized as PPP project companies in which the private partner invests its equity. From a balance sheet perspective, such PPP project companies are classified as standalone organizations with limited equity. It does not examine any PPPs involving forfeiting and waivers of objections or any other project forms.

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In the case of a PPP project, the capital supplied by the project company is therefore the only collateral securing the risk burden. As such, the creditworthiness of a PPP contractual partner (risk coverage capacity assessment of the private partner) is assessed at two different levels. Financial capacity is assessed on the basis of the equity ratio, and the sustainability of the PPP is assessed on the basis of its long-term cash flow generation. The financial capacity and sustainability of the private partner (PPP Project company) are thereby determined in two phases in line with the different degrees of risk assumption. 4.3.1 Assessment of sustainability—normal and stress coverage capacity Sustainability is measured in terms of the level of potential for generating positive cash flow surpluses. Normal coverage capacity is derived from the annual cash flow and serves to cover normal load risks that occur on an ongoing basis (minor risks). PPP RCov Norm = CFa,Surplus

(5)

PPP where RCov Norm = normal coverage capacity; CFa,Surplus = annual cash flow surplus of the PPP project. Stress coverage capacity is derived from the accumulated cash flow surplus over several years and serves to cover stress load risks (moderate risks).

RCov Stress =

n 

ppp

CFa,Surplus

(6)

1

 PPP where RCov CFa,Surplus Stress = stress coverage capacity; = cash flow surplus of the PPP project accumulated over several years. 4.3.2 Assessment of financial capacity—crash coverage capacity Financial capacity is measured in terms of the equity capacity of the project company. This equity portion provides financial security against risks that cannot be covered from the accumulated cash flow surplus. The cash flow surplus increases consistently from year 1 through to year n. The increased occurrence of major risks is assumed to only happen over the course of the contract as the infrastructure ages. Crash coverage capacity is derived from the project company’s equity in the project and serves to cover crash load risks (major risks).

Generally, this equity can only be liquidated to a limited extent. For example, this may be the case if the private partner withdraws as operator. It is therefore crucial for a contractual agreement to stipulate that some of the annual cash flow is accumulated as a surplus on a shared blocked account in order to cover risk costs, especially stress risks (major operative risks). 4.3.3 Risk coverage capacity The risk coverage principle is illustrated in Figure 5. The risk coverage capacity is therefore calculated as follows: ⎧ Cov ⎫ ⎪ ⎨ RNorm ⎪ ⎬ ppp (8) = f (CFSurplus ; EQPC ) (RCov ) = RCov Stress ⎪ ⎩ Cov ⎪ ⎭ RCrash

 where RCov = risk coverage capacity vector; PPP = cash flow surplus from the PPP project; CFSurplus PC EQ = equity of the PPP. 4.4

Module 4—test of PPP risk-bearing ability

The multi-dimensional risk allocation model is based on the following three decision-making dimensions (base variables of risk allocation as defined by Girmscheid and Busch (2008a): – Players’ ability to influence the risks – Players’ ability to minimize the impacts of the risks – Risk coverage capacity of the risk taker and combines these into a holistic whole. This model tests whether the normal, stress and crash coverage capacities defined in module 2—‘‘Initial risk allocation of the project company’’ can be adopted (Fig. 6). If this is the case, the sustainably optimized allocation of risks is completed and the risk-bearing ability assured. If the risk coverage capacity of the project 95%

100%

60% 60% 40% 30%

Accumulated surplus

Annual surplus

CF

∑ CF

PC,allo

Project company’s equity

PC,allo

Not covered

PC

EQ

Normal risk Stress risk

PC RCov Crash = EQ

Crash risk

(7)

PC where RCov = Crash = crash coverage capacity; EQ equity of the PPP project company.

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Total project company risk

Figure 5. pany.

Risk coverage principle of the PPP project com-

Module 2: Risk load dimension

Module 3: Risk coverage capacity

Agg R Norm < RCov Norm

No

New Allocation

Agg R Stress < RCov Stress Agg R Crash < RCov Crash

Yes

No

Analysis of economic efficiency

No Go?

No

PPP < PSC?

Yes

Yes Self-performance

Implement PPP project

Figure 6. Module 3— Test of risk-bearing ability of the PPP project company.

company is lower, however, the risk allocation process must be repeated in a cybernetic process. If optimal risk allocation is not possible, an analysis of economic efficiency (Girmscheid et al. 2008) must then be performed to examine whether the ‘‘possible’’ risk allocation still renders the PPP alternative the cheaper solution compared with the PSC of self-performance and self-operation. 5

CONCLUSIONS

The new concepts for the provision of high quality PPP services in a long-term PPP commitment to partnership require clear analysis of the risks and risk costs. Long-term partnership commitment can only produce a win-win situation if the uncertainties surrounding risk are largely addressed right from the tender and bidding phase, and the risks are optimally allocated in line with the respective competences and influencing abilities of the partners in order for the public sector to generate the economic benefits of a PPP. Municipalities can use this risk identification and allocation model (PPP-RIA-Model) to identify, evaluate and assess the long-term risks and to systematically optimize the allocation of these risks. The tool for evaluating risk allocation provides objective support for the public sector in – realistically assessing the intended risk allocation, – formulating its requirements for the private partner, and – assessing the bidding company’s project vehicle to ensure sufficient risk coverage capacity. Optimal risk allocation using this PPP-RIA-Model enables maximum benefit to be gained from the possible synergies of a PPP thus securing the long-term

partnership and economic efficiency of a PPP for both public sector and private partner. REFERENCES Akintoye, A., Beck, M., and Hardcastle, C. (2003). PublicPrivate Partnerships—Managing risks and opportunities, Blackwell Science, Oxford. Bertalanffy, L.v. (1973). General system theory: foundations, development, applications, G. Braziller, New York. Boulding, K.E. (1956). ‘‘General Systems Theory-The Skeleton of Science.’’ Management Science, 2(3), 197–208. Boussabaine, A. (2007). Cost Planning of PFI and PPP Building Projects, Taylor & Francis, Abingdon, UK. Dilthey, W. (1900). Die Entstehung der Hermeneutik, Verlag von B.G. Teubner, Leipzig. Elbing, C. (2006). Risikomanagement f ür PPP-Projekte, Eul, Lohmann. European-Commission. (2002). Guidelines for successfull Public-Private-Partnerships, Brüssel. Girmscheid, G. (2007a). Forschungsmethodik in den Baubetriebswissenschaften, ETH Zürich, Zürich. Girmscheid, G., and Busch, T.A. (2008a). Projektrisikomanagement in der Bauwirtschaft, Bauwerk, Berlin. Girmscheid, G., and Busch, T.A. (2008b). Unternehmensrisikomanagement in der Bauwirtschaft, Bauwerk, Berlin. Girmscheid, G., Lindenmann, H.P., Dreyer, J. and Schiffmann, F. (2008). Forschungsbericht ASTRA 2003/07: Kommunale Strassennetze in der Schweiz: Formen neuer Public Private Partnership (PPP)—Kooperation für den Unterhalt, Bundesamt für Strassen ASTRA, Bern. Glasersfeld, E.v. (1998). Radikaler Konstruktivismus: Ideen, Ergebnisse, Probleme, Suhrkamp, Frankfurt a.M. Gleissner, W., and Füser, K. (2003). Leitfaden Rating Basel II, Vahlen, München. Grimsey, D., and Lewis, M.K. (2002). ‘‘Evaluating the risks of public private partnerships for infrastructure projects.’’ International Journal of Project Management, 20(2), 107–118. Heidegger, M. (1938). Holzwege, Klostermann, Frankfurt am Main. HM-Treasury. (2004). The Orange Book—Management of Risk—Principles and Concepts, London. Mayring, P. (2002). Einführung in die qualitative Sozialforschung, Beltz-Verl., Weinheim [u.a.]. Merna, A., and Lamb, D. (2003). Project Finance: The Guide to Value and Risk Management in PPP Projects, Euromoney Books, Oxford. Piaget, J. (1973). Erkenntnistheorie der Wissenschaften vom Menschen, Ullstein, Frankfurt a. M. PPP Task Force NRW. (2007). Anleitung zur Prüfung der Wirtschaftlichkeitsuntersuchung von PPP-Projekten im öffentlichen Hochbau. PPP-Initiative NRW. Schierenbeck, H. (2003). Risiko-Controlling und integrierte Rendite-/Risikosteuerung, Gabler, Wiesbaden. Standard & Poor’s. (2006). Corporate Ratings Criteria (2006), The McGraw-Hill Companies. Weber, M. (1980). Wirtschaft und Gesellschaft: Grundriss der verstehenden Soziologie, Mohr, Tübingen. Yin, R.K. (2003). Case study research: design and methods, Sage, Thousand Oaks, Calif. [u.a.].

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Privileges and attractions for private sector involvement in PPP projects A.P.C. Chan, P.T.I. Lam, D.W.M. Chan & E. Cheung Department of Building and Real Estate, The Hong Kong Polytechnic University, Hung Hom, Kowloon, Hong Kong Special Administrative Region, China

Y. Ke Department of Construction Management, Tsinghua University, Beijing, China

ABSTRACT: Public Private Partnership (PPP) projects often involve a costly and lengthy procurement process. Hence many of these projects tend to be very large public works to be worthwhile. Often the private sector has been known to contribute in terms of financial support, technical skills, innovation, technology advances, specialist knowledge and efficiency. For the public sector it is often advantageous to involve the private sector but whether the private sector is willing to participate is another question. This paper therefore looks at the privileges and attractions for the private sector to involve with PPP projects. Due to the rapid development of urban infrastructure in China, this country was believed to be of interest to many investors. Although the Hong Kong Special Administrative Region (referred to as Hong Kong from here onwards) is now part of China, it has often been known as the gateway to China for western investors due to the ‘‘one country, two systems’’ arrangement. 1

INTRODUCTION

Although Hong Kong is part of China, under the ‘‘one country, two systems’’ policy, the practice and experience of conducting PPP projects in these places are quite different. Hong Kong has been governed by the British for a long duration. And during this time the western practices of running projects proactively have been assimilated by the local government. Mainland China on the other hand has always adopted a more conservative Asian approach to procuring projects. One major similarity between the two administrative systems is that both have had a strong interest in procuring more public projects by the PPP model. Stepping into the 21st Century, the bottleneck effect of infrastructure shortage for the Chinese economy emerged and imposed budgetary pressure on the Mainland Chinese government. The investment in infrastructure development could not be completed by the Mainland Chinese government alone (Sachs et al., 2007) which provides a good business opportunity for the private investors. In Beijing alone, some of the recently implemented PPP projects include Metro Line 4 Project, Lugouqiao Sewage Treatment Plant Phase 1 Project, Gaoantun Waste-to-Energy Plant, National Stadium Project, and the Concession Project of natural gas in the East New District of YiZhuang Road (Beijing Municipal Commission of Development and Reform 2005).

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Hong Kong has secured a long history of launching PPP projects. The first and most famous PPP project in Hong Kong is the Cross Harbour Tunnel which was delivered by Build-Operate-Transfer model in the late sixties (Chan et al. 2007a). Although this project experienced immediate success, a few other less successful attempts suggested that this model was not easy to follow. Hence the government slowed down as there was never any desperate urge to adopt PPP anyway. In recent years, PPP has been popularly used worldwide. Apart from the obvious financial advantages of adopting PPP, other drivers of this relatively new approach were also observed. As such, the Hong Kong government has been increasingly more interested in pursuing public projects via PPP scheme. Recently, a number of massive public sector projects have already been confirmed that the PPP model would be used for their procurement. For example, the Shatin to Central rail link and the Kwun Tong rail extension. The new metro line will consist of nine stations. Construction will start in 2010 and the two phases of the line will be completed by 2015 and 2019 (Information Services Department 2008). The interest over PPP proves that it has its own attractiveness and privileges. The findings presented in this paper examine specifically what these are according to Mainland Chinese and Hong Kong respondents. This study is part of a research study looking at developing a best practice framework for PPPs in Hong Kong (Chan et al. 2007b).

2

were sent to 103 target respondents in Mainland China and 95 target respondents in Hong Kong. It was anticipated that some of these target respondents would have colleagues and personal connections that would be knowledgeable in the area of PPP to participate in this research study as well; hence some of the respondents were dispatched five blank copies of the survey form. A total of 53 completed questionnaires from Mainland China and 34 from Hong Kong were returned representing response rates of 52% and 36%, respectively. The higher response rate in Mainland China compared to Hong Kong was anticipated. There have not been that many PPP projects in Hong Kong hence the number of people involved in PPP projects would be less. Mainland China on the other hand has been involved in many more PPP projects recently in comparison with Hong Kong. Also, the population size in Mainland China is much higher than Hong Kong. Mainland China has a booming population size of 1.32 billion as recorded in March 2008 (China Population Development and Research Center 2008), and although Hong Kong is densely populated for a city of its size, its population is much smaller than Mainland China at only 6.96 million at the end of 2007 (Census and Statistics Department 2008).

PREVIOUS RESEARCH ON PRIVELEGES AND ATTRACTIONS OF PPP

The questionnaire template designed by Li (2003) was adopted for this study. Although the authors could have developed their own research questionnaire, there were several advantages foreseeable to adopt Li’s (2003) survey questionnaire rather than designing a new template. Firstly, the value of Li’s (2003) questionnaire has already been recognized by the industry at large. His publications as a result of the research findings derived from the questionnaire are evidence of its worthiness. Secondly, there would be no added advantage to reinvent the work that has previously done by other researchers. And thirdly by administering Li’s (2003) questionnaire again but in different administrative systems would be of interest for comparison purposes in the future. Therefore Li’s (2003) questionnaire was adopted for the survey as presented in this paper with prior permission obtained from the author Dr. Li Bing and his doctoral research supervisor, Prof. Akintola Akintoye who is currently the Head of the School of Built and Natural Environment, University of Central Lancashire, United Kingdom.

3

RESEARCH METHODOLOGY

4

An empirical questionnaire survey was undertaken in both Mainland China and Hong Kong from October 2007 to December 2007, to compare and contrast the attractions and privileges of PPP in these two similar and yet different administrative systems. In this study, the target survey respondents of the questionnaire included all industrial practitioners from the public, private and other sectors. These respondents were requested to rate their degree of agreement against each of the identified attractions and privileges according to a five-point Likert scale (1 = Least Important and 5 = Most Important). Target respondents were selected based on their direct hands-on involvement in PPP projects. Survey questionnaires

Table 1.

DISCUSSION OF SURVEY RESULTS

The attractions and privileges of PPP were assessed from different perspectives of the Mainland China and Hong Kong respondent groups. The means for each administrative system were calculated and ranked in descending order of importance as shown in Table 1. 4.1

Ranking of attractions and privileges of PPP

The mean values for the attractions and privileges as rated by Mainland Chinese respondents ranged from 3.40 to 4.02. For those rated by respondents from Hong Kong the mean values ranged from 3.36 to 3.82. This observation has reflected that the variations in their

Mean scores and rankings for the privileges and attractions of PPP. Mainland China and Hong Kong

a. Government sponsorship b. Government assistance in financing c. Government guarantee d. Tax exemption or reduction e. Incentive of new market penetration

Mainland China

Hong Kong

N

Mean

Rank

N

Mean

Rank

N

Mean

Rank

86 86 86 86 86

3.48 3.78 3.86 3.76 3.74

5 2 1 3 4

53 53 53 53 53

3.40 4.02 4.02 4.00 3.70

5 1 1 3 4

33 33 33 33 33

3.61 3.39 3.61 3.36 3.82

2 4 2 5 1

* N = Number of survey respondents.

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Table 2.

Results of Kendall’s concordance analysis for the privileges and attractions of PPP.

Number of survey respondents Kendall’s coefficient of concordance (W ) Chi-square value Degree of freedom (df) Asymptotic significance

Mainland China and Hong Kong

Mainland China

Hong Kong

86 0.026 9.039 4 0.06

53 0.077 16.413 4 0.003

33 0.041 5.446 4 0.245

responses are small, only 0.62 and 0.46 for Mainland China and Hong Kong respectively. Another observation which can be made from the mean calculation is that three out of the five attractions and privileges were rated slightly higher by respondents in Mainland China compared to those in Hong Kong and vice versa for the remaining ones. As such, it must be noted that the means were interpreted directly. The differences observed do not indicate that the attractions and privileges were statistically significant. Categorically speaking, those three attractions and privileges rated higher by respondents in Mainland China were attractions and privileges offered by the government including: 1. Government assistance in financing (joint first position); 2. Government guarantee (joint first position); and 3. Tax exemption or reduction. For those two attractions and privileges observed to be higher for Hong Kong respondents, they were not focused on a particular aspect: 1. Incentive of new market penetration; and 2. Government sponsorship. As the respondents were asked to rate the 5 attractions and privileges according to a Likert scale from 1 to 5 (1 = Least Important and 5 = Most Important), a value above ‘3’ would represent that the attraction and privilege is of importance. It was found that all were rated above the mean value of ‘3’ by Mainland Chinese and Hong Kong respondents. 4.2 Agreement of respondents within Mainland China and Hong Kong As shown in Table 2 the Kendall’s coefficient of concordance (W) for the rankings of attractions and privileges was 0.026, 0.077 and 0.041 for ‘Mainland China and Hong Kong’, ‘Mainland China’ and ‘Hong Kong’ respectively. The computed W for Mainland China was significant with p = 0.003 showing consistency in the responses. Whereas, for the Hong Kong respondents p = 0.245 showing that the responses were not significantly consistent.

753

Table 3. Results of Spearman rank correlation test between respondents from Mainland China and Hong Kong for the privileges and attractions of PPP. Comparison

rs

Significance

Ranking of Mainland China ranking vs Ranking of Hong Kong

−0.395

0.511

4.3

Agreement of respondents between Mainland China and Hong Kong

The next stage of the analysis was to test whether there is any substantially similar agreement amongst the respondents between the two places which is determined by the Spearman rank correlation coefficient (rs ) again using the SPSS statistical package. The correlation coefficient of the rankings on attractions and privileges was −0.395 which is statistically significant at a 0.511 level. The results show that the two sets of data are not correlated (Table 3). Furthermore, the independent 2-sample t-test was undertaken to examine if there was any significant difference in mean value responses between the two respondent groups for each of the five attractions and privileges of PPP discussed. When the calculated significance level was below the allowable value of 0.05 for a certain attraction and privilege, a large variation was detected between the views of the respondents from Mainland China and Hong Kong. A significance level below 0.05 was used because this degree of significance has been commonly used by other researchers in similar studies. The population means are unknown as it would be impossible to know exactly how many industrial practitioners are involved with PPP projects in Mainland China and Hong Kong. Amongst the t-test results for the five attractions and privileges between Mainland China and Hong Kong respondents, two fell below a significance level of 0.05 (Table 4), the others were not statistically significant. For the attractions and privileges ‘Government assistance in financing’ and Tax exemption or reduction’, the significance levels showed that the respondents from Mainland China and Hong Kong shared very different views on their importance.

Table 4. Results of independent 2-sample t-test for privileges and attractions of PPP as identified by Mainland Chinese and Hong Kong respondents.

3a

3b

3c

3d

3e

5

Government sponsorship

Government assistance in financing Government guarantee

Tax exemption or reduction

Incentive of new market penetration

Equal variances assumed Equal variances not assumed Equal variances assumed Equal variances not assumed Equal variances assumed Equal variances not assumed Equal variances assumed Equal variances not assumed Equal variances assumed Equal variances not assumed

Levene's test for equality of variances

t-test for equality of means

F

Significance

t

Degree of freedom

Significance (2-tailed)

0.146

0.703

0.817

84

0.416

0.800

63.269

0.427

–2.499

84

0.014

–2.339

54.207

0.023

–1.592

84

0.115

–1.564

64.157

0.123

–2.718

84

0.008

–2.584

57.266

0.012

0.469

84

0.640

0.481

73.781

0.632

7.843

0.855

1.951

1.458

CONCLUSIONS

In this paper the results from a questionnaire survey looking at the attractions and privileges of PPP have been presented. In general the results found that all respondents rated the five pre-defined attractions and privileges importantly. The Mainland Chinese respondents rated three attractions and privileges offered by the government higher than the Hong Kong respondents. On the other hand those attractions and privileges rated higher by the Hong Kong respondents showed significant differences between mean values of the two survey groups. The results presented in this paper have formed a foundation for further study into the topic of implementing PPP projects in Mainland China and Hong Kong. ACKNOWLEDGEMENTS The work described in this paper was fully supported by a grant from the Research Grants Council of the Hong Kong Special Administrative Region, China (RGC Project No. PolyU 5114/05E). Sincere thanks go to Dr. Bing Li and Professor Akintola Akintoye for permitting the research team to adapt their survey questionnaire template. Special gratitude is also extended to those industrial practitioners from both Mainland China and Hong Kong, who have kindly participated in the questionnaire survey reported in this paper from October 2007 to December 2007.

0.006

0.358

0.166

0.231

REFERENCES Beijing Municipal Commission of Development and Reform (2006). 11th Five-Year Plan for the infrastructure sector of Beijing. October 8, 2006. Census and Statistics Department (2008). Hong Kong Special Administrative Region Government. (March 7, 2008). Chan, A.P.C. (2000). ‘‘Evaluation of enhanced design and build system—a case study of a hospital project.’’ Construction Management and Economics, 18(7), 863–871. Chan, A.P.C., Chan, D.W.M., and Ho, K.S.K. (2003). ‘‘An empirical study of the benefits of construction partnering in Hong Kong.’’ Construction Management and Economics, 21(5), 523–533. Chan, A.P.C., Sidwell, T., Kajewski, S., Lam P.T.I., Chan D.W.M. and Cheung E. (2007a). ‘‘From BOT to PPP—A Hong Kong Example.’’ Proceedings of the 2007 International Conference on Concession Public/Infrastructural Projects (ICCPIP), Dalian University of Technology, China, August 24–26, 2007, 9:010–018. Chan, A.P.C., Lam, P.T.I., Chan D.W.M., Sidwell, T., Kajewski S. and Cheung E. (2007b). ‘‘A Research Framework for Investigating Public Private Partnerships (PPP) in Hong Kong.’’ Proceedings of the 4th International Conference on Construction in the Twenty First Century (CITC IV), Gold Coast, Australia, July 11–13, 2007, 334–341. Chan, D.W.M., and Kumaraswamy, M.M. (1996). ‘‘An evaluation of construction time performance in the building industry.’’ Building and Environment, 31(6), 569–578. Chen, P.Y. and Popovich P.M. (2002). Correlation: Parametric and nonparametric measures. Thousand Oaks, California, Sage Publications.

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China Population Development and Research Center (2008). (March 7, 2008). Higgins, J.J. (2004). An introduction to modern nonparametric statistics. Pacific Grove, California, Thomson, Brooks/Cole. Information Services Department (2008). Shatin-Central link construction set for 2010, Hong Kong Special Administrative Region Government. (March 12, 2008). Keller, G. (2005). Statistics for Management and Economics. Seventh Edition. Thomas Brooks/Cole.

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Li, B. (2003). Risk management of construction public private partnership projects, Ph.D. Thesis, Glasgow Caledonian University, United Kingdom. Sachs, T., Tiong, R.L.K. and Wang, S.Q. (2007). ‘‘Analysis of political risks and opportunities in public private partnerships (PPP) in China and selected Asian countries.’’ Chinese Management Studies, 1(2): 126–148. Siegel, S. and Castellan, N.J. (1988). Nonparametric Statistics for the Behavioral Sciences. McGraw-Hill, Inc. SPSS (2002). SPSS 11.0 Statistical Algorithms, SPSS Incorporation, United States.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Training of skills and thinking in structural timber design L. Ozola Latvia University of Agriculture, Jelgava, Latvia

ABSTRACT: The current work expresses the author’s opinion on teaching of structural enginering courses and timber especially the objective of which is to develop the undergraduate student’s knowledge, understanding and skills by emphasizing the significance of the course project module. It is typical that the timber design incorporates a wide range of topics from structural analysis, and this training is a very significant one for new engineer even if he/she chooses for timber construction branch or another one. 1

2

EXPECTED QUALIFICATION FOR STRUCTURAL ENGINEER

We hear some opinion in the architectural and construction management circles that with the development of sophisticated software the role and responsibility of structural engineers decreases. It seems that while preparing the input data only the designers do not need to know the details of structural analysis. Is it so? No, just the reverse—the contemporary structural engineer must be able to perceive and trace the software-based design progress as he always remains the creator of the structure and the expert for structural design model treated. He must take a decision over a more reliable system composition and to assess adequately the assumptions on the behaviour of elements, joints and support conditions, and to take other critical choices as regards the longterm effects for the service life of the structure. The structural engineer must have a clear view for the future at present (Huntzinger et al. 2007). The following attributes have been expected to be relevant for an exceptional engineer (Dym et al. 2005): – Knowledge base to generate effective structures – Possessing of professional responsibility, ethics and ability for a critical evaluation before the acceptance of the design concept, the solution and/or a symplified model advanced or the first reasonable explanation proposed – Ability to employ the experience accumulated in engineering practice over the years – Permanent intellectual development and awareness to understand the impact of his/her decisions – Ability to express his/her opinion and to communicate effectively.

757

AIMS OF THE COURSE TRAINING

These are university years used effectively for studying the former knowledge and the experience including the errors committed and accidents happened. Otherwise the useless efforts and waste of time and funds may be expected. Also the theoretical background is significant. Every new idea should be generated on the theory basis. But there is no availability for the theoretical models perfectly proper to real structures. And it is the task for an engineer to analyse as far as the behaviour of structural system to be designed may be represented by conventional models, and to do the distinction between essentially influencing factors and those of lesser importance. Therefore the serious studies in the related fields are necessary. The aim of course training is to provide knowledge, understanding and necessary skills for the structural timber design including the choice of materials, defining design models, the calculations of internal forces as regards the loads in some more unfavorable combinations, checking of strength and stability conditions, evaluation of final displacements, design of connections, and to ensure the overall stability of the system. At the end of this course the student should be able to solve the task for covering some tens of meters for the span. He should understand the mechanical behaviour of elements and different connections used in timber structures, and to evaluate the strength and weakness of timber structures as regards long term loading effects in the certain service conditions. To achieve these goals, the following educational objectives have been established for the department of structural engineering: to provide a theoretical background for the design of timber structures, to provide a thorough training in the methods of analytical and

experimental analysis for the behaviour of structural elements, including the model formulation; to provide some instruction in the fundamentals of the design process according to Eurocode 5 conditions. 3

backward students. Hence the intellectual capacity is often not developed and employed worthwhile. Also, the education is affected by the fluctuations in the national economy both in quantity and quality sense when survival conditions are put in the forefront.

PREREQUISITES AND LEARNING STYLES 4

Often in comprehensive undergraduate education system the more democratic entrance conditions are accepted and the entrance examinations have been abated for the greatest part of applicants. Hence, the undergraduate students undertake the civil engineering program from different starting points and have some distinctive abilities and preliminary knowledge. Consequently the success of the conventional educational process may differ essentially from one person to another. In this situation the teaching strategy, i.e., the methods, requirements and assessment mode in educational complex need improvements. The structural engineering courses are rather difficult, the subjects comprise a big information package, complicated models to be studied, not easy perceived at the first sight, which requires the teaching methods provided by the course instructor aimed to activating the thinking and strong prerequisites and a good choice of learning style from students’ side. The developed cooperation and conversation between the teaching and learners’ sides become more critical for the course success, which may be measured not only in the terms of the students’ test scores, but also by personal satisfaction. Both the cognitive (evaluation, synthesis, analysis) and learning styles (Matrisciano & Belfiore 2003) are preferable for the studies of engineering subjects in mixed groups having distinctive prerequisites and motivation between the students. The students having the worst score results can obtain some knowledge and skills by a learning style which comprises many options for the combinations of learning dimensions (perception, input, organization, processing, understanding) and categories (sensory, intuitive, visual, auditory, inductive, deductive, active, reflective, sequential, global). Every learning style, excluding swotting up, may lead to some positive result in the developing process of thinking. It is another question whether the level of intellectual autonomy in engineering concepts and judgments will be attained or the intellectual humility will only take place when the professional is acting in full accord with the prescriptive codes, technical specifications, and/or instructions provided by the authorities. Both qualities of educated persons may be employed successfully if only they are in proper positions in the related construction branch. The teaching in mixed groups cover some significant deficiencies. The stagnation in the development of more capable students is observed when the most efforts are devoted to training middling and

SYLLABUS OF TIMBER STRUCTURES

The course of timber structures form three percents of total credit point volume acquired by a student studying the civil engineering programme. The course is constructed by three interacting teaching modules. The theoretical background has been provided by the lectures including more significant concepts and design methods. The second module comprises the laboratory works (tests of structural elements and joints under static loading) and tutorials for the applying of calculation methods. The main emphasis has the last module—course project which is the individual work and comprises the design of timber frame system for the given span size (from 10 up to 36 meters). Specific features and a large variety of structural solutions enable to define the different levels for studying and understanding of timber engineering course corresponding to the prerequisites and motivation of students. There is a possibility to play with span sizes to be covered, the complexity of structures selected, the section and material types, and the fasteners used in joints, see Table 1. Some graphical representations used for explanations are shown in Figures 1, 2. The computer programs (Statics, MS Excel, MathCAD) enable to perform the calculations and drawings effectively.

Table 1.

Course training levels in timber design. Description of training levels

Design topic

Lower

Higher

Choice of type of structure Determination of loads Definition of design model Calculation of internal forces Design of elements

Defined by instructor Dead, imposed, snow

Choice argumented by student + wind pressure

Simplified

Correct

Manual Satisfaction of EC5 conditions Use of specified fastener Bracings formal CAD format Not completed

Using software Optimal design Optimal joint design Bracing system analysed CAD format Checking of deformed state, 2nd order analysis

Design of joints Stability of system Drawings Assessment of design for longterm service life

758

A zona

pe

c = -0.5 F: c pe= +0.7

B zona

1

1

a

E zona: cpe = -0.54

c = -0.5 F: c pe= +0.7

a

pe

e/4= 4450

B: cpe = -0.8

A: cpe =-1.2

e/10= 1780

d= 15400

a

L

1 - 1; 2 - 2

a

a

z = 7418

L

wd1= 2.7 kN/m

h+u

w d2

=2

.5

w d3 =

/m 1 kN

wd4 = 0.5

e = 2h = 17800 mm wd5 = 0.3 kN /m

w d6

=0

.46

h = 8900

L

h

a

6400

h

D zona: cpe = +0.8

Wind direction (Θ = 0°)

wd7 = 1.1 kN/m

L

c = -0.5 G:c pe= +0.7 pe

a

I: cpe = -0.42 cpe = -0.06 b = 35600

H: cpe = -0.33 cpe = +0.18

c = -0.96 J: c pe= +0.02 pe

2

h

2

a

Figure 1. Some visualization tools used for explanations in undergraduate studies for calculation of loads and definition of design model.

5

DISCUSSION

In the best meaning the structural timber design is a complex process in which the designers generate, evaluate, and specify the concepts for the system composition based on their erudition, knowledge, creativity,

759

working abilities and ability to a critical judgement of the affecting factors and conditions changing during the service life of the structure. These are university years when the basic skills for this qualification should be trained. There are many difficulties for both course instructor and the student

t t hd

t

N c,d B

Nt,d

dc a3

hb

a2b

a2d

a4

a4

a1b a3

a1d

D

Vd

t2

t2

hd

N c,d

a2

a3

a1d

b

α

Vd

a1d

hb

a4

a2b

a2c

Figure 2.

t

a1b

a3

nrb

a4

nrd

Nt,d

t1

t1

Some visualization tools used for explanations in undergraduate studies for structural timber design.

as regards complexity of stress states, fastener types and connection capacities to be analysed and design conditions to be satisfied. It is the training experience accumulated by many years which proves that it is not possible to define unique effective pedagogical methods and styles to attain the expected ends as the case in point is an intellectual and professional

development of the personality having his/her individual will, motivation, talent and capacity for work. Always the discussion on promoting of the creativity has been opened when deals with teaching methods. Keeping out of new ideas put up by the student lacks support from any up-to-date accepted teaching method. Notice only that it is significant to gain a point

760

for student’s understanding that in current stage of engineering developments the knowledge is the main pillar to be set as foundation for creativity. Otherwise abundant of mental energy may be wasted for ‘‘invention of bicycle’’, what may be useful for intellectual development as such but not bringing considerable contribution in growth of the professional qualities. The developed ability to analyse the affecting factors (what requires the sufficient knowledge base) has been stable foundation for sustainable engineering practice. Another actual problem has been advanced while teaching methods are dicussed is the teamwork training. Really, the structural course projects are of great time- and labour-consuming ones. The teamwork would be more effective when the design task has been performed conscientiously by any member and mutual communication takes place for discussing of problems, besides the finding and correction of errors in due time may be expected more often. Yet the real situation often is far from idealized. Usually some leader of team carries out the bulk of work. At the same time another members have not forced for studying of severe things. What is more, let us allow for independent development of individual at this first stage of professional education! Good results may be expected when some individual course projects are presented by authors before group. 6

SUMMARY

It is the most significant task to foster the consciousness and motivation of the student for training the structural analysis and design in order to reach the independence and selfreliance in his/her learning process.

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Permanent development of individual skills in structural engineering is more important as only the strong individuals will be able to join with the capable problem-solving workgroups in the future. Also, the developed oral and written communication skills have been the particular concern of a new engineer. It is important the permanent intellectual development for both instructor and student. REFERENCES Dym, C.L., Agogino, A.M., Eris, O., Frey, D.D., Leifer, L.J. 2005. Engineering Design Thinking, Teaching, and Learning. Journal of Engineering Education January: 103–120. Huntzinger, D.N., Hutchins, M.J., Gierke, J.S., Sutherland, J.W. 2007. Enabling Sustainable Thinking in Undergraduate Engineering Education. International Journal of Engineering Education Vol. 23, No. 2: 218–230. Matrisciano, A., Belfiore, N.P. 2003. An investigation on the students’ learning styles in an Advanced Applied Mechanics Course. Proceedings of the Second International Conference on Structural and construction Engineering: 2129–2135. A.A. Balkema Publishers. Niewoehner, R.J. 2006. Applying a critical thinking model for engineering education. World Transactions on Engineering and Technology Education Vol. 5, No. 2: 341–344.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Use of alternative dispute resolution in construction: A comparative study S.O. Cheung & P.S.P. Wong Construction Dispute Resolution Research Unit, Department of Building and Construction, City University of Hong Kong

P. Kennedy School of the Built and Natural Environment, Glasgow Caledonian University, Glasgow, Scotland

ABSTRACT: The quest for more efficient construction dispute resolution has prompted major policy decisions in relation to the use of ADR. This paper presents a comparison of the construction dispute handling approaches in five common law based jurisdictions: Australia, Hong Kong, New Zealand, Singapore and the United Kingdom. Based on a literature review, it is found that the right to refer construction disputes (of any type) to adjudication is supported by statute in the United Kingdom and New Zealand. In contrast, in Australia and Singapore, only payment-related disputes are singled out for statutory adjudication. In Hong Kong, while the use of ADR techniques remains voluntary, mediation is the preferred choice. Four reasons that may have led the above differences are suggested. They are: (1) the contracting parties’ right to employ ADR techniques, (2) the consensus to champion statutory adjudication, (3) the scope of legislation and (4) the ambition to achieve final settlement through the use of ADR.

1

INTRODUCTION

Arbitration and litigation are conventional methods to resolve construction disputes. Nevertheless, these processes have been criticized for being costly, lengthy and adversarial (Jahren & Dammeier 1990; Lim 1993; Cheung 1999). The use of Alternative Dispute Resolution (ADR) has been advocated to address the above criticism (Hibberd & Newman 1999). During the last two decades, the quest for greater use of ADR in construction has prompted its statutory use in some jurisdictions (Gaitskell 2007; Kennedy 2008). The first piece of legislation in this regard is the Housing Grants, Construction and Regeneration Act (HGCRA) that was passed in 1996 in the United Kingdom (UK). The statutory use of adjudication for construction disputes was brought into force in 1998 through a piece of secondary legislation (Scheme for Construction Contracts) which sets out the minimum standards of payment terms (to protect the sub-contractors’ cash flow) and the procedure to be followed in an adjudication. Since then, the use of adjudication has become available as a statutory right open to all parties in all construction contracts. Nevertheless, if the parties are not happy with the adjudicator’s decision, they can refer the dispute to arbitration in accordance with their contract—usually upon practical completion of the project (Anderson 2008; Kennedy 2008). Following this UK initiative, similar statutory use of ADR had been promulgated in Australia, New Zealand and Singapore (Gaitskell 2007). Moreover, ADR movement in Hong

763

Kong has been confined at the contractual level (Chau 2007). Voluntary mediation has been introduced in the standard forms of construction contract for public projects since early 1990s (Cheung 1999). In 2005, the private sector followed the Government’s footsteps and voluntary mediation was for the first time included in the Joint Contract Working Committee’s Standard Form of Building Contract (Leung 2007). More significantly, the Hong Kong Government has recently made a policy decision to promote a greater use of mediation to enhance speedy and cost effective construction dispute resolution (Wong 2008). The fact that different approaches are used to tackle the same problem makes it interesting to study the underlying reasons. For this purpose, this paper presents a comparison of the approaches taken in five major jurisdictions in handling construction disputes. These five jurisdictions are: Australia, Hong Kong, New Zealand, Singapore and the UK. The comparison also provides valuable information on the difference in emphasis towards construction dispute resolution. 2

CONSTRUCTION DISPUTE HANDLING APPROACHES

Based on a literature review, construction disputes in the five jurisdictions examined have typically been handled by two approaches as presented in Figure 1. Under the Type A approach, a dispute will be resolved firstly by ADR. If this fails, the dispute

Payment related disputes Hong Kong

United Kingdom, New Zealand, Australia, Singapore

Figure 1.

Nonpayment related disputes Australia, TYPE A APPROACH: Singapore, Hong Kong

United Kingdom, New Zealand

TYPE B APPROACH:

Dispute handling approaches of the five jurisdictions.

will then be referred to arbitration upon practical completion of the project. This approach has been embodied in a number of standard forms of contract. Take Hong Kong as an example, mediation has become an integral part of the dispute resolution clause in the major Government General Conditions of Contracts for viz.: Airport Core Program 19921 , Civil Engineering Works 19992 , Building Works 19993 and Design and Build Contracts 19994 and the latest version of the private forms of building contract published by the Joint Contract Working Committee5 . Similarly, in the UK, mediation and adjudication have been introduced as an optional dispute resolution approach as stipulated in the Joint Contracts Tribunal (JCT) Standard Building Contract 20056 and Standard Design and Build Contract 20057 . It is noteworthy that, under the contractual framework, the use of ADR techniques before referring a dispute to arbitration is voluntary. If either of the contracting parties refuses, the use of the ADR techniques can be bypassed (Cheung & Yeung 1998). Furthermore, the prescribed voluntary ADR procedures typically involve appointing an independent neutral to give expert opinion. Nevertheless, in contrast to arbitration and litigation, the expert’s recommendations are typically not binding on the parties (Jones 2006). Contrary to the Type A approach, with the Type B approach a dispute could firstly be referred to statutory adjudication. The arrangement has been considered effective to tackle two major deficiencies of the conventional contractual dispute resolution regime in construction: [1] the parties’ right to bypass ADR before proceeding to arbitration and [2] the enforcement of the non-binding experts’ determination (Jones 2006). Through legislation, parties now have the right to refer a dispute to adjudication. Furthermore, the decision of the adjudicator is binding unless and until the dispute

has been settled by agreement, by litigation or an award made in a subsequent arbitration (Gaitskell 2007). In Hong Kong, construction disputes have typically been handled by the Type A approach (Chau 2007; Leung 2007). In New Zealand and the UK, in contrast, Type B approach has been used to handle construction disputes (Gaitskell 2007). In Australia and Singapore, Type A approach applies to non-payment related disputes like the disputes about claims arisen from the extension of time, delay and disruption, personal liability, while Type B approach applies in handling payment related disputes (including progress payment, one-off payment and final payment) (Jones 2006). 3

FACTORS AFFECTING THE USE OF DISPUTE HANDLING APPROACH

Based on a literature review, four major reasons that may have led to the above differences in dispute handling approaches are identified as follows: 3.1

Right to employ ADR techniques

The Type A approach has been adopted in Hong Kong for handling all types of construction dispute. In particular, voluntary mediation has been incorporated contractually (Chau 2007; Leung 2007). For example, a reference to ADR has been set as a condition precedent to ‘early’ arbitration. It is stipulated in Clause 41.5(1) of the Standard Form of Building Contract for use in the Hong Kong Special Administrative Region, Private Edition (with quantities) 2005 that if the contracting parties have attempted to negotiate through designated representatives and mediation but the payment-related dispute remains unsettled, they are allowed to commence arbitration before practical completion of the

764

Work. Nevertheless, the use of voluntary mediation remains relatively narrow in Hong Kong (Hill & Hall 2008). Furthermore, having considered the UK experience in voluntary mediation before trial, the Judiciary of Hong Kong proposes the use of cost sanction to penalize a disputant who unreasonably refuses to mediate (Turner 2004; Wong 2006). Despite this, the Type A Approach has been regarded as being ineffective in reducing the umber of arbitration and litigation cases in construction. One of the major reasons is that it is neither a breach of contract or law in Hong Kong if the parties fail to refer a matter to mediation (Hill & Hall 2008). Unless parties are legally or contractually bound to use mediation, their reliance on arbitration and litigation to reach final settlement of construction disputes shall remain unchanged (Hill & Hall 2008). Unsurprisingly, jurisdictions that opted for Type B Approach have their own piece of legislation supporting the statutory use of adjudication (Table 1 refers). Moreover, the adjudication procedures as stipulated in the Act would take precedent over contract terms if they were unclear or designed to provide less protection for the rights of the parties. For example, as stipulated in s108(5) of the UK Act (1996), the adjudication procedures detailed in the Scheme for Construction Contracts (England and Wales) Regulations 1998 shall be incorporated in the construction contracts with which provisions failed to comply with the rules as stipulated in s108(2) (Teo 2008; Anderson 2008). Similar provisions can be found in s9 of the New Zealand Construction Act 2002 that ‘applies to every construction contract . . . related to carrying out construction work in New Zealand’ (NZCA 2002). In Australia and Singapore, their respective Building and Construction Table 1.

Industry Security of Payment Acts shall take effect if the payment terms in the construction contracts were silent or contradictory to the purpose of securing the sub-contractor’s payments (Teo 2008). Positive outcomes derived from providing parties’ statutory right to adjudicate have been reported in a number of countries. This includes a significant drop of arbitration and litigation cases and the mounting successful adjudication cases (Kennedy 2006; Kennedy & Milligan 2007; Gould & Linneman 2008). 3.2

Consensus to champion statutory adjudication

Reported in a number of review studies, the legislation on compulsory adjudication had typically been preceded with a consensus view to deal with the cash flow problems of the sub-contractors (Anderson 2008, Dancaster 2008; Hill & Hall 2008). For jurisdictions that have been employing the Type B Approach to handle disputes, legislation on statutory adjudication were typically triggered by a quest for practical solutions against payment being denied or made contractually subject to pay-when-paid clauses (Gaitskell 2007). Nevertheless, it appears that the voice pressing the legislature to ban pay-when-paid clauses or impose statutory adjudication is less resounding in Hong Kong. In this aspect, Hill & Hall (2008) pinpointed three reasons: (1) Contrary to other jurisdictions, the construction community in Hong Kong is relatively small. To sustain the business network with the suppliers and sub-contractors, main contractors are prudent in withholding payments; (2) in Hong Kong the Employment Ordinance (Cap. 57) statutorily protects against the contractors’ nonpayment

Legislation related to compulsory adjudication in different countries/regions.

Location

United Kingdom

New Zealand

Singapore

Legislation

The Housing Grants, Construction and Regeneration Act 1996 (described as the ‘UK Act 1996’ hereafter)

Construction Contracts Act 2002 (described as the ‘NZ Act 1999’ hereafter)

Building and Construction Industry Security of Payment Act 2004 (described as the ‘Singapore Act 2004’ hereafter)

Time for decision

28 calendar days (additional 14 days subject to agreement between both parties) 8 No

20–30 working days (or subject to agreement between both parties) 9

7–14 working days (or subject to agreement between both parties) 10 Yes

Restrict the right of adjudication in relation to payment only?

No

765

Australia—New South Wales, Victoria, Queensland Building and Construction Industry Security of Payment Act (described as the ‘NSW Act 2002’, ‘Victoria Act 2002’ and ‘Queensland Act 2004’ hereafter) 10 working days (or subject to agreement between both parties) 11 Yes

to employees. This dampens the demand to legislate against withholding contractors’ payment; and (3) while the outcomes of mediations are confidential, it is widely understood that there has been a long history of adopting mediation successfully on public projects in Hong Kong. 3.3

Scope of legislation

The scope of legislation may direct the dispute handling approach. For example, in the Building and Construction Industry Security of Payment Acts in Australia and Singapore, contractors are statutorily entitled to progress payments. The Acts not only render pay-when-paid provision of no effect12 , but also empower contractors a right to adjudicate the payment claims13 . The adjudicator is required to make a decision, which is binding unless and until the dispute is finally determined by litigation (Table 1 refers). The above indicates that adjudications are restricted to disputes concerning the contractors’ and sub-contractors’ right to progress payment. As such, the legislation in Australia and Singapore has framed adjudication as an adjunct mechanism to deal with payment-related disputes (Teo 2008; Jones 2006). This may help to explain why payment-related disputes in Australia and Singapore are typically handled by Type B approach. However, it is worth noting that the primary purpose the UK and New Zealand legislations is to provide parties to a construction contract a ‘right to refer a dispute to adjudication’14 . s108(1) of the UK Act 1999 and s25(1) of the NZ Act 2002 stipulate that all types of dispute are statutorily admissible to be referred to adjudication (Jones 2006). In this regard, the UK and New Zealand construction dispute resolution regimes are Type B. 3.4 Ambition to achieve final settlement by using ADR

being supportive to the adjudicator’s decision, they have the right to set aside the adjudicator’s determination if there has been a substantial denial of natural justice (Gaitskell 2007). The above findings indicate the difficulties to achieve final settlement by using adjudication in Australia and Singapore. Some may argue that in terms of the enforceability of the adjudicators’ determination, the situation in the UK is indeed similar to that in Australia and Singapore. It is evidenced by s108(3) of the UK Act that, ‘‘ . . . the decision of the adjudicator is binding until the dispute is finally determined by legal proceedings, by arbitration . . . or by agreement’’ and the parties may agree to ‘accept the decision of the adjudicator as finally determining the dispute’ It seems indicating that the parties’ agreement on the adjudicator’s decision is not part of the statutory scheme. Nevertheless, having considered that the UK Act introduces a statutory right of a party to refer all types of dispute to adjudication, adjudicators are empowered to handle dispute cases related to the final contract price (Jones 2006). The chance of achieving construction dispute settlement by using adjudication is therefore higher in the UK because of the wider power of the adjudicator. As stipulated in the 2002 NZ Act, the adjudicators’ determination is enforceable for payment-related disputes but unenforceable per se on matters about rights and obligations16 . under s61 of the 2002 NZ Act, the court in dealing with non-payment disputes, must have regard to, but is not bound by, the adjudicator’s determination 17 . Regarding the adjudicators’ decision about rights and obligations, the successful party can enforce it through bringing the case to court proceedings. The above indicates that the use of Type B approach may be affected by the prospect of the adjudicators’ decision to become a final determination (Jones 2006). 4

The construction dispute handling approaches in Australia and Singapore are different from that in the UK and New Zealand. This may be due to the difference in intention behind the scheme. The Building and Construction Industry Security of Payment Acts in Australia and Singapore stipulate that the matters being subject to statutory adjudication are restricted to ‘payment claims’13 . The adjudicator appointed under these Acts is responsible ‘to determine the amount of the progress payment . . . , the date on which any such amount becomes payable . . . and the rate of interest payable on any such amount’ only15 . In other words, it is outside the scope of the Acts for the adjudicator to determine disputes concerning non-payment related matters. Moreover, the sub-contractors have been given statutory rights to sue for the amount determined by the adjudicator as a debt-due (Jones 2006). Furthermore, despite the courts in Australia and Singapore

CONCLUDING REMARKS

This study compares and contrasts the construction dispute handling approaches in five jurisdictions: Australia, Hong Kong, New Zealand, Singapore and the UK. Two approaches are identified: Type A approach embodies a voluntary use of ADR techniques in the dispute handling process; while Type B approach focuses on the party’s right to adjudicate. By adopting Type B approach, disputes shall be subjected to statutory adjudication. It is found that, except Hong Kong, the other four jurisdictions employ a Type B approach to handle payment-related disputes. In Hong Kong the contracting parties’ right to employ ADR techniques has yet to be statutorily protected. This may due to the fact that a consensus view on the solution to deal with the unfair construction contract terms in Hong Kong has yet been reached. The disputes handling approach

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being adopted in Hong Kong appears to echo with the intention of the Government to promote greater use of voluntary mediation. Nevertheless, the success of such a dispute resolution approach relies on the parties’ cooperativeness and willingness to reach final settlement. Without the parties’ genuine desire to resolve the dispute, the use of voluntary ADR may well just another cost and time consuming exercise (Jones 2006). Perhaps restricted by the ultimate goal of imposing statutory adjudication, Australia and Singapore employ the Type B approach to handle payment-related disputes only. The Type B approach also applies to the handling of non-payment-related disputes in the UK and New Zealand. Having considered that the legislation in these jurisdictions confers on a party a statutory right to seek adjudication as the way to settle disputes, it is believed that the use of construction dispute handling approaches may vary with the parties’ statutory right to adjudicate. Furthermore, the findings suggest that the Type B approach applies in jurisdictions that enable utilizing the adjudicators’ decision as the final determination. ACKNOWLEDGEMENTS The work described in this paper was fully supported by a grant from CityU (Project No. 7002498). ENDNOTES 1. See Clause 92 of the Government of Hong Kong General Conditions of Contracts for the Airport Core Program 1992. 2. See Clause 86 of the Government of the Hong Kong Special Administrative Region General Conditions of Contract for Engineering Works 1999. 3. See Clause 86 of the Government of the Hong Kong Special Administrative Region General Conditions of Contract for Building Works 1999. 4. See Clause 86 of the Government of the Hong Kong Special Administrative Region General Conditions of Contract for Design and Build Contracts 1999. 5. See Clause 41 of the Standard Form of Building Contract for use in the Hong Kong Special Administrative Region, Private Edition (with quantities) 2005. 6. See Clauses 9.1 and 9.2 of the JCT Standard Building Contract 2005. 7. See Clauses 9.1 and 9.2 of the JCT Standard Design and Build Contract 2005. 8. See s108(2) of the UK Act 1999. 9. See s46(2) of the NZ Act 2004. 10. See s17(1) of the Singapore Act 2004.

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11. See s21(3) of the NSW Act 2002, s22(4) of the Victoria Act 2002, and s25(3) of the Queensland Act 2004. 12. See s12(1) of the NSW Act 1999, s13(1) of the Victoria Act 2002, s16(1) of the Queensland Act 2004 and s9(1) of the Singapore Act 2004. 13. See s17 (1) of the NSW Act 1999, s18(1) of the Victoria Act 2002, s21(1) of the Queensland Act 2004 and s13(1) of the Singapore Act 2004. 14. See s108(1) of the UK Act 1996 and s25(1) of the NZ Act 2002. 15. See s22(1) of the NSW Act 1999, s23(1) of the Victoria Act 2002, s26(1) of the Queensland Act 2004 and s17(2) of the Singapore Act 2004. 16. See s48, s58 and s59 of the NZ Act 2002. 17. See s61 of the NZ Act 2002.

REFERENCES Anderson, R.N.M. 2008. Adjudication in the United Kingdom: Constitutional implications. Journal of Professional Issues in Engineering Education and Practice, ASCE 134(3): 309–314. Chan, C.F. 2006. Security of Payment Legislation-Case of a Blunt but Practical and Equitable Instrument, Journal of Professional Issues in Engineering Education and Practice, ASCE 133(3): 248–257. Chau, K.W. 2007. Insight into Resolving Construction Disputes by Mediation/Adjudication in Hong Kong. Journal of Professional Issues in Engineering Education and Practice, ASCE 133(2): 143–147. Cheung, S.O. 1999. Critical factors affecting the use of alternative resolution processes in construction. The International Journal of Project Management 17(3): 189–194. Cheung, S.O. & Yeung, Y.W. 1998. The Effectiveness of the Dispute Resolution Advisor System: A Critical Appraisal. The International Journal of Project Management 16(6): 367–374. Dancaster, C. 2008. Construction adjudication in the United Kingdom: Past, present and future. Journal of Professional Issues in Engineering Education and Practice, ASCE 134(2): 204–208. Gaitskell, R. 2007. International statutory adjudication: its development and impact, Construction Management and Economics 25(7): 777–784. Gould, N. & Linneman, C. 2008. Ten years on: Review of adjudication in the United Kingdom. Journal of Professional Issues in Engineering Education and Practice, ASCE 134 (3): 298–301. Hibberd, P. & Newman, P. 1999. ADR and Adjudication in Construction Disputes. London: Blackwell Science Ltd. Hill, T. & Hall, C.J. 2008. Adjudication: Temporary binding and tiered dispute resolution in construction and engineering: Hong Kong experience. Journal of Professional Issues in Engineering Education and Practice, ASCE 134(3): 306–308. Jahren, C. & Dammeier, B. 1990. Investigation into Construction Disputes. Journal of Management in Engineering, ASCE 6 (1): 39–46.

Jones, D. (2006) Adjudication: Evolutionary step or revolutionary idea? In Problem solving: rules, roles and regulations—Canadian College of construction lawyers 9th Annual conference, 1–4 June 2006, Fox Harbour, Nova Scotia. Kennedy, P. 2008. Evolution of statutory adjudication as a form of dispute resolution in the U.K. construction industry. Journal of Professional Issues in Engineering Education and Practice, ASCE 134(2): 214–219. Kennedy, P. & Milligan, J.L. 2007. Research analysis of the progress of adjudication based on returned questionnaires from adjudicator nominating bodies (ANBs), Report No. 8, November 2007. U.K.: Adjudication Reporting Centre. Leung, H.F. 2007. The Role of the Judiciary and Statutes in Relation to Mediation. Asian Dispute Review October 2007: 128–131.

Lim, L.Y. 1993. Alternative Dispute Resolution in Singapore’s Construction Industry. Australian Dispute Resolution Journal 4(2): 114–124. Teo, P.J.P. 2008. Adjudication: Singapore perspective. Journal of Professional Issues in Engineering Education and Practice, ASCE 134 (2): 224–230. Turner, P. 2004. Enforceability of ADR Clauses, The Quarterly Publication of the Hong Kong Mediation Council 19 (February): 3–4. Wong, Y.L. 2006. The Benefits of Mediation, Asian Dispute Review July 2006: 100–102. Wong, Y.L. 2008. The use and development of mediation in Hong Kong. Asian Dispute Review April 2008: 54–56.

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Construction maintenance and infrastructure

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Construction safety in the repair and maintenance sector A.P.C. Chan, F.K.W. Wong, M.C.H. Yam, D.W.M. Chan & C.K.H. Hon Construction Safety Research Group, Research Centre for Construction and Real Estate Economics, Department of Building and Real Estate, The Hong Kong Polytechnic University, HKSAR, China

D. Dingsdag University of Western Sydney, Australia

H. Biggs Queensland University of Technology, Australia

ABSTRACT: Statistics indicate that the percentage of fatal industrial accidents arising from repair, maintenance, minor alteration and addition (RMAA) works in Hong Kong was disturbingly high and was over 56% in 2006. This paper provides an initial report of a research project funded by the Research Grants Council (RGC) of the HKSAR to address this safety issue. The aim of this study is to scrutinize the causal relationship between safety climate and safety performance in the RMAA sector. It aims to evaluate the safety climate in the RMAA sector; examine its impacts on safety performance, and recommend measures to improve safety performance in the RMAA sector. This paper firstly reports on the statistics of construction accidents arising from RMAA works. Qualitative and quantitative research methods applied in conducting the research are discussed. The study will critically review these related problems and provide recommendations for improving safety performance in the RMAA sector. 1

INTRODUCTION

Safety has been recognized as one of the besetting problems with the Hong Kong construction industry. The Hong Kong Construction Industry Review Report (HKCIRC 2001) highlighted safety to be one of the six major areas for improvement. With shrinkage of the construction market and reduced volume of new works, repair, maintenance, minor alteration and addition (RMAA) works are playing an increasingly important role in the local construction industry. RMAA works refers to those minor works such as construction projects for village-type houses in the New Territories, minor alterations, repairs, maintenance and interior decoration of existing buildings (Labour Department 2008b). Research on safety performance of RMAA works, however, has been rather limited. 1.1

RMAA works in the construction industry

The construction industry contributed 2.7% to the GDP (at current factor cost) of Hong Kong in 2006 (Census and Statistics Department 2008a). The number of persons directly engaged in the construction industry was 135,337 in 2006 (Census and Statistics Department 2008b). For the types of construction activities, construction work in RMAA is becoming more and more important as it contributed to 53.2%

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of the total construction market in 2007. As shown in Table 1, the proportion of RMAA works to total construction market has been doubled from 1997 to 2007. Importance of RMAA works to the whole construction industry is expected to increase continuously in the foreseeable future. To counteract the disastrous effect of the global financial crisis to the construction industry, the HKSAR Government is trying to assist the construction industry by launching more minor works to create immediate employment opportunities (Development Bureau, HKSAR 2008). The Development Bureau of the HKSAR has just announced that HK$ 8.56 billion (approx. US$ 1.10 billion) will be spent on minor works in 2009/2010 to create 1600 jobs. The projects include refurbishment of the exterior of 50 government buildings, renovation of aged protective surfaces of 500 slopes, the installation and retrofitting of energy-efficient facilities for various government departments, and provision of green roofs on 40 government buildings (The Standard on 14 January 2009). 1.2

Safety performance of RMAA works

Referring to Table 2, safety performance of the construction industry has been improved tremendously. Accident rates per 1000 workers fell from 247.9 in 1998 to 60.6 in 2007, representing a remarkable drop

Table 1.

Gross value of construction work at current market prices (1997–2007). (Unit: HK$ million at current prices; US$1 = HK$ 7.8) 1997

1998

1999

2000

2001

2002

2003

2004

Residential (A) 36,633 48,761 56,225 51,920 41,774 36,503 28,612 Non-residential (B) 32,392 33,866 20,455 17,407 16,026 16,502 18,243 Civil Engineering (C) 29,957 19,349 16,873 20,583 24,491 21,358 20,710 Total Construction Investment (A+B+C) 98,982 101,975 93,553 89,910 82,290 74,362 67,564 Repair, Maintenance, Minor alteration and Addition (D) 32,518 31,341 32,884 32,161 31,696 31,638 31,468 Total Construction Market (A+B+C+D) 131,500 133,316 126,437 122,071 113,986 106,000 99,032 Percentage of RMAA Works to Total Construction Market (%) 24.7 23.5 26.0 26.3 27.8 29.8 31.8

2005

2006

2007

20,085 16,945 15,518 16,064 17,425 17,060 14,161 17,289 19,044 14,686 12,311 10,123 56,553 48,691 41,990 43,476 36,618 42,160 48,240 49,390 93,171 90,851 90,230 92,866

39.3

46.4

53.5

53.2

Source: Report on the Quarterly Survey of Construction Output, Table 1A and Table 3, C&SD, The HKSAR Government. Table 2.

Industrial accidents of the construction industry.

(a) All reported construction accidents** (b) Accident rate per 1000 workers (c) All reported accidents in RMAA Works** (i) No. of reported accidents in RMAA Works in public sector sites** (ii) No. of reported accidents in RMAA Works in private sector sites** Percentage of RMAA accidents to all reported construction accidents (c)/(a)]

1998

1999

2000

2001

2002

2003

2004

2005

2006

2007

19,588 (56)

14,078 (47)

11,925 (29)

9206 (28)

6239 (24)

4367 (25)

3833 (17)

3548 (25)

3400 (16)

3042 (19)

247.9 3510 (7)

198.4 3328 (10)

149.8 3402 (12)

114.6 2582 (4)

85.2 1925 (10)

68.1 1485 (8)

60.3 1454 (6)

59.9 1509 (12)

64.3 1697 (9)

60.6 1524 (6)

466 (0)

449 (3)

475 (1)

331 (2)

250 (2)

158 (2)

104 (0)

64 (2)

60 (5)

50 (1)

3044 (7)

2879 (7)

2927 (11)

2251 (2)

1675 (8)

1327 (6)

1350 (6)

1445 (10)

1637 (4)

1474 (5)

17.9%

23.6%

28.5%

28.0%

30.9%

34.0%

37.9%

42.5%

49.9%

50.1%

** Figures in the brackets denote the number of fatalities. Source: Labour Department HKSAR (2008b) Accidents in the Construction Industry of Hong Kong (1998–2007).

of 75.6% (Labour Department 2008b). However, we should note that the accident rate per 1000 workers has reached a plateau after 2003. The challenge to the industry would be how to further drive down accident rates to achieve continuous improvement in safety performance. It is noteworthy that there has been a remarkable increase in the percentage of accidents on RMAA worksites over the past ten years, from 17.9% of 1998 to 50.1% of 2007. Admittedly, this change may be due to the increase in proportion of RMAA works to the whole construction industry; however, it still provides a valid ground for our investigation on safety of RMAA worksites.

1.3

Safety climate

Zohar (1980) defined safety climate as ‘a summary of molar perceptions that employees share about their work environments . . . a frame of reference for guiding appropriate and adaptive task behaviors’ (p.96). Griffin and Neal (2000) suggested that employees’ perceptions of the policies, procedures, and practices relating to safety comprise the safety climate. As stated by Zohar (2003), safety climate reflects the true perceived priority of safety in an organization. Some researchers define safety climate as a current-state reflection of the underlying safety culture (Mearns et al. 2001, 2003).

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Zohar (2003) separated safety climate into two dimensions: level and strength. The level of safety climate reflects shared perceived priority of safety whereas strength of safety climate is homogeneity of perceptions of the importance of safety. Safety systems and polices do not automatically generate safety; it is the true priority of safety that is consensually perceived by people that affects their safety behavior. For example, a company has imposed overt safety policies and management systems; however, when safety and time come into conflict, the manager gives the message that time overrides safety. People inside the organization will project a low priority for safety, i.e. a low level of safety climate. Safety climate influences one’s behavior through behavior-outcome expectancies (Zohar 2003). Low safety climate implies that people assign lower weight to safety but greater value to short-term gains, e.g. finishing the work faster. Under a low safety climate, people also underestimate the likelihood of possible injury. It is believed that expectancies influence prevalence of safety behavior which in turn influences company safety performance. 1.4 Significance and value This study aims to investigate the causal relationship between safety climate and safety performance of RMAA works. As supported by statistics, safety legislation and related policies can effectively drive down accident rates to a certain level, but these will reach a plateau sometime. This tendency is reflected in other developed industrialized societies, notably Australia where current research shows that legislative compliance on its own is not sufficient for continuing reduction in incidence and severity. For example, in the State of NSW, Australia, compensated manual handling injuries were consistently within the 30–35 percent of total compensated injuries between 1997/98 and 2006/07, notwithstanding that the overall number of compensated injuries has fallen over the last decade or so (WorkCover 2007). It is believed that positive safety culture is a possible way to further improve safety performance. This research on safety climate will contribute to providing insightful ways on how to change people’s safety behaviors. Besides, safety climate is a prevalent issue that interests practitioners in the construction industry. The Occupational Safety and Health Council (OSHC) of the HKSAR has recently developed and promoted the Safety Climate Index (SCI) to the industry. SCI is a survey tool for measuring level of safety climate. Industry practitioners have practical reasons to know more about safety climate so as to make better use of safety climate scores. Examples include the meaning of high/ low levels of safety climate, implications to organizational policies and management, the way to

further improve safety performance and to create safe working environments. 2 2.1

OVERALL RESEARCH FRAMEWORK Literature review

To begin the study, a systematic literature review has been done on keywords like safety climate, safety culture and safety performance in two academic electronic databases, ISI Web of Knowledge and Scopus. A total of 78 articles were identified in ISI Web of Knowledge and 38 articles were identified in Scopus, confirming results of each other. Seven key journals have been identified as well. Follow up searches were done on individual journals and reference lists of individual papers. Another search was done on top-tier journals in the construction industry. 2.2

Questionnaire Survey

A questionnaire survey instrument is an effective method to seek a large sample size for quantitative data analysis (Hox et al. 2008). The survey instrument will focus on identifying the relationships between the safety climate factors and safety performance. A safety climate survey tool, Safety Climate Index (SCI), developed by the Occupational Safety and Health Council, HKSAR (OSHC, 2008) will be employed and supplemented with other relevant questions. Comments gained from the pilot questionnaire will be incorporated to refine the research questionnaire. RMAA contractors listed under the Hong Kong General Building Contractors’ Association, the Voluntary Subcontractor Registration Scheme, and the

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Figure 1.

Overall research framework for the study.

Census & Statistics Department will be invited to participate in this questionnaire survey. It is likely that the responses on the safety climate factors may be affected by their working position hence the respondents will be stratified according to their organizational status, namely, managerial, supervisory and site workforce. At least 100 samples from each of these organizational levels will be obtained to ensure a valid analysis. 2.3

Structured interviews

Ten structured interviews will be conducted, including both large and small scale contractors which undertake RMAA works. Ten questions have been set to capture experienced practitioners’ perceptions on the causes of accidents, current safety practices of RMAA works and their implementation obstacles. These questions are in place to achieve the primary purpose of getting a holistic view of the safety practices in the RMAA sector. Interviewees will be asked to assess the safety climate of their company on RMAA works by giving ratings from 1 to 5 on seven prescribed safety climate factors. 2.4

Case studies

To gain an in-depth understanding of the causes of accidents in the RMAA sector, fatal accidents in RMAA works will be selected and analyzed. These cases will be analyzed for common features, characteristics, failures and problems. Major factors studied include: – Time and date of accident—to detect if there is any noticeable pattern in terms of time, day of a week, month and season that RMAA accidents may happen. – Age of victim—to detect if a certain age group is more prone to RMAA accidents. Other demographics such as gender, cultural origin; i.e. Mainland China/ Nepal, level of education/ literacy will also be studied. – Trade of worker—to detect if there are any traceable trades which are more prone to RMAA accidents. – Length of experience—to detect if the length of experience affects RMAA accidents. – Place of accident—to study whether there are any common locations resulting in RMAA accidents. – Agency involved—to detect if there are any agents such as certain types of tools or technical processes which are more likely to be involved in RMAA accidents. – Type of work involved—to detect if there are any types of work which are more prone to RMAA accidents. – Unsafe condition/action—to study whether there are any common unsafe conditions/actions resulting in RMAA accidents.

– Safety education and training—to trace whether the victims have received any safety education and training prior to RMAA accidents. – Use of safety equipment—to determine whether the victims have used any safety equipment in RMAA accidents. – Employment terms and conditions—to identify whether the mode of employment (e.g. permanent, day or hourly labour) affects RMAA accidents. 2.5

Focus group meetings

Focus group meetings will be organized to solicit RMAA safety related information from a much wider audience. This is particularly important for RMAA works because most of these projects are carried out by contractors of small to medium scale. Vaughn et al. (1996) asserted that focus groups should possess two core elements: (1) a trained moderator who sets the stage with prepared questions or an interview guide; and (2) the goal of eliciting participants’ feelings, attitudes and perceptions about a selected meeting. At least four focus group meetings will be organized in this study: two for the RMAA supervisory and management staff, and two for front-line RMAA workers. Each focus group will be moderated by a research team member, who will first make a short presentation on the RMAA accident statistics and setting the stage with prepared questions. The participants will then be divided into groups of six to ten to sustain lively and active discussion. Other research team members will be assigned to sit in each group to trigger discussions and responses. Each group will then be asked to nominate a rapporteur to present the group’s findings and recommendations at the end of the meeting.

3

DATA ANALYSIS

Advanced statistical techniques such as Factor Analysis and Multiple Linear Regression Analysis will be applied to analyze the raw data obtained from the questionnaire survey to identify the impacts of safety climate on safety performance in the RMAA sector. Factor Analysis is a statistical technique used to identify a relatively small number of factors that can be used to represent relationships among sets of many interrelated variables (Hair et al. 2006). Principal factor extraction with promax rotation will be performed through SPSS FACTOR program on the safety climate attributes. The purpose of this test is to compute the degree of item loading on their corresponding factors. The chi-square test statistic of Bartlett’s Test of Sphericity will also be performed to test whether the correlations are significantly different from zero.

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Multiple Linear Regression (MLR) Analysis is considered as the most suitable technique to derive the relationships between the variables (Hair et al. 2006). The relative influence of the safety climate factors on the safety performance of RMAA projects extracted from factor analysis will be explored by multiple stepwise regression analysis. A stepwise selection procedure with a significance level of 5% will be used to select statistically significant variables to be incorporated into the model. Data variables will be added one at a time and the regression model re-run noting the changes at each step in the coefficient of determination (R2 ) value and, more importantly, the significance level of variables. Only those variables with a P-value of less than 10% will be included in the final regression equations. The coefficient (R2 ) indicates how much variation in the dependent variable is explained by a group of independent variables; and the higher its value, the more powerful the model. Adjusted R2 , which attempts to more realistically reflect the goodness of fit of the model, will be used to compare and identify the best-fit regression model. 4

VALIDATION OF THE RESULTS

Research data and analyses will be triangulated from multiple sources to help improve the credibility of the findings. Results derived from the questionnaire survey, structured interviews, focus group meetings and case studies will be cross-referenced to complement each other. Workshops will be used to discuss (preliminary) conclusions with industry practitioners involved in the study to help understand the relevance of the findings in the context of changing circumstances over the period studied. Three workshop sessions have been included in the research programme. 5

CONCLUSIONS

To conclude, this paper has briefly illustrated the research framework of how to investigate the relationship between safety climate and safety performance of the RMAA sector. The research team expects that RMAA works will become more and more important to the local construction industry. Their safety problems have to be tackled if we would like to see a continuous improvement in safety performance of the construction industry. Previous research, existing safety management systems and legislation, and plateauing reductions in injury and fatality rates, suggest that we can do little else to drive down accident rates. It is proposed that developing a positive safety climate is one of the effective ways to improve safety performance. A positive safety climate should lead to increasing safe behaviors and in turn improve safety

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performance. In this study, this proposition will be tested by looking at whether safety climate affects people’s safe behaviors and then propose effective safety measures and recommendations to the industry for consideration of implementation.

ACKNOWLEDGEMENT The work described in this paper is fully supported by a grant from the Research Grants Council of the Hong Kong Special Administrative Region, China (RGC Project Ref. No. B-Q05M)

REFERENCES Census and Statistics Department, HKSAR. 2008a. Gross Domestic Product (GDP) by Economic Activity— Percentage Contribution to GDP at Current Factor Cost (Table 036). Census and Statistics Department, HKSAR. 2008b. Report on 2006 Annual Survey of Building, Construction and Real Estate Sectors. Census and Statistics Department, HKSAR Government. Development Bureau, HKSAR. 2008. Immediate measures proposed to assist construction industry. Publications and Press Releases on 25 November 2008. Griffin, M.A. & Neal, A. 2000. Perceptions of safety at work: A framework for linking safety climate to safety performance, knowledge, and motivation. Journal of Occupational Health Psychology 5(3): 347–358. Hair, J.F., Anderson, R.E., Tatham, R.L. & Black, W.C. 2006. Multivariate Data Analysis. Upper Saddle River: Prentice Hall. Hong Kong Construction Industry Review Committee. 2001. Construct for excellence. Report of the Construction Industry Review Committee, Hong Kong SAR. Hox, J.J., de Leeuw, E.D. & Dillman, D.A. 2008. The cornerstones of survey research. In E.D. de Leeuw, J.J. Hox & D.A. Dillman (eds), International Handbook of Survey Methodology: 1–17. USA: Taylor and Francis Group. Labour Department, HKSAR. 2008a. Occupational Safety and Health Statistics 2007. Occupational Safety and Health Branch, Labour Department, HKSAR Government. Labour Department, HKSAR. 2008b. Accidents in the Construction Industry of Hong Kong (1998–2007). Occupational Safety and Health Branch, Labour Department, HKSAR Government. Mearns, K. Whitaker, S.M. and Flin, R. 2001. Benchmarking safety climate in hazardous environments: A longitudinal, interorganizational approach. Risk Analysis 21(4): 771–786. Mearns, K. Whitaker, S.M. and Flin, R. 2003. Safety climate, safety management practice and safety performance in offshore environments. Safety Science 41(8): 641–680. Occupational Safety and Health Council, HKSAR. 2008. Construction Industry Safety Climate Index Software.

The Standard (14 January 2009). Hong Kong, Kai Tak takes off. Vaughn, S., Schumm, J.S. & Sinagub, J. 1996. Focus Group Interviews in Education and Psychology. United States: Sage Publications. WorkCover, New South Wales. 2007. New South Wales Workers Compensation Statistical Bulletin 2006/07.

Zohar, D. 1980. Safety climate in industrial ogranizations: Theoretical and applied implications. Journal of Applied Psychology 65(1): 96–102. Zohar, D. 2003. Safety climate: Conceptual and measurement issues. In J.C. Quick & L.E. Tetrick (eds), Handbook of Occupational Health Psychology: 123–142. Washington, D.C.: American Psychological Association.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Challenges of a substation and infrastructure upgrade in an urban downtown setting M.L. Cochrane Stantec Consulting, Edmonton, AB, Canada

C.D. Wagner EPCOR, Edmonton, AB, Canada

ABSTRACT: EPCOR, along with the design consultant, Stantec Consulting, faced a number of challenges when taking on a major infrastructure upgrade project for the supply of power to the downtown area of the City of Edmonton. During the design phase, these challenges included the high price of land, severe space constraints, designing around existing underground infrastructure, and working with future city infrastructure upgrade plans. Construction presented more barriers including field design changes due to existing conditions, managing numerous scheduling restrictions, managing multiple contractors in a very small working area, public consultation, safety for work occurring in a highly populated area, and sequencing of work to minimize interruption to existing service while relocating and replacing existing equipment. Although the list of challenges was long, the project team was able to deliver a successful project by planning for anticipated challenges and being flexible enough to adapt the execution plan as new obstacles arose. 1

presented unique challenges to this infrastructure upgrade in a downtown setting:

INTRODUCTION

The project included modifications to an existing substation and the expansion of a second, consisting of a 240 kV/72 kV 450 MVA transformer, new control building, and relocation of two 72 kV High Pressure Fluid-Filled (HPFF) underground transmission cables installed in a restricted footprint. A 10.5 km, 480 MVA, 240 kV underground Cross-Linked PolyEthylene (XLPE) transmission cable system was also constructed between the two stations. The cables are installed in a concrete encased ductbank from north Edmonton to downtown. This project was requested by the Alberta Electrical System Operator (AESO) and assigned to EPCOR to build and modify the existing facilities as required. There are three main driving factors behind this project: – Improve the power supply to the downtown core and safeguard the power grid – Allow a downtown generating station to be decommissioned – Expand the available power supplied due to load growth in the downtown core. Many of the challenges faced on this project spanned both design and construction phases, with initial management of issues during the design stage and further management and adaptation during the construction stage. Following are the key areas that

777

1. 2. 3. 4. 5. 6. 7.

Space constraints Public consultation Project planning and staffing of project team City planning Underground unknowns Sequencing of work Field changes and design change support

2

SPACE CONSTRAINTS

Many alternative locations were considered by the AESO when selecting the site of the substation to be expanded. The downtown substation selected for expansion is located on three lots in an older area that has been neglected, but is currently undergoing a revitalization and a construction boom with numerous high rise condominium developments in the surrounding area. Land prices jumped considerably in the year prior to project commencement. Originally two empty city lots were purchased across the street from the existing substation. Public consultation and complex engineering requirements led to the purchase of two additional lots adjacent to the existing substation. Buildings located on these lots were demolished prior to the start of construction. The two extra lots were used for construction laydown and storage, site trailers, and wash cars. They

EPCOR’s public consultation team took a proactive approach to dealing with public inquiries. A preplanning session helped determine some questions from the public. Stakeholder meetings were conducted in person with a project team member and the major stakeholders, as identified in the pre-planning session. Key stakeholders included area hospitals, schools, businesses, and community leagues. Two open houses were held for the public to address concerns and questions. In addition, the public consultation group set up mailings, handouts, regular project updates, signs along the route, inquiry hotline, and a web site. Transparent planning facilitated smooth consultation without objections from the public.

would have been an asset throughout the project to review documents and respond to questions from the consultants and contractors, especially where timely responses were directly related to maintaining the project schedule. As an example, the responsibility of material handling on site was not assigned to a project team member; therefore, the laydown area was accessible to all contractors and was not controlled. While this did not raise significant issues, assigning this role to an individual as a part time duty could have increased the efficiency of construction activities. Construction site staff completed a daily work log. While all staff contributed to this useful reference tool throughout the project, having an electronic log would have been an improvement to this process, allowing for easier referencing and information searches in the future. Weekly status reports were completed for the project manager to pass on to senior management. Numerous photos were taken throughout the project stages for documentation. These photos were used for as-built purposes, financial discussions, safety discussions, and engineering support. The photos proved to be a valuable resource; however, they could have been more useful if a detailed filing system had been developed to log the thousands of photos. The Province of Alberta and the City of Edmonton experienced an economic boom due to the price of oil at the time of this project. Contractors had difficulty staffing well-trained and qualified personnel in the specialized fields required for this project. This presented challenges in maintaining the schedule and ensuring quality standards were met. In one case, a contractor was awarded two separate contracts for different aspects of the project. During construction, the contractor was found to be competing within their organization for resources and experienced personnel to complete each contract.

4

5

were then sold at project completion. An additional lot was purchased adjacent to the two new construction lots part way through construction for access and laydown, then converted to a permanent parking lot for employee access to the substation, and is also available for future expansion. During design, special consideration was required to ensure maintenance of minimum electrical clearances, while minimizing the overall footprint of the station to fit into the two compact lots. Alternatives to the final design such as a gas insulated indoor station were considered, but discarded due to cost. A solid dielectric bus was used instead of traditional underground cable bus to accommodate the small space and existing underground duct banks. A 5 m tall brick wall around the substation enabled equipment to be placed closer to the perimeter and also acted as a sound and safety barrier. The final equipment layout design considered future operations and maintenance activities to maximize access. 3

PUBLIC CONSULTATION

PROJECT PLANNING AND STAFFING OF PROJECT TEAM

The owner’s main project team was formed in June of 2005, and consisted of a project manager, substation project coordinator, cable project coordinator, finance, document control, purchasing, scheduling, regulatory affairs, and public consultation groups. This team evolved as the project progressed and members from safety, environmental, legal, and engineering support became involved. When the project progressed into construction, a construction manager, quality assurance supervisor, safety supervisors, construction engineers, and site administration were also added. The size of the team was small for the scale of the project. Clearly defined roles for the project team were required early on to avoid duplication of effort and to ensure nothing was overlooked. Additional staff

CITY PLANNING

EPCOR anticipated that permitting and City approvals would be extensive, and began coordinating and planning with the City from the very early stages of the project. Meetings began prior to the design contract being awarded to determine initial requirements and to map out the best plan for working with the City. Preliminary planning involved determining what permits and City review stages were required and developing key contacts with various City departments. Early consultation with the City also allowed them to share their plans for the expansion of the Light Rail Transit (LRT) system, which resulted in changes to the planned transmission line route. Initial consultation with the City revealed that the installation methods used for the construction of the underground duct line would have to be considered in

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the design stage of the project. Selection of the exact routing of the transmission line required evaluation of the construction obstacles that would be faced, as well as the impact of construction on the surrounding neighborhoods. The route selected minimized disruption to roadways by constructing several kilometers of the transmission line through an old rail line that is now used as a multi-use recreation corridor. Careful selection of the route still posed construction challenges. It was identified that one construction method could not be used to construct the entire 10.5 km route. The majority of the transmission line route was constructed using open trenching to install the concrete encased duct bank. At two roadway crossing locations, the City would not allow the road to be open trenched due to the high traffic volume. At these locations, the jack and bore method was used to cross under the roadway without impact to traffic. In another location, the transmission line had to cross under a large railway yard and a major arterial highway that runs through Edmonton. Again, in this case, open trenching was not an option. At this location, a 520 m Horizontal Directional Drill (HDD) was used to complete the crossing. In both the jack and bore and HDD installations, a casing pipe was first installed, the duct line conduit was then pulled through the casing pipe, and finally the casing pipe was filled with concrete. In Edmonton, a Utility Line Assignment (ULA) is required for all newly constructed utility routes. In order to start construction without delay, it was determined what information was needed for ULA submission and the drawings were prepared before the Issued for Construction (IFC) package. This allowed the ULA permits to be applied for in advance of the IFC drawings being issued, giving time for ULA review and approval to be completed in advance of the scheduled start of construction. Late in the ULA process it was identified that the location of the planned LRT had recently changed, resulting in a conflict with the planned routing of the underground transmission line. Stantec and EPCOR had to work quickly to revise the design and move a segment of the underground transmission cable route from one side of a roadway to the opposite side. This resulted in a significant effort to re-issue drawings and then re-submit them for ULA approval. Due to the contacts that were developed early on with the City and the contractor’s plans to not commence construction at this location until a later date, the revision time was minimal and did not impact the overall schedule of the construction project, as well as eliminated construction rework. Due to ongoing redevelopment of the area surrounding the substation, plans were taken to a City redevelopment committee. Long term closures of the road and alley adjacent to the substation also required special permits and planning to ensure scheduling was acceptable.

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6

UNDERGROUND UNKNOWNS

In an urban environment, there are often a variety of utility services including water, sewer, drainage, power distribution, natural gas, telephone cable, and television cable running underground. These services are found at different locations and at different elevations. As well, each utility has a required minimum clearance that must be maintained—no other underground facility may be constructed within a minimum distance from an existing underground utility. There are clearance requirements for both separation of a line crossing, as well as the separation required between parallel lines. Unlike above ground services such as overhead power lines, one cannot easily determine the exact location of underground facilities. In this case, City records of the location of all utilities must be relied upon. Some locations can be verified by measurements taken at manholes or other access points, however, these verifications are limited to point locations only and do not verify elevations of the entire line. Not all records provided complete data, and in many cases, exact elevations at a particular point where a utility was crossing the duct line were unknown, so elevations had to be extrapolated between two known points. The underground transmission line needed to bend under and over a maze of existing lines; however, the transmission cable could not twist and bend at the will of the route designer. Due to the size of the cable, the design required careful consideration of the minimum bend radius and the maximum pulling tension of the cable allowed between vaults. Even with the possibility of snaking the route between existing lines, the cable would not be able to bend through a duct line route with numerous horizontal or vertical bends. In addition, the cable also required a minimum depth to ensure there was enough ground cover to dissipate the heat that will be generated by the operating cable. All design clearances and avoiding conflicts with an existing underground utility were based on the data in the City records. This design was IFC to the contractor without verification of the actual locations of existing underground utilities. Upon award of the construction work, the contractor was informed of the method by which the design was produced and advised of the risk of error, inaccuracy, or lack of information in the City records. At the commencement of construction, the contractor located and exposed utility crossings only as needed for construction. This approach proved to be inefficient due to differing elevations between actual and provided records. The difference in information resulted in cases where construction was delayed to determine alternate routing, and in one instance, part of the already constructed duct line had to be removed to adjust the duct location.

After seeing these delay issues arise very early in the project, Stantec, EPCOR, and the contractor worked together to determine how to minimize delays and rework. It was determined that the contractor would work ahead with an additional crew to physically locate all utility crossings on the duct line route. Locations were discovered using a hydrovac truck and survey crew to provide the detailed, accurate measurements. Working ahead to determine exact locations proved to be an excellent decision for the progress of construction. In some cases, major differences in elevations were found, which resulted in significant design effort being required to re-plan the cable routing in these locations. By having the information ahead of time, the field support staff and design staff were able to evaluate options and prepare a new design without delaying construction. This project would have benefited further if the work to determine exact locations and elevations at all utility crossings was done prior to the design stage. Accurate design data would have resulted in IFC drawings that required significantly less re-work during the construction phase and less updating effort during the as-built stage. 7

SEQUENCING OF WORK

The underground duct line was constructed with three or more crews working on different areas of the route. Much of the route was constructed along roadways that required lane closures for construction. The City traffic standards outline regulations on allowable road closure times including limitations during peak traffic hours (morning and afternoon rush hours), evenings, and weekends. In addition, a number of events occurring in close proximity to the areas of construction imposed restrictions on allowable dates and times of road closures. To ensure the contractor performing construction of the underground duct line could plan appropriately for the allowable road closure dates and times, this information was collected during the design phase of the project and provided in the technical specifications of the duct line construction package. This information was collected through meetings with various groups at the City to ensure closure information covered both regular road closure regulations, as well as coordination with City construction activities and special events. During construction, further meetings and consultations were necessary to review requirements and establish plans for specific closures for each location. Permitting and signage requirements had to be considered several days to weeks ahead of when a closure was required. This required special attention by the construction contractor to ensure that all work crews’ sequencing of work was well planned and that alternate

options were available to ensure work crews were not stalled by limitations set by the various road closure constraints. Signage for the road closures was coordinated with the City. Clear marking of construction zones were important in slowing traffic to protect the workers and the public. At Victoria Terminal Substation, the space to construct the substation was very limited and presented many design difficulties in trying to fit the required equipment onto two city lots. The underground space constraints presented even more impediments to determining how to fit a building basement, transformer foundation, containment pit, numerous piles, drainage system, underground cable trays, the new transmission line duct line, and two relocated existing underground transmission cables, all under two city lots. The transformer foundation and building foundation were constructed first. Then the piles were installed. After piling was complete, construction of the underground transmission line duct bank was started. Due to the minimum bend radius of the cables, the duct line in the substation was approximately five meters deep to allow the cable to make the 90 degree bend from running underground to rising vertically out of the ground. Excavation of the duct line to five meters would be challenging in any location, but as the duct bank ran very close to a number of piles and the transformer foundation in the substation yard, the task was even more complex. To ensure that the integrity of the piles were not compromised during the duct bank excavation, a series of steel rods were designed to span between piles, connecting to piles that were not exposed in the duct bank excavation. This ensured that the excavated piles did not move in any direction, which was critical in ensuring clearance requirements for electrical equipment was maintained. After construction of the duct bank, backfill techniques had to be evaluated to ensure the effectiveness and location of the piles was not compromised. Traditional compaction techniques could result in damage and movement of the piles. In some locations, fillcrete (a low strength concrete) was used as it does not require compaction and has suitable friction properties. In other locations, backfill was compacted with a smaller compaction device to minimize vibrations around the piles. Once the site was backfilled to a level where the piles were considered stable, standard backfill and compaction techniques were used. All work inside of EPCOR’s energized substations must be performed under qualified safety supervision. Chain link fencing on concrete jersey barriers was used to separate the new construction site from the existing energized yard. This reduced the amount of supervision required and allowed the contractors to work without restriction in the areas of new construction. This fence was easy to move and reconfigurable to accommodate construction access.

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It was intended that the construction of the substation was to be completed in several stages. The civil construction was to be completed prior to the general electrical contractor being on site. Due to construction delays, there were two main contractors and multiple sub-contractors on site at the same time. Construction coordination meetings were formally held weekly and informally on a daily basis to determine work fronts, discuss safety on a small work space, and minimize conflict on work zones. Due to the small construction zone, the contractors often shared construction equipment such as back hoes and manlifts. Prior to starting some of the more complex construction activities, constructability reviews were held to discuss sequence of work, safe work planning, and equipment and materials required. These reviews could have been more beneficial if they were held earlier in the project and more frequently to ensure designs were constructible and all materials were available. 8

FIELD CHANGE AND DESIGN CHANGE SUPPORT

It is impossible to eliminate all changes during construction, so proper planning to manage changes was critical. Changes during construction can range from a design modification due to material or equipment substitution, changes to construction methodology, opportunities for cost savings, and encountering unexpected field conditions. When a situation arises where a change is required, it can often lead to review and re-design by the design consultant or the project engineering support team working in the field. Typical process for issue management would begin by a construction crew identifying an issue, notifying the field engineering support team, review by the field engineering support team, review and reporting of the issue to the design consultant, return of a revised design from the consultant, and the field engineering team communicating this change back to the construction contractor. This process is typical and straightforward. The challenge is managing this process to ensure all changes are properly documented and recorded, and ensuring that the process occurs in a timely manner to minimize delays in construction. To ensure information was clear and properly documented, a simple Field Change Notice (FCN) system was developed. The system was based on a single page form that documented issues and provided instructions on changes required to address the issue. The FCN system was effective because it was managed within a clearly defined process for reporting, review, and approval of all information. This ensured that the critical information in each FCN was communicated to

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the right people in a time efficient manner. The system was not complex and was easily developed; it worked because of its simplicity. For the underground transmission line, Stantec provided a field engineer to support EPCOR’s construction management team. The engineer worked out of the construction management office and was in the field daily to respond to issues arising during construction. Having this resource in the field allowed EPCOR and Stantec to react quickly to issues that arose during construction. In many instances, the field engineer could respond to an issue and work with EPCOR’s engineer to resolve it without delay, ensuring proper FCN documentation was generated to record the change. In other cases, where the field engineer was not able to resolve the issue on site, she was able to immediately access the design team at Stantec to determine a resolution. This direct link to the design team eliminated a step in the communication chain and allowed EPCOR to benefit from minimal delays in construction while issues were being resolved. For the construction of the substations, EPCOR placed one of their own engineers and an Engineer In Training (EIT) in the construction management office to provide a similar field engineering support role. In the case of the substation work, the new construction was being integrated with existing infrastructure, making the presence of EPCOR engineers with knowledge of the existing sites an important element in the field engineers’ positions. Unlike the underground transmission line that was new construction, the substation construction took place beside and integrated with existing equipment. Field support for the substations was managed through a similar process of having the field engineers available and on site to evaluate issues and determine how to manage them. The EPCOR engineer and EIT were both electrical engineers, with extensive experience in substations. However, their experience and knowledge regarding civil and structural construction was limited, given that this was not their area of specialty. EPCOR field engineers quickly determined that the optimal process for handling field construction issues depended upon the nature of the issue. In the case of civil or structural issues, EPCOR field engineers contacted Stantec’s engineers or designers for consultation. Often, Stantec’s personnel would visit the site to review the issue and determine a resolution. In the case of electrical or equipment issues, the EPCOR engineers were often able to resolve the issue on their own, or in the case of requiring consultation from Stantec, they could communicate enough detail that site visits from Stantec’s electrical personnel were often not required. This combined process of addressing issues ensured that adequate experience and knowledge was used to address issues, while also ensuring a problem was resolved as quickly as possible.

Where design changes were required to address issues encountered in the field, sketches were used, where appropriate, to communicate the changes. A design change may require new information on dimensions or other information that requires a drawing to clearly communicate this information. To revise and re-issue a drawing that has been IFC, the process can be slow and is not conducive to the quick turn around required to minimize delays in construction. The processes of quality review, stamping, and permitting a drawing takes time and involves a number of people, regardless of the change made. By issuing sketches instead of revising issued drawings, Stantec and EPCOR were able to work together to ensure design changes were clearly communicated, but communicated in the most efficient time possible. The sketches were used only where appropriate, but provided a valuable system for making changes and then tracking these changes to ensure the information was included during the red-line mark up of construction drawings. For the substations, IFC drawings were issued at various stages to allow construction to begin without delays. Given the staged issuance of drawings, careful attention was required to ensure that the correct revisions of drawings were being used at the time of construction. It was determined that the contractors often did not issue new revisions of the drawings to their field crews. This was managed by the owner’s on site quality assurance supervisor continually inspecting the drawings on site. If the revision was not correct, the contractor was ordered to stop construction until the proper revision was produced.

9

CONCLUSIONS

The new substation and underground transmission cable were successfully energized in October 2008, meeting an important milestone in the successful completion of the project. As required by the AESO, the project was completed on schedule and within budget. Meeting the schedule and budget commitments on this project was a key measure of success for the project. Despite a long list of challenges, the project team was able to deliver a successful project by planning for anticipated challenges and being flexible enough to adapt the project execution plan as new obstacles arose. During the design phase, the team worked to find alternative design methods and equipment options to work within the limited space of the new substation. Construction changes were addressed with up front planning to anticipate problems and deal with them early to reduce delays. Some of the lessons learned on this project were gained through careful planning and anticipation. In these cases, measures to deal with these challenges were put in place early. However, no matter how well any project team plans for a project, unexpected challenges are always going to arise and the true measure of a successful project is the ability for that project team to be flexible enough to manage these issues as they arise and find solutions that keep the project on track.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Construction of concrete embedded, direct fixation, ballasted, LVT and special trackwork K.H. Dunne, N. Slama & K. Wong URS/Washington Division, USA

ABSTRACT: This Paper addresses the Owner/Contractor design/build interfaces necessary to complete the four different types of trackwork required to design and build the most cost effective and productive trackway system on the 6.2 mile Eastside Extension of the LA Metro Gold Line LRT System. The trackwork encompasses 2.2 mile long high attenuation low profile Low Vibration Track (LVT Blocks) in the 21 –0 diameter TBM bored tunnels under East Los Angeles, Embedded Track on 3.5 miles of surface trackway, and 0.5 miles of Ballasted Track at the western termination of the Eastside Extension at Union Station in Los Angeles. 1

INTRODUCTION

The Eastside Light Rail Extension of the Gold line is a $624-million design-build lump sum project. This project was awarded to Eastside Light Rail Transit Constructors (ELRTC), a joint venture between URS Washington Division and Obayashi Corporation, in 2004 to construct the six-mile-long project. It is an extension of the existing Pasadena Gold Line, which runs 13.7 miles from Pasadena to Union Station in downtown Los Angeles. The scope of this project consists of many facets of civil and structural engineering design and construction work throughout the project alignment. Utility relocation and street improvements, retaining walls and MSE walls, retrofit of the I-710 bridge and 1st Street Bridge, construction of six at-grade stations, placing over 60,000 c.y. of concrete for two underground stations, 2.2 miles of 21-ft-diamater TBM bored twin tunnels, and laying out a total of 12.4 miles of continuously welded rail (CWR). 2

TRACKWORK

This project is unique because it utilizes four different types of trackwork to design and build the most cost effective and productive trackway system on the Eastside Extension. The different trackwork types consist of 0.3 miles of Ballasted, 0.2 miles of Direct Fixation, 2.2 miles of Low Profile Low Vibration (LVT), and 3.5 miles of embedded trackwork. 2.1 Ballasted trackwork Ballasted trackwork is the most common type of trackwork that comes to people’s minds when railroads are mentioned (Fig. 1).

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Figure 1. wall.

Ballasted trackwork—timber ties along MSE

The principle of ballasted track is to restrict rail movement by fastening both left and right rail to a single tie, concrete or timber, and locking the tie in place by compacting ballast rock around it with the use of a tamper machine. The construction process on the Eastside Extension consists of the following procedures: 1. The existing grade is fine graded to 2 –10 5/8 below top of rail and compacted to 90%. The reason for 2 –10 5/8 below top of rail is to allow room for the 8 of sub-ballast, 12 of ballast, the height of the concrete/timber tie, elastomeric bearing pad and 115RE rail. 2. Once the area is graded the overhead cantenary system (OCS) cast-in-drill-hole (CIDH) foundation is placed. 3. With the foundations placed, the train control system duct-bank, and track drains are installed. 4. After the underground utilities have been installed, 8 of sub-ballast is placed top of the sub grade in

Figure 2. wall.

5.

6. 7. 8. 9. 10.

Figure 3.

Direct fixation—parts and materials.

Figure 4.

Direct fixation—formwork.

Ballasted trackwork—concrete ties along MSE

4 lifts and compacted to 95% density. The subballast consists of a fine crush aggregate base (CAB). After the sub-ballast meets specification from quality control testing, 10 of ballast rock is placed top of the sub-ballast in 4 lifts and compacted until no further movement of the ballast is evident. For this project, American Railroad Engineering and Maintenance-Of-Way Association (AREMA), No. 4 ballast (1 clean crushed rock) was used. With the Ballast in place, the concrete ties are distributed along the alignment at 30 o.c. spacing. Next, two ‘strings’ of Continuously Welded Rail (CWR) is pulled over the ties and fastened on with the use of special Pandrol E-clips. Elastomeric bearing pads and shoulder insulators keep the rail isolated and abate stray current. With the CWR’s firmly attached to each tie, the track is then raised using track jacks to its final alignment. Once the rail is set to its proper alignment, the track is flooded with ballast. With the use of a tamper machine we can push and compact the ballast to consolidate the ballast completely around the tie. After this is complete, the track jacks are removed. With the rail in place a regulator will then come through to dress the ballast shoulders.

2.2 Direct Fixation (DF) trackwork Direct Fixation known as DF track is mainly found on aerial structures due to its relatively light dead load and vibration attenuation. The principle of DF track is that each individual CWR is clipped to a DF fastener that is then anchored to a concrete plinth that restricts the rail from movement. The following construction procedures were used by ELRTC for the construction of the Direct Fixation Trackwork on existing plinths. 1. The rail is set to proper alignment, gauge dimension (the distance between the inside of the two parallel

rails), and super-elevation using Iron Horse Gauge Support Fixtures (GSF) spaced on 10 − 0 centers. 2. With the rail aligned and suspended, DF fasteners are clipped to the rail at 30 spacing on center using temporary retainer clips. Slobber plates, a plastic plate with slotted holes, are installed underneath the fasteners to mitigate air pockets from forming due to improper consolidation when concrete is placed (see Figure 3). 3. With the DF pads clipped, slobber plates are installed, and anchor inserts firmly attached to the rail. Then, iron workers begin to tie rebar for the concrete plinth underneath, and carpenters begin to build their formwork (see Figure 4). The width of the plinth varies at different locations. The tangent track width is 2 –2 , and the super elevated curve plinths widen out to 4 –0 . The height of the concrete plinth is determined by the profile of the rail. Once the rebar is installed and formwork constructed the concrete is now ready to be placed. Concrete placement for rail is usually scheduled from 10:00 am to 11:00 am to ensure there is no movement from expansion due to change in temperature. During this process the rail is constantly monitored for alignment, gauge dimension, and cross-level. After the concrete is cured for 3 hrs, the temporary retaining clips are loosen to prevent rail movement throughout the remainder of the day so that the anchor inserts are not disturbed while in the concrete. Drainage chases were constructed at every 20 –0 along the plinths.

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Figure 5.

Direct fixation—I-101 bridge aerial structure.

Figure 6.

Embedded track—installation.

Figure 7. bridge.

Embedded track–finish product on 1st street

4. After the concrete has reached its designed strength, the Iron Horse GSF’s, slobber plates, and formwork is removed so that only the anchor bolt insert are embedded in concrete. If air pockets greater than 5% of the entire bearing surface are present, concrete patchwork is applied to eliminate the voids. 5. Finally, the rail is lowered onto the DF fastener and special Pandrol E-clips along with nylon insulators installed to prevent any stray current from bleeding into the ground. The use of DF needs extensive formwork, and must consider the time restraint between placing concrete, curing time and loosening the temporary retaining clips from the rail. This process can incur an increase in man-hour/cost to construct. In fact ELRTC executed an engineering change order to switch over to LVT trackwork inside the twin 2.2-mile long tunnel segments originally designed for the DF system. 2.3

Embedded trackwork

Embedded trackwork is an excellent choice for LRT systems, given the common R.O.W. constraints found on many Projects, considering traffic control required. The concept behind embedded track is that both running rails are cast-in-place in concrete restricting the rail from any movement. 1. The guideway is excavated to a depth of 2 –0 below top of rail and the sub-grade is compacted to 90%. The reason for this depth is to allow for 6 of crushed material base (CMB), 11 of 1st lift invert concrete, and 7 of 2nd lift colored concrete, in the case of Eastside. 2. After the trackway alignment is graded, the overhead catenary system foundations are installed. 3. Next, a 6 lift of CMB is placed on top of the subgrade and compacted to 95% density. 4. Once the CMB has been compacted to meet specifications, 2 layers of welded wire fabric (WWF) reinforcement are installed.

5. With the WWF installed, the CWR is roughly placed into position on wood blocks or a minimum number of steel ties (see Fig. 6). 6. A rail boot is placed around the rail and taped to the CWR to keep the rail electrically isolated from stray current. 7. Embedded track ties are attached to the rail every 10 as a construction aide to get the rails to the correct alignment and gage cross-level. 8. Formwork is then set on both sides to create a 7 −6 track guideway, with construction joints installed every 300 –0 . 9. 4 concrete dobes are installed to separate the 2 layers of WWF. 10. A high voltage electrical test (holiday test) is scheduled to test the rail boot for any punctures or any discontinuities. If a hole is found, the boot is patched and retested. 11. The 1st lift of concrete invert is placed a thickness of 11 , just below the rail. As the concrete is placed, the rail is surveyed and checked for horizontal and vertical alignment, as well as the gage dimension. 12. During concrete curing, the gauge dimension and cross-level are checked again. 13. When the rail is in final alignment, rail tape is used to prevent any debris from entering the rail boot.

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A wooden flange way form is attached to the head of the rail to block out concrete to create the flange way area. Embedded Track is expensive, but was chosen for the Eastside Project for aesthetics, and the ability for pedestrian vehicles to traverse across intersections with the track running parallel and perpendicular to the street. The finish trackwork provides an aesthetic appearance without electrical isolation equipment on both sides of the rail. 2.4

Low Profile Vibration Track (LVT)

Figure 8.

LVT trackwork—concrete placement.

Figure 9.

LVT trackwork—before.

Figure 10.

LVT trackwork—after.

LVT was used inside tunnels, primarily because of its ease of construction in confined work areas. Individual LVT blocks are clipped on the CWR strings with rail fasteners, and set on the tunnel invert. The gage and elevation are checked and surveyed for design compliance, and then concrete is placed. The trackwork construction process on Eastside consists of the following procedures after the tunnels had already been constructed, and As-Built drawings and survey data collected, during the first phase of the Contract. 1. Strings of CWR are pulled on rail rollers and set inside the length of the entire tunnel. 2. Next, LVT blocks are transported inside on a LowRailer, and distributed every 30 spaced on center along with elastomeric bearing pads that seat between the base of the rail and the LVT block. Two types of LVT blocks are used, depending upon the area, if noise level requirements are present. At the tunnel transitions to the underground station platforms, High Attenuation (HA) LVT blocks were used. These blocks are 18-mm wider than the standard 260-mm. 3. With the use of a rail threader, entire lengths of rail can be lifted and set on top of the LVT blocks with the proper gage dimension. 4. Once the rail is seated on the blocks, Pandrol E clip rail fasteners and shoulder insulators are installed to prevent stray current from bleeding into the concrete and keep the rail electrically isolated. 5. Iron horse ties are a type of rail leveling beam, installed at every 10 –0 . These are used to raise the rail to its final position, gage dimension, and super-elevation. 6. In order to place the monolithic concrete, and allow access to remove the iron horse ties since they are not a permanent installed material, ELRTC had to decide on the constructability of the two methods, and which is the most cost efficient. 7. The first method is called ‘‘cow-pie’’. This process is used to eliminate the use of extensive form work by placing concrete around and underneath individual blocks to create a bearing surface tain

proper horizontal and vertical alignment. The second method is to build wood formwork around the iron horses—this is to prevent and block-out concrete so that the iron horses can be removed at a later time once the monolithic concrete slab has been placed. The second method introduced the need to go back, once the iron horses were removed, to place concrete to close the block-outs. ELRTC decided on the second method. 8. With the block-out forms in place, the tunnel was then flooded with concrete to the top of the rubber LVT boot.

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9. Once concrete has cured, formwork was stripped and moved ahead to the next 1000 Track Feet to setup. Similar to embedded track, the LVT reinforced concrete blocks are embedded in concrete to create a homogenous slab but without the need for embedded track parts and materials. The advantages to LVT trackwork is the flexibility and option to choose between LVT and HA-LVT in areas sensitive to noise and vibration. The use of LVT also meant that no steel components were to be used that may corrode overtime. The main advantage is the high production rate—ELRTC on average constructed approximately 1000 track feet per 10 hr shift. 3

Figure 11.

DF crossover—installation.

Figure 12.

DF crossover—finish product.

SPECIAL TRACKWORK

The Eastside Extension consist a total of 6 crossovers, two #10 embedded double crossover, a #10 embedded left hand crossover, a #10 embedded right hand crossover, a #8 ballasted double crossover, and a #10 ballasted right hand crossover. The construction method for the ballasted crossover is nearly identical to the ballasted track—the wood ties were pre-plated and assembled before we connected to the existing ballast tracks. Embedding the DF crossover created a problem with regards to corrosion protection and electrical isolation on the embedded crossovers because a rail boot cannot be installed due to the complexity of the each assembly. To resolve this issue, ELRTC decided to isolate the rail by placing a ballast mat all around the crossover creating a bathtub effect to mitigate stray current bleeding into surrounding area. ELRTC used several methods to construct the crossover. The first is termed ‘‘bottom up construction’’ because the invert slab was placed first. Next, ELRTC cored holes in the concrete and epoxied the DF anchor inserts. The second method is the ‘‘top down construction method’’. It became evident that this method is a lot faster, more cost efficient, and resulted in a better quality product. The construction procedure for the DF embedded crossover is as follows:

6.

7. 8.

9.

1. The area was excavated similar to step one of the embedded track and duct bank installation. 2. A 6 concrete mat was placed to allow for a stable working surface for the ballast mat to be installed. 3. The ballast mat was unrolled and stapled together to create the bathtub effect. 4. Next, WWF (welded wire fabric) reinforcement was installed. The crossover rail is placed on the wire mesh and temporarily jointed together with rail joint bars. 5. W14 × 146 beams were spanned across the excavation perpendicular to the rail at each end of the

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4

earthen box for the switch, and at each frog. With the help of coil rods, the crossover was raised to proper alignment. Embedded ties were used to ensure the rail did not sag as a result of spanning from beam to beam. With the rail set to proper elevation, DF fasteners were attached to the rail. Elastomeric insulating pads was glued the DF fasteners and the attached anchor inserts. Necessary block-outs were installed for train control equipments. With the formwork installed and the rail in final alignment, the 1st lift of the invert concrete is placed to the height of the bottom DF fastener. After the concrete cured, the switch machines, fouling bonding, and switch rods are installed for train control purposes, as well as the flange way formwork. Finally, the 2nd lift of the colored concrete was placed. Once the concrete is cured, forms are stripped, extra lengths of ballast mat cuffed over the crossover was trimmed 1/2 bellow concrete and sealed with a special sealant.

RESTRAINING RAIL

Restraining rail was installed on curves with a radius less than 500 and curves with a radius less than 300 required the rail to be pre-curved at the factory before delivery.

Other specifications for the curved rail include radius between 500 to 250 the gauge dimensions is to be 56 3/3 instead of 56 1/2 at tangent track. The reason behind the widen gauge is to reduce premature wear on the train wheels and prevent derailment. There are 5 locations on the Eastside Extension project that required restraining rail, 3 are in embedded track and the rest on the DF track. 5

The quality of the project has been excellent and the constructor owner relationship has been very open and positive. The biggest accomplishment of the project will have to be its safety record. As of June 2009, the Eastside project has successfully achieved over 4 million man-hours without a lost-time injury accident.

CONCLUSIONS

The Eastside Gold Line Extension was a very successful project completed under budget and delivered on time.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Design issues of the Palmdale Water Reclamation Plant expansion K. Monroe, J. Stanton, P. Wong & S. Maguin County Sanitation Districts of Los Angeles County, USA

ABSTRACT: The Palmdale Water Reclamation Plant is undergoing expansion to provide improved secondary and new tertiary treatment for 45.4 million liters a day of wastewater. Construction will occur between 2008 and the end of 2011, at an approximate cost of $114,000,000. During the design of this expansion a number of issues arose related to code editions, seismic loads, and construction sequencing to accommodate plant down time limitations. 1

INRODUCTION

The County Sanitation Districts of Los Angeles County (Sanitation Districts) are a partnership of 24 independent special districts serving approximately 5.3 million people in Los Angeles County. The Sanitation Districts’ service area covers approximately 2100 square km and encompasses 78 cities and unincorporated territory within the County. The Sanitation Districts construct, operate, and maintain facilities to collect, treat, recycle, and dispose of wastewater and industrial wastes. The Sanitation Districts’ 2200 km of main trunk sewers and 11 wastewater treatment plants convey and treat approximately 1900 million liters per day, 750 million liters a day of which are available for reuse in the dry Southern California climate. The Palmdale Water Reclamation Plant (WRP) was originally constructed in 1953 with a capacity of 2.8 million liters a day. The WRP has undergone a number of incremental expansions over the years and is currently located on two sites in the unincorporated County area adjacent to the City of Palmdale. The current permitted capacity of the treatment plant is 56.8 million liters a day. An aerial photo of the current Palmdale WRP is shown in Figure 1. 2

PURPOSE

California has recently adopted the first new building code in 10 years. The new California Building Code (CBC) is based on the 2006 International Building Code (IBC), while the previous CBC is based on the 1997 Uniform Building Code (UBC). The new CBC became effective for projects starting on January 1, 2008. This project went out for bid on December 27, 2007. Due to the timing of the project, structural calculations for the major hydraulic structures were performed using both codes.

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Figure 1.

The Palmdale WRP.

Also, the project is located 4.2 km north of the San Andres fault. Recently the USGS has predicted that this portion of the fault has a 59% chance of a magnitude 6.7 or larger earthquake within the next 30 years. Water retaining structures are designed for hydrodynamic loads and lateral seismic earth pressures. Results of the side-by-side seismic analyses are presented. The potential for seismically induced settlement of the soil was also encountered. The method used for mitigating this soil settlement is presented. In addition to the Code and seismic issues, the existing water reclamation plant will remain in service during the construction of this project. The project specifications limit plant shut downs to four hours maximum. While this constraint does not affect the majority of the construction, it does affect the upstream and downstream ends of the new project at the tie-in structures. At the downstream end, all of the plant effluent will flow through a new downstream junction structure. The new junction structure will replace an existing downstream manhole, which will be demolished. The new junction structure is needed for two

reasons; first, it must allow flows from the new construction to enter into the existing plant effluent line; and second, the new flows will cause the junction structure to be pressurized during high flow events. The existing manhole was not designed to be pressurized. The sequence of construction for the new junction structure that minizes plant shutdowns is presented.

BUILDING CODE ISSUES

On January 1, 2008, the State of California changed from a building code based on the 1997 UBC to a new building code based on the 2006 IBC. This was the first new building code adopted in a decade in California. This change in building codes results in significant changes in how seismic loads are calculated. Since the Palmdale WRP project went out for bid on December 27, 2007, it could technically still be designed under the previous California building code. However, at the beginning of the project it was decided to perform parallel sets of calculations using both codes for the major water retaining structures. Table 1. Seismic parameters for UBC based calculations. Parameter

Value

Zone Seismic source type Distance to fault Soil profile Importance factor Ca Na Cv Nv

4 A 4.2 km D 1.00 0.56 1.28 1.09 1.71

4

HYDRODYNAMIC LOADS

The water retaining structures are designed for service loads in accordance with ACI 350-06 ‘‘Code Requirements for Environmental Engineering Concrete Structures and Commentary.’’ This is the code that governs the design of water retaining concrete structures. For the determination of Seismic loads, ACI 350.3-01 ‘‘Seismic Design of Liquid-Containing Concrete Structures’’ was used. However, the seismic parameters used in the hydrodynamic calculations differed depending on the Code used. The hydrodynamic calculations of ACI 350.3 are based on the work of Professor George W. Housner, and include the effects of impulsive and convective forces of water motion. See Figure 2. In practice the impulsive and convective forces are combined into a resultant force, which is then used to determine the pressure to be applied in the design of the tank wall. See Figure 3. The Aeration Tanks are a series of parallel tanks with inside dimensions of 8.2 m wide by 79.25 m long. The operating water depth is 4.6 m. Hydrodynamic loads were calculated in both the long and short directions. Table 3 provides the results of the calculations in the long direction of the tank. Table 4 provides the results of the calculations in the short direction tank. The results of the parallel calculations in the long direction of the tanks show little difference in most of

Table 2. Seismic parameters for IBC based calculations.

Undisturbed Water Surface Parameter

Value

Ss S1 Distance to fault Site classification Fa Fv SDS SD1 Seismic use group

1.605 0.885 4.2 km D 1.00 1.50 1.07 0.885 II (Wastewater Treatment Plant) 1.25 E

Oscillating Water Surface

Importance factor Seismic design category

Pc

Wi Figure 2.

790

hc

Wc Pi hi

3

Tables 1 and 2 present the seismic parameters for the UBC and IBC, respectively, used in the calculations. For a one story concrete masonry unit (CMU) building, the base shear calculated from the UBC is 0.31 W. The base shear calculated from the IBC is 0.30 W. This result was typical for many of the structures on the site. The UBC uses near fault factors in calculating the base shear, while the IBC uses contour mapping of acceleration parameters, which account for known faults. It would be surprising if the two codes gave significantly different results for structures in California.

Hydrodynamic model of a liquid containing tank.

the design parameters. For those values which have the largest impact in the design of the structure, P’ and d, there are no significant differences. The actual wall design is based on the UBC values which resulted in similar force, P’ and a dimension d of 152 mm higher than the IBC calculations. In the short direction, the

Definition of variables in Tables 3 and 4. Pc Pi MIBP MEBP

Undisturbed Water Surface Oscillating Water Surface

hc hi

w1

w1 w2 P

P’ d

d

Table 5.

w2 Figure 3. pressure.

Table 3. Comparison of hydrodynamic forces in the long direction of the aeration tanks. UBC

IBC

h37.8 30.2 5399 50 0.042 2.3 1.7 4.4 21.1 70.1 2.1

5.3 37.4 1447 33 0.012 2.3 1.7 2.4 23.2 70.1 2.0

wall design is based on the IBC calculations since the resultant force is higher than the UBC resultant. The calculations for the other water retaining structures had similar results. The UBC and IBC calculations resulted in similar values with the actual design being based on whichever resulted in the most conservative design.

5

Pc (kN) Pi (kN) MIBP (kN-m) MEBP (kN-m) Convective acceleration hc (m) hi (m) w1 (kN/m2 ) w2 (kN/m2 ) P’ (kN/m) d (m)

IBC

15.1 27.6 107 31 0.30 2.7 1.7 3.3 16.3 54.0 2.1

14.7 32.9 111 33 0.23 2.7 1.7 2.9 18.8 59.8 2.1

SEISMIC SOIL LOADS

The Palmdale WRP is located in Los Angeles County north of the San Gabriel Mountains, 4.2 km north of the San Andreas Fault. In addition to the hydrodynamic forces, the design accounts for seismic soil loads for the buried or partially buried structures. Due to the proximity of the San Andreas Fault, these seismic loads governed the design in many cases. Design soil loads are presented in Table 5. Note that the seismic soil load of 512 kg/m3 is approximately the same as the active soil pressure of 480 kg/m3 . A service load combination of the two results in a nearly uniform load on cantilevered retaining walls. As a result, for many of the tanks, the design of the exterior walls was controlled by the load combination that includes seismic soil loads. The exterior walls are designed first for conditions that assume that the tank is full of water, but there is no soil backfill.

Table 4. Comparison of hydrodynamic forces in the short direction of the aeration tanks. UBC

Design soil loads.

Lateral soil pressure Unrestrained (Active): 480 kg/m3 Restrained (At Rest): 881 kg/m3 Seismic: 512 kg/m3 Inverted Triangle

Resultant hydrodynamic force & design wall

Pc (kN) Pi (kN) MIBP (kN-m) MEBP (kN-m) Convective acceleration hc (m) hi (m) w1 (kN/m2 ) w2 (kN/m2 ) P (kN/m of wall length) d (m)

Convective force Impulsive force Moment at the level of the base of the tank including the bottom pressure. Used in calculating overturning moments. Moment at the level of the base of the tank excluding the bottom pressure. Used in calculating wall water pressures Height above the base of wall to the resultant convective force Height above the base of wall to the resultant impulsive force Hydrodynamic water pressure at top of wall Hydrodynamic water pressure at base of wall Service level resultant force, (impulsive + convective) on the wall Height above base of wall to P’

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The next condition to be designed for is an empty tank with soil backfill. Seismic load cases are similar. The tank is assumed to be full but no backfill is in place is one load case, the next is an empty tank with soil backfill. A typical exterior wall with the water loads is shown in Figure 4. Figure 5 shows a typical exterior tank wall with the design soil loads. For calculation purposes, the water loads are independent of the soil loads. The service loadings will match actual conditions. For example, it will occur that the tank will be empty at some time when there is backfill in place. The project specifications for the water retaining structures require the contractor to perform a water tightness test prior to backfilling. The soil seismic load case is checked assuming the tank is empty. This reflects the possibility that during its life, the tank could be taken out of service for an extended period. The hydrodynamic load case is checked with no soil loading on the exterior of the wall. This is done not because it is assumed that the earthquake could occur when the tank is being excavated—a very unlikely event—but because for this load case the soil loads would be resisting the hydrodynamic loads. The geotechnical engineers provide soil loads that are conservative as loads but would not be conservative if used to resist other loads. Also, the resistance would be dependent on the actual field compaction of the backfill. Field compaction less than specified could result in an unconservative design. Although designing the exterior wall for hydrodynamic loads independent of

the soil loads, does not match what actually occurs, it does result in a conservative design. 6

The soils at the site are subject to seismic induced soil settlement. This occurs when relatively soft or loose soils are compacted during earthquake shaking. The soils at the site are alluvial deposits that are predominantly poorly graded sand, silty sand, and clayey sand with occasional silt and clay seams. A thin blanket of artificial fill soil and alluvium in the upper 1.5 meters to 1.8 meters are generally loose to marginally medium dense. The compaction of the upper 1.5 meters to 1.8 meters of soil during an earthquake will result in as much as 70 mm of settlement for structures with shallow foundations. This amount of settlement could significantly damage foundations and walls. In order to mitigate this settlement, and reduce building damage, the structures with shallow foundations will have the sub grade over excavated to 1.8 m below the existing grade. Then soil will be backfilled at a minimum of 90% of maximum compaction to the required finished grade. In some cases, where the finished grade is above existing grade, this requirement results in 1.8 m of excavation, and 3.6 m of backfill. This requirement affects all of the buildings and many other structures in the Palmdale WRP expansion. However, most the water retaining structures are not affected by the seismic induced soil settlement since their bottom slabs are set more than 1.8 m below existing grade.

7

Hydrostatic Load Figure 4.

Water loads on an exterior wall of a typical tank.

Seismic Soil Surcharge Load Load

Figure 5.

Hydrodynamic Load

Static Soil Load

Soil loads on an exterior wall of a typical tank.

SEISMICALLY INDUCED SOIL SETTLEMENT

CONNECTING THE EXISTING PLANT TO THE EXPANSION STRUCTURES

The existing WRP will remain in service during the construction of the expansion project. The structures associated with the expansion are separate from the existing, and tie-ins occur at only two locations— the upstream and downstream ends of the expansion project. At the downstream end, all of the plant effluent will flow through a new downstream junction structure. The new junction structure will replace an existing downstream manhole, which will be demolished. The new junction structure is needed for two reasons; first, it must allow flows from the new construction to enter into the existing 1.1 m reinforced concrete pipe (RCP) plant effluent line; and second, the new flows will cause the junction structure to be pressurized during high flow events. The existing manhole was not designed to be pressurized. The new junction structure was designed to be constructed around the existing manhole while limiting plant shutdowns to four hours maximum.

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The new junction structure is larger than the existing manhole. The first step in constructing the new junction structure will be to excavate to the base of the existing manhole. Both an existing lined drainage channel and a 610 mm encased RCP are within the excavation limits. The channel lining will have to be demolished and rebuilt after the excavation is backfilled. The contractor will have to support the RCP in place. The contractor is required to provide details and calculations for the supporting system to be used. After excavation, the bottom slab and walls of the new junction structure will be constructed. The new bottom slab will be doweled into the existing manhole base. The new walls will be placed around the existing 1.1 m RCP. A hydrophilic waterstop is used to provide water tightness at the pipe to wall joint. At this point a full flow-through inflatable plug will be used to temporarily contain the plant effluent flow allowing the contractor to demolish the existing manhole and complete construction of the new structure in the dry. The only plant shut downs required will be during the installation and removal of the full flow-through plug. This can be easily accomplished during the allowed four-hour shutdown. 8

CONCLUSIONS

The expansion of the Palmdale Water Reclamation Plant provided the County Sanitation Districts of Los Angeles County their first opportunity to compare the new IBC based California Building Code to the

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previous UBC based edition of the CBC in an actual design. Structural design calculations demonstrate minimal differences for this project. In practice, the two codes resulted in similar designs. This outcome was expected since water-retaining structures have performed well during past earthquakes, so there has been no reason to change the code requirements. Other aspects of the project provide insight into practical methods of mitigating seismic induced soil settlement and into the proper inclusion of hydrodynamic and soil seismic loads in the design of a water retaining structure. Finally a practical construction sequencing method of providing a tie-in to an existing plant effluent line while maintaining service and minimizing plant down time is discussed. REFERENCES American Concrete Institute 2006. Code Requirements for Environmental Engineering Concrete Structures and Commentary (ACI 350-06). American Concrete Institute 2006. Seismic Design of LiquidContaining Concrete Structures and Commentary (ACI 350.3-06). California Building Standards Commision 2007. California Building Code—California Code of Regulations—Title 24, Part 2. International Code Council 2006. International Building Code. International Conference of Building Officials 1997. Uniform Building Code.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Fuzzy logic based diagnostic tool for management of timber bridges S. Ranjith, S. Setunge & R. Gravina RMIT University, Melbourne, Australia

S. Venkatesan Victoria University, Melbourne, Australia

ABSTRACT: Australia has a large number of timber bridges currently in need of new and effective management practices. Timber bridges have a high maintenance cost and are affected significantly by a number of deterioration mechanisms which are not fully understood. Early diagnosis of possible deterioration scenarios can lead to effective management strategies. This paper presents development of an expert system which aids the nonexpert to diagnose the cause of the distress in timber bridges using the cause-and effect diagrams and fuzzy logic approach. The inputs to the system are linguistic variables such as the type of element, the visual symptoms, the environmental conditions, method of construction and the location of the bridge. The expert system executes fuzzy inference to evaluate the cause of the distress using these input data and built in rules.

1

INTRODUCTION

Timber bridge management is a major issue for many Engineers in Local Government. The level of expertise available for managing timber bridges is not adequate in some councils. Timber bridges require high maintenance cost and good management practices. Identifying distress mechanisms and early signs of predicting deterioration are essential for good management practice. This tends to avoid ‘‘band-aid’’ management practices that react to emergencies, but helps to plan bridge maintenance and repairs strategically. The aims of the timber bridge diagnostic tool is to help engineers without sufficient experience to make appropriate decision on repair strategies and program maintenance schedules by identifying the distress mechanisms. This approach can capture the complexity of relating different types of defects and the corresponding possible causes in a simple algorithm. The input to the system is usually in the form of natural language that contains intrinsic imprecision and uncertainty, therefore a reasoning mechanism that can deal with uncertain and imprecise information is used in the proposed expert system. This has suggested the application of fuzzy sets theory that can provide a basis for handling various types of input information within a formal mathematical frame work. Cause and effect diagrams have been used to classify the relationships between defects and their causes and to develop the rule-bases. Fuzzy models have previously been used in the area of deterioration diagnosis ex: crack formations in reinforced concrete structures (Chao & Cheng, 1998), the impacts of design factors on bridge deterioration

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(Chao & Chen 2002), detecting distress mechanisms in reinforced structures exposed to aggressive environment (Venkatesan et al. 2007) and assessment of damage in bridge decks (Furuta et al. 1990). 2

CAUSE AND EFFECT DIAGRAM

The cause and effect diagram is a tool for discovering all the possible causes for a particular effect, sometimes called as Ishikawa or ‘‘fishbone’’ diagram (Fig. 1). It graphically illustrates the relationship between a given outcome and all the factors that influence V3

V1

V5 V31 V52

V11

V12 V14

V51 V53

V54

V13

V55 V32

V15

distress mechanism

V21

V41 V22

V23 V24

V2

V62

V42

V4

Mk

V61

V63

V6

Figure 1. Cause-and-effect diagram of distress mechanism affecting timber bridges.

Table 1.

Causes of defects of primary level.

Table 3.4.

Region (V24 ).

Cause (p)

Characteristic

Cause (t)

Characteristic

V1 V2 V3 V4 V5 V6

Elements Environment Construction Characteristic of deterioration Symptoms Element position or location

V241 V242

Marine Mountainous

Table 2.

Elements (V1 ).

Cause (s)

Characteristic

V11 V12 V13 V14 V15 V16

Deck Girder Stringer Pile Kerbs Bituminous surface

Construction (V3 ).

Table 4. Cause (s)

Characteristic

V31 V32

Whether timber is treated or not Over loading Whether timber is treated or not (V31 ).

Table 4.1. Cause (t)

Characteristic

V311 V312

Timber is treated Timber is not treated

Environment (V2 ).

Table 4.2.

Over loading (V32 ).

Cause (s)

Characteristic

Cause (t)

Characteristic

V21 V22 V23 V24

Temperature Moisture Rainfall Regions

V321 V322

Yes No

Table 3.

Table 5. Table 3.1.

Temperature (V21 ).

Cause (t)

Characteristic

V211 V212 V213

Below 20◦ C 20◦ C−25◦ C Above 25◦ C

Table 3.2.

Moisture (V22 ).

Cause (t)

Characteristic

V221 V222

Below 20% Above 20%

Table 3.3.

Rainfall (V23 ).

Cause (t)

Characteristic

V231 V232

Low rainfall High rainfall C

this outcome. Cause and effect diagram combined with fuzzy logic approach have been used to identify the defects in concrete structures. This approach can be used to identify the distress mechanisms in timber bridges as well.

Characteristics of deterioration (V4 ).

Cause (s)

Characteristic

V41 V42

Location of the deterioration Rate of deterioration

Table 5.1.

Location of the deterioration (V41 ).

Cause (t)

Characteristic

V411 V412 V413 V414 V415 V416 V417 V418 V419 V41-10

End of the member Top of the member Inside of the member Mid span of the member Inside and along the member Beneath the cross head Along the nail At the exposed ends and interfaces with kerbs External surface Where the timber is restrained by bolted steel plate or other type of fastening Ground contact

V41-11

The cause parameters can be divided into ‘‘Primary’’ (p), ‘‘Secondary’’ (s) and ‘‘Territory’’ (t) levels. Primary level parameters are denoted by V = {V1 , V2 , . . . Vi . . ., Vn }, secondary level parameters are denoted by Vi = {Vi1 , Vi2 , . . .Vij . . . , Vim } and

796

Table 5.2.

Table 6.

tertiary level parameters are denoted by Vij = {Vij1 , Vij2 , . . .Vijk . . ., Vijl }.

Rate of deterioration (V42 ).

Cause (t)

Characteristic

V421 V422

Gradual Rapid

3

This section describes how the fuzzy logic approach is utilized to determine the distress mechanisms affecting timber bridges.

Symptoms (V5 ).

Cause (s)

Characteristic

V51 V52 V53 V54 V55

Checks Split Decay Sign Loss of timber section material alone Loss of timber section material and appearance of termite and/or mud walled tubes called galleries on the outside surface of the member and/or inside the member Narrower and shorter tunnels on the surface and a wasted appearance or hourglass effect Crushing Crack Small 5 to 10 mm diameter holes in the piles Narrower and shorter tunnels on the surface of the pile

V56 V57 V58 V59 V5-10

MODELING APPROACH

Table 6.1.

Decay sign (V53 ).

Cause (s)

Characteristic

V531 V532 V533 V534

Discoloration Cube or Alligator Patterns Fungi or fruiting bodies (conks) Stain

Table 6.2.

Crack (V58 ).

Cause (t)

Characteristic

V581 V582 V583 V584

Crack across the grain Large longitudinal cracks Transverse crack Longitudinal crack

3.1

A Fuzzy Set is any set that allows its members to have different grades of membership (membership function) in the interval [0, 1]. The degree of membership is expressed as its membership function. Fuzzy sets and linguistic variables are used to quantify the concepts used in natural language which can then be manipulated. Linguistic variables must have valid syntax and semantics which can then be specified by fuzzy sets or rules. A syntactic rule defines the wellformed expressions in T(L), where the term T(L) is a set of linguistic variable and L is the set of values it may take. For example, T(Age) = {very_young, young, middle_age, old, very_old}, where each of these values may itself be a linguistic variable that can take on values that are fuzzy sets. The membership function could be defined as the S function μold (x) = S(x; 60, 70, 80). The degree of status of the variables in rule bases for different distress mechanisms Mk , for matching user input variables with the variables in rule-bases which is defined as a set A = {very_low, low, medium, ˜ high, very_high}. Then, the fuzzy set is defined as C, x = {0, 0.1, 0.2, . . ..1.0}, and the membership function (Fig. 2) is defined as μà (x) = S(x; 0, 0.5, 1.0), μA , μ C 1.0 0.9 0.8

Table 7.

Reasoning engine

Element position or location (V6 ).

0.7

Cause (t)

Characteristic

0.6

V61 V62 V63

Under water Above ground Below ground

0.5

very_high high

medium

0.4 0.3

low

0.2

Table 7.1.

Under water (V61 ).

Cause (t)

Characteristic

V611

Zone between bed level and mean low tide level Ground or normal water level

V612

0.1

very_low

0

X 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9

1.0

Figure 2. Membership functions of linguistic variables Set A & Set C.

797

4

μB

DEMONSTRATION OF THE APPLICATION OF THE METHOD

-0.5

high

The following case study was taken from the Journal of Waterway, Port, Coastal and Ocean Engineering. Anido et al. (2004).

medium

4.1

very high -0.4

-0.3

Case study example: The Custom House Wharf

The Custom House Wharf is an earth-filled pier structure with wooden-timber and a steel crib bulkhead, wood piles, and an asphalt paved wood deck. There are several marine-related businesses operating on the pier (Maine DOT 1986). Several piles had reduced cross-sections in the tidal zone between low and high

-0.2 low -0.1 very low_ 0

X 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9

1.0

Table 8. Figure 3. Set B.

User input.

Membership functions of linguistic variables Variable (Vi / Vij )∗ Description

x € X, where X is the Status space. The degree of status of the variables in rule bases for different distress mechanisms Mk , for not-matching user input variables with the variables in rule-bases which is defined as a set B = {very_low, low, medium, high, ˜ very_high}. Then, the fuzzy set is defined as B(x), x = {0, 0.1, 0.2, . . .1.0}, and the membership function (Fig. 3) is defined as μB˜ (x) = S(x; 0, 0.5, 1.0), x € X, where X is the Status space. The degree of ˜ confirmation of the fuzzy set is defined as C(x), x = {0, 0.1, 0.2, . . ..1.0}, and the membership function (Fig. 2) is defined as μC˜ (x) = S(x; 0, 0.5, 1.0), x € Y, where Y is the confirmation space. In the user interface users are asked to select the element under deterioration. It is compulsory to select an element in order to proceed to the next screen. Once the user selects the element and enter the requested information, all rule-bases related to that element type are taken up for assessment. Then, a fuzzy vector is generated based on the membership function values μà (x) and μB˜ (x). Then, each of the fuzzy vectors is compared with the subsequent vector such that if F(i) < F(i+1) then the vector F(i+1) is selected. The above step is duplicated until all the vectors are match and the vector with the maximum positive rating is selected as the closest possible match. If the difference between any two vectors is zero then three matching solutions are provided with a degree of confirmation. The degree of confirmation is determined by the following equations. Confirmation γ = [(Fi /Fi ) ∗ μC (x)]. In the cases of two or three mechanisms with close confirmation limits, the matching mechanisms are listed with the advice that further precise information is required or re-run the software in order to evaluate the dominant distress mechanism.

V1 V2 V21 V22 V24 V3 V31

V4 V41 V5

V6 V61

Element Environment Temperature# Moisture# Region Construction Whether timber is treated or not# Characteristic of deterioration Location of the deterioration Symptoms

Element position or location Under water

User input (Vij /Vijk )

Degree of confirmation (μC (x))∗∗

Pile (V14 )

Very high

Below 20◦ C (V211 ) Above 20% (V222 ) Marine (V241 )

High High Very high

Timber is treated (V311 )

Medium

External surface (V419 )

Very high

Narrower and Very high shorter tunnels on the surface and a wasted appearance or hourglass effect (V56 )

Zone between bed level and mean low tide level (V611 )

Very high

* The variables are selected from the required variables to define all rule-bases of element (pile). More variables may be added when more rule-bases are created. * Degree of confirmation is provided by the user according to user’s confidence. # The most possible input for ‘temperature’, ‘moisture’ and ‘whether timber is treated or not’ are assumed in this example, since these information are not available in the case study.

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Table 9.

Rule base 1—Teredo attack (M1 ).

Variable Description (Vi ) V1 V2

V3 V4

V5

V6

Element Environment Temperature (V21 ) Moisture (V22 ) Region (V24 ) Construction Whether timber is treated or not (V311 ) Characteristic of deterioration Location of the deterioration (V41 ) Symptoms

Element position or location Under water (V61 )

Table 11. Rule base 3—Sand movement with the tides attack (M3 ). Variable (Vi )

Characteristic

V1 V2

Below 20◦ C Above 20% Marine

V3

Timber is not treated

V4 External surface Small 5 to 10 mm diameter holes in the piles below water and there may not be any sign of attack on the timber surface

V5

V6

Rule base 2—Crustuceans attack (M2 ).

Variable (Vi ) Description V1 V2

V3 V4

V5

V6

Element Environment Temperature (V21 ) Moisture (V22 ) Region (V24 ) Construction Whether timber is treated or not (V311 ) Characteristic of deterioration Location of the deterioration (V41 ) Symptoms

Element position or location Under water (V61 )

Characteristic

Element Environment Temperature (V21 ) Moisture (V22 ) Region (V24 ) Construction Whether timber is treated or not (V311 ) Characteristic of deterioration Location of the deterioration (V41 ) Symptoms

Pile

Element position or location Under water (V61 )

Below 20◦ C Above 20% Mountains Timber is not treated External surface Wear away (Diameter of the pile is smaller than the original) Ground or bed level

Zone between bed level and mean low tide level Table 12.

Table 10.

Description

Pile

Variable status matrices for input. M1

M2

M3

Characteristic

Variable*

Yes

No

Yes

No

Yes

No

Pile

V1 V21 V22 V24 V311 V41 V5 V6

0.9 0.5 0.5 0.9 0.75 0.9 0.9 0.9

−0.45 −0.25 −0.25 −0.45 −0.375 −0.45 −0.45 −0.45

0.9 0.5 0.5 0.9 0.75 0.9 0.9 0.9

−0.45 −0.25 −0.25 −0.45 −0.375 −0.45 −0.45 −0.45

0.9 0.1 0.1 0.9 0.5 0.9 0.9 0.9

−0.45 −0.05 −0.05 −0.45 −0.25 −0.45 −0.45 −0.45

Below 20◦ C Above 20% Marine Timber is not treated

* Some input variables are more compulsory to define a particular rule base than other. For example ‘‘whether timber is treated or not’’ is essential to determine the distress mechanism of weathering, but not for the distress mechanism of over loading.

External surface Narrower and shorter tunnels on the surface and a wasted appearance or hourglass effect Zone between bed level and mean low tide level

tide. Other piles had extensive visible damage at the butt (reduction in cross section), as well. One wood pile was measured at two locations: The diameter at the butt was 254 mm, and the diameter at the mud line

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Table 13.

Comparison of user input with Rule base-1.

Variable

Matching with user input

Membership value set A

Membership value set B

V1 V21 V22 V24 V311 V41 V5 V6

Yes Yes Yes Yes No Yes No Yes

0.9 0.5 0.5 0.9 − 0.9 − 0.9

− − − − −0.375 − −0.45 −

Table 14.

Comparison of user input with Rule base-2.

the fuzzy vector F˜ 2 is:

Variable

Matching with user input

Membership value Set A

Membership value Set B

γ =

V1 V21 V22 V24 V311 V41 V5 V6

Yes Yes Yes Yes No Yes Yes Yes

0.9 0.5 0.5 0.9 − 0.9 0.9 0.9

− − − − −0.375 − − −

Table 15.

Comparison of user input with Rule base-3.

So the distress mechanism is Crustaceans attack with a degree of confirmation of 90%.

Variable

Matching with user input

Membership value Set A

Membership value Set B

5

V1 V21 V22 V24 V311 V41 V5 V6

Yes Yes Yes No No Yes No No

0.9 0.1 0.1 − − 0.9 − −

− − − −0.45 −0.25 − −0.45 −0.45

Table 16.

Comparison of Fuzzy vectors.

Variable

F˜ 1

F˜ 2

F˜ 3

V1 V21 V22 V24 V311 V41 V5 V6

0.9 0.5 0.5 0.9 −0.375 0.9 −0.45 0.9

0.9 0.5 0.5 0.9 −0.375 0.9 0.9 0.9

0.9 0.1 0.1 −0.45 −0.25 0.9 −0.45 −0.45

level (1.83 m below the butt) was only 165 mm. This loss of cross section represents about a 50% reduction in the cross sectional area. To assess the condition of a wood pile below the mud line, a hole of approximately 130 mm in depth was excavated in the surrounding soil. Visual inspection indicated that the wood pile had no reduction in cross section or any apparent damage below the mud line. Fuzzy vectors for the above three rule bases are shown below for comparison (Table 16).   F˜ 1 − F˜ 2 = −1.35 < 0, F˜ 2 is selected   F˜ 2 − F˜ 3 = 4.725 > 0, F˜ 2 is selected Rule base-2 is selected as the best matching distress mechanism; The degree of confirmation for



[(F˜ i /



˜ ∗ μC (x)] Fi)

= [(0.9/5.125) ∗ 0.9 + (0.5/5.125) ∗ 0.75 + (0.5/5.125) ∗ 0.75 + (0.9/5.125) ∗ 0.9 − (0.375/5.125) ∗ 0.5 + (0.9/5.125) ∗ 0.9 + (0.9/5.125) ∗ 0.9 + (0.9/5.125) ∗ 0.9] = 0.90

CONCLUSIONS

This paper attempted to develop an expert system to identifying the distress mechanisms in timber bridges using the cause-and effect diagrams and fuzzy logic approach. Identifying distress mechanisms and early signs of predicting deterioration are essential to establish an efficient repair and maintenance program. The work completed will assist the infrastructure managers to manage the bridges proactively rather than reactively. The expected deliverables includes improved program maintenance schedule, earlier diagnosis, improved decision making on repair strategies, reduced risk and minimized the life cycle cost. The following conclusions are drawn from the work: 1. Cause and effect diagrams can be effectively used to identify the main variables and sub variables that are causing the distress mechanisms. 2. Accuracy of the expert system is dependent on quality of the data and availability of the necessary data. 3. The proposed expert system can be used by both experts and non-experts to identify the distress mechanism in timber bridges. 4. Uncertainty involved in data input such as visual inspection can be handled using fuzzy logic. 5. The rule bases can be improved by consultation with experts and/or when further inspection data becomes readily available. REFERENCES Anido, R. et al. 2004. Assessment of wood pile deterioration due to marine organisms. Journal of waterway, Port, Coastal and Engineering. Chao, S. & Cheng, F. 1998. Fuzzy pattern recognition model for diagnosing cracks in RC Structures. Journal of computing in Civil Engineering. Crews, D.K. Managing timber bridges—without over stressing. Australian Small Bridge Conference, 2007.

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Furuta, H. et al. A Fuzzy expert system for durability assessment of bridge decks. Proceedings of First International Symposium on Uncertainty modeling and analysis, 1990. Maine Department of Transportation (DOT). 1986. Port inventory and Evaluation, Rep., Vol. 1 Augusta, Me. Muller, W. Timber girder inspection using ground penetrating radar. Proceeding of International Symposium on NonDestructive Testing in Civil Engineering, 2003. Principal Bridge Engineer’s Section, Design Department, Vic Roads. 1995. VIC Roads Bridge Inspection Manual. Setunge, S. et al. Management of reinforced concrete bridges exposed to aggressive environments. Concrete Conference, 2007. Transport Technology Division, Queensland Department of Main Roads. 2000. Bridge Inspection Manual.

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Venkatesan, S. et al. Evaluation of distress mechanisms in bridges exposed to aggressive environments. Clients Driving Innovation: Moving Ideas into Practice Conference, 2006. Venkatesan, S. et al. Diagnostic process based on fuzzy logic for management of bridges exposed to aggressive environments. Clients Driving Innovation: Benefiting from Innovation Conference, 2007. Zadeh, L.A. 1975. The concept of a linguistic variable and its application to approximate reasoning-I. American Elsevier Publishing Company, Inc. Zhao, Z. & Chen, C. 2001. Concrete bridge deterioration diagnosis using fuzzy inference system. Oxford; Elsevier Science Ltd.

Organizational behavior

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Behaviors of leadership in architectural offices E. Kasapo˘glu Istanbul Kultur University, Department of Architecture, Istanbul, Turkey

ABSTRACT: Leadership can be defined simply as collecting a group of people around definite objectives and having the ability of their achieving these objectives. In architectural offices, as all groups, design group also needs a leader. If the members of the design team do not trust and believe in the owner of the architectural office, this is a false beginning for construction management process that without being a team achievement of the project can be affected negatively. This study is based on the research in 2008. Data were collected through a questionnaire directed to 40 architects from different architectural offices and discussed through a new model to be effective for the management of the architectural offices. Behaviors of leadership are gathered into two main groups which are delegation of authority and orientation. The common behaviors of the leader architects are aimed to be measured through subgroups of these main groups.

1

INTRODUCTION

Architectural offices are one of the important parts of construction process as production begins with architectural project. Every project has a different function and content of the task changes with a new one which requires a new design process. It is a hard and troubled process consisting of creativity that needs a team working. Although it is easy to be successful with a group, it is not easy working with a group, trusting and believing in each of the team members, collecting them around team objectives and creating their desire to work. If personal goals of every group member are united with the group objectives, every member of the group will be more desirous to achieve the group objectives. As all of the groups need a leader, design group also needs a leader. The leader can be the head of the project team and this leader is the owner (s) of the offices in Turkey. If the members of the design team do not trust and believe in the owner of the architectural office, being owner of the office will not be enough to be the leader of the office. This is a false beginning for construction management process that without being a team achievement of the project can be affected negatively. 2

LEADERSHIP

People are social living being that live in groups. They need leaders to manage these groups and achieving objectives. Leadership had been in every age of history and one of the most popular research topics among organizational behavior subjects and since the days of Greek philosophers, has also contemplated. The Greek philosopher Plato is defined leadership as prudence,

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courage, temperance, and justice. Although leadership is one of the most observed phenomena on the earth, it is the least understood one (McShane & Von Glinow 2000). Whereas leaders and leadership situations differs, has caused numerous leadership definitions. While some of the definitions of leadership based on leader characteristics, some of the other based on behaviors, and another on situations. A definition of leadership needs to sufficiently comprise different theories through applications. Leadership is generally defined as influencing others to work willingly toward achieving objectives. Leaders simultaneously fill many roles, interacting with and motivating subordinates, leading groups whose members are interacting and in which conflicts may arise. Leadership means crystallizing a direction for subordinates, and then tapping into all the authority, charisma and traits the leader can muster to make the subordinates want to follow the leader in achieving the leaders’ goals (Dessler 2001). All leaders of effective groups share four characteristics. According to Warren Bennis, who devoted decades to researching leadership issues concludes that; all effective leaders provide direction and meaning to the people they are leading. They remind people what is important and why, what they are doing makes an important difference by this way. The second characteristic is; they generate trust. As they are proactive and willing to risk failing in order to succeed; the third one is their favor of action and risk taking. In both tangible and symbolic ways they reinforce the notion that success will be attained; they are also purveyors of hope. These four characteristics are not enough to describe leadership. Leadership does not take place in a vacuum. There are three important variables every

leader must deal. The people who are being led; the task the people are performing and the environment in which people and the task exist. These three variables are different in every situation. It is also differs for all leaders, what is expected and needed, as every leader is in different situation (Ivancevich et al. 2005). 3

BEHAVIORS OF LEADERSHIP IN ARCHITECTURAL OFFICES

Behaviors of leadership in architectural offices are considered as two main groups through the level of authority delegation and orientation. In architectural offices delegation of authority reflects the level of freedom of the employee architects in the office and gathered into three subgroups as authoritarian, participative and free-rein styles of leadership. Orientation shows in which direction is the leader architect oriented which was gathered also into three subgroups as achievement oriented, employee oriented and task oriented behaviors of leaderships. 3.1

Behaviors of delegation of authority

Authoritarian leadership, which is the first group of delegation of authority, solves the problem and make the decision alone, using the information available at the time (Dessler 2001). Authoritarian leaders schedule work, maintain performance standards and let subordinates know what is expected of them (Johns 1992). They take full authority, assume full responsibility, and structure the complete work situation for their employees, who are expected to do what they are told (Newstrom & Davis 1993). The leader is competent and a good coach clarifies performance goals, the means to reach those goals, and the standards against which performance will be judged. The employee is motivated to learn a new skill. The situation is a new environment for the employee. In some of the appropriate conditions, if the time is short and the leader has all the information to solve the problem; using authoritarian leadership can be suitable as a right behavior. In some of the appropriate conditions, especially for a new employee who is just learning the job, authoritarian leadership may suit as a right behavior in an architectural office. The leader tells her/his employees what she/he wants done and how she/he wants it done, without getting advice of her/his team members. If the leader has some of the information and the team members have some of the information, participative style can be the right behavior used. This allows all parties to become a part of the team and allows them make a better (McShane & Von Glinow 2000). Participative leaders decentralize authority (Newstrom & Davis 1993), consult with subordinates about work-related matters and consider their opinions (Johns 1992). Participative decisions are not unilateral, as with the authoritarian,

because they arise from consultation with followers and participation by them. The leader and group are acting as a social unit (Newstrom & Davis 1993). With a team of members who know their job, using a participative style can be suitable as a right behavior for an architectural office. The leader is friendly and approachable, makes the work more pleasant and his/her behaviors encourage and facilitates employees involvement in decisions beyond their normal work activities. The leader consults with the employees, asks for their suggestions, and before making a decision takes these ideas into serious consideration. If there is no short of time and the leader wants to gain more commitment and motivation from the employee, it can be the right behavior using participative style. The employees know their jobs and want to become part of the team and the situation allows time. In this type of style, the leader and one or more employees are on the decision making process that is determining what to do and how to do it. However, the leader maintains the final decision making authority. Freerein leaders delegate the authority completely to the employee performing the task. They avoid power and responsibility. Employee perform the task and responsible from all of the decisions about the project. Freerein leaders depend largely upon the group establish its own goals and work out its own problems; they play only a minor role and delegate the authority completely to the employee performing the task. In this type of style, employee perform the task and responsible from all of the decisions about the project (Newstrom & Davis 1993). 3.2

Behaviors of orientation

The second group of behaviors of the leader architects is through their orientation. A leader may be oriented towards achievement, employee or the task. It depends on the orientation of the leader architect. Some leaders may be achievement oriented that they transfer the task and they are not interested in which way the employee followed. A leader can not perform the entire task in the office and needs to share responsibility. In some situations, the leader also might need to be in other places performing other tasks. Achievement oriented behaviors encourage employees to reach their peak performance, allows the employees to make the decision, and they are only interested in achieving (McShane & Von Glinow 2000). Achievementoriented leaders, increase creativity, motivation and spiritually of their followers and encourage subordinates to exert high effort and strive for a high level of goal accomplishment and express confidence that these goals can be attained (Johns 1992). In some conditions, if the team members know more about the job than the leader, using achievement oriented style can be the right behavior for an architectural office. The leader in an architectural office sets goals,

806

expects employees to perform at their highest level, seeks improvement in employee performance continuously, and shows a high degree of confidence that employees will assume responsibility and accomplish attaining goals. However, the leader is still responsible from the decisions. Achievement oriented leadership can be used when the team members are able to analyze the situation and determine what needs to be done and how. A leader in architectural office should be aware of the fact that he/she can not do everything and it is impossible going further without sharing responsibility. A leader should set priorities and delegate certain tasks. Task oriented leaders assign employee to specific tasks, clarify their work duties and procedures, and ensure that they follow company rules, and push them to reach their performance capacity (McShane & Von Glinow 2000). While person concerned leaders do best in moderately favorable conditions, task oriented leaders do best in situations that are highly unfavorable for exerting influence (Cohen et al. 2001). Employee orientation is concerning about the human needs of the employees, built teamwork, help employees with their problems and provides psychological support when needed. Leaders with a strong employee oriented style do personal favors for employees, support their interest when required and treat employees as equals. Employee oriented leaders have compassion that they show concern for the suffering or welfare of others and mercy to others. Leaders make an effort to understand the needs of their employees and take steps to address those needs and concerns (Sarros et al. 2007). Addressing to personal and future interests of their followers is one of the main behaviors of employee oriented leaders.

4 4.1

A STUDY OF LEADERSHIP IN ARCHITECTURAL OFFICES Data collection

A questionnaire was designed to solicit information about leadership behaviors of architects. The first part of the questionnaire comprises demographics like gender and age of the architect, size and age of the office. The second part of the questionnaire comprises the questions through measuring behaviors of leadership. The research examines 40 architects who are owner or partner of architectural offices. The sample used in the research was drawn from the data base of the architectural centre of Turkey, Arkitera. A cover letter and a questionnaire form were sent to 150 architects, by e-mail for three times. The cover letter comprises the promise of secrecy and using the results for only academically purposes. The key informant used in this study was the owner or the partner of each office. The data collection comprises the first half of 2008 and 40 of the architects were replied. The sample comprises

807

different sizes and ages of architectural offices that many of them are well known architectural offices. Although the research comprises all of the architectural offices from Turkey, most of the architectural offices were in Istanbul. 4.2

The sample

The gender proportions of the respondent architects were 7 women and 33 men, which mean nearly 18% of the architectural offices in Turkey, have women owner or partner. The questionnaire was sent all of the architectural offices, which can be reached by e-mail, without taking into consideration of the gender of the responded. A general evaluation of these results shows also the low ratio of women architects in Turkish architecture as owner or partner. As the questionnaire were sent all of the architectural offices, which can be reached, without taking into consideration of the city where architectural offices take place, 29 of the respondents were from Istanbul. 8 of the respondents were from Ankara and 3 were from other parts of Turkey. The age of the architectural offices were measured by asking respondents to indicate the number of years the office had been in existence. The age of the interval between 11 and 15 years and greater than 21 years are the two main groups. According to these results, the architectural offices in Turkey established between 11 and 15 years earlier are at a rate of 35%. The ratio of the offices established twenty one and more years is nearly 40%. There is a gap between the ages of between 11–15 years and 21 and more years that, which might have been a direct relationship with the economic conditions of Turkey. The size of the architectural offices was measured by asking respondents to indicate the total number of their employees. The size distributions of architectural offices, the number of employee in the interval of 1–5 and 6–10 employees are the two main groups. According to these results, the ratio of the number of employees between 1 and 5 is 30% and 6 and 10 is 32.5%. Architectural offices in Turkey are small organizations that the rate of the offices having more than ten employees is 37.5%. Architects were asked ten questions to measure their tendency of behaviors of leadership. The respondents were given a five category response scale ranging from always, often, sometimes, rarely and never. The behaviors of leadership were divided into two main categories, which are delegation of authority and orientation, through the view of leadership behaviors in architectural offices. The first group of five questions was through the level of delegation of authority that the questions were asked to measure which styles of leadership architects behave usually, authoritarian, participative or free-rein. The first set of questions was for measuring

the level of freedom and punishment in the office, participating of the employee architects to the decision taking process, the way of problem solving and explaining the reasons of the behaviors to the employee architects. 5 5.1

RESULTS AND DISCUSSION Behaviors of delegation of authority

The first question was about the level of restrictions in the office. Architects were asked to indicate the level of freedom of the employee architects in their work. According to the results, more than half of the respondents set free the employee architects in their work. Figure 1 show that a very limited group of architects depended on their leader architect. The responses are gathered into two main groups as sometimes and often. The second question of the first group was about the level of problem solving of the employee architects. Architects were asked if they preferred employee architects solving their problems about the task before all else. Figure 2 show the responds of the architects to the second question. According to the results, although nearly half of the respondents prefer solving the problems of their employees, a limited group at a rate of 15% objects. The responses are distributed between always, often and sometimes. The third question was about the level of explaining the reasons of their behaviors to the employee architects. Figure 3 show the responds of the architects to the first question. According to the results, most of the architects prefer to explain the reasons of their behaviors to the employee architects at a rate of 32% and

29% which means architects explain the reasons of their behaviors sometimes and rarely. The fourth question to measure the level of delegation of authority of the leader architects were the level of participation of the employee architects to the decision about the architectural projects. The architects were asked to indicate the level of participation of the employee architects, when it is needed to have a decision about the project. According to the results, leader architects generally make decision with the employee architects (Figure 4). The responses are gathered largely on often. The fifth question of the first group of questions to measure the level of delegation of authority was about the level of punishment. Architects were asked to indicate the level of firing the employee architects when they make a big mistake. Figure 5 shows the responds of the architects. According to the results leader architects, generally do not fire employee architects if they make a big mistake.

always 2% never 23%

rarely 2%

rarely 33%

Figure 3.

sometimes 33%

always

never 0% always 25%

always often sometimes

often 47%

never

Figure 4.

never

Participation.

Freedom. never rarely 5% 10%

sometimes 28%

Figure 2.

never

rarely

rarely

Figure 1.

rarely

sometimes 30%

sometimes

often 44%

often

Behavior.

sometimes 23%

often

always sometimes

rarely 5%

never 3% always 18%

often 12%

Problem solving.

always 28%

always

never 30%

often

always 5% often 12%

often 29%

rarely 33%

never

Figure 5.

808

Fault.

often sometimes

sometimes rarely

always

sometimes 20%

rarely never

5.2

Behaviors of orientation

Second group of the leadership behaviors were through orientation of the leader architects which are achievement oriented, task oriented or employee oriented styles of leaderships. The second group of five questions was asked to measure in which behavior of the leader architects oriented. A leader can be oriented to achievement, which is transferring work to the employee and leaving all the responsibility to the employee, as they are interested in achievement, but not in which way to be followed. On the other hand, a leader can focus on the individuality and personality needs of their employees and emphasize building good interpersonal relationships. The third type of orientation behavior is task oriented leadership; focus on production and technical aspects of the task. Closely supervising, constantly checking up employees and telling them what to do are the main behaviors of task oriented leaders. The first question for measuring the direction of the orientation of the Turkish leader architects was about the level of administration of the work. Architects were asked if they are responsible from achieving continuity of work. Figure 6 shows the responds of the architects to the third question. According to the results, most of the architects at a rate of 41% and 38% always and often achieve the continuity of work. The second question was about the level of work transfer in the office. The architects were asked to indicate the level of freedom after transferring work to the employee architects during project phase. Figure 7

rarely 3% sometimes 18%

never 0%

always sometimes

always often sometimes

rarely

rarely never

often 57%

Administration. Figure 8.

never 10% rarely 18% sometime s 37%

Figure 7.

never 0% always 20%

never

often 38%

Figure 6.

rarely 5% sometimes 18%

often always 41%

show the responds of the architects to the third question. According to the results, sometimes at a rate of 37% leader architects can leave the work to the employee architects, but not completely. The third question of the second group of questions was about the level of participation of the employees to take a decision. Architects were asked to indicate the level of their deciding what will be done and how in the office. Figure 8 show the responds of the architects to the second question. According to the results, leader architects decide often what to do and how in the office at a rate of 57% in the office. The fourth question was about the level of rewarding of the leader architects. Architects were asked to indicate the level of rewarding when the employee architects completed their work successfully. According to the results, more than 60% of the respondents generally reward the employee architects. Figure 9 shows the responds of the leader architect. The responses are distributed between always, often and sometimes. The fifth question of the second group of questions was about the level of work distribution. Leader architects were asked to indicate the level of their distributing the work to the employee architects. Figure 10 show the responds of the architects to the fifth question. According to the results most of the architects prefer always and often distributing the work to the employee architects at a rate of 54% and 33%. A general evaluation from the last question can be accepted as a main indication that Turkish architects behave task oriented generally.

always 10%

Decision.

rarely 5%

always often

often 25%

sometime s 28%

sometimes rarely never

Work transfer.

Figure 9.

809

Reward.

never 3% always 32% often 32%

always often sometimes rarely never

rarely 3%

never 5%

always often

sometimes 5% often 32%

Figure 10.

6

sometimes rarely

always 55%

never

Work distribution.

CONCLUSIONS

Leadership involves the social relationship between the manager and the members of the group and represents the effectiveness of interpersonal relationships between leaders and members (Luria 2008). In architectural offices, the importance of leadership is equal to the other types of organizations. Behaviors of leaders affect directly to the performance of the employee architects. According to the results, although Turkish architects are participative leaders, their intension is behaving towards participative to authoritarian. Although Turkish architects indicate employee architects are free in their work, the level of participation is high and the level of firing is low, they also indicate that their preference is not at the same level with employee architects solving the problems in their work before all else and explaining the reasons of their behaviors to the employee architects. According to the results, leader architects have a tendency to behave authoritarian. The second set of questions was asked to measure the orientation of Turkish leader architects. According to the results, although they have a tendency to behave employee oriented, the results of Figure 9 shows that they have a rewarding system which means they have a tendency through behaving achievement oriented to task oriented. When the results of the other questions evaluated the results show that they have a tendency through close supervision. In an architectural office, participative leadership can be the right behavior in some situations, but it can be wrong in some other situations. Generally, authoritarian style of leadership can be used when a leader should tell the team members that a method for achieving the project is not working properly and a new one is needed. In some other situations participative style of leadership can be the right behavior. A leader can ask for the views of the team members and participating in creating a new method. If the situations allow,

achievement oriented style of leadership can be used to introduce the new method. It all depends on the situation, if the relationships are based on respect and trust or on disrespect; if the employees have enough experience to do the job; if the leader or the employee has the information. It all depends on also, the levels of internal conflicts and stress, the type of task and if the task is structured, unstructured and complicated or simple. Being the project restricted by the laws or established procedures is also affects the way of leader behaves. A leader in an architectural office can use authoritarian leadership for the new employee. If the employee has enough experience, the leader can choose participative style. If the employee has worked long enough and the leader trusts him/her for doing his/her best, he can choose achievement oriented style. Consequently, the employees with an internal locus of control, believe that they have control over their work environment, prefer participative and achievement oriented leadership behaviors and may feel embarrassed with an authoritarian style. In contrast, people with an external locus of control believe that, their performance is due luck and fate, so they tend to be more satisfied with authoritarian leadership. It is significant also that, limited employee oriented leadership behavior can also affect the level of achievement of employee architects positively, as appreciation has a great impact on everybody. REFERENCES Cohen, A.R. Fink, S.L. Gadon, H. Willits, R.D. 2001. Effective Behavior in Organizations. USA: McGraw-Hill/ Irwin. Dessler, G. 2001. Management, Leading People and Organizations in the 21st Century. USA: Prentice-Hall. Ivancevich, J.M. Konopaske, R. Matteson, M.T. 2005. Organizational Behavior and Management. USA: McGrawHill. Johns, G. 1992. Organizational Behavior, Understanding Life at Work. USA: HarperColins. Luria, G. 2008. Controlling for Quality: Climate, Leadership and Behavior. The Quality Management Journal. 15(1): 27–40. McShane, S. & Von Glikow, M. 2000. Organizational Behavior. USA: McGraw-Hill. Newstrom, J.W. & Davis, K. 1993. Organizational behavior, human behavior at work. USA: McGraw-Hill. Sarros, J.C. Cooper, B.K. Santora, J.C. 2007. The character of leadership. Ivey business journal online, May/June.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Gendered behavior: Cultures in UK engineering and construction organizations B. Bagilhole Loughborough University, Leicestershire, UK

ABSTRACT: Over the past 20 years, numerous initiatives in the UK have attempted to redress the under-representation of women in engineering and construction (E&C), with limited impact. Modest increases in the proportion of women studying E&C have failed to translate into an equivalent increase in E&C professionals. The problematic experiences of women in E&C professions are a concern for academic researchers, although, more recently, research has moved attention away from concentrating on increasing the supply of women to the impact of institutional structures, systems and cultures that disadvantage women. This paper analyses cultures in E&C organizations as dynamic processes that impact on the beliefs, values and behaviors of their members. Gender is fundamental to this, particularly in male dominated professions. Specifically then, this paper investigates the subjective symbolic association between masculinity and femininity, cultural norms, and the working male model prevalent in E&C organizations, including: the inconsistent relationship between policy and culture; the long-hours culture; the conflict between family and work; gender stereotyping; socialization and identity; and networking and the career ladder.

1

WOMEN AND ENGINEERING AND CONSTRUCTION (E&C)

HESA (2007) statistics show that in the UK in2005/ 2006 33% of SET undergraduate students were women, significantly lower than the average across all subjects (57%). Only 14% of women undergraduate students in HE are studying SET related subjects, compared to 37% of men students. The in-crease in women students has failed to translate into an equivalent increase in women SET professionals, with figures suggesting that in 2006 women only ac-count for 13% of science, engineering and ICT professionals compared to an average 49% across all occupations (ONS 2007). The problematic experiences of women in the SET professions are a common concern for academic researchers. Research such as the ETAN report (European Commission 2000) have moved attention on from just increasing the supply of women in SET sectors to the impact of institutional structures, cultures and systems that disadvantage women. Studies have shown, for ex-ample, that women are not driven away from technology because of their lack of ability, but rather because of ‘an atmosphere of dominant masculinity’ (Sagebiel 2003). They find it much more difficult to cope with engineering values, systems and performance criteria which have been established by men for men, and not for women (Bagilhole 2002). Research on preconceptions about the nature and cultures of engineering and construction (E&C)

811

professions has mainly focused on school-age student’s perceptions of subjects and professions. However, it is important to note how preconceptions are formed—through a complex dialogue between the dominant cultural image and subsequent effects, thus reinforcing and creating the context in which E&C occupations are ‘understood’. Historically, E&C’s image has been tough, heavy and dirty and associated with machinery. This is not only because the workforce is predominantly male, but because the prevailing powerful cultures and ethos of E&C industries appear to be extremely masculine, helping to reproduce the perception that engineering careers are unsuitable for women. Within the technology sector the ideological resonance of ‘skill’ and technical ability is deeply entrenched in traditional masculine identity, and in strict opposition to traditional femininity. The established relationship between E&C and traditional notions of masculinities and the common-sense discourses that surround the E&C professional, both in society at large and within the sector in particular, highlights the deep contradictions that women E&C professionals face in organizational cultures. 2

GENDERED CULTURES

This paper adopts Brown’s (1995: 9) definition of organizational cultures as ‘the pattern of beliefs, values

incorporate equal opportunities or work life balance policies into the cultures of the organization.

and learned ways of coping with experience that have developed [and continue to develop] during the course of an organization’s history, and which tend to be manifested in its material arrangements and in the behavior of its members.’ It is important to emphasize the link between the cultures of an organization, informal and formal structures and the accepted/non-accepted behaviors of employees. Importantly, McIlwee and Robinson (1992: 5) suggest that workplace cultures are the medium in which gender behaviors fundamentally interact with opportunities created by organizational structure. Itzin (1995) powerfully characterized single-gender cultures as being hierarchical, patriarchal, sex-segregated, sexually divided, sex-stereotyped, sex-discriminatory, sexualized, sexist, misogynist, resistant to change, and with gendered power structures. The combination of these features forms a workplace where traditional masculinity is a dominant element of corporate cultures and the activities the work involves are associated with only one gender. The behaviors most valued and rewarded in E&C organizations are reflective of those typically associated with traditionally masculinity. As Hofstede states; ‘Women are not considered suitable for jobs traditionally filled by men, not because they are technically unable to perform these jobs, but be-cause women do not carry the symbols, do not correspond to the hero images, do not participate in the rituals or foster the values dominant in the men’s culture’ (Hofstede 2003: 16).

3

4

LONG HOURS CULTURE

Long hours culture is prevalent in E&C. Lingard (2004) found that in the construction industry sitebased employees work longer hours than employees based in the head or regional office, which can act as a major deterrent for women pursuing careers in these roles. Project-based working is argued to be particularly time-crucial, where E&C professionals work on a project until successful completion, giving as much time is needed (Davis 2001). Carter and Kirkup (1990) found that because engineering work is often task-oriented rather than time-oriented, there is often pressure to work long hours and for work to spill over into private time. This can be particularly significant for women, who often have more domes-tic responsibilities than men. The demands made on employees to be available at weekends and during the evenings, and the expectation that E&C professionals will work long-hours place pressure on individuals who may have commitments outside of the workplace.

5

ORGANISATIONAL POLICIES AND CULTURES

Within E&C professions an inconsistent relationship between organizational policies and cultures is marked. Despite the implementation of equal opportunities policies at an organizational level there exists a gap between reality and organizational rhetoric (Elvitigala et al. 2006). Paying ‘lip-service’, and the overt promotion of a liberal approach, to equal opportunities may be crucial in maintaining the dominant cultures (Bagilhole 2006). Work-life balance policies and practices have the potential to enhance opportunities for women in the workplace, but are often undermined by workplace cultures (Lewis 2001). Women perceive that to take up the policies would put them at a distinct disadvantage in comparison to their men colleagues, who rarely used such policies. Research in this area suggests that there is insufficient employer commitment to work-life balance; career progression is affected by balancing care and work; available work-life balance options are not always appropriate; and, organizational culture does not always permit the implementation of initiatives. It is therefore crucial for E&C organizations to

FAMILY VERSUS WORK

It is with regards to maternity leave and the return to work that the family/work conflict becomes particularly problematic for women in E&C professions. Childcare responsibilities are perceived to undermine career commitment regardless of the differences between women; non-mothers are conceptualized as a ‘risk’ and mothers as a ‘problem’ (Devine 1992). The Womeng (2006) study indicates that those who employ women have to justify their decision more strongly than if they had hired a man. Often women, who are only too aware of an existing ‘anti-family’ culture within organizations, will downplay their desire for a family when asked in interviews as ‘they knew they would not be employed if they challenged the interviewer on his assumptions’ (Devine 1992: 566). Women still experience clear discrimination surrounding the issue of maternity leave and the return to work; requests for part-time contracts are often agreed alongside some kind of demotion of position within the company. This lack of flexibility in the organization of contracts has long-term effects on women’s career progression. Thus there exists a ‘take-up gap’ between work-life balance initiatives and the barriers that discourage the utilization of these opportunities (Womeng 2006). The lack of flexible working, the low status and negative consequences associated with changing to part-time work was cited as one of

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the main reasons women decide to leave, or think of leaving E&C industries. 6

GENDER STEREOTYPING

Women are perceived as unsuitable for technical careers because of the dominant association between traditional notions of masculinity and technology (Webster 2005). It is claimed they are better communicators and directed towards the ‘soft’ side of E&C professions, such as sales, personnel, and deskbased work. However, Bennett et al. (1999) found that women in construction were pushed into the specialist technical skills rather than managerial roles. Within E&C industries ‘ghettoisation’ occurs; where men and women are concentrated in particular areas, and those areas that are dominated by women are culturally understood as lower status. Women in E&C professions are, in most cases, reluctant to openly question the stereotype of the association between traditional notions of masculinity and technology, despite the clear paradox their existence in the sector poses. In engineering, there are clear distinctions made between ‘real’ engineering (technicist) and other work in the sector (Faulkner 2005a) often utilizing the conceptual framework of the masculine/feminine continuum. Presumptions are made about the hands-on ‘tinkering’, technical abilities of professionals which form an important membership issue in how one belongs (Faulkner 2005a) and is part of the deeply gendered occupational cultures in E&C. On a more positive note, women who succeed in SET professions can experience a great sense of pride and status as a result of making it in a ‘man’s world’, which is expressed as a pleasurable and empowering experience. Thus, the decision to engage in E&C professions ‘‘should be understood within the context of the relative status and value attached to ‘men’s work’ as compared with typical ‘women’s work’’’ (Henwood 1996: 210). 7

OCCUPATIONAL SOCIALIZATION AND GENDER IDENTITY

Faulkner (2005a) suggests that the socialization processes that women experience communicates a clear way of ‘becoming and belonging’ as an engineer that often brings to the fore the question of gender authenticity that hangs over women engineers. Miller (2002) suggests that the strategies women develop to survive often involve adapting to the dominant masculine culture, rather than trying to change or challenge it. She suggests that the assimilation strategy used by the majority of women was ‘muted’; ‘there is an unawareness of the masculine nature of

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the context’ (Miller 2002: 157). Women what types of behavior are rewarded. Dryburgh (1999) terms this ‘professionalization’; internalizing the values, norms and symbols of the professional cultures. However, Miller (2002) highlights the fact that while women can learn masculine rules and behaviors, they cannot directly mirror them. Thus, while the coping strategies adopted by women may be extremely successful on a short term, individual basis, they serve to reinforce the gendered system, leaving little hope for long-term change. Etzkowtiz et al. (2000) found that women are typically viewed as ‘honorary men’ or ‘flawed women’ for attempting to participate in a traditionally man’s realm. Numerous research studies indicate that women who seek entry into cultures dominated by men either have to act like men in order to be successful, leave if they are not adaptable to the cultures, or remain in the industry without behaving like men but maintaining unimportant positions. Often women, when talking of their career decisionmaking and experiences with regard to E&C occupations, assert that it is their individual identity, rather than gender identity, that holds most influence (Carter and Kirkup 1990). For example, problems in confidence are individualized and attributed to personal failings. Henwood (1996) found her interviewees wished to avoid speaking explicitly of problems faced within the ‘boys culture’ and confusion about the concept of ‘equality’. Miller (2002) also found that Canadian women engineers conformed to beliefs and values consistent with a masculine value system. Sinclair suggests that, ‘these women enjoy the company of men, share interests and aspirations that are typically characterized as masculine, and perhaps seek their approval’ (2005, p.139). Maupin and Lehman (1994) also found that it was necessary to suppress or eliminate attitudes and behaviors that would identify individuals as ‘typical women’. Dryburgh maintains that this cultural adaptation is likely to include defining sexist behavior as exceptional, working hard to show solidarity with men colleagues and accepting uncritically the masculine cultures into which they are entering. Powell et al. (2006) found that women engineers often held traditionally stereo-typical views of women outside engineering and suggest that adopting an ‘antiwoman’ approach is a strategy adopted in order to succeed in the work-place. However, any career success among such women is unlikely to promote the interests of women in the sector generally. It also raises questions about the concept of a ‘critical mass’: the idea that once there is a sufficient proportion of women in engineering, the traditionally masculine cultures will no longer prevail. As Sinclair points out, by the time women achieve positions of formal power, they have learned and share similar influencing strategies to their men colleagues: ‘they have become

enculturated’ (2005: 110). However, despite ‘enculturation’, the women in Miller’s (2002) study still de-scribed feeling like an ‘outsider’. In spite of women engineers destabilizing gender roles by acting ‘like men’, at some point the salience of the perception that they are women takes precedence. 8

GENDERED NETWORKING AND THE CAREER LADDER

Despite the need to ‘fit in’ with traditionally masculine workplace cultures, informal networks within E&C organizations make this difficult for women. Social networks often revolved around traditionally masculine activities such as sport or drinking, spheres. Faulkner suggests that it can be difficult for women (and marginal men) to gain access to men’s networks, ‘not least because they bond through shared interests, humor, etc. at the golf course or over drinking sessions’ (2006: 12). The existence of ‘old boys’ networks’ is not unusual, based upon self-promotion, ‘game-playing’ and unwritten rules constructed by men. Also, because of the time that in-formal networking demands, women with family responsibilities are particularly disadvantaged. 9

CONCLUSIONS

Although significant inroads have been made in attracting women to HE courses in some E&C disciplines in the UK, women’s representation remains low across the E&C occupations. The problematic experiences of women in the E&C professions have been well researched over the last two decades, and it is suggested that women’s experiences have changed little over this period. The complex inter-play of gendered cultures which combine to shape women’s career opportunities continues. The cultures of E&C professions are still seen as a problematic arena for women to develop their careers within. The competitive nature of industry often means that arguments for increasing women’s entry to E&C has been based solely on business needs rather than a move towards inclusive cultures, and for women to progress they have to make accommodations which many find unpalatable. The dominance of traditionally masculine cultures in E&C professions is a key theme. While gender can be considered as only one aspect of culture, it is fundamental. The symbolic association between the masculine/feminine and various roles and modes of working have consequences for the people working in E&C organizations. They affect the formal and informal organizational structures, communicate ‘acceptability’ of particular identities and ways of being and reward or punish accordingly.

At the same time, there have been, some positive changes for women in E&C, in some workplace cultures. Faulkner (2005b), for example, identified a number of gender inclusive dynamics in engineering, including: respectful interactions between women and men engineers; wide-ranging and inclusive topics of conversation and humor; mixed-sex socializing and close friendships, and; care taken to avoid or challenge potentially offensive jokes and language. Some companies are making a real impact. However, despite some positive steps, the overriding conclusion is that career paths for women in SET organizations continue to be problematic. The established relationship between SET and traditional notions of masculinities and the discourses that surround the SET professional, both in society at large and within the sector, highlights the deep contradictions that women SET professionals face. The coping mechanisms women have been shown to adopt tend to be individualistic strategies, whereby the management of gender is seen to lie in women’s own hands, but such coping strategies have pre-dominantly failed to challenge the persisting cultures and structures in SET. Concerns over whether SET professions should be marketed to women given the barriers that they face have re-emerged throughout this period as increasing numbers of women have failed to prevent the reproduction of masculine cultures. Whilst the extent to which cultures can be consciously manipulated is contested ground, it is clear that without fundamental change the SET professions seem certain to remain problematic arenas for women to develop their careers within.

REFERENCES Auster, E.R. & Ekstein, K.L. 2005 Professional women’s mid-career satisfaction: an empirical exploration of female engineers. Women in Management Review 20(1): 4–23. Bagilhole, B. (2002), Women in non-traditional occupations: Challenging Men, Palgrave Macmillan, London. Bagilhole, B. (2006) ‘Family Friendly Policies and Equal Opportunities: A contradiction in terms?’ British Journal of Guidance and Counseling, Vol 34(3), 327–343. Beauvoir, S. (1949) The Second Sex. Translated and edited by H.M. Parshley (1993) London: David Campbell Publishers. Bennett, J.F., Davidson, M.J. and Gale, A.W. (1999) Women in Construction: A comparative investigation into the expectations and experiences of female and male construction undergraduates and employees. Women in Management Review 14(7): 273–91. Brown, A. (1995) Organizational Culture. 2nd Edition. Essex: Pearson Education Limited. Butler, J. (1990) Gender Trouble. London: Routledge. Carter, R. and Kirkup, G. (1990) Women in Engineering: A good place to be? Basingstoke: Mac-millan. Cole, L. (2000) (Ed.) Encyclopedia of Feminist Theories. London: Routledge.

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Davis, K.S. (2001) Peripheral and Subversive: Women Making Connections and Challenging the Boundaries of the Science Community. Science education. 85: 368–409. Devine, F. (1992) Gender Segregation in the Engineering and Science Professions: A Case of Continuity and Change. Work, Employment and Society 6: 557–75. Dryburgh, H. (1999) Work Hard, Play Hard: Women and professionalization in engineering—adapting to the culture, Gender and Society, 13 (5): 664–82. Elvitigala, G., Amaratunga, D., Haigh, R. and Ingirige, B. (2006) Women’s’ Entry into Construction: Does industry culture act as a barrier? Construction in the XXI Century: Local and global challenges, Rome, Edizionu Scientifiche Italiane. Equal Opportunities Commission (2007) The Gender Equality Duty. Available at: [Accessed: May 2007]. Etzkowitz, H., Kemelgor, C. and Uzi, B. (2000) Athena Unbound: The advancement of women in science and technology. Cambridge, Cambridge University Press. European Commission (2000) Science Policies in the European Union: Promoting excellence through mainstreaming gender equality. A Report from the ETAN Expert Working Group on Women and Science. Luxembourg: Office for Official Publications of the European Communities. Available at: [Accessed: August 2006]. Evetts, J. (1998) Managing the technology but not the organization: women and career in engineering. Women in Management Review, 13 (8): 283–90. Evetts, J. (1994) Women and Careers in Engineering: Continuity and change in the organization. Work, Employment and Society 8(1): 101–12. Faulkner, W. (2006) Genders In/Of Engineering: A research report. Edinburgh: University of Edinburgh. Available at: http:// extra.shu.ac.uk/ nrc/ section_2/ publications/ reports/Faulkner_Genders_in_Engineering_Report.pdf [Accessed January 2007]. Faulkner, W. (2005a) Becoming and Belonging: Gendered Processes in Engineering, pp. 15–25. In: J. Archibald, J. Emms, F. Grundy, J. Payne and E. Turner (Eds.) The Gender Politics of ICT. Mid-dlesex, Middlesex University Press. Faulkner, W. (2005b) Engineering Workplace Cultures: Men’s spaces and in/visible women? Public webcast lecture for the launch of ‘Science, Engineering and Technology: A course for women re-turners.’ 3 November 2005, The Open University, UK. Fox, M.F. (1998) Women in Science and Engineering: Theory, Practice, and Policy in Programs. Signs 24(1): 201–25. Gherardi, S. (1994) The Gender We Think, The Gender We Do in Our Everyday Lives, Human Relations, 47(6): 591–610. Grant, L., Kennelly, I. and Ward, K.B. (2000) Revisiting the Gender, Marriage, and Parenthood Puzzle in Scientific Careers. Women’s Studies Quarterly, 28(1/2): 62–85. Greenfield, S., Peters, J. Lane, N. Rees, T. and Samuels, G. (2002) SET Fair: A Report on Women in Science, Engineering and Technology from Baroness Greenfield CBE to the Secretary of State for Trade and Industry. Available at: [Accessed: May 2007].

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Henwood, F. (1996). WISE Choices? Under-standing Occupational Decision-making in a Climate of Equal Opportunities for Women in Science and Technology. Gender and Education 8(2): 199–214. Higher Education Statistics Agency (2007) All HE Students by Subject of Study, Domicile and Gender. London: HESA. Hofstede, G. (2003) Cultures and Organizations. Software of the Mind: Intercultural Cooperation and Its importance for Survival. First published in 1991. London: Profile Books. Itzin, C. (1995) The Gender Culture. In: C. Itzin and J. Newman (Eds.) Gender, Culture and Organizational Change: Putting Theory into Practice. Routledge, London, pp. 30–53. Kreinberg, N. and Lewis, S. (1996) The Politics and Practice of Equity: Experiences from Both Sides of the Pacific. In: L.H.Parker, L.J. Rennie and B.J. Fraser (Eds.), Gender, Science and Mathematics: Shortening the Shadow. Dordrecht, The Netherlands: Kluwer Academic Publishers. Lewis, S. (2001) Restructuring Workplace Cultures: The Ultimate Work-Family Challenge? Women in Management Review 16: 21–9. Lingard, H. (2004) The Work-life Experiences of Office and Site-based Employees in the Australian Construction Industry. Construction Management and Economics 22(9): 991–1002. Lingard, H. and Sublet, A. (2002) The Impact of Job and Organizational Demands on Marital or Relationship Satisfaction and Conflict Among Australian Civil Engineers. Construction Management and Economics 20(6): 507–21. McIlwee, J.S. and Robinson, J.G. (1992) Women in Engineering: Gender, power and workplace culture. Albany, State University of New York Press. Marshall, J. (1993) Patterns of Cultural Awareness: Coping strategies for women managers. In: B.C. Long and S.E. Kahn (Eds.) Women, Work and Coping. Montreal and Kingston, ON.: McGill-Queen’s University Press. Miller, G.E. (2002) The Frontier, Entrepreneurial-ism and Engineers: Women coping with a web of masculinities in an organizational culture. Culture and Organization, 8 (2): 145–60. Office for National Statistics (2007) Quarterly Labour Force Survey 2006. London: HMSO. Avail-able at: [Accessed May 2007]. Powell, A., Bagilhole, B. and Dainty, A. (2006). The Problem of Women’s Assimilation into UK Engineering Cultures: Can Critical Mass Work? Equal Opportunities International 25(8): 688–699. Rosser, S.V. (1998). Applying Feminist Theories to Women in Science Programs. Signs 24(1): 171–201. Sagebiel, F. (2003) Masculinities in organizational cultures in engineering: Study of departments in institutions of Higher Education and perspectives for social change. Presented at Gender and Power in the New Europe, 5th European Feminist Research Conference, 20–24 August, Lund, Sweden. Sinclair, A. (2005) Doing Leadership Differently. Melbourne: Melbourne University Press. UK Resource Centre for Women in Science, Engineering and Technology (2006) The Gender Equality Duty in Science, Engineering and Technology and How to Implement It.

Good Practice Guide. Bradford: UKRC. Available at: [Accessed: May 2007]. Wajcman, J. (1991) Feminism Confronts Technology. Cambridge: Polity Press. Wallsgrove, R. (1980) Towards a radical philosophy of science. In: The Brighton Women and Science group (Eds.) Alice through the microscope. London: Virago. Webster, J. (2005) Why are women still so few in IT? Understanding the persistent under-representation of women in

the IT professions. In: J. Archibald, J. Emms, F. Grundy, J. Payne and E. Turner (Eds.) The Gender Politics of ICT. Middle-sex, Middlesex University Press, pp. 3–14. Womeng Consortium (2006) Creating Cultures of Success for Women Engineers: Synthesis report. Available at: [Accessed: May 2007].

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Knowledge management (KM): ‘Integrating past experiences’ A. Weippert & S. Kajewski Queensland University of Technology, Brisbane, Australia

ABSTRACT: Architectural, Engineering, and Contractor (AEC) industry stakeholders have identified and/or customised various Knowledge Management (KM) solutions to meet their unique project needs and individual business objectives. Some have done so successfully, others not. Increased numbers of AEC industry leaders are starting to realise the importance of KM, but are still ‘confused’ by the mystifying array of KM solutions available to them. In an attempt to demonstrate leadership in facilitating the successful implementation and application of innovative KM initiatives within the highly fragmented and project-based AEC industry sector, this paper provides a summary of an extensive literature investigation into the current state of play of innovative KM applications and initiatives in both public and private AEC industry practices. This is the first major deliverable of an ongoing research investigation into developing an innovative AEC industry specific, value-adding and sustainable ‘KM Framework’. 1

INTRODUCTION

When it comes to knowledge management (KM)— commonly referred to as being highly controversial; hard to pin down; and meaning different things to different people—it is no surprise that today’s AEC industry leaders and stakeholders urgently seek best practice guidance, clarification, and know-how on a number of KM related issues. Further requesting timely responses to ‘when, how, where, who, what, and why’ they need to invest their valuable resources in implementing an innovative KM initiative—i.e. ‘what’s in it for me?’ This paper provides a select summary of an extensive literature investigation into the current state of play of innovative KM applications and initiatives in both public and private AEC and other leading industry sectors, and is the first major deliverable of an ongoing research investigation into developing an ‘innovative AEC industry-specific, value-adding and sustainable KM Framework’. The main objective of this ongoing international research undertaking is to: 1. Establish private and public case study projects that will foster the expansion of KM applications in the AEC sectors. 2. Demonstrate leadership in facilitating the use of innovative KM initiatives within AEC organisations and projects. 3. Identify appropriate KM solutions that will improve resource management; support and integrate total project life cycle considerations; and increase efficiencies on projects. 4. Promote KM benefits and efficiencies—thereby encouraging their wider adoption and promoting

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a much needed ‘knowledge sharing philosophy’ within the AEC industry. 5. Examine the cultural and sub-cultural ‘dynamics’ that challenge the implementation of any form of change within the AEC industry.

2

RELATIONSHIP BETWEEN DATA, INFORMATION AND KNOWLEDGE

‘‘The only source of knowledge is experience . . . everything else is just information’’ Einstein in Amin et al. (2001, p. 52) and Kazi (c2005, p. vii). To help better understand the ‘management of knowledge’, one needs to first differentiate and recognise the relationship between ‘data’, ‘information’ and ‘knowledge’. Knowledge, information and data are fluid and in a constant state of flux, and although they are inter-related with very distinct features and hierarchical relationships, it is not uncommon to see a ‘blurring’ of meanings between these (Collison & Parcell 2004; Dieng-Kuntz & Matta 2002; Mertins et al. 2001; Orna 2005). Figure 1 illustrates one way as to how these three relate to each other. Data, for example have no meaning or significance in themselves, whereas information is essentially data that has been processed to be ‘useful’ by aiming to provide answers to the questions: ‘Who?’, ‘What?’, ‘Where?’, and ‘When?’ Yet, this is not always the case. That is to say, by merely aggregating or collecting data and identifying certain relationships between variables does not guarantee its most effective utilisation or application (Sensky 2002).

Data

Managing Value

Understanding Relations

Information

Data Processing

Data Management Data Analysis

Information Management

Understanding Patterns

Knowledge

High

in social interaction among people (Nonaka & Takeuchi 1995)

SUPPORTING TECHNOLOGIES

Low

Understanding Principles

Figure 1. Relationships between data, information and knowledge.

Knowledge on the other hand is built-up from data and information as well as prior experiences, with the aim to answer the question ‘How?’ This inturn develops ‘new’ knowledge (from what already exists) and provides further answers to the question ‘Why?’ The entire knowledge process therefore hinges on the assumption that seeking, using and learning information and experiences is part of a holistic approach to making better sense of the world, driven by both external and internal factors that have seemingly infinite social, technical and economic implications. However, while knowledge is a prerequisite for understanding, the availability of appropriate knowledge does not necessarily guarantee understanding (Sensky 2002). ‘‘Information gets stale [yet] knowledge gains in strength [therefore] knowledge, unlike natural resources and other physical capital, is not depleted when it is used . . . it is expanded and open to further growth, refinement and marketability’’ (Stapleton 2003; Hari et al. 2005, p. 53). Although the terms information and knowledge are often used interchangeably, there is still a clear distinction between the two: 1. Knowledge, unlike information (referred to as the ‘flow of messages’), is about the beliefs and commitment of its holder; a function of a particular stance, perspective, or intension; whilst anchored to the very flow of information itself 2. Knowledge, unlike information, is about action— i.e. knowledge is essentially related to human action 3. Knowledge, like information, is about meaning— i.e. it is context specific and relational in that it depends on the situation and is created dynamically

Figure 1 further attempts to distinguish KM from data management (DM) and information management (IM), whilst illustrating their reliance on various supporting technologies—i.e. overlapping tools that support the management of data, information and knowledge, including, computer hardware and software, storage, indexing and retrieval systems, etc. Recognising the profound distinction between DM and IM is deemed important as it helps highlight the central role of methodologies and technologies that support the process of managing knowledge (KM) (Sensky 2002; Amin et al. 2001). 3

BRIDGING THE GAP BETWEEN KM AND DM

‘‘Without data management [DM] knowledge management [KM] is impossible’’ (Amin et al. 2001, p. 49). In order to successfully bridge the all to common gap (Fig. 2) that tends to exist between the technologies and processes associated with DM; and people and their collective ability to rapidly, willingly, and efficiently interact and collaborate their unique experiences (knowledge), requires managerial commitment and employee dedication from all private and public AEC industry stakeholder organisation levels to build a comprehensive and sustainable ‘knowledge-sharing culture’. 3.1

Confusing KM with IM

There is still a lot of confusion in defining the term KM and how it differs from traditional IM (Shukla & Srinivasan 2002; Kazi c2005.). Although KM means exactly that: managing of knowledge (Tiwana 2000), the better understanding and acceptance of KM is hindered by there not being a single definition that has (or even can be) universally agreed upon. Based on the findings of the in-depth literature investigation of this PhD study, the authors propose

Technology

Data Management (DM)

Process

Bridge the Gap Knowledge-sharing Culture

People

Knowledge Management (KM)

Collaboration

Figure 2. Bridge the gap between KM and DM.

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4

the following working/operational definition for the term KM—underpinning the development of the proposed innovative AEC industry-specific, value-adding and sustainable ‘KM Framework’: Knowledge Management is a multifunctional approach in achieving business objectives, creating business value and generating an enhanced competitive advantage, by making best use of knowledge and experience (knowhow); by focusing on innovative processes of acquiring, creating and sharing knowledge (in real time); whilst considering and aligning unique social (cultural) and technical dynamics more effectively and efficiently. IM on the other hand generally refers to one or all of the processes used for managing information within an organisation or project environment. IM is also used interchangeably to represent the management of information resources, information and communication technologies (ICT), and information processes—i.e. processes that enable the creation, production, organisation, storage, retrieval and distribution of information resources to support business decision-making (Standards Australia 2003). As a result, organisations have invested a considerable amount of IM resources over the last three decades, all of which contribute to today’s confusion and misrepresentation of KM:

As large generators and consumers of information, the AEC industry is becoming more and more aware of the information overload factor and that a large portion of corporate and project information assets are no longer under central control—i.e. it is not always clear how authoritative or legitimate the information is. In response to these validity challenges, certain industry leaders recognise that one productive solution would be to better access and manage the great amounts of knowledge and unique experiences created within the boundaries of their organisations and projects. Hence, the term Knowledge Management is increasingly being introduced to the various industry sectors of the globe (KMSciences 2006) and driven by the following potential benefits: 1. KM both makes and saves money: Research indicates that the sums of money saved are significant—i.e. hundreds of millions of Pounds, Dollars or Euros every year. 2. KM puts technology into perspective: It allows individuals and organisations to develop innovative systems and processes that are meaningful and relevant. 3. KM is about people: Establishing itself as the ‘‘management discipline of the decade . . . drawing attention to aspects which previously have often been neglected’’. Employees benefit from these innovative business processes—i.e. by enjoying work more, they contribute more, as well as learn and understand more—‘‘where a knowledgefriendly culture increasingly determines the success of the company as a whole’’ (Bahra 2001; Rollett 2003).

1. KM and IM employ many of the same ICT solutions: – Local Area Networks (LAN), Internet, Intranets, email, bulletin boards, etc. 2. Vendor Confusion: – Vendors still sell typical tools used for IM as KM tools—e.g. search engines could be used for a KM program, but not necessarily for the same purpose as in IM. – Confusion extends to vendors promoting any networking/ICT tool as a KM tool. 3. Confusing ‘databases’ with ‘knowledge bases’: – A typical database would seek to provide the right information to the right person at the right time at the right cost and with minimum effort in retrieval. – A knowledge base focuses on transfer of knowledge across different individuals or units; and application of knowledge across different contexts (Shukla & Srinivasan 2002). The above experiences are echoed in (Handzic 2004), confirming that certain ICT manufacturers and suppliers are going as far as simply renaming existing IM tools as KM tools in an attempt to support KM activities and industry requirements.

KM IN THE AEC INDUSTRY

The KM concept may be considered new to certain members of the AEC industry, but the importance, use, and sharing of knowledge has always been practiced—e.g. ‘‘a bricklayer passing on his knowledge to an apprentice, so that the skills continue into the next generation’’(SAI-Global 2004; c2003; Christensen 2003; Carrillo & Anumba 2002; Carrillo et al. 2000). Today’s senior AEC executives are starting to become more aware of innovative/non-traditional principles and associated KM philosophies (with a more ‘holistic approach’ to performance), and accepting these as being ‘‘central to the sustainability of the construction industry [where] its importance is reflected through the role of human and social capital’’ (Robinson et al. 2006). 5

SENSE OF URGENCY

‘‘Project-based industries, especially the construction industry, are under growing pressure to compete

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in new ways [hence] the development of measuring techniques capable of helping construction organisations move towards a knowledge culture is vital for today’s knowledge economy’’ (Egbu 2004, p. 308; Kululanga & McCaffer 2001, p. 353). Unfortunately, what is continually going astray is convincingly selling the KM concept/process to industry members, due to the lack in providing a proven methodology for implementing the idea to help organisations gain competitive advantage; and effectively measuring intellectual capital. What is required is for the AEC industry (as a whole) to undergo a paradigm and cultural shift in relation to measuring success in today’s competitive economy— moving it from concentrating purely on tangible (monetary) assets, to focussing more on intangible assets (knowledge, experience, etc.) (Carrillo et al. 2000). It is therefore no surprise why there is a current sense of obligation felt among KM leaders and AEC industry researchers to better understand how to access and mange the untapped wealth of taken-for-granted knowledge that is embedded in people’s minds (Hasan & Handzic 2003; Orange et al. 2000; Collison & Parcell 2004; Knight & Howes 2003).

the need to look after people (Rethinking Construction 2000). Over the last decade there is an emerging importance placed on the ‘people factor’ within the AEC industry—also referred to as the ‘‘most difficult resource for construction organisations to manage’’. This is mainly due to the intrinsic characteristics of the industry employing an extremely diverse range of people made up of a wide range of occupational cultures and professional backgrounds, all of which challenge the successful implementation of a ‘knowledge sharing philosophy’ (Pathirage et al. 2007). ‘‘Managing the knowledge assets of an organisation is just as much about managing people as it is about managing products and processes that are a result of their know-how or capability’’ (Bishop and Business-Excellence-Australia 2002, p. 3). AEC organisations who therefore fail to improve their negative attitude and lack of performance towards ‘respecting’ their own people and others, will fail to recruit and retain the best talent (knowledge and expertise) and business partners. This view is further echoed in (Linowes 1999) where holding on to good people is regarded as one of today’s most significant management challenge.

7 6

THE NEED FOR A KNOWLEDGE SHARING PHILOSOPHY

The AEC industry is predominantly made up of hierarchical structures and multi-disciplinary teams that that is generally perceived as not having a strong approach towards valuing its employees or their individual and collective contributions. This perception (be it accurate or not) makes it more difficult for AEC industry organisations and/or project team leaders to promote a ‘knowledge sharing philosophy’ amongst their employees, especially when sharing their tacit knowledge, due to it being regarded as personal property rather than organisational property. In earlier times, when people remained with their employers throughout their working lives, the preservation and dissemination of knowledge was not necessarily an issue. Yet in today’s dynamic employment arena, this is becoming increasingly challenging, with people working for a number of different employers throughout their lifetime, thereby increasing the loss of knowledge and experiences within organisations, inturn reducing overall productivity and effectiveness (competitive advantage) (Debowski 2006). ‘‘Knowledge not shared will not be fully used and may eventually be lost . . . [resulting in] wasted time and money, in addition to disgruntled staff’’ (SAIGlobal 2004, p. 3). The most urgent business challenges currently facing the industry are therefore not the implementation of innovative solutions, but rather

KNOWLEDGE SHARING BARRIERS

Successfully transforming organisations or project teams from a ‘knowledge is power’ culture to a ‘knowledge sharing is power’ culture (Skyrme 1998) is at the forefront of many of today’s AEC industry change leaders. The efficient distribution of the right knowledge from the right people to the right people at the right time is yet another ongoing challenge the AEC industry faces in its attempts to promote a ‘knowledge sharing philosophy’. This is further hindered by organisations reportedly rewarding ‘counterproductive’ behaviours of ‘knowledge hoarding’, rather than rewarding the productive behaviours of ‘knowledge sharing’ (Rollett 2003). Additional knowledge-sharing barriers that tend to originate from individual behaviours, perceptions, and actions, which then get intertwined within groups, teams or between business-functions of an organisation or project include: 1. The perceived or actual lack of time—to share knowledge or to identify colleagues in need of specific knowledge—including lack of contact time and interaction between knowledge sources and recipients. 2. Fear that sharing may reduce or jeopardise people’s job security—taking ‘ownership’ of intellectual property due to fear of not receiving ‘just’ recognition or accreditation from managers and colleagues.

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3. Limited awareness or understanding of the potential value and benefit gained from sharing knowledge with others. 4. Existence of a strong hierarchy (dictatorship), or position-based status (formal power to ‘pull rank’). 5. Insufficient or no procedures in place to effectively capture, evaluate, and disseminate past experiences (successes and failures) that would enhance individual and organisational learning. 6. Difference in experience and education levels. 7. Poor verbal/written communication and interpersonal skills of employees. 8. Age and gender differences—perceived generation gap. 9. No social networking skills or opportunities. 10. Lack of trust in people because they may misuse knowledge or take unjust credit for it. 11. Differences in national culture or ethnic background and the values and beliefs associated with it (including language, beliefs, habits, etc.) (Riege 2005). 12. Having the ‘not invented here’ syndrome— usually fuelled by a lack of trust in the accuracy or credibility of knowledge.

leadership in facilitating the use of innovative KM initiatives within the highly fragmented public and private AEC industry organisations and projects, future research undertakings (presented in separate papers) include: 1. Further identify and examine appropriate KM solutions (frameworks, processes, systems, etc). 2. Explore the relevant KM ‘dynamics’ associated to the AEC industry sector. 3. Establish case study projects to test, field trial and/or evaluate the use of innovative KM solutions and demonstrate the benefits and efficiencies (if any) obtained. 4. Further examine the AEC industry’s most valuable and influential knowledge resource—i.e. people and the complex dynamics of organisational and project team cultures and sub-cultures. 5. Promote a new ‘knowledge-based philosophy’ as an integral part of developing an ‘innovative AEC industry-specific, value-adding and sustainable KM Framework’.

9 8

FUTURE RESEARCH

‘‘The future belongs to those endowed with knowledge’’ (Nonaka & Takeuchi 1995, p. 7). The AEC industry has always excelled at managing complex programmes, often involving groups of people brought together for one-off projects and working in hazardous or inhospitable locations. As a result, the industry has developed both flexibility and superior skills in thinking on their feet and problem solving. What it is not so good at, however, is planning for the future (Foresight 2000). By presenting innovative KM solutions, frameworks and processes in a form that is ‘understandable’ to the AEC industry, will inevitably enhance the enthusiasm and willingness of industry members in recognising what positive results can be achieved on future projects through adopting sustainable KM initiatives (Jennex 2005). To achieve this, AEC industry members need to unlearn their traditional ways of doing things and grasp the importance of getting out of the ‘‘old mode of thinking that knowledge can be acquired, taught, and trained best through manuals, books, or lectures’’. Instead they need to focus more on the less formal and systematic side of gaining knowledge—i.e. by concentrating on highly subjective insights, intuitions and hunches that are gained through experience, pictures, and metaphors, etc (Nonaka & Takeuchi 1995). Consequently, to meet the aims and objectives of this research project and in an attempt to demonstrate

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CONCLUSIONS

If properly executed, developing an ‘innovative AEC industry-specific, value-adding and sustainable KM Framework’ that supports the entire knowledge lifecycle of a project, will arguably bring about significant benefits to AEC organisations by increasing productivity levels (through the efficient distribution of the right knowledge from the right people to the right people at the right time); providing better client service and enhanced satisfaction (through the collective learning and capture of historic project knowledge, lessons learned, and experiences gained); and helping solve wilful problems (by connecting/bridging relevant expertise and experiences). On completion, AEC industry organisations are expected to not only embrace the innovative and sustainable concept of the author’s newly proposed KM framework, but also be able to apply its relevant and interdependent principles and processes as a new way of doing business and managing of assets (knowledge and experiences) in order to facilitate continuous improvement and enhanced organisational performance (competitiveness). Finally, by fostering the proposed paradigm and cultural shift within the AEC industry, moving it towards embracing a ‘knowledge sharing philosophy’, will ensure that each employee (regardless of geographic location) is empowered to contribute towards, assimilate, and draw on a central collection of relevant and validated knowledge and experiences. This inturn will integrate the knowledge-based AEC industry and its members in a unique way.

REFERENCES Amin, A., S. Bargach, et al. 2001. ‘‘Building a Knowledge Sharing Culture’’ 2007, from www.slb.com (white paper). Bahra, N. 2001. Competitive Knowledge Management, Basingstoke Palgrave. Bishop, K. and Business-Excellence-Australia 2002. New Roles, Skills and Capabilities for the Knowledge-focused Organisation, Sydney Standards Australia International. Carrillo, P.M. and C.J. Anumba 2002. ‘‘Knowledge Management in the AEC Sector: An Exploration of the Mergers and Acquisitions context.’’ Knowledge and Process Management p. 149 9/3(Jul/Sep): 149–161. Carrillo, P.M., C.J. Anumba, et al. 2000. ‘‘Knowledge Management Strategy for Construction: Key IT and Contextual Issues’’ Construction Information Technology 2000. Taking the construction industry into the 21st century (June 28–30). Christensen, P.H. 2003. Knowledge management: perspectives and pitfalls. Copenhagen, Copenhagen Business School Press. Collison, C. and G. Parcell 2004. Learning to Fly: Practical Knowledge Management from some of the World’s Leading Learning Organizations, Chichester, West Sussex Capstone. Debowski, S. 2006. Knowledge Management, Milton, Qld, John Wiley & Sons Australia. Dieng-Kuntz, R. and N. Matta 2002. Knowledge Management and Organizational Memories, Boston Kluwer Academic Publishers. Egbu, C. 2004. ‘‘Managing Knowledge and Intellectual Capital for Improved Organizational Innovations in the Construction Industry’’ Engineering, Construction and Architectural Management 11(5): 301. Foresight 2000. Constructing the Future: Making the Future Work for You, UK, Foresight, Built Environment and Transport Panel, Construction Associate Programme. Handzic, M. 2004. Knowledge Management: Through the Technology Glass, New Jersey; Singapore World Scientific. Hari, S., C. Egbu, et al. 2005. ‘‘A Knowledge Capture Awareness tool: An Empirical Study on Small and Medium Enterprises in the Construction Industry’’ Engineering, Construction and Architectural Management 12(6): 533. Hasan, H. and M. Handzic 2003. Australian Studies in Knowledge Management. Wollongong, N.S.W., University of Wollongong Press. Jennex, M.E. 2005. Case studies in Knowledge Management. Hershey PA Idea Group Pub. Kazi, A.S. c2005. Knowledge Management in the Construction Industry: A Socio-technical Perspective. Hershey PA Idea Group Pub. Keyes, J. 2006. Knowledge management, Business Intelligence, and Content Management: The IT Practitioner’s Guide. Boca Raton, FL, Auerbach Publications. KMSciences 2006. ‘‘Knowledge and Collaboration for Project Managers.’’ from http://www.findwhitepapers. com/ and www.kmsciences.com Knight, T. and T. Howes 2003. Knowledge Management— A blueprint for Delivery: A Programme for Mobilizing Knowledge and Building the Learning Organization. Oxford, Butterworth-Heinemann. Kululanga, G.K. and R. McCaffer 2001. ‘‘Measuring Knowledge Management for Construction Organizations’’ Engi-

neering Construction and Architectural Management 8 (5–6): 346–354. Linowes J.G. 1999. ‘‘Invest in the Best, Communications.’’ Journal of Management in Engineering (November/ December). Mertins, K., P. Heisig, et al. 2001. Knowledge Management: Best Practices in Europe. New York Springer. Nonaka, I. and H. Takeuchi 1995. The Knowledge-creating Company: How Japanese Companies Create the Dynamics of Innovation. New York Oxford University Press. Orange, G., A. Burke, et al. 2000. ‘‘Organisational Learning in the UK Construction Industry: A Knowledge Management Approach.’’ 2007, from http://csrc.lse.ac.uk/asp/ aspecis/20000202.pdf Orna, E. 2005. Making Knowledge Visible: Communicating Knowledge through Information Products. Aldershot Gower. Pathirage, C.P., D.G. Amaratunga, et al. 2007. ‘‘Tacit Knowledge and Organizational performance: Construction Industry Perspective.’’ Journal of Knowledge Management 11(1): 115–126. Rethinking Construction 2000. A Commitment to People: ‘‘Our Biggest Asset’’ UK, Report from the Movement of Innovation’s working Group on Respect for People, Rethinking Construction, Department of Trade & Industry. Riege, A. 2005. ‘‘Three-dozen Knowledge Sharing Barriers Managers Must Consider.’’ Journal of Knowledge Management 9(3): 18–35. Robinson, H.S., C.J. Anumba, et al. 2006. ‘‘STEPS: A Knowledge Management Maturity Roadmap for Corporate Sustainability.’’ Business Process Management Journal 12(6): 793–808. Rollett, H. 2003. Knowledge Management: Processes and Technologies. Boston Kluwer Academic Publishers. SAI-Global 2004. Introduction to Knowledge Management in Construction, Sydney, Standards Australia International. Sensky, T. 2002. ‘‘Knowledge Management’’ Advances in Psychiatric Treatment 8: 387–396. Shukla, A. and R. Srinivasan 2002. Designing Knowledge Management Architecture: How to Implement Successful Knowledge Management Programs. New Delhi; Thousand Oaks, CA Response Books. Skyrme, D. 1998. ‘‘Know Why.’’ 2007, from http://www. skyrme.com/pubs/kwhyhow.htm Standards Australia, I. 2003. Knowledge Management: Interim Australian Standard. Sydney Standards Australia International. Stapleton, J.J. 2003. Executive’s Guide to Knowledge Management: The Last Competitive Advantage. Hoboken, N.J., J. Wiley & Sons. Tiwana, A. 2000. The Knowledge Management Toolkit: Practical Techniques for Building a Knowledge Management System. Upper Saddle River, NJ, Prentice Hall PTR. Von Krogh, G., I. Nonaka, et al. 2000. Enabling Knowledge Creation: How to Unlock the Mystery of Tacit Knowledge and Release the Power of Innovation, New York Oxford University Press. Yeh, Y., S. Lai, et al. 2006. ‘‘Knowledge Management Enablers: A Case Study.’’ Industrial Management & Data 106(6): 793–810.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Managing innovative change within organisations and project team environments A. Weippert & S. Kajewski Queensland University of Technology, Brisbane, Australia

ABSTRACT: In an attempt to enhance the efficiency, productivity and competitiveness of today’s Architectural, Engineering, and Contractor (AEC) industry, this paper summarises the current status of an ongoing PhD research investigation in developing a sustainable AEC industry specific best-practice ‘Innovation-driven Change Framework’—more specifically a summation of the ‘fourth interrelated dynamic’ (culture). Leveraging off the outcomes of a two year industry and government supported Cooperative Research Centre for Construction Innovation (CRCCI) research project, as well as referring to recent internationally renowned case studies and related literature investigations, this research investigation includes further identifying, processing, analysing and categorising various culture change methods, models, frameworks and processes utilized within the AEC and other industry sectors, and incorporating these findings in developing an AEC industry-specific ‘Innovation-driven Change Framework’.

1

INTRODUCTION

‘If the construction industry is to build core competencies, maintain capability, and benefit from innovation, it has to change from an adversarial and blame cultures to a sharing culture’ (Egbu 2004, p. 313). Changing the AEC industry’s culture and subculture (beliefs, attitudes, values, etc) is perhaps one of the last available ‘mechanisms’ left for it to enhance and maintain its current levels of competitiveness within the International construction arena. Consequently, there is an urgent need for AEC industry leaders to recognise the various social dynamics that are likely to influence the successful implementation of a sustainable innovative-driven change initiative within organisations, groups or project team environments. Unfortunately, this ‘transformation’ of AEC industry ‘personalities’ and traditional work practices is not easy. When the implementation of a new innovation-driven change initiative, solution or process urges the need for changing an organisation’s culture, senior management have to realise (from the outset) that hierarchical-imposed solutions usually do not work well when sub-cultural differences and conflicting assumptions are involved. Instead, new intercultural processes are to be developed, permitting better communication between the sub-groups, and allowing the strengths of each to interact to form an integrative and new implementation solution (Schein 1997). Yet, if this process is not undertaken and managed correctly, then the old (traditional) and the new

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practices will superficially and temporarily co-exist, resulting in the organisation’s original ‘way of doing things’ to eventually resurface (Palmer et al. 2001). In an attempt to enhance the efficiency, productivity and competitiveness of today’s AEC industry, and to meet the objectives of this PhD research undertaking, this paper discusses the current status of an ongoing and in-depth investigation in developing a sustainable AEC industry specific best-practice ‘Innovation-driven Change Framework’ that comprises of six key, interconnected and interdependent dynamics that the authors believe are to be considered when any form of change or new way of ‘doing things’ is proposed within an organisation or project team environment (Fig. 1)—i.e. a framework that can (a) strategically be employed within AEC industry project-based organisations, business plans and project team environments to help determine their current levels of ‘readiness’ and ‘adaptability’ towards the sustainable implementation, leadership and management of an innovation-driven change initiative; (b) provide AEC industry leaders with a set of best practice guidelines that can assist in identifying, assessing and potentially overcoming the ‘deeply embedded’ cultural and sub-cultural ‘threats’ that challenge the successful uptake of a innovation-driven change initiative, whilst simultaneously recognising efficient and value-adding ways on how to ‘embrace’ the myriad of potential ‘opportunities’ a new way of ‘doing things’ may offer—i.e. by looking at organisation and team member values, attitudes, perceptions and beliefs, etc. towards the implementation of an innovation-driven change initiative; and (c) assist

CHANGE

1

TRAINING AND EDUCATION

6

LEADERSHIP

5

‘KEY AND INTERDEPENDENT DYNAMICS THAT SHOULD BE PART OF THE DECISION-MAKING PROCESS IN THE INTRODUCTION OF ANY FORM OF INNOVATIONDRIVEN CHANGE WITHIN AN ORGANISATION OR TEAM’

2

3

INNOVATION

IMPLEMENTATION

4 CULTURE

Figure 1.

Innovation-driven change framework. Figure 2.

change leaders in identifying and quantifying the need, expectations, preparedness, willingness, and sustainability of accurately ‘matching’ the implementation of an innovation-driven change initiative with that of the uniquely interwoven and profoundly entrenched cultures and sub-cultures within AEC industry organisations, groups and project team environments. NOTE: Due to the restrictions imposed on the length of this paper, the authors will only focus on providing a select summation pertaining to the ‘fourth interrelated dynamic’ of the ‘Innovation-Driven Change Framework’—that of culture.

2

The Iceberg of culture.

a fundamental part of any decision-making process (Duarte & Snyder 2001).

3

WORKING DEFINITION FOR THE TERM ‘CULTURE’

Based on the findings of the in-depth literature investigation of this PhD study, the author proposes the following working/operational definition for the term ‘culture’—underpinning the fourth interconnected dynamic of the proposed sustainable AEC industry specific best-practice ‘Innovation-driven Change Framework’ (Fig. 1):

SO WHAT IS CULTURE?

In the case of humans, the term ‘culture’ is often the primary way in which one group (organisation, team, etc) differentiates itself from another. Based on various literatures identified during this PhD investigation, the term ‘culture’ has a wide range of similar and even contradictory definitions. One way of portraying the characteristic patterns of an individual’s or group’s behaviour and cultural elements of its culture is by referring to a diagram of an ‘iceberg’ (Fig. 2). Briefly, the ‘Iceberg of Culture’ illustrates that the behaviour, attitudes, and values of members is dependent upon the sets of both conscious and unconscious beliefs that individual members possess, and that these beliefs are seen as a ‘key element’ of organisational culture (Williams et al. 1993). Another way of interpreting the Iceberg of Culture is as ‘hidden scripts’ created by repeated interactions between members of a group, which are used (consciously and unconsciously) to guide their behaviours. These, over time, become ‘invisible’ and ‘second nature’, serving as ‘shortcuts’ for guiding daily actions, creating perceptions or assumptions, and forming

Culture is a collection of common experiences, standards, assumptions, perceptions, morals, beliefs, and ‘ways of thinking’ that both ‘represent’ and ‘influence’ the way ‘things are done’ by members within an organisation, group, or team environment.

4

NEED FOR CULTURE CHANGE

The relevance and importance of organisational culture and the need for it to be changed has increased over the last decade or so. Statements by Williams et al. (1993) on culture and the need for change confirms many organisations decide to change their existing culture based on the need to implement a strategy-driven change initiative, usually due to a certain ‘crises’ or ‘opportunity’ being identified. Many organisations are therefore driven to change due to business demands (crises or opportunity), which inturn insist on the need to change the existing culture itself. ‘If you do not see a truck racing towards you, you are unlikely to jump out of the way . . . likewise . . . if you do not realise that you are standing on a treasure

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of gold, you are unlikely to bend down and pick it up . . . [therefore] . . . if people fail to see the need for change (whether threat or opportunity driving it), they will not change’ (Black & Gregersen 2002, p. 20). The above extract reinforces the importance for organisations, groups and project teams to realise and create a need for change, before the act of change can take place. Unfortunately, to convince people of the need for change is easier said than done, because (a) people tend not to see even the most obvious threats and opportunities, and (b) being ‘blinded’ by the ‘way we have always done things around here’. 5

CULTURE CHANGE METHODS

beliefs and attitudes within the organisation or project team—i.e. promote cultural change. In this case, recruitment, selection and redundancy are frequently part of the change process, but because employee commitment and ‘positive’ culture is recognised as being essential to the long-term survival of a company, it is suggested to do this only once—i.e. make one large ‘cut’ or ‘reshuffle’ rather than a series of small ones. 2. Changing People’s Beliefs and Attitudes: Due to beliefs of individuals directly being influenced by, or formed through observation, interaction, participation, and ‘persuasive’ communication, one or more of the following methods for changing the beliefs, attitudes and values of employees are recommended:

Literature investigations identified and assessed a number of alternate methods and approaches of employing culture change initiatives within organisational or project team environments. The most relevant to this PhD study are extracted, dissected and then reintroduced into the development process of an AEC industry-specific ‘Innovationdriven Change Framework’. Only one of these culture change methods will be discussed for the purpose of this paper. 5.1 Six key methods of changing culture The ‘culture change mechanism’ by Williams et al. (1993), comprises of six key methods (Fig. 3) the author of this PhD study believes is relevant and worth considering by change leaders when attempting changing the culture of an organisation: The following provides a brief description of the above six key methods and their relevance to successfully changing the culture of an organisation: 1. Changing People: By changing people (particularly those in key positions or with more ‘uncompromising’ attitudes) one may change the pattern of

i.

CHANGING People

Through: Use of role models (Champions) Participation Use of formal communication Counselling Management Education

ii. CHANGING Beliefs and Attitudes

vi. CHANGING Corporate Image

CHANGING ORGANISATION CULTURE v. CHANGING Structures, Systems and Technology

iii. CHANGING Behaviour

iv. CHANGING Places

Figure 3.

Six key methods of changing culture.

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– Through use of role models/leaders (Champions): By recognising the importance of senior/ key individuals, acting as role models or champions to achieve the desired attitudes and behaviours of employees. Simply through observation, for example, people are likely to ‘imitate’ those behaviours that they believe are likely to lead to individual and/or overall success. – Through participation: Formalised group discussions, such as morning meetings, team briefings, etc., are alternate methods for developing shared beliefs and attitudes. Certain organisations may find formalised discussions (when held in an appropriate climate) increase the involvement of employees—i.e. (a) encouraging identification and commitment to a team, task or organisation; (b) improving organisational communication and control; (c) requiring group experience in problem solving activities; and finally (d) promoting participative management and decisionmaking practices. – Through use of formal communication: Although a common process, attempts in changing employee beliefs and attitudes can be used more effectively to better ‘communicate’ the organisation’s culture to their employees and local community from which they recruit the majority of their employees. This can, for example, be done through (a) in-house or external corporate advertising media groups, and (b) by publishing articles entitled ‘protecting customer investment’; ‘putting the customer first’; or ‘responding to change demands’. – Through counselling: When, for example, there is a need for an organisation to make significant reductions in staff in order to cut costs and improve its profitability, it is difficult to promote a proactive and positive culture at the same time. Therefore, once the process of informing management is complete, and there is complete commitment to the change from the most senior

level, it is suggested each level of management carries out one-to-one interviews with their employees—before any changes are made or publicly announ- ced, explaining in detail the intended changes and implications for the individual concerned, and defining what would be expected of those working in the ‘new’ organisation. – Through Management Education: Educating management is a central strategy for many organisations to achieve a cultural change. Where, for example, the top two or three levels of management are sent on external and professionally run courses, and where external consultants are then brought in to advise and run a ‘customised’ internal program to help ‘cascade’ this newly acquired knowledge and management process down to the rest of the staff and project team members. 3. Changing Behaviour: The above confirms changing culture is a matter of changing values and attitudes, rather than teaching people new techniques or replacing old procedures or processes with new ones. However, research indicates that training new skills is likely to change people’s behaviour, beliefs and attitudes towards their capabilities in, for example producing a new product or outcome, or accepting the ‘new way of doing things’. 4. Changing Places: Sub-cultures within an organisation develop around differences in functions, roles, and levels of its members. Therefore, in order to promote the existing and overall culture of an organisation, one can ‘reshuffle’ or ‘rotate’ groups and/or individuals with different knowledge and learning experiences and move them into key positions within other sub-cultures (groups, teams, etc)—e.g. certain key personnel/culture change champions (managers, project leaders, etc.) who have been in a certain position for a certain period, agree to be moved/rotated to other sections, departments or projects. These changes can result in improved performances, attitudes, beliefs and values in not only the department or project team they are relocated to, but also the one they left behind. 5. Changing Structures, Systems and Technology: Changing the structure of an organisation or project team will usually make an ‘unpredictable’ impact on its culture—e.g. influencing existing work groups and communication networks. However, by revising and improving existing reward, appraisal, and incentive programs, and by better monitoring budgeting and control systems (which are said to be linked to specific individual and group behaviours) will inturn increase the chances of changing people’s beliefs and attitudes towards performing in a ‘required way’.

6. Changing the Corporate Image: Developing a corporate image typically develops a positive attitude among both internal and external clients and employees whilst enhancing their commitment towards the organisation, projects, etc.—also described as ‘optimistic expectations’. A positively influencing corporate image can be achieved through a name, a logo, a slogan, unique advertising, publication of successes, promoting and facilitating social events, encouraging employee and family involvement, etc. Interestingly, many organisations and project teams who successfully change their culture and sub-cultures initially do not have ‘culture change’ as their main aim or objective. Instead their main objective, in many cases, is to for example: successfully implement a new business strategy or introduce a new company policy. This process consequently and very much coincidently, also changes employee beliefs and attitudes and thus the culture of the organisation itself. 6

RESISTANCE TOWARDS CHANGE

Literature confirms that the most recognised and welldocumented findings from studying individual and organisational behaviour (culture), is their unequivocal resistance towards change (Robins 1998). The following provide an outline of three sources/ types of resistance towards the implementation of innovation-driven change initiatives experienced within organisations and project team environments— i.e. AEC industry (in general), individual, and organisational. 6.1

AEC industry resistance towards change

The industry has to realise that investing in an innovation-driven change initiative, such as the implementation of a new technology, is no longer viewed as simply purchasing a new piece of hardware or software, but more so as a potential and long term investment in the process of change itself (Cleveland 1999; Buch & Wetzel 2001). Unfortunately, the nature of the industry’s constructed products, and its project-based organisations and processes, limit the successful implementation of technology-led or innovation-driven change (Gann 1997). This inherent form of resistance exists due to numerous unique and AEC industry-specific factors, including: 1. The site-based and geographically dispersed nature of projects. 2. The processes of erecting, assembling and installation of its unique ‘products’ and services. 3. The need for durable and sustainable products and processes, which inturn signifies a preference to

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Table 1.

Five individual sources of resistance to change.

Five sources of resistance Fear of unknown Selective information processing

Economic factors

Security Habit

Individual characteristics Employees/project team members dislike uncertainty within their immediate and overall work environment. Therefore, if the reasons and benefits of implementing an innovation-driven change initiative is not clearly articulated (clouded or disguised), they may develop a negative attitude or behave ‘dysfunctional’ toward the newly way of ‘doing things’. Individuals tend to shape their world through their perceptions of their environment. Once created, this ‘perceived world’ tends to ‘protect’ itself by resisting any form of change—i.e. employees/project team members ‘selectively’ process information in order to keep their perceptions intact, by for example: Only hearing what they ‘want’ to hear, Ignoring any arguments, statistics or potential benefits the implementation of a change initiative will provide them or may challenge the ‘world they’ve created’ around them. Similar to the above security factor, the threat that the implementation of a innovation-driven change initiative may changes current job tasks or established work routines, which in turn may lower employee income is yet another source of individual resistance. These ‘economic fears’ surface due to employees being concerned that they won’t be able to ‘perform the new tasks or routines to their previous standards, especially when pay is closely tied to productivity’. Certain people have a ‘high need for security’ and are likely to resist change as it ‘threatens their sense of safety.’ Fearing that the implementation of a technology-led solution or innovation-driven change initiative may challenge their current employment status, need of expertise or level of skills. Humans are habitual in nature—‘creatures of habit.’ Life’s ‘complexities’ are challenging enough— being faced with hundreds of decisions to make every day. To cope with these ‘complexities’ and stressful work environments, employees and project team members all rely on numerous ‘habits’ and ‘programmed responses’ that have developed and become entrenched within them over time. When confronted with change, the tendency to respond in ones ‘habitual’ ways itself becomes a source of resistance. By simply ‘changing places’ can disrupt an individual’s habits—e.g. when a department decides to move to a new office building or when a team moves onto the next project, inevitably means employees are likely to have to change certain basic yet established ‘habits’, such as: Building new working relationships—i.e. identifying new personal/professional ‘boundaries’ and ‘commonalities.’ Taking a new street to work (traffic, etc). Finding a new parking space (availability, cost, etc).

continually use (what the construction industry considers to be) ‘tried and tested’ (traditional) techniques, tools, systems, processes, etc. 4. Buildings and structures becoming more complex—often involving the integration of expensive, untried and incompatible systems, products, processes and services. 5. Legacy of ‘sunken costs’ when investing in the implementation of technology-led or innovationdriven change initiatives. ‘Resistance to change is a concern even in organisations where innovation and change are part of the culture’ (White & Bruton 2007, p. 101). 6.2

Individual resistance towards change

From an individual’s perspective (employees, project team members, etc.), resistance towards innovationdriven change resides in the ‘basic human characteristics’, which includes ones ‘perceptions, personalities and needs’. To help illustrate this point, Table 1 provides five reasons as to why individuals resist any form of change.

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7

CONCLUSIONS

Culture is one of the most difficult and complex dynamics to understand let alone change. This is mainly due to culture being defined and perceived in so many different and sometimes conflicting ways. Culture is also described as being both influencing and being influenced by various conscious (visible) and unconscious (hidden) factors that determine the unique way in which members of an organisation, group or project team address and interact with one another. Senior managers and change leaders therefore need to seriously consider the various features, characteristics and qualities of its culture and sub-cultures, and make it an integral part of their decision-making process when implementing an innovation-driven change initiative into an existing work environment. Unfortunately, this ‘transformation of personalities’ is not easy, as it is hindered by the AEC industry’s exceptionally fragile level of trust, its reluctance to share knowledge and experiences, its inherent and deeply entrenched resistance to doing things differently, and its multiple levels of sub-cultures,

organisations continue then the application of even the most optimum or fail-proof implementation strategy may in many cases be worthless. That is to say, if an organisation’s most valuable resource (people) is not properly aligned with, and fully supportive of the implementation strategy of an innovation-driven change initiative or ‘new way of doing things’, then the success and sustainability of the initiative his highly likely to either stall or eventually fail (Schneider 2000).

each with their own unique perceptions, beliefs, value sets and attitude towards a proposed innovation-driven change initiative. The development of the author’s proposed AEC Industry-Specific ‘Innovation-Driven Change Framework’ is based on the premise that should one of the six interdependent dynamics (Fig. 1) be influenced or changed, it will affect at least one of the other five dynamics. That is to say, should any of the six interdependent AEC industry-specific innovation-driven change dynamics be ignored, disobediently followed, mislead/ managed/monitored, or simply overlooked, then the implementation process of a proposed innovationdriven change initiative or ‘new way of doing things’ is unlikely to succeed. For example, senior management and change leaders can not simply decide to implement a new business strategy or process without simultaneously:

REFERENCES

1. Confirming the organisation’s ability to change and willingness to adapt to a new or innovative way of ‘doing things’. 2. Considering the most suitable implementation strategy. 3. Correctly identifying and fully understanding the various dynamics of their inherent culture and sub-cultures, including individual beliefs, values, attitudes, assumptions, etc. towards the proposed change; developing unique motivation strategies, rewards and incentive packages for managers, employees and project team members alike; and promote a trusting and sharing environment that is willing to commit and fully embrace the proposed change. 4. Clearly identifying effective leaders and committed champions who are capable of achieving the implementation goals and objectives and who facilitate appropriate working environments that encourage innovative thinking; an ‘atmosphere of creativity’ where staff enjoy the freedom to solve complicated problems in non-traditional ways; the desire to institutionalise innovation as standard business practice; and the support for collaborative efforts. 5. Facilitating both internal and external training incentives that encourage lifelong learning, ongoing development, and active creation and sharing of knowledge to help ensure the successful implementation of innovation-driven change initiatives within existing and future work environments.

Bate P. 1996. Strategies for Cultural Change, Oxford, Butterworth Heinemann. Black J.S. and Gregersen H.B. 2002. Leading Strategic Change: Breaking Through the Brain Barrier. New Jersey, Pearson Education Inc. Buch K. and Wetzel D.K. 2001. ‘‘Analysing and Realigning Organizational Culture.’’ Leadership & Organization Development Journal 22(1). Cleveland Jr. A.B. 1999. ‘‘Knowledge Management: Why It’s Not an Information Technology Issue?’’ Journal of Management in Engineering (November/December). Duarte D.L. and Snyder N.T. 2001. Mastering Virtual Teams: Strategies, Tools, and Techniques that Succeed, San Francisco, Jossey-Bass Inc. Egbu, C. 2004. ‘‘Managing knowledge and intellectual capital for improved organizational innovations in the construction industry’’ Engineering, Construction and Architectural Management 11(5): 301. Gann D. 1997. Technology and Industrial Performance in Construction, Brighton, UK, Prepared by University of Sussex, for OECD Directorate for Science, Technology and Industry. Palmer I., Dunford R., et al. 2001. ‘‘Changing Forms of Organizing: Dualities in Using Remote Collaboration Technologies in Film Production.’’ Journal of Organizational Change Management 14(2). Robins, S.P. 1998. Organisational Behaviour: Concepts, Controversies and Applications. New Jersey, Prentice Hall Inc. Schein E.H. 1997. Organizational Culture and Leadership, San Francisco, Jossey-Bass Inc. Schneider W.E. 2000. Why Good Management Ideas Fail: The Neglected Power of Organisational Culture. Strategy & Leadership. White M.A. and Bruton G.D. 2007. The Management of Technology and Innovation: A Strategic Approach. Canada, Thomson South-Western. Williams A., Dobson P., et al. 1993. Changing Culture: New Organisational Approaches, London, Institute of Personal Management (IPM).

In conclusion, should the lack of attention to the human aspects (culture and sub-cultures) of today’s highly competitive AEC industry and project-driven

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Personality types of civil engineers and their roles in team performance K. Gautam & A. Singh Department of Civil and Environmental Engineering, University of Hawaii at Manoa, Honolulu, USA

ABSTRACT: Though the Myer’s-Brigg’s Type Indicator (MBTI) has been a popular tool for studying personality types and traits of personnel in many professions for many decades, its application to construction management and civil engineering has been limited. Most of the few researchers who studied personality types of civil engineers argued that the dominant personality types of engineers were introversion, sensing, thinking and judging (ISTJ). Results of their research were compared and it was seen that ISTJ may not always be the dominant personality type in large engineering groups. This has significance in understanding personnel behavior on projects so as to assign the right people to the right jobs, and ensure improved communication and project success. 1

INTRODUCTION

This paper dwells on the personality types of civil engineers and attempts to explore how personality types play roles in cross functional teams. The major objective of this paper is to identify common personality types of civil engineers by synthesizing the findings of various studies. In addition, two major departments of a State Public Agency studied by Johnson & Singh (1998) are analyzed to understand the influence of personality types in a cross-functional set up. Finally, this paper focuses on rationalization of personality types of civil engineers for improved team performance in construction organizations. Several studies have shown that personality traits correlate with the performance of individuals in a team (Carr et al. 2002; Johnson & Singh 1998). Today’s civil engineers are a diverse group of people drawn from a broad range of global socio-cultural backgrounds. Their performance in a team pivots around their behavioral styles. Moreover, the chances of successful team work increases as the individuals understand each other. The performance of the organization is further dependant on the capacity of management to capitalize on the personality traits and types of each and every team member (Culp & Smith 2001). Katherine Briggs started research on typology in 1917 and enhanced Carl Jung’s theory on personality. Along with her daughter, Isabel Myers, Katherine Briggs developed a tool, popularly known as the Myers Briggs Type Indicator (MBTI), around the time of World War II. The MBTI enabled to distinguish four personality preferences and 16 personality types. MBTI is a popular modern psychological tool, with more than 2,000,000 people undertaking the test each year (Culp & Smith 2001). Federal and other agencies in USA have widely embraced MBTI as a tool

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for personality measurement in an attempt to improve their management performance (Rutzick 2007). MBTI measures four personality preferences, each subdivided into two personality traits as outlined briefly in the following section. 1. Interaction Preference reveals the personality traits of Extraversion (E) or Introversion (I). 2. Information Gathering Preference is described by personality traits of Sensing (S) or Intuition (N). 3. Decision-Making Preference characterizes what a person relies on in making decisions, either logic i.e., Thinking (T) or emotion, i.e., Feeling (F). 4. Living Preference relates to how people handle issues in their life—whether they are Perceptive (P) or dismissive and Judgmental (J). The combinations of information-gathering and decision-making preferences are known as the problem solving style. Likewise, the combinations of interaction and living preferences are defined as the environmental style. The overall combination of the styles gives the personality type. Thus, there are four levels of personality. For example, a person can be of the extroverted, intuitive, thinking, and perceptive (ENTP) type. Combinations of four personality preferences (eight traits) generate 16 unique personality types (Keirsey & Bates 1984; Johnson & Singh 1998; Culp & Smith 2001; and Hirsh & Kise 2006). 2

STUDIES ON PERSONALITY TYPES OF CIVIL ENGINEERS

Published research on the use of MBTI, or any other personality tests, has been far and few between for construction and civil engineers, although it is emerging

Table 1. Distribution of personality traits of various professions (source: Culp & Smith 2001). Group Engineers Life Insurance Agents Basketball officials Management Consultants Human-resources personnel Business Manager U.S. Population

Introvert Sensors Thinkers Judgers (I), % (S), % (T), % (J), % 62 26

54 83

75 63

67 71

34

97

42

33

Not 81 available 62 59

41

38

61

61

55

76

46.7

75

50

73

40

54

slowly. Research in this area can be more or less narrowed to Carr et al. (2002), O’Brien et al. (1998), Johnson & Singh (1998), Rosati (1998), Raymond & Hill (1988), and Varvel et al. (2004). A study carried out by Macdaid et al. (1986) surveyed more than 60,000 individuals of different professions, with engineers among them. Although their study did not specify the engineering disciplines, the results indicated that engineers are generally of the ISTJ type. Rosati (1998), at a Canadian university, found that the majority of the best engineering students who were able to graduate in a minimum time of four years were ISTJ type. Another similar study revealed that ISTJ types were quite prominent among undergraduate engineering students at a US university (O’Brien et al. 1998). Varvel et al. (2004) studied 193 senior design engineering students and found that the majority of the graduating students were ISTJ type. Raymond & Hill (1988) found that engineers, in general had predominant traits of I-S-T-J, but that certain disciplines, such as mining, even had E-S-F-P. He did not survey civil engineers. Culp & Smith (2001) surveyed engineers employed by consulting engineering firms in USA, and also found that they were predominantly of the ISTJ type. Table 1 shows the comparative distribution of personality traits of engineers against other selected professions as well as against the general US population. The table has some interesting data that serves to explain the behavior of members of different professions.

3

Table 2. SPA.

Distribution of individual personality types at Civil Engineers at SPA (Johnson & Singh (1998)

Type

Construction %

Design %

Total %

Engineers surveyed by Cub & Smith (2001) %

ENFJ ENFP ENTJ ENTP ESFJ ESFP ESTJ ESTP INFJ INFP INTJ INTP ISFJ ISFP ISTJ ISTP

12% 4% 0% 0% 8% 8% 4% 4% 8% 0% 0% 0% 4% 20% 16% 12%

0% 17% 0% 0% 8% 0% 8% 0% 0% 8% 0% 8% 25% 25% 0% 0%

8% 8% 0% 0% 8% 5% 5% 3% 5% 3% 0% 3% 11% 22% 11% 8%

2% 4% 7% 5% 4% 1% 8% 5% 2% 5% 14% 6% 5% 2% 23% 6%

Figure 1.

Overall personality traits of civil engineers at SPA.

Figure 2. at SPA.

Overall personality types of civil engineers

RESULTS OF THE SPA STUDIES

Johnson & Singh (1998) surveyed 65 civil engineers at State Public Agency (SPA) in Hawaii. Among them 43 were in the construction department and 22 in

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the design department. Forty-eight out of 65 civil engineers (74% of population) returned the survey, composed of 31 construction engineers (72% of construction population) and 17 design engineers (77% of design population). The results of personality types are summarized in Table 2. The overall distribution of personality traits and personality types of all civil engineers at SPA are illustrated respectively in Figure 1 and Figure 2.

4 4.1

ANALYSIS AND DISCUSSION Design engineers

Design engineers had predominant traits of introversion (I), sensing (S), feeling (F) and judging (J). As shown in Table 2, 66 % of the design engineers were introverted, 66% sensors, 83% feelers, and 41% were judgers. ‘Thinking’ was the weakest trait, and only 16% of the design engineers demonstrated such a style. It could be reasoned that ‘feeling’ is required to imagine original designs, which is why design engineers are naturally high in it. ‘Introversion’ is necessary to think quietly, while ‘sensing’ helps to collect factual design data rather than relying on intuition. Finally, being ‘judgmental’ helps put a stop to design perceptions since at the end of the day, standards and codes have to be fulfilled, and imaginations must come to a stop so that a certain design can be produced within a stipulated time. As a group, design engineers displayed I-S-F-J traits. However, ISFJ and ISFP were the two main personality types among the design engineers. Seven personality types were observed in the design department. Among them, 50% of the design engineers were either of the ISFJ or ISFP type, both types being equally predominant (25% each). The second largest personality type was ENFP at 17%. The remaining minorities were ESFJ, ESTJ, INFP, and INTP, all distributed at 8% each. Out of 16 possible personality types, 9 types— ENFJ, ENTJ, ENTP, ESFP, ESTP, INFJ, INTJ, ISTJ, and ISTP—were missing, which is not astounding given the small population and sample size. These personalities carry important qualities such as originality, interest in discovering causes and effects, ability to solve problems on the spot, and tactfulness in people interactions found missing among the design engineers. However, the distribution of personality types provides useful insights into the organization. The absence of ISTP types indicates that there are no ‘mechanically oriented’ personalities interested in finding causes and effects, an important aspect in engineering design teams. The absence of ESTP indicates the lack of individuals good at solving problems. Similarly, the lack of ENFJ and INFJ types indicate the lack of individuals who strive for originality,

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defend their design principles, and deal with others tactfully. 4.2 Construction engineers Similar to design engineers, introversion (I), sensing (S), feeling (F) and judging (J) was the predominant trait among construction engineers. As shown in Table 2, 60% of the construction engineers were introverted, 76% were sensors, 52% judgers and 64% feelers. Intuition was the weakest trait, with only 24% of the construction engineers demonstrating such a preference. It can be reasoned that construction engineers are strong on ‘sensing’ because they need to see construction in place before they can exercise project controls. They are largely introverted because that is part of the overall engineering personality, in general, as nurtured during their education. Their tendency towards ‘feeling’ might come from their necessity to understand the many different people they meet on the job. As a group, like the design engineers, construction engineers displayed I-S-F-J traits. Construction engineers exhibited only one predominant personality type, ISFP, with 20% being this type. ISTJ was the second largest personality type with 16%. ENFJ and ISTP were the third largest at 12%. Among 16 possible personality types, only ENTJ, ENTP, INFP, INTJ, and INTP were not found among construction engineers, but again, this can be attributed to the small sample size. Among the construction engineers, personalities with scientific and idealistic pursuits were missing. This might explain why construction engineers tend to be practical and realistic rather than theoretical. The lack of personalities such as INTJ and INTP indicates that there are not enough thinkers and visionaries in the department with sufficient theoretical and scientific pursuits for thorough and logical followup. Similarly, the lack of INFP personalities means that enthusiasts and loyalists are missing, which is why the workplace could be a dull place, and which could also lead to a situation of individualistic activities focused on selfish behavior instead of enhancing institutional credibility. Both of these observations were confirmed through qualitative surveys from an earlier study (Singh 1997). 4.3

Similarities between the design and construction engineers

Despite their differences, design and construction engineers demonstrated remarkable similarities in their personality preferences. From Table 3, correlation of the mean value of bipolar scores of preferences between design and construction engineers was rb = 0.96, which is significant; the correlation on

Table 3. Distribution of environmental and problem solving styles of the civil engineers at SPA (source: Johnson & Singh 1998). Preference Style Environmental EJ EP IJ IP Problem Solving ST SF NT NF

Construction Engineers

Design Engineers

Total

n 7 4 9 9

n 2 2 5 5

n 9 6 14 14

10 10 1 6

1 8 1 3

11 18 2 9

Of the 16 possible personality types, only three types, ENTJ, ENTP and INTJ, were missing among the civil engineers at SPA as a whole. Since the NT’s are largely missing, SPA lacks visionary personalities among engineers. Owing to the lack of ENTJ’s, they don’t have engineers good at reasoning. The absence of INTJ’s implies they are without people with a scientific bent of mind. The lack of ENTP types implies that there are not adequate personalities who are capable of solving new and challenging problems. It is interesting to note that SPA being a public agency lacks confident executives of the ENTJ type who could be good at public speaking. That said, it must be remembered the sample size is small, which may lead to unsymmetric observations. 4.6

environmental style was particularly high at re = 0.91 (Johnson & Singh 1998). 4.4

Differences between the design and construction engineers

Statistical analyses indicated that the personality types of design and construction engineers are dissimilar. The main difference was in their problem solving style. The correlation of decision-making preference between them was low at rd = 0.64. See data in Table 3. Only 19% of designers were thinkers compared to nearly double (36%) of the same among construction engineers. From Table 2, the correlation coefficient of overall personality types between the two groups indicated that their correlation is low at rt = 0.47. Hence, as a group, there are differences between the construction and design engineers. This means that effort is required to make them work in teams together. 4.5

Civil engineers at SPA

Though, as a whole, SPA engineers displayed I-S-FJ traits, the predominant personality type was ISFP, with nearly every fifth engineer falling under this type. Sensing was most predominant in the organization as a whole, with more than 70% of the civil engineers possessing such a trait. This indicates that they rely mainly on real data and facts, which is an asset when high quality products are desired, either it be for producing detailed drawings by design engineers or converting those details into reality by construction engineers. Over 60% of the SPA engineers were introverted, preferring to focus their energies on the internal world of ideas and experience. On the decision-making preference, feeling was clearly dominant over thinking. Intuition (N) and Thinking (T), were the weakest preferences among SPA’s civil engineers.

Comparison with a study on consulting engineers

Though the Johnson & Singh (1998) study yielded ISFJ as an overall personality style, literature review revealed that, by and large, engineers are predominantly of the introverted, sensory, feeling, and judging (ISTJ) type (Culp & Smith 2001; O’Brian et al. 1998; Rosati 1998; Varvel, et al. 2004). So, the study on consulting engineers carried out by Culp & Smith (2001) was compared with the results at SPA’s civil engineers. Correlation and proportion tests were performed assuming that personality preferences and personality types follow the normal distribution of probability in populations. Using data from Table 1, the proportion test indicates that the claim (H0 : p1 = p2 ) that the civil engineers’ predominant personality traits include introversion (I) or judging (J) should not be rejected at a level of significance (α) up to 23%. However, the tests showed that the null hypothesis (H0 : p1 = p2 ) should be rejected for the personality traits for the categories of sensing (S) and thinking (T) for α less than 2.44%. Furthermore, and using the data in the last two columns of Table 2, tests indicated that the proportion of each type was equal at α = 5% except for ENFJ, ESFP, ISFP and INTJ. These differences occur predominantly around the trait ‘F.’ Correlation tests indicated that personality traits and types of the engineers between these two studies are poorly correlated. The correlation coefficient on personality traits was nearly zero at rp = −0.03, indicating that they don’t match. Next, the correlation on personality types also showed a poor matching at rt = −0.09. This means that there can be variances in results in different geographical locations of different sizes. The studies carried out by O’Brien (1998), Rosati (1998), and Varvel et al. (2004) on engineering students revealed that the first half of the personality type was

832

predominantly I-S, as found in this study (Table 2). This similarity can probably be explained by the professional orientation of engineers and engineering students, since their work largely demands the I-S orientation. However, O’Brien (1998), Rosati (1998), and Varvel et al. (2004) found the second half of the personality type to be T-J, versus the F-P found in the study. We can perhaps explain this difference as being due to cultural reasons, since the population mix in the mainland and Canada is different to that in Hawaii. Hence, it may be quite appropriate to comment that the first half of personality types is profession-dependent, while the second half is culture-dependent, though this could be another item for further study. 5

SIGNIFICANCE OF WORK AND FINDINGS

The outcome of this short case study is that it should no longer be assumed that the dominant personality type for construction and design engineers is ISTJ. It is clear from the outcomes of the Johnson & Singh (1998) study that around 20% of engineers are actually of the ISFP type in contrast to other studies that found ISTJ to be the predominant type. This is important knowledge for Human Resource professionals who need to ensure that they recruit the right people into the right jobs, with the right personalities that go with those jobs. Ideally, organizations may attempt to distribute equally all personality types in their spider graphs, so that all personalities are well balanced in the organization, such that the maximum benefits of each type can be cultivated. However, such a situation is rarely achievable. Moreover, by understanding the personality traits of individuals, we can now better anticipate their communication needs, which is a major key to improved project performance and success. Indeed, without understanding the behavior of individuals on projects, it becomes difficult to manage them, motivate them, or fulfill fundamental project objectives to ensure project success. Traditionally, civil engineers are traditionally expected to be technically competent and their individual personality types are not considered as important as their technical expertise. Since engineering education is highly standardized and the decision making is usually influenced by measurements and numbers, subjective notions of personality were considered to be greatly irrelevant or a great mystery for civil engineers (Fleetham & Griesmer 2006). However, this study has demonstrated that organizational behavior elements, such as engineering decision making preferences, is highly influenced by the composition of professionals’ personality types. For a balanced performance, it will be ideal to expect that all types of personalities and traits are present in any large organization.

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6

CONCLUSIONS

A lot can be learned of an organization and its people by studying the personality traits and types of its employees. The case study at SPA and comparison with other studies came up with the following conclusions: – The MBTI survey indicated that the predominant personality traits of civil engineers at SPA were introversion (I), sensing (S), feeling (F) and judging (J). However, the predominant personality type in SPA was ISFP, i.e., introversion, sensing, feeling and perceiving. Nearly every fifth engineer is of the ISFP type. – Design and construction engineers share similar personality traits; they correlate well on the environmental style. – The main difference between the design and construction engineers was the decision-making preference, where the percentage of construction engineers using the thinking style was nearly double the design engineers’. – The SPA’s design department lacks engineers who are tactful, interested in discovering causes and effects, or good at solving problems on the spot. – The results of SPAS personality type were at odds with other studies. Proportion tests indicated that while introversion (I) and judging (J) are predominant personality traits among civil engineers, there may be differences in and sensing (S) and thinking (T). – When carefully applied, MBTI personality tests may assist in construction and engineering management to achieve team effectiveness, and overall organizational productivity.

REFERENCES Carr, P.G., Garza, J.M. and Vorster, M.C. (2002) Relationship between personality traits and performance for engineering and architectural professionals providing design services. J. of Management in Engineering, 18(1), Oct. 2002. Culp, G. and Smith, A. (2001) Understanding psychological type to improve project team performance. J. of Management in Engineering, 17(1), Jan. 2001. Fleetham, C. and Griesmer S.K., (2006) Leveraging personality for business success. J. of Leadership and Management in Engineering 6(4) pp. 143–174. Hill, R.E. and Summers, T.L. (1988) Project Teams and Human Group, in David Cleland and William King (eds.) Project Management Handbook, Van Nostrand Reinhold Co., NY. Hirsh, S.K. and Kise J.A.G. (2006) Work it Out—Using Personality Type to improve team Performance. Davis-Black Publishing, Mountain View, CA.

Johnson, H.M. and Singh A. (1998) Personality of civil engineers. J. of Management in Engineering, 14(4), 45–56. Keirsey, D. and Bates, M. (1984) Please Understand Me: Character and Temperament Types, Prometheus Nemesis Book Company, Del Mar, CA. Macdaid, G.P., McCaulley, M.H. and Kainz, R.I. (1986) Atlas of type tables. Gainesville: Center for Application of Psychological Type. Myers, I.B., McCaulley, M.H., Quenk, N.L. and Hammer, A. L. (1998) MBTI Manual: A guide to the development and use of the Myers-Briggs type indicator. Consulting Psychologists Press, 3rd ed. O’Brien, T.P., Bernold, L.E. and Akroyd, D. (1998) MyersBriggs Type Indicator and academic achievement in engineering education. International J. of Engineering Education, 14(5), 311–315.

Rosati, P. (1998) Academic progress of Canadian engineering students in terms of MBTI personality type. International J. of Engineering Education, 14(5), 322–327. Rutzick, K. (2007) Personality Test, Government Executive, 39(9), 22–23. Singh, A. (1997), Analysis of General Management Questionnaire at Oahu District, HDOT, National Technical Information Service, Springfield, VA, May 1997, 54 pp. Varvel, T., Adams, S.G., Pridie, S.J. and Ulloa, B. (2004) Team effectiveness and individual Myers-Briggs personality dimensions. J. of Management in Engineering, 20(4), Oct. 2004.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

System service oriented cooperation – lessons for the construction industry D. Lunze & G. Girmscheid Institute for Construction Engineering and Management, ETH Zurich, Switzerland

ABSTRACT: Life-cycle oriented service provisions in the construction industry need the entire competence spectrum of the key firms and planners. In this regard, the question for construction enterprises is whether cooperation is a beneficial approach for system oriented construction services. A qualitative empirical study has been performed to capture the lessons learned from industries that have successfully implemented cooperation structures. The results of the qualitative empirical study are decision-making and configuration elements as well as the success factors of strategic system service oriented cooperations that occur across different industries and different production repetition types. Along the generic cooperation process the results have been conceptualized in a phenomenological explanatory model. The results in the phenomenological explanatory model have the potential to be transferred as structural elements in a constructivist actional generic-deductive configuration model in the form of a cooperative business model for life-cycle service provision in the construction industry.

1

INTRODUCTION

In German-speaking countries the majority of construction contracts are basically being awarded on the basis of the initial investment costs of buildings (Girmscheid 2007b). This focus disregards the lifecycle costs of a building; instead the focus centers only on design and construction costs (Girmscheid & Lunze 2008). In order to remedy this grievance, clients are demanding a paradigm shift that incorporates the entire life-cycle of a building into the analysis of economic efficiency (Girmscheid 2006). Together with the Swiss construction industry, the Chair of Construction Process and Enterprise Management at the Institute for Construction Engineering and Management at ETH Zurich is developing a cooperative business model for life-cycle service provision in response to the demand for a paradigm shift. In this model the competences that are necessary for life-cycle service provision are to be combined in a strategic cooperation. System service oriented cooperation (SSOC) has been discovered to be an effective approach for various, highly competitive industries, such as the automotive industry etc., to elude from pure price competition (Girmscheid 2005). The aim of this system service oriented cooperation (SSOC) is to release synergy potentials between cooperating companies that allow for customer oriented services and/or products with greater customer benefits. These customer oriented services/products enable companies to pursue a differentiation strategy (Porter 1980) against their competitors. The construction process in German-speaking countries is still highly fragmented with respect to the life-cycle phases and the different construction trades

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(Girmscheid 2005). It has not been possible to establish cooperation successfully within the construction industry. This applies to both dimensions: cooperation along the value creation chain and cooperation between different trades. The cooperative business model which is under development at the Institute for Construction Engineering and Management at ETH Zurich contributes toward overcoming the fragmentation of the construction process in both dimensions. The following approach to developing the cooperative life-cycle business model has been chosen: – Collecting the lessons learned from branches and companies which actively exercise SSOC to increase their success in their business environment. – Developing a cooperative life-cycle business model for construction firms and transforming and implementing the lessons learned from other industries into the specific business environment of the construction industry. This qualitative empirical study was performed to capture the lessons learned from industries that have successfully implemented SSOC structures into their branch structure.

2

STATE OF RESEARCH

In the context of business management research, interorganizational cooperation is an important mode of coordinating economic activity (Sydow & Möllering 2004). Cooperation is considered to be a hybrid variant between the alternatives hierarchy (make) and market (buy) (Sydow & Möllering 2004; Powell

1990; Williamson 1985). The research on interorganizational cooperation in both the Germanspeaking and international research community is broadly positioned. There are research works that cover generic aspects of inter-organizational cooperation with emphasis on formation, development and management of inter-organizational cooperation structures (Lorange & Roos 1993; Segil 1996; Sydow 1999; Backhaus & Piltz 1990; Friedli 2000). Particularly the SSOC. (Heinz 1996) is considered to be an essential approach for the cooperation in technological development projects (Håkansson 1989; Andressen 2006). Another focus of research in inter-organizational cooperation examines certain aspects of international and/or global strategic alliances (Contractor 2002; Lorange & Roos 1991). Whereas Brockmann (2007) contributes by revealing the success factors of international construction joint ventures. The branch-specific mechanisms for successful cooperation in certain industries have been evaluated amongst others in the automotive industry (Scholta 2005; Royer 2000; Schindele 1996), in the railway industry (Matthews 2000) and in the shipbuilding industry (Shigo et al. 2004). The construction industry differs considerably from these above mentioned industries, particularly in respect of its production repetition type (PRT) of producing unique products/buildings. None of the studies conducted on inter-organizational cooperation takes the different PRTs of created services/products into account. Therefore the aim of the presented qualitative empirical study was the multistage approximation of the production repetition type (PRT) in the analyzed industries to the production of unique products/buildings in the construction industry. The empirical case study was conducted to find out if there are criteria in reference to strategic decision-making, configuration elements and success factors of SSOCs which are dependent or independent of the PRT and across branches. If there are independencies, the criteria identified in the case study could be implemented in the specific condition of the construction industry.

3 3.1

RESEARCH METHODOLOGY Methodological frame

The hermeneutic science program (HSP) which has been implemented into construction management research by Girmscheid (2007a) is interpreting and constructing the socio-technical world. The interpretativist research approach which is derivated from the HSP, is applied to generate a phenomenological explanatory model for SSOCs. The epistemological foundation of the interpretativist research paradigm is the phenomenological hypothesis for interpreting

reality. Building on the findings of the phenomenological explanatory model, the constructivist research approach (Guba & Lincoln 1994) will be applied to develop and structure an actional generic-deductive configuration model in the form of a cooperative business model for life-cycle service provisions (Girmscheid 2007a). The qualitative empirical analysis presented in this paper represents the first step of the problem solving process developed on the basis of the stated methodological frame. 3.2

Elements of the study

The empirical study ‘‘Lessons-learned in SSOCs’’ was subdivided into a literature study and a multi-case study on the basis of semi-structured problem-centered interviews to achieve highest information density. The literature study (e.g. Bronder & Pritzl 1992) revealed the following structural elements for configuring an SSOC (Fig. 1): – Strategic decision-making criteria for SSOCs – Configuration of the operative structure of the SSOCs – Configuration elements to prevent opportunistic behavior in SSOCs – Success factors of SSOCs 3.3

Research design and research logic

The case study was conducted in the following branches with experience in SSOCs: automotive industry, rail vehicle industry and shipbuilding industry. With respect to their production repetition type (PRT), the branches were selected in order to incorporate the complete spectrum from large-scale serial production to the production of individual products. This approach was chosen to converge to the production of unique products/buildings in the construction industry. In addition to the PRT, the sampling was based on the following analogy criteria: transaction characteristics, project business, system service orientation and the scope of decision and configuration of the respective cooperation (Lunze & Girmscheid 2008). The companies in these branches were selected by means of best-practice criteria (market leadership, economic health, innovativeness and adaptability) (Peters & Waterman 1984). The key aspects that were identified in the literature study were used to create a raster to structure the interviews. According to Yin’s typology of four possible basic types of designs for case studies (Yin 1994), the embedded multiple case study design was identified as the appropriate research design for the scientific objectives. The SSOC within a specific branch with a particular production repetition type constituted one of multiple cases. The prescribed replication logic (Yin 1994) was applied by analyzing the interesting phenomena within their relevant branch-specific contexts.

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et

a

nd

mp co

o eti

m en iron

Scope of decision-making and configuration

t

Branch structure

Corporate structure

Configuration elements Success factors of system service oriented Kooperation (AC2)

Sales market ...

Project delivery Implementation of configuration focus

Decision-makingelements

M

ar k

Context of system service oriented cooperation

nv ne

Decision criteria For the selection of the cooperation partners (AC1)

Decision criteria In favour of a system service oriented cooperation (AC1)

Supply market

Mechanisms To prevent opportunistic behaviour (AC3)

...

Configuration and selection of the cooperation partners

Strategic decision Characteristics of transaction

Pr

Realization /management of the system service oriented cooperation

System service oriented cooperation od

uc

ti o

n

Life-cycle orientation

re pe

t i ti

on

typ

System products/ services

Pricing

... Procurement strategy

e

Figure 1. The system service oriented cooperation in the context of its scope of decision-making and configuration—mapping of the elements of the multi-case study.

By applying literal and theoretical replication (Yin 1994), similar and contrasting findings from the different cases are used to generate a holistic view of the alternative and/or complementary decision-making and configuration elements as well as success factors of SSOCs. The data was collected using semi-structured problem-centered interviews (Mayring 1999) and recorded and codified for the descriptive conceptualization to extract the phenomenological hypothesis by means of a cross-case. The scientific quality of the qualitative empirical analysis was ensured through internal and external validation by means of (Yin 1994; Mayring 1999; Girmscheid 2007a): – methodological procedure according to state-ofthe-art and state of research, – documentation of methodological procedures, – codification of the given statements, – testing and elimination of inconsistent statements, – comparable conditions during the interviews, – retrospective verification of the codified statements and the deductive conceptualized hypothesis by interviewees, – ensuring comparability and reproducibility by means of a case selection with adequate selection criteria.

4

RESULTS—PHENOMENOLOGICAL EXPLANATORY MODEL

A strategic system service oriented cooperation is formed, developed and managed in the following

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phases of the generic cooperation process (Bronder & Pritzl 1992): – Strategic decision – Configuration and selection of the cooperation partners – Realization respectively management of the strategic cooperation The cross-case revealed that the conceptualized interpretativist findings comply with the literal replication (Yin 1994) across all analyzed branches and production repetition types. The case selection across the complete range of PRTs enabled the generation of a holistic image of the similar and complementary decision-making and configuration elements as well as success factors of SSOCs. Consequently the findings can be transferred to branches producing unique products/services, such as the construction industry. The results are presented in a conceptualized phenomenological explanatory model along the generic cooperation process. 4.1

Strategic decision

The decision to initiate a strategic SSOC is usually made on the initiative of the system leader. The core competences of the system leader enable it to decide which economic and technical activities to perform in its enterprise (make). All other economic and technical activities needed to satisfy customers’ requirements for a system service are potential candidates for outsourcing. These services/products can be integrated into the system service by supply on the market (buy) or by cooperation (cooperate).

The following essential criteria are applied when opting for a system service oriented cooperation: – The possibility to release synergy potentials – The complexity as well as the interactivity and the integrativity of the requested (sub-)system service in reference to the overall system – The specificity of the requested (sub-)system service – The supply risk associated with the (sub-)system service System suppliers also make an intentional decision regarding the participation in an SSOC. For those suppliers it is the decision between: – offering standard services/products with standardized interfaces to the overall system to follow a cost leadership strategy in a pure price competition environment or – alternatively, involvement in (sub-)system services with undefined interfaces to the overall systems to follow a differentiation strategy. Due to the system complexity and the amount of interdependencies in terms of undefined system interfaces with the second mentioned option the need for development can only be satisfied within a cooperative, interactive, integrative transaction situation. 4.2 Selection of the cooperation partners The selection of the cooperation partners is also subject to the performance competition on the market. From the system leader’s point of view, the system suppliers are selected on the basis of a value benefit analysis. The weighted rating of the following criteria leads to the selection of the cooperation partners and the favorable overall package: – Quality – Delivery time, adherence to schedules – Experience in previous cooperations between system leader and system supplier (supplier rating) – Special (sub-)system know-how – Capacities, resources, performance potential – Specific technical solution concept development by the (sub-)system supplier – Price The weighting basically results from the motives behind opting for an SSOC (see chapter 4.1). In respective market segments the system suppliers also have the possibility to select between various system leaders by means of the following selection criteria: – Overall system competence – References – Reputation with regard to business conduct, paying habits and fairness – Technology and innovation leadership

The qualitative empirical study showed that only system leaders with a reputation for being cooperative in regard to all above mentioned criteria attract system suppliers that perform excellently in their market segment. 4.3

Realization/management of the strategic cooperation

The following configuration elements and success factors have been identified for the realization respectively management of an SSOC: Cooperation constitutive success factors: – Compatibility in the intention to cooperate – Sense-making of the cooperation as a means of obtaining the cooperation’s objectives – Understanding of mutual positions – Balanced benefits for all cooperation partners – Selection of the appropriate partners – Trust – Commitment Market-related success factors: – Competition in partner and solution selection – Limited dependencies among the partners – Enterprise size/financial strength of the cooperation partners – Intra-cooperative paying habits – Risk distribution according to risk influence and bearing capacity Hierarchy-related success factors: – Common understanding that the cooperative relationship between system leader and system suppliers is still a transaction between purchaser (principal) and contractor (agent) – Concentration of the corporate resources of the involved partners on the cooperation and its synergetic objectives – Transparent mutual review of the cooperation partners Process factors:

respectively

product-related

success

– Functional call for tenders for the (sub-) system services – Mutual openness for innovations in the (sub-) systems and the overall system – Capability to be innovative – Cooperative contribution according to the respective competences of the cooperation partners – Dependability – Transparent supervision of the mutual services – Quality – Flexibility of the cooperation partners to adapt to changing process-related conditions

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Leadership-related success factors: – Top management support – Personnel to maintain the cooperation relationship – Reliable decisions with short decision-making processes The following mechanisms (processes and measures) were identified to prevent opportunistic behavior within the cooperation as far as possible: Inter-organizational linkage of the objectives: – Formulation of the corporate objectives of the participating partners within their respective market and competitive environment – Educing of common strategic objectives for the system service oriented cooperation – Cost framework due to competition in the selection phase as well as internal open books for the cooperation works Trust building: Building trust reduces the social complexity within the cooperation thus avoiding opportunistic behavior (Brockmann 2007). Measures that increase trust are the mutual experience of the partners with each other, the implementation and acceptance of a transparent control system including an objective authority, and the reputation of the partners within their market and competitive environment. Inter-organizational transparency: Transparency between the partners avoids information asymmetries as a cause for bounded rationality. It is ensured on different instances of a cooperation: – Transparent communication within the SSOC – Transparency in the objectives of the SSOC and the motives of its partners – Process-related transparency with respect to disclosure of the service/production processes and emerging problems in the processes, with the aim of identifying common solution concepts directly Integrative cooperation management: A management is installed to maintain the cooperation relationship to ensure compliance with the cooperation-related interests of the partners. The management personnel should be fitted out as follows: – Sufficient decision-making and configuration authority – Adequate personnel and financial resources – Technological understanding of the (sub-) systems Competitive contribution by the partners to the cooperation is ensured by: – Prior to selection for the cooperation, and regularly during the cooperation, the system service contributions of the cooperation partners have to compete against external competitors. An alternative are open books

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– Mutual dependencies should be limited to a minimum – A competitive advantage should be generated by means of differentiation through innovations and continuous improvement Multi-dimensional customer orientation: – end customers to the overall system – system leader as the customer of the system suppliers Multi-dimensional customer orientation is realized by implementing a multi-stage requirements management which incorporates, on the one hand, the product requirements for the end customers (overall system) and, on the other hand, those service and process requirements between the system leader and the (sub-)system supplier in the cooperation.

5

CONCLUSIONS

The success of a strategic SSOC is the result of a cooperation development process. The mutual experience within the cooperation determines the direction of future cooperation endeavors. The aspects that are linked with the success factors of former cooperations therefore lead to future strategic decisions about cooperations. At the same time, the success of a cooperation depends also on the realization of the expectations stipulated in the selection criteria. The cooperation process is not stringently linear, but an interactive, iterative and therefore dynamic process that develops across multiple project experiences. The qualitative empirical study showed that decision-making and configuration elements as well as success factors for strategic SSOCs exist in different industries and different production repetition types. In regard to the applied criteria, the findings have the potential to be transferred in the specific context of the construction industry. In a further step, the findings will be merged under consideration of the specific conditions in the construction industry in a constructivist, actional, generic-deductive cooperative business model for life-cycle service provisions in the construction industry.

REFERENCES Andressen, T. 2006. System Sourcing—Erfolgspotenziale der Systembeschaffung—Management und Controlling von Kooperationen. Wiesbaden: Deutscher UniversitätsVerlag. Backhaus, K. & Piltz, K. 1990. Strategische Allianzen. Düsseldorf etc.: Verlagsgruppe Handelsblatt.

Brockmann, C. 2007. Erfolgsfaktoren von Internationalen Construction Joint Ventures in Südostasien. Zürich: Eigenverlag des IBB an der ETH Zürich. Bronder, C. & Pritzl, R. 1992. Developing strategic alliances: A conceptual framework for successful co-operation. European Management Journal 10(4): 412–421. Contractor, F.J. 2002. Cooperative strategies and alliances. Amsterdam: Pergamon. Friedli, T. 2000. Die Architektur von Kooperationen. St. Gallen. Girmscheid, G. 2005. Partnerschaften und Kooperationen in der Bauwirtschaft—Chance oder Irrweg? Bauingenieur 80(Nr.2): S.103–113. Girmscheid, G. 2006. Risikobasiertes probabilistisches LCNPV-Modell—Bewertung alternativer baulicher Lösungen. Bauingenieur 81(Nr. 9): S.394–405. Girmscheid, G. 2007a. Forschungsmethodik in den Baubetriebswissenschaften. Zürich: Eigenverlag des IBB an der ETH Zürich. Girmscheid, G. 2007b. Projektabwicklung in der Bauwirtschaft—Wege zur Win-Win-Situation für Auftraggeber und Auftragnehmer. Berlin: Springer. Girmscheid, G. & Lunze, D. 2008. Paradigmawechsel in der Bauwirtschaft—Lebenszyklusleistungen. Bauingenieur 82(Nr.2): S.87–97. Guba, E.G. & Lincoln, Y.S. (1994). Competing paradigms in qualitative research. In: Handbook of Qualitative Research. N.K. Denzin & Y.S. Lincoln. Sage, Thousand Oaks: 105–118, chapter 6. Håkansson, H. 1989. Corporate Technological Behaviour— Co-operation and Networks. London etc.: Routledge. Heinz, I. 1996. Die Entwicklung zum Systemanbieter auf neuen Märkten: Ein Beispiel für den fundamentalen Wandel von Grossunternehmen. St. Gallen. Lorange, P. & Roos, J. 1991. Why Some Strategic Alliances Succeed and Others Fail. Journal of Business Strategy 12(1): 25. Lorange, P. & Roos, J. 1993. Strategic alliances—Formation, implementation, and evolution. Cambridge, Mass. etc.: Blackwell. Lunze, D. & Girmscheid, G. 2008. Erfolgsfaktoren strategischer systemgeschäftlicher Kooperationen— Zwischenbericht Phase A: Qualitativ empirische Untersuchung (Multi-Case-Studie) zur Evaluation von Entscheidungs- und Gestaltungselementen sowie Erfolgsfaktoren strategischer systemgeschäftlicher

Kooperationen in baufremden Branchen. Zürich: Eigenverlag des IBB an der ETH Zürich. Matthews, R.A. 2000. A healthy supply industry needs cooperation. Railway Age 201(8): S.10. Mayring, P. 1999. Einführung in die qualitative Sozialforschung: Eine Anleitung zu qualitativem Denken. Weinheim: Beltz-Psychologie Verlags Union. Peters, T.J. & Waterman, R.H. 1984. In search of excellence: Lessons from America’s best-run companies. New York: Warner Books. Porter, M.E. 1980. Competitive strategy—techniques for analyzing industries and competitors. New York: Free Press. Powell, W.W. 1990. Neither Market nor Hierarchy: Network Forms of Organization. Research in Organizational Behavior 12: 295–336. Royer, S. 2000. Strategische Erfolgsfaktoren horizontaler kooperativer Wettbewerbsbeziehungen: Eine auf Fallstudien basierende erfolgsorientierte Analyse am Beispiel der Automobilindustrie. München: Rainer Hampp Verlag. Schindele, S. 1996. Entwicklungs- und Produktionsverbünde in der deutschen Automobil- und -zulieferindustrie unter Berücksichtigung des Systemgedankens. Aachen: Shaker. Scholta, C. 2005. Erfolgsfaktoren unternehmensübergreifender Kooperation am Beispiel der mittelständischen Automobilzulieferindustrie in Sachsen. Chemnitz: Institut für Print- und Medientechnik (Zentrale Vervielfältigung) der TU Chemnitz. Segil, L. 1996. Intelligent business alliances—How to profit using today’s most important strategic tool. New York: Times Business. Shigo, N., Erkens, E., et al. (2004). Die Entwicklung von integrierten logistischen Netzwerken im deutschen Schiffbau. In: Logistik Management—Prozesse, Systeme, Ausbildung. T. Spengler. Heidelberg, Physica-Verlag: 425–439. Sydow, J. 1999. Strategische Netzwerke—Evolution und Organisation. Wiesbaden: Gabler. Sydow, J. & Möllering, G. 2004. Produktion in Netzwerken: Make, Buy & Cooperate. München: Franz Vahlen. Williamson, O.E. 1985. The economic institutions of capitalism: firms, markets, relational contracting. New York: Free Press etc. Yin, R.K. 1994. Case study research design and methods. Thousand Oaks etc.: SAGE publications.

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Sustainability and energy conservation

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Building environmental assessment tool S. Vilˇceková, E.K. Burdová & I. Šenitková Technical University of Kosice, Institute of Building and Environmental Engineering, Kosice, Slovakia

ABSTRACT: In recent years the building environmental assessment and rating systems are in process of development in East Block countries. The existing systems and tools are the base of development for new system available in Slovak conditions. The aim of building environmental assessment is a sustainable building design, which demands the cooperation among civil engineers, architects, environmentalists and other experts from different areas of building environmental assessment. Building environmental assessment methods have emerged as a widely adopted way to evaluate the performance of buildings across a broad range of environmental considerations. Eight models used world wide in relation to environmental assessment of buildings, were compared on the basis of their covered. The summary of sustainable building assessment systems will be presented in this paper. The proposal of building environmental assessment system applicable in Slovak conditions and method of indicators evaluation are also presented in the paper.

1 1.1

BUILDING ENVIRONMENTAL ASSESSMENT Characterization

In the past decade, building environmental assessment methods or systems have been developed and used in different countries for evaluating the sustainable and environmental performance. Building environmental assessment and certification systems are intended to foster more sustainable building design, construction and operations by promoting and making possible a better integration of environmental concerns with cost and other traditional decision criteria. Different building assessment systems approach this task from somewhat different perspectives, but they have certain elements in common. Most, if not all, deal in one way or another with site selection criteria, the efficient use of energy and water resources during building operations, waste management during construction and operations, indoor environmental quality, demands for transportation services, and the selection of environmentally preferable materials. The assessment of building environmental performance covers a wide range of issues and may involve not only a number of environmental, but also economical, social and cultural factors. Building environmental assessment is a specific complex of proceedings oriented to systematic and objective evaluation of building performance. These processes lead to design, construction and operation of buildings with respect to criteria of sustainable development. Building environmental assessment is not only tool of control, but also tool of sustainable building design. The assessment systems are

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based on the building’s life cycle: pre-design, new buildings, existing buildings, and renovation. Since 1990s building environmental assessment systems, methods and tools are developed for evaluation of building performance in many countries. Different building environmental assessment systems and methods have the same or similar issues of evaluation. Most of them deal with project management, site selection, energy performance, waste and water management, indoor environmental quality, environmentally preferable materials, etc. These comprehensive assessments of buildings require multidisciplinary and multi-criterion approach. The purposes of building assessment from environmental aspects are the determination of real buildings state from viewpoint of safety and reliability, the possibility of buildings comparison, the find of environmental buildings potential and the proposal of improvements resulting in sustainable buildings. Recently the building environmental assessment is also discussed theme in Slovakia at which time such assessment system is in process of development. The systems and tools used in many countries are the base of new system development for Slovak conditions (Senitkova 2004; Vilcekova 2007).

1.2

State of the art

The most significant building environmental assessment systems used over the world are BREEAM, Green Globes, LEED, SBTool, CASBEE, HK-BEAM, NABERS, LEnSE, etc. Building Research Establishment’s Environmental Assessment Method (BREEAM) as the building environmental assessment method was developed in the

United Kingdom in 1990. Ecohomes was a version of BREEAM for homes. It provided an authoritative rating for new, converted or renovated homes, and covers houses, flats and apartments. In April 2007 the Code for Sustainable Homes replaced Ecohomes for the assessment of new housing in England. The Code became operational in April 2007 in England, and having a Code rating for new build homes mandatory, from 1st May 2008. The Code for sustainable homes covers 9 categories of sustainable design including: energy and CO2 emissions; water; materials; surface water run-off; waste; pollution; health and wellbeing; management and ecology (BREEAM 2008). To monitor and assess green buildings in Canada, a rating and assessment system called Green Globes was established. The Green Globes system emerged from more than eleven years of research with inputs from a range of experts and international organizations. The origin of the Green Globes system lies in the Building Research Establishment’s Environmental Assessment Method (BREEM) which was published in 1996 by the Canadian Standards Association. In the year 2000, the system became an online assessment and rating tool under the name Green Globes for Existing Buildings. In the same year efforts to develop the system for the Design of New Buildings also began. Today Green Globes is used in Canada and the United States. In Canada, the Green Globes version for Existing Buildings is owned and operated by Building Owners and Managers Association of Canada (BOMA) under the brand name ‘‘Go Green’’. All other Green Globes products are owned and operated by ECD Energy and Environment Canada. In the United States, the Green Building Initiative (GBI) owns the license to promote and develop Green Globes. In 2005 GBI became the first green building organizations to be accredited as a standards developer by the American National Standards Institute (ANSI). The Green Globes program is an online system that is based on a questionnaire. A report based on the questionnaire is automatically generated which provides the Green Globes rating, achievements and also recommendations. The certification is achieved through a third-party verification undertaken by regional verifiers. The Green Globes system is applicable to all types of buildings of any size including small and large office buildings, multifamily housing structures, schools, universities and libraries. The Green Globes system has assessment tools for: Design for New Buildings or Retrofits, Management and Operations of existing buildings, Green Globes Building Emergency Management Assessment (BEMA), Building Intelligence and Fit-Up (Green Globes 2008). LEED (Leadership in Energy and Environmental Design) was developed and piloted in the U.S. in 1998 as a consensus-based building rating system based on the use of existing building technology. The development of LEED has been through the U.S. Green

Building Council member committees. LEED is a third-party certification program and the nationally accepted benchmark for the design, construction and operation of high performance green buildings. LEED gives building owners and operators the tools they need to have an immediate and measurable impact on their buildings’ performance. LEED promotes a whole-building approach to sustainability by recognizing performance in five key areas of human and environmental health: sustainable site development, water savings, energy efficiency, materials selection and indoor environmental quality. LEED certification provides independent, third-party verification that a building project meets the highest green building and performance measures. All certified projects receive a LEED plaque, which is the nationally recognized symbol demonstrating that a building is environmentally responsible, profitable and a healthy place to live and work (LEED, 2008). SBTool is the software implementation of the Green Building Challenge (GBC) assessment method that has been under development since 1996 by a group of more than a dozen teams. The GBC process was launched by Natural Resources Canada, but responsibility was handed over to the International Initiative for a Sustainable Built Environment (iiSBE) in 2002 (SBTool 2008). Comprehensive Assessment System for Building Environmental Efficiency (CASBEE) was developed in Japan, beginning in 2001. The family of assessment tools is based on the building’s life cycle: pre-design, new construction, existing buildings, and renovation. CASBEE presents a new concept for assessment that distinguishes environmental load from quality of building performance. By relating these two factors, CASBEE results are presented as a measure of eco-efficiency or BEE (Building Environmental Efficiency) (CASBEE 2008). HK-BEAM is the Hong Kong industry’s initiative to measure, improves, certify and label the whole-life environmental sustainability of buildings. HK-BEAM is a comprehensive standard and supporting process covering all building types including residential, commercial institutional buildings and mixed use complexes, both new and existing. HK-BEAM means by which to benchmark and improve performance in the planning, design, construction, commissioning, operation and management of buildings. HK-BEAM was adopted since 1996. Aims of the HK-BEAM is stimulate demand for more sustainable buildings in Hong Kong, giving recognition for improved performance and minimizing false claims; provide a common set of performance standards that can be pursued by developers, designers, architects, engineers, contractors and operators; reduce the environmental impacts of buildings throughout the planning, design, construction, management and demolition life cycle; and increase awareness in the building community, and ensure

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that environmental considerations are integrated right from the start rather than retrospectively (HK-BEAM 2007). National Australian Building Environmental Rating System (NABERS) project commenced in April 2001 is being designed to assess many types of new and existing buildings—particularly commercial and residential—and to enable the building owner or operator to undertake the rating annually with or without the need to hire independent assessors (NABERS 2008). Methodology Development towards a Label for Environmental, Social and Economic Buildings (LEnSE) is project of the 6th Framework Program that suggests a common European methodology for assessment and/or labeling of environmental, social and economic impacts of buildings. LEnSE is a European research project that responds to the growing need in Europe for assessing a building’s sustainability performance. The project draws on the existing knowledge available in Europe on building assessment methodologies. The main objective of LEnSE was to develop a methodology for the assessment of the sustainability performance of existing, new and renovated buildings, which is broadly accepted by the European stakeholders involved in sustainable construction. This methodology will allow for future labeling of buildings, in analogy with the Energy Performance Directive (Lupisek 2008).

2 2.1

BUILDING ENVIRONMENTAL ASSESSMENT SYSTEM PROPOSAL The base of new system development

The fields and indicators of building environmental assessments are proposed on the bases of available information analysis from particular fields of building environmental assessment and also on the base of our experimental experiences. The base of development system is mainly SBTool, also LEED, CASBEE and HK-BEAM. The proposed indicators respect Slovak standards and rules. The proposed fields of building environmental assessment of applicable in Slovak conditions are building site and project management, building constructions, indoor environment, energy performance, water management and waste management.

Table 1.

Field and indicator A A1 A1.1 A1.2 A1.3 A1.4 A1.5 A1.6 A1.7 A1.8 A2 A2.1 A2.2 A2.3 A3 A3.1 A3.2 A3.3 A3.4 A3.5 A3.6

Building site and project management

Site selection issues include transportation and travel distances for building occupants, impacts on wildlife corridors and hydrology, energy supply and distribution limitations. Decisions made during site selection and planning impact on the surrounding natural habitat, architectural design integration, building energy consumption, occupant comfort and occupant

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Building site and project management Site selection Selection of ecologically valuable or sensitive land Selection of land vulnerable to flooding Selection of land close to water endangered contamination Selection of brownfield lands Distance to commercial and cultural facilities Distance to public green space Distance to engineering networks Distance to road-traffic infrastructure Project Planning Assessment of renewable feasibility Preparation of environmental impact assessment report Applicable orientation to maximize passive solar potential Project development Development density Possibility change building purpose Relationship of design with existing streetscapes Policies governing use of private vehicles Use of trees for solar shading and sequestration of CO2 Maintenance or development of wildlife corridors

productivity. Maximize sustainability opportunities in the site selection and site planning process it is need to integrate site issues into the pre-design process. Some opportunities continue through design development and to a more limited extent, through facility and landscape construction. Integrate the building with the site in a manner that minimizes the impact on natural resources, while maximizing human comfort and social connections. The development footprint should enhance the existing biodiversity and ecology of the site by strengthening the existing natural site patterns and making connections to the surrounding site context. Table 1 summarizes the proposed sub-fields and indicators of building site and project management field. Indicator from sub-field ‘‘Site selection’’ is related to selection of land vulnerable to flooding. This indicator introduced in the Table 2 is assessed according to height above 100-year flood plain. 2.3

2.2

Building site and project management.

Building constructions

The quality of the built environment too affects its inhabitants in many ways and is dependent not only on the architectural form and specification, but also on the quality and nature of materials used, the care taken in construction, the quality of building services design and components, and the timely and effective maintenance of the building fabric and support systems. The risks of diseases are also increased when the dwelling’s

Table 2.

Selection of land vulnerable to flooding.

Table 4.

Eco-labeling.

A1.2

Selection of land vulnerable to flooding

B1.10

Eco-labeling

Purpose

To discourage the selection of land for building where there is a substantial risk that the site may be flooded. Height above 100-year flood plain as defined in official documentation or assessment by component authorities.

Purpose

To encourage production and consumption of products with less adverse effects on the environment, to inform consumers about the environmental characteristics of products. Use of environmentally friendly building products

Indicator

The height of the minimum elevation of the site above the elevation of the 100-year flood is: Negative 1.0 m Acceptable 1.3 m Good 2.0 m Best 2.5 m Table 3.

Indicator

Score −1 0 3 5

Negative Acceptable Good Best

The percentage, by weight, of building environmentally friendly product is: 3% 15% 51% 75%

Score −1 0 3 5

Building constructions.

B

Building construction

Table 5.

B1 B1.1 B1.2 B1.3 B1.4 B1.5

Materials Certified building products Use of cement substitutes in concrete Use of materials that are locally produced Use of recycled materials Non-renewable primary energy embodied in construction materials Radioactivity building materials Creation hazardous substances during production building materials Selection low—emission building materials Constructions limiting migration pollutions between occupations rooms Eco-labeling LCA Dismountable, reuse and recycling LCA impact on cost Analysis LCA Renewable

C

Indoor environment

C1 C2 C3 C4 C5

Thermal comfort in heating season Thermal comfort in cooling season Ventilation Noise attenuation through the exterior envelope Noise isolation between primary occupancy areas Daylighting Shading and blind Artificial lighting Interior materials Pollutant migration between occupancies

B1.6 B1.7 B1.8 B1.9 B1.10 B2 B2.1 B2.2 B2.3 B2.4

barriers against insect and rodent vectors are inadequate or poorly maintained. Environmentally friendly building materials and constructions are aimed to reduce of energy and material flows during whole building life cycle. The evaluation is focused on assessment of consumption and depletion of material resources, especially nonrenewable resources, to minimize life-cycle impact of materials on the environment and enhance indoor environmental quality and also focused on evaluation of energy flows through building constructions. Table 3 summarizes the proposed sub-fields and indicators of building constructions field. In Table 4, there is presented indicator ‘‘Ecolabeling’’ from sub-field ‘‘Materials’’. The evaluation of this indicator is according to the percentage, by weight, of building environmentally friendly products that are inbuilt in rating building.

C6 C7 C8 C9 C10

2.4

Indoor environment.

Indoor environment

In the last several years, monitoring of indoor environmental quality has indicated that the air within buildings can be more seriously polluted than the outdoor air in even the largest and most industrialized cities. Indoor pollution sources that release gases or particles into the air are the primary cause of indoor air quality problems in residential and non-residential buildings. Inadequate ventilation can increase indoor pollutant levels by not bringing in enough outdoor air to dilute emissions from indoor sources and by not carrying indoor air pollutants out of the building. High temperature and humidity levels can also increase concentrations of some pollutants. The relative importance of any single source depends on how much of a given pollutant it emits and how hazardous those emissions are. In some cases, factors, such as how old the source is and whether it is properly maintained is significant. Some sources, such as building materials, furnishings, and household products like air fresheners, release pollutants more or less continuously. Other sources, related to activities carried out in the buildings, release pollutants intermittently. These include smoking, the

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Table 6.

Thermal comfort in heating season.

Table 8.

Energy for heating.

C1

Thermal comfort in heating season

D1.1

Energy for heating

Purpose

To ensure thermal comfort in heating season. Designed value of operative temperature is in accordance with requirements of relevant standards (EN 15251:2007).

Purpose Indicator

To determine energy for heating. Class of energy for heating according standards related to energy performance of buildings. Energy for heating is in lower class as C. Energy for heating is in class C. Energy for heating is in class B. Energy for heating is in class A.

Indicator

Negative Acceptable Good Best

Table 7.

In 95% of building volume the opera tive temperature is: θ0 < 19◦ C 19 ≤ θ0 < 20◦ C 20 ≤ θ0 < 21◦ C θ0 ≥ 21◦ C

Negative Acceptable Good Best

Score −1 0 3 5

Table 9.

Energy performance.

D

Energy

D1 D1.1 D1.2 D1.3 D1.4 D1.5 D2 D2.1 D2.2 D2.3

Operation energy Energy for heating Energy for domestic hot water Energy for mechanic ventilation and cooling Energy for lighting Energy for appliances Active systems on using renewable energy sources Solar system Photovoltaic technology Heat recuperation

D3 D3.1 D3.2

Maintains energy Energy management Operation and maintains

use of unventilated or malfunctioning stoves, furnaces, or space heaters, the use of cleaning products and so on. High pollutant concentrations can remain in the air for long periods after some of these activities. Health effects from indoor air pollutants may be experienced soon after exposure or, possibly, years later. Immediate effects may show up after a single exposure or repeated exposures. The investigations conducted at Institute of Building and Environmental Engineering focused on selected chemicals occurrence in nonresidential buildings of various types have shown that the office buildings have the highest level of pollution related to interior materials. These results were included also in development of this building environmental system for office buildings. Table 5 summarizes the proposed indicators of indoor environment field (Vilcekova 2008). Next indicator presented in Table 6 is evaluated according to requirements of European standard (EN 15251:2007).

Score −1 0 3 5

Water management.

E

Water

E1 E2 E3 E4

Reduction and regulation water flow in water systems Surface water run-off Drinking water supply Using filtration ‘‘grey water’’

Table 10. E2

Surface water run-off. Surface water run-off

To ensure that surface water is managed within site boundaries and is re-injected into the aquifer. Indicator The quality of a surface water management plan. Score Negative A credible general plan has not been developed for the management of surface water. −1 Acceptable A general plan has been developed for the man agreement of surface water and its percolation into the ground within site boundaries, including at least 80% of natural surface water courses, paved and landscaped areas. 0 Good A detailed plan has been developed for the management of surface water and its percolation into the ground within site boundaries, including at least 90% of natural surface water courses, paved and landscaped areas. 3 Best A detailed plan has been developed for the management of surface water and its percolation into the ground within site boundaries, including 100% of areas. 5

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Purpose

2.5

Energy performance

Goal is reduce total building energy consumption and peak electrical demand, reduce air pollution, contributions to global warming, and ozone depletion caused by energy production, slow depletion of fossil fuel reserves and achieve energy cost and related

savings due to upgrades to infrastructure. In the last few years, there is devoted a considerable attention to energy performance of buildings in Slovakia. Energy performance of buildings is related to energy consumption for heating, cooling, domestic hot water, ventilation and lighting. In the Table 7 are presented the sub-fields and their indicators about energy performance (Vilcekova 2008). In Table 8, there is presented indicator ‘‘Energy for heating’’. The evaluation of energy needs for heating is according to valid standards and rules (Act No. 555/2005). 2.6

Water management

Goal is preserve site watersheds and groundwater aquifers, conserve and reuse storm water, maintain appropriate level of water quality on the site and in the building, reduce potable water consumption and reduce off-site treatment of wastewater. Table 9 summarizes the proposed indicators of water management field. Table 11.

Waste management.

F

Waste

F1 F1.1

Solid waste Measures to minimize solid waste resulting from building construction Measures to minimize solid waste resulting from building operations Composting Emission Measures to minimize gas waste from building construction Measures to minimize gas waste from building operation

F1.2 F1.3 F3 F3.1 F3.2

Table 12. Measures to minimize solid waste resulting from building operations. F 1.2 Purpose

Indicator

Negative Acceptable Good Best

Measures to minimize solid waste resulting from building operations To minimize the amount of waste off the site by encouraging the development and implementation of a construction waste management program, with sorting, re-using and recycling measures. The development of a credible construction waste management plan. The percentage, by weight, of construction waste to be re-used or re-cycled, as predicted in the construction was management plant, is: Score 3% −1 15% 0 51% 3 75% 5

In Table 10, there is presented indicator ‘‘Surface water run-off’’. 2.7

Waste management

Goal is minimize waste generated from construction, renovation, and demolition of buildings, minimize waste generated during building occupancy and encourage better management of waste. Table 11 summarizes the proposed sub-fields and indicators of waste management field. In Table 12, there is presented indicator ‘‘Measures to minimize solid waste resulting from building operations’’. The evaluation of this indicator is according to development of a credible construction waste management plan.

3

CONCLUSIONS

In this paper is introduced the proposal of building environmental assessment system applicable in Slovak conditions. The base of assessment development is systems and methods used in many countries. The main building environmental assessment fields are building site and project management; building constructions; indoor environment; energy performance; water and waste management. There are presented one indicator from each field and method of its evaluation with respects to standards and acts valid in Slovakia.

ACKNOWLEDGEMENTS The authors are grateful to the Ministry of Education of Slovak Republic for financial support in frame of project NFP OPVaV with ITMS code 26220120018 and of the project VEGA 1/0585/09.

REFERENCES BREEAM. Code for Sustainable Homes. Technical Guide. Department for Communities and Local Government. April 2008. www. Communities.gov.uk Green Globes. Assessment and Rating System. Program Summary and Users Guide. Green Building Initiative, Oregon, 2008 LEED. Green Building Rating System for New Construction and Major Renovations. Version 2.2, Washington, DC, April 2008. SBTool. An Overview of SBTool, September 2008. CASBEE. CASBEE for New Construction —Technical Manual. Institute for Building Environment and Energy Conservation, March 2008. HK-BEAM. An environmental assessment for existing Office building. 2007. NABERS. NABERS Office Building Trial, June 2008.

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Lupíšek, A.: LEnSE —European Methodology of Buildings Assessment. In 10th Presfessional Conference of Postgraduate Students: Juniorstav 2008, Brno, Czech Republic, p. 4 Šenitková, I. & Vilˇceková, S. 2004. Indoor Air Quality Auditing, In Proc. of XXXII IAHS International Congress: Sustainability of the Housing Projects, Trento, Italy, p. 6. Šenitková, I. & Vilˇceková, S. 2004. Indoor Thermal Comfort Auditing, In Proc. of XXXII IAHS International Congress: Sustainability of the Housing Projects, Trento, Italy, p. 6. Vilˇceková, S., Burdová, E. & Šenitková, I. 2007. Sustainable building assessment systems summary. In Interrupted operation. Journal of Lviv National University, Lviv Ukraine, No 600, p. 559–567.

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Vilˇceková, S. & Burdová, E. 2008. Indoor environmental quality in assessment system of buildings. In Proc. of 11th International Conference on Indoor Air Quality and Climate: Indoor Air 2008, Copenhagen, Denmark, p. 8. Vilˇceková, S. & Burdová, E. 2008. Energy performance and environmental assessment of buildings in Slovakia. In ENERGODOM 2008: Low-energy buildings, Cracow, Poland, 443–448. Vilˇceková, S., Krídlová Burdová, E. & Šenitková, I. 2008. Indoor environmental quality assessment. In Selected Scientific Papers: Journal of Civil Engineering, Košice, Slovakia, vol. 3, No. 2, p. 25–33.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Building passive design and hotel energy efficiency B. Su School of Architecture, UNITEC Institute of Technology, Auckland New Zealand School of Architecture, Shenyang Jianzhu University, Shenyang, China

Q. Wang School of Architecture, Shenyang Jianzhu University, Shenyang, China

ABSTRACT: This pilot study seeks to identify the relationships between real energy consumption data and building design data and introduces a method to use real energy consumption data of a large hotel to calculate the extra energy use related to winter indoor thermal conditions, which roughly represents the space heating energy. This study not only presents and identifies relationships between the increasing or decreasing trend in space heating energy and the increase of building design data of the sample hotels, but also establishes the start point and feasibility for further study with a large number of sample hotels to identify, in more detail, the quantitative relationships between building design data and space heating energy data for further developing passive design guides for hotel energy efficiency.

The first and best place to consider building energy efficiency is during the design of the building, not when the building has been completed and is in operation. Previous study suggests that the better design of new buildings would result in a 40–75% reduction in their energy consumption to compare with 2000 levels. Large hotels are big energy users in a city. To minimize the influence of differences in hotels’ facilities, the study randomly collected the monthly real energy consumption data, monthly occupancy and building design data of a number of 4–5 stars large hotel (over 100 guestrooms) in Auckland city of New Zealand in a temperate climate with mild winter and in Shenyang city of China with a cold climate. Figure 1 and Figure 2 show monthly minimum, maximum and mean temperatures of Auckland and Shenyang. Figure 3 shows monthly mean energy consumptions per room per day of Auckland sample hotels. Figure 4 shows monthly mean energy consumptions per unit volume (m3 ) of indoor space per day. To compare the energy used in the seven large hotels the mean energy used per room per day in Hotel 6 is lower than the other hotels but the mean energy used per m3 per night could be higher than the other hotels. Using different units to compare the energy used in different hotels may lead to different results. The question is what energy unit is appropriate to be used to present or compare energy consumptions or energy efficiency related to the different hotel building designs, especially to compare the energy used for internal space heating and cooling. The national and international energy surveys for

851

the hotel sector commonly use kWh/room/year as the units to present the energy used in the hotels.The mean energy data use per room in the national energy survey can be used to show the general profile of energy used in the hotel business. The mean energy used for space heating and cooling per unit of the volume actually used in the hotel (not including the volume of the vacant rooms) is more appropriately used for the comparison of building design. Max Tem. Temperature (deg.C)

INTRODUCTION

Mean Tem.

Min Tem.

30 25 20 15 10 5

Jan Feb Mar Apr May Jun Jul Aug Sep Oct Nov Dec

Figure 1.

Monthly temperatures in Auckland. Max. tem.

Temperature (deg.C)

1

30 25 20 15 10 5 0 -5 -10 -15 -20

Figure 2.

Mean tem.

Min. tem.

Jan Feb Mar Apr May Jun Jul Aug Sep Oct Nov Dec

Monthly temperatures in Shenyang.

150

Hotel 2

30% Ratio of heaing energy to winter energy

Hotel 1

Hotel 3 100

Hotel 4

50

Hotel 5

0

Hotel 6 Dec

Nov

Oct

Sep

Aug

Jul

Jun

May

Apr

Mar

Feb

Jan

Mean energy (kWh/room/day)

200

15% 10% 5% 0%

2

1

4

3

5

6

7

Hotels

Figure 5. Ratio of heating energy to winter energy of Auckland sample hotels.

Ratio of heating energy and winter energy

Hotel 1 Hotel 2 Hotel 3 Hotel 4 Hotel 5

80% 70% 60% 50% 40% 30% 20% 10% 0% 1

Hotel 6 Dec

Nov

Oct

Sep

Aug

Jul

Jun

May

Apr

Mar

Feb

Jan

Mean energy (kWh/m3/day)

20%

Hotel 7

Figure 3. Monthly mean energy consumptions per room per day of Auckland sample hotels.

0.8 0.7 0.6 0.5 0.4 0.3 0.2 0.1 0

25%

2

3

Hotel 7

4

5 6 Hotels

7

8

9

10

Figure 6. Ratio of heating energy to winter energy of Shenyang sample hotels.

Figure 4. Monthly mean energy consumptions per m3 per day of Auckland sample hotels.

During the winter, a hotel mainly uses its energy on water heating, space heating, refrigeration, cooking, lighting and other building services. Comparatively, the energy used for the indoor space heating is more closely related to the hotel building thermal performance and its indoor thermal conditions. For a hotel with a central air conditioning system, the space heating can be supplied by a boiler. The boiler also supplies the hot water for the whole hotel. It is difficult to identify how much energy is only used for space heating according to energy consumption data from the meters in the current central air conditioning system and it is time consuming and expensive to install meters on the existing central air conditioning systems to only record the actual energy used for the space heating. The study introduces a method to use real monthly energy consumption data of a hotel to calculate the extra energy consumption resulted from the impact of winter indoor thermal conditions of the sample hotel, which can be used to roughly present space heating energy and compare different designs of hotels for energy efficiency. The study uses the difference between mean daily energy usages per unit volume of occupied indoor space (kWh/m3 day) of a hotel in the winter months and the other months of the year as the basic and mean energy consumption unit, which mainly comprises space heating energy and other extra energy for hot water heating and all appliances, which are impacted by the winter indoor thermal conditions of a hotel. The smaller difference between mean daily usage in winter months and the other months can

roughly represent the better indoor space thermal conditions responded to the winter climate conditions. Figure 3 and Figure 4 show ratios mean space heating energy to winter energy of Auckland and Shenyang sample hotels. The study uses the following main architectural features as sample design data to investigate the relationships to winter indoor space heating energy: – – – –

Ratio of building surface to volume Ratio of total window to wall area Ratio of north (south) wall area to building volume Ratio of total north (south), east, west wall area to building volume – Ratio of total wall area to building volume – Building volume – Building height

2 2.1

DATA ANALYSIS Ratio of building surface to volume

The ratios of building surface to volume of sample hotels in Auckland and Shenyang are in the ranges of 0.09 to 0.24 and 0.10 to 0.15 respectively. An increasing trend in differences between mean daily energy usages per unit volume of occupied indoor space (kWh/m3 day, roughly present mean space heating energy) of Auckland and Shenyang sample hotels in the winter months and the other months of the year are both associated with increasing in ratios of building surface to volume (see Figs 7–8). A building with a low

852

0.2 0.18 0.16 0.14 0.12 0.1 0.08 0.06 0.04 0.02 0 0.08

0.1

0.12 0.14 0.16 0.18 0.2 0.22 Ratio of building surface to volume

0.11 0.12 0.13 0.14 0.15 Ratio of building surface to volume

0.16

0.18

Space heating energy Space heating energy

1.6 1.4 1.2 1 0.8 0.6 0.4 0.2 0 0.08

0.2 0.18 0.16 0.14 0.12 0.1 0.08 0.06 0.04 0.02 0 0.14

853

0.14

0.32

0.17 0.20 0.23 0.26 0.29 Ratio of window area to wall area

0.32

Figure 11. Ratio of window area to wall area and space heating energy of Auckland sample hotels. 1.6 1.4 1.2 1 0.8 0.6 0.4 0.2 0 0.1

Figure 9. Ratio of total wall area to building volume and space heating energy of Auckland sample hotels.

0.09 0.13 0.1 0.11 0.12 Ratio of total wall area to building volume

Figure 10. Ratio of total wall area to building volume and space heating energy of Shenyang sample hotels.

Space heating energy 0.06 0.08 0.1 0.12 0.14 0.16 Ratio of total wall area to building volume

Ratio of total window area to wall area

The ratios of total building window area to wall area of sample hotels in Auckland and Shenyang are in the ranges of 0.15 to 0.31 and 0.14 to 0.38 respectively. An increasing trend in mean space heating energy of Auckland and Shenyang sample hotels are both strongly associated with increasing in ratios of total window area to wall area (see Figs 11–12). The windows are commonly the weak elements for building thermal performance. For a large hotel design in Auckland and Shenyang to increase the ratio of total window area to wall area will increase space heating energy when the rest of the design data are unchanged. Generally a small ratio of total window area to wall

Space heating energy 0.1

Figure 8. Ratio of building surface to volume and space heating energy of Shenyang sample hotels. 0.2 0.18 0.16 0.14 0.12 0.1 0.08 0.06 0.04 0.02 0 0.04

2.2

0.24

Figure 7. Ratio of building surface to volume and space heating energy of Auckland sample hotels. 1.6 1.4 1.2 1 0.8 0.6 0.4 0.2 0 0.09

Auckland and Shenyang sample hotels are both associated with increasing in ratios of total wall area to building volume (see Figs. 9–10).

Space heating energy

Space heating energy

ratio of building surface to volume has a small external surface area per unit of indoor space from which to lose heat to the outdoors, and uses less energy for space heating, hot water and other appliances, which can be affected by indoor thermal conditions during the winter. The ratio of building surface to volume of the multi-storey residential building with the permanent heating should be 0.3 or less for saving the energy for space heating (Liu 2000). Normally the building with smaller ratio of building surface area to building volume uses less energy for space heating. Wall of a multi-storey hotel building is more important than roof to impact building thermal performance and indoor thermal conditions to compare with a house. The ratios of total wall area to building volume of sample hotels in Auckland and Shenyang are in the ranges of 0.06 to 0.17 and 0.09 to 0.14 respectively. An increasing trend in mean space heating energy of

0.15 0.2 0.25 0.3 0.35 Ratio of window area to wall area

0.4

Figure 12. Ratio of window area to wall area and space heating energy of Shenyang sample hotels.

area is good for saving the space heating energy in the Auckland and Shenyang hotels. 2.3

Ratio of total north (or south), east, west wall area to building volume

Space heating energy

Ratios of total North (or south), West and East wall areas to building volume of Auckland sample hotels and ratios of total north (or south), west and east wall areas, which can get direct solar radiation heat, to building volume of Shenyang sample hotels are in the ranges of 0.04 to 0.11 and 0.06 to 0.10 respectively. A decreasing trend in mean space heating energy of Auckland and Shenyang sample hotels are both associated with increasing in ratios (see Figs 13–14). During winter, the hotel, with the orientation facing to the equator and a bigger ratio, potentially receive more direct sun. The more area of walls is exposed to the sun; the less the energy is used for the space heating in the Auckland and Shenyang hotels. 0.2 0.18 0.16 0.14 0.12 0.1 0.08 0.06 0.04 0.02 0 0.04

2.4

The ratios of north (or south) wall area to building volume of sample hotels in Auckland and Shenyang are in the ranges of 0.015 to 0.053 and 0.02 to 0.28 respectively. A hotel with good orientation usually has a high ratio of north wall area to building volume. Good orientation should improve indoor thermal conditions and energy efficiency, but the increase of ratios of north (south) wall area to building volume is not associated with an decreasing trend of space heating energy of Auckland (Shenyang) sample hotels (see Figs 15–16). Windows of Auckland sample hotel are single-glazed and the glazed window areas of hotels are commonly low R-value area. The mean ratio of north-facing window area to north wall area and the mean ratio of south-facing window area to south wall area for Auckland and Shenyang sample hotels are higher than the ratios of east-facing and west-facing windows. The negative effect of increasing the ratio of north-facing window area (south-facing wall area) to north wall area (south-facing wall area) could be stronger than the positive effect of increasing the ratio of north wall area (south wall area) to building volume on winter indoor thermal conditions of Auckland (Shenyang) sample hotels. 2.5

0.05 0.06 0.07 0.08 0.09 0.10 0.11 0.12 Ratio of N, E, W wall area to building volume

Ratio of north (or south) wall area to building volume

Building volume

Building volume of sample hotels in Auckland and Shenyang are in the ranges of 51290 to 130530 m3

Space heating energy

Figure 13. Ratio of N, E, W wall area to building volume of Auckland sample hotels. 1.6 1.4 1.2 1 0.8 0.6 0.4 0.2 0 0.05

0.06 0.08 0.07 0.1 0.09 Ratio of S, E, W wall area to building volume

0.11

Figure 16. Ratio of south wall area to building volume of Shenyang sample hotels. Space heating energy

Space heating energy

Figure 14. Ratio of S, E, W wall area to building volume of Auckland sample hotels. 0.2 0.18 0.16 0.14 0.12 0.1 0.08 0.06 0.04 0.02 0 0.01

0.02 0.03 0.04 0.05 Ratio of north wall area to building volume

40

0.06

Figure 15. Ratio of north wall area to building volume of Auckland sample hotels.

0.2 0.18 0.16 0.14 0.12 0.1 0.08 0.06 0.04 0.02 0 60

100 120 80 Building volume (1000m3)

140

Figure 17. Building volume and space heating energy of Auckland sample hotels.

854

Space heating energy

increase of the design datum can still be identified. This preliminary study forms the basis of, and confirms the feasibility for, a further study using a much larger sample, which can identify the quantitative relationships between the space heating energy and a design datum. With a sufficient number of sample buildings, the gradient of the trend line of the design datum’s variation could be used to evaluate the strength of impact on space heating energy, and estimate the increase or decrease of space heating energy when a design datum is changed within a range and the other design data also impact the extra energy consumption differently and simultaneously. If the building code or design handbook can indicate the relationship between the design datum and the space heating energy, architects could make a significant difference to building energy efficiency through building passive design and take responsibility for it.

1.6 1.4 1.2 1 0.8 0.6 0.4 0.2 0 40

60

80

100 120 140 160 180 200 220 Building volume (1000 m3)

Figure 18. Building volume and space heating energy of Shenyang sample hotels.

and 45040 to 217800 m3 respectively. An decreasing trend in mean space heating energy of Auckland and Shenyang sample hotels are both associated with increasing in the ratios of building surface to volume (see Figs 17–18). 3

REFERENCES

CONCLUSIONS

This study introduced a method to use actual 12months energy consumption, occupancy, building design data to calculate their extra energy consumptions related to the winter indoor thermal conditions, which can roughly represent space heating energy, for comparing different hotel building designs. This method could also be applied to other building types and other climate conditions. For a climate with a hot summer and comfortable winter such as a hot-humid climate, the difference between mean daily electricity usage in the summer months and the other months of the year can be used to evaluate and compare different building designs focusing on the summer thermal performance. For a climate with both stressful summer and winter, the difference between mean daily electricity usage in the summer months or the winter months and the other months excluding the winter months or the summer months of the year can be used to evaluate and compare different building designs focusing on both the summer and the winter thermal performances. Although different design data related to the main architectural features can affect the extra energy consumption related to winter indoor thermal conditions differently and simultaneously, this study shows that the relationship between the increasing trend or decreasing trend in space heating energy and the

Clarke, J. 2001. Energy simulation in building design. UK: Butterworth Heinemann. Isaacs, N. and Crocker, N. 1996. Commercial Building Energy Survey: Hotels, Centre for Building Performance Research, Victoria University of Wellington. Becken, S. 2000. Energy Use in the New Zealand Accommodation Sector—report of a survey, Lincoln University. Ministry of Commerce, 2000. Energy data file. Wellington: Ministry of Commerce. Liu, J.P. 2000. Architecture Physicals. Beijing: China Construction Industry Publication. Energy Efficiency and Conservation Authority, 1996. EnergyWise Monitoring Quarterly: hotel sector. Issue 4, June 1996. Su, B. 2004. Architectural Design of Large Hotel and Energy Use for Internal Space Thermal Control. Proceedings of the 1st International Conference on Sustainability Engineering and Science, Auckland, July 6–9, 2004. Su, B. 2004. Mean Energy Used for Central Air-conditioning System related to Hotel Building Design. Proceedings of the 38th Annual Conference of the Architectural Science Association and the International Building Performance Simulation Association—Australasia Conference, Launceston, 10–12 November 2004. Su, B. & Aynsley, R. 2006. A case study on roof thermal performance of naturally ventilated houses in hot-humid climates under summer condition. Architectural Science Review 49(4): 399–407. Su, B. 2007. Building passive design and housing energy efficiency. Architectural Science Review 51(3): 277–286.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Climatic effects on building facades R. Miniotaite Kaunas University of Technology, Kaunas, Lithuania

ABSTRACT: The damage of external walls depends on a high moisture content, which in turn depends on high water absorption during driving rain. One example of such damage is the damage due to direct water penetration in homogeneous walls. The other negative effects of a high moisture content are impaired by heat insulation and accelerated degradation. The investigation of the external layer of walls and durability of different paints is carried out in this article. In case of bi-laminar system ‘‘paint film—the wall being painted’’ two opposite processes take place: water flow rate from outside towards the wall, and water vapour flow rate of the wall to outside. For the wall to be painted, optimum selection of paint is necessary. Investigation on the durability of the building walls external surfaces paints by modeled complex effects in a climatic chamber is purposive only after intermediate investigations and measurements of the substrate physical and mechanical properties that aid in predetermining durability. 1

INTRODUCTION

In the literature much information is given on paint and coatings, including the physical and chemical nature of paints, structure formation, and the results of investigations of physical and mechanical values (Lentinen 1996; Freitas et al. 1996; Carmeliet & Roels 2001; Hale 1976). However, the data of complex investigations of the surfaces already coated are insufficient. Durability of the paints applied to external surfaces of walls has been investigated by evaluation of the number of complex effects withstood in a climatic chamber in modelled cycles simulating outside climate variations (Brocken 1998; Miniotaite 1998a, 1999a, b, 2001; Ramos & Freitas 2006). The following factors of climate effects were evaluated: temperature, frequency and amplitudes of temperature variation; wind velocity, the amount of mineral admixtures containing precipitation falling on the walls, and ultraviolet radiation dose; the number of thaws and the influence of construction factors increasing or decreasing the energy of the above effects. Deterioration degree of the paints was evaluated by examination of physical-mechanical essence of adhesion and by determination of paint-substrate adhesion. Stage-by-stage methodics intended for analysis of different insufficiently investigated physical and mechanical values were prepared, followed by development of aggregate-complex investigation methods (Miniotaite 1999a, b, 2001). Functional link between physical and mechanical values of the paint used and the walls painted asked for substantial amendment of the data on water vapour permeability of the formed two-layer surface, and on water sorption coefficient. Detailed investigations of moisture-caused

857

deformations and the processes of opposite direction sorption-desorption as well as of ‘‘rain permeability’’—water vapour escape from the walls—enabled to compile the data bank necessary both for further investigation and practical application. The influence of counteraction of the water penetrating the surface layer of painted wall and of the escaping water vapour upon durability of the paint was evaluated according to a worked out program and following durability tests carried out in a climatic chamber in accordance with the program. Adequate results of finishing layers durability can be obtained only following classification of the paints according to their nature and of the substrates— according to their microstructure (Miniotaite 1999b, 2001). Summarized durability data on finishing layers of various origin applied to different substrates is reported in the present paper together with the indication of acceptable values of paints physical parameters, taking into account the origin of the paint. Investigation on durability and adhesion of the paints provided the possibility to determine the nature of paint destruction and to decide which of the paints required increase in fastness and which ones—in adhesion with the substrate (wall). 2

INVESTIGATIONS ON THE DURABILITY OF PAINTS

Some data on the physical and mechanical properties of materials intended for finishing, which predetermine the durability of such materials (Miniotaite 1999b, 2001). These data are often insufficient and

fragmentary. In order to properly investigate the durability of paints, it is necessary to compose a methodical chain—the scheme of stage-by-stage and chambertype consecutive investigations. The methods of generalized complex investigations were designed on the basis of the results of investigations carried out according to the stage-by-stage methods (Fig. 1). It was found that the specific physical and mechanical properties of individuals for paints might change when applied to different ‘‘paint—substrate’’ combinations. The comparative results of durability were obtained by the classification of coatings in three groups according to their composition. Specifically, these were classified as paints formed of: 1) aqueous polymeric dispersions; 2) polyacrylate also silicone solutions in organic solvents, or silicone dispersions; 3) silicates. In the case of a bi-laminar system ‘‘paint film—wall painted’’ we encounter water (i.e., rain) flowing from outside towards the wall and water vapour migrating from the wall to outside. The optimum selection of the paint is necessary for the wall to be painted. The water vapour may accumulate in the wall, when precluded from escaping through a very dense (vapour—tight) film, which might in turn cause blisters or result in delamination of the whole film or at least some sections. Several hundreds of combinations of coating and substrate are possible considering the many different surfaces to be coated with a selection of various paints. In this study, brick, cement plaster, lime cement plaster was the substrate chosen for investigation. The

choice was predetermined by the advantages of the silicate brick surface and its homogeneous capillary structure in order to have the results less scattered. The moisture diffusion processes in the silicate brick are comparatively constant across the surface (Miniotaite 2001, Brocken et al. 1998). The silicate bricks, lime cement plaster, cement plaster was coated using paints of different origin total 26 compositions. Analyses of compositions of the paints indicate that vapour permeability depends on the paint used, polarity of film-makers, and bonding agents used. Water vapour permeability coefficient was determined in 23◦ C environment according to requirements of the EN ISO 12572:2001 (EN ISO 12572 2001). Measurements were performed using 3 specimens of 100 mm diameter and 25 mm thickness of the uncoated materials (bricks, plasters) and 3 specimens with surfaces painted for each of the 26 coatings in the study. The painted specimens were fixed on a cup, paint facing down (cup method). Water vapour resistance Zp , [m2 · h · Pa/mg] is in reverse proportion to vapour permeability δ p , [mg/(m · h · Pa)]: Zp =

dx , δp

where dx = the thickness of samples of bricks, plasters, in meters. The specimens used for determination of vapour permeability were also used for determination of the surface water sorption coefficient by DIN 52 617 (DIN 52 617 1987). The specimens oriented with the paint facing downward were soaked in a water bath maintained at 20◦ C temperature. Water sorption coefficient w, [kg/(m2 · h0.5 )] is calculated: m w= √ , t

Figure 1. methods.

The principal scheme of complex research

(1)

(2)

where m = a mass of absorbed water related to 1 square meter of sample, in kg/m2 ; t = duration of soaking, in hours. After vapour permeability coefficient and surface water sorption were determined the new stage of methodical investigation in climate chamber took place where the value of resistance to complex effects was detected in the modelled cycles (Miniotaite 1999b). The necessary data on vapour permeability (vapour resistance) of the materials and the surface layer formed by such materials, ‘‘rain-penetration’’ (surface water absorption and water sorption coefficient) and coatings’ adhesion were determined in parallel with durability investigations carried out in a climatic chamber divided into two sections by masonry partition.

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RESULTS OF COMPLEX RESEARCH DEALING WITH PAINTED SURFACES OF MASONRY WALLS AND ANALYSIS

0,9

160

0,8

140

0,7

120

0,6

100

0,5

80

0,4

60

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40

0,2

20

0,1

w, kg/(m 2 .h 0.5 )

mg/(m.h.Pa).10 -3 p,

C, cycles;

1

180

0

0

D1

D2

D3

D4

D5

D6

D7

D8

D9

Paints mark Resistance to complex effects C Water vapour permeability coefficient Water sorption coefficient w

Figure 2. Resistance to climate effects of aqueous polymeric disperse paints on brick walls considering water sorption coefficient and vapour permeability. Notes: paints mark 0—non-painted brick. 1,2

200 180

1

160 140

0,6

100 80

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60 40

w, kg/(m 2 .h 0.5 )

0,8

120

C, cycles;

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20 0

859

D3

D2

D4

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D7

D5

D8

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Paints mark Resistance to complex effects C Water vapour permeability coefficient Water sorption coefficient w

Figure 3. Resistance to climate effects of aqueous polymeric disperse paints on cement plaster considering water sorption coefficient and vapour permeability. 1,6

200 180

1,4

160 C, cycles; p, mg/(m.h.pa).10-3

Physical properties (i.e., vapour resistance and water sorption coefficient) of the surface layer (0.025 m) of bricks, cement plaster and lime cement plaster coated with aqueous polymeric dispersion paints are compared in Figures 2, 3, 4 (Miniotaite 1999b). Their vapour resistance Zp = (0.66–0.84) m2 · h · Pa/mg [δp = (0.038–0.030) mg/(m · h · Pa)] is on the average 35% higher than those of subgroup ‘‘a’’ (paints D1, D2, D3, D4) (Fig. 3). As was foreseen before beginning these investigations the theoretical characterization of the paints (low vapour resistance—low rain penetration—‘‘good’’; high vapour resistance—high rain penetration— ‘‘bad’’) can be insufficient to evaluate a paint’s durability. Therefore, during the investigations changes in the surface layer resulting in decreases in adhesion between paint and the brick, were fixed. After the exposures in the climate chamber were finished and the results were analyzed, the reliability of the theoretical statement was verified. The value of vapour resistance due to complex effects created by application of the coating on the bricks was compared with the resistance of nonpainted bricks surface for the selected cycles.

0

0

3.1 Paints formed of aqueous polymeric dispersions

1,2 140 1

120

0,8

100 80

0,6

w, kg/(m2.h0.5)

3

200

0

p,

– in the warm part of the chamber room temperature is automatically maintained at θ i = (18 ± 2)◦ C and RH φ = (50–70)%; – an automatic climatic regime was maintained in the cold part of the chamber; – room temperature during 15 hours freezing down to θ e = −(15 ± 5)◦ C; – the temperature of a protective finished layer of the wall θse = (15–20)◦ C during 8 hours reheating; – UV light lamp was used during the last hour of heating; irradiation intensivity 600 W/m2 . – In the cold part of the chamber, water-spray equipment was installed. During a one hour water-spray operation, the finish of the wall had to be covered by a uniform water film (with a spray intensity = 1L/m2 min, temperature θ = (7–12)◦ C, and water pressure = 0.15 MPa); – air circulation at the velocity of v = (2–4) m/s was maintained by a ventilating device installed in the cold part of the chamber.

mg/(m.h.Pa).10 -3

The equipment of climatic effects was installed at one side of the partition, at the other—the equipment for maintenance of the indoor air condition. The basic conditions and means used for climatic tests were as follows:

60 0,4 40 0,2

20 0

0 0

D4

D1

D2

D3

D5

D7

D8

D6

D9

Paints mark Resistance to complex effects C Water vapour permeability coefficient Water sorption coefficient w

Figure 4. Resistance to climate effects of aqueous polymeric disperse paints on lime cement plaster considering water sorption coefficient and vapour permeability.

0,7

120

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w, kg/(m

2

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mg/(m.h.Pa).10 -3

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0

S1

S2

S3

S4

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S6

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Paints mark Resistance to complex effects C Water vapour permeability coefficient Water sorption coefficient w

Figure 5. The comparison of physical values of brick walls and durability of paints made out of polyacrylates and silicone solutions in organic solvents or silicone dispersions. 1,2

200 180

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100 80

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60 40

w, kg/(m

2

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S7

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Paints mark Resistance to complex effects C Water vapour permeability coefficient Water sorption coefficient w

Figure 6. The comparison of physical values of cement plaster and durability of paints made out of polyacrylates and silicone solutions in organic solvents or silicone dispersions. 200

mg/(m.h.Pa).10 -3

1,4

160 1,2 140 1

120 100

0,8

p,

Hardened films of polyacrylates’ or silicones constitute uniformity of the paints. They have no emulsifiers. Adhesion of paint—substrate is ensured by intermolecular interaction between bonding agent of paint and substrate. Because of the organic polymers of silicons are highly resistant, their films are stronger, more elastic and more resistant to temperature. In respect of a silicate brick, the paints of the group concentrate on the scale of low and average water sorption moisture (u24 < 2,5%). Macrostructure of the surface is uniform, not textured. The durability of the paints containing pigment of the above group is extremely sensitive to vapour resistance. With increase of Zp > 0,94 m2 · h · Pa/mg, negative influence of vapour resistance grows fast (Figs 5, 6, 7). Appearance of blebs indicate reduced adhesion.

1,6

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80

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w, kg/(m 2.h 0,5 )

Paints based of polyacrylate also silicone solutions in organic solvents, or silicone dispersions

C, cycles;

3.2

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0

C, cycles;

The tests carried out in the chamber indicated that the paints of identical vapour resistance could be compared even though the nature of deterioration and ageing as well as protective significance of such paints for the painted surface were different. The mechanism of such differences is explained taking into account additionally the complex effects of the moisture-caused deformations of the substrate and physical and mechanical values of the paints Durability of the aqueous polymeric disperse paints is described (considering two basic physical properties) in Figures 2, 3, 4. In the all cases the vapour resistance increase was influenced by using the acrylic bonding agent in appropriate proportions. The increase of vapour resistance is permissible and does not reduce durability of a properly selected composition of the paint. In the case of subgroup ‘‘a’’ permissible water sorption coefficient w < 0.65 kg/(m2 · h0.5 ) and the highest value of vapour resistance Zp < 0.62 m2 · h · Pa/mg [δ p > 0.041 mg/(m · h · Pa)] are suitable with respect to durability. In the case of subgroup ‘‘b’’ all physical parameters are high enough (Fig. 2). The reason for classifying two paints D5 and D9 as non-durable considers the bonding agent: specifically, the amount of bonding agent was insufficient. The interaction between paint and brick was inadequate. Durability of paints in the case of insufficiently stabilized compositions of the aqueous polymeric dispersions is defined as C < 80 cycles irrespective of the values of the water sorption and vapour permeability coefficients. Destruction of the paint specimens in the study is manifest through fast wrinkling of the film, mould formation, loss in the adhesion (small blisters) and washing off after 50–80 cycles.

60 0,4 40 0,2

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S1

S5

S2

S3

S7

S8

S9

S4

S6

Paints mark Resistance to complex effects C Water vapour permeability coefficient Water sorption coefficient w

Figure 7. The comparison of physical values of lime cement plaster and durability of paints made out of polyacrylates and silicone solutions in organic solvents or silicone dispersions.

860

200

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180 1

160 140

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w, kg/(m2.h0,5)

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The nature of destruction is close to lamination: occurrence of tiny blebs—their merging and bursting; lamination or ‘‘scale’’ type cracking of paint. Chemically instable paints are spotty. In some cases the spots are already observed following 100–110 testing cycles. Appearance of spots preceding mechanical disintegration of paints is not analysed in the article, even though the spots have aesthetic depreciation sense. Aesthetic destruction is also typical of some compositions with acceptable mechanical durability.

0,4

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C7

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Paints mark Resistance to complex effects C Water vapour permeability coefficient Water sorption coefficient w

3.3 Paints formed of silicates

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C4

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Paints mark Resistance to complex effects C Water vapour permeability coefficient Water sorption coefficient w

Figure 8. The comparison of physical values of brick walls and durability of silicate paints.

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Figure 9. The comparison of physical values of cement plaster and durability of silicate paints. 1,6

200 180

1,4

160 1,2 140 1

120 100

0,8

80

0,6

w, kg/(m2.h0,5)

C, cycles; p,mg/(m.h.Pa).10-3

As can be seen in Figures 8, 9, 10 negative influence of water sorption and positive influence of vapour permeability upon silicate paints durability prevail. This influence especially distinct for durability of paint (C8): water sorption coefficient distinctly decreased and vapour permeability distinctly increased. A high water sorption coefficient [w = 0.94 kg/ (m2 · h0.5 ) and w = 0.89 kg/(m2 · h0.5 )] has negative influence for indurable paints (e.g.: C1 and C2). These paints are formed of derivatives of salts of silicic acid and fillings inside the film. Adhesion with substrate is assured by forces of electrostatic interaction between surface groups. Van der Waals forces and polar interaction exist also. The paints of the above group get polarized on the scale of increased sand-lime brick’s water sorption moisture u24 = (6–8)%. A stiff framework is typical of surface macrostructure. Therefore the durability of paints of this group is negligibly influenced by moisture caused deformations of substrate. Parameters of surface water sorption and vapour resistance have approximately identical influence upon durability of paints. Vapour resistance should be Zp ≤ 0.7 m2 · h · Pa/mg and water sorption coefficient—w < 0.85 kg/(m2 · h0.5 ). Reducing water sorption coefficient to w = 0.5 kg/(m2 · h0.5 ) and vapour resistance

60 0,4 40 0,2

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0

0

0

C4

C6

C2

C8

C3

C1

C7

C5

Paints mark Resistance to complex effects C Water vapour permeability coefficient Water sorption coefficient w

Figure 10. The comparison of physical values of lime cement plaster and durability of silicate paints.

to Zp = 0.5 m2 · h · Pa/mg might increase durability (15–20)%. Durability of properly hardened paints of the above group should be attributed to the category of acceptable durability (C = 115–140 cycles). One hundred and fifty accelerated cycles in the climatic chamber correspond to 12 years at an average natural ageing. The resistance of the non-painted silicate bricks surface was found to be about 180 cycles. Following the effect of 170–180 cycles, the hydrosilicate crystalline structure of the silicate bricks surface thin layer (0.05–0.2 mm) underwent deterioration. Ground sand particles together with hydrosilicate deterioration products, dirt and other adhered aerosol inclusions crumbled off or were washed away comparatively easily. By analysing the results of grouped paints it was found that peculiar nature of vapour permeability and water sorption was typical of each group; distribution of the paint destruction is different. The nature

of paint destruction depends on the moisture-caused deformation of the substrate on the different level. 4

CONCLUSIONS

Investigation on the durability of the building walls external surfaces paints by modeled complex effects in a climatic chamber is purposive only after intermediate investigations and measurements of the substrate physical and mechanical properties that aid in predetermining durability, that is, sorption-desorption and moisture-caused deformations. Proper determination of paint durability is possible only in the case of simultaneous investigation of the wall surface layer. Influence of moisture deformations upon degradation of coatings depends on the porosity of materials of the surface being coated and on the origin and macrostructure of the coating. The intensity of surface destruction is non-proportional to the number of testing cycles. The increase of the number of modelled cycles accelerates destruction processes. The nature and signs of deterioration and ageing of paints of the separate groups develop in a different way. The intensity of surface destruction is nonproportional to the number of testing cycles. Durability of the paints formed out of aqueous polymeric dispersions is distributed in the direction of fast decrease of water sorption coefficient. Silicate paints resistance to climate effects is more dependent on surface water absorption and less dependent upon vapour permeability. Water absorption of non-durable paints is high; vapour resistance has no significant influence. Surface water absorption of usable paints decreases faster than vapour permeability. Resistance of paints made out of polyacrylates and silicone solutions in organic solvents or silicone dispersions to climate effects depends mined by investigations. There are no lamination and no cracks in case of nonpigment silicone paints. Hydrophobic properties of such paints decrease with time, however silicate bricks surface is protected quite well and for a sufficiently long time. Silicone paint containing pigment does not suit for painting of silicate brick walls. Durable coatings should be considered to be those which have withstood from 150 to 180 modeled complex cycles of climate effects. The coatings of acceptable (permissible) durability are those withstanding

110–150 cycles. Non-durable coatings are those withstanding less than 100–110 cycles—their aesthetic depreciation usually starts following 30–50 testing cycles. The coatings which begin to deteriorate after 80–100 cycles are not recommended either. REFERENCES Brocken, H.J.P., Spiekman, M.E., Pel, L., Kopinga, K. & Larbi, J.A. 1998. Water extraction out of mortar during brick laying: A NMR study, Materials and Structures 31 (5): 49–57. Carmeliet, J. & Roels, S. 2001. Determination of the isothermal moisture transport properties of porous building materials, Journal of Thermal Envelope & Building Science 24: 183–210. DIN 52 617. 1987. Determination of the water absorption coefficient of building materials. EN ISO 12572. 2001. Building materials; Hydrothermal performance of building materials and products. Determination of water vapour transmission properties. Freitas, V.P., Abrantes, V. & Crausse, P. 1996. Moisture migration in building walls—analysis of the interface phenomena. In Building and Environment 31(2): 99–108. Hale, D.K. 1976. The physical properties of composite materials. Journal Material Science, 11: 2105–2141. Lentinen, T. 1996. Capillary moisture transfer in combined porous building materials. In Proc. 4th Symposium on Building Physics in the Nordic Countries, 2: 483–490. Finland. Miniotaite, R. 1998. The durability of finish layers of external walls of buildings. In Proc. Conference on the subject of Construction and Architecture: 248–253. Kaunas: Technology. Miniotaite, R. 1999a. Compatibility of finishing layer and external surface of buildings’ walls from the standpoint of durability. Summary of the Thesis for a doctor’s degree. Lithuania, Kaunas: Technology. Miniotaite, R. 1999b. Compatibility of finishing layer and external surface of buildings’ walls from the standpoint of durability. Doctoral Dissertation, Lithuania: Technology. Miniotaite, R. 2001. The durability of finishing layer external surface of buildings’ walls. Monograph. Lithuania: Technology. Ramos, N. & Freitas, V. 2006. Evaluation strategy of finishing materials contribution to the hygroscopic inertia of a room. Research in Building Physics and Building engineering—Proceedings of the Third International Building Physics Conference, Concordia University, Montreal, Canada: 543–548.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Energy consumption related to winter housing thermal performance B. Su School of Architecture, UNITEC Institute of Technology, Auckland, New Zealand

ABSTRACT: Although different design factors related to the main architectural features, building elements and building materials can affect the extra energy consumption related to winter indoor thermal conditions differently and simultaneously, the previous study shows that the relationship between the increase in winter extra energy consumptions and the trend of the design datum’s variation (such as the ratio of window to wall variation) can still be identified (Su 2007). This study, with 101 sample houses in Auckland, is to identify the gradient of the trend line of the design datum’s variation associated with the increase in extra energy consumptions related to winter indoor thermal conditions. The gradient could be used to evaluate the strength of impact on extra energy consumption, and estimate the increase or decrease of extra energy consumption when a design datum is changed within a range and the other design data also impact the extra energy consumption differently and simultaneously. 1

INTRODUCTION

‘‘It is not hyperbole to suggest that the better design of new buildings would result in a 40–75% reduction in their energy consumption relative to 2000 levels, and that appropriate intervention in the existing stock would readily yield a 30% reduction’’ (Clarke 2001). There are a number of recent researches related to housing thermal performance and energy efficiency. Some focus on improving house thermal performance with insulation to save energy (Geoffrey & Boardman 2000; Verbeech & Hens 2005; Lloyd et al. 2008) and to improve indoor health conditions (Howden-Chapman et al. 2005; Su 2002 & 2006; Gilbertson et al. 2006; Bullen et al. 2008). Some developed a database of low energy homes and low energy techniques applied on them around world (Hamada et al. 2003). Some investigated major life-cycle energy inputs including embodied, operational and on-site construction energy (Pullen 2000; Mithraratne & Brenda 2004) and heating energy (French et al. 2007). Others have focussed on saving household appliance energy for housing energy efficiency (Lu 2004; Lopes et al. 2005; Hart & Dear 2005). For housing energy efficiency design, computer simulations are becoming available design tools (Smeds & Wall 2008; Caldas 2006; Karlsson & Moshfegh 2006). Some studies combine computer simulations and field study data for housing energy efficiency design or improve housing thermal performance (Simonson 2005; Wall 2006; Schuler, Weber & Fahl 2007; Tommerup Rose & Svendsen 2007). This study focuses on the impacts of main design factors related to architecture features, building elements and building materials on housing energy efficiency. This study not only identify impacts of main design factors

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on housing energy efficiency (the relationship between the increase in winter extra energy consumptions and the trend of the design datum’s variation) but also identify the strength of design factor on housing energy efficiency (the gradient of the trend line of the design datum’s variation associated with the increase in extra energy consumptions). Auckland does not normally need air conditioning or a ceiling fan for cooling during summer and only needs temporary heating during winter. In Auckland, the design of a building should focus more on its indoor thermal conditions and thermal performance related to winter conditions for building energy efficiency (Su 2004). On average of New Zealand houses, space heating is the largest single end-use (34%), followed by hot water (29%), appliances (13%), refrigeration (10%), lighting (8%) and cooking (6%). Heating to raise low temperatures is the main use of household energy (63%), comprising space heating (34%) and hot water heating (29%) which are closely related to the winter indoor thermal conditions of a house (Isaacs et al. 2006). To minimize the influence of differences in housing facilities and climates the study randomly selected 101 sample houses with sufficient insulation in their walls and roofs, which are in accordance with the New Zealand Standard: NZS 4218, in the Auckland region which had been using electricity as their only energy resource for space heating, hot water, cooking, lighting, refrigeration and other appliances. The electricity consumption data for each sample house are for the same period of twelve months. Mean annual energy, mean winter energy and mean othermonths energy of total sample houses are 0.06117, 0.07770 and 0.05591 kWh/m3 day. Mean winter extra energy of total sample houses is 0.02177 kWh/m3 day.

2 2.1

DATA ANALYSIS Ratio of building surface to volume

The ratios of building surface to volume of sample houses are 0.45 to 0.96. A house with a low ratio of building surface to volume has a small external surface area per unit of indoor space from which to lose heat to the outdoors, and uses less energy for space heating, hot water and other appliances, which can be affected by indoor thermal conditions during the winter. An increase in mean winter extra energy usage is associated with an increasing trend in the ratios of building surface to volume (see Fig. 1). The mean winter extra energy usage is not only related to and impacted by a particular design datum but also by other design data, both differently and simultaneously. The positions of those real data points are also impacted by other design factors. The gradient of the trend line of the design datum’s variation could be appropriately used to measure the strength of a design datum’s impact. The gradient of trend line is 0.4. For the Auckland houses under the local climate conditions, Equation 1 shows the quantitative relationship

Ratio of surface/volume

Winter extra energy is 8.9% of annual energy and 28% of winter energy of sample houses. Design data for the study were obtained from copies of the sample house plans provided by Auckland City Council. The study used real energy consumption data of a house to calculate the winter extra energy consumption resulted from the impact of winter indoor thermal conditions of the sample houses, which can be used to compare different designs of houses for energy efficiency and quantitatively indicate the difference of winter indoor thermal conditions between sample houses. The study uses the difference between mean daily electricity usage in the winter months (June, July and August) and the other months of the year to represent the extra energy usage related to winter indoor thermal conditions of the sample houses. The difference between mean daily electricity usage in the winter months and the other months can roughly represent the extra winter energy consumption, which mainly comprises space heating, extra energy for hot water heating and all appliances, which are impacted by the winter indoor thermal conditions of a house. The smaller difference between mean daily usage in winter months and the other months can roughly represent the better indoor space thermal conditions responded to the winter climate conditions. This study uses the mean daily electricity usage per unit volume of house indoor space (kWh/m3 day) as the basic energy consumption unit, because the extra energy usage is mainly related to the indoor thermal conditions. Winter extra energy usages of sample houses for this study are 0.00166 to 0.05852 kWh/m3 day.

1 0.9 0.8 0.7 0.6 0.5 0.4 0.3 0

0.01 0.02 0.03 0.04 0.05 0.06 Mean energy difference of winter and rest months

Figure 1. Ratios of building surface to volume and mean winter extra energy usages (kWh/m3 day) of sample houses.

between the increase or decrease of winter extra energy consumption and the decrease or increase of ratio of building surface to volume and impact strength of ratio of building surface to volume on winter extra energy usage. EWE = 0.4RSV

(1)

where EWE = the increase or decrease of mean winter extra energy consumption, kWh/m3 day; RSV = the increase or decrease of ratio of building surface to volume. 2.2

Impact of windows

Ratios of total window area to total wall area of sample houses are 0.071 to 0.421. Ratios of total window area to total indoor space volume of sample houses are 0.036 to 0.149. Ratios of window area on north wall to north wall area of sample houses are 0 to 0.547. The trends of ratios of window area to wall area, ratios of window area to internal space volume and window area on north wall to north wall area of sample houses increase with the increase of their mean winter extra energy usages (see Figs. 2–4). For the Auckland houses under the local climate conditions, Equation 2 and Equation 3 show impact strength of ratio of window area to wall area and ratio of window area to indoor space volume. EWE = 1.2RWW

(2)

where EWE = the increase or decrease of mean winter extra energy consumption, kWh/m3 day; RWW = the increase or decrease of ratio of window area to wall area. EWE = 1.8RWI

(3)

where EWE = the increase or decrease of mean winter extra energy consumption, kWh/m3 day; RWI = the increase or decrease of ratio of window area to wall area.

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Ratio of window/wall

0.5 0.4 0.3 0.2 0.1 0 0

0.01 0.02 0.03 0.04 0.05 Mean energy difference of winter/rest months

0.06

Ratio of window area/ indoor space volume

Figure 2. Ratios of window area to wall area and mean winter extra energy usages (kWh/m3 day) of sample houses.

0.2 0.18 0.16 0.14 0.12 0.1 0.08 0.06 0.04 0.02 0 0

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0.03

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Mean energy difference of winter/rest months

Ratio of N window area/ N wall area

Figure 3. Ratios of window area to internal space of sample houses and mean winter extra energy usages (kWh/m3 day) of sample houses.

0.6 0.5 0.4 0.3 0.2 0.1 0 0

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0.02

0.03

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Mean energy difference of winter/rest months

Figure 4. Ratios of window area on north wall to north wall area and mean winter extra energy usages (kWh/m3 day) of sample houses.

Ignoring one design factor could damage the energy efficiency of an entire building. Large single-glazed windows on north walls are simply a traditional local design convention heating based on older houses constructed without insulation materials for daytime passive solar. For an old house without insulation, when the R-value of both the walls and the single-glazed windows are low, the more radiant solar heat that comes into the indoor space, the greater may be the positive impact on indoor thermal conditions and the reduction of energy consumption for space heating (Su 2008). Nowadays those singleglazed windows create wall areas with low R-value in houses with good insulation but cold indoor surface areas. A house with good insulation loses a lot of heat through these ‘‘cold holes’’, which damages the entire housing passive design in terms of energy efficiency and indoor thermal comfort. If the R-value of a window can match or come close to the insulation level of a wall—for example, a double glazed window— a large window on the north wall can truly and positively impact on the energy efficiency and the indoor thermal conditions of a local building. Successful building passive design for energy efficiency should combine local design traditions with new building materials and new design concepts for building energy efficiency. The occupants of a house with insulation may enjoy passive solar heating through large single-glazed windows during the short winter daytime, but during the longer winter night time, even when heaters are on and the mean indoor air temperature is at a comfortable level, suffer unbalanced bodily heat loss to the cold indoor surfaces through radiation. With current building codes imposing no limitation on the ratio of north-facing window area on north walls to total north wall area, some local new houses have a large portion of north wall area covered by single-glazed windows simply because installing single-glazed windows is cheaper per square metre than building solid walls. 2.3

Windows are commonly weak elements of building thermal performance. The windows of all sample houses for this study are all single glazed windows. The thermal resistance (R-value) of a single glazed window is very low to compare with wall and roof with insulation. With the single glazed windows, increase in the ratio of window and wall of an Auckland house can impact the winter internal space thermal conditions negatively and significantly and increase the space heating energy. Successful building passive design for energy efficiency should take different design factors related to architectural features, building elements and building materials into consideration as a whole.

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Impact of walls

An Auckland house with good orientation will usually have high ratios of north wall area to total wall area or to indoor space volume. Good orientation should improve indoor thermal conditions and energy efficiency, but an increase in mean winter extra energy usage is not associated with a decreasing trend in ratios of north wall area to total wall area and to indoor space volume for the sample houses with insulation (see Figs. 5–6). All windows in the sample houses are single-glazed and the mean ratio of northfacing window area to north wall area (0.28) of sample houses is higher than the ratios of south-facing (0.18), east-facing (0.20) and west-facing windows (0.22). Therefore increasing the ratio of north wall area to total

0.7 Ratio of roof area to indoor space volume

Ratio of N wall area/total wall area

0.8 0.7 0.6 0.5 0.4 0.3 0.2 0.1 0

0.4 0.3 0.2 0.1 0

0

0.01 0.02 0.03 0.04 0.05 0.06 Mean energy difference of winter and rest months

Figure 5. Ratios of north wall area to total wall area and mean winter extra energy usages (kWh/m3 day) of sample houses. Ratio of N wall area/ indoor space volume

0.6 0.5

0

0.01

0.02 0.03 0.04 Mean winter extra energy

0.05

0.06

Figure 7. Ratios of roof area to indoor space volume and mean winter extra energy usages (kWh/m3day) of sample houses.

0.3 0.25

RRI = the increase or decrease of ratio of roof area to indoor space volume.

0.2 0.15 0.1 0.05

3

CONCLUSIONS

0 0 0.01 0.02 0.03 0.04 0.05 0.06 Mean energy difference of winter and rest months

Figure 6. Ratios of north wall area to indoor space volume and mean winter extra energy usages (kWh/m3 day) of sample houses.

wall area also significantly increases the north-facing single-glazed window area with a very low R-value compared with the insulated walls. The negative effect of increasing the ratio of north-facing window area to north wall area may be stronger than the positive effect of increasing the ratio of north wall area to total wall area for the houses with insulation and single-glazed windows. The energy efficiency of a house with sufficient insulation, good orientation and building form can be negatively impacted by single-glazed windows with low R-value. 2.4

Impacts of roof

As a New Zealand house loses about 40% of its heat through the ceiling and roof during the winter, an increase in the ratio of roof area to building volume will increase the total heat loss and the space heating energy consumption of a house. An increase in mean winter extra energy usage is associated with an increasing trend in ratios of roof area (excluding eaves) to indoor space volume of the sample houses (see Fig. 7). Equation 4 shows impact strength of ratio of roof area to indoor space volume. EWE = 1.8RRI

(4)

where EWE = the increase or decrease of mean winter extra energy consumption, kWh/m3 day;

Generally the Auckland summer is comfortable. Normally an Auckland house does not need the energy for internal space cooling during the Auckland summer and only needs temporary heating during the winter. The difference between mean daily electricity usage in the winter months and the other months can roughly represent the extra winter energy consumption, which mainly comprises space heating, extra energy for hot water heating and all appliances, which are impacted by the winter indoor thermal conditions of a house. Winter extra energy is 8.9% of annual energy and 28% of winter energy of sample houses. Winter extra energy is a major portion of energy consumption of an Auckland house. Auckland house passive design for energy efficiency should focus more on the winter house thermal performance and reduction of winter extra energy. There are no universal housing passive design guides for the different locations and climates. The housing passive design guidelines should be related to the major thermal problems of local climate conditions and can derived from the study of local housing energy consumption data and housing design data. With a large number of sample houses, the study not only can identify the relationships between the winter extra energy usage (or the summer extra energy) and design factors but also can identify the impact strength of design factor on the winter extra energy usage. The gradients of increasing or decreasing trend lines of design factors to the increase of winter extra energy (summer extra energy) have the quantitative value to develop passive design guides for building energy efficiency for local new housing development. The gradient could be used to evaluate the strength of impact on extra energy consumption, and estimate the increase or decrease of extra energy consumption

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when a design datum is changed within a range and the other design data also impact the extra energy consumption differently and simultaneously. REFERENCES Bullen, C., Kearns, R.A., Clinton, J., Laing, P., et al 2008. Bringing health home: Householder and provider perspectives on the healthy housing programme in Auckland, New Zealand. Social Science & Medicine 66(2008): 1185–1196. Caldas, L. 2006. Generation of energy-efficient architecture solutions applying GENE_ARCH: An evolutionbased generative design system. Advanced Engineering Informatics 22(2008): 59–70. French, L.J., Camilleri, M.J., Isaacs, N.P., & Pollard, A.R. 2007. Temperatures and heating energy in New Zealand houses from a nationally representative study—HEEP. Energy and Buildings 39(2007): 770–782. Clarke, J. 2001. Energy simulation in building design. UK: Butterworth Heinemann. Gilbertson, J., Stevens, M., Stiell, B. & Thorogood, N. 2006. Home is where the hearth is: Grant recipients’ views of England’s Home Energy Efficiency Scheme (Warm Front). Social Science & Medicine 63(2006): 946–956. Hamada, Y., Nakamura, M., Ochifuji, K., Yokoyama, S., et al. 2003. Development of a database of low energy homes around the world and analyses of their trends. Renewable Energy 28(2003): 321–328. Hart, M. & Dear, R.D. 2005. Weather sensitivity in household appliance energy end-use. Energy and Buildings 36(2004): 161–174. Howden-Chapman, P., Crane, J., Matheson, A. Viggers, H. et al. 2005. Retrofitting houses with insulation to reduce health inequalities: Aims and methods of a clustered, randomised community-based trial. Social Science & Medicine 61(2005): 2600–2610. Isaacs, N. et al. 2006. Report on the year 10 analysis for the household energy end-use project (HEEP). Wellington: Building Research Association of New Zealand. Karlsson, J.F. & Moshfegh, B. 2006. Energy demand and indoor climate in a low energy building—changed control strategies and boundary conditions. Energy and Buildings 38(2006): 315–326. Lloyd, C.R., Callau, M.F., Bishop, T., & Smith I.J. 2008. The efficacy of an energy efficient upgrade program in New Zealand. Energy and Buildings 40 (2008): 1228–1239. Lopes, L., Hokoi, S., Miura, H. & Shuhei, K. 2005. Energy efficiency and energy saving in Japanese residential

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buildings—research methodology and surveyed results. Energy and Buildings 37(2005): 698–706. Lu, W. 2004. Potential energy saving and environmental impact by implementing energy efficiency standard for household refrigerators in China. Energy Policy 34(2006): 1583–1589. Merbeech, G. & Hens, H. 2005. Energy savings in retrofitted dwellings: economically viable? Energy and Buildings 37 (2005): 747–754. Milne, G., & Boardman, B. 2000. Making cold homes warmer: the effect of energy efficiency improvements in low-income homes. Energy policy 28 (6–7): 411–424. Pullen, S.F. 2000. Energy used in the construction and operation of houses. Architectural Science Review 43(2): 87–94. Schuler, A., Weber, C. & Fahl, U. 2007. Energy consumption for space heating of West-German households: empirical evidence, scenario projections and policy implications. Energy Policy 28(2000): 877–894. Simonson, C. 2005. Energy consumption and ventilation performance of a naturally ventilated ecological house in a cold climate. Energy and Buildings 37(2005): 23–35. Smeds, J. & Wall, M. 2008. Enhanced energy conservation in houses through high performance design. Energy and Buildings, 39(2007): 273–278. Standards New Zealand (2004). New Zealand Standard 4218–2004: Energy Efficiency—Small building envelope. Wellington: SNZ Su, B. 2002. A field study of mould growth and indoor health conditions in Auckland dwellings. Architectural Science Review 45(4): 275–284. Su, B. 2006. Prevention of winter mould growth in housing. Architectural Science Review 49(4): 385–390. Su, B. 2004. Mean energy used for central air-conditioning system related to hotel building design. Proceedings of the 38th Annual Conference of the Architectural Science Association, Launceston: ANZAScA. Su, B. and Aynsley, R. 2006. A case study on roof thermal performance of naturally ventilated houses in hot-humid climates under summer condition. Architectural Science Review 49(4): 399–407. Su, B. 2007. Building passive design and housing energy efficiency. Architectural Science Review 51(3): 277–286. Tommerup, H., Rose, J. & Svendsen, S. 2007. Energyefficient houses built according to the energy performance requirements introduced in Denmark in 2006. Energy and Buildings 39(2007): 1123–1130. Wall, M. 2006. Energy-efficient terrace houses in Sweden simulations and measurements. Energy and Buildings 38(2006): 627–634.

Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Green energy and indoor technologies for smart buildings F. Vranay, Z. Vranayova & D. Ocipova Technical University of Kosice, Institute of Building and Environmental Engineering, Slovakia

D. Lukasik Honors, Inc., Liptovsky Mikulas, Slovakia

ABSTRACT: The renewable energy sources are domestic sources of energy that help to enhance the safety of energy supplies and the diversification of energy sources. More and more, the renewable energies contribute to the three pillars of the sustainable development in the economy, environment and the society. An increased utilization of renewable energy sources in the heat and electricity generation is one of priority tasks of the Slovak Republic to boost the use of domestic energy potential and thus to decrease the Slovakia’s dependence on imported fossil fuels. Heat pumps offer the most energy-efficient way to provide heating (central and water heating) and cooling in many applications, as they can use renewable heat sources in our surroundings. This article determines applicability and effectiveness of this system on the case study performed by the experimental workstation of applied research and development of RSE in Kosice, Slovakia. 1

INTRODUCTION

Analyses show the fact that the process of climatic changes is closely tied to the increase of greenhouse emissions in the atmosphere. Depletion of planet’s sources is exceeding 30% of its natural ability of recovery (WWF 2008). These facts support the motivation to decrease energetic costs of energy primary sources in combination with reduction of greenhouse gases emissions. It is the primary ecological debt of the planet in combination with climatic changes that creates the danger of generation of future costs ranging from 5 to 20% GDP (Stern 2006). The report developed by commission administered by Lord Stern also shows the fact that yearly investments of 1% GDP during next 20 years will enable volume stabilization volume of emissions in atmosphere, and therefore also the increase of temperature and climatic changes to acceptable level. The aim is to reduce the greenhouse gas emissions by 80% by year 2050. From the point of view of all consumed energy, which is the main producer of emissions, the buildings contribute by 40%. One way of achieving this goal is reduction of energetic costs of primary energetic sources of buildings for delivery of heating and cooling. Together with reduction we focus on the technique of production of energy and its consumption by intelligent systems of buildings. 1.1

Description of the aim

The experimental workplace: Economic Research Centre for Renewable Energy Sources and Distribution systems was founded with the purpose of

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investigating possibilities to reduce the energetic costs of buildings tied to economy. The Centre operates in the region of Slovakia, which is placed in Central Europe, in mild climatic area. Its founders are the most significant Slovak Universities. Two of them are of technical orientation and one of them is a university of economics orientation together with enterprising subject. The aim of this association is to create conditions for active solutions of economical and legal questions related to implementation of renewable resources and sources of energy with low level of emission production. Current situation is focused on sustaining emissions on limited figures by allocating quotas by the European Union Commission with exact assignment to the producer. Emissions which are within quota limits are funded by the state. The costs related to elimination of damage caused by emissions are not included in the price of energy. In this aspect, the renewable resources are economically handicapped in comparison to fossil fuels. The level of handicap is created by the value of emissions produced by particular type of energy source to energy unit (EC 2006; Florenz 2008). The aim of the Centre is: – To define unified market of renewable energy sources (RSE) and energy produced by fossil fuel. – To handle the priority of RSE input into energy distribution system. – To define legal conditions for investments and investors’ investments protection. – The review of types of RSE based on geographical and other characteristics and set their mutual priorities.

– To ascertain principles for review of influence of forced character of energy source in a particular distribution system. – To solve elimination of RSE implementation legal barriers. – To consider the lack of European Union’s own energy resources. – To understand the coherence of solutions and implementation of projects in practice and legislation.

Energy consumption for heating CO2

(GJ/year)

t/year

205

107

Thermostatisation = 36%

75

1080

1500

Heat pump = 12%

2007 -

1250

1670

1490

1 500

1630

2005 -

1520

2 000

= 100%

Object Insulation = 52%

1997 -

1930

2 500

1996 - Energy consumption for heating

1460

Capillary mats = 8%

2008 -

68

46

500

260

380

1 000

260

Energy consumption

3 000

2220

3 500

3200

4 000

2009 - Co-generation = 8%

17

0

1996 1997 1998 1999 2000 2001 2002 2003 2004 2005 2006 2007 2008 2009 Measurements in years

2 2.1

THE DESCRIPTION AND OBJECTIVES OF EXPERIMENTAL WORKSTATION

Figure 1. Energy consumption. Process of primary energy consumption for heating the experimental laboratory and the amount of CO2 emissions per year.

The aim of the project

By founding experimental workstation we conduct applied research and development of RSE—heat pump and indoor technologies.

3 3.1

– To cut down CO2 emissions to less than 90% of current level. – To ensure that the building is supplied by at most 20% of fossil fuel energies and that at least 80% energies have come from renewable sources. – To minimalize operation costs. – To secure standard working environment with the aim of increasing the standard by use of summer cooling. – To ensure economic return on investments. – To provide economic justification for symbiosis use of sophisticated fossil fuel—natural gas with RSE energy carriers. – To propose legislative outputs. – To ensure the connection between the pedagogical practice with applied research and implementation of outputs.

Original condition

The object was without changes to circuit constructions and the heating system since its construction until 1996. The supply of energy came from central distribution system. Heating plant—transmission station—object. The source was the city’s heating plant burning up coal and natural gas in combination with electrical energy production. Heat consumption as measured at its entry to the object of heating was 3200 GJ (=100%). These initial data will be used to compare the effects of taken subsequent measures. 3.2

Object insulation and change of windows

In 1997 change of windows and insulation of circuit constructions of the object took place. The average consumption of heat in years 1997–2004 was 1678 GJ (=52% of original consumption)

The experimental laboratory is an administrative building with space for laboratory and training (lectures). It was built in 1975 according to then valid standards and principles. The characteristics of constructions and heating system were influenced by the need for energy and heat supply. This object did not contain air condition or cooling system. This building was modified by long-running premeditated process to reduce energy supplied for operation of the object. These can be divided into two phases:

3.3

Thermostatisation

In 2005, the average consumption was reduced to 1165 GJ (=36%) by hydraulic regulation and installation of thermostatic valves. Installation of thermostatic valves on heating bodies enabled local regulation of each heating body. There is also the possibility to consider heat gain in each room by reduction of necessary heating output.

1. Constructional and technical modifications of the object effecting the reduction of heating consumption. 2. Change of energy and fuel source, to secure increase in efficiency of primary energy use and minimalization of CO2 emissions. Fallout of these modifications in both phases 1 and 2 in years 1996–2006 can be seen in Figure 1.

DESCRIPTION OF MEASURES, PHASE NO.1

4 4.1

DESCRIPTION OF MEASURES PHASE NO. 2 Heat pump (HP)

The main change in the system of heat supply happened after installation of RSE—Heat pump—in the object in 2007 (see Figure 2). The original source works as peak and backup source. Heat energy is taken

870

GAS

5

6

1

3

2

3

4

10 10

9 7

11-13

10 8

CGE electricity

distribution net for electricity

Figure 2. Legend: 1 Heat pump (HP), 2 Co-generative element (CGE), 3 Accumulation tank, 4 Heated object, 5 Supporting and top source, 6 Heat exchanger, 7 Absorbing well, 8 Suction well, 9 Well pump, 10 Circulatory pumps—engine room, 11–13 Circulatory pumps—object.

heat pump is electrical energy. The CO2 emission coefficient for production of electrical energy in Slovakia is declared to be 0.640 Kg/kWh. The CGE will be installed to the heat pump system. It will produce the electrical energy to drive heat pump by burning up the gas and it will replace supply of electricity from distribution net. Simultaneously, heat produced by burning will be used to increase the temperature of heating body emerging from heat pump. The CO2 emission coefficient for combined production of heatelectricity in co-generative element is 0.23 kg/kWh. The benefit of CGE is cheaper electrical energy produced directly at the place of consumption and also significant decrease in produced CO2 emissions. It has no significant effect on the amount of supplied heat (=8%). The produced CO2 emissions calculated in tones per year can be seen in Figure 1, the right column.

from groundwater. The energy gained by its cooling transforms heat pump to more efficient mode. It is driven by electric energy. By using heat pump, coefficient of performance (COP) 1:3 was reached. In numbers, this means that the amount of electric energy is 380 GJ (decrease to 12%) If we compare years 2007 and 2006 (to find out the effect of renewable resource only), the result of the source change is the decrease to (380/1080=35%), which stands for 65% saving.

4.2

Capillary mats

Installation of capillary mats to the ceilings anticipates reduction of necessary primary energy needed to supply the heat pump to 260 GJ. The reason of heating system change from conventional heat conducting heating units (bodies) to emanating low temperature way is the increase of heat production effectiveness. COP of heat pump production is expected to be COP 1:4.3 instead of 1:3. However, if we compare original levels in year 1996, the consumption of primary energy is at the level of just (8%). The expected consumption of primary energy when compared with year 2006 means reduction to (260/1080 = 26%). The change of system brings on new quality in form of the possibility to conduct summer cooling in this object. It is enabled by parameters of water in the well in summer (15 degrees) and the demand for cooling water of 18 degrees. The heat pump is completely excluded from the process of cooling and we only need pumping labor for circulation of cooling water. The proportion of pumping labor will be determined by COP of coolness production on level 1:2, which means 7 times increase in effectiveness.

4.3

Co-generative element (CGE)

In the process of production we monitor the CO2 emissions parameter. The source of energy for the drive of

871

5

HEAT AND COOLNESS SOURCE AFTER IMPLEMENTED CHANGES

The system enables production and supply of heat for heating and cooling to ensure the desired climate in summer. It is made of wells (suction and absorbing types), heat pump, co-generative element and heating-cooling system with help of capillary mats. The energetic potential is contained in ground water. After pumping, this water is cooled by approx 4 degrees. Gained energy is transformed by heat pump to higher level (the demand of capillary mats is approx 35 degrees). Water is then heated by energy produced by burning the gas in the co-generative element and will be transferred via accumulation and regulation system into heating system. Heating system is made of ceiling capillary mats (so there is no heat wasted). Electrical energy produced in CGE will be used as fuel for heat pump and other supporting electrical devices. In the cycle of production and consumption, the interesting part is geothermal energy gained from primary energy in the form of gas, which makes substantial part of operating costs. 5.1

Operation in winter

Main source is the renewable low potential geothermal energy (groundwater of approx 12 degrees). Transformation of heat from low potential level into higher (approx 35 degrees for heating with capillary mats) is happening in heat pump, propulsion of which is done via electrical energy produced by burning up the gas in CGE. Production efficiency of electrical energy in CGE by burning up the gas is 30–35%. The rest of heating energy will be used for raising the temperature of heating water, which enables the efficiency of CGE of 85–90%. Produced electrical energy will

be used for propulsion of heat pump and supporting electrical devices (circulatory pumps). Evaluation of electrical energy for production and delivery of heat is defined by heating factor COP, which has the value of 4.3 during heating season. This means 4.3 times higher amount of produced heat than that delivered in electrical energy. The overall energetic utilization of gas via production of heat, production of electricity and the labor of heat pump and delivery of geothermal energy at 200%. 5.2

heat pump is 6.7, then the proportion of energies is as follows: 85% of energy is from geothermal source and 15% of energy is from electrical energy. Overall efficiency of co-generative element is 90%.Calculating primary energy 15%/0.9 = primary energy = 16.7%. Proportion of geothermal energy will be decreased to 100%–16.7% = geothermal energy = 83.3%. The rest of produced heat from co-generative element will be used in distribution net for other consumers for heating and heating of hot water. The realized project creates real environment for effective implementation research of technologies in laboratory and operative conditions: technologies of co-generative elements, heat pumps, thermal capillaries, and technologies in field of measurement and regulation. The solution is the project with possibility to repeat it on other similar applications as well as the utilization of experience and determination of economical expedience of researched technologies implementation. The quality of buildings’ environment influences health, efficiency and comfort of building’s users. Current studies show that the costs of internal environment quality improvement are often higher (for company, employer and building’s owners) than costs of energy consumed in the building. Energetic declaration without declaration of relation to quality of inner environment therefore does not make sense and is not sufficient. Specification and consideration of design criteria for parameters of inner environment are needed for design of buildings, energetic calculations, operation and regulation of buildings. Next phase of the research will be evaluation of operative behavior of the building, interaction with building constructions and study of inner climate parameters.

Operation in summer

The energy of water from the well is used for cooling during the summer. In case of cooling by capillary mats, the temperature of water in capillaries is 18 degrees. Groundwater sustains temperature of approx 15 degrees in summer. Its temperature is lower than the temperature demanded for cooling. It is not necessary to produce coolness; the necessary temperature of 18 degrees will be produced by mixing in the returned cooling water which is already heated. The demand for energy for production and distribution of coolness will consist of just pumping labor for circulation of cooling water via capillary mats. Heat pump doesn’t have to be in operation during cooling. Electrical energy produced by CGE will be used as propulsion of circulatory pumps and the heat produced by burning the gas will be used for heating up the water. The production and delivery of coolness via COP is 14. If we use the gas burned in CGE together with heat to heat up the water to produce electrical energy, the utilization of gas is 480%. Energetic utilization of gas for heating, cooling and heating up the water, during the whole year is 270%. Yearly balance shows the proportion of production of heat and coolness 3/1. COP of heating is 4.3 and COP of cooling is 14, yearly balance of whole system is 6.7. 6

ACKNOWLEDGEMENTS This contribution was written as a solution of ITMS ‘‘26220120018’’ and as contribution of EcoFund of company SPP (Centre 2009).

CONCLUSIONS

REFERENCES

The balance of measured and calculated figures of operation of the building confirms the correctness of assumptions and accuracy of projected balance. From the point of view of energetic balance, the consumption of primary energy (geothermal energy is not being considered) was decreased from original 3200 GJ in year 1996 to 260 GK. This means decrease by 92%. The structure of resource and fuel base was changed. The coverage of consumption has significantly lower amount of CO2 emissions. The decrease from original 205 tones per year to 17 tones per year makes 92%. The structure of fuel for production of primary energies in year 1996 consists of 100% fossil fuel. After adjustments, it consists of geothermal energy and electrical energy produced by burning gas in year 2009. If eventual COP for yearly operation of

Commission of the EC. Communication from the Commission to the Council, the European Parliament, the European Economic Committee and the Committee of the Regions. Putting knowledge into practice: A broad-based innovation strategy for the EU. Brusel, 13.9.2006. Economic Research Centre for Renewable Energy Sources and Distribution systems. Green zone of Kosice as technical and economic symbiosis of RES and natural gas in energetical smart building. Kosice, 30.11.2008. Florenz, K. Interim report on the scientific facts of climate change: findings and recommendations for decision-making. Temporary Committee on Climate Change. Brusel, 22.1.2008. Stern, N. 2006. Review on the Economics of Climate Change In http://en.wikipedia.org/wiki/Stern_Review. WWF. 2008. Living planet report. In http://assets.panda. org/downloads/living_planet_report_2008.pdf.

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Challenges, Opportunities and Solutions in Structural Engineering and Construction – Ghafoori (ed.) © 2010 Taylor & Francis Group, London, ISBN 978-0-415-56809-8

Indoor air quality, distribution systems and energy simulations R. Nagy & I. Šenitková Technical University of Košice, Institute of Building and Environmental Engineering, Slovakia

ABSTRACT: The ventilation can be provided by either natural or mechanical system. The control of ventilation rates is better in a mechanical system, which usually also has a lower energy consumption. The field measurement confirmed that the indoor air quality in the schools is generally unacceptable by lower ventilation rates because of not respecting the occupancy density. Fresh air can be taken directly to the room without causing drafts even in a cold climate, if the airflow from a single properly designed opening does not exceed 10 l/s, and the air velocity is 2–3 m/s at the inlet. The ventilation system shouldn’t ensure only thermal comfort and hygienic level of indoors, but also air distribution to occupancy zone in order to pollutants removal efficiency, pollutants displacement or pollutants dilution. The non-uniformity pollution related to selected distribution system was investigated using the CO2 concentration. 1

CHARACTERIZATION

The objective of this research is to evaluate the performance of natural ventilation systems in school buildings and to perform the experimental measurements in order to analyze possible ways for indoor air quality improvement respecting energy in the schools. Several types of air distribution were selected for presented experiment within the frame of conventional mixing ventilation system and existing natural ventilation system. The distribution systems were installed in naturally ventilated school building in identical classrooms. The carbon dioxide concentrations were studied under indoor climate parameters (temperature, relative humidity and air movement). Three different categories of indoor environment are specified for indoor ventilated spaces. The category I correspond to a high level of expectation and lead to a highest percentage of satisfied occupants in respect of indoor environment, category II a medium level of expectation and category III to a moderate level of expectation. The designer may also select different levels using the same principles related with specific thermal loss. Recommended values of indoor CO2 concentration for ventilated buildings are estimated as maximum CO2 concentration above the outdoor concentration. Maximum CO2 concentration is 350 ppm (category I), 500 ppm (category II) and 800 ppm (category III) above background outdoor concentration for expected indoor air quality (STN EN 15251:2007).

2

respecting of 5 air distribution schemes (1 natural ventilation, 4 mixing ventilation) in 24 measuring points. Consequently the energy demand calculations were realized for selected distribution system.

Figure 1.

Experimental model room 3D.

Figure 2.

Experimental model room 3D.

METHODS

Presented experimental measurements deal with non-uniformity distribution of CO2 concentration

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Figure 3. Experimental model room measuring points location (dimensions 10.7 × 5.8 m).

Table 1.

Characteristics of steady state conditions.

Ventilation system

Total ventilation rate qTOT [l/s]

Natural Mixing

17 55

Average supply air temperature [◦ C]

supply air humidity [%]

Average CO2 concentration [ppm]

Number of distributions

Average surface temperature [◦ C]

 as

θ oa

ϕ as

Co

Ci

Air velocity [m/s]

1 4

18.5 18.5

20 ± 2◦ C 22.0 ± 2◦ C

15.5 15.5

50 50

360 360

378 378